Cyclic loading tests on dual-steel beam-to-column welded ﬂange-bolted web connections were conducted to quantify their moment resistance, plastic deformation and energy dissipation capacities. The test program consisted of ﬁve one-sided connection specimens, where Q355 grade beams and Q690 grade columns were used, to study the e ects of beam-to-column welded ﬂange connection details and panel zone shear strength on the connection seismic performance. Two types of welded connection details were considered, one was the traditional complete-joint-penetration (CJP) welded connections by use of backing bars, and the other type was those with the bottom backing bar further reinforced by a ﬁllet weld. Three column web thicknesses were designed to arrive at strong, intermediate and weak panel zones, respectively. It was found that the CJP weld with the backing bar reinforced only under the bottom beam ﬂange produced satisfactory performance and accommodated plastic rotations larger than 0.03 rad, while a maximum plastic rotation of 0.04 rad could be developed in the high-strength steel panel zone before fracture occurred. These results evidenced that those dual-steel connections could still sustain high seismic deformation demands. Keywords: Joint, High-strength steel, Dual-steel, Joint, Welded ﬂange-bolted web connection, Cyclic, Experiment Corresponding author Email address: hufx@scut.edu.cn (Fangxin Hu) Preprint submitted to Engineering Structures May 24, 2023 1 Introduction 1 High-strength steels with yield strength of at least 460MPa have become an economical 2 alternative to conventional-strength steels, since the former steels enable increased 3 cross-sectional strength, and thus smaller cross sections to be used for the same loading condition 4 [1–6]. This performance allows for more economical and ecological construction. High-strength 5 steels have distinct material properties from the conventional ones. In addition to the increased 6 yield and ultimate tensile strength, decreased ductility is expected in high-strength steels [7–9]. 7 This matters if the plastic deformation capacity is concerned, such as in plastic design and 8 seismic design. 9 For typical steel moment-resisting frames, ductile performance of beam-to-column 10 connections is critical to assure structural safety under severe earthquakes. Several studies have 11 been conducted on the seismic behavior of high-strength steel beam-to-column connections. The 12 ﬁrst study maybe dates back to the 1990s, when Kuwamura and Suzuki [10] investigated the 13 low-cycle fatigue resistance of beam-to-column fully welded joints of a new, though at that time, 14 type of heat-treated 600MPa tensile-strength grade high-strength steel with yield ratios not more 15 than 80%. They found that the joints had an enough safety margin against a strong earthquake 16 motion of the ultimate intensity prescribed in then Japanese seismic design code, in terms of 17 average and cumulative ductilities. Dubina et al. [11] carried out an extensive testing program to 18 evaluate the monotonic and cyclic performance of moment-resisting joints of high-strength steel 19 and mild carbon steel components, including fully welded connections between S235 beams and 20 S355/S460 columns. Sti eners were used to strengthen the connection zone near the ﬂange 21 welds. Therefore, the connections developed substantial ultimate rotations of nearly 0.1 rad and 22 0.06–0.08 rad under monotonic and cyclic loadings, respectively, before ﬁnal fracture at the 23 sti ener in tension. Those rotations were almost all contributed by panel zones in the 24 connections. Oh and Park [12] studied the deformation capacity of beam-to-column fully welded 25 2 connections using HSA800 grade steel with a target tensile strength of 800 MPa (yield strength 26 of 650–770 MPa, and a yield-to-tensile ratio of 0.85 or less) under cyclic loading. It was found 27 that the traditional welded-type connection with weld access holes and backing plates for groove 28 welds at the end of beam ﬂanges sustained limited plastic rotation (less than 0.01 rad), while the 29 alternative without weld access holes and backing plates, yet reinforced by a ﬁllet weld at the root 30 of each groove weld, survived a plastic rotation of more than 0.01 rad. They also studied the 31 reinforced connections they proposed using horizontal sti eners of di erent shapes to widen the 32 beam ﬂange at the connections, which developed plastic rotations of 0.03 rad or higher. Liao 33 et al. [13] studied the seismic performance of Q460 high-strength steel fully welded cruciform 34 beam-to-column connections, and the e ects of weld access holes and locally widened beam 35 ﬂange at the connection. They reported that the connections exhibited ultimate displacements of 36 80–90 mm (corresponding to story drift angles of 0.053–0.06rad), demonstrating good 37 deformation capacity, although the eventual failure all belonged to cracking and fracture at the 38 beam ﬂange for the unreinforced connections, and cracking between the panel zone and the 39 column ﬂange for the only reinforced connection. Kim et al. [14] investigated experimentally the 40 cyclic behavior of fully welded and extended end-plate beam-to-column connections made of 41 SHN490 and SM490 high-strength steels (measured yield strength of about 450 MPa or above). 42 For the fully welded connections, very large rotations (0.05 rad and 0.06 rad for SHN490 and 43 SM490 specimens, respectively) were recorded even at the maximum moment resistance, before 44 the severe shear instability of the panel zone and the ﬁnal complete fracture of the ﬂange. Wang 45 et al. [15] conducted an experiment on the fully welded connection between a Q355 grade beam 46 and a Q460 grade column, and this connection specimen developed a maximum story drift angle 47 of over 0.035 rad but it was not tested to failure in the end. Liu et al. [16] investigated the 48 low-cycle fatigue fracture behavior of welded ﬂange-bolted web beam-to-column connections 49 made of Q460C high-strength steel and compared the performance among di erent welding 50 details and weld access holes. They found that the connections developed plastic rotations of 51 3 0.02–0.03 rad before the fracture at the beam ﬂange. Lu [17] examined the cyclic behavior of 52 welded ﬂange-bolted web connections made of Q460GJ, Q550GJ and Q690GJ high-strength 53 steels, as part of a series of experimental tests on high-strength steel connections of various types. 54 However, very limited plastic rotations (less than 0.01 rad) were achieved for these connections. 55 In addition to the above unreinforced connections, some studies have also been conducted on 56 cover-plate and ﬂange-plate reinforced connections using high-strength steels in China [17–23], 57 and these studies evidenced satisfactory performance of these connections in terms of ultimate 58 rotation capacity, which was basically above 0.04 rad. Besides, Gir ˜ao Coelho et al. [24] evidenced 59 by a series of experiments that S690 and S960 steel web shear panels were able to undergo shear 60 distortions in the range of 0.05 rad to over 0.1 rad, depending on the panel slenderness, aspect ratio 61 and axial load level. This research qualiﬁed those high-strength steel shear panels as a high-ductile 62 connection component. This superior inelastic shear deformation capacity (above 0.1 rad) was also 63 conﬁrmed by Jord ˜ao et al. [25, 26], although their attention was mainly drawn to the assessment of 64 the plastic resistance of column web panels subjected to shear and load introduction (compression 65 and tension). Luo et al. [27] showed by an experimental study that the H-shape beam-to-column 66 joint panels made up of H-SA700B high-strength steel were able to develop plastic distortions of 67 0.025–0.035 rad under cyclic loading before the ﬁnal weld fracture between the beam and column 68 ﬂanges. The real deformation capacity of this type joint panel should be more than 0.035 rad, 69 since the panels alone were not tested to failure. 70 Those previous studies clearly shed some light on the inelastic deformation capacity of 71 high-strength steel beam-to-column connections under monotonic and cyclic loadings. But the 72 rotation information on individual high-strength steel connection components, such as the plastic 73 hinge formed at the beam end, and the shear panel, is still very limited, especially for seismic 74 consideration. High-strength steels are expected to be applied in columns with a higher priority 75 than in beams, as strong column-weak beam (SCWB) is to be assured in seismic design. In this 76 way, dual-steel beam-to-column connections, where the beam and column are made up of 77 4 conventional-strength and high-strength steels, respectively, should be a promising connection 78 type in seismic resistant steel building frames, as pointed out by Dubina et al. [28]. As mentioned 79 above, the available research concerning this type dual-steel connections is limited to 460 MPa 80 yield-strength grade steel and fully welded type [11, 15], and most of the relevant studies did not 81 succeed to evaluate the real cyclic deformation capacity of joint components, in particular, of 82 high-strength steel shear panels. More research on dual-steel beam-to-column welded 83 ﬂange-bolted web connections, which may be the most widely used connection type in steel 84 construction due to its convenient fabrication and erection, is in urgent need. 85 Therefore, cyclic tests on typical dual-steel welded unreinforced ﬂange-bolted web (WUF-B) 86 connections, where Q355 grade beams and Q690 grade columns were used, were carried out in 87 this paper to evaluate their moment strength, deformation and energy dissipation capacities. The 88 cyclic test program is presented, followed by test results. Discussions are then presented regarding 89 the validity of present seismic provisions for this connection design. The goal of this study is 90 to obtain real plastic rotation capacity and develop seismic design recommendations on this type 91 connections. 92 2 Test program 93 2.1 Design of specimens 94 The T-shape assembly which represents an exterior beam-to-column joint in a moment frame 95 structure was selected for investigation in this paper. As shown in Figure 1, the beam and column 96 sizes of the subassemblage were determined by a proper design of the prototype three-bay six-story 97 plane moment frame. The dead (D) and live (L) loads are 6 kN =m2and 2 kN=m2, respectively. The 98 story height ( H) is 3000 mm, and the in-plane and out-of-plane column spacings ( L) are both 6000 99 mm. When subjected to lateral loads such as earthquakes, such a prototype frame is expected to 100 exhibit reverse curvature bending both in its columns and beams, with inﬂection points occurring 101 near the mid-span of the beams and mid-height of the columns. This assumption is reasonable 102 5 when the gravity load is small in comparison to the seismic load. In the T-shape specimen, the 103 column inﬂection points are simulated with load pins at both the top and bottom of the column, 104 while the free end of the beam at the point of actuator attachment simulates the beam inﬂection 105 point, as described in the following section. 106 T-shape specimen6000 6000 6000 3000 3000 3000 3000 3000 3000 30001500 1500Dead loads: 6kN/m2 Live loads: 2kN/m2 Nc F Figure 1. Extraction of the T-shaped assembly (unit: mm) Beams and columns in the prototype frame were made of Q355 (conventional-strength) and 107 Q690 (high-strength) grade steels, respectively. They were checked for both the ultimate and 108 serviceability limit states. As for the former limit state, a factored non-seismic (or say, 109 fundamental) load combination, 1 :3D+1:5L, was taken into account to check the beam strength, 110 column strength and stability. Strength and stability were also satisﬁed under another factored 111 seismic load combination, 1 :2(D+0:5L)+1:3Ehk1, where E hk1represents the design seismic 112 action of the frequently-occurred 1st-group earthquake in a intensity-8 region of ground type II, 113 according to Chinese Seismic Code [29]. Generally, it is required that the beams in 114 earthquake-resistant frames be braced adequately to avoid lateral-torsional buckling, and thus the 115 instability of beams was not considered. To cater for the latter limit state, the maximum beam 116 deﬂection was within the code rated limit, 1 /400 of the beam span, under a nominal (or say, 117 characteristic) non-seismic load combination, D +L. Besides, the maximum story drift angle 118 under the above-mentioned frequently-occurred earthquake was required not to exceed 1 /250. As 119 6 a result, a built-up H-shaped section, H320 Ö160Ö8Ö14, was used for the Q355 steel beams. 120 However, for the Q690 steel columns, three built-up H-shaped sections, i.e, H240 Ö160Ö8Ö12, 121 H240Ö160Ö10Ö12 and H240 Ö160Ö12Ö12 that are di erent only in the web thickness but all 122 satisfying the strong column-weak beam (SCWB) capacity design, were designed to facilitate the 123 evaluation of the impact of panel zone strength. 124 Five specimens were designed as summarized in Table 1. Each connection specimen had 125 complete joint penetration (CJP) welds connecting the beam ﬂanges to the column ﬂange, and an 126 erection plate (or shear tab) of the same steel grade and thickness to the beam web, shop-welded 127 to the column ﬂange with ﬁllet welds and bolted to the beam web for transfer of shear force. 128 Three web connection bolts used in each specimen were all class 10.9s M20 high-strength bolts 129 with pretension force of 155kN [30]. Continuity plates as required, were also used in this study 130 with the same steel grade, width and thickness to the corresponding beam ﬂanges, to prevent 131 local damage to the column ﬂange and web and to help assure uniform stress in the beam ﬂanges. 132 In addition to the varying column web thickness, two types of CJP welded connection details 133 between the beam and column ﬂanges as shown in Figure 2 were included in this investigation 134 for comparison. The type adetail with arc-shaped weld access holes, as shown in Figure 2(a), is 135 representative of the typical practice that has been commonly used before the Northridge 136 earthquake in the United States due to its simplicity [31], but many connections of this type 137 su ered from weld brittle fracture, especially at the bottom ﬂange of the beam. Therefore, the 138 type bdetail has been developed where a reinforcing ﬁllet weld is placed under the bottom 139 backing bar while the top backing bar is not reinforced [31], and the access hole holes are 140 machined with improved shape speciﬁed in AISC Prequaliﬁed Connections [32] to alleviate 141 stress concentrations in the transition region between the beam ﬂange and the drilled hole, as 142 shown in Figure 2(b). This type bdetail is currently recommended in Chinese Seismic Code [29] 143 and Technical Speciﬁcation [33]. As such, the specimen labels shown in Table 1 begin with the 144 beam and column steel grades, followed by the panel zone thickness, and end with aorb 145 7 indicating the beam-to-column ﬂange welded connection detail. 146 Backing bar left in placeBacking bar left in place635º 303466200 35º3310 33 610.9s M24 boltsShear tab Beam Q355 Column Q690Continuity plate Q355 Continuity plate Q35566 (a) Type a Backing bar left in place and reinforced by a fillet62535º 303466200 35º 63310 33 426Backing bar left in place 10.9s M24 boltsShear tab Beam Q355 Column Q690Continuity plate Q355 Continuity plate Q35566 (b) Type b Figure 2. Connection details Table 1. Test specimens Specimen labelBeam ColumnWelding type Steel gradeCross-section size (mm)Steel gradeCross-section size (mm) B355-C690-PZ12 a Q355 H320 Ö160Ö8Ö14 Q690H240Ö160Ö12Ö12a B355-C690-PZ12 b b B355-C690-PZ10 aH240Ö160Ö10Ö12a B355-C690-PZ10 b b B355-C690-PZ8 b H240Ö160Ö8Ö12 b Manual gas shielding arc welding with matching weld materials was used in this study. An 147 E761T1-K3C electrode, which ﬁts to Q690 steel grade as stipulated in newly published Chinese 148 Design Standard for High-Strength Steel Structures [34] was used for the ﬁllet welds in producing 149 the welded H-sections for the Q690 columns. The other weldments, including the CJP welds 150 between the beam and column ﬂanges, were implemented using E49 type electrodes that ﬁt to 151 Q355 steel [35]. 152 8 2.2 Material properties 153 Material testing was performed on all the steel plates used for the test specimens. Tensile 154 coupon tests were conducted to characterize stress–strain responses of the steel plates. Results of 155 the coupon tests were compared to the requirements for high-strength steel plates in the 156 corresponding codes, so as to ensure those steel plates in this study were qualiﬁed. Three 157 coupons were tested for each plate thickness. Full-thickness coupons were used, with a gauge 158 length and width of 50mm and 20mm, respectively. Several quantities were obtained for each 159 plate thickness, including the modulus of elasticity ( E), the yield strength ( fy) on the yield 160 plateau or at a 0.2% o set strain when the plateau is absent, the strain at the end of the yield 161 plateau (or at the onset of the strain hardening branch of the stress–strain curve, st) if any, the 162 ultimate strength ( fu) and its corresponding ultimate strain ( u), yield-to-tensile strength ratio 163 (fy=fu) and percent elongation after fracture based on the speciﬁed parallel length ( ) [36], as 164 summarized in Table 2, where each value represents the average of three coupons. As a example, 165 the engineering stress–strain curves of Q690 grade steel coupons are shown in Figure 3. Simple 166 tensile testing on the class 10.9s M20 high-strength bolts used in this study was also undertaken 167 to obtain the modulus of elasticity ( E), the ultimate strength ( fu) and ultimate strain ( u), as also 168 shown in Table 2. Table 2. Material properties Steel gradePlate thickness (mm)E (GPa)fy (MPa)st fu (MPa)u fy=fu Q3558 205.0 406 0.014 539 0.153 0.75 25% 14 203.5 368 0.020 522 0.200 0.70 29% Q6908 208.3 723 — 822 0.100 0.88 20% 10 189.4 794 — 902 0.106 0.88 21% 12 205.8 775 0.018 816 0.060 0.95 16% 16 219.8 811 0.022 840 0.055 0.97 17% 10.9s M20 206.0 — — 1135 0.110 — — 169 9 00.050.10.150.20.250.3 Strain01002003004005006007008009001000Stress (MPa) Q690-PL8 Q690-PL10 Q690-PL12 Q690-PL16Figure 3. Stress–strain curves of Q690 coupons 2.3 Test setup 170 A load frame assembly was used for testing conducted in the Structures Laboratory of South 171 China University of Technology. It consists of a system of members designed to form a planar 172 frame to transfer the forces from a jack and another actuator, which are installed within the load 173 frame, to the strong ﬂoor of the laboratory. Figure 4 shows the features of the load frame. The 174 T-shape specimen was placed within the load frame and was attached at the top and bottom of 175 the column to large load pins, which in turn, were connected to the crosshead of the 300 t jack at 176 the top and to the large beam at the bottom of the load frame tied to the strong ﬂoor. In this way, 177 free rotation of the column ends during loading was allowed, simulating inﬂection points at the 178 mid-height of the columns of each story. The pins were fabricated from thick plate material and 179 solid steel dowels. Loading was applied on the top of column by the jack and then to the beam tip 180 by the MTS hydraulic actuator, which is capable of 300 kN at a stroke of 250 mm. The actuator 181 was attached to the beam by a pair of end plates jointed with high-strength threaded steel rods. 182 The top of the actuator was attached tightly to the top beam of the load frame. 183 Out-of-plane movement of the beam due to lateral-torsional buckling was restricted by a pair 184 of brackets attached to an additional pair of columns tied down to the strong ﬂoor. The brackets 185 were designed with rollers so that the e ects of potential friction between the brackets and the 186 10 Jack Pin Δ2N S Δ3 Δ1 Δ5Δ4 PinSpecimen Displacement transducersActuator 3000 Units: [mm]Lateral bracing(a) Scheme (b) On site Figure 4. Test setup beam were negligible. 187 2.4 Loading protocol 188 Before the start of cyclic loading, a constant axial compression force was imposed on the 189 top of the column to represent seismic weight. This introduced an axial compression ratio of 190 0.3 with respect to the nominal axial capacity of the column, which remained the same in all 191 the specimens to ensure consistency. At the cyclic loading stage, while the axial load on the 192 column top remained unchanged, the loading history in AISC Seismic Provisions [37] was used to 193 ensure results could be compared to numerous other tests conducted during and after the SAC Joint 194 Venture investigations in the Unites States [31]. This cyclic loading history at the beam tip is based 195 on the story drift angle instead of plastic rotation levels that may be employed popularly prior to 196 the Northridge earthquake. The story drift angle is deﬁned as the lateral story displacement divided 197 by the story height. However, the specimens were loaded actually by applying displacements to 198 the tip of the beam. Thus, the story drift angle in this study is deﬁned as the ratio of the beam tip 199 displacement to the distance ( L=2, where Lis the column spacing in the prototype frame) between 200 11 the beam tip loading point and the column centerline, which is 3000 mm. Figure 5 gives the 201 prescribed loading history, where positive story drift angles indicate downward displacements at 202 the beam tip. As the stroke of the actuator is 250 mm, this indicates that the maximum story drift 203 angle that can be reached is 8%. If there is no apparent strength degradation observed after two 204 cycles at 8%, additional cycles at this amplitude would be continued until failure of the specimen 205 or further signiﬁcant strength degradation occurs. 206 -10-8 -6 -4 -2 0 2 4 6 8 10 Story drift angle (%)0.375% 0.5%0.75% 1%1.5%2%3%4%5%6%7%8% Figure 5. Loading protocol 2.5 Instrumentation 207 The actuator and the jack were equipped with load cells to measure the magnitude of the 208 applied load at the beam tip and on the column top, respectively. As shown in Figure 4, 209 displacement transducers were used to isolate the inelastic rotations which developed in 210 individual connection components of each specimen, i.e., the shear distortion of the panel zone 211 and the plastic hinge rotation in the beam end. The primary displacement transducer labeled 212 DT-1 measured the displacement at the beam tip. This instrument was used as the 213 displacement-control signal to the actuator. Displacement transducers DT-2 and DT-3 were 214 arranged diagonally to measure the average shear deformation of the panel zone, since shear 215 distortion of the panel zone contributes at a large proportion to the total story drift angle 216 12 especially for specimens designed with weak panel zone. Another two transducers DT-4 and 217 DT-5 attempted to monitor the horizontal displacements at the center of pins connected to the top 218 and bottom of the column. In this way, taking into account the possible rigid rotation of the entire 219 specimen, the story drift angle, , is: 220 =1 L=245 H(1) where 1,4and5are readings from the corresponding labeled displacement transducers, Lis 221 the distance (6000 mm) explained above and His the story height between the pin centerlines 222 (3000 mm). The panel zone shear distortion, pz, is: 223 pz=23 2q b2 pz+h2 pz bpzhpz(2) where 2and3are readings from the diagonal transducers, bpzandhpzare the width and height of 224 the panel zone, which are taken as distances between the column ﬂange centerlines and continuity 225 plate centerlines, respectively. The contribution of the panel zone shear distortion to the beam tip 226 displacement, or equivalently, to the story drift angle relative to the column centerline, is given as 227 [38]: 228 pz= 1hb H! pz (3) where hbis the beam depth. The rotation contribution of the plastic hinge in the beam end, if 229 developed, could be found by subtracting the contribution of the panel zone from the total story 230 drift angle. 231 13 3 Test results 232 3.1 Failure modes 233 Depending on the welding type and panel zone strength, the specimens exhibited various 234 failure modes, as summarized in Figure 6. Basically, fracture occurred due to the compact cross 235 sections used in the beams and columns, but the location of fracture di ered among them. If the 236 panel zone had intermediate strength or was even strong, the beam ﬂanges near the column face 237 developed cracking and ﬁnal fracture along the welding fusion face, for example, in Specimens 238 B355-C690-PZ12 aand B355-C690-PZ10 a, when the story drift angle reached 1:7% in the 239 second negative excursion to 6%, and3:5% in the ﬁrst negative excursion to 5%, 240 respectively (see Figures 6(a) and 6(c)). If the panel zone was weak enough, shear fracture 241 appeared in the column web panel as in Specimen B355-C690-PZ8 bwhen the story drift angle 242 reached 6% in the positive excursion of the fourth cycle of 8% (see Figure 6(e)). 243 It should be borne in mind that the ﬁrst tested specimen is B355-C690-PZ12 b, which 244 eventually failed by the weld fracture between the column ﬂange and continuity plates in cycles 245 at the story drift angle of 5% (see Figure 6(b)). Fracture was initially developed in the welds 246 between the top continuity plates and the column ﬂange under positive loading when the story 247 drift angle arrived at 5% for the ﬁrst time. This fracture was associated with cracking between the 248 column ﬂange and web close to the fractured continuity plate welds. When the loading was 249 continued in the negative direction and the story drift angle reached 5%, the bottom continuity 250 plate welds to the column ﬂange fractured in a similar pattern. After inspection, it was found that 251 the ﬁllet welds were mistakenly used to join the continuity plates and the column ﬂanges, since it 252 is speciﬁed in the code that the more expensive CJP groove welds be used instead. This 253 mistakenly used ﬁllet welds had an inadequate thickness of only about 6 mm, which should be 254 responsible for the premature weld fracture. Since the ﬁllet welding rather than the CJP welding 255 had been used in fabrication of all the specimens, it was decided that those continuity 256 14 Beam flange Column flange(a) Specimen B355-C690- PZ12 a Beam Panel zoneColumn(b) Specimen B355-C690- PZ12 b Beam flange Column flange(c) Specimen B355-C690- PZ10 a Column Beam (d) Specimen B355-C690- PZ10 b BeamColumn Panel zone(e) Specimen B355-C690- PZ8b Figure 6. Failure modes plate-to-column ﬁllet welds in the rest specimens were reinforced to a thickness of 12 mm by 257 overlay welding. Results of the rest specimens evidenced the adequate performance of this type 258 of reinforced ﬁllet welds in seismic conditions. 259 It should be also explained that Specimen B355-C690-PZ10 bdeveloped notable 260 lateral-torsional buckling of the beam in cycles of the 7% story drift angle (see Figure 6(d)). This 261 was because the span of lateral bracing was not broad enough to accommodate such a large 262 ﬂexural deformation of the beam at this amplitude. 263 15 3.2 Hysteresis curves 264 Moment–story drift angle hysteresis curves are shown in Figure 7 for all the specimens, where 265 the moment corresponds to the column face and was calculated as the product of the distance 266 between the loading point at the beam tip to the column face, i.e., 3000 mm minus half of the 267 column depth, and the reaction force at the beam tip. The yielding and full plastic moment levels 268 of the beam cross section, My;bandMp;b, determined from the measured material properties shown 269 in Table 2, are also plotted for comparison. Besides, the moment–shear distortion hysteresis curves 270 for all the panel zones in those specimens are shown in Figure 8, where the moment is the same 271 to the above, while the shear distortion was calculated using Eq. (2). The yielding moment of 272 the panel zone in an average sense considering the e ect of axial compression and column shear, 273 which is indicated in Figure 8 for comparison, is 274 My;pz=s 1 Nc fycAc!2fycp 3bpzhpztpzLhc LH Hhpz(4) where Ncis the axial compression force on the column, fy;c,hcandAcare the material yield 275 strength, cross-sectional height and area of the column, respectively, tpzis the panel zone thickness. 276 Solid triangular symbols were added in these ﬁgures to indicate the instant of fracture. 277 As expected, the hysteresis curves of both the entire specimens and the panel zones were full 278 before the ﬁnal failure, and the sti ness and strength remained stable through large inelastic 279 deformations, except for Specimen B355-C690-PZ10 b. This particular specimen failed by 280 lateral-torsional buckling, resulting in somewhat pinching and unsymmetrical responses in 281 positive and negative loading directions. Deterioration of the hysteresis curves of this specimen 282 was noted caused by the lateral-torsional buckling, when the story drift angle exceeded 6%. All 283 the specimens developed the yielding capacity of the beam, and even its full plastic capacity. But 284 there was high risk for exception when the type aconnection detail was employed. For example, 285 Specimen B355-C690-PZ10 afailed to develop the full plastic beam moment when loaded in the 286 16 -8 -6 -4 -2 0 2 4 6 8 Story drift angle (%)-500-2500250500Moment (kNm)M p,b=321kNm -M p,b=-321kNmM y,b=280kNm -M y,b=-280kNm(a) Specimen B355-C690-PZ12 a -8 -6 -4 -2 0 2 4 6 8 Story drift angle (%)-500-2500250500Moment (kNm)M p,b=321kNm -M p,b=-321kNmM y,b=280kNm -M y,b=-280kNm (b) Specimen B355-C690-PZ12 b -8 -6 -4 -2 0 2 4 6 8 Story drift angle (%)-500-2500250500Moment (kNm)M p,b=321kNm -M p,b=-321kNmM y,b=280kNm -M y,b=-280kNm (c) Specimen B355-C690-PZ10 a -8 -6 -4 -2 0 2 4 6 8 Story drift angle (%)-500-2500250500Moment (kNm)M p,b=321kNm -M p,b=-321kNmM y,b=280kNm -M y,b=-280kNm (d) Specimen B355-C690-PZ10 b -8 -6 -4 -2 0 2 4 6 8 Story drift angle (%)-500-2500250500Moment (kNm)M p,b=321kNm -M p,b=-321kNmM y,b=280kNm -M y,b=-280kNm (e) Specimen B355-C690-PZ8 b Figure 7. Moment–story drift angle hysteresis curves 17 -5 -3 -1 1 3 5 Shear distortion (%)-500-2500250500Moment (kNm)M y,pz=388kNm -M y,pz=-388kNm(a) Specimen B355-C690-PZ12 a -5 -3 -1 1 3 5 Shear distortion (%)-500-2500250500Moment (kNm)M y,pz=388kNm -M y,pz=-388kNm (b) Specimen B355-C690-PZ12 b -5 -3 -1 1 3 5 Shear distortion (%)-500-2500250500Moment (kNm)M y,pz=332kNm -M y,pz=-332kNm (c) Specimen B355-C690-PZ10 a -5 -3 -1 1 3 5 Shear distortion (%)-500-2500250500Moment (kNm)M y,pz=332kNm -M y,pz=-332kNm (d) Specimen B355-C690-PZ10 b -8 -6 -4 -2 0 2 4 6 8 Shear distortion (%)-500-2500250500Moment (kNm)M y,pz=240kNm -M y,pz=-240kNm (e) Specimen B355-C690-PZ8 b Figure 8. Moment–shear distortion hysteresis curves 18 positive direction. 287 As for the panel zone responses, both Specimens B355-C690-PZ12 band B355-C690-PZ10 b 288 exhibited unauthentic results because their failure modes were unexpected. Both should have 289 been avoided by correct seismic design. The panel zone yielding moment in Eq. (4) seemed to 290 lead to overestimation for Specimens B355-C690-PZ10 a(see Figure 8(c)) and B355-C690-PZ12 a 291 (see Figure 8(a)), but this might be more likely resulted from the variation in material properties, 292 because a satisfactory agreement was still noted in Specimen B355-C690-PZ8 bin Figure 8(e). It 293 was interesting to ﬁnd that even the Q690 high-strength steel panel zone was capable of very large 294 inelastic distortions, as evidenced by this specimen. Considerable plastic hardening was observed 295 in this panel zone, because of the material strain hardening and the participation of the column 296 ﬂanges in shear transfer after the initial yielding of the web panel. This single test conﬁrmed that 297 panel zone yielding resulted in reliable energy dissipation, even in case of high-strength steel used. 298 3.3 Strength, deformation and energy dissipation 299 Based on the hysteresis curves in Figure 7, the elastic sti ness, Ke, was determined by linear 300 ﬁt to the data within an amplitude of 1.5% and is shown in Table 3. The ultimate (or maximum) 301 moment, Mu, was identiﬁed and is also summarized in Table 3, where Mp;b,My;bandMy;pz(see 302 Eq. (4)) are the full plastic and yielding moments of the beam, and the yielding moment of 303 the panel zone, respectively, determined based on the measured material properties. Due to the 304 variation in the measured material strength from the nominal grade values, the specimens with 305 panel zones of thickness of 12mm and 10mm all behaved as connections of strong panel zone 306 design, because their measured My;pzvalues are even larger than Mp;bvalues. On the contrary, 307 Specimen B355-C690-PZ8 bbehaved as expected with a much lower My;pzthan My;b, and its panel 308 zone did dominate its performance. 309 Overstrength with respect to My;min, the minimum of My;bandMy;pz, and to the full plastic 310 moment of the beam, Mp;b, denoted as Mu=My;minandMu=Mp;b, respectively, are included in Table 311 3. The overstrength values show apparent disparity between di erent connection details. In the 312 19 Table 3. Sti ness and strength Specimen label Ke (kNm)Mp;b (kNm)My;b (kNm)My;pz (kNm)My;min (kNm)Mu (kNm)Mu My;minMu Mp;b B355-C690-PZ12 a13632 321 280388 280327 1.17 1.02 B355-C690-PZ12 b13326 364 1.30 1.13 B355-C690-PZ10 a13271332318 1.14 0.99 B355-C690-PZ10 b13404 368 1.32 1.15 B355-C690-PZ8 b 12553 240 240 321 1.34 1.00 specimens with actually strong panel zones (those except for B355-C690-PZ8 b), the overstrength 313 relative to the yielding moment, Mu=My;min, was about 1.15 when the type aconnection detail 314 was used, while it was about 1.30 when the type bconnection detail was used. The overstrength 315 relative to the full plastic moment of the beam, Mu=Mp;b, also shows the same trend, and was 316 about 1.0 and 1.15 for the connection types aandb, respectively. This demonstrates the better 317 performance of the type bover type ain delaying connection failure, so that more strain hardening 318 could be developed. In the specimen with weak panel zone (B355-C690-PZ8 b), signiﬁcant strain 319 hardening was developed since its Mu=My;minreached nearly 1.35, slightly higher than that in the 320 specimens of strong panel zones and the same connection detail (1.30 for B355-C690-PZ12 band 321 1.32 for B355-C690-PZ10 b). Such high overstrength even led to the full plastic state of the beam 322 (Mu=Mp;bof 1.0), which indicates that some inelastic rotation should be expected at the beam end. 323 Quantities related to deformation and energy dissipation capacities are summarized in Table 324 4. The yield story drift angle, y, was calculated as My;min=Ke. The ultimate story drift angle, u, 325 was counted only if a least one complete cycle of the target story drift angle was completed 326 before fracture occurred [31]. The plastic story drift angle (or plastic rotation, relative to the 327 column centerline) of the entire connection, p, was identiﬁed as the plastic component of u. 328 Table 3 also contains the plastic rotation component in the beam, p;b, and that in the panel zone, 329 p;pz. The plastic rotation of the panel zone ( p;pz) was deduced from the total rotation of the panel 330 zone (pz) shown in Eq. (3), while the plastic rotation of the beam ( p;b) was obtained by 331 20 subtracting p;pzfromp. In this way, the ductility factor, , was computed as the ratio of utoy. 332 Cumulative plastic rotations, p, and energy dissipation, A, of the entire specimens were 333 determined based on the method in ATC-24 [39, 40], as shown in Figure 9, and they were 334 counted until the last successful excursion (or half-cycle) before the fracture, as summarized in 335 Table 4 as well as their normalized values by the yielding quantities, p=yandA= My;miny . 336 It should be borne in mind that, Specimen B355-C690-PZ12 bunderwent an unexpected and 337 unreasonable fracture at the continuity plate welds, and so this specimen would not be discussed 338 in the following evaluation. Specimen B355-C690-PZ10 bexperienced also unexpected 339 lateral-torsional buckling, which should have been avoided, and so the results of this specimen 340 could be deemed as conservative. 341 RotationMomentpii-th excursion pi+1 i+1-th excursionA i A i+1 Figure 9. Deﬁnition of cumulative plastic rotation and energy dissipation in ATC-24 [39] Table 4. Deformation and energy dissipation Specimen label y (rad)u (rad)  p (rad)p;pz (rad)p;b (rad)pp yA (kJ)A My;miny B355-C690-PZ12 a0.020 0.06 3.0 0.033 0.003 0.030 0.558 27.1 167.2 27.9 B355-C690-PZ12 b0.020 0.04 2.0 0.016 0.003 0.013 0.201 9.5 51.5 8.4 B355-C690-PZ10 a0.021 0.04 1.9 0.017 0.005 0.012 0.249 11.8 66.0 10.7 B355-C690-PZ10 b0.021 0.07 3.3 0.041 0.004 0.037 1.021 48.9 312.0 51.2 B355-C690-PZ8 b 0.019 0.08 4.2 0.051 0.039 0.012 1.638 85.7 455.7 95.3 21 It is clear that, all the specimens developed large story drift angles and very ductile 342 performance was achieved, if a story drift angle of 0.04 rad is deemed su cient for Special 343 Moment Frames (SMF) in AISC Seismic Provisions [37]. Due to the small sections of the beams 344 and columns used in this study, the yielding drift angles of about 0.02 rad were nearly two times 345 the traditionally expected value (about 0.01 rad). That means, if instead, a plastic rotation of 346 0.03rad is deemed su cient, connections using the type aconnection detail may be judged to 347 have inadequate rotation capacity, such as shown by Specimen B355-C690-PZ10 a. The other 348 specimens with type bconnection detail performed well beyond this plastic rotation limit. 349 Moreover, it is shown that the weak panel zone brought further increase in the plastic rotation. 350 This is impressive in that it has been traditionally deemed that excessive deformation in panel 351 zone is detrimental to rotation capacity due to the “kink” formed in column ﬂanges, which results 352 in large local stress or strain concentrations near the ﬂange welds. But this adverse e ect was not 353 observed in this study. On the contrary, the panel zone of Q690 grade in the present study 354 sustained as large a plastic shear rotation as about 0.04 rad before its shear fracture, as shown by 355 Specimen B355-Q690-PZ8 b. This panel zone rotation quantity provides valuable reference in 356 seismic performance evaluation, because none of the previous studies have collected this quantity 357 as real as possible. Besides, the cumulative quantities also demonstrate ductile performance. The 358 cumulative plastic rotations were at least 10 times larger than the amplitude of plastic rotation in 359 case of connection type a, were nearly 50 times in case of connection type b, and over 80 times if 360 panel zone rotation was dominant. A similar comparison could be made on the cumulative energy 361 dissipation. Those capacities revealed by this study are much higher than the conventional 362 anticipated seismic demand, where the cumulative plastic deformation is ﬁve to eight times larger 363 than the maximum plastic deformation [40, 41]. 364 22 4 Discussions on current seismic provisions 365 4.1 Welded connection details 366 In the current Chinese codes relevant to seismic design of steel structures [29, 33, 35], the 367 type bconnection detail is required for the CJP weld between the beam and column ﬂanges, and 368 the traditional type aconnection detail is not allowed any more. This requirement was conﬁrmed 369 in the present study because Specimen B355-C690-PZ10 awith connection type asustained a 370 plastic rotation of less than 0.02 rad, which may be insu cient in ductile structures as mentioned 371 above. In fact, nowadays the welding requirements in the United States are even more strict than 372 the above type b. According to AISC Prequaliﬁed Connections [32] for Special Moment Frames 373 (SMF) as well as Intermediate Moment Frames (IMF), the fully welded (welded ﬂange-welded 374 web) connection is recommended. The CJP weld with the backing bar reinforced by a ﬁllet is 375 speciﬁed at the top beam-to-column ﬂange connection, while the CJP weld without a backing bar 376 (or removed after welding) is speciﬁed at the bottom. Therefore, the present study revealed that the 377 welded ﬂange-bolted web connection with the slightly reinforced type bconnection detail might 378 still be adequate to develop a plastic rotation of not less than 0.03 rad, even when high-strength 379 steel is employed in the column. 380 As for the continuity plate welds, the present study also shed some light. The 381 above-mentioned Chinese codes and AISC Seismic Provisions generally require CJP groove 382 welds between continuity plates and column ﬂanges, while ﬁllet welds are allowed between 383 continuity plates and the column web. In this study, although the ﬁllet welds were unintentionally 384 used in all the specimens for the connection of continuity plates to column ﬂanges, they 385 performed well throughout the tests, as long as those ﬁllet welds had adequate size. This 386 advantage of using more economical ﬁllet welds than CJP groove welds has been evidenced as 387 well for welded ﬂange-welded web connections made up of conventional-strength steel [42, 43]. 388 23 4.2 Panel zone 389 Since the e ect of panel zone strength was investigated in this study, it would be necessary to 390 compare the impact of required panel zone strength in di erent design codes, in order to provide 391 a clear picture of whether the current panel zone design method is adequate or should be revised. 392 Therefore, the present design equations are summarized as follows. In Chinese Standard [35], the 393 moment strength of the panel zone should satisfy 394 4 3fy;pzp 3hpzbpztpz pzWp;bfy;b (5) In the United States, it is required by AISC Seismic Provisions [37] that 395 0:6fy;pz0:95hctpzhb 1+3bct2 fc hbhctpz! min fy;b+fu;b 2fy;b;1:2! Wp;bfy;b (6) It should be mentioned that after the SAC Joint Venture investigations, the panel zone strength is 396 suggested within the lower and upper bounds in the FEMA report [31] as 397 (0:9)0:55fy;pzhctpzWy;bfy;b hb L Lhc! Hhb H! (7a) (0:6)0:55fy;pzhctpzWy;bfy;b hb L Lhc! Hhb H! (7b) Lastly, Eurocode 8 [44] requires that 398 0:9fy;pzp 3(hc2tfc)hpztpz +2Mp;fc+min 2Mp;fc;2Mp;st Wp;bfy;b (8) In the above equations, fy;pzis the material yield strength of the panel zone, fy;bandfu;bare the 399 material yield and tensile strength of the beam, respectively, Wy;bandWp;bare the elastic and 400 plastic section moduli of the beam, respectively, bcandtfcare the width and thickness of the 401 column ﬂange, respectively, Mp;fc=bct2 fcfy;cf=4 and Mp;st=bct2 stfy;st=4 are plastic moment 402 24 resistances of the column ﬂange and the pair of continuity plates, respectively, where tstandfy;st403 are the thickness and material yield strength of the continuity plates, respectively, pzis a 404 coecient taken as 0.95 for one-sided connections by Chinese Standard [35], is another 405 coecient, also called transformation parameter in Eurocode 3 [45], to account for the e ect of 406 column shears and it should be determined based on the internal force equilibrium. Panel zone 407 strength ratios of the left-hand side to the right-hand side of all the above design equations based 408 on the measured material strength are summarized in Table 5. It is apparent that the weakest 409 panel zone is designed using Chinese Standard, while the strongest using Eurocode 8, as already 410 demonstrated by a previous comparative study [19]. Both the AISC Seismic Provisions and 411 Eurocode 8 rated panel strength lie in the range given by the FEMA bounds. All the specimens 412 satisﬁed Chinese Standard (ratios larger than 1), but only Specimen B355-C690-PZ8 bwas not 413 qualiﬁed with enough panel strength according to AISC Seismic Provisions or Eurocode 8 (ratios 414 less than 1). However, it was this specimen that demonstrated the best plastic rotation and energy 415 dissipation capacities. This indicates that the criteria governing the minimum strength of the 416 panel zone may be relaxed in the AISC code and Eurocodes. Table 5. Comparison of panel zone strength ratios by Eqs. (5–8) Specimen labelChinese AISC Seismic FEMAEurocode 8 Standard Provisions Lower bound Upper bound B355-C690-PZ12 a1.64 1.28 1.36 0.91 1.17B355-C690-PZ12 b B355-C690-PZ10 a1.40 1.11 1.16 0.77 1.01B355-C690-PZ10 b B355-C690-PZ8 b 1.02 0.83 0.85 0.56 0.75 417 5 Conclusions 418 An experimental study on the cyclic behavior of dual-steel beam-to-column welded ﬂange- 419 bolted web connections, i.e., whose columns and beams were made of Q690 and Q355 grade 420 25 steels, respectively, was undertaken in this paper. The experimental program consisting of ﬁve one- 421 sided connection specimens was intended to investigate the inﬂuence of beam-to-column ﬂange 422 welded connection details and panel zone strength. The following conclusions are drawn: 423 1) All the specimens with intermediate and strong panel zones exhibited weld fracture, as 424 expected, at the beam-to-column ﬂange connections, and this weld fracture was largely 425 a ected by the welded connection detail used. The specimen with a weak panel zone 426 exhibited shear fracture of the panel zone. 427 2) The strong-panel specimens as traditionally welded with backing bars, developed about 15% 428 larger maximum moment resistances than the beam yielding moment, close to the beam full 429 plastic moment; they sustained plastic rotations of 0.01–0.03 rad and cumulative plastic 430 rotations of more than 10 times larger than the yielding rotation. 431 3) The strong-panel specimens with traditional backing bars, but further reinforced with ﬁllet 432 welds underneath the bottom ﬂange, developed 30% larger maximum moment resistances than 433 the beam yielding moment, or nearly 15% larger than the beam full plastic moment; they 434 sustained plastic rotations of at least 0.03 rad and cumulative plastic rotations of about 50 435 times larger than the yielding rotation. 436 4) The weak-panel specimen with the ﬁllet-reinforced backing bar underneath the bottom ﬂange 437 exhibited signiﬁcant strain hardening in the panel zone by developing a 30% larger maximum 438 moment resistance than the yielding moment of the panel zone. This specimen was able to 439 achieve a maximum plastic rotation of 0.05 rad and a cumulative plastic rotation of over 80% 440 larger than the yielding rotation. Panel zone yielding did not produce adverse e ect on the 441 plastic rotation capacity. 442 5) The welded connection detail with backing bars and reinforced by a ﬁllet weld only underneath 443 the bottom beam ﬂange is adequate for highly ductile seismic demand with a maximum plastic 444 rotation of 0.03rad, while the traditional one with unreinforced backing bars is not. The high- 445 strength steel panel zone is also very ductile with a maximum plastic rotation capacity of 0.04 446 26 rad. 447 It is inferred from this study that those welded ﬂange-bolted web connections with 448 high-strength steel columns generally exhibit excellent inelastic performance with desirable 449 energy dissipation characteristics, if their bottom backing bars are reinforced by ﬁllet welds, and 450 their continuity plate welds and lateral bracing are adequate. Fracture is expected to occur only 451 after plastic rotations larger (sometimes signiﬁcantly) than 0.03rad. It must be noted that the 452 experiments in this study used to justify this superior performance were on beams with a depth of 453 only 320mm. Further study on connections of larger size beams, especially those with deeper 454 sections, is needed. 455 Acknowledgments 456 The authors would like to acknowledge the ﬁnancial supports for this work by the National 457 Natural Science Foundation of China (Grant Nos. 51638009, 51978279 and 52108145). 458