# Performance of a Subgrade-Embankment-Seawall System Reinforced by Drainage PCC Piles and Ordinary Piles Subjected to Lateral Spreading

Authors: Zhixiong Chen; Buxin Wang; Chenglong Wang; Yonghong Wang; Peng Xiao; Kangyin Li
Journal: Geofluids
 (2023)

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## Abstract

Coral sand is widely distributed in coastal areas and used as a soil foundation in an increasing number of projects. Historical records show that the liquefaction and lateral spreading of coral sand foundations occurred many times during earthquakes, resulting in serious geological and engineering disasters. Based on a rigid drainage pile, a PCC pile (Large Diameter Pipe Pile by using Cast-in-place Concrete) and drainage body are innovatively combined into a new drainage PCC pile, to reduce the harm caused by liquefaction and lateral spreading of the coral sand foundation. In this study, the seismic response of the subgrade-embankment-seawall system on the coral sand liquefaction site was simulated using a shaking table. The excess pore water pressure ratio, acceleration, average foundation settlement, seawall displacement, embankment deformation, and pile body bending moment of the ordinary pile foundation and new drainage PCC pile foundation under seismic loads ofPGApeakgroundacceleration=0.05g, PGA=0.1g, and PGA=0.2g on the coral sand site were analyzed and compared. Compared with ordinary pile foundations, the excess pore water pressure ratio, pile bending moment, seawall displacement, foundation settlement, and embankment deformation of the new drainage PCC pile foundation were markedly reduced, while acceleration increased, which showed that the new drainage PCC pile had a positive anti-liquefaction effect on the coral sand foundation.


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## Body

## 1. Introduction

Coral sand is the remains of biological corals after death in the ocean, and it is widely distributed across the sea between 30 degrees of south and north latitude. There have also been many liquefaction earthquakes on coral reef soil sites throughout history. Many studies have shown that the lateral spreading of liquefaction is the primary cause of structural damage during the earthquake [1–9]. For example, the earthquake with a magnitude of 8.1 in Guam in 1993 led to the liquefaction of hydraulic fill coral sand in the coastal area of APRA port [10]. The liquefaction led to hundreds of meters of ground fissures in the naval base, lateral sliding, and uneven settlement, resulting in the damage of more than half of the buildings on the island. The earthquake with a magnitude of 7.0 in Haiti in 2010 led to serious liquefaction of the international port of Port-au-Prince [11], resulting in sand jetting, water gushing of the foundation soil, and serious expansion. These processes caused serious damage to the road, wharf pile foundation, and surrounding buildings, inducing serious economic losses.The lateral spreading of liquefied foundation soil is primarily caused by a failure to eliminate the excess pore water pressure of foundation soil sufficiently quickly. Therefore, scholars have performed many experimental studies to develop drainage piles to reduce excess pore water pressure. A sand compaction pile was the first drainage pile used to treat liquefaction sites. Ohbayashi et al. [12] analyzed a lot of foundation property data on an earthquake site and concluded that a sand compaction pile could improve the liquefaction resistance of the foundation. Harada et al. [13] proposed a layer ring structure for a steel pipe pile that is composed of multilayer rings. During an earthquake, the excess pore water can enter the drainage device through the gap between the rings and thus be drained out of the soil. Many studies have shown that drainage piles have good liquefaction resistance, but most studies have investigated flexible piles, and the problem of insufficient bearing capacity remains poorly understood.Liu et al. [14] proposed rigid drainage pile technology, which can consider the bearing capacity and drainage function. Chen et al. [15] studied the liquefaction resistance of a single rigid drainage pile using a small-scale shaking table test. Yang et al. [16] also performed an experimental study on the liquefaction resistance of a rigid drainage pile group. Results show that the rigid drainage pile can effectively dissipate the excess pore water pressure of the soil around the pile, maintain the effective stress of the soil, and markedly reduce the possibility of liquefaction, regardless of if a single or group pile is present.However, these experimental studies primarily focused on the pore pressure response in the soil, and the study of the liquefaction resistance of rigid drainage piles on the lateral spreading site of coral sand foundations is limited. In addition, there is a lack of research on the subgrade-embankment-seawall system.According to previous experience [17–25], a shaking table model can accurately reproduce seismic phenomena. Therefore, in this study, a vertical subgrade-embankment-seawall system on a lateral spreading site of coral sand foundations was simulated, and the seismic responses of the subgrade-embankment-seawall system reinforced by the new drainage PCC pile and the ordinary pile were compared through a series of shaking table tests. To simulate earthquakes of different intensities, three amplitudes of seismic waves were applied: small earthquakes (PGA=0.05g), moderate earthquakes (PGA=0.1g), and large earthquakes (PGA=0.2g).


## 2. Test Setup and Model Preparation

### 2.1. Shaking Table Facility

The test was performed in the shaking table laboratory of the geotechnical experiment building of Chongqing University, China. The power unit uses the American ANCO small shaking table, and its primary parameters are shown in Table1. Based on the results of many previous studies and design experience [26–31], a laminar shear box was used in this test to effectively reduce the influence of the boundary effect. Its length, width, and depth were 0.95 m, 0.85 m, and 0.75 m, respectively.Table 1
Parameters of the shaking table system.
ParametersFigurePlatform size1.2m×1.2mMaximum load1000 kgMaximum horizontal acceleration2.0 gThe range of frequency0~50 HzMaximum horizontal speed0.5 m/sMaximum horizontal displacement100 mmAccording to the previous studies [32–34], the Bockingham π theorem is used in this test to design the scale model. The similarity ratio of the elastic modulus is SE=0.05, and the geometric similarity ratio is SL=0.05. Because actual engineering sand (coral sand) is used for the test, the density similarity ratio is Sρ=1. The similarity ratio of other parameters can be calculated according to the Bockingham π theorem, as shown in Table 2.Table 2
Similitude ratios.
ParametersPhysical symbolsConversion formulaSimilarity ratioLengthLSL0.05DensityρSρ1Elasticity modulusESE0.05AccelerationaSa=SESL‐1Sρ−11FrequencyfSf=SE0.5SL−1Sρ−0.54.472QualitymSm=SL3Sρ1.25×10−4StrengthFSF=SL2SE1.25×10−4Normal stressσSσ=SE0.05TimeTST=SLSE−0.5Sρ0.50.233Cross sectionASA=SL22.5×10−3Inertia momentISI=SL46.25×10−6


### 2.2. Model Pile and Subgrade-Embankment-Seawall System

The test model is made of plexiglass. In the test, the length of the PCC model pile is 600 mm. The outer diameter and inner diameter are 50 mm and 36 mm, respectively. Different from ordinary piles, a drainage filter was fixed on both sides of the drainage pile, as shown in Figure1.Figure 1
Schematic diagram of the pile: (a) picture of the model pile and (b) cross section.
(a)(b)The test model was a vertical subgrade-embankment-seawall system. The subgrade was set as a horizontal site with a height of 0.6 m, an ocean depth of 0.2 m, and a seawall height of 0.7 m. The seawall in the test group was composed of PCC pipe piles and plexiglass plates, while the seawall in the control group was composed of 35 mm diameter solid piles and plexiglass plates.A base was designed to simulate the bearing layer, and the subgrade pile and seawall pile were placed first in the base. Then, they were placed on the impermeable plastic film and put into the laminar shear box together. After the base was installed, the foundation pile group and seawall pile were fixed in the corresponding base, as shown in Figure2. The relative position between piles was fixed temporarily with a fixed bracket so that the piles would not have a large relative displacement when preparing the foundation soil using the pluviation deposition method.Figure 2
Schematic diagram of the model box.The coral sand is used as the raw material to prepare the embankment. The thickness of each layer of the embankment model was 30 mm, and reinforced geogrids with a tensile strength of approximately 15 kN/m were used to wrap it. The embankment layout direction was perpendicular to the load vibration direction, and the embankment center coincided with the pile group center. The upper and bottom widths of the embankment were 220 mm and 460 mm, respectively. The height was 120 mm, and the slope of the embankment was 45o, as shown in Figure 3.Figure 3
Picture of the model box after soil preparation.


### 2.3. Foundation Preparation

The coral sand used in this test was taken from an island reef. The specific physical parameters of coral sand and standard sand are shown in Table3. The grain size distribution curves of the two types of sand are shown in Figure 4. The particle content of coral sand with particle diameters larger than or equal to 0.25 mm exceeded 50% of the total mass, thus categorizing this sand as medium coarse.Table 3
Physical properties of the sand.
The types of sandD50/mmCuGSρd, max/(g·cm-3)ρd, min/(g·cm-3)Coral sand0.482.672.81.481.15Standard sand0.604.52.631.641.35Figure 4
Grain size distribution curves of coral sand and standard sand.The laminar shear box was divided into two parts along the vibration direction using a wooden partition. The left part was used to prepare the new drainage PCC pile model foundation (the test group, case 1), and the right part was used to prepare the conventional pile model foundation (the control group, case 2), as shown in Figure3. The preparation process was as follows. In the first step, the partition was fixed on the laminar shear box before the experiment so that the laminar shear box was divided into two parts. In the second step, the inner part of the laminar shear box was wrapped with plastic film that had a thickness of 0.12 mm because the laminar shear box can be permeable to water and was marked every 5 cm in height to facilitate the subsequent preparation of foundation soil by the pluviation deposition method. In the third step, the fixed base was assembled and placed at the bottom of the laminar shear box, and the foundation pile group and the seawall pile were welded and fixed in the base successively. Then, the position of the pile group was fixed with a temporary fixing bracket. In the fourth step, the foundation soil was prepared by the pluviation deposition method. According to the height mark of the plastic film, water with a depth of 5 cm was poured in and backfilled with sand each time [35]. The operation process was repeated, and accelerometers and pore water pressure transducers were placed at the design heights. In the fifth step, pebbles were scattered across the top 5 cm of the test group as a transverse drainage channel. In the sixth step, the embankment was prepared. The preparation process of the embankment is shown in Section 2.2. The model foundations after the above six steps are shown in Figure 3.


### 2.4. Layout of the Sensors

In the model foundation, the sensor layout positions of the test group and the control group were the same, and a schematic diagram of sensors layout is shown in Figures5(a)–5(c). To determine the bending moment response of the piles, strain gauges were symmetrically arranged on piles 1, 2, 3, 4, 5, 6, 7, and 8. The accelerometers were arranged in three layers. In order to explore the response of excess pore water pressure at different depths, the excess pore water pressure transducers of the test group and the control group were arranged in three layers, and three excess pore water pressure transducers were placed in each layer. Four displacement transducers were used to collect the displacement data of the embankment and seawall, and were fixed on the shaking table through a rigid bracket.Figure 5
Instrument layout: (a) plan view, (b) 1-1 cross section, and (c) 2-2 cross section.
(a)(b)(c)


### 2.5. Test Conditions and Loading Scheme

The test conditions are shown in Table4. The loading scheme included three stages: PGA=0.05g (first stage), PGA=0.1g (second stage), and PGA=0.2g (third stage). In the first stage, to explore the influences of different seismic waves on the liquefaction resistance of piles, white noise, Northbridge (far-field), Loma (far-field), ChiChi (far-field), SanFernando (far-field), Loma (near-field without pulse), Northbridge (near-field without pulse), Loma (near-field with pulse), Northbridge (near-field with pulse), and sinusoidal wave (5 Hz) were input subsequently with seismic intensities of PGA=0.05g. The loading sequences in the second and third stage were the same to that in the first stage. The variation of a sinusoidal wave is simpler than the other seismic waves, which is more convenient to analyze the acceleration and excess pore water pressure in experimental research. Therefore, many scholars [36–38] adapted input sinusoidal waves in shaking table experiments. This test will compare and analyze the seismic response of the test group and the control group under sinusoidal wave excitation with seismic intensities of PGA=0.05g, PGA=0.1g, and PGA=0.2g. Figure 6 shows the acceleration time histories of sinusoidal waves with different seismic intensities.Table 4
Test conditions.
CaseType of pileRelative compactnessGauge heightTypes of sand1Rigid drainage pile70%570 mmCoral sand2Ordinary pile70%570 mmCoral sandFigure 6
Time-history curves of the input acceleration: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)


## 2.1. Shaking Table Facility

The test was performed in the shaking table laboratory of the geotechnical experiment building of Chongqing University, China. The power unit uses the American ANCO small shaking table, and its primary parameters are shown in Table1. Based on the results of many previous studies and design experience [26–31], a laminar shear box was used in this test to effectively reduce the influence of the boundary effect. Its length, width, and depth were 0.95 m, 0.85 m, and 0.75 m, respectively.Table 1
Parameters of the shaking table system.
ParametersFigurePlatform size1.2m×1.2mMaximum load1000 kgMaximum horizontal acceleration2.0 gThe range of frequency0~50 HzMaximum horizontal speed0.5 m/sMaximum horizontal displacement100 mmAccording to the previous studies [32–34], the Bockingham π theorem is used in this test to design the scale model. The similarity ratio of the elastic modulus is SE=0.05, and the geometric similarity ratio is SL=0.05. Because actual engineering sand (coral sand) is used for the test, the density similarity ratio is Sρ=1. The similarity ratio of other parameters can be calculated according to the Bockingham π theorem, as shown in Table 2.Table 2
Similitude ratios.
ParametersPhysical symbolsConversion formulaSimilarity ratioLengthLSL0.05DensityρSρ1Elasticity modulusESE0.05AccelerationaSa=SESL‐1Sρ−11FrequencyfSf=SE0.5SL−1Sρ−0.54.472QualitymSm=SL3Sρ1.25×10−4StrengthFSF=SL2SE1.25×10−4Normal stressσSσ=SE0.05TimeTST=SLSE−0.5Sρ0.50.233Cross sectionASA=SL22.5×10−3Inertia momentISI=SL46.25×10−6


## 2.2. Model Pile and Subgrade-Embankment-Seawall System

The test model is made of plexiglass. In the test, the length of the PCC model pile is 600 mm. The outer diameter and inner diameter are 50 mm and 36 mm, respectively. Different from ordinary piles, a drainage filter was fixed on both sides of the drainage pile, as shown in Figure1.Figure 1
Schematic diagram of the pile: (a) picture of the model pile and (b) cross section.
(a)(b)The test model was a vertical subgrade-embankment-seawall system. The subgrade was set as a horizontal site with a height of 0.6 m, an ocean depth of 0.2 m, and a seawall height of 0.7 m. The seawall in the test group was composed of PCC pipe piles and plexiglass plates, while the seawall in the control group was composed of 35 mm diameter solid piles and plexiglass plates.A base was designed to simulate the bearing layer, and the subgrade pile and seawall pile were placed first in the base. Then, they were placed on the impermeable plastic film and put into the laminar shear box together. After the base was installed, the foundation pile group and seawall pile were fixed in the corresponding base, as shown in Figure2. The relative position between piles was fixed temporarily with a fixed bracket so that the piles would not have a large relative displacement when preparing the foundation soil using the pluviation deposition method.Figure 2
Schematic diagram of the model box.The coral sand is used as the raw material to prepare the embankment. The thickness of each layer of the embankment model was 30 mm, and reinforced geogrids with a tensile strength of approximately 15 kN/m were used to wrap it. The embankment layout direction was perpendicular to the load vibration direction, and the embankment center coincided with the pile group center. The upper and bottom widths of the embankment were 220 mm and 460 mm, respectively. The height was 120 mm, and the slope of the embankment was 45o, as shown in Figure 3.Figure 3
Picture of the model box after soil preparation.


## 2.3. Foundation Preparation

The coral sand used in this test was taken from an island reef. The specific physical parameters of coral sand and standard sand are shown in Table3. The grain size distribution curves of the two types of sand are shown in Figure 4. The particle content of coral sand with particle diameters larger than or equal to 0.25 mm exceeded 50% of the total mass, thus categorizing this sand as medium coarse.Table 3
Physical properties of the sand.
The types of sandD50/mmCuGSρd, max/(g·cm-3)ρd, min/(g·cm-3)Coral sand0.482.672.81.481.15Standard sand0.604.52.631.641.35Figure 4
Grain size distribution curves of coral sand and standard sand.The laminar shear box was divided into two parts along the vibration direction using a wooden partition. The left part was used to prepare the new drainage PCC pile model foundation (the test group, case 1), and the right part was used to prepare the conventional pile model foundation (the control group, case 2), as shown in Figure3. The preparation process was as follows. In the first step, the partition was fixed on the laminar shear box before the experiment so that the laminar shear box was divided into two parts. In the second step, the inner part of the laminar shear box was wrapped with plastic film that had a thickness of 0.12 mm because the laminar shear box can be permeable to water and was marked every 5 cm in height to facilitate the subsequent preparation of foundation soil by the pluviation deposition method. In the third step, the fixed base was assembled and placed at the bottom of the laminar shear box, and the foundation pile group and the seawall pile were welded and fixed in the base successively. Then, the position of the pile group was fixed with a temporary fixing bracket. In the fourth step, the foundation soil was prepared by the pluviation deposition method. According to the height mark of the plastic film, water with a depth of 5 cm was poured in and backfilled with sand each time [35]. The operation process was repeated, and accelerometers and pore water pressure transducers were placed at the design heights. In the fifth step, pebbles were scattered across the top 5 cm of the test group as a transverse drainage channel. In the sixth step, the embankment was prepared. The preparation process of the embankment is shown in Section 2.2. The model foundations after the above six steps are shown in Figure 3.


## 2.4. Layout of the Sensors

In the model foundation, the sensor layout positions of the test group and the control group were the same, and a schematic diagram of sensors layout is shown in Figures5(a)–5(c). To determine the bending moment response of the piles, strain gauges were symmetrically arranged on piles 1, 2, 3, 4, 5, 6, 7, and 8. The accelerometers were arranged in three layers. In order to explore the response of excess pore water pressure at different depths, the excess pore water pressure transducers of the test group and the control group were arranged in three layers, and three excess pore water pressure transducers were placed in each layer. Four displacement transducers were used to collect the displacement data of the embankment and seawall, and were fixed on the shaking table through a rigid bracket.Figure 5
Instrument layout: (a) plan view, (b) 1-1 cross section, and (c) 2-2 cross section.
(a)(b)(c)


## 2.5. Test Conditions and Loading Scheme

The test conditions are shown in Table4. The loading scheme included three stages: PGA=0.05g (first stage), PGA=0.1g (second stage), and PGA=0.2g (third stage). In the first stage, to explore the influences of different seismic waves on the liquefaction resistance of piles, white noise, Northbridge (far-field), Loma (far-field), ChiChi (far-field), SanFernando (far-field), Loma (near-field without pulse), Northbridge (near-field without pulse), Loma (near-field with pulse), Northbridge (near-field with pulse), and sinusoidal wave (5 Hz) were input subsequently with seismic intensities of PGA=0.05g. The loading sequences in the second and third stage were the same to that in the first stage. The variation of a sinusoidal wave is simpler than the other seismic waves, which is more convenient to analyze the acceleration and excess pore water pressure in experimental research. Therefore, many scholars [36–38] adapted input sinusoidal waves in shaking table experiments. This test will compare and analyze the seismic response of the test group and the control group under sinusoidal wave excitation with seismic intensities of PGA=0.05g, PGA=0.1g, and PGA=0.2g. Figure 6 shows the acceleration time histories of sinusoidal waves with different seismic intensities.Table 4
Test conditions.
CaseType of pileRelative compactnessGauge heightTypes of sand1Rigid drainage pile70%570 mmCoral sand2Ordinary pile70%570 mmCoral sandFigure 6
Time-history curves of the input acceleration: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)


## 3. Experimental Results and Analysis

### 3.1. Excess Pore Water Pressure

In this study, the excess pore water pressure ratio (Ru) is used to represent the change in pore pressure, and the excess pore water pressure ratio (Ru) is defined as the ratio of the excess pore water pressure to the total stress. In previous studies, the excess pore water pressure ratio was often defined as the ratio of excess pore water pressure to the effective stress, and 1.0 was taken as the liquefaction standard [39, 40]. During the liquefaction process, the total stress was generally larger than effective stress. Therefore, in this test, an Ru of 0.8 could be regarded as the liquefaction standard.The layout diagram of the test pore water pressure transducers is shown in Figure7. P1, P4, and P7 are referred to as Row A measurement points; P2, P5, and P8 are referred to as Row B measurement points; and P3, P6, and P9 are referred to as Row C measurement points. P1, P2, and P3 are called deep measurement points; P4, P5, and P6 are called the middle depth measuring points; and P7, P8, and P9 are called shallow depth measuring points.Figure 7
Layout of the pore water pressure transducers.Figure8 shows the time history curve of the excess pore water pressure ratio. In the figure, vibration excitation was input from 5 s and lasted for 10 s; thus, only the stage from 5 s to 15 s in the figure was analyzed. In Figure 8, the excess pore water pressure ratio of the test group was markedly lower than that of the control group overall, which indicated that the anti-liquefaction effect of the drainage pile was significant. When PGA=0.05g, the peak excess pore water pressure ratio is shown in Table 5. According to Figure 8(a) and Table 5, as PGA=0.05g, the peak excess pore water pressure ratio of the test group and control group were 0.07 and 0.081, respectively, and both were located at the P7 position. The two groups both showed low value of Ru, which indicated that the foundation soil was not liquefied and nearly retained its original effective stress. As the PGA increased to 0.1 g, the peak excess pore water pressure ratio is shown in Table 6. According to Figure 8(b) and Table 6, Ru changed markedly. However, the maximum excess pore water pressure ratio of the two groups still occurred at the P7 position. Specifically, at the P7 position, the Ru of the control group reached 0.94, thereby indicating that the soil at the top of the control group was liquefied, while the Ru of the test group had just reached 0.8. As the PGA further increased to 0.2 g, the excess pore water pressure ratio of the foundation soil increased again. The peak excess pore water pressure ratio is shown in Table 7. The Ru of shallow observation points at the positions of P7, P8, and P9 in the control group were 1.66, 1.02, and 1.10, respectively, indicating that the shallow foundation in the control group was liquefied. The Ru of the shallow observation points at the positions of P7, P8, and P9 in the test group were 1.40, 0.90, and 0.75, respectively, indicating that the shallow foundation of the test group was nearly liquefied.Figure 8
Time-history curve of excess pore water pressure ratio: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)Table 5
Peak excess pore water pressure ratio (PGA=0.05g).
Measure pointP1P2P3P4P5P6P7P8P9Case 10.0190.0130.0160.0180.020.020.070.0520.046Case 20.0280.0240.0220.0460.0460.0290.0810.0670.07Table 6
Peak excess pore water pressure ratio (PGA=0.1g).
Measure pointP1P2P3P4P5P6P7P8P9Case 10.240.140.270.240.280.220.800.530.55Case 20.330.300.270.540.470.390.940.600.70Table 7
Peak excess pore water pressure ratio (PGA=0.2g).
Measure pointP1P2P3P4P5P6P7P8P9Case 10.260.180.310.370.340.341.400.900.75Case 20.360.340.330.680.490.361.661.021.10Figures9(a)–9(c) show the changes in the peak excess pore water pressure ratio along the depth of three Rows A, B, and C under the action of different sinusoidal waves. The figure shows that under the seismic load of three intensities, the peak excess pore water pressure ratio of each row increased with decreasing burial depth, and the peak value at the top was much larger than that at the middle and bottom. Therefore, the shallow layer of the foundation was more prone to liquefaction under seismic loading.Figure 9
Peak excess pore water pressure ratio along the depth: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)Figures10(a)–10(c) show the variation of the peak excess pore water pressure ratio along the horizontal direction under the action of different sinusoidal waves. As shown in Figure 10, the peak excess pore water pressure ratio in the test group was lower than that in the control group along the horizontal direction overall. In the test group, the peak excess pore water pressure ratio of the shallow foundation soil showed a decreasing trend from left to right overall. The reason for this phenomenon was likely due to the existence of the drainage body, which allows the excess pore water in the foundation soil at Row B and Row C to more easily drain out of the foundation, resulting in the excess pore water pressure in the foundation soil of Row B and Row C being markedly lower than that of Row A. In the middle or deep foundation, the peak excess pore water pressure ratios in each row at the same depth were similar. However, the peak excess pore water pressure ratios of some observation points in Row B or C were larger than those in Row A. This was probably because the deep and middle layers are far from the surface, and the vertical discharge of excess pore water was more difficult than that in the horizontal. Therefore, when an earthquake load occurs, the excess pore water in the foundation soil at Row A will continue to flow laterally to Rows B and C and then be discharged out of the foundation through the drainage body or channel. This process directly led to an increase in the excess pore water pressure in Rows B and C.Figure 10
Variations of the peak pore water pressure ratio along the horizontal direction: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)


### 3.2. Acceleration

In this section, the acceleration response of the foundation soil of the test group and the control group under the action ofPGA=0.1g will be compared and analyzed. The layout of the test accelerometers is shown in Figure 11. The layout of the test group is the same as that of the control group, but the data of A2 and A7 of the control group are lost.Figure 11
Layout of the accelerometers.To briefly describe the seismic response of the foundation soil, only the time history curve under the action ofPGA=0.1g is shown in Figure 12, and the duration of the seismic load is 10 s. Considering A6 as an example, from 0 s to 1 s, the vibration excitation rose rapidly to the peak value. There was no marked difference in the acceleration response of the foundation soil between the test group and the control group. The development law was similar to the excitation load in Figure 6(b). Neither of the test group and the control group underwent liquefaction and maintained a large shear transfer capacity. From 1 s to 2 s, the excess pore water pressure ratio of the foundation soil continued to increase rapidly and gradually changed from a non-liquefied state to a liquefied state. The acceleration of the model foundation decreased rapidly, and the shear transfer capacity decreased rapidly. The attenuation amplitude of the test group is less than that of the control group. From 2 s to 10 s, the model foundation was in a state of liquefaction. The acceleration of the foundation soil in the test group increased slightly in the stable state until the vibration excitation stopped. The reason for this phenomenon was that under vibration excitation, the drainage channel in the new drainage PCC pile could discharge the excess pore water in time, which increased the shear transfer capacity to some extent.Figure 12
Time-history curve of the acceleration (PGA=0.1g): (a) A1, (b) A3, (c) A4, (d) A5, (e) A6, and (f) A8.
(a)(b)(c)(d)(e)(f)The acceleration amplification factor which is defined as the ratio of the peak value of the acceleration at the measuring point to the peak value of the acceleration input from the shaking table was used to further analyze the acceleration response. By comparing and analyzing the acceleration amplification factors of measuring points A1, A4, A6, and A8 in the test group and the control group under different intensity seismic loads, the distribution of acceleration amplification factors between the two groups was explored. According to the acceleration time history curve, the foundation soil was liquefied after approximately 2 s. The acceleration amplification factor before and after liquefaction is shown in Figure13. As PGA=0.05g, 0.1 g, and 0.2 g, the acceleration amplification factor of foundation soil before and after liquefaction gradually increased from the bottom to the top. Especially near the top part, the growth was significant. Figure 13(a) shows that as PGA=0.05g, the acceleration amplification factors of the test group and the control group before and after liquefaction were similar. The reason for this phenomenon is that as PGA=0.05g, the foundation soil in the test group and the control group had no liquefaction phenomenon, and the shear transfer capacities were similar. Figures 13(b) and 13(c) show the amplification factor of foundation soil under the action of PGA=0.1g and PGA=0.2g, respectively. The difference in acceleration amplification factors between the test and control groups before liquefaction was small, and both maintained good shear transfer capacity. When the soil was liquefied, the acceleration amplification factors of the two groups were significantly different, and the amplification factor of the test group was obviously higher than that of the control group. These phenomena likely occurred because the drainage pile could discharge part of the excess pore water from the soil quickly. Thus, the drainage pile foundation had a higher shear transfer capacity than the ordinary pile foundation.Figure 13
Acceleration amplification factors along the depth: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)


### 3.3. Pile Bending Moment

To explore the pile bending moment response under seismic load in different working conditions, the measuring points (B1-B5) of the seawall piles are shown in Figure14. According to the Euler-Bernoulli beam theory [41], the bending moment M can be obtained in the following form:
(1)M=EI·εt−εcD,where EI is the bending stiffness of the pile, the elastic modulus (E) of plexiglass is 2.8 GPa, εt and εc represent the tensile and compressive strains measured by the strain gauges on both sides of the pile body, and D is the outer diameter of the pile body.Figure 14
Measuring points of the bending moment.Figures15(a)–15(c) show the peak bending moment of the seawall pile. The figure shows that the peak bending moment at each measuring point of the pile in the test group was less than that in the control group as PGA=0.2g, 0.1 g, and 0.05 g, indicating that the new drainage PCC pile could effectively maintain the stability and safety of the seawall. The pile bending moment increased with increasing depth generally, and as PGA=0.2g and 0.05 g, the pile bending moment at the location of B2 was larger than that at the location of B1. Figure 15(d) shows the ratio of the peak bending moment of the test group to that of the control group (RB). At the B5 position near the pile top with various earthquake intensities, the value of the ratio (RB) was high, indicating that the difference of the peak bending moment between the test group and the control group was small. From B4 to B1, the ratio (RB) increased gradually with depth under three seismic loads. With increasing earthquake intensity, the ratio of the peak bending moment (RB) also increased, particularly for a larger depth. In general, as the soil depth and earthquake intensity increased, the difference of peak bending moment for the test group and the control group decreased.Figure 15
Peak pile bending moment: (a)PGA=0.05g, (b) PGA=0.1g, (c) PGA=0.2g, and (d) ratio of peak bending moment between the test group and control group.
(a)(b)(c)(d)


### 3.4. Displacement and Settlement

Figure16 shows the time history curve of the seawall displacement under three earthquake intensities. Vibration excitation was input from the 5 s in the figure and lasted for 10 s. The figure shows that with various earthquake intensities, the displacement time history curve of the seawall shows a continuous growth trend. As PGA=0.05g, the excess pore water pressure of the foundation soil was small, and the seawall produced only small displacements under the action of PGA=0.05g. The displacements of the test group and control group were 0.52 mm and 1.17 mm, respectively. As PGA=0.1g, the excess pore water pressure ratio of the foundation soil increased rapidly, and the surface layer of the foundation soil was near complete liquefaction. The effective stress of the middle and lower foundation soils had also been seriously weakened. After loading stopped, the produced displacements of the seawall in the test group and the control group reached 18.4 mm and 28.7 mm, respectively. As PGA=0.2g, the excess pore water pressure continued to increase. However, due to the large displacement of the foundation soil after a moderate excitation load (PGA=0.1g), the foundation soil tended to become more stable; thus, the produced displacement was smaller than that with PGA=0.1g. Because the seawall displacement of the control group was greater than that of the test group after a moderate excitation load (PGA=0.1g), the compactness of the soil in the control group was larger after the moderate earthquake. Finally, the displacements of the seawall showed no significant differences between the test group and the control group, which were 12.7 mm and 13.1 mm, respectively. As shown in Figure 16, the peak displacement and residual displacement of the test group were less than those of the control group, indicating that the new drainage PCC pile could effectively reduce the lateral spreading of soil liquefaction and played an important role in protecting the stability of the seawall.Figure 16
Time-history curve of the seawall displacement: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)Figure17 shows the variation of the height of the soil foundation surface for the test group and the control group. The height of test group was significantly larger than the control group, which indicated that the foundation settlement of the test group was smaller than that of the control group. The average settlements of the test group and the control group reached 39.4 mm and 57.8 mm, respectively. The soil foundation surface settlement of the former was approximately 32% less than that of the latter. This shows that the new drainage PCC pile could effectively reduce the settlement of the coral sand foundation during the liquefaction process.Figure 17
Variation of the height of the soil foundation surface.Figure18 shows the time history curve of the embankment displacement. The embankment located upon the soil foundation surface. Because the displacement data were lost as PGA=0.05g, only the embankment response under the action of PGAs of 0.1 g and 0.2 g was analyzed. The figure shows that with PGA=0.1g and 0.2 g, the displacement of the test group changed slightly, and curve trend presented a stable state in general. Differently, the control group showed an increasing trend, and the peak and residual displacement were significantly larger than those of the test group. After the seismic load stopped, the residual displacements of the control group as PGA=0.1g and 0.2 g were approximately 14.4 mm and 16.0 mm, respectively, and those of the test groups were near 0 mm. This implied that the drainage pile could effectively ensure stability of the embankment under earthquake loads.Figure 18
Time-history curve of the embankment displacement: (a)PGA=0.1g and (b) PGA=0.2g.
(a)(b)Figures19(a) and 19(b) show the picture of the foundation deformation under the action of PGA=0.1g and 0.2 g, respectively. As shown in the figure, under the action of PGA=0.1g, the control group experienced a large seawall displacement and foundation settlement. Compared with the control group, the soil settlement and seawall displacement in the test group were small. Moreover, the bottom layer of the embankment was still in good contact with the foundation soil, and the deformation was small. After a large excitation load (PGA=0.2g), the seawall moved more, and the subgrade also settled. At this time, the deformation of the bottom layer of the embankment in the control group further intensified and was completely submerged by water, while the bottom layer of the test group was in a semi-submerged state.Figure 19
Picture of the foundation deformation: (a)PGA=0.1g and (b) PGA=0.2g.
(a)(b)


## 3.1. Excess Pore Water Pressure

In this study, the excess pore water pressure ratio (Ru) is used to represent the change in pore pressure, and the excess pore water pressure ratio (Ru) is defined as the ratio of the excess pore water pressure to the total stress. In previous studies, the excess pore water pressure ratio was often defined as the ratio of excess pore water pressure to the effective stress, and 1.0 was taken as the liquefaction standard [39, 40]. During the liquefaction process, the total stress was generally larger than effective stress. Therefore, in this test, an Ru of 0.8 could be regarded as the liquefaction standard.The layout diagram of the test pore water pressure transducers is shown in Figure7. P1, P4, and P7 are referred to as Row A measurement points; P2, P5, and P8 are referred to as Row B measurement points; and P3, P6, and P9 are referred to as Row C measurement points. P1, P2, and P3 are called deep measurement points; P4, P5, and P6 are called the middle depth measuring points; and P7, P8, and P9 are called shallow depth measuring points.Figure 7
Layout of the pore water pressure transducers.Figure8 shows the time history curve of the excess pore water pressure ratio. In the figure, vibration excitation was input from 5 s and lasted for 10 s; thus, only the stage from 5 s to 15 s in the figure was analyzed. In Figure 8, the excess pore water pressure ratio of the test group was markedly lower than that of the control group overall, which indicated that the anti-liquefaction effect of the drainage pile was significant. When PGA=0.05g, the peak excess pore water pressure ratio is shown in Table 5. According to Figure 8(a) and Table 5, as PGA=0.05g, the peak excess pore water pressure ratio of the test group and control group were 0.07 and 0.081, respectively, and both were located at the P7 position. The two groups both showed low value of Ru, which indicated that the foundation soil was not liquefied and nearly retained its original effective stress. As the PGA increased to 0.1 g, the peak excess pore water pressure ratio is shown in Table 6. According to Figure 8(b) and Table 6, Ru changed markedly. However, the maximum excess pore water pressure ratio of the two groups still occurred at the P7 position. Specifically, at the P7 position, the Ru of the control group reached 0.94, thereby indicating that the soil at the top of the control group was liquefied, while the Ru of the test group had just reached 0.8. As the PGA further increased to 0.2 g, the excess pore water pressure ratio of the foundation soil increased again. The peak excess pore water pressure ratio is shown in Table 7. The Ru of shallow observation points at the positions of P7, P8, and P9 in the control group were 1.66, 1.02, and 1.10, respectively, indicating that the shallow foundation in the control group was liquefied. The Ru of the shallow observation points at the positions of P7, P8, and P9 in the test group were 1.40, 0.90, and 0.75, respectively, indicating that the shallow foundation of the test group was nearly liquefied.Figure 8
Time-history curve of excess pore water pressure ratio: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)Table 5
Peak excess pore water pressure ratio (PGA=0.05g).
Measure pointP1P2P3P4P5P6P7P8P9Case 10.0190.0130.0160.0180.020.020.070.0520.046Case 20.0280.0240.0220.0460.0460.0290.0810.0670.07Table 6
Peak excess pore water pressure ratio (PGA=0.1g).
Measure pointP1P2P3P4P5P6P7P8P9Case 10.240.140.270.240.280.220.800.530.55Case 20.330.300.270.540.470.390.940.600.70Table 7
Peak excess pore water pressure ratio (PGA=0.2g).
Measure pointP1P2P3P4P5P6P7P8P9Case 10.260.180.310.370.340.341.400.900.75Case 20.360.340.330.680.490.361.661.021.10Figures9(a)–9(c) show the changes in the peak excess pore water pressure ratio along the depth of three Rows A, B, and C under the action of different sinusoidal waves. The figure shows that under the seismic load of three intensities, the peak excess pore water pressure ratio of each row increased with decreasing burial depth, and the peak value at the top was much larger than that at the middle and bottom. Therefore, the shallow layer of the foundation was more prone to liquefaction under seismic loading.Figure 9
Peak excess pore water pressure ratio along the depth: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)Figures10(a)–10(c) show the variation of the peak excess pore water pressure ratio along the horizontal direction under the action of different sinusoidal waves. As shown in Figure 10, the peak excess pore water pressure ratio in the test group was lower than that in the control group along the horizontal direction overall. In the test group, the peak excess pore water pressure ratio of the shallow foundation soil showed a decreasing trend from left to right overall. The reason for this phenomenon was likely due to the existence of the drainage body, which allows the excess pore water in the foundation soil at Row B and Row C to more easily drain out of the foundation, resulting in the excess pore water pressure in the foundation soil of Row B and Row C being markedly lower than that of Row A. In the middle or deep foundation, the peak excess pore water pressure ratios in each row at the same depth were similar. However, the peak excess pore water pressure ratios of some observation points in Row B or C were larger than those in Row A. This was probably because the deep and middle layers are far from the surface, and the vertical discharge of excess pore water was more difficult than that in the horizontal. Therefore, when an earthquake load occurs, the excess pore water in the foundation soil at Row A will continue to flow laterally to Rows B and C and then be discharged out of the foundation through the drainage body or channel. This process directly led to an increase in the excess pore water pressure in Rows B and C.Figure 10
Variations of the peak pore water pressure ratio along the horizontal direction: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)


## 3.2. Acceleration

In this section, the acceleration response of the foundation soil of the test group and the control group under the action ofPGA=0.1g will be compared and analyzed. The layout of the test accelerometers is shown in Figure 11. The layout of the test group is the same as that of the control group, but the data of A2 and A7 of the control group are lost.Figure 11
Layout of the accelerometers.To briefly describe the seismic response of the foundation soil, only the time history curve under the action ofPGA=0.1g is shown in Figure 12, and the duration of the seismic load is 10 s. Considering A6 as an example, from 0 s to 1 s, the vibration excitation rose rapidly to the peak value. There was no marked difference in the acceleration response of the foundation soil between the test group and the control group. The development law was similar to the excitation load in Figure 6(b). Neither of the test group and the control group underwent liquefaction and maintained a large shear transfer capacity. From 1 s to 2 s, the excess pore water pressure ratio of the foundation soil continued to increase rapidly and gradually changed from a non-liquefied state to a liquefied state. The acceleration of the model foundation decreased rapidly, and the shear transfer capacity decreased rapidly. The attenuation amplitude of the test group is less than that of the control group. From 2 s to 10 s, the model foundation was in a state of liquefaction. The acceleration of the foundation soil in the test group increased slightly in the stable state until the vibration excitation stopped. The reason for this phenomenon was that under vibration excitation, the drainage channel in the new drainage PCC pile could discharge the excess pore water in time, which increased the shear transfer capacity to some extent.Figure 12
Time-history curve of the acceleration (PGA=0.1g): (a) A1, (b) A3, (c) A4, (d) A5, (e) A6, and (f) A8.
(a)(b)(c)(d)(e)(f)The acceleration amplification factor which is defined as the ratio of the peak value of the acceleration at the measuring point to the peak value of the acceleration input from the shaking table was used to further analyze the acceleration response. By comparing and analyzing the acceleration amplification factors of measuring points A1, A4, A6, and A8 in the test group and the control group under different intensity seismic loads, the distribution of acceleration amplification factors between the two groups was explored. According to the acceleration time history curve, the foundation soil was liquefied after approximately 2 s. The acceleration amplification factor before and after liquefaction is shown in Figure13. As PGA=0.05g, 0.1 g, and 0.2 g, the acceleration amplification factor of foundation soil before and after liquefaction gradually increased from the bottom to the top. Especially near the top part, the growth was significant. Figure 13(a) shows that as PGA=0.05g, the acceleration amplification factors of the test group and the control group before and after liquefaction were similar. The reason for this phenomenon is that as PGA=0.05g, the foundation soil in the test group and the control group had no liquefaction phenomenon, and the shear transfer capacities were similar. Figures 13(b) and 13(c) show the amplification factor of foundation soil under the action of PGA=0.1g and PGA=0.2g, respectively. The difference in acceleration amplification factors between the test and control groups before liquefaction was small, and both maintained good shear transfer capacity. When the soil was liquefied, the acceleration amplification factors of the two groups were significantly different, and the amplification factor of the test group was obviously higher than that of the control group. These phenomena likely occurred because the drainage pile could discharge part of the excess pore water from the soil quickly. Thus, the drainage pile foundation had a higher shear transfer capacity than the ordinary pile foundation.Figure 13
Acceleration amplification factors along the depth: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)


## 3.3. Pile Bending Moment

To explore the pile bending moment response under seismic load in different working conditions, the measuring points (B1-B5) of the seawall piles are shown in Figure14. According to the Euler-Bernoulli beam theory [41], the bending moment M can be obtained in the following form:
(1)M=EI·εt−εcD,where EI is the bending stiffness of the pile, the elastic modulus (E) of plexiglass is 2.8 GPa, εt and εc represent the tensile and compressive strains measured by the strain gauges on both sides of the pile body, and D is the outer diameter of the pile body.Figure 14
Measuring points of the bending moment.Figures15(a)–15(c) show the peak bending moment of the seawall pile. The figure shows that the peak bending moment at each measuring point of the pile in the test group was less than that in the control group as PGA=0.2g, 0.1 g, and 0.05 g, indicating that the new drainage PCC pile could effectively maintain the stability and safety of the seawall. The pile bending moment increased with increasing depth generally, and as PGA=0.2g and 0.05 g, the pile bending moment at the location of B2 was larger than that at the location of B1. Figure 15(d) shows the ratio of the peak bending moment of the test group to that of the control group (RB). At the B5 position near the pile top with various earthquake intensities, the value of the ratio (RB) was high, indicating that the difference of the peak bending moment between the test group and the control group was small. From B4 to B1, the ratio (RB) increased gradually with depth under three seismic loads. With increasing earthquake intensity, the ratio of the peak bending moment (RB) also increased, particularly for a larger depth. In general, as the soil depth and earthquake intensity increased, the difference of peak bending moment for the test group and the control group decreased.Figure 15
Peak pile bending moment: (a)PGA=0.05g, (b) PGA=0.1g, (c) PGA=0.2g, and (d) ratio of peak bending moment between the test group and control group.
(a)(b)(c)(d)


## 3.4. Displacement and Settlement

Figure16 shows the time history curve of the seawall displacement under three earthquake intensities. Vibration excitation was input from the 5 s in the figure and lasted for 10 s. The figure shows that with various earthquake intensities, the displacement time history curve of the seawall shows a continuous growth trend. As PGA=0.05g, the excess pore water pressure of the foundation soil was small, and the seawall produced only small displacements under the action of PGA=0.05g. The displacements of the test group and control group were 0.52 mm and 1.17 mm, respectively. As PGA=0.1g, the excess pore water pressure ratio of the foundation soil increased rapidly, and the surface layer of the foundation soil was near complete liquefaction. The effective stress of the middle and lower foundation soils had also been seriously weakened. After loading stopped, the produced displacements of the seawall in the test group and the control group reached 18.4 mm and 28.7 mm, respectively. As PGA=0.2g, the excess pore water pressure continued to increase. However, due to the large displacement of the foundation soil after a moderate excitation load (PGA=0.1g), the foundation soil tended to become more stable; thus, the produced displacement was smaller than that with PGA=0.1g. Because the seawall displacement of the control group was greater than that of the test group after a moderate excitation load (PGA=0.1g), the compactness of the soil in the control group was larger after the moderate earthquake. Finally, the displacements of the seawall showed no significant differences between the test group and the control group, which were 12.7 mm and 13.1 mm, respectively. As shown in Figure 16, the peak displacement and residual displacement of the test group were less than those of the control group, indicating that the new drainage PCC pile could effectively reduce the lateral spreading of soil liquefaction and played an important role in protecting the stability of the seawall.Figure 16
Time-history curve of the seawall displacement: (a)PGA=0.05g, (b) PGA=0.1g, and (c) PGA=0.2g.
(a)(b)(c)Figure17 shows the variation of the height of the soil foundation surface for the test group and the control group. The height of test group was significantly larger than the control group, which indicated that the foundation settlement of the test group was smaller than that of the control group. The average settlements of the test group and the control group reached 39.4 mm and 57.8 mm, respectively. The soil foundation surface settlement of the former was approximately 32% less than that of the latter. This shows that the new drainage PCC pile could effectively reduce the settlement of the coral sand foundation during the liquefaction process.Figure 17
Variation of the height of the soil foundation surface.Figure18 shows the time history curve of the embankment displacement. The embankment located upon the soil foundation surface. Because the displacement data were lost as PGA=0.05g, only the embankment response under the action of PGAs of 0.1 g and 0.2 g was analyzed. The figure shows that with PGA=0.1g and 0.2 g, the displacement of the test group changed slightly, and curve trend presented a stable state in general. Differently, the control group showed an increasing trend, and the peak and residual displacement were significantly larger than those of the test group. After the seismic load stopped, the residual displacements of the control group as PGA=0.1g and 0.2 g were approximately 14.4 mm and 16.0 mm, respectively, and those of the test groups were near 0 mm. This implied that the drainage pile could effectively ensure stability of the embankment under earthquake loads.Figure 18
Time-history curve of the embankment displacement: (a)PGA=0.1g and (b) PGA=0.2g.
(a)(b)Figures19(a) and 19(b) show the picture of the foundation deformation under the action of PGA=0.1g and 0.2 g, respectively. As shown in the figure, under the action of PGA=0.1g, the control group experienced a large seawall displacement and foundation settlement. Compared with the control group, the soil settlement and seawall displacement in the test group were small. Moreover, the bottom layer of the embankment was still in good contact with the foundation soil, and the deformation was small. After a large excitation load (PGA=0.2g), the seawall moved more, and the subgrade also settled. At this time, the deformation of the bottom layer of the embankment in the control group further intensified and was completely submerged by water, while the bottom layer of the test group was in a semi-submerged state.Figure 19
Picture of the foundation deformation: (a)PGA=0.1g and (b) PGA=0.2g.
(a)(b)


## 4. Conclusion

In this study, a series of shaking table tests were performed to investigate the seismic response of a subgrade-embankment-seawall system reinforced by a new drainage PCC pile and ordinary pile on a coral sand liquefaction spreading site. The primary conclusions of this study are as follows:(1)
Under the action of different earthquake intensities, the peak excess pore water pressure ratio in the test group was lower than that in the control group, indicating that the new drainage PCC pile could effectively reduce the excess pore water pressure ratio of foundation soil and the possibility of liquefaction. The peak excess pore water pressure ratio of the foundation soil increased with increasing height from the bottom. In practical engineering applications, attention should be given to the treatment of the liquefaction of soil surface layer(2)
Compared with ordinary pile, the foundation soil reinforced with new drainage PCC pile would maintain a higher shear stress and induce higher acceleration of the soil. These effects became more significant with increasing earthquake strength. With a larger vibration excitation, the acceleration amplification factors of soil for the two groups were significantly different(3)
The peak bending moment of the seawall pile generally increased with increasing depth, and the peak bending moment of the test group was much lower than that of the control group, indicating that the drainage pile could effectively protect the stability of the seawall. However, with increasing seismic intensity, the difference of peak bending moment for the test group and the control group decreased(4)
Under seismic load, drainage piles could reduce the excess pore water pressure of soil foundation and weaken the liquefaction lateral spreading of the soil to effectively reduce the settlement of foundation soil, the deformation of the embankment, and the displacement of the seawall. According to the test data, the foundation settlement of the test group was approximately 32% less than that of the control group



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*Source: 4489478-2023-01-23.xml*