Four Q690 high-strength steel beam-to-column moment connections, which use welded joints between beam and column ﬂanges as well as a bolted beam web to the column ﬂange, were tested under cyclic loading. The e ects of beam-to-column welding details and panel zone strength were studied. Two pairs of welded ﬂange details were incorporated, including the Chinese code-speciﬁed complete-joint-penetration (CJP) welded connections with backing plates where the bottom backing plate is reinforced by a ﬁllet weld, and enhanced CJP welded connections with backing plates removed, weld roots backgouged, and further reinforced by ﬁllet welds. Three panel zone thicknesses were designed to characterize strong, intermediate and weak panel zones, respectively. The test results show that backing plates should be removed in these moment connections to prevent brittle weld fracture, but still rendering only limited plastic hinge rotations in the order of 0.01–0.02 rad in the beam end before ductile fracture of the beam ﬂange. The panel zone, however, survived even after a plastic shear rotation of 0.03 rad, demonstrating much higher plastic deformation capacity than the beam plastic hinge. This suggests that ductility can be exploited in 690 MPa high-strength steel panel zones, but not reliably in 690 MPa steel beams. Keywords: High-strength steel, Beam-to-column connection, Moment connection, Welded ﬂange-bolted web connection, Cyclic, Experiment Corresponding author Email address: hufx@scut.edu.cn (Fangxin Hu) Preprint submitted to Thin-Walled Structures December 4, 2023 1 Introduction 1 High-strength steels with a nominal yield strength no less than 460 MPa have a great deal of 2 beneﬁts over conventional-strength steels. By using high-strength steels, the increased 3 cross-sectional strength results in reduced structural weight for the same loading condition [1–6], 4 and thereby allows for more economical and ecological construction. The increased elastic 5 deformation capacity of high-strength steels makes them suitable for structural members to 6 remain elastic under strong earthquakes, which means these steels have a great potential to be 7 used in seismic resilient steel structures, such as in self-centering structures [7–9]. In spite of 8 their much higher yield and ultimate tensile strength, high-strength steels have much lower 9 ductility, in terms of the fracture strain or elongation, than conventional steels [10–12]. This 10 greatly impacts the inelastic behavior of structural members or connections. It is well known that 11 ductility of moment-resisting connections is important for satisfactory seismic performance of 12 steel moment frames, but these moment connections made of high-strength steels have not been 13 scrutinized. Regarding inelastic behavior under monotonic or cyclic loading, a limited number of 14 studies have been conducted on high-strength steel welded ﬂange-welded web connections 15 [13–16], welded ﬂange-bolted web connections [17–21], bolted extended /ﬂush end-plate 16 connections [14, 22–33], cover-plate /ﬂange-plate /sti ener/haunch reinforced connections 17 [15, 16, 18, 19, 34–41] as well as column web panels subjected to shear [42–44]. Among the 18 above various connection types, beam-to-column welded-ﬂange connections, including welded 19 ﬂange-welded web and welded ﬂange-bolted web connections, are still the most popular 20 connections in steel construction, both owing to their fully restrained property and convenient 21 fabrication as well as erection in practice. 22 Kuwamura and Suzuki [13] tested seven beam-to-column connections subjected to cyclic 23 loading of di erent constant displacement amplitudes. These connection specimens were made 24 of 600MPa tensile-strength grade high-strength steel with yield ratios less than 80%. The 25 2 beam-to-column ﬂanges were grooved and complete-joint-penetration (CJP) welds were used 26 with ﬂux backing plates. It should be noted that these backing plates were removed after welding 27 and reinforcing ﬁllet welds were applied. No scallops or weld access holes were needed as these 28 groove welds were completed in the fabrication shop using a ﬂux-core arc welding (FCAW) 29 process. Two levels of heat input (20 kJ /cm and 60 kJ /cm) were examined. The results show that 30 these connections with a small-size beam (H-shaped section of H200 Ö100Ö9Ö9 mm) sustained 31 substantial cumulative plastic deformation. Kuwamura and Suzuki [13] also concluded that the 32 connections had an enough safety margin against the prescribed strong earthquake motion in 33 Japan by comparing the average and cumulative ductilities revealed in the tests with the demands 34 from seismic analysis. 35 Dubina et al. [14] evaluated eight types of beam-to-column moment connections through both 36 monotonic- and cyclic-loading tests, among which one type was the fully welded connection 37 constructed from a mild steel (S235) beam and high-strength steel (S460) column, strengthened 38 by vertical sti eners outside the beam ﬂanges. The beam and column ﬂanges were connected 39 using CJP groove welds without access holes or backing plates, while the beam web was ﬁllet 40 welded to the column. Metal active-gas welding (MAG) was used. After buckling of the sti ener 41 and beam ﬂange under compression, and some shear deformation of the panel zone, this 42 specimen ﬁnally failed by weld crack initiation at the sti ener in tension. It developed large 43 rotation exceeding 0.06 rad under cyclic loading, and nearly 0.1 rad under monotonic loading. 44 These rotation capacities were all contributed by the panel zone. 45 Oh and Park [15] assessed eight beam-to-column connections made of HSA800 grade steel 46 (tensile strength above 800 MPa and yield strength between 650 and 770 MPa) under cyclic 47 loading, among which one connection had traditional weld details (CJP groove welds at the beam 48 ﬂanges with backing plates and weld access holes, and two-side ﬁllet welds at the beam web), 49 and another connection used the non-scallop (no-access-hole) welding method that removed 50 backing plates and reinforced the CJP weld root with ﬁllet welds. They observed that the 51 3 traditional welded connection developed very limited plastic rotation less than 0.01 rad, while the 52 non-scallop connection sustained nearly 0.02 rad. 53 Liao et al. [16] tested four beam-to-column cruciform welded ﬂange-welded web connections 54 made of Q460D (460 MPa) steel in China under cyclic loading. One of the connections used the 55 traditional weld details and weld access hole geometry speciﬁed in Chinese Technical 56 Speciﬁcation [45], where backing plates were used for CJP groove welds at both beam ﬂanges 57 but only the backing plate below the bottom beam ﬂange was reinforced by a ﬁllet weld beneath. 58 Another two of the connections used the suggested improved access hole geometries in Chinese 59 Technical Speciﬁcation [45] and FEMA-350 [46], respectively, while their weld details were the 60 same to the previous. It was reported that these connections sustained ultimate displacements of 61 80–90 mm, or story drift angles of 0.053–0.06 rad, demonstrating substantial deformation 62 capacity. No signiﬁcant e ect of the weld access hole geometry in these connections was found. 63 Nie et al. [19] tested four beam-to-column one-sided moment connections made of 64 Q690GJNHE (690 MPa) steel in China under cyclic loading, among which two were welded 65 ﬂange-bolted web connections and di ered in cross-sectional slenderness of the beam and 66 column. The beam ﬂanges were CJP groove welded to the column using ceramic backing plates 67 which could be easily removed after welding. The improved weld access hole details in Chinese 68 Technical Speciﬁcation [45] were followed and rich argon gas-shielded metal arc welding 69 (GMAW) process was used. The beam web was connected to the column through a pair of shear 70 tabs by four high-strength bolts. The two specimens experienced fracture in heat-a ected zone 71 (HAZ) of the beam groove weld, and developed plastic deformation of about 20 mm, 72 corresponding to a plastic rotation of about 0.012 rad. The specimen with more slender beam and 73 column sections was found to sustain slightly lower plastic deformation that the other one. 74 Qiang et al. [20] evaluated four beam-to-column welded ﬂange-bolted web connections 75 through monotonic-loading tests under ambient and elevated temperatures. Two specimens, made 76 of Q690 and Q960 high-strength steels in China, respectively, were tested monotonically under 77 4 an ambient temperature. It seems that the beam ﬂanges were connected to the column using CJP 78 groove welds without any backing plate, but the details on weld access holes were not provided. 79 Interestingly, the Q690 specimen failed by fracture between the column ﬂange and the column 80 web and continuity plate, while the Q960 specimen underwent fracture in the beam 81 ﬂange-to-column groove weld. The column ﬂange-to-continuity plate fracture in the previous 82 specimen seemed to be caused by the misuse of ﬁllet welds, whereas current codes generally also 83 require CJP groove welds for continuity plates. Nevertheless, the Q690 specimen developed large 84 connection rotation of almost 0.06 rad, while the Q960 specimen developed only a half. 85 Liu et al. [17] tested four Q460C (460 MPa) steel beam-to-column welded ﬂange-bolted web 86 connections subjected to cyclic loading. These connections had di erent welding details. Two 87 weld access hole shapes (the standard shape in Chinese Seismic Code [47] and improved shape in 88 FEMA-350 [46]) were examined, as well as four details related to backing plates (i.e., steel 89 backing plates left in place, backing plates removed and weld roots backgauged and 90 ﬁllet-reinforced, backing plates reinforced by ﬁllet welds beneath, ceramic backing plate 91 removed after welding). All specimens underwent crack initiation at the termination of beam 92 ﬂange groove weld and ﬁnally fracture through the beam ﬂange. They developed considerable 93 plastic rotations of 0.02–0.03 rad. As expected, the connection using ceramic backing plates 94 sustained the largest cumulative plastic deformation, followed by the connection with backing 95 plates removed after welding. The other two connections with backing plates, reinforced by ﬁllet 96 welds or not, behaved very similarly. 97 Lu [18] tested eight (four one-sided and four cruciform) moment connections under cyclic 98 loading, among which the connections made of Q460GJ (460 MPa), Q550GJ (550 MPa) and 99 Q690GJ (690 MPa) steels, respectively, had the typical welded ﬂange-bolted web detailing. For 100 the CJP groove welds between the beam and column ﬂanges, backing plates were used and left 101 in place. No ﬁllet welds were added for reinforcement. These specimens had a strong panel 102 zone so that the plastic hinge was expected in the beam end. Except for the Q460GJ specimen, 103 5 which developed a plastic rotation of about 0.012 rad, the Q550GJ and Q690GJ specimens hardly 104 developed any plastic deformation since they fractured upon loading into the ﬁrst signiﬁcant post- 105 yielding cycle. 106 The panel zone plays an important role in the seismic behavior of beam-to-column moment 107 connections. Previous research have evidenced that moderate yielding in the panel zone may 108 promote the total plastic rotation capacity, but too much shear distortion of the panel zone may be 109 detrimental [41, 48]. Then it deserves to be examined how much shear distortion could be 110 sustained by high-strength steel panel zones. Gir ˜ao Coelho et al. [42] indicated through 111 monotonic-loading tests that S690 and S960 shear panels could sustain shear distortions 112 exceeding 0.05 rad, even above 0.1 rad, and this deformation capacity highly depended on the 113 panel slenderness, aspect ratio and axial load level. Jord ˜ao et al. [43] also conﬁrmed this superior 114 inelastic performance. Further, Luo et al. [44] evaluated eight beam-to-column connections under 115 cyclic loading, among which two connections had H-shape beams connected to H-shape columns 116 using welded ﬂange-welded web details. The CJP groove welds at the beam ﬂanges used backing 117 plates but weld access holes were not used. The connections developed plastic distortions of 118 0.025–0.035 rad in the panel zone before ﬁnal fracture in the CJP welds. The real deformation 119 capacity of this steel panel zone should be somewhat higher since the panel zones in the 120 specimens remained intact. Interestingly, the premature fracture of the CJP weld, which used a 121 matching weld ﬁller material, occurred but the beam in the above specimens developed little 122 plastic deformation. 123 Recently, the authors [21] tested ﬁve welded ﬂange-bolted web connections, in which beams 124 and columns were made of Q355 (355 MPa) conventional steel and Q690 high-strength steel, 125 respectively. The authors discovered that for both traditional [45] and improved [46] weld access 126 holes, the CJP groove weld connecting the Q355 beam ﬂange to the Q690 column with a backing 127 plate, where the weld metal matched with the beam steel grade, could develop substantial plastic 128 rotation reaching 0.03 rad. This rotation even qualiﬁes for special moment frames in AISC Seismic 129 6 Provisions [49], which may partly be attributed to the relatively shallow beam (320 mm in depth) 130 used in the tests compared to the United States practice. More interestingly, it was found that the 131 Q690 panel zone experienced a plastic rotation of 0.04 rad before its shear fracture. This data 132 provides the reference rotation capacity of the panel zone made of 690 MPa high-strength steel. 133 In spite of the previous studies, limited evidence is available in China for high-strength steel 134 welded ﬂange-welded /bolted web moment connections, especially those connections using 690 135 MPa or higher grade steel. Note that, several tests have indicated quite limited rotation capacity 136 of 690 MPa steel beam plastic hinge formed at the column face [15, 18, 19, 44], while the 690 137 MPa steel panel zone seems to behave in a very ductile manner [21, 44]. Therefore, to provide 138 further evidence, an experimental program was conducted to develop conﬁdence and 139 fundamental data related to the rotation capacity of welded unreinforced ﬂange-bolted (WUF-B) 140 web moment connections made of Q690 steel in China. Four specimens were tested under cyclic 141 loading to examine their seismic performance in this paper, to supplement previous tests on 142 dual-steel moment connections [21]. The plastic rotation and energy dissipation capacities 143 revealed by this study are aimed at developing design guidelines for this connection and 144 evaluating seismic performance of corresponding moment frames. 145 2 Test program 146 2.1 Design of specimens 147 This paper focuses on investigating the T-shaped subassemblage, which represents a 148 connection between an exterior column and beam in a moment frame structure. The dimensions 149 of the beam and column in the subassemblage were determined based on an appropriate design of 150 a three-bay six-story plane moment frame prototype, as illustrated in Figure 1. The applied dead 151 (D) and live (L) loads are 6 kN =m2and 2 kN=m2, respectively. The height of each story ( H) is 152 3000 mm, and both the in-plane and out-of-plane column spacing ( L) are 6000 mm. Under lateral 153 loads like earthquakes, it is anticipated that the prototype frame will experience reverse curvature 154 7 bending in both the columns and beams, with inﬂection points occurring near the mid-span of the 155 beams and the mid-height of the columns. This assumption holds true when the seismic load is 156 signiﬁcantly larger than the gravity load. In the T-shaped specimen, load pins were employed at 157 the top and bottom of the column to simulate the column inﬂection points, while the beam 158 inﬂection point was simulated by the free end of the beam where the actuator is attached. Further 159 details can be found in the subsequent section. 160 T-shape specimen6000 6000 6000 300030003000300030003000 30001500 1500Dead loads: 6kN/m2 Live loads: 2kN/m2 Nc F Figure 1. Extraction of the T-shaped assembly (unit: mm) Both beams and columns in the prototype frame were fabricated using high-strength (grade 161 Q690) steel in China. Both ultimate and serviceability limit states were considered for these 162 members. Regarding the ultimate limit state, beam strength, column strength, and stability were 163 assessed using a factored non-seismic load combination of 1 :3D+1:5L. Additionally, strength 164 and stability requirements were met under a di erent factored seismic load combination of 165 1:2(D+0:5L)+1:3Ehk1, where E hk1represents the design seismic action of the minor earthquake 166 (1st-group, intensity-8 and ground type II), as per the regulations outlined in Chinese Seismic 167 Code [47]. Typically, it is essential to provide su cient bracing for beams in earthquake-resistant 168 frames to prevent lateral-torsional buckling. Therefore, beam instability was not considered in 169 this analysis. For the serviceability limit state, the maximum beam deﬂection was checked 170 against the code-speciﬁed limit of 1 /400 of the beam span under a nominal non-seismic load 171 8 combination of D +L. Furthermore, the maximum story drift angle imposed by the minor 172 earthquake mentioned earlier was required to remain below 1 /250. Consequently, a built-up 173 H-shaped section with dimensions of H280 Ö160Ö8Ö10 was selected for the beams, while two 174 design outcomes with built-up H-shaped sections, namely H220 Ö160Ö12Ö16 and 175 H220Ö160Ö16Ö16, for the columns were achieved. Further, another design was implemented to 176 reinforce the column section H220 Ö160Ö12Ö16 with a 8-mm-thick doubler plate in 177 beam-to-column connections, indicating a total thickness of 20 mm for the panel zone. These 178 sections di er only in terms of the panel zone thickness (including the doubler plate, if 179 appropriate) but all satisfy strong column-weak beam (SCWB) capacity design, thereby enabling 180 the evaluation of the panel zone strength impact. 181 Four specimens were designed for the purpose of this study. The design details are 182 summarized in Table 1. The connection specimens were constructed with CJP groove welds, 183 which e ciently connect the beam ﬂanges to the column ﬂange. Additionally, an erection plate 184 (or shear tab) of the same steel grade and thickness as the beam web was shop-welded to the 185 column ﬂange using ﬁllet welds, and bolted to the beam web to transfer shear force. In each 186 specimen, three class 10.9s M20 high-strength bolts were utilized, with a pretension force of 155 187 kN [50], for the web connection. To ensure structural integrity, continuity plates of matching 188 steel grade, width, and thickness to the beam ﬂanges were incorporated in the investigation. 189 These continuity plates serve two purposes: protecting the column ﬂange and web from local 190 damage and ensuring uniform stress distribution in the beam ﬂanges. In addition to the varying 191 panel zone thickness, two types of CJP groove weld details between the beam and column 192 ﬂanges, as depicted in Figure 2, were analyzed for comparison. The ﬁrst type, labeled as “ b” 193 (Figure 2(a)), represents the current requirement in China (i.e., suggested in both Chinese 194 Seismic Code [47] and Technical Speciﬁcation [45]) and commonly employed in practice. In this 195 type, a reinforcing ﬁllet weld is applied under the bottom backing plate, while the top backing 196 plate remains unreinforced. The weld access hole is machined according to the improved shape 197 9 proposed in FEMA-350 [46] and further suggested in AISC Prequaliﬁed Connections [51]. This 198 hole shape has been shown to alleviate stress concentrations in the transition region between the 199 beam ﬂange and the drilled hole. The second type, labeled as “ c” (Figure 2(b)), is a suggested 200 welded-ﬂange connection detail in AISC Prequaliﬁed Connections [51] after comprehensive 201 investigations by SAC Joint Venture in the United States. The backing plates underneath both 202 beam ﬂanges are removed, and the weld root is backgouged and further reinforced by a ﬁllet 203 weld. In spite of its relatively expensive process compared to Type b, this connection detail can 204 largely improve its deformation performance and the connection rotation capacity can be 205 enhanced substantially. Consequently, the specimen labels shown in Table 1 comprise the beam 206 and column steel grades, followed by the panel zone thickness, and conclude with either borc 207 indicating the beam-to-column ﬂange welded connection detail. 208 Backing bar left in place and reinforced by a fillet62535º 303466200 35º 63310 33 426Backing bar left in place 10.9s M24 boltsShear tab Beam Q355 Column Q690Continuity plate Q355 Continuity plate Q35566 (a) Type b Backing bar removed and backgougedBacking bar removed and backgouged62535º 3034 66200 35º33 10 33 42615 1510.9s M24 boltsShear tab Beam Q355 Column Q690Continuity plate Q355 Continuity plate Q3556666 Backing bar removed and backgougedBacking bar removed and backgouged62535º 3034 66200 35º33 10 33 42615 1510.9s M24 boltsShear tab Beam Q355 Column Q690Continuity plate Q355 Continuity plate Q35566 (b) Type c Figure 2. Connection details The research employed manual gas-shielded metal arc welding (GMAW) process in fabrication 209 by use of matched weld ﬁller materials with Q690 high-strength steel in China. An E761T1-K3C 210 electrode was utilized, as speciﬁed in the recently published Chinese Design Standard for High- 211 10 Table 1. Test specimens Specimen labelBeam section (mm)Column section (mm)Doubler plate (mm)Welding type B690-C690-PZ12 c H280Ö160Ö8Ö10H220Ö160Ö12Ö16 None c B690-C690-PZ16 b H220Ö160Ö16Ö16 None b B690-C690-PZ16 c H220Ö160Ö16Ö16 None c B690-C690-PZ20 c H220Ö160Ö12Ö16 8 c Strength Steel Structures (JGJ /T 483) [52], for all the welding work, including the ﬁllet welds in 212 the welded H-shaped beam and column sections, the CJP groove welds and the ﬁllet welds for 213 reinforcement, the ﬁllet welds attaching the shear tab to the column ﬂange, as well as the welds 214 between continuity plates and the column. This electrode has proof (yield) and ultimate strength 215 of 701 and 800 MPa, respectively, and its average Charpy v-notch toughness is about 57 J under a 216 temperature of20C, according to the mill report. 217 2.2 Material properties 218 All steel plates utilized in the experiment, which have also been used in a previous study on 219 dual-steel moment connections by the authors [21], underwent tensile coupon testing to analyze 220 their stress-strain behavior. These test results were then compared to the requirements outlined 221 in the relevant codes governing high-strength steel plates. This validation process ensured the 222 qualiﬁcation of the steel plates used in this study. For each plate thickness, three coupons were 223 subjected to testing. The coupons employed were of full thickness, possessing a gauge length 224 of 50mm and a width of 20mm at the reduced portion. Several parameters were determined for 225 each plate thickness, including the modulus of elasticity ( E), the yield or proof strength ( fy), the 226 strain at the end of the yield plateau (or at the initiation of the strain hardening phase of the stress- 227 strain curve, st) if applicable, the ultimate strength ( fu), the corresponding ultimate strain ( u), 228 the yield-to-tensile strength ratio ( fy=fu), and the percentage elongation after fracture based on the 229 speciﬁed parallel length ( ) [53]. These values, which represent the averages of three coupons, 230 are summarized in Table 2. Figure 3 illustrates the full-range engineering stress-strain curves of 231 11 the Q690 grade steel coupons, which have been reported in the previous study by the authors 232 [21]. Additionally, simple tensile testing was conducted on the class 10.9s M20 high-strength 233 bolts utilized in the study, providing information regarding the modulus of elasticity ( E), ultimate 234 strength ( fu), and ultimate strain ( u). These results can also be found in Table 2. Table 2. Material properties Steel gradePlate thickness (mm)E (GPa)fy (MPa)st fu (MPa)u fy=fu Q6908 208.3 723 — 822 0.100 0.88 20% 10 189.4 794 — 902 0.106 0.88 21% 12 205.8 775 0.018 816 0.060 0.95 16% 16 219.8 811 0.022 840 0.055 0.97 17% 10.9s M20 206.0 — — 1135 0.110 — — 00.050.10.150.20.250.3 Strain01002003004005006007008009001000Stress (MPa) Q690-PL8 Q690-PL10 Q690-PL12 Q690-PL16 Figure 3. Stress–strain curves of Q690 coupons [21] 235 2.3 Test setup 236 The testing was conducted in the Structures Laboratory of South China University of 237 Technology and it utilized a load frame assembly. This assembly, shown in Figure 4, is comprised 238 of interconnected members that form a planar frame. It serves the purpose of transferring forces 239 from a 300t jack and another MTS hydraulic actuator, both of which were installed within the 240 12 load frame, to the sturdy ﬂoor of the laboratory. The T-shaped specimen was positioned within 241 the load frame and secured at the top and bottom of the column using large load pins. These pins 242 were then connected to the crosshead of the jack at the top and to a substantial beam at the bottom 243 of the load frame, which was tied to the strong ﬂoor. Such arrangement allowed for unrestricted 244 rotation of the column ends during loading, thereby simulating inﬂection points at the midpoint 245 of the columns on each story. The load pins were constructed from thick plate material and solid 246 steel dowels. Loading was applied to the top of the column through the jack, and subsequently to 247 the beam tip by the actuator. This actuator, capable of delivering 300 kN of force with a stroke 248 length of250 mm, was attached to the beam using a pair of end plates connected by 249 high-strength threaded steel rods. The top of the actuator was ﬁrmly fastened to the top beam of 250 the load frame. To prevent any out-of-plane movement of the beam caused by lateral-torsional 251 buckling, a pair of brackets were employed. These brackets were a xed to an additional pair of 252 columns, which were securely fastened to the strong ﬂoor. In order to minimize the impact of 253 potential friction between the brackets and the beam, the brackets were designed with rollers. Jack Pin Δ2N S Δ3 Δ1 Δ5Δ4 PinSpecimen Displacement transducersActuator 3000 Units: [mm]Lateral bracing (a) Scheme [21] (b) On site Figure 4. Test setup 254 13 2.4 Loading protocol 255 Prior to initiating cyclic loading, an axial compression force was applied to the top of the 256 column to represent seismic weight. This resulted in an axial compression ratio of 0.3 in relation 257 to the nominal axial capacity of the column. This ratio remained consistent across all specimens 258 to maintain uniformity. During the cyclic loading phase, the axial load on the column top 259 remained unchanged. The loading history speciﬁed in AISC Seismic Provisions [49] was 260 employed to ensure comparability with numerous other tests conducted during and after the SAC 261 Joint Venture investigations in the United States [48]. This cyclic loading history, based on the 262 story drift angle, deviates from the commonly used approach of employing plastic rotation levels 263 prior to the 1997 Northridge earthquake. The story drift angle is deﬁned as the lateral 264 displacement of the story divided by the height of the story. In this study, the specimens were 265 subjected to displacements at the tip of the beam. Hence, the story drift angle in this context 266 represents the ratio of the displacement at the beam tip to the distance of 3000 mm (equal to L=2, 267 where Lsigniﬁes the column spacing in the prototype frame) between the loading point at the 268 beam tip and the centerline of the column. Figure 5 illustrates the prescribed loading history, with 269 positive story drift angles indicating downward displacements at the beam tip. Since the stroke 270 length of the actuator is 250 mm, the maximum attainable story drift angle is 8%. If no 271 signiﬁcant strength degradation is observed after two cycles at this magnitude, additional cycles 272 at the same amplitude will be conducted until the specimen fails or further substantial strength 273 reduction occurs. 274 2.5 Instrumentation 275 To measure the applied load magnitude at the beam tip and column top, load cells were 276 installed on the actuator and the jack, respectively. Figure 4 illustrates the utilization of 277 displacement transducers to isolate the rotation contributions occurring in speciﬁc components of 278 each specimen’s connection. These components include the shear distortion of the panel zone and 279 14 -10-8 -6 -4 -2 0 2 4 6 8 10 Story drift angle (%)0.375% 0.5%0.75% 1%1.5%2%3%4%5%6%7%8%Figure 5. Loading protocol [49] the plastic hinge rotation in the beam end. 280 The primary displacement transducer, labeled as DT-1, was employed to measure the 281 displacement at the beam tip and served as the displacement-control signal for the actuator. 282 Diagonally arranged displacement transducers, DT-2 and DT-3, were utilized to measure the 283 average shear deformation of the panel zone. This is particularly signiﬁcant for specimens 284 designed with a weak panel zone, as the shear distortion of the panel zone contributes 285 signiﬁcantly to the total story drift angle. Two additional transducers, DT-4 and DT-5, were 286 positioned to monitor the horizontal displacements at the center of pins connected to the top and 287 bottom of the column. Accounting for the possible rigid rotation of the entire specimen, the story 288 drift angle, , was calculated by [21]: 289 =1 L=245 H(1) where 1,4and5correspond to the readings from the respective labeled displacement 290 transducers, Lis the distance explained earlier as 6000 mm, and Hsigniﬁes the story height 291 between the pin centerlines, which is 3000 mm. The shear distortion of the panel zone, denoted 292 15 by pz, was calculated by [21]: 293 pz=23 2q b2 pz+h2 pz bpzhpz(2) where 2and3represent readings from the diagonal transducers, bpzandhpzrepresent the width 294 and height of the panel zone, measured as distances between the column ﬂange centerlines and 295 continuity plate centerlines, respectively. The contribution of the panel zone shear distortion to 296 the displacement at the beam tip, or equivalently, to the story drift angle relative to the column 297 centerline, was determined by [21]: 298 pz= 1hb H! pz (3) where hbdenotes the beam depth. By subtracting the contribution of the panel zone from the total 299 story drift angle, the rotational contribution of the plastic hinge in the beam end, if present, could 300 be evaluated. 301 3 Test results 302 3.1 Failure modes 303 Failure modes of the specimens are summarized in Figure 6. Although the beam section of 304 H280Ö160Ö8Ö10 is classiﬁed as an elastic (or non-compact) section according to Eurocode 3 305 (Class 3) [54] and Chinese Standard (Class S4) [55], based on its ﬂange slenderness, all specimens 306 eventually fractured before signiﬁcant local buckling in the beam end. The only specimen welded 307 with backing plates, Specimen B690-C690-PZ16 b, experienced premature fracture at the bottom 308 CJP groove weld when the story drift angle reached 3:1% in the ﬁrst negative excursion of 4% 309 (see Figure 6(a)). The loading continued and fracture occurred at the top CJP groove weld when 310 the story drift angle approached 3% in the second positive excursion of this amplitude (see Figure 311 16 6(b)). Both fracture phenomena were characterized by the total tear out of the weld metal from 312 the column ﬂange surface. This is a common failure mode for CJP welds with backing plate, since 313 the backing plate hides weld ﬂaws resulting from the weld root pass and provides an initial surface 314 crack depth [48]. 315 In contrast to Specimen B690-C690-PZ16 b, the three specimens with backing plates removed 316 (i.e., with Type cdetail) fractured as well but in the beam ﬂange outside the heat a ected zone 317 near the weld access hole. Apparently, this kind of fracture in the ﬂange steel allowed for larger 318 deformation than the premature weld fracture mentioned above. To be speciﬁc, Specimen 319 B690-C690-PZ16 c, which had an intermediate panel zone thickness among the three specimens, 320 fractured at the story drift angle of -3.4% of the second cycle of 6% amplitude (see Figure 6(c)). 321 Due to the rapid crack growth across the beam ﬂange, the fracture ﬁnished all of a sudden and 322 brought an impact to the web bolts. Hence, the lowest bolt, as shown in Figure 6(c), also 323 fractured in shear. The comparison between Specimens B690-C690-PZ16 cand 324 B690-C690-PZ16 bevidenced the e ectiveness of removing backing plates to improve 325 connection performance by making the fracture mode more ductile. The specimen with a weaker 326 panel zone, B690-C690-PZ12 c, survived until the second cycle of 7%, and fracture in the top 327 beam ﬂange occurred when the story drift angle approached 1% in the positive excursion of that 328 cycle (see Figure 6(c)). On the other hand, the specimen of a stronger panel zone with a doubler 329 plate, B690-C690-PZ20 c, su ered fracture in the beam bottom ﬂange when the story drift angle 330 reached5:5% at the ﬁrst negative excursion of 6% (see Figure 6(e)). The panel zone of this 331 specimen was so strong that plastic demand was expected to be concentrated to the beam end. In 332 addition to the fracture observed, slight local buckling was noted in the beam top ﬂange under 333 compression, as shown in Figure 6(e). 334 3.2 Hysteretic curves 335 Figure 7 displays the hysteretic curves of all specimens, illustrating the relationship between 336 moment and story drift angle. The moment was calculated as the product of the distance between 337 17 Column flange Beam flangeBeam web Weld fracture (a) Specimen B690-C690- PZ16 b(bottom fracture) BeamColumn Panel zoneWeld fracture (b) Specimen B690-C690- PZ16 b(top fracture) Beam web Beam flangeShear tab Column flange Flange fractureBolt fracture(c) Specimen B690-C690- PZ16 c BeamPanel zoneFlange fracture (d) Specimen B690-C690- PZ12 c BeamPanel zoneColumn Flange buckling Flange fracture (e) Specimen B690-C690- PZ20 c Figure 6. Failure modes the loading point at the beam tip and the column face—speciﬁcally, 3000 mm minus half of the 338 column depth—and the reaction force at the beam tip. To provide a basis for comparison, the 339 yielding moment ( My;b) and full plastic moment ( Mp;b) of the beam cross section were determined 340 using the measured material properties presented in Table 2 and plotted on the ﬁgure. 341 Furthermore, Figure 8 presents the moment-shear distortion hysteretic curves for the panel 342 zones in all specimens. The moment is that at the column face mentioned earlier, while the shear 343 distortion was determined using Eq. (2). For comparison, the yielding moment of the panel zone 344 (My;pz), taking into account the inﬂuence of axial compression and column shear, is indicated in 345 18 Figure 8. This yielding moment was calculated using [21]: 346 My;pz=s 1 Nc fycAc!2fycp 3bpzhpztpzLhc LH Hhpz(4) where Ncrepresents the axial compression force on the column, fy;crefers to the material yield 347 strength, hcis the column depth, Acis the cross-sectional area of the column, and tpzcorresponds 348 to the thickness of the panel zone. Note that, the ratio Nc=fycAcwas maintained as 0.3 in all 349 specimens. 350 As anticipated, the hysteretic curves exhibited by all specimens were plump without any 351 obvious pinching, sti ness or strength degradation prior to the ultimate fracture, except for 352 Specimen B690-C690-PZ16 b. Notably, this speciﬁc specimen failed due to premature weld 353 fracture. Consequently, inelastic rotation was hardly developed by this specimen, as indicated by 354 the hysteretic response still within the beam yielding moment capacity (Figure 8(a)). The 355 specimens with intermediate and strong panel zones as well as Type cbacking plate detail, 356 B690-C690-PZ16 cand B690-C690-PZ20 c, exhibited yielding moment capacity of the beam but 357 did not attain its full plastic capacity (Figure 8(b) and 8(d)). This is also under expectation 358 because of the non-compact beam section used in this study. However, in Specimen 359 B690-C690-PZ12 c, because of the weak panel zone, the ultimate moment could not reach the 360 beam yielding capacity (Figure 8(c)). The panel zone of this specimen yielded before the beam. 361 With regard to the responses of the panel zones, the specimen with the weakest panel zone, 362 B690-C690-PZ12 c, developed a plump and full hysteretic curve. Notable agreement with the 363 yielding moment predicted by Eq. (4) was observed in this specimen, as depicted in Figure 8(c). 364 It is intriguing to observe that the Q690 high-strength steel panel zone demonstrated remarkable 365 inelastic shear distortions, as evidenced by this specimen. Noteworthy plastic hardening was 366 additionally observed within this panel zone, attributable to the strain hardening of the material 367 and the involvement of the column ﬂanges in shear transfer subsequent to the initial onset of 368 19 -8 -6 -4 -2 0 2 4 6 8 Story drift angle (%)-500-2500250500Moment (kNm)M p,b=441kNm -M p,b=-441kNmM y,b=397kNm -M y,b=-397kNm(a) Specimen B690-C690-PZ16 b -8 -6 -4 -2 0 2 4 6 8 Story drift angle (%)-500-2500250500Moment (kNm)M p,b=441kNm -M p,b=-441kNmM y,b=397kNm -M y,b=-397kNm (b) Specimen B690-C690-PZ16 c -8 -6 -4 -2 0 2 4 6 8 Story drift angle (%)-500-2500250500Moment (kNm)M p,b=441kNm -M p,b=-441kNmM y,b=397kNm -M y,b=-397kNm (c) Specimen B690-C690-PZ12 c -8 -6 -4 -2 0 2 4 6 8 Story drift angle (%)-500-2500250500Moment (kNm)M p,b=441kNm -M p,b=-441kNmM y,b=397kNm -M y,b=-397kNm (d) Specimen B690-C690-PZ20 c Figure 7. Moment–story drift angle hysteretic curves 20 -5 -3 -1 1 3 5 Shear distortion (%)-500-2500250500Moment (kNm)M y,pz=424kNm -M y,pz=-424kNm(a) Specimen B690-C690-PZ16 b -5 -3 -1 1 3 5 Shear distortion (%)-500-2500250500Moment (kNm)M y,pz=424kNm -M y,pz=-424kNm (b) Specimen B690-C690-PZ16 c -5 -3 -1 1 3 5 Shear distortion (%)-500-2500250500Moment (kNm)M y,pz=303kNm -M y,pz=-303kNm (c) Specimen B690-C690-PZ12 c -5 -3 -1 1 3 5 Shear distortion (%)-500-2500250500Moment (kNm) M y,pz=490kNm-M y,pz=-490kNm (d) Specimen B690-C690-PZ20 c Figure 8. Moment–shear distortion hysteretic curves 21 yielding in the web panel. This particular test provides experimental support for the reliability of 369 energy dissipation through panel zone yielding, even when employing 690 MPa high-strength 370 steel. It should be noted that the maximum inelastic shear distortion that was achieved by 371 Specimen B690-C690-PZ12 cis about 0.03 rad, but this is not the real deformation capacity of the 372 panel zone since no shear failure was observed. In fact, the previous testing on dual-steel 373 connections showed that the 690 MPa steel panel zone could sustain as large a plastic shear 374 distortion as nearly 0.04 rad [21]. The other specimens with intermediate or strong panel zones 375 exhibited basically elastic responses in their panel zones. 376 3.3 Strength, deformation and energy dissipation capacities 377 The determination of the elastic sti ness, denoted as Ke, was based on the hysteretic curves 378 represented in Figure 7. Linear regression was employed to ﬁt the data within a 2% amplitude 379 range, and the resulting values are presented in Table 3. It is important to highlight that Ke380 represents the sti ness of the entire beam-to-column assembly, rather than the rotational sti ness 381 of the connection itself. The ultimate (or maximum) moment, denoted as Mu, was determined by 382 averaging maximum positive and negative moments and is also summarized in Table 3. 383 Specimens featuring panel zones with thicknesses of 16 mm or 20 mm demonstrated robust panel 384 zone design, as their My;pzvalues surpassed the Muvalues. In contrast, the specimen with a 385 12-mm-thick panel zone, B690-C690-PZ12 c, exhibited a lower My;pzvalue compared to My;b, 386 indicating that its performance was primarily inﬂuenced by the panel zone. The level of 387 overstrength, deﬁned as the ratio of the ultimate moment Muto the yield moment My(the 388 minimum between My;bandMy;pz), is included in Table 3. With the exception of the specimen 389 featuring Type bconnection detail (B690-C690-PZ16 b), which experienced premature fracture, 390 the other specimens with Type cdetail demonstrated maximum moments ( Mu) exceeding the 391 yield moment ( My), although this overstrength remained within 10%. This suggests that some 392 degree of inelastic rotation should be anticipated at the beam end or in the panel zone. The 393 specimen with the weakest panel zone (B690-C690-PZ12 c) exhibited moderate strain hardening, 394 22 as indicated by its Mu=My(orMu=My;pz) ratio of 1.1, slightly higher than that of the other 395 specimens with stronger panel zones (B690-C690-PZ16 cand B690-C690-PZ20 c). It is important 396 to note that the ultimate moments ( Mu) observed in all specimens were below the full plastic 397 moments of the beam ( Mp;b), due to the utilization of non-compact beam sections in these 398 specimens. 399 Table 3. Sti ness and strength Specimen label Ke (kNm)Mp;b (kNm)My;b (kNm)My;pz (kNm)My (kNm)Mu (kNm)Mu MyMu Mp;b B690-C690-PZ16 b 9933 441 397424 397323 0.81 0.73 B690-C690-PZ16 c 9987 404 1.02 0.92 B690-C690-PZ12 c 9469 303 303 335 1.10 0.76 B690-C690-PZ20 c 9442 490 397 429 1.08 0.97 Table 4 provides a summary of various parameters pertaining to deformation and energy 400 dissipation capacities. The yield story drift angle, denoted as y, was determined by calculating 401 the ratio of the yield moment, My, to the elastic sti ness, Ke. The ultimate story drift angle, u, 402 was considered valid only if at least one complete cycle of the target story drift angle was 403 achieved prior to fracture [48]. The plastic story drift angle, p, which represents the plastic 404 rotation relative to the column centerline, was determined as the plastic component of u. 405 Additionally, Table 4 presents the plastic rotation component in the beam, p;b, as well as in the 406 panel zone, p;pz. The plastic rotation of the panel zone, p;pz, was deduced from the total rotation 407 of the panel zone, pz, deﬁned in Eq. (3). On the other hand, the plastic rotation of the beam, p;b, 408 was obtained by subtracting p;pzfromp. Consequently, the ductility factor, , was computed as 409 the ratio of utoy. Furthermore, cumulative plastic rotations, p, and energy dissipation, A, 410 for the entire set of specimens were evaluated using the approach described in ATC-24 [56, 57] 411 and depicted in Figure 9. It should be noted that these quantities were considered up until the last 412 successful excursion (or half-cycle) prior to fracture, and their normalized values were also 413 included in Table 4, denoted as p=yandA= Myy , respectively. 414 23 RotationMomenti-th excursion i+1-th excursion" 3piAi " 3pi+1 Ai+1Figure 9. Deﬁnition of cumulative plastic rotation and energy dissipation in ATC-24 [56] Table 4. Deformation and energy dissipation Specimen label y (rad)u (rad)  p (rad)p;pz (rad)p;b (rad)pp yA (kJ)A My;miny B690-C690-PZ16 b0.040 0.03 0.8 0.002 0.001 0.001 0.028 0.7 6.0 0.4 B690-C690-PZ16 c0.040 0.06 1.5 0.021 0.002 0.019 0.252 6.3 86.6 5.3 B690-C690-PZ12 c0.032 0.07 2.2 0.035 0.030 0.005 0.515 16.1 146.1 14.5 B690-C690-PZ20 c0.042 0.05 1.2 0.010 0.002 0.008 0.142 3.4 49.1 2.8 It is evident that, with the exception of Specimen B690-C690-PZ16 b, all specimens displayed 415 substantial story drift angles and achieved satisfactory performance, considering a story drift 416 angle of 0.04 rad as an acceptable criterion for Special Moment Frames (SMF) according to 417 AISC Seismic Provisions [49]. However, due to the utilization of high-strength steel materials 418 and relatively small beam and column sections in this study, the yield drift angles of these 419 specimens reached nearly 0.04 rad, approximately four times the traditionally expected yield drift 420 angle (around 0.01 rad) for moment connections made of conventional-strength steel. 421 Consequently, if a plastic rotation of 0.03 rad is deemed necessary for SMF, only the specimen 422 featuring the weakest panel zone and Type cdetail (B690-C690-PZ12 c) is considered to possess 423 sucient rotation capacity. The other two specimens with stronger panel zones and Type cdetail 424 (B690-C690-PZ16 cand B690-C690-PZ20 c) demonstrated plastic rotations ranging from 0.01 to 425 24 0.02 rad, which are only suitable for Intermediate Moment Frames (IMF) [49]. 426 Remarkably, when comparing the three specimens with Type cdetail but varying panel zone 427 thicknesses, it is apparent that the weak panel zone contributes to an increased overall plastic 428 rotation capacity. This is noteworthy because it has conventionally been believed that excessive 429 deformation in the panel zone adversely a ects rotation capacity due to the formation of a “kink” 430 in the column ﬂanges, leading to signiﬁcant local stress or strain concentrations near the ﬂange 431 welds. However, this adverse e ect was not observed in the present study. The Q690-grade panel 432 zone in this study sustained a plastic shear rotation as large as approximately 0.03 rad, and there 433 was no failure in the panel zone even at this magnitude, as demonstrated by Specimen B690- 434 C690-PZ12 c. This shear rotation quantity aligns with a previous test conducted by the authors 435 [21], which revealed an ultimate plastic rotation capacity of around 0.04 rad for a Q690 panel 436 zone of similar size to the one investigated in this study. 437 Additionally, the cumulative quantities indicate a more ductile behavior in the panel zone 438 compared to the beam end. In the case of plastic shear in the panel zone, the cumulative plastic 439 rotation is 16 times greater than the yield rotation, whereas for plastic bending in the beam end, it 440 is only 3 to 6 times larger. A similar comparison can be made regarding the cumulative energy 441 dissipation. Normalizing the cumulative plastic rotation based on the maximum plastic rotation, 442 the normalized values range from 12 to 15 for the three specimens with Type cweld detail. These 443 seismic capacities exceed the conventionally anticipated seismic demand, which suggests that the 444 cumulative plastic deformation should be 5 to 8 times larger than the maximum plastic 445 deformation [57, 58]. 446 4 Discussion 447 4.1 CJP groove weld details 448 In current Chinese seismic design codes pertaining to steel structures [45, 47, 55], the 449 welding detail denoted as Type bin this paper is prescribed for connection of beam and column 450 25 ﬂanges by means of CJP welds. In contrast, the United States imposes more stringent weld detail 451 requirements for ductile moment frames. For instance, AISC Prequaliﬁed Connections [51] for 452 SMF or IMF recommend a welded ﬂange-welded web connection. In this connection, a CJP weld 453 with backing plate that is reinforced by a ﬁllet weld is executed at the upper beam-to-column 454 ﬂange connection, while a CJP weld without a backing plate or one that is removed after the 455 welding process is speciﬁed for the lower connection. 456 Nonetheless, the ﬁndings of the current investigation demonstrate that Type bweld detail fails 457 to ensure ductile performance, as evidenced by premature fracture observed in Specimen B690- 458 C690-PZ16 bat their CJP welds. Moreover, it is worth noting that even the AISC-prescribed 459 weld details for welded ﬂange-welded web connections may prove insu cient, given that the CJP 460 weld with a backing plate, although reinforced by a ﬁllet weld at the bottom beam ﬂange, still 461 experienced fracture in the above specimen. In contrast, the specimens employing Type cdetail 462 exhibited relatively ductile behavior, characterized by fracture occurring in the beam ﬂange. 463 In fact, a prior study conducted by the authors [21] had previously a rmed the reliability of 464 employing Type bweld detail in dual-steel connections, speciﬁcally for connecting 465 conventional-strength steel beam ﬂanges to a high-strength steel column. This disparity in 466 performance is likely attributable to the diminished material ductility inherent in Q690 467 high-strength steel when compared to its conventional-strength counterparts. Consequently, it is 468 concluded that in the case of Q690 high-strength steel welded ﬂange-bolted web connections, the 469 CJP welds should be executed by removing backing plates, performing backgouging, and further 470 reinforcing the weld root by a ﬁllet weld. While this weld detail, denoted as Type cin this study, 471 is considerably costlier than Type b, its adoption is imperative to avert the risk of brittle fracture 472 in high-strength steel connections, as elucidated in this research. 473 In the context of continuity plate welds, this study has provided noteworthy insights. Current 474 Chinese design codes and AISC Seismic Provisions commonly stipulate the use of CJP groove 475 welds in the connection between continuity plates and column ﬂanges. However, they permit the 476 26 utilization of ﬁllet welds between continuity plates and the column web. In this investigation, 477 an unintentional use of ﬁllet welds was in all specimens for continuity plate-to-column ﬂange 478 connections. Interestingly, these ﬁllet welds demonstrated commendable performance, provided 479 they were of su cient size. This observation, along with the previous test results by the authors 480 [21], underscores the advantage of employing cost-e ective ﬁllet welds rather than more expensive 481 CJP groove welds even in case of high-strength steel, a beneﬁt that has also been substantiated in 482 conventional-strength steel welded ﬂange-welded web connections [59, 60]. 483 4.2 Balanced design between the beam and panel zone 484 A major focus in this study is to evaluate the e ect of panel zone strength on connection 485 performance. Hence, the existing panel zone design methodologies were compared. Capacity-to- 486 demand ratios for the panel zones in all specimens have been consolidated and presented in Table 487 5. These ratios elucidate the degree to which the panel zones can withstand the speciﬁed demand 488 levels, as stipulated in the design codes originating from China, the United States, and Europe, in 489 accordance with the following expressions [21]: 490 4 3fy;pzp 3hpzbpztpz pzWy;bfy;b (5) 0:6fy;pz0:95hctpzhb 1+3bct2 fc hbhctpz! min fy;b+fu;b 2fy;b;1:2! Wy;bfy;b (6) (0:9)0:55fy;pzhctpzWy;bfy;b hb L Lhc! Hhb H! (7a) (0:6)0:55fy;pzhctpzWy;bfy;b hb L Lhc! Hhb H! (7b) 27 0:9fy;pzp 3(hc2tfc)hpztpz +2Mp;fc+min 2Mp;fc;2Mp;st Wy;bfy;b (8) Eq. (5) is from Chinese Standard [55], Eq. (6) is from AISC Seismic Provisions [49], Eqs. (7) 491 denote lower and upper bounds for the panel zone resistance suggested by FEMA [48], and Eq. 492 (8) is from Eurocode 8 [61]. In the above equations, fy;pzis the yield strength of the panel zone 493 material; fy;band fu;bare yield and tensile strength of the beam material, respectively; Wy;bis 494 the elastic section modulus of the beam; bcandtfcare the width and thickness of the column 495 ﬂange, respectively; Mp;fc=bct2 fcfy;cf=4 and Mp;st=bct2 stfy;st=4 are plastic moment resistances of 496 the column ﬂange and a pair of continuity plates, respectively, where tstandfy;stare the thickness 497 and material yield strength of the continuity plates, respectively; pzis a coe cient taken as 0.95 498 for one-sided connections; is another coe cient called transformation parameter to account for 499 the e ect of column shears and it should be determined based on the internal force equilibrium. 500 Note that, as the beam sections are non-compact in this study, the elastic section modulus ( Wy;b) 501 rather than the plastic one ( Wp;b) is used to estimate the beam strength. 502 Evidently, the panel zones exhibit a varying degree of capacity depending on the design code 503 employed, with the weakest panel zone, characterized by the largest capacity-to-demand ratio, 504 emerging under Chinese Standard design practices. Conversely, the most robust panel zone, 505 denoted by the smallest ratio, is observed when adhering to Eurocode 8, as has been corroborated 506 in earlier investigations [21, 35]. It is noteworthy that the panel strength prescribed by Eurocode 507 8 aligns closely with the bounds deﬁned by FEMA. In contrast, the panel strength requirements 508 stipulated by Chinese and AISC codes fall below the lower FEMA bound. 509 All specimens met the panel zone strength criteria outlined in Chinese Standard, as evidenced 510 by their capacity-to-demand ratios exceeding 1. Similarly, compliance with AISC Seismic 511 Provisions was also achieved across all specimens. However, only Specimen B690-C690-PZ12 c 512 failed to meet the panel strength criterion in Eurocode 8, as indicated by its ratio less than 1. 513 28 Interestingly, this particular specimen exhibited superior plastic rotation and energy dissipation 514 capacities. These ﬁndings suggest that the criterion in Eurocode 8 for minimum panel zone 515 strength may warrant reconsideration and relaxation in high-strength steel connections, since the 516 advantageous role played by high-strength steel panel zones in dissipating energy is a rmed 517 through a direct comparison of their plastic rotation capacity (exceeding 0.03 rad) with that of the 518 beam end (ranging from 0.01 to 0.02 rad), as documented in Table 4. Table 5. Capacity-to-demand ratios of panel zones Specimen labelChinese AISC Seismic FEMAEurocode 8 Standard Provisions Lower bound Upper bound B690-C690-PZ12 c 1.04 1.00 0.76 0.51 0.75 B690-C690-PZ16 b1.46 1.34 1.06 0.71 1.02B690-C690-PZ16 c B690-C690-PZ20 c 1.69 1.52 1.23 0.82 1.17 519 5 Conclusions 520 This paper reports on an experimental investigation into the cyclic response of Q690-grade 521 high-strength steel beam-to-column connections, speciﬁcally focusing on welded ﬂange-bolted 522 web connections. A set of four one-sided connection specimens was tested, with the primary 523 objective of assessing the impact of beam-to-column ﬂange weld details and panel zone strength. 524 The study yields the following conclusions: 525 1) The connection using backing plates and reinforced with a ﬁllet weld underneath the bottom 526 beam ﬂange, exhibited brittle fracture at CJP welds without development of any plastic 527 rotation. The other connections with all backing plates removed and CJP weld roots 528 backgouged and further reinforced by ﬁllet welds, exhibited ductile fracture at beam ﬂanges. 529 Hence, the weld detail not using backing plates or removing them after the welding process is 530 strongly recommended for 690 MPa high-strength steel welded ﬂange-bolted web 531 connections. 532 29 2) The strong-panel-zone connections without backing plates developed very slight strain 533 hardening, and their maximum moment resistances were very close to the beam yielding 534 moment. These connections sustained plastic rotations of 0.01–0.02 rad in the beam end, and 535 cumulative plastic rotations of about 3–6 times larger than the yielding rotation. 536 3) The weak-panel-zone connection without backing plates exhibited slight strain hardening by 537 developing a 10% larger maximum moment resistance than the yielding moment of the panel 538 zone. This connection achieved a plastic rotation of 0.03 rad in the panel zone, and a cumulative 539 plastic rotation of about 16 times larger than the yielding rotation, until fracture in the beam 540 end. Yielding of the panel zone did not show adverse e ect on the total plastic rotation capacity 541 of the connection. 542 It should be noted that, in this experimental study, the connections made with relatively 543 shallow beams (280 mm in depth) were evaluated. Further study on high-strength steel moment 544 connections with larger size beams, especially those with deeper beam sections, is needed. 545 Acknowledgments 546 The authors would like to acknowledge the ﬁnancial supports for this work by the National 547 Natural Science Foundation of China (Grant Nos. 51638009, 51978279 and 52108145), 548 Guangdong Basic and Applied Basic Research Foundation (Grant Nos. 2021A1515010610 and 549 2023A1515010047) and Fundamental Research Funds for the Central Universities (Grant No. 550 2023ZYGXZR098). 551