U.S. Department of Transportation Publication No. FHWA-NHI-16-009

Federal Highway Administration FHWA GEC 012 – Volume I 

July 2016

NHI Courses No. 132021 and 132022

Design and Construction of

Driven Pile Foundations – Volume I 

Developed following:

AASHTO LRFD Bridge Design Specifications, 7th Edition, 2014,with 2015 Interim.

AASHTO LRFD Bridge Construction Specifications, 3rd Edition, 2010, with ‘11, ‘12, ’13, ‘14, and ‘15 

Interims.

NOTICE

The contents of this report reflect the views of the authors, who are responsible for the facts 

and accuracy of the data presented herein. The contents do not necessarily reflect policy of 

the Department of Transportation. This report does not constitute a standard, specification, or 

regulation. The United States Government does not endorse products or manufacturers. Trade 

or manufacturers′ names appear herein only to illustrate methods and procedures, and are 

considered essential to the objective of this document. 

Technical Report Documentation Page

1. REPORT NO.

FHWA-NHI-16-009

2. GOVERNMENT ACCESSION NO. 3. RECIPIENT'S CATALOG NO.

4. TITLE AND SUBTITLE

Geotechnical Engineering Circular No. 12 – Volume I 

Design and Construction of Driven Pile Foundations

5. REPORT DATE

July 2016

6. PERFORMING ORGANIZATION CODE

7. AUTHOR(S)

Patrick J. Hannigan, PE, Frank Rausche, PhD, PE, 

Garland E. Likins, PE, Brent R. Robinson, PhD, PE, and 

Matthew L. Becker, EI.

8. PERFORMING ORGANIZATION 

REPORT NO.

9. PERFORMING ORGANIZATION NAME AND ADDRESS

Ryan R. Berg & Associates, Inc.

2190 Leyland Alcove

Woodbury, MN 55125

10. WORK UNIT NO.

11. CONTRACT OR GRANT NO.

DTFH61-11-D-00049

12. SPONSORING AGENCY NAME AND ADDRESS

National Highway Institute

U.S. Department of Transportation

Federal Highway Administration, Washington, DC 20590

13. TYPE OF REPORT & PERIOD 

COVERED

Final Report

14. SPONSORING AGENCY CODE

15. SUPPLEMENTARY NOTES

FHWA COTR: Heather Shelsta

FHWA Technical Working Group: Naser Abu-Hejlah, PhD, PE; Scott Anderson, PE; and

Silas Nichols, PE

16. ABSTRACT

This document presents information on the analysis, design, and construction of driven 

pile foundations for highway structures. This document updates and replaces FHWA 

NHI-05-042 and FHWA NHI-05-043 as the primary FHWA guidance and reference 

document on driven pile foundations. The manual addresses design aspects including 

subsurface exploration, laboratory testing, pile selection, aspects of geotechnical and 

structural limit states, as well as technical specifications. Construction aspects including 

static load tests, dynamic tests, rapid load tests, wave equation analyses, dynamic 

formulas and development of driving criteria, as well as pile driving equipment, pile 

driving accessories, and monitoring of pile installation inspection are also covered. Step 

by step procedures are included for most analysis procedures and design examples.

17. KEY WORDS

Driven pile foundations, foundation economics, site 

characterization, geomaterial properties, axial compression 

resistance, axial tension resistance, lateral resistance, pile 

groups, specifications, nominal resistance determination

tests, construction monitoring and quality assurance. 

18. DISTRIBUTION STATEMENT

No restrictions.

19. SECURITY CLASSIF.

Unclassified

20. SECURITY CLASSIF.

Unclassified

21. NO. OF PAGES

517 

22. PRICE

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

Notes: 

A compliant version can be found here: https://www.fhwa.dot.gov/publications/convtabl.cfm

1) The primary metric (SI) units used in civil engineering are meter (m), kilogram (kg), second (s), Newton (N), and Pascal (Pa=N/m2). 

2) In a "soft" conversion, an English measurement is mathematically converted to its exact metric equivalent.

3) In a "hard" conversion, a new rounded metric number is created that is convenient to work with and remember.

PREFACE

The purpose of this manual is to provide updated, state-of-the-practice information 

for the design and construction of driven pile foundations in accordance with the 

Load and Resistance Factor Design (LRFD) platform. Engineers and contractors 

have been designing and installing pile foundations for many years. During the past 

three decades, the industry has experienced several major improvements including 

newer and more accurate methods of predicting and measuring geotechnical 

resistance, vast improvements in design software, highly specialized and 

sophisticated equipment for pile driving, and improved methods of construction 

control. Previous editions of the FHWA Design and Construction of Driven Pile 

Foundations manual were published 1985, 1996, and 2006 and chronicle the many 

changes in design and construction practice over the past 30 years. This two 

volume edition, GEC-12, serves as the FHWA reference document for highway 

projects involving driven pile foundations. 

Volume I, FHWA-NHI-16-009, covers the foundation selection process, site 

characterization, geotechnical design parameters and reporting, selection of pile 

type, geotechnical aspects of limit state design, and structural aspects of limits state 

design. Volume II, FHWA-NHI-16-010, addresses static load tests, dynamic testing 

and signal matching, rapid load testing, wave equation analysis, dynamic formulas, 

contract documents, pile driving equipment, pile accessories, driving criteria, and 

construction monitoring. Comprehensive design examples are presented in 

publication FHWA-NHI-16-064. 

Throughout this manual, numerous references will be made to the names of 

software or technology that are proprietary to a specific manufacturer or vendor. 

Please note that the FHWA does not endorse or approve commercially available 

products, and is very sensitive to the perceptions of endorsement or preferred 

approval of commercially available products used in transportation applications. Our 

goal with this development is to provide recommended technical guidance for the 

safe design and construction of driven pile foundations that reflects the current state 

of practice and provides information on advances and innovations in the industry. 

To accomplish this, it is necessary to illustrate methods and procedures for design 

and construction of driven pile foundations. Where proprietary products are 

described in text or figures, it is only for this purpose.
The primary audience for this document is: agency and consulting engineers 

specialized in geotechnical and structural design of highway structures; engineering 

geologists and consulting engineers providing technical reviews, or who are 

engaged in the design, procurement, and construction of driven pile foundations 

This document is also intended for management, specification and contracting 

specialists, as well as for construction engineers interested in design and contracting 

aspects of driven pile systems.

This document draws material from the three earlier FHWA publications in this field; 

FHWA-DP-66-1 by Vanikar (1985), FHWA HI 97-013 and FHWA HI 97-014 by 

Hannigan et al. (1998), and FHWA NHI-05-042 and FHWA NHI-05-043 by Hannigan 

et al. (2006). Photographs without specific acknowledgement in this two volume 

document are from these previous editions, their associated training courses, or 

from the consulting practice of GRL Engineers, Inc.

The following individuals were part of the Ryan R. Berg & Associates internal peer 

review team and are acknowledged for their technical advice and contributions to 

this version of the document: 

Mr. Jerry DiMaggio – Applied Research Associates, Inc.

Mr. Van E. Komurka – Wagner Komurka Geotechnical Group, Inc.

Mr. Billy Camp – S&ME, Inc.

Dr. Brian Anderson – Auburn University
i 

TABLE OF CONTENTS

LIST OF TABLES ...................................................................................... xi

LIST OF FIGURES ....................................................................................xv

1 DRIVEN PILE FOUNDATION MANUAL .............................................. 1

1.1 INTRODUCTION ..................................................................... 1

1.2 PURPOSE OF THE MANUAL................................................... 1

1.3 SCOPE OF MANUAL............................................................... 3

1.4 HISTORY OF DRIVEN PILE FOUNDATIONS............................ 4

1.5 INFORMATION SOURCES ...................................................... 4

REFERENCES........................................................................................... 6

2 OVERVIEW OF PILE FOUNDATION DESIGN & CONSTRUCTION...... 7

2.1 INTRODUCTION ..................................................................... 7

2.2 LIMIT STATES ........................................................................ 7

2.3 LOADS, LOAD COMBINATIONS, AND LOAD FACTORS .......... 9

2.4 NOMINAL AND FACTORED RESISTANCE ............................14

2.5 STRENGTH LIMIT STATES ...................................................15

2.6 SERVICE LIMIT STATES.......................................................15

2.7 EXTREME EVENT LIMIT STATES .........................................16

2.8 CONSTRUCTION OF PILE FOUNDATIONS ...........................16

2.9 FOUNDATION SPECIALIST INVOLVEMENT..........................17

2.10 THE DRIVEN PILE DESIGN AND CONSTRUCTION 

PROCESS ............................................................................18

2.11 COMMUNICATION................................................................39

REFERENCES.........................................................................................42

3 CONSIDERATIONS IN FOUNDATION SELECTION..........................45

3.1 FOUNDATION DESIGN APPROACH .....................................46

3.2 FOUNDATION ALTERNATIVES.............................................47

3.2.1 Shallow Foundations...................................................49

3.2.2 Shallow Foundations with Ground Improvement ...........49

3.2.3 Deep Foundations ......................................................49

3.2.3.1 Driven Piles ..................................................50

3.2.3.2 Drilled Shafts................................................50

3.2.3.3 Micropiles.....................................................50

3.2.3.4 Continuous Flight Auger (CFA) Piles ..............51
ii

3.3 ESTABLISHMENT OF A NEED FOR A DEEP 

FOUNDATION.......................................................................51

3.4 ECONOMIC ASPECTS OF FOUNDATION SELECTION..........54

3.4.1 Foundation Support Cost.............................................54

3.4.1.1 Cost Optimization Example............................55

3.5 OTHER CONSIDERATIONS ..................................................61

3.5.1 Constructability...........................................................62

3.5.2 Consideration of Pile Driving Noise ..............................62

3.5.5 Durability Considerations.............................................68

3.6 UNANTICIPATED OCCURRENCES.......................................72

3.6.1 Fill Stockpile ...............................................................73

3.6.2 Adjacent Construction .................................................74

REFERENCES.........................................................................................75

4 SITE CHARACTERIZATION.............................................................79

4.1 INTRODUCTION ...................................................................79

4.2 SITE CHARACTERIZATION PROGRAM ................................80

4.2.1 Data Collection ...........................................................80

4.2.2 Field Reconnaissance Survey......................................83

4.2.3 Detailed Field Exploration............................................85

4.2.3.1 Geophysical Surveys ....................................85

4.2.3.2 Depth, Spacing, and Frequency of Boring & 

In-Situ Tests .................................................86

4.2.3.3 Soil Boring Methods ......................................88

4.2.3.4 Soil Sampling Methods..................................88

4.2.3.5 Rock Exploration Methods (Coring / Drilling) ...91

4.2.3.6 Groundwater ................................................94

4.2.4 Information Required for Construction ..........................95

REFERENCES.........................................................................................96

5 GEOMATERIAL DESIGN PARAMETERS AND GEOTECHNICAL 

REPORTS.......................................................................................99

5.1 IN-SITU SOIL TESTING.........................................................99

5.1.1 Standard Penetration Test......................................... 102

5.1.2 Cone Penetration Test .............................................. 109

5.1.3 Vane Shear Test....................................................... 112

5.1.4 Other In-Situ Tests.................................................... 113

5.1.4.1 Dilatometer Test ......................................... 114

5.1.4.2 Pressuremeter Tests................................... 114

5.1.4.3 Dynamic Cone ............................................ 114

5.2 SOIL PARAMETERS ........................................................... 115
iii

5.2.1 Soil Classification and Index Properties...................... 115

5.2.2 In-Situ Stress............................................................ 118

5.2.3 Shear Strength ......................................................... 119

5.2.3.1 Laboratory Tests for Soil Shear Strength ...... 121

5.2.3.1.1 Direct Shear ........................................ 121

5.2.3.1.2 Unconfined Compression ..................... 121

5.2.3.1.3 Triaxial Compression Test.................... 122

5.2.3.2 Effective Stress Friction Angle Correlations, 

Cohesionless Soils...................................... 122

5.2.3.3 Fully Drained Shear Strength of FineGrained Cohesive Soils ............................... 127

5.2.3.4 Undrained Shear Strength of Fine-Grained 

Cohesive Soils............................................ 127

5.2.4 Deformation ............................................................. 130

5.2.4.1 Elastic Deformation..................................... 130

5.2.4.2 Primary Consolidation Settlement ................ 132

5.2.5 Electro Chemical Properties ...................................... 134

5.3 ROCK PARAMETERS ......................................................... 134

5.3.1 Rock Index Properties and Classification.................... 135

5.3.2 Rock Mass Shear Strength........................................ 135

5.3.3 Rock Mass Deformation ............................................ 136

5.4 CONSIDERATIONS FOR PILE DRIVABILITY ....................... 136

5.5 SELECTION OF PARAMETERS FOR DESIGN AND 

CONSTRUCTION................................................................ 137

5.5.1 Soil Parameters........................................................ 138

5.5.2 Rock Parameters...................................................... 138

5.5.3 Site Variability .......................................................... 140

5.6 GEOTECHNICAL REPORTING............................................ 142

5.6.1 Geotechnical Data Reports........................................ 142

5.6.2 Geotechnical Foundation Design Reports................... 144

5.6.3 Geotechnical Baseline Reports.................................. 148

REFERENCES....................................................................................... 149

6 PILE TYPES FOR FURTHER EVALUATION................................... 155

6.1 OVERVIEW OF TYPICAL PILE TYPES ................................ 155

6.2 TIMBER PILES.................................................................... 164

6.3 STEEL H-PILES .................................................................. 166

6.4 STEEL PIPE PILES ............................................................. 168

6.4.1 Closed End Steel Pipe .............................................. 168

6.4.2 Open End Steel Pipe ................................................ 170
iv

6.5 MONOTUBE PILES ............................................................. 173

6.6 TAPERTUBE PILES ............................................................ 174

6.7 SPIN FIN PILES .................................................................. 174

6.8 PRESTRESSED CONCRETE PILES.................................... 176

6.9 CONCRETE CYLINDER PILES............................................ 178

6.9.1 Spun-Cast Cylinder Piles........................................... 179

6.9.2 Bed-Cast Cylinder Piles ............................................ 180

6.9.3 Industrial Concrete Products (ICP) Piles..................... 180

6.10 COMPOSITE PILES ............................................................ 181

6.10.1 Precast Concrete - Steel H-pile Composite Piles......... 181

6.10.2 Steel Pipe - H-pile Composite Piles............................ 182

6.10.3 Corrugated Shell - Timber Composite Piles ................ 182

6.10.4 Corrugated Shell - Pipe Composite Piles .................... 183

6.11 PILE TYPES INFREQUENTLY USED ON 

TRANSPORTATION PROJECTS ......................................... 183

6.11.1 Fundex Piles ............................................................ 183

6.11.2 Tubex Piles .............................................................. 184

6.11.3 Pressure Injected Footings (PIF)................................ 184

6.11.4 Mandrel Driven Piles ................................................. 185

6.11.5 Reinforced Concrete Piles ......................................... 185

6.12 DESIGN CONSIDERATIONS IN AGGRESSIVE 

SUBSURFACE ENVIRONMENTS ........................................ 187

6.12.1 Corrosion of Steel Piles............................................. 188

6.12.1.1 Corrosion in Non-Marine Environments ........ 188

6.12.1.2 Corrosion in Marine Environments ............... 192

6.12.2 Sulfate and Chloride Attack on Concrete Piles ............ 193

6.12.3 Bacteria, Fungi, Insect, and Marine Borer Attacks on 

Timber Piles ............................................................. 194

6.12.4 Design Options for Piles Subject to Degradation or 

Abrasion .................................................................. 195

6.13 SELECTION OF PILE TYPE AND SIZE FOR FURTHER 

EVALUATION...................................................................... 196

6.14 HISTORICAL PRICE INFORMATION ................................... 199

6.14.1 California.................................................................. 200

6.14.2 Florida ..................................................................... 201

6.14.3 Indiana..................................................................... 202

6.14.4 Maryland .................................................................. 203

6.14.5 North Carolina .......................................................... 204

6.14.6 Pennsylvania............................................................ 205

6.14.7 Texas....................................................................... 206
v 

REFERENCES....................................................................................... 207

7 GEOTECHNICAL ASPECTS AND LIMIT STATE DESIGN ............... 211

7.1 INTRODUCTION ................................................................. 211

7.1.1 Static Analysis Methods in Limit State Design............. 212

7.1.2 Events During and After Pile Driving .......................... 214

7.1.2.1 Cohesionless Soils...................................... 214

7.1.2.2 Cohesive Soils............................................ 216

7.1.2.3 Additional Soil Resistance Considerations .... 217

7.1.3 Load Transfer........................................................... 219

7.1.4 Effective Stress ........................................................ 221

7.1.5 Resistance Factors ................................................... 223

7.1.6 Interdiscipline Communication and Coordination......... 227

7.2 STRENGTH LIMIT STATES ................................................. 227

7.2.1 Determination of Nominal Resistance for Single Piles . 227

7.2.1.1 General...................................................... 227

7.2.1.2 Static Analysis Overview ............................. 229

7.2.1.3 Nominal Resistance of Single Piles in Soils .. 232

7.2.1.3.1 Nordlund Method – Cohesionless Soils . 232

7.2.1.3.2 α-Method - Cohesive Soils ................... 245

7.2.1.3.3 P2A Method – Mixed Soil Profiles ......... 251

7.2.1.3.4 Effective Stress β-Method – Mixed Soil 

Profiles ............................................... 255

7.2.1.3.5 Brown Method – Mixed Soil Profiles – 

SPT Data ............................................ 258

7.2.1.3.6 Eslami and Fellenius Method – CPT 

Data.................................................... 260

7.2.1.3.7 Nottingham and Schmertmann Method 

– CPT Data ......................................... 262

7.2.1.4 Nominal Resistance of Single Piles to Rock.. 269

7.2.1.4.1 Piles Driven into Soft and Weak Rock ... 271

7.2.1.4.2 Piles Driven to Hard Rock .................... 272

7.2.1.5 Software for Single Pile Nominal Resistance 

Computations ............................................. 274

7.2.2 Resistance of Pile Groups in Axial Compression......... 280

7.2.2.1 Pile Groups in Cohesionless Soils................ 282

7.2.2.2 Pile Groups in Cohesive Soils ...................... 284

7.2.2.3 Block Failure of Pile Groups ........................ 286

7.2.3 Design for Axial Tension Resistance .......................... 288

7.2.3.1 Axial Tension Resistance of Single Piles ...... 288
vi

7.2.3.2 Axial Tension Resistance of Pile Groups ...... 290

7.2.3.2.1 Axial Tension Resistance of Groups in 

Cohesionless Soils .............................. 291

7.2.3.2.2 Axial Tension Resistance of Groups in 

Cohesive Soils..................................... 292

7.2.4 Nominal Axial Resistance Changes after Pile Driving .. 293

7.2.4.1 Relaxation .................................................. 293

7.2.4.2 Soil Setup................................................... 294

7.2.4.2.1 Estimation of Pore Pressures During 

Driving ................................................ 299

7.2.4.3 Implementation of Time Effects During 

Construction ............................................... 300

7.2.5 Nominal Lateral Resistance ....................................... 301

7.2.6 Pile Length Estimates for Contract Documents ........... 301

7.2.7 Groundwater Effects and Buoyancy ........................... 302

7.2.8 Site Dewatering ........................................................ 302

7.2.8.1 Artesian Conditions ..................................... 303

7.2.9 Scour....................................................................... 304

7.2.10 Downdrag ................................................................ 307

7.3 SERVICE LIMIT STATES..................................................... 308

7.3.1 Tolerable Vertical Deformations and Angular 

Distortion.................................................................. 309

7.3.1.1 Load Factor for Vertical Deformations .......... 310

7.3.2 S-0 Concept for Vertical Deformations ....................... 310

7.3.3 Construction Point Concept ....................................... 312

7.3.4 Recommended Procedure for Vertical Deformation 

Analysis ................................................................... 314

7.3.5 Pile Group Settlement ............................................... 316

7.3.5.1 Elastic Compression of Piles........................ 317

7.3.5.2 Group Settlement in Cohesionless Soils ....... 317

7.3.5.2.1 Method Based on SPT Test Data.......... 319

7.3.5.2.2 Method Based on CPT Test Data.......... 319

7.3.5.3 Group Settlement in Cohesive Soils ............. 320

7.3.5.4 Time Rate of Settlement in Cohesive Soils ... 325

7.3.5.5 Group Settlement in Layered Soils ............... 326

7.3.5.6 Group Settlement Using the Janbu Tangent 

Modulus Approach...................................... 328

7.3.5.7 Group Settlement Using the Neutral Plane 

Method....................................................... 334

7.3.6 Settlement Due to Downdrag..................................... 336
vii

7.3.6.1 Recommended Approach for Downdrag ....... 341

7.3.6.2 Methods for Reducing Downdrag and Drag 

Force ......................................................... 348

7.3.7 Horizontal Pile Foundation Deflection......................... 351

7.3.7.1 Pile Head Fixity........................................... 353

7.3.7.2 Lateral Design Methods............................... 355

7.3.7.3 p-y Method ................................................. 356

7.3.7.4 Strain Wedge Method.................................. 362

7.3.7.5 Single Piles ................................................ 364

7.3.7.6 Pile Groups ................................................ 364

7.3.7.6.1 Lateral Resistance Increases Through 

Ground Improvement ........................... 371

7.3.8 Lateral Squeeze of Foundation Soil and Solutions ...... 374

7.3.9 Overall Stability ........................................................ 375

7.4 EXTREME EVENT LIMIT STATES ....................................... 376

7.4.1 Extreme Event Scour During Check Flood.................. 376

7.4.2 Seismic and Seismic Induced Downdrag .................... 377

7.4.2.1 AASHTO Recommendations for the 

Equivalent Static Seismic Force ................... 378

7.4.2.2 Liquefaction................................................ 382

7.4.3 Ice and Collisions ..................................................... 385

7.4.3.1 Ice Loads ................................................... 386

7.4.3.2 Vehicle Collison .......................................... 387

7.4.3.3 Vessel Collision .......................................... 387

7.4.4 Combined Extreme Events ........................................ 388

7.5 DETERMINATION OF MINIMUM PILE PENETRATION......... 388

7.6 DETERMINATION OF Rndr TO ESTABLISH CONTRACT 

DRIVING CRITERIA ............................................................ 390

7.7 DRIVABILITY ANALYSIS ..................................................... 391

7.7.1 Factors Affecting Drivability ....................................... 392

7.7.2 Methods for Determining Pile Drivability ..................... 393

7.7.3 Drivability versus Pile Type........................................ 395

7.8 CONSIDERATIONS FOR BATTER PILE DESIGN AND 

CONSTRUCTION................................................................ 395

7.9 CORROSION AND DETERIORATON................................... 397

7.10 ADDITIONAL DESIGN AND CONSTRUCTION 

CONSIDERATIONS............................................................. 398

7.10.1 Minimum Pile Spacing, Clearance, and Cap 

Embedment.............................................................. 398

7.10.1.1 Special Considerations for Large Pile Sizes .. 399
viii

7.10.2 Identification of High Rebound Soils........................... 400

7.10.3 Soil and Pile Heave................................................... 400

7.10.4 Piles Driven Through Embankment Fills ..................... 402

7.10.5 Effect of Predrilling, Jetting and Vibratory Installation 

on Nominal Resistance ............................................. 403

7.10.6 Densification Effects on Nominal Resistance and 

Installation Conditions ............................................... 404

7.10.7 Plugging of Open Pile Sections.................................. 405

REFERENCES....................................................................................... 411

8 STRUCTURAL ASPECTS AND LIMIT STATES............................... 427

8.1 INTRODUCTION ................................................................. 427

8.2 BASIC STRUCTURAL PROPERTIES OF DRIVEN PILES...... 428

8.2.1 Material Properties.................................................... 428

8.2.2 Pile Section Definitions.............................................. 430

8.2.3 Effective Length and Buckling.................................... 433

8.3 STRUCTURAL CONSIDERATIONS AND RESISTANCE 

FACTORS........................................................................... 434

8.3.1 Depth to Fixity .......................................................... 435

8.3.2 Limiting Slenderness Ratio ........................................ 436

8.3.3 Resistance Factors ................................................... 436

8.4 TIMBER PILES.................................................................... 439

8.4.1 Driving Stresses ....................................................... 440

8.4.2 Structural Resistance ................................................ 441

8.4.2.1 Axial Compression Parallel to Grain ............. 441

8.4.2.2 Flexure....................................................... 443

8.4.2.3 Combined Flexure and Axial Compression ... 444

8.5 STEEL PILES...................................................................... 445

8.5.1 Driving Stresses ....................................................... 445

8.5.2 Structural Resistance ................................................ 445

8.5.2.1 Axial Compression...................................... 445

8.5.2.2 Flexure....................................................... 449

8.5.2.3 Combined Axial Compression and Flexure ... 452

8.5.2.4 Shear......................................................... 453

8.5.3 Example Calculations for H Pile Structural 

Resistance. .............................................................. 454

8.6 CONCRETE PILES.............................................................. 459

8.6.1 Driving Stresses ....................................................... 459

8.6.2 Structural Resistance ................................................ 460

8.6.2.1 Axial Compression...................................... 460
ix

8.6.2.2 Biaxial Flexure ............................................ 461

8.7 COMPOSITE PILES ............................................................ 464

8.7.1 Driving Stress........................................................... 464

8.7.2 Structural Resistance ................................................ 464

8.7.2.1 Axial Compression...................................... 464

8.8 LAYOUT OF PILE GROUPS ................................................ 467

8.9 PRELIMINARY DESIGN OF PILE BENT AND GROUP 

CAPS.................................................................................. 470

8.9.1 Cap Considerations for Large Pile Sizes..................... 481

REFERENCES....................................................................................... 482

A LIST OF FHWA/ NHI RESOURCES RELEVANT TO DEEP 

FOUNDATIONS ........................................................................... 485

B LIST OF ASTM AND AASHTO PILE DESIGN AND TESTING 

SPECIFICATIONS........................................................................ 491

C PILE HAMMER INFORMATION .................................................... 495
x 
xi

LIST OF TABLES

Table 2-1 Limit State, Load Case, and Load Combination (after 

AASHTO 2014)...................................................................... 8

Table 2-2 Load Combinations and Load Factors (after AASHTO 2014)...10

Table 2-3 Load Factors for Permanent Loads (after AASHTO 2014).......11

Table 2-4 Design Stage Communication ...............................................40

Table 2-5 Construction Stage Communication.......................................41

Table 3-1 Foundation Information Sources Provided by the Federal 

Highway Administration ........................................................45

Table 3-2 Foundation Types and Typical Uses (Modified from Bowles 

1977)...................................................................................48

Table 3-3 Pile Length Estimates Based on Field Methods and 

Factored Resistances ...........................................................59

Table 3-4 Compression Factor, α, for Sand Based on Soil Density and 

Level of Driving Energy (after Massarsch and Fellenius 

2014)...................................................................................67

Table 4-1 Subsurface Exploration Phases.............................................81

Table 4-2 Sources of Subsurface Information and Use ..........................82

Table 4-3 Minimum Number of Exploration Points per Substructure 

(modified from Sabatini et al. 2002) .......................................87

Table 4-4 Soil Boring Methods .............................................................89

Table 4-5 Summary of Rock Coring Methods ........................................92

Table 4-6 Rock Quality Designation......................................................94

Table 5-1 Field and Laboratory Tests for Geomaterial Parameter 

Determination .................................................................... 100

Table 5-2 Summary of In-Situ Methods ............................................... 103

Table 5-3 Soil Index Tests used for Driven Pile Design (modified from 

Brown et al. 2010).............................................................. 116

Table 5-4 Typical Values of Sensitivity (after Sowers 1979).................. 117

Table 5-5 AASHTO (2014) Correlation Between SPT (N1)60 values to 

Drained Friction Angle of Granular Soils (modified after 

Bowles 1997)..................................................................... 123

Table 5-6 Correlation Between Relative Density, SPT N value, and 

Internal Friction Angle for Cohesionless Soils in GEC-5 

(2002) (after Meyerhof 1956) .............................................. 123
xii

Table 5-7 Correlation Between Relative Density, CPT Cone 

Resistance, and Angle of Internal Friction for Clean Sands 

(after Meyerhof 1976)......................................................... 125

Table 5-8 Unconfined Compressive Strength of Particles for Rockfill 

Grades in Figure 5-12......................................................... 126

Table 5-9 Empirical Values for Unconfined Compressive Strength, qu, 

and Consistency of Cohesive Soils Based on Uncorrected 

N- Value (after Bowles 1977) .............................................. 129

Table 5-10 Estimating Soil Modulus, Es, Based on Soil Type (after 

AASHTO 2014).................................................................. 131

Table 5-11 Estimating Soil Modulus, Es, from SPT N value (after 

AASHTO 2014).................................................................. 132

Table 5-12 Estimating Soil Modulus, Es, from Cone Resistance, qc 

(after AASHTO 2014) ......................................................... 132

Table 5-13 Usual Types of Soft Rocks (after Kanji 2014). ...................... 139

Table 5-14 Coefficients of Variation for Geotechnical Properties and InSitu Tests (after Duncan and Wright 2005).......................... 141

Table 5-15 Pile Foundation Decisions Influenced by Subsurface 

Information ........................................................................ 143

Table 5-16 Information Included in Geotechnical Data Reports (after 

Brown et al. 2010).............................................................. 144

Table 6-1 Timber Piles Technical Summary ........................................ 157

Table 6-2 Steel H-Piles Technical Summary ....................................... 158

Table 6-3 Steel Pipe Piles Technical Summary ................................... 159

Table 6-4 Precast, Prestressed Concrete Technical Summary ............. 160

Table 6-5 Monotube Pile Technical Summary ..................................... 161

Table 6-6 Tapertube Piles Technical Summary ................................... 162

Table 6-7 Composite Piles Technical Summary................................... 163

Table 6-8 Preservative Retention Requirements (after Collin 2002) ...... 195

Table 6-9 Pile Type Selection Based on Subsurface and Hydraulic 

Conditions. ........................................................................ 198

Table 6-10 Pile Type Selection Based on Pile Shape Effects. ................ 199

Table 6-11 CALTRANS, 2014 Contract Cost Data................................. 200

Table 6-12 FDOT 2015 Historical Cost Information................................ 201

Table 6-13 INDOT 2015 Unit Price Summaries ..................................... 202

Table 6-14 MDSHA Price Index Published July 2015............................. 203

Table 6-15 2015 NCDOT Bid Quantities and Averages.......................... 204

Table 6-16 2015 PennDOT Bid Quantities and Averages....................... 205

Table 6-17 TxDOT 12-Month Project Bid Averages ............................... 206
xiii

Table 7-1 Resistance Factors for Static Analysis Methods Presented 

in this Manual (modified from AASHTO 2014) ...................... 224

Table 7-2 Resistance Factors for Field Determination Methods (after 

AASHTO 2014).................................................................. 226

Table 7-3 Summary of Static Analysis Methods in GEC-12 and 

AASHTO (2014) for Determination of Nominal Resistance .. 228

Table 7-4 Methods of Static Analysis for Piles in Cohesionless Soils .... 230

Table 7-5 Methods of Static Analysis for Piles in Cohesive Soils........... 231

Table 7-6 Design Table for Evaluating Kδ for Piles when ω = 0° and V 

= 0.10 to 1.00 ft3

/ft.............................................................. 240

Table 7-7 Design Table for Evaluating Kδ for Piles when ω = 0° and V 

= 1.0 to 10 ft3/ft.................................................................. 241

Table 7-8 Design Parameter Guidelines for Cohesionless Siliceous 

Soil (after API 1993) .......................................................... 253

Table 7-9 Approximate Range of β and Nt Coefficients (after Fellenius 

2014)................................................................................. 256

Table 7-10 Input Factors for Brown’s Method ........................................ 260

Table 7-11 Cs Values for Eslami and Fellenius Method (after Fellenius 

2014)................................................................................. 261

Table 7-12 CPT Cf VALUES ................................................................ 264

Table 7-13 Published Nominal Shaft Resistance Values in Weak Rock 

Materials............................................................................ 270

Table 7-14 Published Nominal Toe Resistance Values in Weak Rock 

Materials............................................................................ 271

Table 7-15 Summary of Computer Analysis Software for Axial Single 

Pile Analysis ...................................................................... 275

Table 7-16 Soil Setup Factors (after Rausche et al. 1996) ..................... 299

Table 7-17 Hydraulic Design, Scour Design, and Scour Design Check 

Flood Frequencies per HEC-18 (after Arneson et al. 2012) ... 305

Table 7-18 Tolerable Movement Criteria for Bridges.............................. 310

Table 7-19 Time Factors for Settlement ................................................ 326

Table 7-20 Typical Modulus and Stress Exponent Values (after 

Canadian Geotechnical Society 1985) ................................. 333

Table 7-21 Representative Values of ε50 for Clays................................. 358

Table 7-22 Representative Modulus of Subgrade Reaction Values, ks

and kc, for Clays and Sands ................................................ 359

Table 7-23 Laterally Loaded Pile Group Studies.................................... 367

Table 7-24 Summary of Ground Improvement Method and Increase in 

Lateral Resistance (after Rollins and Brown 2011)................ 373
xiv

Table 7-25 Summary of Ground Improvement Method and Associated 

Costs for Increase in Lateral Resistance after (Rollins and 

Brown 2011) ...................................................................... 373

Table 7-26 Site Class Definition ........................................................... 379

Table 7-27 Site Factor Values, Fpga, at Zero Period Acceleration ............ 380

Table 7-28 Site Factor Values, Fa, for Short Period Coefficient Ss........... 380

Table 7-29 Site Factor Values, Fv, for Long Period Coefficient, S1 .......... 380

Table 7-30 Seismic Zones ................................................................... 382

Table 7-31 Blow Count Correction, Ncorr, for the Equivalent Clean Sand 

Blow Count, (N1)60-e ............................................................ 384

Table 7-32 Summary of Load and Resistance Information ..................... 390

Table 8-1 Common Steel Pipe Pile Grades and Yield Stress................ 429

Table 8-2 Common Steel H-pile Grades and Yield Stress .................... 429

Table 8-3 Timber Pile Compression Stress and Elastic Modulus 

Reference Values (after AASHTO 2014).............................. 430

Table 8-4 Rate of Increase of Soil Modulus with Depth for Sands 

(ksi/ft) (after AASHTO 2014) .............................................. 436

Table 8-5 Resistance Factors During Pile Driving ................................ 438

Table 8-6 Structural Limit Resistance Factors for Piles in Compression 438

Table 8-7 Reference Design Values for Timber Piles, ksi (after 

AASHTO 2014).................................................................. 439

Table 8-8 Time Effect Factors (after AASHTO 2014) ........................... 439
xv

LIST OF FIGURES

Figure 2-1 Structural analysis of bridge used to establish foundation 

force effects (modified from Brown et al. 2010).......................12

Figure 2-2 Example of factored load calculation......................................13

Figure 2-3 Driven pile design and construction process...........................19

Figure 3-1 Situations in which deep foundations may be needed 

(modified from Vesic 1977). ..................................................53

Figure 3-2 Soil profile and nominal geotechnical resistances versus 

depth...................................................................................56

Figure 3-3 Pile cost versus pile penetration depth...................................57

Figure 3-4 Nominal resistance of pile support cost versus penetration 

depth...................................................................................57

Figure 3-5 Nominal and factored resistances for axial compression 

loads. ..................................................................................58

Figure 3-6 Nominal resistance and pile support cost...............................60

Figure 3-7 Nominal and factored resistances for axial tension loads.........61

Figure 3-8 Predicted vibration levels for class II and class III soils 

(after Bay 2003). ..................................................................65

Figure 3-9 Settlement area and magnitude around a pile in 

homogeneous sand deposit (after Massarsch and Fellenius 

2014)...................................................................................67

Figure 3-10 Corrosion and resulting buckling of foundation pile 

(courtesy Wisconsin Department of Transportation)................69

Figure 3-11 Concrete pile damage from corrosion effects on concrete 

pile reinforcement (courtesy of Moser 2011)...........................70

Figure 3-12 Concrete surface abrasion and deterioration in tidal zone 

(courtesy of Moser 2011)......................................................71

Figure 3-13 Examples of timber pile deterioration (courtesy Bridge 

Engineering Center at Iowa State University). ........................72

Figure 3-14 Tilting columns adjacent to stockpile (courtesy DelDOT). ........73

Figure 3-15 Cracked pile cap from lateral soil movements (courtesy 

DelDOT). .............................................................................74

Figure 4-1 Typical field reconnaissance form..........................................84

Figure 4-2 Standard split spoon sampler. ...............................................91

Figure 4-3 Shelby tubes (after Mayne et al. 2002)...................................91

Figure 4-4 Rock core samples...............................................................93
xvi

Figure 5-1 Schematic of common of in-situ tests (after Mayne et al. 

2001)................................................................................. 102

Figure 5-2 Standard Penetration Test schematic (after Mayne et al. 

2001)................................................................................. 104

Figure 5-3 Standard Penetration Test hammer types. ........................... 105

Figure 5-4 Instrumented 2 foot long AW rod atop drill string for SPT 

hammer energy measurements........................................... 106

Figure 5-5 Adjacent borings with different SPT hammer types (after 

Finno 1989). ...................................................................... 107

Figure 5-6 Cone penetrometers (after Mayne et al. 2001)...................... 110

Figure 5-7 CPT rig.............................................................................. 110

Figure 5-8 Soil behavior type classification chart based on normalized 

CPT data (after Robertson 1990)......................................... 112

Figure 5-9 Vane shear test schematic (after Mayne et al. 2001)............. 113

Figure 5-10 Typical ranges of friction angle for rockfills, gravels and 

sands (Note: 1 kPA = 0.145 psi) (after Terzaghi et al. 

1996)................................................................................. 126

Figure 5-11 Relationship between φ’ and PI (after Terzaghi et al. 1996)... 127

Figure 5-12 Soil stress strain curve and 50% secant modulus (after 

Brown et al. 2010).............................................................. 131

Figure 5-13 Plot of Void Ratio vs Vertical Effective Stress from 

consolidation test. .............................................................. 133

Figure 6-1 Pile classification chart. ...................................................... 156

Figure 6-2 Timber pile typical illustration. ............................................. 157

Figure 6-3 Steel H-pile typical illustration.............................................. 158

Figure 6-4 Steel pipe piles typical illustration. ....................................... 159

Figure 6-5 Precast, prestressed concrete typical illustration................... 160

Figure 6-6 Monotube pile typical illustration. ......................................... 161

Figure 6-7 Tapertube pile typical illustration.......................................... 162

Figure 6-8 Composite piles typical illustration. ...................................... 163

Figure 6-9 Timber piles. ...................................................................... 164

Figure 6-10 H-piles with driving shoes. .................................................. 166

Figure 6-11 Typical 16 inch closed end pipe pile..................................... 169

Figure 6-12 42 inch diameter, spiral weld, open end, pipe for main river 

pier.................................................................................... 171

Figure 6-13 Large diameter open end pipe piles. .................................... 171

Figure 6-14 Constrictor plate................................................................. 172

Figure 6-15 Constrictor plate installed inside pipe pile............................. 172

Figure 6-16 Tapered Monotube section (right) with add-on sections 

(left). ................................................................................. 173
xvii

Figure 6-17 Tapertube piles (courtesy DFP Foundation Products, LLC). .. 174

Figure 6-18 Spin Fin pile load transfer illustration. .................................. 175

Figure 6-19 30 inch diameter Spin Fin pile. ............................................ 176

Figure 6-20 Typical prestressed concrete piles....................................... 177

Figure 6-21 Square prestressed concrete piles....................................... 177

Figure 6-22 Typical spun-cast concrete cylinder pile section.................... 179

Figure 6-23 Concrete cylinder pile. ........................................................ 180

Figure 6-24 Precast concrete piles each with H-pile stinger..................... 181

Figure 6-25 Steel pipe - H-pile composite piles....................................... 182

Figure 6-26 Soil sampling and testing protocol for corrosion assessment 

of steel piles in non-marine applications............................... 189

Figure 6-27 Procedure for uniform or macrocell corrosion assessment of 

steel piles in non-marine applications................................... 190

Figure 6-28 Procedure for determination of electrochemical testing, 

corrosion monitoring and corrosion mitigation techniques...... 191

Figure 6-29 Loss of thickness by corrosion for steel piles in seawater 

(after Morley and Bruce 1983)............................................. 193

Figure 7-1 Compaction of cohesionless soils during pile driving (after 

Broms 1966). ..................................................................... 215

Figure 7-2 Disturbance of cohesive soils during pile driving (after 

Broms 1966). ..................................................................... 216

Figure 7-3 Situation where two static analyses are necessary due to 

scour. ................................................................................ 218

Figure 7-4 Situation where two static analyses are necessary due to fill 

materials............................................................................ 218

Figure 7-5 Typical load transfer profiles. .............................................. 220

Figure 7-6 Effective stress diagram – water table below ground 

surface. ............................................................................. 222

Figure 7-7 Effective stress diagram – water table above ground 

surface. ............................................................................. 222

Figure 7-8 Nordlund’s general equation diagram for nominal 

resistance. ......................................................................... 234

Figure 7-9 Relationship of δ/φ and pile soil displacement, V, for various 

pile types (after Nordlund 1979). ......................................... 235

Figure 7-10 Design curve for evaluating Kδ for piles when φ = 25 ̊ (after 

Nordlund 1979). ................................................................. 235

Figure 7-11 Design curve for evaluating Kδ for piles when φ = 30 ̊ (after 

Nordlund 1979). ................................................................. 236

Figure 7-12 Design curve for evaluating Kδ for piles when φ = 35 ̊ (after 

Nordlund 1979). ................................................................. 237
xviii

Figure 7-13 Design curve for evaluating Kδ for piles when φ = 40 ̊ (after 

Nordlund 1979). ................................................................. 237

Figure 7-14 Correction factor for Kδ when δ ≠ φ (after Nordlund 1979)...... 238

Figure 7-15 Relationship between maximum unit toe resistance and 

friction angle for cohesionless soils (after Meyerhof 1976)..... 238

Figure 7-16 Chart for estimating αt coefficient and bearing capacity 

factor N’q (after Bowles 1977)............................................. 239

Figure 7-17 Adhesion values for piles in cohesive soils (after Tomlinson 

1979)................................................................................. 248

Figure 7-18 Adhesion factors for driven piles in clay (Tomlinson 1980)..... 249

Figure 7-19 Penetrometer design curves for pile shaft friction in sand 

(after FHWA Implementation Package, FHWA-TS-78-209). . 265

Figure 7-20 Design curve for pile shaft friction in clay (after 

Schmertmann 1978). .......................................................... 265

Figure 7-21 Illustration of Nottingham and Schmertmann Procedure for 

estimating pile toe resistance (FHWA-TS-78-209). ............... 266

Figure 7-22 Bearing capacity factors for foundations on rock (modified 

from Pells and Turner 1980)................................................ 273

Figure 7-23 Soil profile for computer software comparison. ..................... 276

Figure 7-24 Nominal resistance from DRIVEN program using FHWA 

method. ............................................................................. 276

Figure 7-25 Nominal resistance from APILE program using FHWA 

Method. ............................................................................. 277

Figure 7-26 Nominal resistance from UNIPILE program using Beta 

Method. ............................................................................. 277

Figure 7-27 Nominal resistance versus depth for three pile types 

calculated by the APILE program with the FHWA Method. .... 278

Figure 7-28 Design chart of nominal resistance, Rn, and factored 

resistance, Rr, versus depth for one pile type with different 

field control methods. ......................................................... 280

Figure 7-29 Overlap of stress zones for friction pile group (after Bowles 

1988)................................................................................. 283

Figure 7-30 Measured dissipation of excess pore water pressure in soil 

surrounding full scale pile groups (after O’Neill 1983)............ 285

Figure 7-31 Three dimensional pile group configuration (after Tomlinson 

1994)................................................................................. 287

Figure 7-32 Design chart for nominal and factored resistance in axial 

tension. ............................................................................. 290

Figure 7-33 Uplift of pile group in cohesionless soil (after Tomlinson 

1994)................................................................................. 292
xix

Figure 7-34 Uplift of pile group in cohesive soils (after Tomlinson 1994)... 293

Figure 7-35 Excess pore water pressure due to pile driving (after 

Poulos and Davis 1980)...................................................... 300

Figure 7-36 Local and channel degradation scour................................... 305

Figure 7-37 Angular distortion due to non-uniform bridge settlement........ 309

Figure 7-38 Example settlement and angular distortion profile (after 

Modjeski and Masters 2015). .............................................. 311

Figure 7-39 Settlement profile with design differential settlement and 

design angular distortion from S-0 concept (after Modjeski 

and Masters 2015). ............................................................ 312

Figure 7-40 Construction point concept: (a) identification of key 

construction points, (b) estimated load-deformation behavior 

(after Modjeski and Masters 2015)....................................... 313

Figure 7-41 Settlement profile with angular distortion from construction 

point concept (after Modjeski and Masters 2015).................. 314

Figure 7-42 Stress zone from single pile and pile group (after Tomlinson 

1994)................................................................................. 316

Figure 7-43 Values of the bearing capacity index, C', for granular soil 

(data from Hough 1959)...................................................... 318

Figure 7-44 Equivalent footing concept. ................................................. 320

Figure 7-45 Stress distribution below equivalent footing for a pile group 

in firm clay (after Duncan and Buchignani, 1976).................. 321

Figure 7-46 Stress distribution below equivalent footing for a pile group 

(modified from Cheney and Chassie 2000)........................... 322

Figure 7-47 Plot of void ratio vs. vertical effective stress from 

consolidation test. .............................................................. 323

Figure 7-48 Non-linear relation between stress and strain in soil (after 

Fellenius 1990). ................................................................. 329

Figure 7-49 Neutral plane (after Goudrealt and Fellenius 1994)............... 335

Figure 7-50 Calculated load versus depth in 12.75 inch O.D. concrete 

filled pipe pile as a function of time (after MnDOT 

Geotechnical Manual)......................................................... 338

Figure 7-51 Change in neutral plane, negative shaft resistance, and 

drag force during transition to geotechnical strength limit 

(after MnDOT Geotechnical Manual). .................................. 339

Figure 7-52 Conceptual illustration of soil and pile movement (left) and 

resulting neutral plane and pile forces (right) (adapted from 

Siegel et al. 2013). ............................................................. 339

Figure 7-53 Common downdrag situation due to fill weight...................... 340

Figure 7-54 Common downdrag situation due to ground water lowering... 340
xx

Figure 7-55 Plot of static analysis results (after Siegel et al. 2013)........... 342

Figure 7-56 Plot of normalized toe resistance versus toe movement 

(after Siegel et al. 2013)...................................................... 343

Figure 7-57 Axial load and resistance plot of cumulative shaft resistance 

vs. depth (after Siegel et al. 2013)....................................... 344

Figure 7-58 Axial load and resistance plot of unfactored permanent load 

plus cumulative shaft resistance vs depth (after Siegel et al. 

2013)................................................................................. 345

Figure 7-59 Axial load and resistance plot including mobilized 

resistances (after Siegel et al. 2013)................................... 345

Figure 7-60 Axial Load and resistance plot including neutral plane 

location based on mobilized toe resistance (after Siegel et 

al. 2013). ........................................................................... 346

Figure 7-61 Geofoam block approach embankment (courtesy MnDOT). .. 349

Figure 7-62 Over-sized collar for bitumen coating protection. .................. 350

Figure 7-63 Soil resistance to a lateral pile load (after Smith 1989). ......... 352

Figure 7-64 Effect of pile head fixity on translation at service limit state 

(after Wilson et al. 2006)..................................................... 354

Figure 7-65 Effect of pile head fixity on moment in piles at strength limit 

state (after Wilson et al. 2006)............................................ 354

Figure 7-66 Typical lateral analysis pile-soil model. ................................ 357

Figure 7-67 Typical p-y curves for ductile and brittle soil (after Coduto 

1994)................................................................................. 358

Figure 7-68 Graphical presentation of p-y analysis results (after Reese 

1986)................................................................................. 361

Figure 7-69 Comparison of measured and COM624P predicted loaddeflection behavior versus depth (after Kyfor et al. 1992). ..... 361

Figure 7-70 Strain wedge developed in soil (after Ashour et al. 1998a). ... 362

Figure 7-71 Proposed geometry of compound passive wedge (after 

Ashour et al. 1998a)........................................................... 363

Figure 7-72 Soil-pile interaction with multiple soil layers (after Ashour et 

al. 1998a). ......................................................................... 364

Figure 7-73 Illustration of p-multiplier concept for lateral group analysis. .. 366

Figure 7-74 Typical plots of (a) Load versus deflection and (b) Bending 

moment versus deflection for pile group analysis (adapted 

from Brown and Bollman 1993). .......................................... 369

Figure 7-75 Ground improvement treatment areas for increased lateral 

resistance of pile groups in weak soils (Rollins and Brown 

2011)................................................................................. 372

Figure 7-76 Examples of abutment tilting due to lateral squeeze.............. 375
xxi

Figure 7-77 Design response spectrum (after AASHTO 2014)................. 381

Figure 7-78 Correlation between the Residual Undrained Strength 

Ratio, sr / σ’vo and equivalent clean sand SPT blow count, 

(N1)60-e (Idriss and Boulanger 2007). .................................... 383

Figure 7-79 Application of equivalent impact force (after AASHTO 

2014)................................................................................. 388

Figure 7-80 Variation of foundation response depending on group 

configuration and batter (after Wilson et al. 2006)................. 396

Figure 7-81 Balance of forces on pile subject to heave (after Haggerty 

and Peck 1971).................................................................. 402

Figure 7-82 Variation in expected static resistance from APILE API 

method based on plugged, unplugged, and automatic 

calculated plugging............................................................. 408

Figure 7-83 Plugging of open end pipe piles (after Paikowsky and 

Whitman 1990)................................................................... 409

Figure 7-84 Plugging of H-piles (after Tomlinson 1994)........................... 409

Figure 8-1 H-pile section dimensions. .................................................. 430

Figure 8-2 Table of H-pile section properties (modified from Skyline 

Steel 2015). ....................................................................... 431

Figure 8-3 Pipe pile section dimensions. .............................................. 431

Figure 8-4 Effective length factors, K (after AASHTO 2014)................... 434

Figure 8-5 Moment interaction diagram................................................ 463

Figure 8-6 Pile group layout. ............................................................... 469

Figure 8-7 Cap section and notation. ................................................... 472

Figure 8-8 Critical punching (two-way) shear sections around column.... 474

Figure 8-9 Critical section from overlapping critical perimeters (after 

AASHTO 2014).................................................................. 475

Figure 8-10 Pile punching shear for: (a) circular pile; (b) rectangular H 

pile; (c) circular pile near corner of cap; and (d) two nearby 

circular piles....................................................................... 476

Figure 8-11 One-way beam shear critical sections from column............... 478

Figure 8-12 One-way beam shear critical sections from pile(s). ............... 478
xxii
xxiii

LIST OF SYMBOLS

A - Pile cross sectional area (7.3).

Ab - Brown’s regression analysis factor based on soil type (7.2).

Ad - Angular distortion (7.3).

Adm - Modified angular distortion (7.3).

Ap - Pile toe area (7.1); 

Cross sectional area of pile material at pile toe (7.2) (7.10).

App - Cross sectional area of soil plug in open end pipe or H-piles at pile toe 

(7.2); Cross sectional area of pile and soil plug at pile toe (7.10).

Aps - Cross sectional area of prestressing steel (8.6).

As - Pile shaft surface area (7.1) (7.2) (7.10); 

Peak seismic ground acceleration coefficient modified by short-period 

site factor (7.4).

Asi - Pile shaft interior surface area (7.10).

Aso - Pile shaft exterior surface area (7.10).

Ast - Cross sectional area of steel (8.5). (8.7) (9.5).

Astc - Cross sectional area of compression reinforcing steel (8.9).

Astr - Cross sectional area of longitudinal reinforcing steel (8.6) (8.7).

Astt - Cross sectional area of tension reinforcing steel (8.9).

Astv - Area of transverse reinforcement within distance, s (8.9).

A1 - Constant based on soil type and subsurface condition (7.2).

A2 - Constant based on pile type (7.2).

a(t) - Acceleration measured at the gage location (10.4).

B - Width of pile group (7.2) (7.3); 

Bb - Database calibrated regression factor (7.2).

b - Pile width or diameter (3.5) (7.1) (8.4);

Width/ Height of square (8.2);

Depth of beam or width of dimension lumber (8.4);

Width of the compression face (8.9).

bc - Column side for square columns (8.9).

bf - Flange width of pile section (8.2) (8.5).

bo - Critical punching shear perimeter (8.9).

bv - Width of interface (8.9).

C - Perimeter of pier excluding half circles at ends of oblong pier (7.4). 
xxiv

C’ - Dimensionless bearing capacity index, determined from average 

corrected SPT N value, for layer with consideration of SPT hammer type

(7.3).

Ca - Pile adhesion (7.2) (7.10). 

Cc - Compression index (5.2) (7.3).

Cd - Pile perimeter at depth d (7.2);

Deck factor for timber pile structural resistance (8.4).

CF - Correction factor for Kδ when δ ≠ φ (7.2).

CF - Size factor for timber pile structural resistance (8.4).

Cf - Conversion factor for cone tip resistance to sleeve friction (7.2).

Cfu - Flat-use factor for timber pile structural resistance (8.4).

Ch - Pore water pressure dissipation factor (7.2).

Ci - Incising factor for timber pile structural resistance (8.4).

CM - Wet service factor for timber pile structural resistance (8.4).

Cn - Correction factor for SPT N value (5.1).

Cp - Toe correction coefficient for Eslami and Fellenius CPT Method (7.2).

Cr - Recompression index (5.2) (7.3).

Cs - Swell index (5.2); 

Shaft correlation coefficient for Eslami and Fellenius CPT Method (7.2). 

Csm - Elastic seismic response coefficient (7.4).

Cv - Coefficient of consolidation (5.2) (7.3); 

Volume factor for timber pile structural resistance (8.4).

Cα - Secondary compression index (5.2).

Cλ - Time effect factor for timber pile structural resistance (8.4).

C1 - Composite column constant 1 (8.7).

C2 - Composite column constant 2 (8.7).

C3 - Composite column constant 3 (8.7).

C4 - Damping constant for Statnamic test (11.4).

c - Cohesion (5.2) (7.1); 

Distance from centroid to outer edge (8.2); 

Distance between the neural axis and the compressive face (8.9).

c’ - Effective cohesion (5.2).

co - Column diameter (8.9).

c1 - Small column side for rectangular columns (8.9).

c2 - Large column side for rectangular columns (8.9).

D - Pile embedded length (3.5) (7.2) (7.3) (7.10); 

Pile penetration below the rock surface (7.2).

Outside pile diameter (8.2); 

DA - Design angular distortion (7.3).

DWT - Deadweight tonnage (7.4).
xxv

DB - Pile embedded length into bearing stratum (7.3).

Dil - Depth from reference to top of incompressible layer (7.3).

Dnp - Depth from reference to neutral plane (7.3).

Dr - Relative density (5.1) (5.2).

Dδ - Design differential settlement (7.3).

D’ - Effective depth, 2/3 pile embedded length (7.1).

D1 - Inside pile diameter (8.2).

d - Equilibrium depth (7.10).

dc - Clear space (8.9).

df - Depth to fixity below the ground (8.3); Pile embedment into cap (8.9).

ds - Distance to center of steel (8.9).

dsc - Distance from extreme compression fiber to the centroid of the 

compression reinforcement (8.9).

dst - Distance from extreme compression fiber to the centroid of the tensile 

reinforcement (8.9).

dv - Vane diameter (for Vane Shear Test) (5.2);

Effective depth to reinforcement (8.9).

dw - Web depth of pile section (8.5).

E - Elastic modulus of material (Youngs modulus) (5.2) (7.3); 

Elastic modulus of pile material (7.3) (8.3). 

Ec - Elastic modulus of concrete (8.2).

Ee - Modified elastic modulus of steel for composite column (8.7).

Em - Rock Mass Modulus (5.1).

En - Rated hammer energy (5.1).

Eo - Timber reference value for elastic modulus (8.2). 

ER - SPT Hammer efficiency as determined by energy measurements in 

accordance with ASTM D4633 (5.1).

Er - Manufacturers rated hammer energy (3.5); 

Modulus of Intact Rock (5.1).

Es - Elastic modulus of soil (5.1) (5.2) (7.3) (8.3) . 

Esm - Secant modulus (7.3). 

Est - Elastic modulus of steel (8.5); 

Elastic modulus of prestressing steel (8.6).

E50 - 50% secant modulus (5.4).

e - Void ratio (7.3).

eo - Initial soil layer void ratio (5.2.4) (7.3).

F - Vertical force (7.4).

Fa - Short period site factor (7.4); 

Fb - Adjusted timber pile structural flexural resistance (8.4). 

Fbo - Timber reference value for strength in flexure (8.4).
xxvi

Fc - Adjusted timber pile structural axial resistance (8.4).

Fco - Timber reference value for compressive stress parallel to grain (8.2) 

(8.4). 

Fe - Nominal compressive resistance of composite section (8.7).

Fp - Plug mobilization factor (7.2).

FPGA - Zero period site factor (7.4).

Fv - Long period site factor (7.4). 

Fvo - Timber reference value for strength in shear (8.4).

Fvs - Factor for pile driving method (1.0 for impact or 0.68 for vibratory) (7.2).

Fy - Yield stress of steel (8.2) (8.5) (8.7) (8.8).

Fyf - Minimum yield strength of lower strength flange (8.5).

Fyr - Yield stress of reinforcing steel (8.6) (8.7).

Fyw - Yield stress of web (8.5).

fc - Consolidation factor (non-dimensional regression factor) (7.2).

f’c - Ultimate compression strength for concrete, 

Concrete compressive strength at 28 days (8.2) (8.6) (8.7).

fcr - Elastic local buckling stress (8.5).

fn - Unit negative shaft resistance (7.3).

fpe - Effective prestress in concrete (8.6).

fr - Remolding recovery rate (non-dimensional regression factor) (7.2).

fs - Unit sleeve friction; Average unit sleeve friction (5.1) (7.2). 

fs - Unit shaft resistance over the pile surface area (7.1) (7.2); 

Unit positive shaft resistance (7.3).

fsc - Stress in the mild steel compression reinforcement at nominal flexural 

resistance (8.9).

fsi - Interior unit shaft resistance (7.10).

fso - Exterior unit shaft resistance (7.10).

fst - Stress in the mild steel tension reinforcement at nominal flexural 

resistance (8.9).

f1 - Correction factor for undrained shear strength determination (5.2).

G - Shear modulus (5.2) (8.5). 

Gs - Specific Gravity (5.2).

Ho - Initial soil layer thickness (5.2) (7.3).

Hv - Maximum vertical drainage path in cohesive layer (7.3).

h - Total Thickness of sublayers (7.3); 

Distance between flange and centroid for warping torsional constant 

(8.5); Structural depth, thickness of cap less pile embedment (8.9).

hf - Height of embankment fill (7.3).

hi - Thickness of soil strata (5.2).

hv - Vane height (for Vane Shear Test) (5.2).
xxvii

hw - Height of water (pressure head) for calculation of pore water pressure 

(5.2).

I - Moment of inertia (8.2);

Weak axis moment of inertia (8.3).

If - Influence factor for group embedment (7.3).

Ix - Moment of inertia about the major principal axes of cross section (8.5).

Iy - Moment of inertia about the minor principal axes of cross section (8.5).

j - Stress exponent (7.3).

K - Effective length factor for buckling (7.2) (8.2) (8.5) (8.7).

Ko - At rest earth pressure coefficient (5.2) (7.2).

Ks - Ratio of unit pile shaft resistance to cone unit sleeve friction for 

cohesionless soils (7.2).

Kδ - Coefficient of lateral earth pressure (7.2).

kc - Modulus of subgrade reaction for cyclic lateral loading (7.3).

ks - Modulus of subgrade reaction for static lateral loading (7.3).

k1 - Regression factor (0.17 for PSC, 0.12 for CEP and 0.15 for OEP) (7.2).

k2 - Regression factor (0.00044 for PSC piles, 0.00078 for CEP, and 0.00060 

for OEP) (7.2).

L - Total pile length (7.3). 

L% - Pile length subject to heave (7.10).

Le - Effective pile length considering unbraced length (8.4). 

Lo - Embedded pile length at the time of initial driving (7.2).

Ls - Span length (7.3).

Lt - embedded pile length at time “t” after initial driving (7.2).

l - Unbraced length, or laterally unsupported length plus df (8.3) (8.5) (8.7).

Mb - Bending moment (7.3).

Mn - Nominal flexural resistance (structural) (8.4) (8.5) (8.6).

Mp - Plastic moment about the weak axis (8.5).

Mr - Factored flexural resistance (structural) (8.3) (8.5) (8.6).

Mrx - Factored flexural resistance about x-axis (8.5) (8.6).

Mry - Factored flexural resistance about y-axis (8.5) (8.6).

Mu - Factored moment load (structural) (2.3) (8.3).

Mux - Factored moment about x-axis (8.5) (8.6) (8.8).

Muy - Factored moment about y-axis (8.5) (8.6) (8.8). 

mn - Dimensionless modulus number (7.3).

mnr - Dimensionless recompression modulus number (7.3).

ms - Semilog-linear slope of Rs vs t from multiple restrike tests (7.2).

N - Uncorrected field SPT resistance value (5.1) (5.2) (5.4) (5.5) (7.2) (7.3) 

(7.4).

Na - Average SPT N-Value over pile length (7.2).
xxviii

Nc - Dimensionless bearing capacity factor (7.2).

Nk - Bearing capacity factor, typically from 15 to 20 (5.2).

Nq - Dimensionless bearing capacity factor (API Method) (7.2).

N’q - Dimensionless bearing capacity factor (7.2).

Nt - Toe resistance coefficient (7.2). 

N60 - SPT N value corrected for 60% energy transfer (5.1.1) (7.2) (7.3).

(N1)60 - SPT N value corrected for energy and overburden stress (5.1) (5.2) (7.2) 

(7.4) (7.10);

Average corrected SPT N value within a depth B below pile toe (7.3). 

(N1)60e- Equivalent clean sand blow count (7.4).

Ncorr - SPT N value corrected for percent fines (7.4.).

Nγ - Bearing capacity factor (7.2). 

n - Exponent typically equal to 1 in clays (e.g., Olsen 1997) and 0.5 in sandy 

soils (e.g. Lio and Whitman 1986) (5.1); 

Number of piles in pile group (7.2) (7.3) (8.8).

nh - Rate of increase of soil modulus with depth (8.3).

ni - Number of piles whose centers lie inside the two-way shear critical 

section (8.9).

no - Number of piles whose centers lie outside the two-way shear critical 

section (8.9).

Pe - Elastic critical buckling resistance (8.5).

Pe(x) - Equivalent static horizontal seismic force acting on superstructure (7.4).

PGA - Peak ground acceleration coefficient (7.4).

Pm - P-multiplier for p-y curve (7.3).

Pn - Nominal axial resistance (structural) (8.4.2) (8.5) (8.6) (8.7). 

Po - Equivalent nominal axial yield resistance (structural) (8.5) (8.6).

Pr - Factored axial resistance (structural) (3.4.1) (8.3) (8.5) (8.6) (8.7).

Prx - Factored axial resistance determined on the basis that only eccentricity, 

ey, is present (8.6).

Prxy - Factored axial resistance in biaxial flexure (8.6).

Pry - Factored axial resistance determined on the basis that only eccentricity, 

ex, is present (8.6).

Ps - Pile shape factor (7.2); 

Ps - Equivalent static vessel impact force (7.4). 

Pt - Pile base factor (7.2).

Pu - Factored axial load (structural) (8.3.3) (8.6) (8.8) (8.9).

Pui - Maximum single pile axial load (8.8).

Puz - Factored axial load from superstructure/substructure acting upon pile 

cap (8.8).

p - Soil resistance per unit pile length (7.3).
xxix

pa - Atmospheric pressure (5.1) (5.2).

pf - Design foundation pressure (7.3.4).

Q - Total factored load (2.3); Factored axial load (7.3) (7.6); 

Slender element reduction factor (8.5).

Qc - Scour Design Check Flood Frequency (7.2).

Qd - Hydraulic design flood frequency (7.2); 

Unfactored permanent load (7.3).

Qmax - Maximum axial compressive force in the pile (7.3).

Ql - Live load on a pile (7.3).

Qi - Force effect (2.3) (7.3) (7.4).

QL - Lateral pile load (7.3).

Qs - Scour Design Flood Frequency (7.2).

q - Surcharge (7.3).

qc - Cone tip resistance (5.1) (5.2) (7.2). 

qca - Average cone tip resistance within a depth of B below the pile toe (7.3).

qc1 - Average qc over a distance of xb below the pile toe (7.2). 

qc2 - Average qc over a distance of 8b above the pile toe (7.2).

qDMT - Dilatometer test tip resistance (5.5).

qE - Eslami cone stress (7.2).

qEg - Geometric average of the corrected cone tip resistance over the 

influence zone (7.2).

qL - Limiting unit toe resistance (7.2).

qp - Unit toe resistance (7.1) (7.2) (7.1).

qt - Corrected cone tip resistance (5.1) (7.2).

qu - Unconfined compressive strength (5.1) (5.2) (5.3) (5.5) (7.2).

R - Radius of pier (7.4).

Rf - Friction ratio or fs/qt (5.1).

Rn - Nominal resistance (2.4) (3.4) (7.1) (7.2) (7.3) (7.6) (7.10).

Nominal resistance of each individual pile in the group (7.2).

Rndr - Nominal driving resistance (3.4) (7.6). 

Rng - Nominal resistance of pile group (7.2).

Rno - Initial nominal resistance at time “to” of driving (7.2).

Rp - Nominal toe resistance (7.1) (7.2) (7.3).

Rr - Factored resistance (2.4) (7.2) (8.8) (8.9). 

Rrelax - Resistance loss from relaxation (7.6).

Rrg - Factored resistance of the pile group (7.3).

Rs - Nominal shaft resistance (3.4) (7.1) (7.2) (7.3) (12.7).

Rscour - Resistance loss from scour (7.6).

Rso - Initial shaft resistance at “to” of driving (7.2).

Rs

+ - Positive Shaft Resistance (7.3).
xxx

Rs

-

 - Negative Shaft Resistance (7.3).

Rug - Nominal uplift resistance of the pile group (7.2).

rp - Equivalent pile radius (7.2).

rs - Minimum radius of gyration (8.3);

Radius of gyration about axis normal to plane of buckling (8.5) (8.7).

S - Settlement (7.1); Estimated total settlement (7.3); 

Elastic section modulus (8.2) (8.4).

Savg - Average settlement (3.5).

Sd - Differential settlement of the foundation (7.3). 

Sdd - downdrag movement (7.3).

Sc - Settlement from primary consolidation (5.2) (7.3). 

SDS - Csm value with a period of 0.2 seconds = FaSs (7.4).

SD1 - Csm value with a period of 0.2 seconds = FvS1 (7.4).

Sh - Horizontal abutment movement (7.3).

Smax - Maximum settlement (3.5).

Sp - Pile slope (7.3).

Ss - Short period spectral acceleration (7.4).

St - Sensitivity of a cohesive soil (5.2).

Sta - Total foundation settlement (7.3).

Stp - Total foundation settlement before construction (7.3).

Str - Relevant total settlement (7.3).

Sv - Vertical fill settlement (7.3).

Sy - Elastic section modulus about weak axis (8.5).

S1 - Long period coefficient (7.4).

s - Spacing of the transverse reinforcement (8.9). 

sr - Residual shear strength (7.4). 

su - Undrained shear strength (3.4) (5.1) (5.2) (5.5) (7.2) (7.4) (8.3); 

Vane Shear Test undrained shear strength (5.5); 

Average undrained shear strength (7.3); 

Undrained shear strength of soft cohesive soil (7.3) (7.10).

T - Theoretical time factor for percentage of primary consolidation (7.3).

Tm - Period of vibration of mth mode(s) (7.4).

To - Reference period to define spectral shape = 0.2Ts (7.4).

Ts - Corner period when spectrum changes from independent to inversely 

proportional = SD1 /SDS. (7.4).

Tv - Input torque during shear (for Vane Shear Test) (5.2).

t - Time after driving (7.2); 

Time for settlement to occur (7.3); 

t - Pipe pile wall thickness (8.2) (8.5). 

tcap - Thickness of pile cap (8.9).
xxxi

tf - Flange thickness of pile section (8.2) (8.5).

ti - Ice thickness (7.4).

to - Time after driving from which the increase in resistance is linear in 

logarithmic time (days) (typically 0.5 for sand, 1.0 for clay) (7.2).

tsoil - Thickness of compressible soil beneath neutral plane (7.3).

tw - Web thickness of pile section (8.2) (8.5).

U - Displacement (7.3).

u - Pore water pressure (5.1) (5.2) (7.1).

ue - Excess pore water pressure (7.2).

uh - Hydrostatic pore water pressure (7.2).

V - Volume of soil displaced per unit length of pile (7.2); 

Vessel impact velocity (7.4). 

Vc - Nominal shear resistance provided by concrete tensile strength (8.9).

Vn - Nominal shear resistance (8.9).

Vr - Factored shear resistance (structural) (8.3).

Vs - Shear wave velocity (7.4); 

Nominal shear resistance provided by steel reinforcement (8.9). 

Vu - Factored shear load (structural) (2.3) (8.3) (8.9).

W - Pier or Abutment Width (4.2); 

Equivalent weight of the superstructure (7.4).

Wc - Estimated weight of pile cap (8.8).

Wg - Effective weight of the pile/soil block including pile cap weight (7.2).

Wp - Weight of soil plug (7.10).

Ws - Estimated weight of soil above pile cap (8.8).

w - Moisture Content (5.1) (5.2) (7.2) (7.4).

x - Distance along x-axis from the center of the column to each pile

center (8.8).

Yo - Pile head deflection (7.3).

y - Lateral soil (or pile) deflection (7.3); 

Distance along y-axis from the center of the column to each pile center 

(8.8).

Z - Length of pile group (7.2) (7.3).

Zy - Plastic section modulus about weak axis (8.5).

α - Compression factor for settlement (3.4) (3.5); 

Dimensionless adhesion factor (7.2).

α’ - Ratio of pile unit shaft resistance to cone unit sleeve friction for cohesive 

soils (7.2).

αt - Dimensionless factor in Nordlund method (7.2).

β - Bjerrum-Burland beta coefficient (7.2).

βc - Ratio of the long side to the short side of the load (8.9).
xxxii

βm - Mobilized angle for strain wedge analysis (7.3).

β1 - Stress block factor (8.9).

Δ - Elastic compression (7.3).

Δd - Length of pile segment (7.2).

ΔH - Total settlement at Pier or Abutment (7.3).

ΔH100 - Differential settlement over 100 Feet within a pier or abutment, or the 

differential settlement between piers (7.3).

Δum - Maximum excess pore pressure (7.2).

Δσ - Additional stress at mid-point of soil layer from loading (5.2);

Additional pressure from structural loading (7.3).

Δσh - Changes in deviator stress in the direction of loading (7.3).

Δε - Strain from the increase in effective stress (7.3); 

δ - Friction angle between pile and soil (7.2) (7.3); 

ε - Strain (5.2) (7.3). 

εcu - Failure strain of concrete in compression (8.6).

εv - Vertical strain (5.2).

ε50 - Strain at one half the maximum principal stress (7.3) (7.4).

ηg - Pile group efficiency (7.2).

ηi - Load modifier based on ductility, redundancy, or operation classification 

(2.3) (2.4).

γ - Total unit weight of soil (5.1) (5.2) (5.5).

γ' - Buoyant unit weight of soil (5.5).

γd - Dead Load Factor (7.3).

γf - Unit weight of embankment fill (7.3).

γi - Load factor, statistically based multiplier applied to force effect (2.3) (2.4) 

(7.4) (7.3); 

Unit weight of soil strata for calculation of in-situ stress (5.2).

γl - Load factor for force effect due to live loads (7.3).

γp - Load factor for force effect due to permanent loads (2.3) (7.3) (7.4).

γSE - Load factor for force effect due to vertical settlement (7.3).

γTG - Load factor for force effect due to temperature gradient (2.3).

γw - Unit weight of water (5.2).

λ - Normalized column slenderness factor (8.7).

λf - Slenderness ratio for flange (8.5).

λpf - Limiting slenderness ratio for a compact flange (8.5).

λrf - Limiting slenderness ratio for a non-compact flange (8.5).

ν - Poisson ratio (5.2) (5.4).

σ - Normal stress (pressure) on plane of failure, stress (5.2).

σ’ - Effective normal stress (pressure) on plane of failure (σ - u) (5.2).
xxxiii

σ’d - Vertical effective stress at the center of depth increment d (7.2).

σdr - Driving stress (8.4) (8.5) (8.6).

σ’ho - Horizontal effective stress at the sample depth (5.2).

σ’o - Effective stress prior to stress increase (7.3).

σp - Preconsolidation pressure or stress (5.1) (5.2) (7.3).

σ’p - Vertical effective stress at the pile toe (7.2).

σr - Reference stress for settlement with Janbu Tangent Modulus (7.3).

σv - Total stress (7.1). 

σ’v - Vertical effective stress (7.1).

σ’vi - Initial vertical effective stress prior to pile driving (7.2).

σ’vo - Vertical effective stress at the sample depth (5.1.1) (5.2) (7.2);

Vertical effective stress at midpoint of each layer (prior to stress 

increase) (7.3).

σ’1 - Effective stress after stress increase (7.3).

τ - Shear stress at failure (shear strength) (5.2); Shear strength of soil (7.1).

φ - Resistance factor, statistically based multiplier on nominal resistance 

(2.4); Angle of internal friction (2.4) (5.1) (5.2) (5.5) (7.1) (7.2) (7.3).

φ’ - Effective Stress Friction Angle (5.1) (5.5) (7.2.) (7.10).

φc - Resistance factor (pile structural resistance in compression) (8.3) (8.5) 

(8.6).

φb1 - Block Failure (7.1).

φda - Resistance factor (pile structural resistance during driving) (8.3) (8.4) 

(8.5) (8.6).

φdyn - Resistance factor (based on the construction control method) (2.10) (3.4) 

(7.1) (7.2) (7.6).

φf - Resistance factor (pile structural resistance in flexure) (8.3) (8.5) (8.6).

φm - Mobilized angle of internal friction (7.3).

φstat - Resistance factor (based on the static analysis method) (7.1) (7.2). 

φug - Resistance factor for group uplift (based on the uplift analysis method) 

(7.1) (7.2).

φup - Resistance factor (based on the uplift analysis method) (3.4) (7.1) (7.2).

φv - Resistance factor (pile structural resistance in shear) (8.3).

ψ - Ratio of undrained shear strength divided by effective overburden 

pressure,ssuu/σσ′vvvv (7.2).

ω - Angle of pile taper from vertical (7.2).

μ - Correction factor (5.2).
xxxiv

LIST OF ACRONYMS 

AASHTO - American Association of State Highway and Transportation Officials

ASTM - American Society for Testing and Materials

BL - Blast load

BOR - Beginning of Restrike 

BR - Vehicular braking force

CD - Consolidated Drained triaxial test

CE - Vehicular centrifugal force

CED - Closed End Diesel hammer

CEP - Closed End Pipe

CFA - Continuous Flight Auger

COR - Coefficient of Restitution

CPT - Cone Penetration Test

CPTu - Piezo Cone Penetration Test

CR - Force effects due to creep

CT - Vehicular collision force

CU - Consolidated Undrained triaxial test

CV - Vessel collision force

DA - Design Angular Distortion

DC - Dead load components and attachments

DD - Downdrag 

DF - Drag force

DLT - Dynamic Load Test

DMT - Dilatometer test

DW - Wearing surface and utilities

DWT - Deadweight tonnage 

EH - Horizontal earth pressure

EL - Locked-in stress 

EOD - End of Drive

EQ - Earthquake load

ER - SPT hammer efficiency as determined by energy measurements

ES - Earth surcharge 

EV - Vertical earth pressure

FHWA - Federal Highway Administration

FR - Friction load

I.D. - Inner diameter
xxxv

IC - Ice load

IM - Vehicular dynamic load allowance

KE - Kinetic Energy

LL - Liquid Limit; Vehicular live load

LS - Live Load Surcharge

LVDT - Linear Variable Differential Transformer

MUP - Modified Unloading Point Method

NHI - National Highway Institute

O.D. - Outer Diameter

OED - Open Ended Diesel hammer

OEP - Open Ended Pipe

PE - Potential Energy

PGA - Peak Ground Acceleration coefficient

PI - Plasticity Index

PL - Plastic Limit; Pedestrian live load

PS - Secondary forces from post-tensioning

RSA - Residual Stress Analysis

SA - Static Analysis

SE - Force effect due to settlement

SH - Force effects due to shrinkage

SLT - Static Load Test

SPT - Standard Penetration Test

SRD - Soil Resistance to Driving

SUP - Segmental Unloading Point Method 

TG - Force effect due to temperature gradient

TU - Force effect due to uniform temperature

UPM - Unloading Point Method

UU - Unconsolidated Undrained triaxial test

VST - Vane shear test

WA - Water load and steam pressure

WD - Downward traveling wave, Wave Down

WE - Wave Equation

WEAP - Wave Equation Analysis Program

WL - Wind on live load

WS - Wind load on structure

WU - Upward traveling wave, Wave Up

1 

CHAPTER 1

DRIVEN PILE FOUNDATION MANUAL

1.1 INTRODUCTION

In 1985 the Federal Highway Administration (FHWA) published the first edition of 

this manual. Subsequent editions were published in 1997 and 2006. Since the last 

published update, significant changes in pile design, construction, and performance 

requirements have occurred which make it necessary to once again update the 

manual. This 2015 update is primarily dictated by the need to revise and update 

manual content in accordance with the Load and Resistance Factor Design (LRFD) 

methodology which replaced the Allowable Strength Design (ASD) method for the 

design of transportation substructures in 2007. Other significant changes in practice 

addressed by this edition of the manual include:

• emphasis on a rational economic evaluation of the foundation design,

• use of higher strength pile materials and/or larger driven pile sections to 

support greater foundation loads,

• updates in computer programs for pile foundation analysis and design,

• use and quantification of soil setup in pile design and construction,

• improvements in pile installation equipment and equipment performance 

monitoring, 

• increased use of instrumented static load test programs, and

• improvements in QA/QC methods for nominal resistance and pile 

integrity verifications.

1.2 PURPOSE OF THE MANUAL

The purposes of the previous driven pile foundation manual editions remain largely 

unchanged. It is worthwhile to restate the purpose and objectives of the manual.

1. There exists a vast quantity of information on driven pile foundations which 

presently is not compiled in a form which is useful to most practicing 

engineers. There are proven rational design procedures, information on 

construction materials, equipment and techniques, and useful case histories. 
2

Unfortunately, much of this information is fragmented and scattered. 

Standard textbooks and other publications on the subject tend to be 

theoretically oriented and practicing design and construction engineers often 

find them lacking in practical aspects. One of the primary goals of this 

manual is to meet that need for the practicing engineer.

2. Many historical design methods lead to unnecessarily conservative designs 

because they were based solely on experience and tradition with little 

theoretical background. Well established rational design procedures and 

techniques are summarized herein that provide more economical driven pile 

foundation systems that can safely support the applied structural loads.

3. There are opportunities for substantial cost savings on driven pile foundation 

projects through the use of improved methods of design and construction 

technology. A minimum of fifteen percent of the substructure cost can be 

easily saved by utilizing such methods and, in most cases, the savings are 

more significant.

4. Since the adoption of LRFD methodology for all transportation projects, a 

comprehensive driven pile foundation manual has been needed. This manual 

is intended to fulfill that need as well as to establish minimum design 

standards and recommendations.

5. Design criteria for bridges and other structures are becoming more complex 

and sophisticated. Extreme design events such as scour, debris loading, 

vessel impact, and seismic events require that foundation performance be 

evaluated under lateral and uplift loading, group behavior, and substructure - 

superstructure interaction. In addition, deformation performance 

requirements (lateral and vertical deflections) are routinely included in project 

requirements. This new series of performance criteria frequently results in 

foundations which are more costly, more complex to design, and more difficult 

to construct.

6. The loading conditions noted above have can have a substantial impact on 

the structural design of the piles. In the past, driven piles were often 

designed structurally for axial loads only using an allowable stress approach. 

The allowable stresses had been established primarily to assure pile 

drivability. However, the requirement that piles be analyzed for combined 

horizontal and axial loads requires a change in the evaluation procedure. The 

pile top is subjected to both horizontal and axial loads and in a pile group the 
3 

pile resistance to lateral loads varies with each pile row. Of course this 

complicates the geotechnical analysis. It also complicates the pile structural 

analysis. A combined bending and axial load analysis of the structural 

behavior of the pile must be performed. Particularly for concrete piles this 

analysis must be based on an ultimate strength analysis and it is not always 

obvious which pile within a group is the critical one. Comprehensive software 

is now available to perform the necessary analyses and is discussed herein.

7. Alternative contracting methods (ACM), design-build (D-B) and CM/GC 

(construction manager/general contractor), are increasingly replacing the 

conventional design-bid-build (DBB) method as the preferred methods of 

project development and delivery. Among the changes which these ACM’s 

affect is an intentional, but yet aggressive challenge to existing design 

specifications.

8. Final pile design selection should involve a cost evaluation. In the past, such 

evaluations have been implied but they were not a routine part of the design 

process. Methods have been developed to perform cost evaluations of pile 

foundations that include the effects of soil setup. These concepts will be 

presented in this edition of the manual.

9. A larger selection of pile hammer types and sizes, improvements in hammer 

performance, advancements in equipment controllability and installation aids 

allow efficient pile installation in most subsurface conditions.

1.3 SCOPE OF MANUAL

The manual is limited to driven piles and consists of eighteen chapters and four 

appendices. The first half of the manual covers the design aspects of pile 

foundations including cost evaluations, geotechnical data collection and analysis, 

selection of pile type, as well as geotechnical and structural aspects of limit state 

design. The second half of the manual covers methods of nominal resistance 

verification as well as chapters on pile driving equipment, accessories and 

inspection procedures. Theoretical discussions have been included only where 

necessary. Specific recommendations are made where appropriate. Example 

problems are included to provide hands-on knowledge to manual users.

The manual is a standalone document that provides guidance for engineers on the 

design and construction of driven pile foundations. A separate training course will 
4

be used to transfer knowledge in this area and will continue with the original goal of 

updating transportation department practice. Also, new engineers continue to join 

transportation agencies and require expanding their knowledge in the practical 

aspects of pile design and installation.

1.4 HISTORY OF DRIVEN PILE FOUNDATIONS

The detailed history of driven pile foundations has been lost to time. It has been 

postulated that some of the earliest use of driven timber piles dates back to 800 BC 

where the piles were installed with mauls or drop hammers. Reinforced concrete 

piles debuted in Europe in 1897 and in Chicago in 1901. Structural steel piles 

including, pipe, I-beams, and H-piles followed not too long thereafter. Octagonal 

and square, precast, prestressed concrete piles as well as 36 inch and 54 inch 

diameter post-tensioned concrete cylinder piles developed in the 1950’s.

As noted above, the first pile hammers were simple drop hammers. The first 

modern pile driving hammer was a Scottish steam hammer patented by Naysmith in 

1839. In the U.S., steam pile driving hammers were reported between the mid- to 

late 1800’s. In the mid 1920’s, the diesel pile hammer was invented in Germany. 

Vibratory hammers were invented in the Soviet Union in the 1940’s and made their 

way into the U.S. market in the 1960’s. The first hydraulic pile hammers were 

developed in Scandinavia in the 1960’s. In the 1960’s, Bodine developed a resonant 

pile hammer but the hammer had limited use in the market due to mechanical 

reliability issues. The resonant pile driver re-emerged in the 2000’s. Many 

improvements in hammer features and operation have occurred over the years for 

all hammer types. Rausche (2000) summarized the development of pile driving 

equipment along with equipment capabilities and properties.

1.5 INFORMATION SOURCES

The information presented in this manual has been collected from several sources. 

The primary references are the 7th Edition of the AASHTO LRFD Bridge Design 

Specifications (2014) with 2015 Interim Revisions and the 3rd Edition of the 

AASHTO LRFD Bridge Construction Specifications with 2010, 2011, 2012, and 2014 

Interim Revisions. Additional sources of information include: "Evaluation of Soil and 

Rock Properties," GEC-5 by Sabatini et al. (2002), “Implementation of AASHTO 

LRFD Design Specifications for Driven Piles” by Abu-Hejleh et al. (2013), “Drilled 

Shafts: Construction Procedures and LRFD Design Methods” GEC-10 by Brown et 
5 

al. (2010), “LRFD Seismic Analysis and Design of Transportation Geotechnical 

Features and Structural Foundations” GEC-3 by Kavazanjian et al. (2011) as well as 

“Evaluating Scour at Bridges, Fifth Edition” HEC-18 by Arneson et al. (2012).

The information within has been condensed, modified and updated as needed. The 

sources also include state-of-the-art technical publications, manufacturers' literature, 

existing Federal Highway Administration (FHWA), National Highway Institute (NHI) 

and Transportation Research Board (TRB) publications, standard textbooks, and 

information provided by State and Federal transportation engineers. Reference lists 

are provided at the end of each chapter. 

Many of the documents used in the development or updating of this manual, as well 

as useful industry links are available at

https://www.fhwa.dot.gov/engineering/geotech/. 
6

REFERENCES

Abu-Hejleh, N., Kramer, W.M., Mohamed, K., Long, J.H., and Zaheer, M.A. (2013). 

Implementation of AASHTO LRFD Design Specifications for Driven Piles, 

FHWA-RC-13-001. U.S. Dept. of Transportation, Federal Highway 

Administration, 71 p.

American Association of State Highway and Transportation Officials (AASHTO). 

(2014). AASHTO LRFD Bridge Design Specifications, US Customary Units, 

Seventh Edition, with 2015 Interim Revisions. American Association of State 

Highway and Transportation Officials, Washington, D.C., 1960 p.

Arneson, L.A., Zevenbergen, L.W., Lagasse, P.F., and Clopper, P.E. (2012). 

Evaluating Scour at Bridges, Fifth Edition, FHWA-HIF-12-003, Hydraulic 

Engineering Circular (HEC) No. 18. U.S. Dept. of Transportation, Federal 

Highway Administration, 340 p.

Brown, D. A., Turner, J.P. and Castelli R.J. (2010). Drilled Shafts: Construction 

Procedures and LRFD Design Methods, FHWA-NHI-10-016, Geotechnical 

Engineering Circular (GEC) No. 10. U.S. Dept. of Transportation, Federal 

Highway Administration, 970 p.

Kavazanjian, E., Wan, J-N. J., Martin, G.R., Shamsabadi, A., Lam, I., Dickenson, 

S.E., and Hung, C.J. (2011). LRFD Seismic Analysis and Design of 

Transportation Geotechnical Features and Structural Foundations, FHWANHI-11-032, Geotechnical Engineering Circular (GEC) No. 3. U.S. Dept. of 

Transportation, Federal Highway Administration, 592 p.

Rausche, F. (2000). Keynote Lecture: Pile Driving Equipment: Capabilities and 

Properties, Proceedings of the Sixth International Conference on the 

Application of Stress Wave Theory to Piles, A.A. Balkema, pp. 75-90.

Sabatini, P.J., Bachus, R.C., Mayne, P.W., Schneider, J.A., and Zettler, T.E. (2002). 

Evaluation of Soil and Rock Properties, FHWA-IF-02-034, Geotechnical 

Engineering Circular (GEC) No. 5. U.S. Dept. of Transportation, Federal 

Highway Administration, 385 p.
7

CHAPTER 2

OVERVIEW OF PILE FOUNDATION DESIGN & CONSTRUCTION

2.1 INTRODUCTION

As stated by Professor R. B. Peck, “driving piles for a foundation is a crude and 

brutal process.” The interactions among the piles and the surrounding soil are 

complex. Insertion of piles alters the character of the soil and intense strains are set 

up locally near the piles. The non-homogeneity of soils, along with the effects of the 

pile group and pile shape, adds further difficulties to the understanding of soil-pile 

interaction. For piles driven to hard rock or into soft rock, the strength and structure 

of the rock mass including joints, bedding planes, and the degree of weathering 

complicate our understanding of rock-pile interaction.

Broad generalizations about pile behavior are unrealistic. An understanding of the 

significance of several factors involved is required to be successful in the design of 

pile foundations. Because of the inherent complexities of pile behavior, it is 

necessary to use practical semi-empirical methods of design, and to focus attention 

on significant factors rather than minor or peripheral details.

To arrive at the optimum driven pile foundation solution, the foundation engineer 

must have a thorough understanding of the subsurface conditions including soil/rock 

parameters and behavior, the applicable limit states, the factored loads and load 

combinations, project performance requirements, foundation costs, and the current 

foundation design and construction practices where the foundation is located. 

2.2 LIMIT STATES

Four limit states are identified in the AASHTO LRFD Bridge Design Specifications 

(AASHTO 2014): strength, service, extreme event, and fatigue. In addition, these 

limit states have several load combination cases such that up to thirteen limit states 

may require evaluation in a bridge design. There are five strength limit cases, two 

extreme event limit cases, four service limit cases, and two fatigue limit cases as 

described in Table 2-1. 
8 

Limit states that commonly govern driven pile foundation designs include: Strength I, 

Strength IV, Extreme Event I (earthquake), Extreme Event II (ice, vessel, blast, and 

vehicle collision), and Service I. All applicable limit states have equal importance in 

a driven pile foundation design. Service Limit States II, III, and IV and Fatigue Limit 

States I and II are relevant to the behavior of superstructure elements and are not

generally applicable to foundation design.

Table 2-1 Limit State, Load Case, and Load Combination (after AASHTO 2014)

Limit State and 

Load Case Load Combination

Strength I Basic load combination related to normal vehicular use of the bridge without 

wind. 

Strength II Load combination relating to use of the bridge by owner-specified special 

vehicles, evaluation permit vehicles, or both, without wind. 

Strength III Load combination for bridge exposed to wind velocity exceeding 55 mph without 

live loads.

Strength IV Load combination for very high dead load to live load force effect ratios. 

(Typical for bridge spans greater than 250 feet).

Strength V Load combination for normal vehicular use of the bridge with wind velocity of 55 

mph. 

Extreme Event I Load combinations including earthquake. 

Extreme Event II

Load combinations relating to ice load, collision by vessels and vehicles, and 

certain hydraulic events with a reduced live load other than that which is part of 

the vehicular collision load.

Service I Load combinations relating to normal operational use of the bridge with a 55 

mph wind and all loads taken at their nominal values. 

Service II Load combinations intended to control yielding of steel structures and slip of 

slip-critical connections due to vehicular live load. 

Service III

Load combinations relating to tension in prestressed concrete superstructures 

with the objective of crack control and to principal tension in the webs of 

segmental concrete girders. 

Service IV Load combinations intended to control tension in prestressed concrete columns 

with the objective of crack control. 

Fatigue I Fatigue and fracture load combination related to infinite load-induced fatigue life.

Fatigue II Fatigue and fracture load combination related to finite load-induced fatigue life.
9 

2.3 LOADS, LOAD COMBINATIONS, AND LOAD FACTORS

The total factored load, Q, associated with a given limit state is calculated based on 

the applicable force effect, load modifiers, and load factors.

 Eq. 2-1 

Where: 

ηi = load modifier based on ductility, redundancy, or operation 

classification.

γi = load factor, a statistically based multiplier applied to force effect.

Qi = force effect.

A specific load combination applies for each limit state case. A general description 

of the applicable load, load combinations and load factors associated with each limit 

state is presented in Table 2-2. The two letter codes in the second column in Table 

2-2 correspond to permanent loads and the remaining two letter code descriptions 

correspond to transient loads.

For example, the total factored load, Q, for Strength Limit State I is defined as 

follows:

Q = γp DC +γp DD + γp DW + γp EH + γp EV + γp ES + γp EL + γp PS + γp CR + 

γp SH + 1.75LL + 1.75IM + 1.75CE +1.75BR + 1.75PL +1.75LS + WA + 

FR + 1.20TU + γTG TG + γSE SE

The load factor for permanent loads, γp, has maximum and minimum values as 

prescribed in Table 2-3. For permanent force effects, the load factor that produces 

the more critical load combination is selected from Table 2-3. 
10

Table 2-2 Load Combinations and Load Factors (after AASHTO 2014)

Load Combination 

Limit State

DC

DD

DW

EH

EV

ES

EL

PS

CR

SH

LL

IM

CE

BR

PL

LS WA WS WL FR TU TG SE EQ* BL* IC* CT* CV*

Strength I γp 1.75 1.00 - - - - - - 1.00 0.50/1.20 γTG γSE - - - - - - - - - - - - - - - 

Strength II γp 1.35 1.00 - - - - - - 1.00 0.50/1.20 γTG γSE - - - - - - - - - - - - - - - 

Strength III γp - - - 1.00 1.40 - - - 1.00 0.50/1.20 γTG γSE - - - - - - - - - - - - - - - 

Strength IV γp - - - 1.00 - - - 1.00 1.00 0.50/1.20 - - 

- - - - - - - - - - - - - - - - - - - 

Strength V γp 1.35 1.00 0.40 - - - 1.00 0.50/1.20 γTG γSE - - - - - - - - - - - - - - - 

Extreme

Event I γp γEQ 1.00 - - - - - - 1.00 - - - - - 

- - - - 1.00 - - - - - - - - - - - - 

Extreme

Event II γp 0.50 1.00 - - - - - - 1.00 - - - - - 

- - - - - - - 1.00 1.00 1.00 1.00

Service I 1.00 1.00 1.00 0.30 1.00 1.00 1.00/1.20 γTG γSE - - - - - - - - - - - - - - - 

Service II 1.00 1.30 1.00 - - - - - - 1.00 1.00/1.20 - - 

- - - - - - - - - - - - - - - - - - - 

Service III 1.00 0.80 1.00 - - - - - - 1.00 1.00/1.20 γTG γSE - - - - - - - - - - - - - - - 

Service IV 1.00 - - - 1.00 0.70 - - - 1.00 1.00/1.20 - - 

- 1.00 - - - - - - - - - - - - - - - 

Fatigue I -**

LL, IM & CE - - - 1.50 - - - - - - - - - - - - - - - - - 

- - - - - - - - - - - - - - - - - - - 

Fatigue II -**

LL, IM & CE - - - 0.75 - - - - - - - - - - - - - - - - - 

- - - - - - - - - - - - - - - - - - - 

* - Use one of these at a time

** - Load factors only applied to LL, IM and CE

The two letter load descriptions correspond to permanent and transient loads as 

follows:

DC – dead load components and attachments

DD – downdrag

DW – wearing surface and utilities

EH – horizontal earth pressure

EV – vertical earth pressure

ES – earth surcharge

EL – locked-in stress

PS – secondary forces from post-tensioning

CR – force effects due to creep

SH – force effects due to shrinkage

LL – vehicular live load

IM – vehicular dynamic load allowance

CE – vehicular centrifugal force

BR –vehicular braking force 

PL – pedestrian live load

LS – live load surcharge

WA – water load and steam pressure 

WS – wind load on structure

WL – wind on live load

FR – friction load

TU – force effect due to uniform temp.

TG – force effect due to temp. gradient

SE – force effect due to settlement

EQ – earthquake load

BL – blast load

IC – ice load

CT – vehicular collision force

CV – vessel collision force
11

Table 2-3 Load Factors for Permanent Loads (after AASHTO 2014)

Type of Load

Max. 

Load 

Factor, 

γp

Min. 

Load 

Factor, 

γp

DC: Components and Attachments 1.25 0.90

DC: Strength IV only 1.50 0.90

DD: Downdrag (Tomlinson α-Method) 1.40 0.25

DD: Downdrag: (λ Method) 1.05 0.30

DW: Wearing Surface and Utilities 1.50 0.90

EH: Horizontal Earth Pressure (Active) 1.50 0.90

EH: Horizontal Earth Pressure (At-rest) 1.35 0.90

EH: Horizontal Earth Pressure (AEP for anchored walls) 1.35 N/A

EL: Locked-in Construction Stresses 1.00 1.00

EV: Vertical Earth Pressure (overall stability) 1.00 N/A

EV: Vertical Earth Pressure (retaining walls and abutments) 1.35 1.00

EV: Vertical Earth Pressure (rigid buried structures) 1.30 0.90

EV: Vertical Earth Pressure (rigid frames) 1.35 0.90

EV: Vertical Earth Pressure (flexible buried str. - metal box, plate, fiberglass) 1.50 0.90

EV: Vertical Earth Pressure (flexible buried str. - thermoplastic) 1.30 0.90

EV: Vertical Earth Pressure (flexible buried str. - all others) 1.95 0.90

ES: Earth Surcharge 1.50 0.75

Brown et al. (2010) summarized the basic limit state design process for a bridge or 

other structure. For each limit state, the structural engineer determines the 

foundation force effects using a preliminary structural model of the proposed 

structure. This structural model is developed and analyzed under the limit state load 

combinations described in Table 2-2. Factored loads are used in the analyses. 

Figure 2-1 illustrates the reactions at a column-cap joint computed by the structural 

analysis. These are the force effects transmitted to the deep foundation supported 

cap. For driven pile designs, these reactions are resolved into vertical, horizontal, 

and moment components, and are taken as the factored values of axial, lateral, and 

moment force effects, respectively at the top of the pile cap. Multiple iterations are 

typically performed to obtain agreement between deformations and forces calculated 

by the structural analysis and those based on geotechnical analysis at the 

structure/foundation (column-cap) interface. The resulting factored force effects 

must be less than the factored resistance. This is an oversimplified description of 
12

the process but it describes the general procedure by which factored foundation 

force effects are determined for each applicable limit state.

Figure 2-1 Structural analysis of bridge used to establish foundation force effects 

(modified from Brown et al. 2010).

In the structure model, the load factors in Tables 2-2 and 2-3 are varied over the 

specified ranges to determine the load combinations resulting in maximum force 

effects on the foundation. These maximum force effects are then used in the limit 

state checks. 

An example of the final structure loads determined for a 45 foot wide, three span 

structure having a total length of 350 feet and a main span length of 100 feet is 

presented in Figure 2-2. The structure will be subjected to both scour and seismic 

events. At Abutment 1, the highest factored axial loads are 2548 kips for the 

Strength I limit state and 1982 kips for the Service II limit state. The abutment also 

has a lateral shear load of 1980 kips transverse to the bridge for the Extreme Event I 

limit state. At Pier 2, the highest factored axial loads are 5542 kips for the Strength I 

limit state and 4307 kips for the Service II limit state. The pier is also subjected to a 

lateral shear load of 1267 kips longitudinal to the bridge, and a moment of 40,800 ftkips at the Extreme Event I limit state. Each limit state load combination is then 

checked independently based on all appropriate factored loads and moments and 

compared to the corresponding factored resistance and performance criteria 

(tolerable deforations) for each case.
13

Figure 2-2 Example of factored load calculation.
14

2.4 NOMINAL AND FACTORED RESISTANCE

The nominal resistance, Rn, is calculated and then multiplied by the applicable 

resistance factor to determine the factored resistance, Rr. Nominal geotechnical 

resistances and associated resistance factors are discussed in Chapter 7. Nominal 

structural resistances and applicable resistance factors are presented in Chapter 8. 

The factored resistance must be greater than or equal to the sum of all factored 

force effects in all applicable limit states. The basic LRFD methodology equation is 

given as:

∑ φγη =≤ RRQ rniii Eq. 2-2 

Where:

ηi = load modifier based on ductility, redundancy, or operation 

classification, applied to the force effect.

γi = load factor, statistically based multiplier applied to force effect.

Qi = force effect.

φ = resistance factor, statistically based multiplier applied to nominal 

resistance.

Rn = nominal resistance.

Rr = factored resistance.

The load modifiers for redundancy, ηr, described in AASHTO Article 1.3.4 were 

developed for superstructures. Paikowsky (2004) in NCHRP Report 507 defined 

redundancy in driven pile foundation designs based on the number of piles in a pile 

cap with redundant piles defined as 5 or more piles per pile cap and non-redundant 

piles defined as 4 piles or less per pile cap. For non-redundant driven pile 

foundation designs, AASHTO Article C10.5.5.2.3 recommends the resistance factor 

be reduced by 20%.
15

2.5 STRENGTH LIMIT STATES

The strength limit states ensure local and global strength and stability against 

statistically significant load combinations occurring during the structure design life. 

Strength limit state design of driven pile foundations includes an evaluation of the 

nominal geotechnical and structural resistances as well as the loss of lateral and 

vertical support in the design flood event (100 year) due to scour. For driven pile 

designs, strength limit state considerations include:

• axial compression resistance of single piles and pile groups, 

• uplift resistance of single piles and pile groups, 

• lateral resistance of single piles and pile groups,

• bearing stratum punching failure, 

• structural resistance in axial compression, combined axial and flexural 

loading, and shear, and 

• drivability including and driving stresses.

Geotechnical and structural strength limit state considerations are described in detail 

in Chapters 7 and 8, respectively of this manual.

2.6 SERVICE LIMIT STATES

The service limit states provide limits on stress, deformation, and cracking under 

regular service conditions. Service limit state considerations in driven pile 

foundation designs include:

• vertical deformation – settlement,

• horizontal movements,

• rotation,

• overall stability, and

• deformations due to scour at the design flood (100 year event).

Service limit state considerations are further discussed in AASHTO (2014) Articles 

10.5.2.1 through 10.5.2.4 and associated commentary. All applicable service limit 

state load combinations must be evaluated.
16

2.7 EXTREME EVENT LIMIT STATES

Extreme event limit states ensure structural survival of a bridge under unique major 

occurrences such as earthquakes, floods, and vehicle, vessel collisions, or blasts 

with return periods significantly greater than the bridge design life. Extreme event 

limit states for driven pile foundation design include: 

• the check flood (500 year event) for scour, 

• vessel collision, 

• vehicle collision, 

• blast loading,

• seismic loading, and 

• other site-specific situations determined by the design engineer.

A liquefaction assessment is required as part of the design for multi-span bridges if 

the site is classified as Seismic Zone 2, 3, or 4. Extreme event limit states are 

discussed further in Section 7.4 of this manual.

2.8 CONSTRUCTION OF PILE FOUNDATIONS

Construction of a successful driven pile foundation that meets the design objectives 

depends on relating the load requirements to the resistance requirements of the field 

installation and resistance determination method. The means for obtaining such a 

foundation must be explicitly incorporated into the plans and specifications as well 

as adhered to in the construction administration. 

A pile foundation must be installed to meet the limit state requirements for 

compressive, lateral and uplift resistance. This may dictate driving piles for a 

required nominal resistance or to a predetermined penetration depth established by 

the designer to satisfy strength, service and extreme event limit state performance 

requirements. It is equally important to avoid pile damage or foundation cost 

overruns by excessive driving. These objectives can all be satisfactorily achieved by 

use of wave equation analysis, dynamic monitoring of pile driving, and/or static load 

testing. Some agencies have calibrated and/or developed new dynamic formulas for 

nominal resistance verification to replace more unreliable dynamic formulas such as 

the Engineering News formula.

Knowledgeable construction supervision and inspection are the keys to proper 

installation of piles. State-of-the-art designs and detailed plans and specifications 
17

must be coupled with good construction supervision to achieve desired results. Post 

construction review of pile driving results versus predictions regarding pile 

penetration resistances, pile lengths, field problems, and static and/or dynamic load 

test nominal resistances is essential.

These reviews add to the experience of all engineers involved on the project and will 

enhance their skills. In addition, the implementation of LRFD in pile foundation 

design with rationally determined resistance factors makes it possible to use data 

from the post construction review to improve the resistance factors for future 

projects.

2.9 FOUNDATION SPECIALIST INVOLVEMENT

The input of an experienced foundation specialist is essential to produce a 

successful driven pile foundation. A foundation specialist has both a structural and 

geotechnical background in design and construction. The foundation specialist is 

the most knowledgeable person for selecting the pile type, estimating pile length, 

and choosing the most appropriate and cost effective method to determine the 

nominal resistance. In some agencies, the role of the foundation specialist may be 

divided amongst individuals and disciplines. Regardless of how the foundation 

specialist role is fulfilled, geotechnical and structural expertise in both design and 

construction knowledge is essential as design input.

The foundation specialist should be involved from the planning stage through the 

design and construction process. In some project phases (i.e. preliminary 

explorations, preliminary design, and final design), the foundation specialist will have 

significant involvement. In other project phases, such as construction, and post 

construction review, the foundation specialist’s involvement may be more of a 

technical support role. The foundation specialist’s involvement provides the needed 

continuity of design personnel in dealing with design related issues that develop 

throughout the construction stage. The importance of this continuity of knowledge, 

experience, and communication applies to all types of contracting methods.
18

2.10 THE DRIVEN PILE DESIGN AND CONSTRUCTION PROCESS

The driven pile design and construction process includes aspects that are unique in 

all of structural design. Because pile driving characteristics are related to the 

nominal geotechnical resistance for most soils, they can be used to improve the 

accuracy of the estimated nominal geotechnical resistance. In general, the various 

methods of determining nominal geotechnical resistance from dynamic data such as 

penetration resistance with wave equation analysis or with dynamic measurements 

are more accurate than the static analysis methods based on subsurface exploration 

information. It should be clearly understood that the static analysis based on the 

subsurface exploration information has the primary function of providing an estimate 

of the pile length for contractor bidding purposes. 

Pile drivability is a critical aspect of the design process and must be considered 

during the design phase. If the design is completed, a contractor is selected, and 

then the piles cannot be driven, large additional costs can often result. Therefore, it 

is absolutely necessary that driven pile foundation design and construction be linked

to a greater degree than in the design and construction of other foundation types.

During construction, the pile driving criterion is usually a blow count criterion that is 

established at the early stage of field installation, and individual pile penetration

depths may vary depending on the subsurface variability. In some instances, piles 

may need to meet a minimum penetration depth and a blow count criterion. 

Minimum pile penetration depths are sometimes established to satisfy uplift or lateral 

loads, or for serviceability consideration. In other cases, satisfying a required pile 

penetration depth established from static analysis may be the sole pile installation 

criterion.

The driven pile design-construction process is outlined in the flow chart of Figure 

2-3. The design and construction process will be discussed block by block using the 

numbers in the blocks as a reference and will serve to guide the designer through all 

of the tasks that must be completed. The block border depicts whether the structural 

engineer, geotechnical engineer, or construction engineer has the lead role for a 

given step in the design or construction process. This highlights the importance of 

interdisciplinary communication. The foundation specialist may perform some or a 

portion of the outlined structural and geotechnical steps based on their background 

and the organizational structure of the transportation agency.
19

Figure 2-3 Driven pile design and construction process.
20

Figure 2-3 Driven pile design and construction process (continued).
21

Figure 2-3 Driven pile design and construction process (continued).
22

Figure 2-3 Driven pile design and construction process (continued).
23

Figure 2-3 Driven pile design and construction process (continued).
24

Figure 2-3 Driven pile design and construction process (continued).

Figure 2-3 is representative of the key steps in a typical design-bid-build project. 

The contractor in a design-bid-build project would be retained between Blocks 27 

and 28, however the design and construction tasks for projects delivered using 

design-build or construction manager/general contractor contracts will vary. In a 

design-build contract, the contractor would be retained early on in the process once 

the project’s general requirements and preliminary supporting information has been 

established. Hence, the early steps of the flow chart in a design-build contract will 

vary. While the responsibility for the individual tasks will differ in alternative 

contracting methods, the identified key design or construction activities are still 

performed.
25

Block 1: Establish Global Project Performance Requirements.

The first step in the design process is to determine the structure requirements.

1. Is the project a new bridge, a replacement bridge, a bridge renovation, a 

bridge widening, a retaining wall, a noise wall, or sign or light standard?

2. Will the project be constructed in phases or all at one time?

3. What are the general structure layout and approach grades?

4. What are the surficial site characteristics?

5. What are the approximate foundation loads? What are the deformation or 

deflection limits (total settlement, differential settlement, and lateral 

deformations) for the service and extreme event limit states?

6. Will the structure be subjected to any extreme event limit states such as 

seismic, scour, debris loading, vessel or vehicle impact, etc.?

7. Are there any special considerations to be evaluated such as lateral 

squeeze?

8. Are there modifications in the design that may be desirable for the site under 

consideration such as changes in substructure locations or span length?

9. Are there site or surrounding environmental considerations that must be 

considered in the design (low headroom, utility conflicts, aggressive soil 

environments, environmentally impacted soil and/or groundwater, limitations 

on noise, vibrations, etc.)? These factors may influence the selection of 

suitable deep foundation types, the deep foundation installation equipment, 

as well as the need for installation aids such as predrilling for driven pile 

options.

10. Are there other factors influencing bridge span lengths including river 

navigation channel width, road or railroad crossings; or avoidance of previous 

foundations?
26

Block 2: Determine Structure (bridge) Geometry, Substructure Locations and 

Elevations.

The general bridge geometry, probable substructure locations, and the top of 

foundation elevations should be established at this time.

Block 3: Define General Site Geotechnical Conditions, Scour, and Seismicity.

A great deal can be learned about the foundation requirements with even a very 

general understanding of both the site geology and area geotechnical conditions. 

Frequently there is information available on foundations that have been constructed 

in the area. This information can be of assistance in avoiding problems. Both 

subsurface exploration information and foundation construction experience should 

be collected prior to beginning the foundation design. This step is sometimes 

overlooked in practice. 

If scour will be a design consideration, the hydraulic engineer should be consulted to 

determine probable scour depths that may impact the foundation selection.

Block 4: Perform Preliminary Structure Modeling. Determine Preliminary 

Substructure Loads and Tolerable Deformations.

Preliminary structural analysis and modeling of the proposed bridge or structure is 

performed at this time. The strength, service and extreme event limit states load 

demands and performance requirements at the foundation top have been 

established. For major bridge structures such as cable-stayed and balanced 

cantilever bridges, construction stage loads may govern the foundation design.

Many agencies also have total and differential settlement criteria for typical bridges 

which must be satisfied by the design. Lateral deformation limits for the proposed 

structure should also now be known and conveyed to the foundation specialist.

It is imperative that the foundation specialist obtain a completely defined and 

unambiguous set of foundation loads and deformation limits in order to proceed 

through the foundation design process. Accurate load information and deformation 

criteria are essential in the development and implementation of an adequate 

subsurface exploration program for the planned structure. It is important for the 

foundation specialist to address the level of uncertainty in their deformation 
27

predictions and distinguish between total final loads and incremental loading such 

that post-construction deformation can be distinguished from total deformations. Of 

course, post-construction deformations are those of most interest.

Block 5: Develop and Execute Subsurface Exploration and Laboratory Testing 

Program for Feasible Foundation Systems.

Based on the information obtained in Blocks 1-4, it is possible to make decisions 

regarding the necessary information that must be obtained for the technically 

feasible foundation systems at the site. The subsurface exploration program 

including the associated laboratory testing must meet the project needs for design 

and construction at a cost consistent with the size and complexity of the project. 

Depending upon the project size and complexity, it may be advantageous to perform 

the subsurface exploration program in phases, one for preliminary planning and 

general site evaluation and a second phase for final design.

The results of the subsurface exploration soil boring and in-situ testing program 

along with laboratory test results are used to prepare a subsurface profile; define soil 

and rock parameters including strength, compressibility, parameter variation, 

liquefaction susceptibility, and seismic earth pressure parameters; subsurface water 

conditions, as well as identify critical cross sections for design. The design profile 

for each substructure location will be developed from the information gathered in this 

block and used in later blocks. Site characterization and design parameter selection 

are covered in detail in Chapters 4 and 5. 

Block 6: Evaluate Information and Determine Candidate Foundation Systems.

The information collected in Blocks 1-5 must be evaluated and candidate foundation 

systems selected. The question to be answered is what candidate foundation 

systems are appropriate for consideration based on the site conditions. This 

question will be answered based primarily on the strength and compressibility of the 

geomaterials, the proposed loading conditions, the project deformation limits, the 

project schedule, and the foundation cost. Shallow foundations may be determined 

to be technically feasible and the most economical solution provided all project 

performance requirements can be addressed. Ground improvement techniques in 

conjunction with shallow foundations should also be evaluated. Shallow and deep 

foundation interaction with approach embankments and approach slabs must also 

be considered.
28

If the performance of a shallow foundation exceeds the established deformation 

limit, or if excessive scour is a concern, a deep foundation must be used. The 

design of shallow foundations and ground improvement techniques are not covered 

in this manual. Information on design consideration for shallow foundations can be 

found in Munfakh et al. (2001), Kimmerling (2002), Samtani et al. (2010), and in

Abu-Hejleh et al. (2014). The effects of settlement can be more comprehensively 

examined by using a construction point concept and estimating the differential 

settlement between supports as discussed by Modjeski and Masters, Inc, et al. 

(2015) in SHRP 2 Report S2-R19B-RW-1. Information on ground improvement 

techniques can be found in Elias et al. (2006) and at http://www.geotechtools.org. 

The selection of the appropriate foundation system including cost considerations is 

discussed in greater detail in Chapter 3, while Chapter 6 presents pile type selection 

along with advantages and disadvantages.

Block 7: Determine if a Deep Foundation is Required.

Once candidate foundation systems have been identified in Block 6, it has been 

determined that a deep foundation system is required. Therefore, driven piles and 

other deep foundation systems must be further evaluated. These other deep 

foundation systems are primarily drilled shafts, but also include micropiles, 

continuous flight auger (CFA) piles, and other drilled-in deep foundation systems. 

The questions that must be answered in deciding between driven piles and other 

deep foundation systems will center on both the technical feasibility and the relative 

costs of available systems. Foundation support cost can be conveniently calculated 

based on the cost per unit of load carried. The foundation cost analysis should 

address all temporary and permanent requirements (e.g., pile caps, effects of pile 

cap elevation, cofferdams, use of vertical and/or batter piles, load tests, construction 

control tests) for that specific foundation type. In addition, constructability and 

productivity (i.e., schedule) must be considered. This manual is concerned with 

driven piles; therefore other types of deep foundations will not be further discussed. 

Design guidance on drilled shafts can be found in Brown et al. (2010). For micropile 

design guidance refer to Sabatini et al. (2005), and for CFA piles refer to in Brown et 

al. (2007). Economic considerations in the foundation selection process are 

discussed in Chapter 3.
29

Block 8: Select 2 to 5 Candidate Driven Pile Types and Sections for Further 

Evaluation.

At this point on the flow chart, the primary concern is for the design of a driven pile 

foundation. The pile type must be selected consistent with the factored loads 

(compression, tension, and lateral) to be resisted per pile. Consider this problem. 

The general magnitude of the pier load is known from the information obtained in 

Blocks 1 and 4. However, a large number of combinations of pile types and nominal 

geotechnical resistances can satisfy the nominal resistance requirements for the 

factored loads. In the case of axial compression resistance, should twenty piles with 

a nominal resistance of 225 kips be used to support a 4,500 kip factored load, or 

would it be better to use ten piles with a nominal resistance of 450 kips, or five piles 

with a nominal resistance of 900 kips? This decision should consider the nominal 

structural resistance of the piles, the realistic geotechnical nominal resistances and 

deformations for the soil conditions at the site, the cost of the different pile types, 

and the capability of construction contractors to install the piles. Changes in the 

foundation selection from twenty 225 kip piles to five 900 kip piles will also affect 

structure and foundation response and therefore adjust the lateral and axial loads as 

well as pile cap requirements and costs. At this point in the design process, 2 to 5 

candidate pile types and/or sections that meet the general project requirements 

should be selected for further evaluation. Pile type and selection considerations are 

covered in greater detail in Chapter 6.

Approximate loads determined in Block 1 were refined in Block 4. At those stages of 

the design process, the other aspects of the total structural design were probably 

insufficiently advanced to establish the final factored loads. By the time that Block 7 

has been reached, the foundation loads and deformation limits should be closer to 

being finalized. 

If there are extreme event limit states applicable to the structure, the design must 

satisfy those load and deformation requirements. Vessel impact will be evaluated 

primarily by the structural engineer and the results of that analysis will give factored 

loads for that case. There may be stiffness considerations in dealing with vessel 

impact since the design requirement is basically that some vessel impact energy be 

absorbed.

Scour presents a different requirement. AASHTO (2014) Article 3.7.5 requires 

changes in foundation conditions resulting from the design flood (100 year event) be 

evaluated at the strength and service limit states. Foundation condition changes 

from the check flood (500 year event) must be considered and evaluated at the 
30

extreme event limit state. Scour is not a force effect. However, scour can change 

the substructure conditions and alter the consequences of force effects acting on the 

structure. It must be assured that after scour, the pile will still satisfy geotechnical 

and structural resistance demands.

In many locations, seismic design will be an important factor. Since the 1971 San 

Fernando Earthquake, significant emphasis has been placed on the design of 

highway bridges in seismic events. AASHTO (2014), Section 10, Appendix A10 

discusses seismic analysis and design requirements. Additional seismic design 

guidance is available in GEC-3, LRFD Seismic Analysis and Design of 

Transportation Geotechnical Features and Structural Foundations, by Kavazanjian

et al. (2011), as well as in the AASHTO (2011) Guide Specifications for LRFD 

Seismic Bridge Design, 2nd Edition, with 2012, 2014, and 2015 Interims.

Block 9: Calculate Nominal and Factored Structural Resistance.

The maximum nominal and factored structural resistances of all candidate pile types 

in axial compression, bending, and combined axial compression and flexure 

resistance are calculated at this time. Pile structural resistance is covered in 

Chapter 8.

The maximum factored resistance for a given pile type is the lesser of the factored 

structural resistance or the factored geotechnical resistance for that pile.

Block 10 Calculate the Nominal and Factored Geotechnical Resistance, as well as 

Perform a Preliminary Drivability Assessment of Candidate Pile Types.

The maximum nominal and factored geotechnical resistance in axial compression 

and uplift are calculated as a function of pile penetration depth for all candidate pile 

types. A static analysis method(s) appropriate for the pile type(s), the soil 

conditions, and the loading condition should be selected. These methods are 

presented in detail in Chapter 7. Factored geotechnical resistances are most often 

calculated based on the resistance determination method used in the field but can 

also be calculated solely on the basis of an appropriate static analysis method. 

Review of the nominal and factored geotechnical resistances versus depth for 

candidate pile sections and various resistance determination methods assists the 

designer in selecting the foundation type and associated resistance determination 

method.
31

Deposits within the subsurface strata which are unsuitable as load carrying 

geomaterial must be identified. The geotechnical resistance from these layers must 

be determined by static analysis. Unsuitable layers may include urban fills, organic 

deposits, soft and very soft cohesive soils, as well as potentially scourable or 

liquefiable materials.

Preliminary wave equation drivability assessments of candidate pile sections should 

be performed at this time. These analyses must consider the geotechnical 

resistance from suitable as well as unsuitable layers. A commonly available pile 

hammer having a ram weight of 1 to 2% of the required nominal resistance is a good 

initial trial hammer size for preliminary drivability assessments.

Block 11: Estimate the Preliminary Number of Piles, Preliminary Group Size, and 

Resolve Individual Pile Loads at All Limit States.

The preliminary number of piles at a substructure location can be estimated by 

dividing the largest factored axial load by an individual pile’s factored geotechnical 

resistance. Axial compression and axial tension requirements should both be 

considered. In some cases, lateral load requirements may control the design. 

Therefore, similar past experience may also be used to estimate the preliminary 

number of piles, trial group configurations, and pile spacing. Using the factored 

loads with the trial group configurations, determine the limit state reactions on the 

substructure unit and the resulting maximum factored load per pile at the strength, 

service, and extreme limit states for each candidate pile type.

Block 12: Estimate Pile Penetration Depth for Axial Compression Loads and 

Check Group Efficiency in Axial Compression.

For each candidate pile section and each group configuration, determine the 

estimated pile penetration depth where the factored geotechnical resistance 

exceeds the factored load. Note that the factored geotechnical resistance will vary 

based on the resistance verification method. Block 12 establishes an estimated pile 

penetration depth for compression loading for each of the candidate pile types. Note 

this estimated depth is a function of the resistance determination method. Check 

that group axial compression resistance meets design requirements. If the group 

resistance group does not satisfy requirements, modify the design.
32

Block 13: Establish Minimum Pile Penetration Depth for Axial Tension Loads.

Determine the maximum factored axial tension load to be resisted in any limit state. 

From the nominal axial tension resistance versus depth results calculated in Block 

10, determine the minimum pile penetration depth necessary to obtain a factored 

geotechnical resistance greater than the maximum factored axial tension load. 

Block 13 establishes a minimum pile penetration depth for tension loading for each 

of the candidate pile types. Note that this minimum depth is a function of the 

resistance determination method. Check that group axial tension resistance meets 

design requirements. If axial tension requirements cannot be achieved, modify the 

design. Return to Block 8 and select a new pile type or new pile section, or return to 

Block 11 and evaluate new group configurations.

Block 14: Establish Minimum Pile Penetration Depth for Lateral Loads.

Determine p-y models and required geomaterial properties for each layer in the 

subsurface profile. The p-y model and parameters chosen depend on the soil or 

rock response being modelled in the service or extreme event limit case. Results of 

this analysis will be compared to the tolerable lateral deformation requirements. 

Check that the deformation and lateral resistance of the trial group configuration 

meets design requirements. Determine the minimum pile penetration depth 

necessary to resist the maximum applied lateral loads within the permissible 

deformation limit. This depth establishes the minimum pile penetration depth for 

lateral loading for the candidate pile type and group. This topic is covered in greater 

detail in Section 7.3.4.2. If lateral deformation requirements cannot be achieved, 

modify the design. Return to Block 8 and select a new pile type or pile section, or 

return to Block 11 and evaluate new group configurations.

Block 15: Establish Pile Penetration Depths that Satisfy Tolerable Deformations 

Based on Group Settlement Computations.

Preliminary group configurations should now be evaluated for settlement. One of 

the primary objectives at this stage is to determine if a minimum pile penetration 

depth is required and, if so, to determine the required minimum pile toe elevation. In 

some subsurface profiles, piles could attain the requisite nominal resistance near the 

bottom of a dense layer overlying a compressible layer. In order to satisfy tolerable 

deformation limits, it may be necessary to drive through, or otherwise penetrate, the 

higher, suitable layer to preclude large future settlements. Hence, in some 
33

stratigraphy cases, it is necessary to calculate group settlements over a range of pile 

termination depths to determine the minimum (highest pile toe elevation) and, in 

unique stratigraphy cases, the maximum (lowest pile toe elevation) acceptable for 

tolerable group settlements. Any penetration depths that will result in intolerable 

deformations should be identified. Where appropriate, Block 15 establishes a 

minimum pile penetration depth, maximum pile penetration depth, or both. If 

tolerable deformation requirements cannot be achieved, modify the design. Return 

to Block 8 and select a new pile type, or return to Block 11 and evaluate new group 

configurations.

Block 16: Check Pile Drivability.

Identify the maximum pile penetration depth required by each block in Block 12 

through Block 15. Perform a wave equation drivability analysis for each candidate 

section to the maximum pile penetration depth determined in any of these blocks. 

Candidate pile types that cannot be driven to the required nominal resistance and/or 

the minimum pile penetration depth without exceeding material stress limits and 

within a reasonable blow count of 30 to 120 blows per foot with appropriately sized 

driving systems are eliminated at this time. It should be noted that 120 blows per 

foot or 10 blows per inch is considered refusal driving conditions by many hammer 

manufacturers. Therefore, depending on the expected driving conditions (e.g., an 

extended duration of hard driving compared to a quick transition onto hard rock), it 

may more reasonable to assess candidate pile drivability based on an upper limit of 

96 blows per foot or 8 blows per inch. In some cases pile installation aids such as 

jetting or predrilling may be evaluated subject to other design limitations.

The drivability analyses should also consider what influence the group configuration 

(pile spacing) and construction procedures and constraints (i.e., predrilling,

cofferdams, etc.) may have on pile installation conditions. If the selected pile type 

does not meet drivability requirements, a different pile section, pile type, or group 

configuration is required. Return to Block 8 and select a new pile type, or return to 

Block 11 and evaluate new group configurations.

Block 17: Determine Location of Neutral Plane and Magnitude of Drag Force.

The neutral plane location for drag force evaluation is performed in this step. Note 

that the neutral plane location may have already been determined in Block 15 

depending upon the settlement analysis approach selected by the designer. The 
34

neutral plane is defined as the location where the sum of the permanent structure 

load and drag force is balanced by the sum of the shaft and toe resistances

occurring below the neutral plane. The maximum drag force, caused by differential 

movement between the pile and the geomaterials, occurs at the neutral plane. The 

drag force magnitude is related to the pile properties, loading conditions, soil stress 

state, and deformation. The sum of the permanent structure load plus the maximum 

drag force should be less than the pile’s factored structural resistance. If it is not, a 

different pile section or pile type is required, and the design process returns to Block 

8, or a new group configuration is required and the design process returns to Block 

11. Determination of the neutral plane location and drag force is discussed further in 

Section 7.2.10.

It should be noted that the AASHTO (2014) design specifications do not specifically 

address drag force considerations relative to the pile structural resistance. For 

example, the factored structural resistance of an H-pile is the nominal structural 

resistance multiplied by the resistance factor for axial compression, φc. This 

resistance factor is 0.70 for combined axial and flexural resistance of undamaged 

piles, 0.60 for the axial resistance of piles in compression under good driving 

conditions, and 0.50 for the axial resistance of piles in compression subject to 

damage due to severe driving conditions. While not equivocally stated, it follows 

that if the neutral plane is located below the point of fixity and above the depth where 

H-piles are subject to potential damage during driving, then the sum of the 

permanent load plus drag force at the neutral plane is limited to 0.70 of the nominal 

structural resistance.

Block 18: Does the Computed Total Settlement, Differential Settlement, and 

Angular Distortion Satisfy Design Requirements?

In this block, the total settlement should be computed and compared to project 

deformation limits. At the same time, the computed differential settlement between 

the current substructure location and adjacent substructure units should be 

compared to differential settlement and angular distortion limits. The construction 

point concept described in Section 7.3.3 and the recommended procedure for 

vertical deformation analysis in Section 7.3.4 should be followed.

If the total settlement, differential settlement, or angular distortion limits are 

exceeded, the design process returns to an earlier block. If a new pile type or 

section needs to be considered, the design process returns to Block 8. If existing 
35

candidate pile types or sections are considered for the redesign of one or more of 

the substructure units, then the design process returns to Block 11.

Block 19: Evaluate Economics of Candidate Pile Types and/or Sections.

The next step is to determine the pile support cost versus pile penetration depth for 

each candidate pile type and/or section. The support cost, which is the cost per ton 

supported, is the pile cost at a given penetration depth divided by the nominal 

resistance at that penetration depth. The pile cost can be calculated from the unit 

cost per foot multiplied by the pile length between the pile cutoff elevation and the 

pile penetration depth. These plots should be evaluated to identify possible pile 

termination depths to obtain the lowest pile support cost. In some instances, uplift 

load rather than axial compression load may control the design in which case the 

same concept could be used for nominal tension resistance. The pile cost should 

include all costs associated with a given candidate section including material, 

transportation, and installation. In addition, the cost of the pile cap, cofferdams, 

dewatering or other related construction items should be included. This economic 

assessment should be modified appropriately in cases where lateral load demand 

and deformation control the design. The cost of the resistance determination 

method should be considered when evaluating the economics of candidate piles. 

Economic evaluation is presented in detail in Chapter 3.

Block 20: Is the Preliminary Design Complete at All Substructure Locations?

The preliminary design process in Blocks 9 through 19 is completed for each 

substructure location until the preliminary design is complete at all foundation units.

Block 21: Refine Structural Modeling and Loads.

At this point, the structure and foundation response to the loading cases can be 

further refined based on the structural and geotechnical analyses completed in Block 

9 through Block 17. The structural model is now reanalyzed using the results from 

these preliminary substructure analyses to better determine the structure loads at 

the foundation top, and lateral earth pressure loads on the abutments.
36

Block 22: Did the Loads Significantly Change Requiring Reevaluation of the 

Foundation Design?

Based on the refined structural modeling, review the preliminary design based on 

the refined loads. If necessitated by the refined modelling, return to Block 8 and 

modify the design.

Block 23: Re-evaluate Foundation Stiffness and Loads. 

For dynamic elastic analyses in seismic or other dynamic extreme events, the 

structural engineer now re-evaluates foundation stiffness. The structural model will 

be rerun to get new foundation loads.

Block 24: Did the Loads Significantly Change Requiring Reevaluation of the 

Foundation Design?

Review the preliminary design based on the refined loads from the dynamic elastic 

analyses. If loads from lateral pile analysis do not match the new foundation top 

loads within 5%, re-iterate the process until convergence is achieved. See Section 

7.4.2.1 and GEC 3, (Kavazanjian et al. 2011). If necessary, return to Block 8 and 

modify the design.

Block 25: Does the Design Meet All Limit State Requirements? 

The foundation specialist checks the estimated and/or minimum pile toe elevation for 

nominal geotechnical resistance in axial compression and tension, deformation 

under lateral load, serviceability, and drivability. If any limit state checks are not 

satisfied, the design process returns to Block 8.

Block 26: Design Pile Caps and Abutments. 

The final design of the pile cap (and abutment walls) is completed according to 

AASHTO specifications for concrete structures. The size and thickness of the pile 

cap in evaluating trial group configurations should have been previously evaluated

and the cost of the resulting pile cap estimated. Pile cap cost is a key component in 

selecting the most cost effective foundation and should not be overlooked in the 
37

design process. A procedure for preliminary sizing of pile caps is provided in 

Chapter 8.

Block 27: Prepare Plans and Specifications, Set Nominal Resistance Field 

Verification Procedure.

When the design has been finalized, the structural engineer prepares the project 

plans and the geotechnical engineer the project specifications. It is important that all 

of the quality assurance procedures are clearly defined for the bidders to avoid 

claims after construction is underway. The factored pile load should be shown on 

the plans along with nominal resistance, the resistance determination method and 

the method’s resistance factor, φdyn. If changes in the resistance determination 

method are later appropriate, the required nominal resistance with the new method 

can readily be evaluated. Any minimum or maximum pile penetration depth 

requirements should also be clearly identified. This information should also be 

readily apparent in the project specifications so that the contractor has no question 

regarding the installation requirements. Quality assurance procedures should be in 

place that address commonly occurring pile installation issues such as obstructions, 

drivability and construction records. Contract documents and specifications are 

discussed in Chapter 14, and the foundation report is covered in Chapter 5.

Block 28: Perform Evaluation of Contractor’s Equipment Submission.

At this point the engineering effort shifts to the field. The contractor will be required 

to submit a description of the pile driving equipment that is intended to be used on 

the project for the engineer’s evaluation. On most projects, a wave equation 

analysis is performed to estimate the blow count that must be achieved in the field to 

meet the required nominal driving resistance as well as to check if the pile can be 

driven to the required pile penetration depth, if specified. Driving stresses are also 

determined and evaluated. If all conditions are satisfactory, the equipment is 

approved for driving. Some design specifications make this information advisory to 

the contractor rather than mandatory. Chapters 12 and 14 provide additional 

information in this area.
38

Block 29: Set Preliminary Driving Criteria, Drive Test Pile(s), Perform Resistance 

Evaluation Tests, and Evaluate the Nominal Resistance Achieved.

Depending on the specified resistance determination method, the blow count 

needed to achieve the required nominal resistance can be established and used as 

the preliminary driving criteria. With some methods, the preliminary criteria can be 

determined before driving a pile and with other methods a pile must be driven, 

tested, and results evaluated. Preliminary driving criteria are generally established 

using wave equation analysis or a dynamic formula since these can be performed 

before driving a pile. Additional details on wave equation analysis and dynamic 

formulas are presented in Chapters 12 and 13, respectively. 

The test pile(s), if required, are driven to the preliminary criteria. Driving 

requirements may be defined by blow count, minimum penetration depth, dynamic 

monitoring results, or a combination of these conditions. The nominal resistance 

can be evaluated using the observed blow count and hammer performance (e.g., 

stroke, energy readout) in a wave equation analysis or dynamic formula. 

Alternatively, the nominal resistance achieved by the test pile can be evaluated by

more reliable methods such as a static load test, dynamic load test, or rapid load 

test. Details on static load testing are presented in Chapter 9, dynamic load testing

in Chapter 10, rapid load testing in Chapter 11. 

Each field resistance determination method has its own resistance factor, φdyn. 

Therefore, it is important to determine the resistance determination method during 

the design process. The selected resistance determination method determines the 

required nominal resistance. Post-design modification to the resistance 

determination method changes the resistance factor and the required nominal 

resistance. Changing to a more reliable resistance determination method reduces 

the required nominal resistance, which generally shortens the piles. Conversely, 

changing to a less reliable resistance determination method increases the required 

nominal resistance. This may result in the original design no longer meeting design 

or constructability requirements.

On smaller projects, a dynamic formula might be used to evaluate the nominal 

resistance. The modified Gates Formula is frequently used for this purpose but 

some transportation agencies have also developed or adopted other dynamic 

formulas. If a dynamic formula is used, then drivability and hammer selection will be 

based on the blow count given by the formula only, since stresses and drivability to a 

penetration depth are not determined. Dynamic formula usage is covered in 

Chapter 13.
39

Block 30: Adjust Driving Criteria or Design. 

At this stage the final installation criteria can be set or, if test results from Block 29 

indicate the nominal resistance is inadequate, the driving criteria may have to be 

changed. Occasionally, it may be necessary to make changes in the design as far 

back as Block 8. It may also be possible to change to a more reliable resistance 

determination method with a higher resistance factor and thereby reduce the 

required nominal resistance to a nominal resistance that can be achieved without a 

major design change.

In some cases, it is desirable to perform preliminary field testing before final design. 

When the project is very large and the subsurface conditions are difficult, it may be 

possible to achieve substantial cost savings by having results from a design stage 

test pile program, including actual driving records at the site. Such results can be 

used to optimize the design, and reduce contractor contingencies when included as 

part of the bid package.

Block 31: Drive Production Piles with Construction Monitoring.

After the driving criteria is established, the production pile driving begins. Quality 

control and quality assurance procedures have been established and are applied. 

Construction inspection items are discussed in greater detail in Chapter 18. 

Problems may arise and must be handled as they occur in a timely fashion.

Block 32: Post-Construction Evaluation and Refinement of Design. 

After completion of the foundation construction, the design should be reviewed and 

evaluated for its effectiveness in satisfying the design requirements and also its cost 

effectiveness. This review benefits future similar projects constructed in similar 

conditions.

2.11 COMMUNICATION

Good communication between all organizations and disciplines involved in the 

design and construction of a pile foundation is essential to reach a successful 

completion of the project. In the design stage, communication and interaction is 

needed between the structural, geotechnical, geologic, hydraulic, and construction 
40

disciplines, as well as with consultants, drill crews and laboratory personnel. In the 

construction stage, structural, geotechnical and construction disciplines need to 

communicate for a timely resolution of construction issues as they arise. Tables 2-4

and 2-5 highlight some of the key issues to be communicated in the design and 

construction stages.

Table 2-4 Design Stage Communication

Subject Structural Geotechnical Hydraulic Construction Field 

Crews Laboratory

Preliminary 

Structure Loads and 

Performance 

Criteria. 

X X X 

Determination of 

Scour Potential. X X X 

Determination of 

Applicable Extreme 

Events. 

X X X 

Review of Past 

Construction 

Problems in Project 

Area. 

X X X X 

Implement 

Subsurface 

Exploration and 

Testing Programs. 

X X X X X X 

Determination of 

Pile Type, Length 

and Nominal 

Resistance. 

X X X 

Select Construction 

Control and Quality 

Assurance Methods.

X X X 

Effects of Approach 

Fills on Design. X X 

Prepare Plans and 

Specifications. X X X X 
41

Table 2-5 Construction Stage Communication

Subject Structural Geotechnical Construction

Perform Required Methods of Construction Control 

and Quality Assurance. X X X 

Perform Wave Equation Analysis of Contractors 

Driving System to Establish Driving Criteria. X X X 

Perform Static Load Test(s) and/or Dynamic 

Monitoring and Adjust Driving Criteria. X X X 

Resolve Pile Installation Problems / Construction 

Issues.

X X X 
42

REFERENCES

Abu-Hejleh, N., Kramer, W.M., Mohamed, K., Long, J.H., and Zaheer, M.A. (2013). 

Implementation of AASHTO LRFD Design Specifications for Driven Piles, 

FHWA-RC-13-001. U.S. Dept. of Transportation, Federal Highway 

Administration, 71 p.

American Association of State Highway and Transportation Officials (AASHTO). 

(2011). Guide Specifications for LRFD Seismic Bridge Design, 2nd Edition, 

with 2012, 2014, and 2015 Interim Revisions. American Association of State 

Highway and Transportation Officials, Washington, D.C., 331 p.

American Association of State Highway and Transportation Officials (AASHTO). 

(2014). AASHTO LRFD Bridge Design Specifications, US Customary Units, 

Seventh Edition, with 2015 Interim Revisions. American Association of State 

Highway and Transportation Officials, Washington, D.C., 1960 p.

Brown, D.A., Dapp, S.D., Thompson, W.R., and Lazarte, C.A. (2007). Design and 

Construction of Continuous Flight Auger (CFA) Piles. FHWA-HIF-07-03, 

Geotechnical Engineering Circular (GEC) No.08. U.S. Dept. of 

Transportation, Federal Highway Administration, 289 p.

Brown, D. A., Turner, J.P. and Castelli R.J. (2010). Drilled Shafts: Construction 

Procedures and LRFD Design Methods, FHWA-NHI-10-016, Geotechnical 

Engineering Circular (GEC) No. 10. U.S. Dept. of Transportation, Federal 

Highway Administration, 970 p.

Elias, V., Welsh, J.P., Warren, J., Lukas, R.G., Collin J.G., and Berg, R.R. (2006). 

Ground Improvement Methods Volumes I and II, FHWA-NHI-06-019 and 

FHWA NHI-06-020. National Highway Institute, Federal Highway 

Administration, U.S. Department of Transportation, Washington D.C.

Kavazanjian, E., Wan, J-N. J., Martin, G.R., Shamsabadi, A., Lam, I., Dickenson, 

S.E., and Hung, C.J. (2011). LRFD Seismic Analysis and Design of 

Transportation Geotechnical Features and Structural Foundations, FHWANHI-11-032, Geotechnical Engineering Circular (GEC) No. 3. U.S. Dept. of 

Transportation, Federal Highway Administration, Washington, D.C., 592 p.
43

Kimmerling, R.E. (2002). Shallow Foundations, FHWA-IF-02-054, Geotechnical 

Engineering Circular (GEC) No. 6. U.S. Dept. of Transportation, Federal 

Highway Administration, Washington, D.C., 310 p.

Modjeski and Masters, Inc. (2015). Bridges for Service Life Beyond 100 Years: 

Service Limit State Design. SHRP2 Report S2-R19B-RW-1. Transportation 

Research Board, Washington D.C., 268 p. 

Munfakh, G., Arman, A., Collin, J.G., Hung, J.C.-J., and Brouillette, R.P. (2001). 

Shallow Foundations Reference Manual, FHWA-NHI-01-023. National 

Highway Institute, Federal Highway Administration, Washington, D.C., 222 p.

Paikowsky, S.G. (2004), with contributions from Birgisson, B., McVay, M., Nguyen, 

T., Kuo, C., Baecher, G., Ayyub, B., Stenersen, K., O.Malley, K., 

Chernauskas, L., and O’Neill, M., Load and Resistance Factor Design (LRFD) 

for Deep Foundations, NCHRP Report 507. Transportation Research Board, 

Washington, D.C., 76 p.

Sabatini, P.J., Tanyu, B., Armour., P., Groneck, P., and Keeley, J. (2005). Micropile 

Design and Construction, FHWA-NHI-05-039. National Highway Institute, 

U.S. Dept. of Transportation, Federal Highway Administration, Washington, 

D.C., 436 p.

Samtani, N.C., Nowatzki, E.A., and Mertz, D.R. (2010). Selection of Spread 

Footings on Soils to Support Highway Bridge Structures, FHWA-RC TD-10-

001. U.S. Department of Transportation, Federal Highway Administration, 

Washington, D.C., 98 p.
44
45

CHAPTER 3

CONSIDERATIONS IN FOUNDATION SELECTION

A foundation is the interfacing element between the superstructure, substructure,

and the underlying geomaterial (i.e., soil or rock). The loads transmitted by the 

foundation to the underlying geomaterial must not cause shear failure or damaging 

deformations of the superstructure. It is essential to systematically consider various 

foundation types and to select the most appropriate alternative based on the 

superstructure requirements, the subsurface conditions, and foundation cost. 

Foundation types include shallow foundations consisting of spread footing or mat 

foundations with or without ground improvement; or deep foundations consisting of 

driven piles, micropiles, drilled shafts, or continuous flight auger (CFA) piles.

Subsequent chapters of this manual provide guidance on driven pile foundation 

design and construction. FHWA guidance for other foundation solutions is shown in 

Table 3-1 below.

Table 3-1 Foundation Information Sources Provided by the Federal Highway 

Administration

Foundation Solution FHWA Reference Author(s)

Spread Footings GEC-6, FHWA-SA-02-054 Kimmerling (2002) 

Ground Improvement FHWA-NHI-06-019/020 Elias et al. (2006)

Micropiles FHWA-NHI-05-039 Sabatini et al. (2005)

Drilled Shafts GEC-10, FHWA NHI-10-016 Brown et al. (2010) 

CFA Piles GEC-8, FHWA-HIF-07-03 Brown et al. (2007) 

Complete references for the above design manuals are provided at the end of this 

chapter. Information on the availability of these documents is provided at 

https://www.fhwa.dot.gov/bridge/geopub.htm 
46

3.1 FOUNDATION DESIGN APPROACH

The following design approach is recommended to determine the most appropriate

foundation alternative.

1. Determine the foundation loads to be supported, structure layout, limits on 

total and differential settlements, lateral deformations, lateral loads, scour, 

seismic and other extreme event loading conditions, and special requirements 

such as construction phasing and time constraints on construction. A 

complete knowledge of these issues is of paramount importance.

2. Evaluate the subsurface exploration and laboratory testing data. Ideally, the 

subsurface exploration and laboratory testing programs were performed with 

knowledge of the foundation loads and the needed geomaterial resistances. 

If the subsurface information is insufficient, perform a second subsurface 

exploration program, laboratory testing program, or in-situ testing program. 

3. Prepare a final subsurface profile and critical cross sections. Determine 

subsurface layers suitable or unsuitable for spread footings, pile foundations, 

or drilled shaft load transfer. Also consider if ground improvement techniques 

could modify unsuitable layers into suitable bearing layers.

4. Determine the most feasible foundation alternatives. Both shallow 

foundations and deep foundations should be considered. Deep foundation 

alternatives include driven piles, drilled shafts, micropiles, and CFA piles. 

Proprietary deep foundations systems should not be excluded as they may be 

the most economical alternative in a given condition. Consideration should 

be given to the following foundation options.

Shallow Foundations:

a. Spread footings (without ground improvement). 

b. Mat foundations.
47

Deep Foundations:

a. Driven pile foundations.

b. Candidate pile types. 

c. Viable pile sections. 

d. Drilled shafts.

e. Micropile.

f. Continuous Flight Auger (CFA) piles.

5. Prepare cost estimates for technically feasible alternative foundation designs 

including all associated substructure and construction control method costs. 

The cost estimates should consider the concept of foundation support cost 

introduced in Section 3.4.1. The foundation support cost should include all 

associated temporary and permanent substructure costs required for 

foundation construction (e.g. sheeting or cofferdam requirements, concrete 

tremie seal, pile cap requirement and size), the effect of environmental or 

construction limitations (noise, vibration, low overhead, access restrictions), 

as well as any required mitigation procedures (noise shrouds, predrilling, 

bubble nets, etc.).

6. Select the most appropriate foundation alternative. Generally the most 

economical alternative (lowest foundation support cost) should be selected 

and recommended. The ability of the local construction force as well as the 

availability of materials and equipment should also be considered.

For major projects, if the estimated costs of technically feasible foundation 

alternatives (during the design stage) are within 15 percent of each other, then 

alternate foundation designs should be considered for inclusion in the contract 

documents. The most economical foundation design will then be determined by 

construction demand and material pricing rather than subtleties in the design 

estimate. 

3.2 FOUNDATION ALTERNATIVES

To determine the most preferred foundation alternatives, both shallow foundations 

and deep foundations should be considered. Table 3-2 summarizes shallow and 

deep foundation types and uses, as well as applicable and non-applicable 

subsurface conditions.
48

Table 3-2 Foundation Types and Typical Uses (Modified from Bowles 1977)

Foundation Type Use Applicable Soil Conditions

Non-suitable 

or Difficult Soil Conditions

Spread footing, 

wall footings.

Individual columns, walls, 

bridge piers.

Any conditions where bearing 

capacity is adequate for applied 

load. May use on single stratum; 

firm layer over soft layer, or 

weaker layer over firm layer. 

Check immediate, differential and 

consolidation settlements.

Any conditions where 

foundations are supported on 

soils subject to excessive scour 

or liquefaction. 

Mat foundation. Same as spread and wall 

footings. Very heavy 

column loads. Usually 

reduces differential 

settlements and total 

settlements.

Generally soil bearing value is 

less than for spread footings. 

Over one-half area of structure 

covered by individual footings. 

Check settlements.

Same as footings.

Driven pile 

foundations. 

In groups to transfer 

heavy column and bridge 

loads to suitable soil and 

rock layers. Also to resist 

uplift and/or lateral loads.

Poor surface and near surface 

soils. Geomaterials suitable for

load support 15 to 300 feet below 

ground surface. Check 

settlement and lateral 

deformation of pile groups.

Shallow depth to hard stratum. 

Sites where pile driving 

vibrations or heave would

adversely impact adjacent 

facilities. Boulder fields.

Drilled shafts. In groups to transfer

heavy column loads. 

Monoshafts and small 

groups sometimes used. 

Cap sometimes 

eliminated by using drilled 

shafts as column

extensions.

Poor surface and near surface 

soils. Geomaterial suitable for 

load support located 25 to 300 

feet below ground surface.

Caving formations difficult to 

stabilize. Artesian conditions. 

Boulder fields. Contaminated 

soil. Areas with concrete 

delivery or concrete placement 

logistic problems.

Micropiles. Often used for seismic 

retrofitting, underpinning, 

very difficult drilling 

through overburden 

materials, in low head 

room situations, and for 

projects with noise or 

vibration restrictions. 

Any soil, rock, or fill conditions 

including areas with rubble fill, 

boulders, and karstic conditions. 

High slenderness ratio may 

present buckling problems from 

loss of lateral support in 

liquefaction susceptible soils. 

Low lateral resistance. 

Offshore applications. 

CFA Piles. In groups to transfer 

heavy loads to suitable 

geomaterials. Projects 

with noise and vibration 

restrictions. 

Medium to very stiff clays, 

cemented sands or weak 

limestone, residual soils, medium 

dense to dense sands, rock 

overlain by stiff or cemented 

deposits.

Very soft soils, loose saturated 

sands, hard bearing stratum 

overlain by soft or loose soils, 

karst conditions, areas with 

flowing water. Highly variable 

subsurface conditions. 

Conditions requiring long piles 

due to deep scour, liquefiable 

layers, or penetrating very hard 

strata or rock, offshore 

conditions. 
49

3.2.1Shallow Foundations

The feasibility of using spread footings for shallow foundation support should be 

considered in any foundation selection process. Spread footings are generally more 

economical than deep foundations in situations where geomaterial and loading 

conditions are conducive for their use. Favorable conditions include: competent 

soils within shallow depth; foundation width is expected to be relatively small; spread 

footing can be placed at an economical depth (e.g., 10-feet); quality granular fills are 

available (Abu-Hejleh et al. 2014). Spread footings on engineered compacted 

granular fills and mechanically stabilized earth (MSE) granular fills to support bridge 

abutments provide a satisfactory alternative to deep foundations with good 

performance and significant cost savings (Abu-Hejleh et al. 2014). Bridge 

foundations that are subject to scour, liquefaction, or large settlement, as well as 

those with large uplift or lateral load demand are typically not suitable for shallow 

foundations. Additional details on spread footings for highway bridges may be found 

in FHWA GEC-6, Shallow Foundations by Kimmerling (2002).

3.2.2 Shallow Foundations with Ground Improvement

In some situations, ground improvement methods can be used to improve 

subsurface conditions thereby allowing the use of shallow spread footing 

foundations. Ground improvement in conjunction with spread footings should be 

economically evaluated provided they meet strength and service limit state 

requirements as this combination may be more cost effective than deep foundation 

solutions. Additional information on ground improvement may be found in FHWANHI-06-019/020 Ground Improvement Methods (Elias et al. (2006), and at 

http://www.GeoTechTools.org. 

3.2.3 Deep Foundations

Deep foundations are generally needed where the axial compression, axial tension, 

lateral load demand or a combination of the above cannot be satisfied by the near 

surface soil conditions. However, deep foundations should not be used 

indiscriminately for all subsurface conditions and for all structures. There are 

subsurface conditions where a driven pile, drilled shaft, micropile, or CFA pile may 

be very difficult or costly to install. Ground improvement techniques can also be 

used with deep foundations as an economical means to improve lateral resistance in 

weak surficial soils (Rollins and Brown 2011). 
50

3.2.3.1 Driven Piles

Driven piles are the most commonly used deep foundation system for transportation 

projects. Piles are typically installed in groups using an impact pile driving hammer. 

Multiple pile types with various section properties are available to resist almost any 

load demand. Pile lengths for some pile types can be easily adjusted and spliced in 

the field to accommodate variations in subsurface conditions. Further guidance on 

pile selection, pile type advantages and disadvantages, as well as the best pile type 

for a given subsurface condition may be found in Chapter 6.

3.2.3.2 Drilled Shafts

Drilled shafts are frequently used for transportation projects with large axial 

compression or lateral load demand. They are installed by mechanically or 

percussion drilling a hole to the required depth and filling the hole with concrete. 

Sometimes an enlarged base or bell is formed mechanically to increase the toe 

bearing area. Drilling slurry and/or temporary casings can be used when the sides 

of the hole are unstable. Reinforcing steel is installed as a cage inserted prior to 

concrete placement. For complete details on drilled shafts, reference should be 

made to FHWA GEC-10, Drilled Shafts: Construction Procedures and LRFD Design 

Methods by Brown et al. (2010) and AASHTO (2014) design specifications Section 

10.8. 

3.2.3.3 Micropiles

Micropiles are often used for transportation projects in karst areas as well as for 

underpinning, seismic retrofitting, and projects with difficult drilling conditions. 

Micropiles are a small diameter (< 12 inches), reinforced, drilled, and grouted deep 

foundation element. Many different tools and installation methods are available to 

construct micropiles. They are often installed by rotating a casing with a cutting 

edge into the geomaterial or by percussion methods. Cuttings are removed with 

circulating drilling fluid. Reinforcing steel is then inserted and a sand-cement grout 

is pumped through a tremie. Pressurized two-stage grouting techniques are also 

used. The casing can be partially or fully withdrawn. Due to the wide variety of 

micropile types and construction techniques, reference should be made to FHWA05-039, Micropile Design and Construction, Sabatini et al. (2005), and AASHTO 

(2014) design specifications Section 10.9 for complete details. 
51

3.2.3.4 Continuous Flight Auger (CFA) Piles

Continuous Flight Auger (CFA) or alternatively Auger Cast-in-Place (ACIP) piles are 

usually installed by turning a continuous-flight hollow-stem auger into the ground to 

the required depth. As the auger is withdrawn, grout or concrete is pumped under 

pressure through the hollow stem, filling the hole from the bottom up. Vertical 

reinforcing steel is pushed down into the grout or concrete column before it hardens. 

Uplift tension reinforcing can be installed by placing a single high strength steel bar 

through the hollow stem of the auger before grouting. After reinforcing steel is

placed, the pile head is cleaned of any lumps of soil which may have fallen from the 

auger. For complete details on CFA piles, reference should be made to FHWA 

GEC-8, Design and Construction of Continuous Flight Auger Piles by Brown et al. 

(2007).

3.3 ESTABLISHMENT OF A NEED FOR A DEEP FOUNDATION

The first difficult problem facing the foundation designer is to establish whether or 

not the site conditions dictate that a deep foundation must be used. Vesic (1977) 

summarized typical situations in which piles may be needed. These typical 

situations as well as additional uses of deep foundations are shown in Figure 3-1. 

Figure 3-1(a) shows the most common case in which the upper soil strata are too 

compressible or too weak to support heavy vertical loads. In this case, deep 

foundations transfer loads to a deeper competent stratum and act as predominantly 

toe bearing foundations. In the absence of a competent stratum within a reasonable 

depth, the loads must be gradually transferred, mainly through soil resistance along 

the shaft, Figure 3-1(b). An important point to remember is that deep foundations 

transfer load through unsuitable layers to suitable layers. The foundation designer 

must define at what depth suitable soil layers begin in the soil profile.

Deep foundations are frequently needed because of the relative inability of shallow 

footings to resist inclined, lateral, or uplift loads and overturning moments or 

excessive deformations. Deep foundations resist uplift loads by shaft resistance, 

Figure 3-1(c). Lateral loads are resisted either by vertical deep foundations in 

bending, Figure 3-1(d), or by groups of vertical and battered piles, which combine 

the axial and lateral resistances of all piles in the group, Figure 3-1(e). Lateral loads 

from overhead highway signs and noise walls may also be resisted by groups of 

deep foundations, Figure 3-1(f).
52

Deep foundations are often required when scour around footings could cause loss of 

bearing capacity at shallow depths, Figure 3-1(g). In this case the deep foundations 

must extend below the depth of scour and develop their full nominal resistance in the 

support zone below the level of expected scour. FHWA scour guidelines using the 

Hydraulics Engineering Circular No. 18 (Arneson et al. 2012) require the 

geotechnical analysis of bridge foundations to be performed on the basis that all 

stream bed materials in the scour prism have been removed and are not available 

for bearing or lateral support. Costly damage and the need for future underpinning 

can be avoided by properly designing for scour conditions.

Liquefaction and other seismic effects on deep foundation performance must be 

considered for deep foundations in seismic areas. Soils subject to liquefaction in a 

seismic event may dictate that a deep foundation be used, Figure 3-1(h). Seismic 

events can also induce significant lateral loads to deep foundations. During a 

seismic event, liquefaction susceptible soils offer less lateral resistance, reduced 

shaft resistance, and can add drag load to a deep foundation.

Deep foundations are often used as fender systems to protect bridge piers from 

vessel impact, Figure 3-1(i). Fender system sizes and group configurations vary 

depending upon the magnitude of vessel impact forces to be resisted. In some 

cases, vessel impact loads must be resisted by the bridge pier foundation elements. 

Single deep foundations may also be used to support navigation aids.

In urban areas, deep foundations may occasionally be needed to support structures 

adjacent to locations where future excavations are planned or could occur, as in 

Figure 3-1(j). Use of shallow foundations in these situations could require future 

underpinning in conjunction with adjacent construction.

Deep foundations are also used in areas of expansive or collapsible soils to resist 

undesirable seasonal movements of the foundations. Under such conditions, deep 

foundations are designed to transfer foundation loads, including uplift or downdrag, 

to a level unaffected by seasonal moisture movements, Figure 3-1(k). 
53

Figure 3-1 Situations in which deep foundations may be needed 

(modified from Vesic 1977). 
54

3.4 ECONOMIC ASPECTS OF FOUNDATION SELECTION

In many instances, either a shallow or deep foundation alternative is technically 

feasible. Feasible conditions for the use of a shallow spread footing foundation are 

summarized in Abu-Hejleh et al. (2014). Under these circumstances, an evaluation 

of the shallow foundation should include; 1) the dimensions and depth of shallow 

footings based on bearing resistance, 2) the magnitude and time rate of settlement 

under anticipated loads, and 3) detailed cost analysis including such factors as need 

for cofferdams, overall substructure cost, dewatering and foundation seals, 

construction time, construction risk, claims potential, and other project constraints. A 

comparative analysis of feasible deep foundation alternatives should also be 

performed. The cost analysis of feasible alternatives generally has a significant role 

in final selection of the foundation type.

Because this manual deals only with driven pile foundations, other types of 

foundations will not be discussed further.

3.4.1 Foundation Support Cost

A rational comparison of technically viable foundation systems can be made based 

on the foundation support cost of each candidate foundation type. The foundation 

support cost is defined as the total cost of the foundation divided by the factored 

load the foundation supports in tons or kips. Factored loads are used since 

resistance factors vary by foundation type. For driven pile foundation projects, the 

total foundation support cost where axial loading governs can be separated into 

three major components; the pile support cost, the pile cap support cost, and the 

construction control method support cost. The total foundation cost should also 

include all costs associated with a given foundation system including the need for 

excavation or retention systems, dewatering, pile caps and cap size, environmental 

restrictions on construction activities, etc. A detailed discussion on foundation 

support cost applications to driven pile design is presented in Komurka (2015).

It should be noted that the above discussion assumes that the axial compression 

load controls the foundation design. On projects where the axial tension load, lateral 

load, or combined axial and lateral loads govern the foundation design, the support 

cost concept can still be applied. In those situations, the cost per ton of factored 

uplift load or factored lateral load should be used in the selection process. 
55

3.4.1.1 Cost Optimization Example

Figure 3-2 presents a layered soil profile that will be used to illustrate the cost 

optimization process. The first step in the cost optimization of a candidate pile 

section is to perform a static analysis to determine its nominal geotechnical 

resistance versus the pile penetration depth. Static analysis methods are described 

in greater detail in Chapter 7. A static analysis was performed for an HP 14x117 Hpile using the FHWA method in the APILE computer program. The results of that 

static analysis are presented in Figure 3-2, and include the nominal geotechnical 

resistance which is comprised of the nominal shaft resistance and the nominal toe 

resistance at a given pile penetration depth.

Several possible pile penetration termination depths and associated nominal 

resistances are apparent in Figure 3-2. Piles could be driven into the medium dense 

sand layer at a depth of 60 feet for a nominal resistance of 150 kips; seated into the 

medium dense sand layer near a depth of 70 feet for 300 kips; driven through the 

medium dense sand layer and underlying stiff clay layer and into the extremely 

dense sand layer at 110 feet for 1000 kips; or driven to bedrock near 118 feet for 

1700 kips. A rational economic assessment of these potential pile termination 

depths and nominal resistances is needed for cost effective design. For most pile 

types, the pile cost can usually be assumed as linear with depth based on unit price 

as illustrated in Figure 3-3. However, this may not be true for very long concrete 

piles or long, large section steel piles. These exceptions may require the cost 

analysis to reflect special transportation, handling, or splicing costs for the concrete 

piles or extra splice time and cost for steel piles.

Figure 3-4 presents a plot of nominal resistance pile support cost versus pile 

penetration depth. This figure is obtained by dividing the pile cost at a given depth 

from Figure 3-3 by the nominal resistance at that same depth from Figure 3-2. Note 

that Figure 3-4 includes the nominal resistance versus depth in addition to the pile 

support cost. For the H-pile section evaluated, a general conclusion can quickly be 

reached; piles penetrating to more than 60 feet will be more cost effective than 

shorter piles. At the top of the sand layer near 60 feet, the pile has a nominal 

resistance of 150 kips and a nominal resistance support cost of $39 per kip. Near 

the middle of this sand layer at a penetration depth of 70 feet, the pile has a nominal 

resistance of 300 kips and a much lower nominal resistance support cost of $19 per 

kip. The nominal resistance pile support cost drops further to $9 per kip at a depth 

of 110 feet and a nominal resistance of 1000 kips. Finally, the nominal resistance 

pile support cost continues to decrease to $5.50 per kip at a depth of 118 feet and a 

nominal resistance of 1722 kips. 
56

An HP 14x117 H-pile section with a yield strength of 50 ksi has a nominal structural 

resistance of 1720 kips when fully embedded. An example detailing the calculation 

of the nominal structural resistance of an H-pile is provided in Section 8.5.3. The 

maximum factored load that can be placed on the pile is the lesser of the factored 

structural resistance or the factored geotechnical resistance. For the H-pile section 

used, the maximum factored structural resistance is 1032 kips when good driving 

conditions are anticipated based on the subsurface conditions, pile damage is 

unlikely, and a driving shoe is therefore not required. The maximum factored 

structural resistance decreases to 860 kips when severe driving conditions are 

anticipated based on the subsurface conditions, damage is possible, and a driving 

shoe is required. Figure 3-5 includes the nominal and factored structural resistances 

in the strength limit state under axial compression loading. The structural resistance 

for all pile types is discussed in Chapter 8. 

Figure 3-2 Soil profile and nominal geotechnical resistances versus depth. 
57

Figure 3-3 Pile cost versus pile penetration depth.

Figure 3-4 Nominal resistance of pile support cost versus penetration depth. 
58

Figure 3-5 Nominal and factored resistances for axial compression loads.

Figure 3-5 also includes the nominal and factored geotechnical resistances in the 

strength limit state under axial compression loading. Geotechnical aspects of limit 

state design are covered in Chapter 7. The factored geotechnical resistances 

versus pile penetration depth shown in Figure 3-5 are based on construction control 

with the following field verification methods:

φdyn = 0.80 Driving criteria established by successful static load test of at least one 

pile per site condition and dynamic testing with signal matching of at 

least two piles per site condition, but no less than 2% of production piles.

φdyn = 0.75 Driving criteria established by successful static load test of at least one 

pile per site condition without dynamic testing or by dynamic testing with 

signal matching on 100% of the production piles. 
59

φdyn = 0.65 Driving criteria established by dynamic testing with signal matching of at 

least two piles per site condition, but no less than 2% of production piles. 

φdyn = 0.50 Driving criteria established by wave equation analysis based on end of 

drive conditions only.

φdyn = 0.40 Driving criteria established by FHWA modified Gates dynamic formula 

based on end of drive conditions only.

The required nominal driving resistance, Rndr, needed to utilize the full factored 

structural resistance, Pr, of either 1032 or 860 kips with a given resistance 

verification field method is provided in Table 3-3. The required nominal driving 

resistance for full utilization of the factored structural resistance can be calculated by 

dividing the factored structural resistance by the resistance factor of the field 

verification method, φdyn. 

The estimated pile length for the nominal driving resistance, Rndr, can then be 

determined from Figure 3-5 by the depth where the nominal geotechnical resistance, 

Rn equals the nominal driving resistance value for the field method. The 

abbreviations used in the table for the field verification methods are as described 

above with SLT for static load test, DLT for dynamic load test, WEA for wave 

equation analysis and DF for dynamic formula. Note that a drivability strength limit 

check will preclude use of the maximum factored structural resistance, Pr in some 

cases due to the nominal driving resistance, Rndr, required by that field method.

Table 3-3 Pile Length Estimates Based on Field Methods and Factored 

Resistances

Field Method φdyn Rndr

Pr = 1032 kips

Estimated 

Length

Rndr

Pr = 860 kips

Estimated 

Length

SLT & 2% DLT 0.80 1290 kips 116 ft 1075 kips 113 ft

SLT or 100% DLT 0.75 1376 kips 117 ft 1147 kips 115 ft

2% DLT 0.65 1588 kips 118 ft 1323 kips 117 ft

WEA 0.50 2064 kips 118 ft+ Rock 1720 kips 118 ft

FHWA Gates DF 0.40 2580 kips 118 ft+ Rock 2150 kips 118 ft+ Rock

Figure 3-5 also illustrates the benefit of using a more reliable field verification 

method for piles terminated in the geomaterials above bedrock. If an HP 14x73 

section is chosen for this subsurface condition instead of the HP 14x117 section, the 

maximum factored structural resistance for a fully embedded HP 14x73 section in 

severe driving conditions is 520 kips. For illustrative purposes, the static analysis 
60

results are assumed unchanged for a HP 14x73 section compared to the HP 14x117 

section based on roughly the same pile perimeter and the same plugged pile toe 

area. The estimated pile length from Figure 3-5 to geotechnically utilize the 

maximum factored structured resistance is 104 feet for field verification with a static 

load test and dynamic testing (φdyn = 0.80) compared to an estimated length of 116 

feet for field verification with the modified Gates dynamic formula (φdyn = 0.40).

The results of the static analysis on the HP 14x117 H-pile section can be also used 

to evaluate the pile penetration requirements for the axial tension load demand. The 

shaft resistance from Figure 3-2 is presented once again in Figure 3-6 and labeled 

Rs for the nominal shaft resistance. The pile cost at a given depth from Figure 3-3

can be divided by the nominal shaft resistance at that same depth to obtain a plot of 

the pile support cost in dollars per kip of nominal uplift resistance versus pile 

penetration depth. Figure 3-6 presents plots of both the nominal uplift resistance 

pile support cost versus depth and the nominal uplift resistance versus depth. Two 

distinct plateaus in the pile support cost for tension loading are apparent, $90 to 

$100 per kip of nominal uplift resistance above 60 feet and $15 to $25 per kip of 

nominal uplift resistance below 70 feet. 

Figure 3-6 Nominal shaft resistance and pile support cost.
61

Figure 3-7 presents the nominal shaft resistance, Rs, and the factored axial tension 

resistance versus depth, Rr. In Figure 3-7, the nominal shaft resistance, Rs, is 

multiplied by the uplift resistance factors, φup, of 0.60 if the uplift resistance is 

determined with a static load test, 0.50 if obtained from a dynamic load test with 

signal matching, and 0.25 in clay layers and 0.35 in sand layers for uplift resistance 

determined by the α-method and Nordlund static analysis methods.

Figure 3-7 Nominal and factored resistances for axial tension loads. 

3.5 OTHER CONSIDERATIONS

In many instances, several driven pile alternatives may be technically feasible for 

satisfying the project load and deformation requirements. Other considerations such 

as constructability, noise, vibrations, effects on nearby structures and utilities, 

durability, long term maintenance, and other considerations should also be 

evaluated to determine the most appropriate driven pile selection.
62

3.5.1 Constructability

Constructability should also be assessed during the design stage with any 

construction considerations factored into the pile selection and final design. 

Drivability is a key component of constructability and must be evaluated. The 

chosen pile section must be capable of being installed to the specified minimum 

penetration depth dictated by scour, extreme event or deformation considerations. 

The design and drivability assessment should consider the presence of any hard 

layers that must be penetrated to meet design requirements that could necessitate 

use of pile penetration aids such as predrilling or jetting. Constructability 

considerations and drivability analysis should also consider potential densification 

effects of soils constrained within cofferdams or soils that may densify from driving 

large pile groups. Similarly, the influence of soil heave on adjacent structures, 

utilities, or other piles from driving displacement piles in a cohesive deposit should 

be considered along with the need for pile penetration aids to mitigate the heave

potential, if present. The cost of a driven pile foundation will increase on projects 

with limitations or restrictions on head room, work space, and construction access. 

Similarly, environmental restrictions on vibrations, noise, work hours, marine 

mammals, and fish will likely increase foundation costs.

3.5.2 Consideration of Pile Driving Noise

Driven piles are installed by impact hammers. Noise levels associated with typical 

impact pile driving activities depend upon the hammer and pile type used. Noise 

from impact pile driving operations typically ranges from around 80 to 135 dBa. If 

local ordinances dictate allowable noise levels at or below this level, some driving 

equipment may not meet these requirements. Manufacturers of a few diesel and 

hydraulic hammers can provide optional noise suppression devices that may reduce 

the pile driving generated noise by about 10 dBa. Independently manufactured 

devices are also available. Additional information on noise suppression equipment 

is presented in Chapter 15.

In noise sensitive areas, the foundation designer should review any noise 

ordinances to determine if pile driving noise suppression devices would be 

necessary and if so, the impact this may have on the contractor’s equipment 

selection and productivity. If limits on work hours, pile equipment type, or noise 

suppression equipment are required, costs associated with these limitations should 

be considered in the foundation selection process.
63

3.5.3 Pile Driving Induced Vibrations

Since piles are driven by impact or vibratory hammers, ground vibrations of some 

magnitude are always induced into the surrounding soils during pile installation. In 

many situations these vibrations do not pose a significant threat or risk to structures. 

The vibrations created by pile driving depend on the soil type, pile type and section, 

pile hammer, installation techniques, penetration resistance, pile toe penetration 

depth, and distance from the pile. Therefore, the distance from a pile driving 

operation where these variables combine to cause vibration levels that are a threat 

to structures, facility operations, or utilities varies.

The ground vibration level where vibration induced soil densification and settlement 

or structural damage from direct vibrations may occur depends upon the vibration 

magnitude, frequency, and number of cycles as well as the type and condition of the 

existing structure or facility. In addition, nearby structures may house sensitive 

equipment whose operation may be adversely affected by pile driving vibrations. 

Establishing appropriate vibration limits for sensitive equipment can be difficult, 

which makes identifying and addressing such installations all the more important.

For pile driving projects with structures, facilities, or utilities within a zone potentially 

affected by pile driving vibrations, careful evaluation of the pile driving procedures 

and/or monitoring of ground vibrations during pile installations should be performed 

by personnel with vibration monitoring and mitigation experience. Even if a designer 

is satisfied the pile driving vibrations pose no threat to structures or equipment, 

advance notification to neighbors of pile driving activities, pre-construction condition 

surveys, and vibration monitoring plans promotes a “good neighbor” policy which 

can reduce the risk of claims.

Potential damage to nearby structures can result from two mechanisms:

1. Vibrations induced soil densification and settlement. 

2. The effects of vibrations on the structure itself.

Lacy and Gould (1985) found that vibratory induced soil densification settlements 

and resulting structural damage can occur at ground surface peak particle velocities 

much less than 2 inches per second and that soil gradation is an important factor in 

this phenomenon. They reported that significant vibration induced settlements 

occurred at some sites with peak particle velocities measured on the ground surface 

as low as 0.1 to 0.2 inches per second. Sands particularly susceptible to vibratory 
64

densification were late Pleistocene deposits with uniformity coefficients of up to 4 or 

5 and relative densities of up to 50 or 55%.

Wiss (1981) reported typically recommended "safe" ground vibration levels for 

structures between 0.5 and 4 inches per second. In codes, such as NFPA 495 

(2013), the maximum allowable particle velocity to prevent the onset or propagation 

of hairline cracks in plaster or drywall is a function of the vibration frequency. For 

example, a particle velocity of 1 inch per second at 30 Hz would be below NFPA 495 

code limits but would be above code limits if the vibration frequency were 10 Hz.

Bay (2003) summarized relationships between peak particle velocity and the 

distance from the pile as a function of rated hammer energy. These results were 

plotted against typical damage thresholds for various types of structures. Charts for 

Class II and Class III soils were provided and are reproduced in Figure 3-8. Class II 

soils were defined as competent soils with SPT-N values of 5 to 15 blows per foot. 

Class III soils are hard soils with SPT N values of 15 to 50 blows per foot. Bay noted 

that stiff soil crusts near the ground surface can significantly increase the vibration 

levels from those noted in the charts and that other factors that can influence the 

vibration levels include nearby deep excavations, rock outcrops, and shallow 

bedrock. Soil-structure interaction should also be considered in assessing vibration 

levels and damage potential. Therefore, while informative, these charts should not 

be used to eliminate vibration monitoring.

If the potential for damaging ground vibrations is high, pile installation techniques 

should be specified to reduce vibration levels. Specifications could require 

predrilling or jetting, cutoff trenches, as well as use of a different pile type or use of a 

specific type of pile hammer. Since predrilling and jetting influence compression, 

uplift, and lateral pile capacities, a determination of probable vibration levels and 

remediation measures should be evaluated in the design stage. A case history 

illustrating how changing pile installation procedures reduced vibratory induced soil 

densification and off-site settlement damage was reported by Lukas and Gill (1992).

NCHRP Project 20-5, Dynamic Effects of Pile Installations on Adjacent Structures, 

by Woods (1997), provides a synthesis of pile driving induced vibrations and typical 

mitigation practices. This synthesis noted that vibration problem management is the 

key to minimizing vibration damage, delays and claims. Two important elements in 

vibration management are a vibration specification with limits on the frequency

dependent maximum peak particle velocity and a predriving survey of surrounding 

structures. An example vibration specification that details the requirements of a 

preconstruction survey as well as particle velocity controls is included in the NCHRP
65

Figure 3-8 Predicted vibration levels for class II and class III soils 

(after Bay 2003).
66

synthesis. The predriving survey needs to document conditions within the potential 

effected area. Woods reported that vibration damage is not likely to occur at a 

distance from a driven pile of 50 feet for piles 50 feet long or less, or at a distance 

greater than one pile length away from driving for piles longer than 50 feet.

Piles with low impedances, EA/C, tend to transmit the hammer energy to the soils 

along the pile shaft and thus increase ground vibrations, whereas piles with higher 

impedances tend to more effectively transmit the hammer energy to the pile toe 

resulting in lower ground vibration levels (Woods 1997; Massarsch and Fellenius 

2008). Hence, selection of a stiffer pile section at sites where vibrations are a 

concern may help reduce vibration problems.

Massarsch and Fellenius (2014) describe ground vibration effects on soil settlement. 

In loose sands and silts, vibratory densification beneath adjacent structures can 

occur, therefore causing deformation of the adjacent foundation. As discussed, pile 

dimensions influence the area affected by ground vibrations. Furthermore they 

propose that settlement occurs in an inverted 2(V):1(H) triangular region around the 

pile as depicted in Figure 3-9. Maximum settlement adjacent to the pile and average 

settlement within a distance of 3b + D/2 from the pile are estimated using Equations 

3-1 and 3-2. A compression factor is applied based on in-situ soil density and 

vibration magnitude, as shown in Table 3-4 for sands.

Maximum settlement Eq. 3-1 

Average settlement Eq. 3-2 

Where:

α = compression factor.

D = pile embedded length (feet).

b = pile diameter or width (feet).
67

Figure 3-9 Settlement area and magnitude around a pile in homogeneous sand 

deposit (after Massarsch and Fellenius 2014). 

Table 3-4 Compression Factor, α, for Sand Based on Soil Density and Level of 

Driving Energy (after Massarsch and Fellenius 2014)

Soil Density Low Ground 

Vibration

Medium Ground 

Vibration

High Ground 

Vibration

Very Loose 0.02 0.03 0.04

Loose 0.01 0.02 0.03

Medium 0.005 0.01 0.02

Dense 0.00 0.005 0.01

Very Dense 0.00 0.00 0.005
68

The Pile Driving Contractors Association (2015) while working with The Citadel and 

WPC has a pile driving vibration database with 23 case histories and over 1800 

individual pile-ground vibration measurements. Data has been submitted from a 

number of design and testing firms, with the goal of allowing contractors and 

designers to make reasonable assessments of the potential vibration effects from 

driven pile installations. This database should only be used for general guidance as 

differing site conditions, construction equipment and installation techniques may not 

be applicable in each unique situation.

3.5.4 Condition Surveys and Vibration Monitoring of Adjacent Facilities

Many projects are built in populated urban areas where existing condition surveys 

and vibration monitoring programs are warranted. Prior to starting construction, the 

internal and external condition of structures potentially affected by the construction 

activity should be documented using photos and videos. The condition of any 

existing cracks should be documented and crack monitors installed as needed. This 

documentation provides a preconstruction baseline of the existing conditions that 

can later be compared to any damage claims should they occur. Following 

construction, a similar survey should be performed with photos to document the 

change, if any, in the existing structures. Condition surveys should be performed 

using film rather than digital photography should damage claims be made.

The distance that should be surveyed on pile driving projects depends on the soil 

conditions, the pile type, the pile driving equipment, and the pile driving procedures. 

Many states have adopted a survey distance based on the rated pile hammer 

energy. Others agencies have policies based solely on the distance away from the 

pile driving operation. For most routine cases, a preconstruction and post 

construction survey along with a vibration monitoring program that documents 

conditions within 200 feet of the pile driving operation should be sufficient. However, 

this distance may need to be extended to as much as 500 feet for older structures, 

structures or utilities in poor condition, or highly vibration sensitive equipment.

3.5.5 Durability Considerations

The durability of a pile foundation can be defined by how long it performs 

satisfactorily, without unforeseen high maintenance costs relative to the foundations 

expected service life. AASHTO has mandated a design life of 75 years for new 

ordinary bridges. Complex major bridge structures may have a design life of 100 

years. Hence, structure foundation durability over the design life is an important 

consideration. In general, driven piles have proved very durable in most 
69

environments. However, environments exist for driven piles as well as any other 

deep foundation type where durability is an important design consideration to satisfy 

design life requirements.

In 2013, a pier supported on H-piles at the Leo Frigo bridge carrying I-43 over the 

Fox River in Green Bay, WI settled 2 feet during the night. The sudden settlement of 

the 32 year old structure resulted from severe corrosion and the resultant buckling 

and shearing of the foundation H-piles. Figure 3-10 illustrates the corrosion of the 

foundation piles uncovered at the pier location during the post event investigation. 

In the final investigation report on the event by Michael Baker, Jr. Inc., (2015) the 

corrosion and deterioration was attributed to a highly unusual set of occurrences at 

the pier location. Fill materials included the presence of porous, industrial fly ash in 

contact with the H-piles. The porous material permitted oxygen access to the piles. 

Also the fill material contained high amounts of sulfates and the water and soil 

surrounding the pile section contained high concentrations of chlorides. This 

unusual set of conditions created a highly aggressive environment for steel piles.

Figure 3-10 Corrosion and resulting buckling of foundation pile 

(courtesy Wisconsin Department of Transportation).
70

Gu at al. (2015) reported on the results of a corrosion study performed to evaluate 

the reuse of 34 year old H-pile foundations. The original project was for the Girard 

Avenue Bridge in Philadelphia, PA and piles were installed through a cinder ash fill 

located at the water table in close proximity to a tidal river. This study found that the 

piles had a 12% loss in section over their 34 year old life. This site specific corrosion 

rate was then used to assess future section loss over the rehabilitated structure life 

for both existing piles that were reused as well as new piles. The section loss rate 

was also used to evaluate the nominal pile resistance under axial and lateral 

loading. It was also noted that site specific corrosion rate of 0.0015 inches/year was 

less than published rates. Section 6.13 provides details on steel pile protection 

measures and design considerations for durability in aggressive environments.

Figure 3-11 Concrete pile damage from corrosion effects on concrete pile 

reinforcement (courtesy of Moser 2011).
71

Figure 3-12 Concrete surface abrasion and deterioration in tidal zone 

(courtesy of Moser 2011).

The durability of concrete piles should also be evaluated during design so that the 

foundation satisfies the target design life. Concrete piles can deteriorate due to 

chloride intrusion and resulting reinforcement and prestressing strand corrosion. 

Moser et al. (2011) noted the three main factors contributing to concrete 

deterioration in coastal structures include the high chloride content in seawater, the

potential for high temperature and relative humidity, and the cyclic wetting due to 

tidal and splash action. Figure 3-11 shows two concrete piles with severe damage 

and cracking due to corrosion of the reinforcement steel. Abrasion of concrete piles 

in intertidal zone, as depicted in Figure 3-12, can also accelerate deterioration. 

Concrete abrasion can also occur at the mudline due to littoral drift of bottom 

sediments. Moser (2011) reported that some prestressed concrete piles in coastal 

Georgia have severe corrosion damage in the splash zone after less than 25 years 

of service which is well under the target design life.

Holland et al (2014) summarized multiple deterioration mechanisms affecting the 

concrete piles on a 38 year old I-95 bridge over Tuttle Creek. Their forensic 

investigation determined damage occurred from chloride-induced corrosion, severe 

deterioration of concrete from sulfate attack, and coarse aggregate attack from 

Cliona boring sponges.
72

To increase concrete pile durability in aggressive environments, high performance 

concrete mixes are being design and studied. The dense, impermeable, concrete 

that results with high performance concrete mixes should better protect 

reinforcement against the ingress of chlorides, sulfates, and seawater, as well as 

resist freeze/thaw and chemical attacks.

Timber piles also have durability considerations. Bigelow et al. (2007) evaluated 

several timber pile supported bridges in Iowa as part of a structure life assessment. 

Figure 3-13 illustrates the durability issues they documented as a result of a break 

down in the preservative barrier including a) mechanical damage, b) abrasion or 

debris damage, c) fire damage, and d) weathering. One of the most common

durability issues occurred on piles in the stream channel as abrasion or debris 

damage from floating debris and/or ice. The study found that creosote treated 

timber piles up and away from the stream channel lased 60 to 70 or more years and 

those in the stream channel have a life expectancy of 40 to 50 years.

Figure 3-13 Examples of timber pile deterioration 

(courtesy Bridge Engineering Center at Iowa State University).

3.6 UNANTICIPATED OCCURRENCES

The long term performance of a foundation is generally assessed by its total and 

differential settlement within and between substructure locations. Unanticipated 

occurrences can result from outside influences, such as a fill stockpile can affect a 

foundation’s long term performance as shown in Figure 3-14 and Figure 3-15. 

Construction of a new pile foundation can also result in unanticipated outside 

influences as described in Section 3.6.2.
73

3.6.1 Fill Stockpile

During the summer of 2104, a portion of the 40 year old elevated section of I-495 

near Wilmington, Delaware was noticed to have substantially tilted over a relatively 

short time period. The tops of some of the 50 feet tall columns supporting the 

elevated roadway were noted to be almost 2 feet out of alignment from their base as 

shown in Figure 3-14. Engineers determined a roughly 50,000 ton stockpile of fill 

material had been placed near the four affected piers. The weight of this fill had 

caused lateral movement of soil beneath the bridge, deformed the H-piles supporting 

the piers, and cracked the pile caps as illustrated in Figure 3-15. This caused four of 

the piers supporting the elevated roadway to tilt. In the above case, the structure 

was performing as expected based on its previous inspection report until being 

affected by the stockpile of fill material. 

Figure 3-14 Tilting columns adjacent to stockpile (courtesy DelDOT).
74

Figure 3-15 Cracked pile cap from lateral soil movements (courtesy DelDOT).

3.6.2 Adjacent Construction

In 2015, the five span bridge carrying I-65 over Wildcat Creek was being widened 

from two to three travel lanes. Due to scour considerations, the new construction 

was supported on a driven H-pile foundation rather than spread footings. Shortly 

after production pile installation, the pier supported by the adjacent spread footing 

foundation settled as much as 10 inches and rotated 7 inches, ENR (2015). This 

resulted in closing the I-65 bridge to stabilize the pier. The movements were 

attributed to the release of artesian pressures and sand beneath the spread footing 

supported pier foundation.
75

REFERENCES

Abu-Hejleh, N., Kramer, W.M., Mohamed, K., Long, J.H., and Zaheer, M.A. (2013). 

Implementation of AASHTO LRFD Design Specifications for Driven Piles, 

FHWA-RC-13-001. U.S. Dept. of Transportation, Federal Highway 

Administration, 71 p.

Arneson, L.A., Zevenbergen, L.W., Lagasse, P.F., and Clopper, P.E. (2012). 

Evaluating Scour at Bridges, Fifth Edition, FHWA-HIF-12-003, Hydraulic 

Engineering Circular (HEC) No. 18. U.S. Dept. of Transportation, Federal 

Highway Administration, 340 p.

Bay, J.A. (2003). A Summary of the Research on Pile Driving Vibrations. 

Proceedings of the Pile Driving Contractor’s Association 7th Annual Winter 

Roundtable, Atlanta, 13 p.

Bigelow, J., Clausen, C., Lebow, S., Greimann, L. (2007). Field Evaluation of Timber 

Preservation Treatments for Highway Applications, CRTE 06-252. Iowa 

Highway Research Board, Iowa Dept. of Transportation, Ames, IA, 104 p.

Bowles, J.E. (1977). Foundation Analysis and Design. Second Edition, McGraw-Hill 

Book Company, Blacklick, OH, 750 p.

Brown, D.A., Dapp, S.D., Thompson, W.R., and Lazarte, C.A. (2007). Design and 

Construction of Continuous Flight Auger (CFA) Piles. FHWA-HIF-07-03, 

Geotechnical Engineering Circular (GEC) No.08. U.S. Dept. of 

Transportation, Federal Highway Administration, 289 p.

Brown, D. A., Turner, J.P. and Castelli R.J. (2010). Drilled Shafts: Construction 

Procedures and LRFD Design Methods, FHWA-NHI-10-016, Geotechnical 

Engineering Circular (GEC) No. 10. U.S. Dept. of Transportation, Federal 

Highway Administration, 970 p.

Elias, V., Welsh, J.P., Warren, J., Lukas, R.G., Collin J.G., and Berg, R.R. (2006). 

Ground Improvement Methods Volumes I and II, FHWA-NHI-06-019 and 

FHWA NHI-06-020. National Highway Institute, Federal Highway 

Administration, U.S. Department of Transportation, Washington D.C.
76

Engineering News Record (2015). Pile Driving and Poor Soils Prompt Indiana I-65 

Shutdown.

Gu, L., McInnes, S.E., Fenton, J.H., Welsh, S.J., and Mouradian, A.G. (2015). 

Corrosion Study of Existing Steel H-pile Installed Through Cinder-Ash Fill. 

Proceeding of the 2015 International Foundations Congress and Equipment 

Expo, San Antonio, TX, pp. 961-972. 

Holland, R.B., Kurtis, K.E., Moser, R.D. Kahn, L.F., Aguayo, F. and Singh, P.M. 

(2014). Multiple Deterioration Mechanisms in Coastal Concrete Piles, A 

Forensic Case Study, Concrete International, Vol. 36, No. 7, pp. 45-52.

Kimmerling, R.E. (2002). Shallow Foundations, FHWA-IF-02-054, Geotechnical 

Engineering Circular (GEC) No. 6. U.S. Dept. of Transportation, Federal 

Highway Administration, Washington, D.C., 310 p.

Komurka, V. (2015). Foundation Support Cost – Applications to Driven-Pile Design. 

Proceedings of the 2015 International Foundations Conference and 

Equipment Exposition, San Antonio, TX, pp. 990-1005.

Lacy H.S. and Gould, J.P. (1985). Settlement from Pile Driving in Sands, Vibration 

Problems in Geotechnical Engineering. American Society of Civil Engineers 

(ASCE), Special Technical Publication, New York, NY, pp. 152-173.

Lukas, R.G. and Gill, S.A. (1992). Ground Movement from Piling Vibrations. PilingEuropean Practice and Worldwide Trends, Proceeding of the Conference ICE 

Telford House, London, pp. 163-169.

Massarsch, K.M, and Fellenius, G.H. (2008). Ground Vibrations Induced by Impact 

Pile Driving. The Sixth International Conference on Case Histories in 

Geotechnical Engineering, Arlington, VA, 38 p.

Massarsch, K.R., and Fellenius, B.H. (2014). Ground Vibrations from Pile and Sheet 

Pile Driving. Part 1 Building Damage. Proceedings of the DFI-EFFC 

International Conference on Piling and Deep Foundations, Stockholm, 

Sweden, pp. 131-139.

Michael Baker, Jr., Inc. (2015). Investigation Report, I-43 Leo Frigo Memorial Bridge, 

Volume I, 181 p.
77

Moser, R., Holland, B. Kahn, L., Singh, P., and Kurtis, K. (2011). Durability of 

Precast Prestressed Concrete Piles in Marine Environment: Reinforcement 

Corrosion and Mitigation Part 1. GDOT Research Project No. 07-30. Office of 

Materials and Research Georgia Department of Transportation, 243 p.

Munfakh, G., Arman, A., Collin, J.G., Hung, J.C.-J., and Brouillette, R.P. (2001). 

Shallow Foundations Reference Manual, FHWA-NHI-01-023. National 

Highway Institute, Federal Highway Administration, Washington, D.C., 222 p.

National Fire Protection Association, (NFPA). (2013). Explosive Materials Code 

(495) - Chapter 11, Ground Vibration, Air Overpressure, Flyrock, and Gases, 

National Fire Protection Association, Quincy, MA, 58 p.

Pile Driving Contractors Association (PDCA). (2015). National Pile Driving Noise 

and Vibration Database, http://www.piledrivers.org. 

Rollins, K.M., and Brown, D.A., (2011). Design Guidelines for Increasing the Lateral 

Resistance of Highway-Bridge Pile Foundations by Improving Weak Soils. 

NCHRP Report No. 697. Washington, D.C., 98 p.

Sabatini, P.J., Tanyu, B., Armour., P., Groneck, P., and Keeley, J. (2005). Micropile 

Design and Construction, FHWA-NHI-05-039. National Highway Institute, 

U.S. Dept. of Transportation, Federal Highway Administration, Washington, 

D.C, 436 p.

Vesic, A.S. (1977). Design of Pile Foundation, Synthesis of Highway Practice No. 

42. National Cooperative Highway Research Program, Transportation 

Research Board, National Research Council, Washington, D.C., 68 p.

Wiss, J.F. (1981). Construction Vibrations: State-of-the-Art. American Society of Civil 

Engineers (ASCE), Journal of the Geotechnical Engineering Division, , Vol. 

107, No. 2, pp. 167-181.

Woods, R.D. (1997). Dynamic Effects of Pile Installations on Adjacent Structures. 

NCHRP Synthesis 253, National Cooperative Highway Research Program, 

Transportation Research Board, Washington, D.C., 96 p.
78
79

CHAPTER 4

SITE CHARACTERIZATION

4.1 INTRODUCTION

The design of a structure's foundation requires adequate knowledge of the 

subsurface conditions at the construction site. The absence of a thorough 

foundation study or adequate geotechnical data often leads to (1) a foundation 

system with a large safety margin which is generally a more expensive foundation 

and in some cases one that may be difficult to construct, or to (2) an unsafe 

foundation, or to (3) construction disputes and claims.

Site characterization consists of a subsurface exploration program and subsequent 

testing to determine geomaterial parameters for foundation design. This information 

is also used in construction feasibility and planning decisions. For reliability based 

design, reducing uncertainty with respect to geomaterial properties may lead to more 

economical foundations, therefore a carefully implemented soil exploration and 

testing program is beneficial. This chapter will focus on site characterization based 

on the assumption that a driven pile deep foundation has been selected for design.

An experienced geotechnical engineer is of great value to the design team and 

assists with planning the subsurface exploration, ordering field and laboratory tests 

of geomaterials, determining the design subsurface profile, and recommending 

technically feasible and appropriate foundation systems. Subsurface exploration 

and/or laboratory testing results are often relayed in technical documents from 

specialty firms directly to the geotechnical engineer who then compiles this 

information into a geotechnical design report. This document is used by multiple 

engineering disciplines during the foundation design and construction phases. In 

some agencies, design memoranda and transmittals are used instead of 

geotechnical reports. In these situations, regular communication between 

engineering disciplines is essential for effective foundation design and construction.

Guidance on site characterization and the evaluation of geomaterial properties for 

geotechnical design is provided in several FHWA reference documents including 

FHWA-NHI-01-031, Mayne et al. (2002); GEC-5, Sabatini et al. (2002); and FHWANHI-06-088, Samtani and Nowarzki (2006). Laboratory and in-situ testing are also
80

integral to a site characterization program. These topics will be discussed further in 

Chapter 5 of this manual.

4.2 SITE CHARACTERIZATION PROGRAM

Site characterization encompasses the gathering and reporting of site conditions 

required to safely and economically design foundations and other earth structures, 

as well as to address constructability considerations. The information necessary to 

adequately characterize the site will vary depending upon the site geologic 

conditions, foundation types being considered for the structure, the structure 

performance requirements, the acceptable risk level for the structure, 

constructability, as well as owner preferences.

The site exploration should be carefully organized, and consists of three main 

phases. These are (1) planning the exploration program and data collection (office 

work), (2) completing a field reconnaissance survey, and (3) performing a detailed 

subsurface exploration program (boring, sampling, and in-situ testing). Each phase 

should be planned so that a maximum amount of information can be obtained at a 

minimum cost. Each phase also adds to, or supplements, the information from the 

previous phase. Table 4-1 lists the purpose of each exploration phase.

Depending upon the project size and contract delivery method, a single phase or 

staged subsurface exploration program may be appropriate. A single phase 

program is common for small project and staged subsurface exploration programs 

are often used for large projects. Staged programs are common for design-build 

projects where the owner performs a preliminary subsurface exploration program 

and the design-build team performs the final subsurface exploration program.

4.2.1 Data Collection

The purpose of this phase is to obtain information about the proposed structure and 

general information on subsurface conditions. The structural information can be 

obtained from studying the preliminary structure plan prepared by the bridge design 

office and by meeting with the structural designer. Approach embankment 

preliminary design and performance requirements can be obtained from the roadway 

office. General information about the subsurface conditions can be obtained from a 

variety of sources, some of which are listed in Table 4-2. Additional sources of 

historical data are listed in GEC-5, Sabatini et al. (2002). The data collection phase 
81

Table 4-1 Subsurface Exploration Phases

Phase Activity Purpose Remarks

1. Planning the 

Exploration

(Office Work).

A. Obtain structure information and 

determine:

1. Type of structure.

2. Preliminary location of piers and 

abutments.

3. Loading and special design events.

4. Allowable differential settlement, lateral 

deformations, and other performance 

criteria.

5. Any special features and requirements.

B. Obtain drilling records for nearby 

structures and from local well drillers.

C. Perform literature reviews including 

maintenance records, pile driving records, 

scour history, etc.

Obtain overall picture of subsurface conditions 

in the area.

See Table 4-2 for 

sources of 

information.

2. Field 

Reconnaissance 

Survey. 

Verify information gained from the office 

phase and plan the detailed subsurface 

exploration.

A. Observe, confirm and collect information 

regarding:

1. Topographic, geologic, and hydrologic 

features.

2. New and old construction in the area 

including utilities. Performance of 

existing structures.

3. Drilling equipment required, cost, and 

access for the equipment.

B. If appropriate, conduct geophysical testing 

to obtain preliminary subsurface 

information.

Field reconnaissance 

is often conducted by 

a multi-disciplined 

team.

3. Detailed 

Subsurface 

Exploration. 

Develop a preliminary boring plan based on 

phases 1 and 2. Conduct a preliminary 

evaluation for viable foundation systems 

including ground improvement. Determine 

subsurface requirements for all of the viable 

foundation systems. The boring plan should 

be modified if needed as the borings are 

performed and detailed subsurface 

information is obtained.

The subsurface exploration should provide the 

following:

1. Depth and thickness of strata 

(subsurface profile).

2. In-situ field tests to determine soil 

design parameters.

3. Samples to determine soil and rock 

design parameters.

4. Groundwater levels including perched, 

regional, and any artesian conditions.

For major structures, 

the pilot boring 

program is often 

supplemented with 

control and 

verification boring 

programs.
82

Table 4-2 Sources of Subsurface Information and Use

Source Use

Preliminary structure plans prepared by 

the bridge design office.

Determine:

1. Type of structure.

2. Preliminary locations of piers and

abutments.

3. Footing loads and special design 

events.

4. Allowable differential settlement, 

lateral deformations, and 

performance criteria.

5. Any special features and 

requirements.

Geotechnical databases. Nearby soil boring logs, laboratory test 

results, and in-situ test results. 

Construction plans and records for 

nearby structures.

Foundation type, old boring data, 

construction information including 

construction problems.

Topographic maps prepared by the 

United States Coast and Geodetic 

Survey (USC and GS), United States 

Geological Survey (USGS) and State 

Geology survey.

Existing physical features shown; find 

landform boundaries and determine 

access for exploration equipment. Maps 

from different dates can be used to 

determine topographic changes over 

time.

County agricultural soil survey maps 

and reports prepared by the United 

States Department of Agriculture 

(USDA).

Boundaries of landforms shown; 

appraisal of general shallow subsurface 

conditions.

Aerial photos prepared by the United 

States Geological Survey (USGS) or 

satellite imagery such as Google Earth.

Detailed physical relief shown; gives 

indication of major problems such as old 

landslide scars, fault scarps, buried 

meander channels, sinkholes, or scour; 

provides basis for field reconnaissance.

Well drilling record or water supply 

bulletins from state geology or water 

resources department.

Old well records or borings with general 

soils data shown; estimate required 

depth of explorations and preliminary 

cost of foundations.

Geologic maps and Geology bulletins. Type, depth and orientation of rock 

formations.

prepares the engineer for the field reconnaissance survey, and identifies possible 

problems and areas to scrutinize.
83

4.2.2 Field Reconnaissance Survey

Site visits by the geotechnical engineer and other members of the design team are 

necessary to properly characterize the site. Field reconnaissance presents an

opportunity to locate and record many constructability issues such as site access, 

headroom restrictions, and other working conditions that may be otherwise 

overlooked from reviewing site or aerial photos and satellite images. This phase 

enables the engineer to substantiate the information gained from the office phase 

and to plan the detailed site exploration program.

The field reconnaissance for a structure foundation exploration includes, but is not 

limited to:

a. Inspection of nearby structures to determine their performance with the 

particular foundation type used.

b. Inspection of existing structure footings and stream banks for evidence 

of scour (for stream crossings) and movement. Large boulders in a 

stream are often an indication of obstructions which may be encountered 

in pile installations.

c. Recording of the location, type and depth of existing structures which 

may be affected by the new structure construction.

d. Visual examination of terrain for evidence of landslides.

e. Location of trees and other vegetation, as well as surface water.

f. Relating site conditions to proposed boring operations. This includes 

recording the locations of both overhead and below ground utilities, site 

access, private property restrictions, water depth and access points for 

marine borings, and other access restrictions or obstructions.

g. Recording of any feature or constraint which may impact the 

constructability of potential foundation systems.

Figure 4-1 contains an example of a field reconnaissance form modified from the 

AASHTO Foundation Investigation Manual (1978) for recording data pertinent to a 

site. At conclusion of the data collection and field reconnaissance phases, the 
84

Figure 4-1 Typical field reconnaissance form. 
85

engineer should have in mind geotechnical loading requirements, which would 

therefore dictate the planning of the field exploration program.

4.2.3 Detailed Field Exploration

The purpose of any field exploration program is to obtain representative information

on subsurface conditions, to recover disturbed and undisturbed soil samples, to 

perform in-situ testing, and to determine groundwater levels. This information 

provides factual basis upon which all subsequent steps in the pile design and 

construction process are founded. Its quality and completeness are of paramount 

importance. Each step in the process directly or indirectly relies on this data. It is 

assumed that a preference to a driven pile foundation is given. Therefore, the field 

exploration plan will focus on items related to the design of driven pile foundations.

The first step in this phase is to prepare a preliminary boring, sampling, and in-situ 

testing plan. For major structures, pilot borings are usually performed at a few select 

locations during the preliminary planning stage. These pilot borings establish a 

preliminary subsurface profile and thus identify key soil strata for testing and 

analysis in subsequent design stage borings. Then, during the design stage of 

major structures, geophysical surveys and a two phase boring program are 

recommended. First, control borings are performed at key locations identified in the 

preliminary subsurface profile to determine what, if any, adjustments are appropriate 

in the design stage exploration program. Following analysis of the control boring 

data, verification borings are then performed to fill in the gaps in the design stage 

exploration program.

A preliminary structural design plan should also be established at this point such that 

borings are properly located with respect to structural or earthen elements. 

Geophysical surveys and in-situ testing are included in this category. Information on 

developing a site exploration program may be found in Sabatini et al. (2002).

4.2.3.1 Geophysical Surveys

Geophysical surveys aide in subsurface characterization over a wide area, and 

supplement borings and other invasive tests. Primarily, seismic and electrical 

methods are used to measure depth to groundwater and bedrock and to resolve 

intermediate soil strata thickness. These methods may identify and locate 

subsurface anomalies requiring further evaluation by soil borings or in-situ tests. 

Geophysical surveys may also locate voids, debris or buried objects as well as 

measure divisions in soil stiffness.
86

Seismic surface wave methods involve using an applied impact or simply ambient 

earthborn noise (traffic, wind, microtremors, etc.) to produce surface waves or 

vibrations that travel through the ground and reflect off changes in subsurface 

materials. These reflected waves are generally measured with geophones located 

on the surface. The collected data can be analyzed to compute soil stiffness, 

allowing the bedrock depth, groundwater depth, and other soil strata to be located. 

Additional details on seismic surface wave methods can be found in Wightman et al. 

(2003), Sirles (2006), Rosenblad and Li (2009), and Sirles et al. (2009).

Ground penetrating radar is a common geophysical option and is useful to detect 

near surface voids in the 20 to 30 feet depth range, buried objects, and to locate the 

water table. However, this technique decreases in effectiveness with depth, and is 

not effective in saturated clayey soils.

Electrical test methods employ electrical currents to measure resistively which may 

be correlated to locate soil strata changes, as well as to detect voids or buried 

objects. DC resistivity testing does produce soil resistivity results below the 

groundwater table, as this medium conducts electrical current, Lucius et al. (2007). 

Results from this test can be modeled to show coarse versus fine grained material, 

as well as the groundwater table location, and bedrock.

It should be noted that geophysical surveys are often performed in the preliminary 

site exploration phase to provide a larger sense of site geology, and should further 

model soil stratigraphy. They should and always be performed in combination with 

direct exploration methods such as soil borings or in-situ testing.

4.2.3.2 Depth, Spacing, and Frequency of Boring & In-Situ Tests

The cost of a subsurface exploration program is comparatively small in relation to 

the foundation cost. For example, the cost of one 2.4 inch diameter boring is less 

than the cost of one 12 inch diameter pile. However, in the absence of adequate 

boring data, the design engineer must rely on extremely conservative designs with 

high safety margins. At the same time, the designer assumes enormous risk and 

uncertainty during the project's construction.

The number of borings required, their spacing, and sampling intervals depend on the 

uniformity of soil strata and loading conditions. Erratic subsurface conditions require 

closely spaced borings while more uniform soil profiles may require less frequent 

exploration. Structures sensitive to settlements or subjected to heavy loads require 

more detailed subsurface knowledge, and in this case, borings should be closely 
87

spaced. Minimum guidelines for depth and quantity of borings for given structures 

are provided by Sabatini et al. (2002) and AASHTO (2014). For deep foundations, 

Table 4-3 provides minimum quantity guidelines.

Table 4-3 Minimum Number of Exploration Points per Substructure 

(modified from Sabatini et al. 2002)

Pier or Abutment Width, W Exploration Points per 

Substructure1

W ≤ 100-ft 1 

W > 100-ft 2 

1

Additional exploration points should be provided if erratic subsurface conditions are 

encountered.

Minimum exploration depths should be met as well. Exploration points should 

extend a minimum of 20 feet below the expected pile toe elevation, or at least two 

times the pile group width. If piles will bear on rock, at least 10 feet of rock coring 

should be performed. Although these minimum values are provided by AASHTO 

(2014), additional boring and in-situ testing depths may be required due to site 

specific geology. Adjustments to the exploration program may be required as 

information becomes available, and likewise, additional rock coring length may be 

required if highly variable bedrock is present. Where unsuitable soil strata such as 

peat, highly organic materials, uncontrolled fills, soft fine grained soils, and loose 

coarse-grained soils are encountered, structure borings should extend through these 

deposits to reach and characterize suitable foundation support materials.

In-situ test spacing and frequency should be used to determine geomaterial design 

parameters based on project requirements. These tests generally supplement other 

aspects of the site exploration and will be performed after preliminary borings have 

been performed. In-situ tests are often done where undisturbed sampling is not 

easily performed. This includes both vane shear and pressuremeter testing among 

others. For in-situ testing and soil borings, many times a standard sampling interval 

is used. However, a customized soil boring and in-situ testing program may be 

formulated if general site geology or loading conditions warrant alternative intervals. 

This could aid with locating the contact between two soil strata, as well as 

discovering a thin layer of unsuitable material. Liquefaction and aggressive 

environment studies may also dictate appropriate sampling intervals.
88

4.2.3.3 Soil Boring Methods

Soil borings are the most frequently used soil exploration technique for projects in 

the public and private sector. Soil borings offer the ability to collect disturbed and 

undisturbed samples from various depths, as well as to perform in-situ tests. 

Common boring methods include augered borings, wash borings, and rotary drilling 

in soil. These methods, as well as sonic drilling, are presented in Table 4-4. 

Continuous flight augers are either solid stem or hollow stem and are both used to 

bore into the subsurface and allow for sampling and/or in-situ testing. A solid stem 

auger is essentially a solid rod with flights, and must be removed in order to test 

native soil. They are therefore not useful in soft soils, sands or areas of high 

groundwater due to the probability of borehole collapse and are best used when soil 

sampling only at relatively shallow depths is required. Hollow stem augers act as a 

casing that stays in place during drilling, whereby subsequent sampling and testing 

is performed through the bottom of the hollow stem. This auger type is therefore 

practical for a variety of subsurface conditions and utilizes a plug when advancing 

the auger. Hollow stem auger diameters range from 2.25 inch I.D. up to 12.25 inch 

I.D with larger sizes reserved for more complex sampling and testing techniques.

Rotary drilling may be performed in soil or rock and involves inserting a drill bit to cut 

and grind the material. Water or drilling mud flushes out the cuttings, and provides 

borehole stability. Air has also been used to force out cuttings in lieu of drilling mud, 

however borehole stability issues remain. To sample soil or perform an in situ test, 

the drill bit is removed and a sampling device is then inserted to collect material. 

Wash borings are advanced by the chopping action of a light bit in combination with 

the jetting action of water or drill fluid coming through the bit. Casing may be used to 

maintain an open bore hole, although typically a bentonite or drilling mud of similar 

properties is adequate. Drilling mud can contaminate recovered soil samples as well 

as add difficulty when trying to classify soil stratigraphy from wash cuttings. In 

addition, heavier particles such as gravel or cobbles may be left at the bottom of the 

hole if the wash system is undersized. Wash borings are infrequently used in the 

United States. 

4.2.3.4 Soil Sampling Methods

Soil samples may be collected via disturbed or undisturbed methods. Disturbed soil 

samples contain representative material that may be used for visual classification 

and more routine laboratory testing which is further described in Chapter 5. These
89

Table 4-4 Soil Boring Methods 

Method Depth Advantages Disadvantages Remarks

Auger 

Borings. 

Most equipment 

can drill to 

depths of 100 to 

200 feet. 

Boring advanced 

without water or 

drilling mud.

Hollow stem auger 

acts as a casing.

Difficult to detect 

change in material.

Heavy equipment 

required.

Water level must be 

maintained in boring 

equal to or greater 

than existing water 

table to prevent 

sample disturbance.

Boring advanced by 

rotating and 

simultaneously 

pressing an auger 

into the ground 

either mechanically 

or hydraulically.

Rotary 

Drilling. 

Most equipment 

can drill to 

depths of 200 

feet or more. 

Suited for borings 4 to 

6 inches in diameter.

Most rapid method in 

most soils and rock.

Relatively uniform hole 

with little disturbance 

to the soil below the 

bottom of hole.

Experienced driller 

can detect changes 

based on rate of 

progress.

Drilling mud if used 

does not provide an 

indication of material 

change as the wash 

water does.

Use of drilling mud 

hampers the 

performance of 

permeability tests.

Borehole advanced 

by rapid rotation of 

drilling bit and 

removal of material 

by water or drilling 

mud.

Rock coring is 

performed by rotary 

drilling.

Liquefaction 

evaluation methods 

require mud rotary 

drilling methods.

Wash 

Boring. 

Most equipment 

can drill to 

depths of 100 

feet or more.

Borings of small and 

large diameter.

Equipment is relatively 

inexpensive.

Equipment is light.

Wash water provides 

an indication of 

change in materials.

Slow rate of 

progress.

Not suitable for 

materials containing 

stones and 

boulders.

Boring advanced by 

a combination of 

the chopping action 

of a light bit and 

jetting action of the 

water coming 

through the bit.

Sonic 

Coring. 

Up to 1000 feet. Generally a 100% 

recovery rate. 

Can collect samples in 

soils and bedrock.

Expensive for large 

amounts of dense 

rock sampling. 

Borehole advanced 

in 10 feet cased 

sections.

Can collect any soil 

type with minimal to 

no fluids.
90

soil samples are not suitable for strength or compressibility testing as the sampling 

disturbance alters their condition. When in-situ particle arrangements, water content 

and other properties must be preserved for laboratory testing, undisturbed samples 

are taken. These samples are collected in devices designed to minimize sample 

disturbance. However, even while using utmost care for removal and transport, no 

sample will be completely undisturbed.

Disturbed soil samples may be collected with a split barrel sampler, sonic cores, or 

through test pits. Split barrel samplers typically range from 1.5 to 2.5 inches in 

diameter and are 18 or 24 inches long. Although multiple sizes exist, the standard 

split spoon sampler used for the Standard Penetration Test (SPT) has an inner 

diameter of 1.375 inches and outer diameter of 2 inches (ASTM D1586). A 

photograph of a standard SPT sampler is presented in Figure 4-2. This test is 

further discussed in Section 5.1.1.

Relatively undisturbed soil samples may be collected with thin walled tubes (Shelby 

tubes), a piston sampler or other specialized means. Thin walled tubes are 

produced in various sizes, and are typically used to collect fine grained soils as 

illustrated in Figure 4-3. These tubes are pushed into the soil and removed after a 

brief swelling of the soil occurs. ASTM D1587 provides detailed guidance on this 

sampling technique.

Piston samplers were developed to prevent soil from entering the sampling tube 

before the sample depth and to reduce sample loss during tube extraction. They are 

basically a thin wall tube sampler with a piston, rod, and a modified sampler head. 

There are numerous types of piston samplers; free or semi-fixed piston samplers, 

fixed-piston samplers, and retractable piston samplers.

The Pitcher sampler is a core barrel sampler that may be used for sampling a broad 

range of materials including undisturbed samples of stiff to hard clays, soft rocks and 

cemented sands. This sampler consists of a rotating outer core barrel with an inner 

thin walled sampling tube. The sampling tube leads the core barrel when sampling 

soft soils and the core barrel leads the sampling tube when sampling hard materials. 

This makes the Pitcher sampler particularly attractive for sampling materials with 

alternating hard and soft layers.
91

Figure 4-2 Standard split spoon sampler.

Figure 4-3 Shelby tubes (after Mayne et al. 2002).

4.2.3.5 Rock Exploration Methods (Coring / Drilling)

Rock drilling and coring is typically performed at the end of a soil boring once 

bedrock is encountered. Special drilling bits are used to cut and grind through rock, 

so that sampling can be performed. Fluid or air is circulated to flush out cuttings, 
92

and an inner core barrel collects rock samples. The recovered rock core is classified 

by rock type, and the core recovery length and recovery percentage noted as a 

percentage of the core run. Table 4-5 summarizes several rock coring methods.

Conventional rotary coring is primarily used to collect rock samples. For this, a 

circular diamond or tungsten drill bit grinds rock so that a core is cut from the 

surrounding rock. To collect intact rock samples, the drill barrel typically has an 

inner and outer tube, of which the inner tube remains stationary during drilling. The 

inner tube is removable such that intact samples may be brought to the surface, 

while the outer tube acts as casing, and remains in place. However, several barrel 

types exist, which are discussed further in Sabatini et al. (2002).

Wire line rotary coring is similar to conventional rotary coring however; the inner 

barrel is connected to a wire retrieval line as opposed to a rod or tube. This retrieval 

method significantly speeds up sampling at deeper depths.

Table 4-5 Summary of Rock Coring Methods

Method Depth Type of 

Samples 

Taken

Advantages Disadvantages Remarks

Conventional 

Rotary Coring 

in Rock. 

Most 

equipment can 

drill to depths 

of 200 feet or 

more. 

Continuous 

Rock Cores. 

Helps 

differentiate 

between 

boulders and 

bedrock.

Can be slow 

and fairly 

expensive.

Several types 

of core barrels 

are used.

Wire Line 

Rotary Coring

in Rock.

Most 

equipment can 

drill to depths 

of 200 feet or 

more. 

Continuous 

Rock Cores. 

Improved 

recovery rate 

of fractured 

rock. Typically 

faster drilling 

rate when 

coring rock at 

deep depths. 

Fairly 

expensive.

Inner tube 

retrieved by 

cable

suspended 

lifting device.

Preferred 

method for 

depths greater 

than 80 feet. 

Rotary Coring 

of Swelling 

Clay or Soft 

Rock. 

Most 

equipment can 

drill to depths 

of 200 feet or 

more. 

Plastic tube 

encased 

sample. 

Inner plastic 

tube protects 

soil and soft 

rock.

More complex 

equipment. 

Smaller 

sample due to 

coring 

equipment. 

A specialized barrel may be used when sampling swelling soil and soft rock. This 

barrel contains a third inner liner which contains the swelling soil or soft rock and is 
93

more easily removed due to the 5% to 10% difference in liner and drill bit size (Acker 

1974). A smaller diameter sample results from this barrel design, as opposed to 

samples collected via rotary coring. However, this collection method may exhibit 

less soil disturbance than thin walled sampling for some soil types.

Rock cores should be stored in a structurally sound box designed for the size of core 

collected as illustrated in Figure 4-4. Cores should be set carefully to retain natural 

bedding and fractures, and handled such that accidental breaks do not occur after 

retrieval. Features should be noted including bedding, fractures, and weathered 

zones, and photographs should be taken, always using a visible reference scale 

such as a tape measure. Sometimes wetting the rock surface can increase color 

contrast for photos. Labels identifying the project, borehole, depth interval and 

number of core runs should be written on the core box, in addition to standard field 

notes.

Rock cores are also given a rock quality designation (RQD). The RQD index test 

allows for a description of rock mass quality and should be performed in the field 

upon core recovery. The RQD index test, described in ASTM D6032, is used to 

provide estimates of overall rock quality. Deere and Deer (1989) proposed the RQD 

to be equal to the sum of the length of sound core pieces, 4 inches or greater in 

length, divided by the total length of core run. Core lengths should be measured 

from the axis centerline, while the RQD is expressed as a percentage. RQD 

designations are shown in Table 4-6, with a further discussion of rock quality index 

parameters provided in Section 5.4.

Figure 4-4 Rock core samples. 

RQD values may be indicative of the pile penetration into the rock needed to satisfy 

resistance requirements when reviewed in combination with additional test results. 

However, the RQD value alone should not be used to correlate rock strength. In 
94

addition to the coring methods presented here, various drill bits exist as well as core 

barrel types, which influence ease of drilling, core size and recovery rate. These 

should be selected accordingly based on experience and recovery requirements. 

Further descriptions of these may be found in Acker (1974), ASCE (2001), and 

ASTM D2113. Core samples should be retained for the duration of the project and 

beyond if needed due to any construction claims.

Table 4-6 Rock Quality Designation

Rock Mass Description RQD

Excellent 90-100

Good 75-90

Fair 50-75

Poor 25-50

Very Poor < 25

4.2.3.6 Groundwater

Accurate ground water level information is needed for the estimation of soil 

densities, determination of effective soil pressures and for the preparation of 

effective stress diagrams. Water levels will also indicate the construction difficulties 

which may be encountered in excavations and the dewatering effort required.

In most structure foundation explorations, water levels should be monitored during 

drilling of the boring, upon completion of the boring, and 24 hours after the 

completion of boring. More than one week may be required to obtain representative 

water level readings in low permeability cohesive soils or in bore holes stabilized 

with some drilling muds. In these cases, an observation well or piezometer should 

be installed in a boring to allow long term ground water monitoring. An observation 

well is typically used to monitor changes in the water level in a select aquifer 

whereas a piezometer is used to monitor changes in the hydrostatic pressure in a 

confined aquifer or specific stratum.

An observation well is usually a slotted section of small diameter PVC pipe installed 

in a bore hole. The bottom section of the slotted PVC pipe is capped and solid PVC 

sections are used to extend the observation well from the top of the slotted PVC 

section to a height above grade. The annulus between the slotted section and the 

sides of the bore hole is backfilled with sand. Once the sand is above the slotted 

PVC section, a bentonite seal is placed in the annulus sealing off the soil stratum in 
95

which the water table fluctuations will be monitored. Grout or auger cuttings are 

used to backfill the remaining void, while a locking removable cover may be placed 

over the annulus top. The water level reading in the observation well will be the 

highest of the water table in any soil layer that the slotted section penetrates.

Piezometers are generally used to monitor hydrostatic pressure changes in a 

specific soil stratum. Piezometers may be either pneumatic or vibrating wire 

diaphragm devices and are installed in a sand pocket with a bentonite seal similar to 

an observation well. Single and multiple piezometers can be installed in a single 

bore hole using a cement-bentonite grout. Additional piezometer information is 

available in FHWA-NHI-01-031, Subsurface Investigations – Geotechnical Site 

Characterization (Mayne et al. 2002).

4.2.4 Information Required for Construction

The subsurface exploration and subsequent soil and rock testing program provides 

essential information for driven pile design and construction. Engineers and 

contractors use this information to select the most appropriate pile type for the 

loading requirements, size pile driving hammer equipment, determine if pile 

installation aids are needed to meet design objectives, prepare cost estimates and 

bid documents, establish construction control methods and generate construction 

schedules. To reduce the risk of unplanned cost and claims, a well-defined and 

executed site characterization program is vital.

As described in Chapter 6, a variety of driven pile types and sections exist and may 

derive their nominal resistance from end bearing, shaft resistance or both. The pile 

type is generally selected based on local practice, structural requirements and 

subsurface conditions, while size and length are contingent upon soil properties and 

bearing layers. Pile drivability and other construction considerations are affected by 

the overall site characterization including permeability, relative density, layering and 

soil response to pile driving.
96

REFERENCES

Acker, W.L. III, (1974). Basic Procedures for Soil Sampling and Core Drilling, Acker 

Drill Company, Inc., Scranton, PA, 246 p.

American Association of State Highway and Transportation Officials (AASHTO). 

(1978). Manual on Foundation Investigations Second Edition. AASHTO 

Highway Subcommittee on Bridges and Structures, Washington, D.C., 196 p.

American Association of State Highway and Transportation Officials (AASHTO). 

(2014). AASHTO LRFD Bridge Design Specifications, US Customary Units, 

Seventh Edition, with 2015 Interim Revisions. American Association of State 

Highway and Transportation Officials, Washington, D.C., 1960 p.

ASTM D1586-11. (2014). Standard Test Method for Standard Penetration Test 

(SPT) and Split-Barrel Sampling of Soils. Annual Book of ASTM Standards, 

Vol. 4.08, ASTM International, West Conshohocken, PA, 9 p.

ASTM D1587-12. (2014). Standard Practice for Thin-Walled Tube Sampling of Soils 

for Geotechnical Purposes. Annual Book of ASTM Standards, Vol. 4.08, 

ASTM International, West Conshohocken, PA, 4 p.

ASTM D2113-14. (2014). Standard Practice for Rock Core Drilling and Sampling of 

Rock for Site Investigation. Annual Book of ASTM Standards, Vol. 4.08, 

ASTM International, West Conshohocken, PA, 20 p.

ASTM D6032-08. (2014). Standard Test Method for Determining Rock Quality 

Designation (RQD) of Rock Core. Annual Book of ASTM Standards, Vol. 

4.09, ASTM International, West Conshohocken, PA, 5 p.

ASTM Vol 4.08. (2014). Soil and Rock I, Vol. 4.08, ASTM International, West 

Conshohocken, PA, 1826 p.

ASTM Vol 4.09. (2014). Soil and Rock II, Vol. 4.09, ASTM International, West 

Conshohocken, PA, 1754 p.
97

Brown, D. A., Turner, J.P. and Castelli R.J. (2010). Drilled Shafts: Construction 

Procedures and LRFD Design Methods, FHWA-NHI-10-016, Geotechnical 

Engineering Circular (GEC) No. 10. U.S. Dept. of Transportation, Federal 

Highway Administration, 970 p.

Cheney, R.S. and Chassie, R.G. (2000). Soils and Foundations Workshop 

Reference Manual. FHWA HI-00-045, U.S. Department of Transportation, 

National Highway Institute, Federal Highway Administration, Washington, 

D.C., 358 p. 

Federal Highway Administration (FHWA). (1996). Geotechnical Engineering 

Notebook DT-15. Differing Site Conditions. U.S. Dept. of Transportation, 

Federal Highway Administration, Washington, D.C., 36 p.

Louie, J, N. (2001). Faster, Better: Shear-wave Velocity to 100 Meters Depth from 

Refraction Microtremor Arrays: Bulletin of the Seismological Society of 

America, V. 91.02, Alexandria, VA, pp. 347-364.

Lucius, J.E., Langer W.H. and Ellefsen, K. (2007). An Introduction to Using Surface 

Geophysics to Characterize Sand and Gravel Deposits. U.S. Geological 

Survey Circular, Reston, VA, 1310, 33 p.

Mayne, P.W., Christopher, B., Berg, R., and DeJong, J. (2002). Subsurface 

Investigations (Geotechnical Site Characterization), FHWA NHI-01-031, U.S. 

Dept. of Transportation, National Highway Institute, Federal Highway 

Administration, Washington, D.C., 300 p.

Sabatini, P.J., Bachus, R.C., Mayne, P.W., Schneider, J.A., and Zettler, T.E. (2002). 

Evaluation of Soil and Rock Properties, FHWA-IF-02-034, Geotechnical 

Engineering Circular (GEC) No. 5, U.S. Dept. of Transportation, Federal 

Highway Administration, 385 p.

Samtani, N.C. and Nowatzki, E.A. (2006). Soils and Foundations: Reference 

Manual, Vol. 1, FHWA-NHI-06-088, U.S. Dept. of Transportation, National 

Highway Institute, Federal Highway Administration, Washington, D.C., 462 p. 

Sirles, P.C. (2006). Use of Geophysics for Transportation Projects, NCHRP 

Synthesis 357, Transportation Research Board of the National Academies, 

Washington, D.C., 109 p.
98

Sirles, P., Shawver, J.B., Pullammanappallil, S., and Batchko, Z. (2009). Mapping 

Top-of-Bedrock and Soft-Soil Zones Beneath High-Traffic Areas Using 2D 

REMI, Symposium on the Application of Geophysics to Engineering and 

Environmental Problems (SAGEEP 2009), Environmental and Engineering 

Geophysical Society, Denver, CO, (available on CD; online: 

http://www.eegs.org ). 

Rosenblad B.L., and Li J. (2009). Comparative study of Refraction Microtremor 

(ReMi) and Active Source Methods for Developing Low-Frequency Surface 

Wave Dispersion Curves. Journal of Environmental and Engineering 

Geophysics. 14 (3), pp. 101-113.

Wightman, W., Jalinoos, F., Sirles., and Hanna, K. (2003). Applications of 

Geophysical Methods to Related Highway Problems. U.S. Dept. of 

Transportation, Federal Highway Administration, 716 p.
99

CHAPTER 5

GEOMATERIAL DESIGN PARAMETERS AND GEOTECHNICAL REPORTS

Properties of cohesive soil, cohesionless soil, soft rock, and hard rock affect the 

design and construction of driven piles. Depending on subsurface conditions, 

additional consideration may be given to aggressive soils or areas effecting pile 

integrity. As discussed in chapter 4, in-situ tests are performed during subsurface 

exploration while samples may be collected for further evaluation in a soils 

engineering laboratory. Results of in-situ and laboratory tests define material 

properties that are often correlated to design parameters, if not directly measured. 

Primary design parameters include shear strength and deformation values. This 

chapter will focus on the determination of soil and rock parameters that influence the 

design and construction of driven piles. Table 5-1 provides a summary of field and 

laboratory tests used for geomaterial characterization and determination of design 

parameters.

A discussion of in-situ and laboratory testing to acquire design parameters is 

provided in subsequent sections. For a more in depth discussion on the collection of 

design parameters, please refer to AASHTO (2014), Bowles (1992), Mayne et al. 

(2002), Duncan and Wright (2005) among other sources. In addition, the latest 

version of GEC-5, Evaluation of Soil and Rock Properties, should be consulted for 

updates in practice. The version at the time of this manual revision is Sabatini et al. 

(2002).

5.1 IN-SITU SOIL TESTING

In-situ testing provides soil parameters for the design of structure foundations 

especially in conditions where standard drilling and sampling methods cannot be 

used to obtain high quality undisturbed samples. Therefore, representative strength 

test data is difficult to obtain on these soils in the laboratory. To overcome these 

difficulties, test methods have been developed to evaluate soil properties, 

particularly strength and compressibility, in-situ.
100

Table 5-1 Field and Laboratory Tests for Geomaterial Parameter Determination

Design Parameter or

Information Needed

Cohesionless

Soil

Cohesive

Soil

Rock

Subsurface Stratigraphy

Drilling and Sampling; 

SPT, CPTu, CPT; 

Geophysical methods

Drilling and Sampling; 

SPT, USS, CPTu, 

CPT, DMT; 

Geophysical methods

Drilling and Sampling;

Rock Core Logging

Groundwater Conditions Well / Piezometer Well / Piezometer Well / Piezometer

Classification USCS Group USCS Group Rock Type

Gradation Sieve Analysis Sieve Analysis

Hydrometer Analysis

N/A

Atterberg Limits N/A Liquid Limit

Plastic Limit

N/A

Moisture Content, w Wet and Oven-Dried 

Weights

Wet and Oven-Dried 

Weights

Lab

Unit Weight, γ SPT, DMT USS-Lab USS-Lab

Sensitivity N/A VST, USS-Lab N/A

RQD and GSI N/A N/A Rock Core Recovery

and Logging

Effective Stress Friction Angle, φ’ SPT, CPTu, CPT, 

DMT

CD or CIU triaxial

on USS

Correlate to GSI

Undrained Shear Strength, su N/A UU, CIU and UC on 

USS, VST, CPT N/A

Preconsolidation Stress, σp SPT, CPT Consolidation test,

DMT, CPTu, CPT

N/A

Elastic Modulus of Soil, Es

SPT, CPT, PMT, 

DMT; correlate with 

Index Properties

DMT, PMT, Triaxial 

Tests; correlate with 

Index Properties

N/A

Unconfined Compressive Strength, qu N/A N/A UC on Rock Core

Modulus of Intact Rock, Er N/A N/A Compression Test on 

Rock Core

Rock Mass Modulus, Em N/A N/A Correlate to GSI and 

qu or Er; PMT
101

Table Key:

CD – Consolidated Drained Triaxial Test (Section 5.2) 

CIU – Consolidated Undrained Triaxial test with pore pressure measurements (Section 5.2)

CPT – Cone Penetration Test (Section 5.1.2) 

CPTu – Cone Penetration Test with pore water pressure measurements (Section 5.1.2)

CU – Consolidated Undrained Triaxial Test (Section 5.2)

DMT – Dilatometer Test (Section 5.1)

GSI – Geological Strength Index 

PMT – Pressuremeter Test (Section 5.1)

RQD – Rock Quality Designation Test (Section 4.2)

SPT – Standard Penetration Test (Section 5.1.1)

UU – Unconsolidated, Undrained Triaxial Test (Section 5.2) 

UC – Unconfined Compression Test (Section 5.2)

USS – Undisturbed soil sample (Section 4.2)

USCS – Unified Soil Classification System

VST – Vane Shear Test (Section 5.1.3) 

In-situ test methods can be effective when used to supplement conventional 

exploration programs. In addition, these methods help identify key strata for further 

conventional sampling and laboratory tests. In the absence of an adequate program 

to determine subsurface design parameters, the engineer must use a more 

conservative approach based on less certain data. This situation also results in 

additional uncertainty and risk and is not recommended.

Primary in-situ tests that provide data for foundation design are the Standard 

Penetration Test (SPT), cone penetration test (CPT), the cone penetration test with 

pore pressure measurements (CPTu), the seismic cone penetration test with pore 

pressure measurements (SCPTu), and the vane shear (VST). Other lesser used insitu testing devices include the pressuremeter test (PMT) and flat dilatometer test 

(DMT). Each of these test methods apply different loading conditions (orientation 

and rate of strain) to gain information on strength and or/stiffness such as torsion for 

VST or axial loading for SPT. The net result is that all in-situ tests are index tests 

that require local correlations and where possible should be benchmarked to 

laboratory tests. In-situ test method schematics are depicted in Figure 5-1. 

The applicability, advantages and disadvantages of all the in-situ testing methods 

are briefly summarized in Table 5-2. For a detailed discussion of a particular in-situ 

testing method, the reader is referred to the publications listed at the end of this 

chapter. NHI course 132031, Subsurface Investigations – Geotechnical Site 

Characterization and the accompanying course manual by Mayne et al. (2002) 

provides a thorough coverage of in-situ testing methods.
102

Figure 5-1 Schematic of common of in-situ tests (after Mayne et al. 2001).

5.1.1 Standard Penetration Test

The most common in-situ test in current use is the Standard Penetration Test (SPT) 

(ASTM D1586). During borehole advancement, the standard split spoon is driven 

using a 140 lb hammer from a 30 inch drop height. The sampler tip starts at the 

current borehole bottom and is driven 18 or 24 inches depending on sampler length. 

The number of blows for each 6 inch drive interval is counted, with the second and 

third interval summed to establish an N value. If the sampler penetrates over the 

increment under the weight of the drill rods or after the SPT hammer is set atop the 

rods, a WOR (weight of rods) or WOH (weight of hammer) designation is recorded 

on the boring log. Furthermore, if very dense or hard material is encountered and 

blow counts exceed 50 blows before finishing a 6 inch sampling interval, the SPT 

event is ended and the N value is designated as 50/x where x is the actual sampler 

penetration in inches measured for the 50 blows. This blow count and partial 

penetration increment should be recorded in lieu of an N value on the respective 

boring logs.
103

Table 5-2 Summary of In-Situ Methods

In-Situ Test

Information Obtained

for Pile Foundation Design 

in Appropriates Soil Types

Advantages Disadvantages

Standard 

Penetration 

Test (SPT)

ASTM D1586

Collection of soil samples to 

confirm subsurface soil type. 

Correlations for determination 

of in-situ density, liquefaction 

susceptibility, and friction 

angle of sands, undrained 

shear strength of clays. Best 

suited for sand, silt, and clay 

materials. Not suitable in 

large gravel, rubble, and rock. 

1. Simple test.

2. Can retrieve samples to 

confirm soil type.

3. Equipment is widely 

available.

4. Correlations available for 

estimating soil parameters.

1. Operator and equipment 

dependent.

2. Samples are disturbed.

3. Cannot evaluate in-situ 

pore pressure.

4. Accuracy of estimated soil 

parameters.

Cone

Penetration

Test with Pore

Pressure

Measurements

(CPTu) 

ASTM D5778 

Continuous evaluation of 

subsurface stratigraphy. 

Correlations for determination 

of in-situ density and friction 

angle of sands, undrained 

shear strength of clays, and 

liquefaction susceptibility. 

Best suited for sand, silt, and 

clay materials. Not suitable in 

gravel, rubble, and rock.

1. Cone can be considered as 

a model pile.

2. Quick and simple test.

3. Can reduce number of 

borings.

4. Relatively operator 

independent.

5. Pore pressure 

measurements can be used 

to assess soil setup effects.

6. Can help determine if 

penetration is drained or 

undrained.

1. Does not provide soil 

samples.

2. Should be used in 

conjunction with soil borings 

in an exploration program.

3. Local correlations can be 

important in data 

interpretation.

4. Location and saturation of 

porous filter can influence 

pore pressure 

measurements.

Pressuremeter

Test (PMT)

ASTM D4719

Bearing capacity from limit 

pressure and compressibility 

from pressure meter 

deformation modulus. Best 

suited in sand, silt, clay and 

soft rock. Not suitable in 

organic soils and hard rock.

1. Tests can be performed in 

and below hard strata that 

may stop other in-situ 

testing devices.

2. Tests can be made on nonhomogenous soil deposits. 

1. Bore hole preparation very 

important.

2. Limited number of tests per 

day.

3. Limited application for 

axially loaded pile design.

Dilatometer

Test (DMT)

ASTM D6635

Correlations for soil type, 

earth pressure at rest, over 

consolidation ratio, undrained

shear strength, and 

dilatometer modulus. Best 

suited for low to medium 

strengths sand and clay. Not 

suitable in dense deposits, 

gravel, and rock.

1. Quick, inexpensive test.

2. Relatively operator 

independent.

1. Less familiar test method.

2. Intended for soils with 

particle 

sizes smaller than fine 

gravel.

3. Limited application for 

axially loaded pile design.

Vane Shear

Test

ASTM D2573

Undrained shear strength. 

Best suited in soft to medium 

clays. Not suitable in silt, 

sand or gravel.

1. Quick and economical.

2. Compares well with 

unconfined compression 

test results at shallow 

depths.

1. Can be used to depths of 

only 13 to 20 feet without 

casing bore hole.

Dynamic Cone 

Test

Qualitative evaluation of soil 

density. Qualitative 

comparison of stratigraphy.

Best suited in sand and 

gravel. Not suitable in clay.

1. Can be useful in soil 

conditions where static 

cone (CPT) reaches 

refusal. 

1. An unknown fraction of 

resistance is due to side 

friction.

2. Overall use is limited.
104

In addition to the penetration resistance that is recorded during driving, soil enters 

the hollow sampler, thus providing a disturbed soil sample. A schematic of the 

Standard Penetration Test is presented in Figure 5-2. A description of the SPT 

sampler was previously presented in Chapter 4 along with a photograph of the 

sampler in Figure 4-2. 

The SPT hammer type and operational characteristics have a significant influence 

on the resulting SPT N values. There are two main hammer types currently in use in 

the U.S., the safety hammer and the automatic hammer. A third hammer type, the 

donut hammer, was used almost exclusively prior to about 1970 but not any longer 

due to safety considerations. Figure 5-3 provides illustrations of these SPT hammer 

types.

Figure 5-2 Standard Penetration Test schematic (after Mayne et al. 2001).
105

Figure 5-3 Standard Penetration Test hammer types.

Measurement studies on SPT energy transfer have been performed (e.g. Kovacs et 

al. 1983; Honeycutt et al. 2014). In general, these studies have indicated that the 

typical energy transfer from donut, safety, and automatic hammers are on the order 

of 45%, 60%, and 80% of the SPT test potential energy, respectively. It should not 

be assumed that all SPT hammers of a given type will have the energy transfer 

values noted above. Energy transfer for a given hammer type can and does vary 

according to hammer maintenance, hammer manufacturer, driller, and operating 

procedures. Because of these variations, it is recommended that SPT hammers 

undergo a yearly calibration in accordance with ASTM D4633 to document hammer 

performance. It may be particularly advantageous to conduct these calibrations prior 

to undertaking major projects. A photograph of energy transfer measurements being 

taken during a SPT sampling event is provided in Figure 5-4. 

The use of reliable qualified drillers and adherence to recommended sampling 

practice cannot be overemphasized. Procurement practices should consider the 

drilling quality, sampling, and testing requirements needed for economical driven pile 

foundation design and construction.
106

Figure 5-4 Instrumented 2 foot long AW rod atop drill string for SPT hammer 

energy measurements.

The significance of the SPT hammer type and energy transfer on N values is very 

apparent in a pile capacity prediction symposium reported by Finno (1989). For this 

event, two soil borings were drilled less than 33 feet apart in a uniform sand soil 

profile. Figure 5-5 presents the results of the two borings, one with SPT N values 

obtained using a safety hammer and the other with an automatic hammer. The SPT 

N values from the safety hammer range from 1.9 to 2.7 times the comparable N 

value from the automatic hammer. This significant variation in N values clearly 

indicates that the type of SPT hammer used should be recorded on all drilling logs. 

It is recommended that SPT N values be corrected and reported as N60 values 

whenever possible. Cheney and Chassie (2000) identified ten common errors that 

influence SPT test results which should also be reviewed by designers and boring 

crews. 
107

Figure 5-5 Adjacent borings with different SPT hammer types 

(after Finno 1989).

Although sources of error exist in the SPT test, correction and/or normalization 

factors have been developed to aid designers. The pile design charts and methods 

provided in this manual are the current standard of practice in the United States and 

use SPT N values based on safety hammer correlations, i.e., 60% energy transfer. 

Therefore SPT N values established on the basis of 60% energy transfer are 

referred to as N60. Conversion of field SPT N values to N60 values based on energy 

transfer measurements are as follows:
108





 



 = 60 60

ER NN Eq. 5-1 

Where:

N60 = SPT N value corrected for 60% energy transfer.

N = uncorrected field SPT resistance value. 

ER = hammer efficiency as determined by energy measurements in 

accordance with ASTM D4633.

Typical SPT hammer efficiencies may also be found in geotechnical literature such 

as FHWA GEC-5, Sabatini et al. (2002) as well as Das (2007). The energycorrected N60 value may be normalized for the effects of overburden stress, 

designated (N1)60, before being used in correlations between N values and soil 

properties. The general conversion is shown in Equation 5-2. 

( )1 60 = nNCN 60 Eq. 5-2 

Where:

(N1)60 = SPT N value corrected for energy and overburden stress.

N60 = SPT N value corrected for 60% energy transfer.

Cn = correction factor for SPT N value. 

AASHTO specifications recommend SPT N values be corrected for overburden 

pressure using Equation 5-3 unless otherwise specified by the design method. 

Relationships for this factor have been published in the literature whereas Equation 

5-3 is the recommended correction in AASHTO specifications from Peck et al. 

(1974). An alternate correction approach from Lio and Whitman (1986) is presented 

in Equation 5-4.

0.2 40 log77.0 '  ≤



 



 =

vo

Cn σ

Eq. 5-3 

n

vo

a

n

p C 













 = ' σ

Eq. 5-4 

Where:

pa = atmospheric pressure (ksf).

σ'vo = vertical effective stress at the sample depth (ksf).

n = exponent typically equal to 1 in clays (Olsen 1997) and 0.5 in sandy 

soils (Lio and Whitman 1986). 
109

Corrected N values are used for soil strength parameter correlations presented in 

Section 5.2. Note that correlations between cohesive soil and physical properties 

with N values are crude and, therefore, correction of N values in cohesive soils is 

generally not necessary.

5.1.2 Cone Penetration Test

The cone penetration test (CPT) was first introduced in the U.S. in 1965. By the mid 

1970's, the electronic cone began to replace the mechanical cone, and in the early 

1980's, the piezocone or cone penetration test with pore pressure measurements 

(CPTu) became readily available. A NCHRP synthesis by Mayne (2007) provides a 

good summary of cone penetration testing practices and use by transportation 

agencies in the US and Canada. The CPT/CPTu has developed into one of the 

most popular in-situ testing devices. Part of this popularity is due to the CPT's ability 

to provide large quantities of useful data quickly and at an economical cost. 

Depending upon equipment capability as well as soil conditions, 300 to 1200 feet of 

penetration testing may be completed in one day. 

Current cone penetration testing relies on the use of electronic cone penetrometers 

with pore pressure measurements. In the cone penetration test, a cone 

penetrometer with a cross sectional area of 1.55 or 2.33 inch2 and a 60° conical tip is 

attached to a series of rods and is continuously pushed into the ground. The cone 

also contains a friction sleeve located behind the conical tip. Pore pressure 

measurements are obtained with all modern cones from a pore pressure transducer 

located behind the conical tip and just before the friction sleeve. Pore pressure 

measurements at this location are referred to as the u2 position.

Typically, a hydraulic ram with 10 to 40 kips of thrust is used to continuously 

advance the cone into the ground at a rate of 0.8 in/sec. Built in load cells are used 

to continuously measure the cone tip resistance, qc, the unit sleeve friction 

resistance, fs, and the pore pressure, u, during penetration. The pore pressure 

measurement is used to correct the measured cone tip resistance, qc, for 

geometrical effects that reduce the measured value in proportion to the amount of 

pore pressure that is generated. The resulting corrected tip resistance is referred to 

as qt. The friction ratio, Rf, is the defined as fs/qt and is commonly used in the 

interpretation of test results. Careful porous element and cavity saturation is 

essential to obtain pore pressure measurements. Test procedures may be found in 

ASTM D5778. Both size cones are displayed in Figure 5-6 with a CPT rig shown in 

Figure 5-7. 
110

Figure 5-6 Cone penetrometers (after Mayne et al. 2001). 

Figure 5-7 CPT rig. 
111

Modern cones may be further divided into three main types, the piezocone (CPTu), 

the seismic piezocone (SCPTu), and the resistivity piezocone (RCPTu). The CPTu 

can also be used to monitor pore pressure dissipation. The seismic cone includes a 

horizontal geophone above the friction sleeve which allows calculation of shear 

wave velocity at selected depths. The resistivity cone uses electrical conductivity or 

resistivity which allows detection of the freshwater – salt water interface or profiling 

of contaminated groundwater plumes. Since pore pressure measurements are 

made with all modern electronic cone penetrometers, the abbreviation CPT will be 

used hereafter and assume that pore pressure measurements are included.

Data collected with CPT can provide a continuous profile of the subsurface 

stratigraphy. A soil behavior chart developed by Robertson (1990) for CPT data is 

presented in Figure 5-8. From correlations with CPT data, evaluations of in-situ 

relative density, Dr, and effective friction angle, φ’, of cohesionless soils as well as 

the undrained shear strength, su, of cohesive soils can be made. Correlations for 

determination of other soil properties, liquefaction susceptibility, and estimates of 

SPT N60 values may also be determined. The accuracy of these correlations may 

vary depending upon geologic conditions. Correlation confirmation with local 

conditions is therefore important.

The primary advantage of CPT testing is the ability to rapidly develop a continuous 

profile of subsurface conditions more economically than other subsurface 

exploration or in-situ testing tools. Because the CPT collects continuous data, it can 

delineate fine changes in stratigraphy and characterize site variability better than 

other methods. Engineering properties can be assessed through empirical 

correlations. For cohesionless soils, these empirical engineering property 

correlations are accurate and commonly used. For cohesive soils, engineering 

property correlations are less accurate. CPT results are relatively operator 

independent and highly repeatable. Because of these advantages, the CPT is 

increasingly being used as an equal or superior subsurface exploration method in 

some stratigraphic conditions such as alluvial sites and liquefiable soils. In other 

subsurface conditions, CPT may be used to reduce the number of conventional 

borings needed on a project, or to focus attention on discrete zones for detailed soil 

sampling and testing.

Limitations of CPT testing include the inability to push the cone in dense or coarse 

soil deposits. In general, soils with SPT N values greater than about 30 to 40 are 

difficult to test with CPT equipment. To penetrate dense layers, cones are 

sometimes pushed in bore holes advanced through the dense strata. While CPT 

can be used in marine environments, the soundings must be performed from a fixed
112

Figure 5-8 Soil behavior type classification chart based on normalized CPT data 

(after Robertson 1990). 

platform (e.g., a jack-up barge) with sufficient casing to provide lateral support for 

the rods or specialized submersible equipment is will be required. Another limitation 

is that soil samples are generally not obtained during routine CPT programs for 

confirmation of stratigraphy. Local correlations are also important in data 

interpretation.

5.1.3 Vane Shear Test

The vane shear test is an in-situ test for determining the undrained shear strength of 

soft to medium clays. Figure 5-9 is a schematic drawing of the essential 

components and test procedure. The test consists of forcing a four bladed vane into 

undisturbed soil and rotating it until the soil shears. Two shear strengths are usually 
113

recorded, the peak shearing strength and the remolded shearing strength. These 

measurements are used to determine the sensitivity of clay and allows for analysis 

of the soil resistance to be overcome during pile driving. It is necessary to measure 

skin friction along the steel connector rods which must be subtracted to determine 

the actual shear strength. The vane shear test generally provides the most accurate 

undrained shear strength values for clays when undrained shear strengths are less 

than 1 ksf. A detailed test procedure may be found ASTM D2573.

It should be noted that the sensitivity of clay determined from a vane shear test 

provides insight into the setup potential of the clay deposit. Soil setup may then be 

used for pile drivability and resistance evaluations. However, the sensitivity value is 

a qualitative and not a quantitative indicator of soil setup.

Figure 5-9 Vane shear test schematic (after Mayne et al. 2001). 

5.1.4 Other In-Situ Tests

Other in-situ tests with limited application to pile foundation design include the 

dilatometer, the pressuremeter, and the dynamic cone. These are briefly discussed 

in the following sections.
114

5.1.4.1 Dilatometer Test

The dilatometer test (DMT) is an in-situ testing device that was developed in Italy in 

the early 1970's and first introduced in the U.S. in 1979. Like the CPT, the DMT is 

generally hydraulically pushed into the ground although it may also be driven. When 

the DMT can be pushed into the ground with tests conducted at 8 inch increments, 

100 to 300 feet of DMT sounding may be completed in a day. The primary utilization 

of the DMT in pile foundation design is the delineation of subsurface stratigraphy 

and interpreted soil properties. However, it would appear that the CPT/CPTu is 

generally better suited to this task than the DMT. The DMT may be a potentially 

useful test for design of piles subjected to lateral loads. For axially loaded piles, the 

dilatometer test has limited direct value. The test procedures for DMT are presented 

in ASTM D6635.

5.1.4.2 Pressuremeter Tests

The pressuremeter test (PMT) is an in-situ device used to evaluate soil and rock 

properties, and imparts lateral pressures to the soil, allowing for soil shear strength 

and compressibility to be determined by interpretation of a pressure-volume 

relationship. Deposits such as soft clays, fissured clays, sands, gravels and soft 

rock can be tested with a pressuremeter. A pressuremeter test produces 

information on the elastic modulus of the soil as well as the at rest horizontal earth 

pressure, the creep pressure, and the soil limit pressure. For piles subjected to 

lateral loads, the pressuremeter test is a useful design tool and can be used for 

determination of p-y curves. However for design of vertically loaded piles, the 

pressuremeter test has limited value. Pile design procedures using pressuremeter 

data have been developed and may be found in FHWA-IP-89-008, The 

Pressuremeter Test for Highway Applications, Briaud (1989). Details on test 

procedures may be found in ASTM D4719, Standard Test Method for Pressuremeter 

Testing in Soils.

5.1.4.3 Dynamic Cone

There are two types of dynamic penetrometers with conical points. The dynamic 

cone type that is most often used has a shaft diameter that is smaller than the cone 

diameter. Theoretically, due to the cone being larger than the shaft, the 

penetrometer measures only point resistance. A lesser used cone type has a shaft 

and cone of the same diameter. This type of dynamic cone penetrometer records 

both skin friction and point resistance, but the two components cannot be analyzed 

independently. Equations have been developed for determining the geotechnical 
115

resistance of pile foundations by using the dynamic cone test data. However, these 

are not used extensively and the dynamic cone penetrometer is not recommended 

for final foundation design unless specific local correlations with load tests to 

geotechnical failure have been taken.

5.2 SOIL PARAMETERS

Index properties and engineering properties for geomaterials are derived from field 

and laboratory testing. Differing soil and rock types may be segregated through 

index tests which aid in the division of subsurface strata for design. These 

properties are typically obtainable through routine sampling and testing methods, 

and assist with constructability limitations or preliminary pile selection based on soil 

type and loading conditions. Engineering properties for soil and rock include shear 

strength, compressibility and permeability. Shear strength values can be used to 

determine driven pile geotechnical resistances while compressibility information can 

be used to check service limit states. Permeability is not generally quantified for pile 

design purposes, however certain slow draining soils can influence pile drivability, 

and therefore knowledge of these areas may assist in constructability concerns. 

Further information on laboratory testing of soils may be found in Bowles (1992), as 

well is in GEC-5 by Sabatini et al. (2002) and Samtani and Nowatzki (2006) among 

other sources.

5.2.1 Soil Classification and Index Properties 

Two soil classification systems are in common use for transportation projects. The 

AASHTO soil classification system is used by many transportation agencies for 

pavement features while the Unified Soil Classification System (USCS) is utilized for 

nearly all other geotechnical work. The USCS is better suited to deep foundation 

design and will therefore be solely used in this manual.

Routine index tests for classification include a particle size distribution, Atterberg 

limits and moisture content. Additional index tests include specific gravity and soil 

unit weight. A summary of these index tests is shown in Table 5-3. 

Laboratory tests to determine the fine content and gradation of cohesionless soils 

and the remolded shear strength of cohesive soils are important in assessing pile 

drivability and potential soil setup effects (changes in nominal resistance with time).
116

Table 5-3 Soil Index Tests used for Driven Pile Design 

(modified from Brown et al. 2010)

Test Index Property 

Determined Soil Type Application Standard Test Procedure

Particle Size 

Distribution

(Mechanical and 

Hydrometer)

Grain Size 

Distribution

Sieve on all Soils; 

Hydrometer on 

Minus #200

USCS 

classification 

Aids in evaluation 

of densification 

and liquefaction 

potential for 

cohesionless soils

ASTM D422

AASHTO T88

Atterberg Limits Liquid Limit (LL)

Plastic Limit (PL)

Plasticity index (PI)

PI = LL - PL

Liquidity Index (LI)

LI = (wn - PL) / PI

Minus #40 USCS 

classification

Soil consistency

Engineering 

property 

correlations

ASTM D4318

AASHTO T89

AASHTO T90

Moisture 

Content

Moisture Content, 

wn

Undisturbed 

Samples; SplitSpoon Samples 

with Moisture 

Change

Soil Consistency

May aid in 

determination of 

water table 

elevation

ASTM D2216

AASHTO T265

Unit Weight

(Undisturbed)

Total Unit Weight, γ

or Density, ρ

Fine-grained

(Cohesive)

Aids evaluation of 

underground soil 

stresses

ASTM D7263

AASHTO T38

Specific Gravity Specific Gravity, Gs Minus #4 Aids evaluation of 

soil density and 

compressibility

ASTM D854

AASHTO T100

Soil 

Classification

USCS Group Name 

and Group Symbol;

All Soils Physical 

characteristics 

define soil types

ASTM D2487

The relative density, gradation, and fine content of cohesionless soils provide useful 

information in assessing pile drivability. Soils with a high fine content generally have 

lower angles of internal friction than lower fine content soils of similar density. A 

high fine content can also affect soil drainage and pore pressures during shear, and 

thus, the effective stresses acting on a pile during driving. Depending upon soil 

density, cohesionless soils with high fine contents are also more likely to 
117

demonstrate soil setup than cohesionless soils with little or no fines. The gradation 

and angularity of the soil grains also influences the angle of internal friction. 

Gradation information is also useful in assessing the densification and liquefaction 

potential of cohesionless soils.

Routine laboratory grain size analyses (mechanical and hydrometer) can quantify 

gradation and fine content. With this information, improved assessments of pile 

drivability and soil setup potential in cohesionless soils can be made. Cohesive soils 

may lose a significant portion of their shear strength when disturbed or remolded, as 

during the pile driving process. The ability to estimate the soil strength at the time of 

driving and the resulting strength gain with time or soil setup is a key component of 

economical pile design in cohesive soils. Soil setup is discussed further in Chapter 

7. The sensitivity of a cohesive soil can provide a qualitative but not quantitative 

indication of potential soil setup. Sensitivity determined in-situ with a vane shear 

device provides the best assessment of cohesive soil sensitivity. However, the 

sensitivity of a cohesive soil can also be determined from laboratory tests on 

undisturbed and remolded samples to determine respective shear strength values.

The sensitivity of a cohesive soil, St, is defined as:

 

Eq. 5-5

Table 5-4 contains typical values of sensitivity as reported by Sowers (1979) which 

may be useful for preliminary estimates of remolded shear strength. Terzaghi and 

Peck (1967) noted that clays with sensitivities less than 16 generally regain a portion 

to all of their original shear strength with elapsed time. Based upon typical 

sensitivity values reported by Terzaghi et al. (1996) as well as by Sowers (1979), the 

remolded shear strength of many cohesive soils during pile driving would be 

expected to range from about 1/3 to 1/2 the undisturbed shear strength.

Table 5-4 Typical Values of Sensitivity (after Sowers 1979)

Soil Type Sensitivity

Clay of medium plasticity, normally consolidated 2-8 

Highly flocculent, marine clays 10-80

Clays of low to medium plasticity, overconsolidated 1-4 

Fissured clays, clays with sand seams 0.5-2 
118

Additional information on index testing of soil may be found in GEC-5 by Sabatini et 

al. (2002) as well as in Samtani and Nowatzki (2006). Section 5.4 provides 

additional insight and discussion on drivability considerations.

5.2.2 In-Situ Stress

For the design of driven piles, the in-situ state of stress at a given depth is needed to 

calculate geotechnical resistance. A basic assessment of stress calculation in soil is 

provided in this section. Further detail is provided in the literature including Sabatini 

et al. (2002), Samtani and Nowatzki (2006), and Das (2007) among other sources. 

The reader is directed to these resources for in depth coverage of this topic.

For geotechnical analyses, the effective soil stress is utilized for design. The 

effective stress, σ’, is the difference of total stress, σ, and pore water pressure, u, 

and may be expressed as Equation 5-6.

' σσ −= u Eq. 5-6 

The total stress at a given depth is generally the product of total soil unit weight for 

each soil strata, γi, and strata thickness, hi. Often multiple soils with ranges in unit 

weight exist at a given location. Therefore a summation of these layers accounts for 

the total stress calculation with depth as illustrated in Equation 5-7. Pore water 

pressure is a function of the unit weight of water, γw, and the pressure head, hw, as 

shown in Equation 5-8. 

= ∑ ( ) n

i ii γσ h Eq. 5-7 

ww = γ hu Eq. 5-8 

The vertical effective stress at a given depth may be calculated as follows:

( ) ww

n

i '

vo = ∑ ii − γγσ hh Eq. 5-9 

Above the water table, free draining, coarse grained soils may be assumed dry. 

More likely however, this soil is partially saturated, particularly in the vadose zone 

where negative pore water pressure results from capillary effects. Accurate soil unit 

weights should be therefore be reflected in this effective stress calculation. The 

lateral resistance for deep foundations is affected by the horizontal effective stress. 

Determination of the effective horizontal stress at a given depth, σ’ho, condition may 
119

be evaluated by relating the vertical effective stress and the at rest lateral earth 

pressure coefficient, Ko, using Equation 5-10.

vo

ho Ko '

'

σ

σ

= Eq. 5-10

Selection of effective stress or total stress usage for design is contingent upon the 

structural loading conditions and soil properties. For piles driven in free draining 

clean sands, an analysis would generally include the use of effective stress 

parameters since it may be assumed pore water pressure dissipates almost 

immediately after loading the soil. Conversely for clays or other slow draining soils, 

an increase in pore water pressure will be apparent immediately after loading. A 

total stress analysis may therefore be practical for this condition. The responsible 

engineer should exercise clear judgment involving loading conditions and the 

selection of design parameters.

5.2.3 Shear Strength

Shear strength is an internal resistance per unit area that the geomaterial can 

provide to resist failure along a plane. Compositional, environmental and in-situ 

factors combine to define the overall shear strength of a soil, which contributes to 

the geotechnical strength limit state for design. Geomaterial shear strength is also 

influenced by the manner (static, dynamic, cyclic) and rate of loading.

GEC-5 by Sabatini et al. (2002) notes several issues relevant to shear strength 

evaluations for driven pile designs. These include variation of the soil shear strength 

between the time the pile is driven and when load application occurs. Time

dependent strength increases, referred to as soil setup, are often observed in 

saturated, normally consolidated to moderately over consolidated clays and finegrained material. Conversely, a decrease in strength with time, referred to as 

relaxation, is often observed for heavily over consolidated clays, dense silts, dense 

fine sands, and weak laminated rock. Therefore, shear strength evaluations should 

consider both of these long and short term conditions.

Shear strength may also vary due to changes in site conditions that affect the in-situ 

effective stress state. Site conditions that may increase or decrease soil shear 

strength values include site dewatering, surface loading, and excavations. Granular 

soil strengths may also change over time due to densification from driving. This 

potential shear strength increase should be evaluated relative to pile drivability and 

constructability considerations.
120

The Mohr-Coulomb failure criterion will be the only model discussed herein and is 

shown in Equation 5-11. 

c += tanφστ Eq. 5-11

Where: 

τ = shear stress at failure (shear strength).

c = cohesion.

σ = total normal stress (pressure) on plane of failure. 

φ = angle of internal friction.

Equation 5-11 is used for total stress analysis and uses an undrained friction angle 

and cohesion. For cases where effective stress or drained parameters are 

necessary, this equation is modified to Equation 5-12. 

c += '' tanφστ ' Eq. 5-12

Where:

τ shear stress at failure (shear strength).

c’ = effective cohesion. 

σ' = effective normal stress (pressure) on plane of failure.

 φ' = effective stress angle of internal friction.

In this drained situation, typically cohesionless soils exist, therefore c’ is assumed 

equal to zero. However upon careful evaluation, c’ may be included to account for 

certain conditions including cemented soils, partially saturated soils and heavily over 

consolidated clays. Determination of c’ should be carefully selected on interpretation 

of the test data due to its strong influence on pile design results.

For short-term, undrained loading of cohesive soils, Equation 5-11 can be simplified 

to a total stress approach. Typically, cohesive soils exhibit a friction angle, φ, equal 

to zero, therefore Equation 5-13 reflects this behavior.

u τ == sc Eq. 5-13

Where: 

τ = shear stress at failure (shear strength).

 c = cohesion.

su = undrained shear strength. 
121

In addition to the above mentioned case, rapid loading of slow draining (low 

permeability) soil generates excess pore water pressure that is not easily quantified. 

Moreover, to simplify the shear stress calculation, the cohesion determined via 

testing may be assumed equal to the design shear strength while in-situ states of 

stress, water content, loading rate, and other variables are unknown.

5.2.3.1 Laboratory Tests for Soil Shear Strength

Three test methods are commonly used to measure shear strength in the laboratory. 

A brief description of each test method is provided in the subsequent sections. 

Further details on the test procedure and analysis may be found in GEC-5 by 

Sabatini et al. (2002), as well as the respective ASTM designations. The three test 

methods are as follows:

1. Direct shear test (AASHTO T236 / ASTM D3080).

2. Unconfined compression test (AASHTO T208 / ASTM D2166).

3. Triaxial compression test (AASHTO T234 / ASTM D2850).

5.2.3.1.1 Direct Shear

The direct shear test is performed by placing a sample of soil into a shear box which 

is split into two parts at mid height. A normal load is then applied to the top of the 

sample and one half of the shear box is pulled or pushed horizontally past the other 

half. The shear stress is calculated from the horizontal force divided by the sample 

area and is plotted versus horizontal deformation. A plot of at least three normal 

stresses and their corresponding maximum shear stresses provides the shear 

strength parameters c and φ. Bowles (1977) notes that the φ values determined 

from plane strain direct shear tests are approximately 1.1 times the φ values 

determined from triaxial tests. Direct shear tests are primarily performed on 

recompacted granular soils, which may not represent their in-situ conditions, and are 

generally not recommended for cohesive soils due to limitations on drainage control 

during shear.

5.2.3.1.2 Unconfined Compression

The unconfined compression test is a widely used laboratory test to evaluate the 

shear strength of cohesive soil and rock. In the unconfined compression test, an 

axial load is applied on a cylindrical sample while maintaining a zero lateral or 

confining pressure. Axial loading is increased to failure and the shear strength is 
122

then considered to be one half the axial stress at failure. Unconfined compression 

tests are performed only on cohesive soil and rock samples.

Unconfined compression tests on cohesive soil samples recovered from deeper 

depths or samples with a secondary structure, such as sand seams, fissures, or 

slickensides, can give misleadingly low shear strengths. This is due to the removal 

of the in-situ confining stress normally present. Triaxial compression tests provide 

improved information on soil shear strength in these cases.

5.2.3.1.3 Triaxial Compression Test

The most versatile shear strength test is the triaxial compression test. Triaxial tests 

allow a soil sample to be subjected to three principal stresses under controlled 

conditions. A cylindrical test specimen is encased in a rubber membrane and is then 

subjected to a confining pressure. Drainage from the sample is controlled through 

its two ends as the shearing force is applied axially and increased to failure. A plot 

of normal stress versus shear stress is developed and parameters c and φ are 

determined. In triaxial tests where full drainage is allowed during shear, or in 

undrained tests with pore pressure measurements during shear, the effective stress 

parameters c' and φ' can be determined.

In triaxial compression tests, the drainage, consolidation, and loading conditions are 

selected to simulate the applicable field conditions for the shear strength evaluation. 

Triaxial compression tests are classified according to the consolidation and drainage 

conditions allowed during testing. The three test types normally conducted are 

unconsolidated undrained (UU) (ASTM D 2850), consolidated undrained (CU) 

(ASTM D 4767), and consolidated drained (CD) (ASTM D 7181). Unconsolidated 

undrained tests provide more reliable shear strength values than those obtained 

from unconfined compression tests. Consolidated undrained tests with pore 

pressure measurements provide both effective stress and total stress parameters.

5.2.3.2 Effective Stress Friction Angle Correlations, Cohesionless Soils

For pile design in cohesionless soil, the effective internal friction angle is typically 

correlated from in-situ test results. Usually, the SPT N value or CPT cone resistance 

provides this relationship. Care should be exercised when using any SPT N value to 

soil parameter correlation. Depending upon the date and method, the correlation 

may be based on an uncorrected N value from a certain SPT hammer type or a 

corrected (N1)60. AASHTO (2014) design specifications note that the designer 

should ascertain the basis of the correlation and use (N1)60 or N, as appropriate. The 
123

SPT (N1)60 correlation to φ’ in AASHTO is presented in Table 5-5 and is the basis of 

the AASHTO resistance factors associated with the AASHTO calibrations.

Table 5-5 AASHTO (2014) Correlation Between SPT (N1)60 values to Drained 

Friction Angle of Granular Soils (modified after Bowles 1997)

SPT

Blow Count,

(N1)60

(blows/ft)

Angle of

Internal Friction

Φ’ 

(degree)

< 4 25-30

4 27-32

10 30-35

30 35-40

50 38-43

FHWA GEC-5 (2002) provides a different correlation for effective friction angle 

based upon the SPT N value. This correlation, presented in Table 5-6, was 

originally tabulated in Meyerhof (1956) for relatively clean sands. Sabatini et al. 

(2002) in GEC-5 recommended φ’ be reduced by 5° for clayey sand and increased 

by 5° for gravelly sand.

Table 5-6 Correlation Between Relative Density, SPT N value, and Internal 

Friction Angle for Cohesionless Soils in GEC-5 (2002) (after Meyerhof 1956)

State of Packing

Relative

Density, 

Dr

(%)

SPT

Blow Count,

N 

(blows/ft)

Angle of 

Internal Friction

Φ’ 

(degree)

Very Loose < 20 < 4 < 30

Loose 20-40 4-10 30-35

Compact 40-60 10-30 35-40

Dense 60-80 30-50 40-45

Very Dense > 80 > 50 > 45

Note: In saturated fine sand or silty sand, N = 15 + (N’-15) / 2 for N’ > 15 where N is 

the blow count corrected for dynamic pore pressure effects during the SPT, and N’ is 

the measured blow count.
124

Other correlation methods between corrected N values and laboratory test results 

have been published to provide drained friction angle equations for sands and 

gravels. Each method provides a correlation based upon the respective soil type, 

sampling method, and testing technique used.

Schmertmann (1975) assessed test data from clean sands and provides a 

correlation between friction angle, SPT N values and overburden stress. The data is 

interpreted for depths greater than 7-ft, for which Kulhawy and Mayne (1990) 

approximated a relationship as Equation 5-14.

 

Eq. 5-14

Hatanaka and Uchida (1996) utilized ground freezing to extract undisturbed samples 

of fine and medium sands. After subjecting samples to drained triaxial compression 

tests, an effective friction angle correlation to (N1)60 was established as shown in 

Equation 5-15.

 Eq. 5-15

Kulhawy and Chen (2007) performed strength tests on samples of sand and gravel. 

Although this study was performed to assess the uplift capacity of drilled shafts, a 

variable spread of sands and gravels were analyzed, and compared well with the 

Meyerhof (1956) correlations. For the combined 57 samples, a regression analysis 

(r2

=0.356, n =57) provided the relationship in Equation 5-16.

 Eq. 5-16

Similar to correlations developed for the SPT, the effective friction angle may be 

compared to the CPT/CPTu generally via cone tip resistance, qc. Meyerhof (1976) 

correlations are shown in Table 5-7 for relatively clean sands.
125

Table 5-7 Correlation Between Relative Density, CPT Cone Resistance, and 

Angle of Internal Friction for Clean Sands (after Meyerhof 1976)

State of Packing 

Relative

Density 

Dr

(%)

Cone Tip 

Resistance 

qc

(tons/ft2) 

Angle of Internal 

Friction

Φ’ 

(degree)

Very Loose < 20 < 4 < 30

Loose 20-40 4-10 30-35

Compact 40-60 10-30 35-40

Dense 60-80 30-50 40-45

Very Dense > 80 > 50 > 45

Robertson and Campanella (1983) compared cone resistance to drained triaxial 

compression tests results on clean, quartz sands. Using the cone tip resistance, a 

relationship to effective friction angle was presented by Kulhawy and Mayne (1990). 

This is shown in Equation 5-17.

 Eq. 5-17

Where:

φ' = effective stress angle of internal friction.

qc = cone tip resistance. 

σ'vo = vertical effective stress at the sample depth. 

For layers with cobbles and boulders, SPT N value correlations for the material 

friction angle can be problematic particularly when the material contains particles 

larger than the size of the split spoon sampler. The resultant inflated N values can 

result in overestimation of the friction angle of the rockfill materials. Therefore, the 

drained friction angle of rockfills, gravels, and sands can be estimated from Figure 

5-10 using the appropriate rockfill grade from Table 5-8 in conjunction with the 

effective normal stress. The rockfill grades of A through E identified in Table 5-8 

are based on the unconfined compressive strength of the rockfill particles. Once the 

rockfill grade is determined, a representative drained friction angle can obtained 

from Figure 5-10 using the average effective normal stress in the layer. The secant 

friction angle is based on a straight line constructed from the origin of the Mohr circle 

diagram to the intersection with the strength envelope at the effective normal stress 

as detailed in Terzaghi et al (1996).
126 

Figure 5-10 Typical ranges of friction angle for rockfills, gravels and sands 

(Note: 1 kPa = 0.145 psi) (after Terzaghi et al. 1996).

Table 5-8 Unconfined Compressive Strength of Particles for Rockfill 

Grades in Figure 5-12

Rockfill Grade Particle Unconfined Compressive 

Strength (ksf)

A ≥ 4610

B 3460-4610

C 2590-3460

D 1730-2590

E ≤ 1730
127

5.2.3.3 Fully Drained Shear Strength of Fine-Grained Cohesive Soils

Effective stress analysis methods may be used for fine grained soils under fully 

drained loading. This approach is useful to assess driven pile resistances a 

significant time after pile installation when pore water pressure has stabilized. For 

these cases, laboratory testing is recommended by AASHTO (2014) to produce 

effective stress design parameters φ’ and c’. Consolidated drained direct shear 

tests, consolidated drained (CD) triaxial tests or consolidated undrained (CU) triaxial 

tests with pore pressure measurements may be used to determine these shear 

strength parameters. Consideration for shear rate should be provided during these 

laboratory tests. Furthermore, complete dissipation of excess pore pressure for 

drained tests or equalization of pore pressure throughout the sample in undrained 

tests should be confirmed.

Preliminary estimates for effective friction angle may be interpreted from plasticity 

index (PI) tests. Terzaghi et al. (1996) assessed a range of clayey soils and shale to 

provide correlations between PI and effective friction angle. These results are 

presented in Figure 5-11. For these estimates FHWA GEC-5, Sabatini et al. (2002), 

recommends utilizing a cohesion value equal to zero (c‘ =0).

Figure 5-11 Relationship between φ’ and PI (after Terzaghi et al. 1996).

5.2.3.4 Undrained Shear Strength of Fine-Grained Cohesive Soils

The undrained shear strength, su, of cohesive soils may be measured via laboratory 

testing or through correlations with in-situ tests. However, laboratory testing of 

undisturbed samples is recommended. This may be accomplished through simple 
128

unconfined compression (UC) tests or by triaxial unconsolidated undrained (UU), or 

consolidated undrained (CU) testing. Sampler disturbance, strain rate variations and 

uncertain drainage conditions can affect shear strength results, although more 

conservative values are produced from these effects (Brown et al. 2010). CU triaxial 

tests offer the advantage of reconsolidating the sample to in-situ stress states before 

shearing, thus providing more representative shear strength values. AASHTO 

(2014) specifically recommends the use of CU and UU testing where possible.

For cases where undisturbed sampling is difficult or does not yield well preserved 

samples, in-situ testing may implemented as an alternative. A discussion of in-situ 

test methods is provided in Section 5.1. Common correlations between the CPT, 

SPT, and VST are provided below. Additional correlations may be found in GEC-5 

by Sabatini et al. (2002) among other sources.

A CPT to su correlation is provided in Equation 5-18. Local calibration of Nk is

recommended.

k

c vo

u N

q

s −σ ' = Eq. 5-18

Where:

qc = cone tip resistance. 

σ'vo = vertical effective stress at the sample depth. 

 Nk = cone factor, typically 14 to 16, may vary from approximately 10 to 18.

A preliminary estimate of undrained shear strength can be made from a corrected 

SPT N value in combination with atmospheric pressure, and a correction factor, f1, 

based on the plasticity index. Stroud (1974; 1989) utilized the f1 empirical correction 

as follows:

100

601 a

u

pNf

s = Eq. 5-19

Where:

f1 = 4.5 for PI = 50.

f1 = 5.5 for PI =15.

pa = atmospheric pressure.

N60 = SPT N value corrected for 60% energy transfer.

A preliminary estimate of su can also be made from uncorrected SPT N values 

(blows / ft). Table 5-9 contains an empirical relationship between the uncorrected 

SPT N value, unconfined compressive strength, and saturated unit weight for 
129

cohesive soils. The undrained shear strength is one half of the unconfined 

compressive strength (su = qu / 2). Correlation of N values to the undrained shear 

strength of clays is crude and unreliable for final design and as stated, should only 

be used for preliminary estimating purposes.

Table 5-9 Empirical Values for Unconfined Compressive Strength, qu, and 

Consistency of Cohesive Soils Based on Uncorrected N- Value (after Bowles 1977)

Consistency Very Soft Soft Medium Stiff Very Stiff Hard 

qu, ksf 0-0.5 0.5-1.0 1.0-2.0 2.0-4.0 4.0-8.0 8.0+

Standard

Penetration

N value

0-2 2-4 4-8 8-16 16-32 32+

γ (saturated), lb/ft3 100-120 100-120 110-130 120-140 120-140 120-140

The Vane Shear Test (VST) can be used for soft to medium clays and produces a 

correlation for su from the input torque, Tv, and vane diameter, dv. During the VST 

both peak and residual shear strengths are measured, thus the sensitivity can be 

calculated (Equation 5-5). When the vane height to diameter ratio is equal to two, 

hv/dv = 2, a widely used relationship found in GEC-5 by Sabatini et al. (2002) is 

shown in Equation 5-20. Furthermore, Bjerrum (1972) developed a correction based 

on static equilibrium theory as shown in Equation 5-21.

 For Eq. 5-20

Where:

su = undrained shear strength.

Tv = input torque during shear.

dv = vane diameter.

hv = vane height.

 Eq. 5-21

Where:

μ = correction factor.

PI = plasticity index.
130

5.2.4 Deformation

Deformation is a measure of the soil response to an applied load. For the design of 

driven piles, tolerable deformations relate to a structure’s serviceability limit state. 

These values are obtained through tests that quantify strain to provide modulus and 

consolidation parameters.

5.2.4.1 Elastic Deformation

Modulus, E, is determined through direct measurement of a strain from an applied 

load via Hooke’s Law (Equation 5-22) or through correlations with other laboratory 

and in-situ tests.

= Eεσ Eq. 5-22

Where: 

σ = stress. 

E = elastic modulus of material.

ε = strain.

A stress- strain curve developed for modulus determination is shown in Figure 5-12. 

In this figure, the origin and half the peak stress are designated as points to develop 

the 50% secant modulus, E50. 

Shear modulus, G, is sometimes used in load- deformation calculations and is 

related to the elastic modulus, E, and Poisson ratio, v, using Equation 5-23 below. 

Elastic deformation of soil will be further discussed in Chapter 7.

( ) +ν = 12

E G Eq. 5-23

A compilation of soil elastic modulus and poison ratio values have been made by 

Bowles (1988) and Department of the Navy (1982). These values were tabulated by 

AASHTO (2014) so that preliminary estimates of elastic soil deformation could be 

established. Correlations to corrected SPT N values and CPT cone resistance are 

also included. Testing should be performed to confirm or supersede these tabulated 

values. Table 5-10 to Table 5-12 are modified from AASHTO (2014). 
131

Figure 5-12 Soil stress strain curve and 50% secant modulus 

(after Brown et al. 2010).

Table 5-10 Estimating Soil Modulus, Es, Based on Soil Type (after AASHTO 2014)
132

Table 5-11 Estimating Soil Modulus, Es, from SPT N value (after AASHTO 2014)

Soil Type Es (ksf) 

Silts, Sandy Silts, Slightly Cohesive Mixtures 8 (N1)60

Clean Fine to Medium Sands, Slightly Silty Sands 14 (N1)60

Coarse Sands and Sands with Little Gravel 20 (N1)60

Sandy Gravels and Gravels 24 (N1)60

Table 5-12 Estimating Soil Modulus, Es, from Cone Resistance, qc 

(after AASHTO 2014)

Soil Type Es (ksf) 

Sandy Soils 2 qc 

5.2.4.2 Primary Consolidation Settlement

A one dimensional consolidation test (AASHTO T216 / ASTM D2435) is often 

performed to estimate the amount and rate at which a cohesive soil deposit will 

consolidate under an applied load. This test provides consolidation parameters 

including the coefficient of consolidation, Cv, compression index, Cc, recompression 

or swell index, Cr (alternatively Cs), and secondary compression index, Cα. 

For this test, a saturated soil sample is constrained laterally while being compressed 

vertically. The vertical compression is measured and related to the change in the 

soil void ratio. Loading the cohesive sample increases pore water pressure within 

the soil voids. Over a period of time, as the water is squeezed from the soil, this 

excess water pressure will dissipate resulting in the soil grains (or skeleton) 

supporting the load. The amount of water squeezed from the sample is a function of 

load magnitude and compressibility of soil deposit while the rate of pressure 

dissipation is a function of the permeability of the soil. Test background and analysis 

information may be found in numerous sources such as Bowles (1977), GEC-5, 

Sabatini et al. (2002), and Samtani and Nowatzki (2006).

The results from the test are used to plot void ratio, e, versus vertical effective 

stress, σ’vo, on a semi log scale to determine the preconsolidation pressure, σp, and 

other quantities. An illustration of a typical e-log σ’v curve is presented in Figure 

5-13. A plot of log time versus sample compression is used to determine coefficient 

of consolidation. Consolidation test results can be used to estimate magnitude and 
133

settlement rate of pile foundations in cohesive soils. Consolidation tests can also be 

plotted as the vertical strain, ε versus vertical effective stress, σ’v, on a semi log 

scale. 

Equations 5-24 to 5-26 utilize the results of consolidation testing to determine 

primary consolidation settlement of cohesive soil.

For normally consolidated (σ’vo = σp) cohesive soil layers:

 Eq. 5-24

For overconsolidated soil with (σ’vo + Δσ ≤ σp):

 Eq. 5-25

Figure 5-13 Plot of Void Ratio vs Vertical Effective Stress from consolidation test. 
134

For overconsolidated soil with (σ’vo + Δσ > σp):

 Eq. 5-26

Where:

Sc = settlement from primary consolidation.

eo = initial soil layer void ratio.

Ho = initial soil layer thickness.

σp = preconsolidation stress.

σ'vo = average vertical effective stress at the mid-point of the soil layer 

before loading.

Δσ = additional stress at mid-point of soil layer from loading.

5.2.5 Electro Chemical Properties

The soil and groundwater can contain constituents detrimental to pile materials. 

Electro chemical classification tests can be used to determine the aggressiveness of 

the subsurface conditions and the potential for pile deterioration. For pile design, 

corrosion rates for steel piles or disintegration rates for concrete may be required to 

evaluate long term structural resistance. These electro chemical tests include:

a. pH (AASHTO T289 / ASTM D4972).

b. Resistivity (AASHTO T288 / ASTM D1125).

c. Sulfate ion content (AASHTO T290 / ASTM D4230).

d. Chloride ion content (AASHTO T291 / ASTM D512).

Additional discussions on the influence of environmental conditions on pile section 

selection are presented in Section 6.14 of Chapter 6.

5.3 ROCK PARAMETERS

For the design of driven piles, rock property determination is necessary to assess 

end bearing resistance as well as long term deformation. Shallow rock profiles may 

necessitate predrilling pile locations into these materials to satisfy minimum 

penetration requirements. A designer’s understanding of rock classification and 

associated rock characteristics is therefore essential. To properly classify and rate a 

given quantity of rock, a distinction between intact rock and rock mass should be 
135

noted (Brown et al. 2010). Intact rock is a collection of consolidated and cemented 

particles forming rock material, where rock core samples are used to determine 

index and strength properties. Joints, faults, and other bedding features serve to 

break up intact rock, therefore this inclusive system combines to form a rock mass.

5.3.1 Rock Index Properties and Classification

When designing foundations that bear on rock, proper rock classification is essential. 

Physical and chemical properties of the rock should be noted during the subsurface 

exploration as the rock mass quality may affect design and construction. Rock 

classification should include color, texture, lithology (e.g. igneous, metamorphic, or 

sedimentary), field hardness, weathering, and geologic discontinuities.

The slake durability test (ASTM D4644) is useful to assess if rock will weather and 

deteriorate. The slake durability test is useful for weak rocks including shale, 

mudstone, and siltstone. These deposits are particularly susceptible to weathering 

or fracturing during the pile driving process. Low slake durability index results may 

help identify deposits in which driven piles are subject to relaxation.

The slake durability test requires representative rock fragments be dried to a 

constant weight in a wire mesh drum. After submerging and rotating the drum with 

containing rock fragments for 10-minutes in water, the sample is again dried to 

constant weight. This process is repeated a second time, and the ratio of final to 

initial dry weights results in the slake durability index, ID. An ID of less than 60% 

correlates to a rock prone to rapid degradation.

5.3.2 Rock Mass Shear Strength

Rock shear strength is typically measured in the laboratory through uniaxial 

compression testing where recovered core samples are prepared and subjected to 

loading in accordance with ASTM D7012. As load is applied during this test, axial 

strain is measured and plotted to determine the elastic modulus. The peak load is 

divided by the specimen’s cross sectional area to provide an unconfined 

compressive strength, qu. AASHTO and other methods for determining the nominal 

resistance of end bearing piles on rock utilizes the rock unconfined compressive 

strength.
136

5.3.3 Rock Mass Deformation

Rock deformation is typically elastic. Laboratory uniaxial compression test results 

can provide information on modulus values if rock deformation calculations are 

required. Rock deformation for most rock formations is generally not a design 

consideration as piles typically penetrate into weak rock or terminate on hard rock. If 

heavily loaded piles are terminated just into or above soil-like weak rock formations 

such as marl, rock mass deformation properties may be needed for design. To 

estimate the modulus values for rock, the designer should consider the RQD, GSI, 

and core recovery and use correlations of these parameters with rock mass 

modulus.

As noted in the previous section, the material modulus, E, is measured by plotting 

load versus strain during the uniaxial compression test. Careful axial and 

diametrical measurements also yield the material’s Poisson ratio, ν. The modulus is 

the slope of the stress-strain curve. Figure 5-12 utilizes a 50% secant modulus, E50, 

where a stress value equal to one-half of the maximum deviator stress was selected. 

Test result tables complied by Kulhawy (1978) for numerous rock types have been in 

use to estimate elastic modulus and Poisson ratio. These tables are reproduced in a 

number of manuals including Samtani and Nowatzki (2006) and Table C10.4.6.5-1 

of AASHTO (2014).

5.4 CONSIDERATIONS FOR PILE DRIVABILITY

AASHTO Article 10.7.8 stipulates that an evaluation of pile drivability should be 

performed during the design stage. Pile drivability is defined as the ability to drive 

the pile to the estimated nominal axial resistance anticipated during driving at or 

above the required penetration depth and to achieve penetration requirements within 

the driving stress limits of the pile material. To satisfy this objective, a detailed 

identification of the subsurface conditions and a thorough understanding of the soil 

behavior and its response to pile installation are needed.

Some subsurface issues that influence drivability are fairly obvious such as the 

presence of boulders above the bearing strata that increase the risk of pile damage

or the identification of artesian conditions that can result in reduced geotechnical 

resistances. Other factors such as fine content, angularity, density, as well as high 

soil plasticity can all have a significant influence on drivability.
137

Cases of large soil quakes or high elastic behavior have been reported in dense silty 

sand, hard silty clay, glacial tills and other fine grained saturated soils (e.g., Likins 

1983; Hannigan 1985; Hussein et al. 2006). Furthermore, Cosentino et al. (2010) 

researched cases of high rebound soils for the Florida DOT. High rebound is an 

effect of increased pore water pressure due to driving generally occurring near the 

pile toe. As driving continues, pore water pressure increases and causes a larger 

elastic response and rebound, leading to refusal driving conditions. If pile driving is 

paused, the pore water pressure dissipates and subsequently smaller rebound 

occurs. The research performed on the Hawthorn Group formation, a Florida soil 

formation, determined that high silt content and N value changes within the strata 

appeared to have strong correlations with soil zones of high elastic rebound.

Some sand deposits have also exhibited high dynamic resistances during pile 

driving. Thompson and Goble (1988) summarized details from nine projects that 

had this behavior. Unfortunately no clear relationship was identified between soil 

deposition, geologic, or mineralogical characteristics. This suggests drivability 

assessments should consider a range in dynamic soil properties rather than a 

specific value. Similarly, a range of dynamic soil properties should be evaluated in 

drivability analyses in highly plastic CH and MH soils where conventional values may 

underestimate the dynamic soil resistances encountered in these materials.

For certain weathered rock such as shale and siltstone, competent material is 

typically encountered during initial drive. However, a reduction in pile end bearing 

resistance can occur with time due to the release of locked in horizontal stresses in 

the laminated rock. Driving of adjacent piles to the same depth or below previously 

installed piles can also shatter the weak rock causing a reduction in end bearing 

resistance. Because rock types vary by region, local experience and analysis 

methods are advantageous in this regard.

5.5 SELECTION OF PARAMETERS FOR DESIGN AND CONSTRUCTION

Soil and rock parameters should be carefully selected based on site specific in-situ 

and laboratory testing results and relevant design methods for driven pile 

foundations. In some cases, resistance factors were developed for a specific design 

methodology based on a specific correlation approach between N and φ, between N 

and su, or a design method limit on the maximum friction angle.
138

5.5.1 Soil Parameters

Preliminary estimates of the soil shear strength and density parameters are often 

made from published correlations with SPT N values. Many of these correlations 

were provided in Section 5.2.3.3 and 5.2.3.4. While these correlations are very 

useful for preliminary design assessments, soil parameter confirmation and 

refinement should be obtained through additional in-situ and laboratory tests. 

Parameter variability for each layer must be considered for final soil parameter 

selection and final design. The shear strength, compressibility and index properties 

of each layer should be adequately quantified for design.

In coarse granular deposits, the selection of the design friction angle should be done 

conservatively. A comparison of nominal resistances from static load test results 

with static analysis predictions indicates that static analyses often overpredict the 

shaft resistance in these deposits. This is particularly true for coarse granular 

deposits comprised of uniform sized or rounded particles. Cheney and Chassie 

(1993) recommended limiting the shearing resistance by neglecting particle interlock 

forces. For shaft resistance calculations in gravel deposits, this results in a 

maximum φ’ angle of 32° for gravels comprised of soft rounded particles, and in a 

maximum φ’ angle of 36° for hard angular gravel deposits. The φ’ angle used to 

calculate the toe resistance is determined using normal procedures.

Final design methods are often selected based on a given foundation type or from 

prior satisfactory design experience with the design method in similar subsurface 

conditions. The soil parameters required for method use must be acquired during 

the subsurface exploration program. Necessary parameters may include USCS 

classification, moisture content, density, grain size distribution, and Atterberg limits 

in addition to shear strength and compressibility information. In AASHTO Article 

10.4, resistance factors have been developed for specific design method equations 

using specific procedural guidance. Hence, the soil parameters necessary to 

perform the analyses should be obtained in the prescribed manner. Design methods 

are discussed further in Chapter 7.

5.5.2 Rock Parameters

Design parameters for rock must be determined for piles driven either into soft rock 

or to hard rock. Rock cores should be collected for both soft and hard rock layers 

and the extent of rock weathering and fracturing should also be determined. The 

value of local experience with a given rock mass behavior cannot be overstated, as 

regional rock formations pose their own unique features.
139

A clear definition of soft and hard rock is not available in technical literature, in 

AASHTO, or in other building codes. In many intact rock classification systems, the 

transition between hard soil and soft rock occurs at an unconfined compression 

strength, qu, of around 20 ksf. Similarly, the transition between soft rock and hard 

rock occurs between unconfined compression strength of 200 and 1000 ksf. In 

driven pile design and construction, soft rock is defined as rock that can be 

penetrated by pile driving and hard rock is defined as rock that cannot be 

penetrated. Hence, piles are driven into soft rock and driven to hard rock.

Soft and weak rocks are difficult materials to sample, test, and quantify. They have 

shear strengths greater than that of conventional soils and less than hard rock. Soft 

rock strengths can be too strong to be tested in soil laboratory equipment and too 

weak to be tested in rock mechanics equipment. Soft rocks often exhibit additional 

problematic characteristics such as crumbling, laminations, disaggregation, high 

plasticity, slaking, and variable strengths depending on weathering. All of these 

considerations complicate determination of representative parameter selection for 

pile design in soft rocks. Usual types of soft rock were identified by Kanji (2014) and 

are listed in Table 5-13. 

Table 5-13 Usual Types of Soft Rocks (after Kanji 2014).

Rock Classification Subclasses

Sedimentary Rocks

Mudstones, shales, siltstones, sandstones, 

conglomerates, breccias, marl, limestone, dolomites, 

gypsum, coal, salt rock, carnallite,

Igneous Rocks

Volcanic breccias and lahars, basaltic breccias, 

pyroclastic deposits, volcanic ash, tuff, igimbrite, and 

weathering products of crystalline rocks (granite, gneiss) 

Metamorphic Rocks Slate, phyllite, schists, quartzite little cemented, and 

metavolcanic deposits.

As noted above, piles can be driven into soft and weak rocks. In these materials, 

rock classification, core recovery, RQD, unconfined compression strength, density, 

and slake durability index parameters should be quantified. Piles cannot be driven 

into hard rock as pile toe damage or yielding occurs before hard rock failure. As with 

soft and weak rocks, rock classification, core recovery, RQD, unconfined 

compression strength, and density parameters should be quantified for pile designs 

on hard rock.
140

5.5.3 Site Variability

After a thorough site investigation and characterization of material properties, a 

subsurface profile is then generated for design and constructability purposes. Often 

times, throughout a relatively uniform soil model, results of laboratory tests and in 

situ tests present conflicting information. Soil cohesion, relative density and 

permeability among other properties may vary within a given strata along a 

horizontal plane or with depth. Final selection of material parameters then becomes 

reliant upon the use of engineering judgment.

Chapter 8 of GEC-5, Sabatini et al. (2002), provides a detailed overview of 

incorporating engineering judgment into the selection of design parameters when 

site variability is encountered. When facing situations of this nature, the Engineer 

can reevaluate subsurface parameters by either (1) locating inconsistencies with 

reported test results or (2) using historical information (i.e. experience, published 

data) to assess the inherent site variability.

Site variability relates to the variability within similar site conditions of the same site, 

not between sites. Furthermore, determination of site variability should be 

performed for each substructure location. When pile designs are prepared (Chapter 

7 and 8), they are done so with respect to the encountered subsurface conditions 

which directly correlate to performance criteria for construction and the resulting 

structure. Generally, the nearest boring(s) or in-situ test(s) are utilized for this 

purpose. Judgment may be applied when establishing a subsurface profile. 

However, the following three steps provided in Paikowsky et al. (2004) serve to aide 

this process.

1. Relate each significant bearing layer and determine average parameters used 

for strength analysis at each boring location.

2. Check the Coefficient of Variation (COV) between the average values for 

each identifiable significant layer obtained at each boring location.

3. Categorize site variability based on COV as low, medium, or high.

a. COV < 25% - Low

b. 25% ≤ COV < 40% - Medium

c. 40% ≤ COV - High
141

When analysis results yield a high COV, additional tests, different tests, engineering 

judgement, or further division of the project site into smaller less variable sub-sites 

may be required.

A larger statistical analysis of design parameters can also serve as a means to 

identify variability. For this, a database of similar geomaterials is constructed with 

the goal of assessing variation within a particular property and the effect on the 

project at large. Duncan and Wright (2005) compiled Table 5-14 based on the 

variability of geotechnical properties from in situ tests. They present the COV for a 

select number of soil parameters, which have been compiled from a number of 

sources. It should be noted that these values are obtained from projects covering a 

wide range of material, and this should not be applied to specific cases.

Sabatini et al. (2002) suggests the use of statistics and sensitivity analyses as a 

minimum practice for design parameter selection. Based on reported test results, 

determination of the sample mean and standard deviation is a relatively 

straightforward process. The COV is a statistical measure of the extent of variability 

and is calculated by dividing the sample standard deviation by the sample mean. By 

applying these statistics to a sensitivity study for design methods, the influence of a 

given design parameter can be assessed. Several designs can then be completed 

efficiently with the use of computer software or spreadsheets.

Table 5-14 Coefficients of Variation for Geotechnical Properties and In-Situ Tests 

(after Duncan and Wright 2005)

Property or In-Situ Test COV (%)

Unit Weight (γ) 3-7 

Buoyant Unit Weight (γ’) 0-10

Effective Stress Friction Angle (φ’) 2-13

Undrained Shear Strength (su) 13-40

Undrained Strength Ratio (su/σ’vo) 5-15

Standard Penetration Test N value (N) 15-45

Electric Cone Penetration Test (qc) 5-15

Mechanical Cone Penetration Test (qc) 15-37

Dilatometer Test Tip Resistance (qDMT) 5-15

Vane Shear Test Undrained Strength (su) 10-20
142

5.6 GEOTECHNICAL REPORTING

Upon conclusion of the site characterization program, the geotechnical engineer 

summarizes the factual data and prepares a geotechnical report. Three types of 

geotechnical reports are appropriate for driven piles: 1) geotechnical data reports; 2) 

geotechnical foundation design reports; and 3) geotechnical baseline reports. The 

type of geotechnical report prepared depends on project type and owner 

requirements.

Geotechnical data reports simply document and transmit a record of the exploration 

performed and the collected data. These data reports may be used on conventional 

Design-Bid-Build or Design-Build projects. Other agencies prefer foundation design 

reports that include the results of the subsurface exploration program as well as an 

assessment of the subsurface conditions with supporting design recommendations. 

The third type of report, a geotechnical baseline report, provides a contractual 

understanding of the subsurface conditions to be expected during construction. 

Geotechnical baseline reports are relatively uncommon on driven pile projects but 

are increasing in use particularly on Design-Build projects with highly variable site 

conditions. Dwyre et al. (2010) provides suggested guidelines for the application of 

geotechnical baseline reports to foundation projects. The various methods of 

geotechnical reporting are discussed in greater detail in the subsequent sections. 

Table 5-15 summarizes the decisions in driven pile design and construction that are 

influenced by the subsurface conditions.

5.6.1 Geotechnical Data Reports

A Geotechnical Data Report (GDR) contains only factual information from the 

performed site exploration. This includes a description of the geologic setting, site 

exploration program, and field and laboratory testing program. Also included are 

completed boring logs as well as field and laboratory test results. The GDR does 

not offer subsurface interpretations, or design recommendations, and is typically 

used when a subcontracted consultant performs the field exploration and associated 

soil or rock testing for the foundation designer. Information included in Geotechnical 

Data Reports is summarized in Table 5-16. The Owner’s design team is responsible 

for interpreting the GDR. For design-build projects, the GDR may be used for 

construction bidding and must be included as a Contract Document.
143

Table 5-15 Pile Foundation Decisions Influenced by Subsurface Information

Decisions Subsurface Information

Project Constraints and 

Auxiliary Equipment

Old foundations, boulders, obstructions. 

Site accessibility and terrain. 

Site constraints, overhead and underground utilities, 

adjacent buildings. 

Pile Type and 

Accessory Selection

Subsurface strata and installation conditions. 

Structural resistance needs. 

Geotechnical resistance available. 

Pile lengths. 

Need for pile splices. 

Bedrock type and depth. 

Water table. 

Coatings due to aggressive or abrasive environment. 

Pile driving shoes. 

Pile Driving Equipment 

Selection

Hammer type requirements. 

Hammer size and energy requirements. 

Crane, leads, template, and other equipment. 

Drilling equipment, augers, spuds, jets. 

Construction Control Hammer energy control. 

Resistance determination method. 

Time dependent soil strength changes. 
144

Table 5-16 Information Included in Geotechnical Data Reports 

(after Brown et al. 2010)

Background Information Overview of project (bridge, structure, retaining wall, or other facility. 

General site conditions (geology, topography, drainage, accessibility).

Specific methods used for site exploration.

Scope of Site Exploration Plans showing location of all borings, test pits, and in-situ test locations.

Number, locations, and depths of all borings and in-situ tests.

Types and frequency of samples obtained; standards used.

Types and numbers of laboratory tests; standards used.

Subcontractors performing the work and dates of work.

Data Presentation Final logs of borings and test pits.

Water level readings and other groundwater data. 

Data tabulations and plots from each in-situ test hole.

Summary tables and data sheets for lab tests performed.

Photographs of rock core.

Results of geophysical tests.

Geological mapping data sheets and summary plots.

Existing information from previous site explorations (boring logs, data). 

5.6.2 Geotechnical Foundation Design Reports

A foundation design report is prepared to present the results of the subsurface 

explorations, laboratory test data, analysis, and specific design and construction 

recommendations for the foundation system of a structure. A foundation design 

report includes all the information noted in Table 5-16. In addition, the foundation 

design report includes an assessment of the subsurface conditions, identifies the soil 

and rock stratigraphy, provides design parameters for each soil and rock layer, 

presents analyses performed and their results, and provides recommendations for 

foundation design and construction. The report should make a clear distinction 

between factual and interpretive information. Foundation design reports are referred 

to frequently during the design and construction period as well as in resolving post 

construction issues such as claims. It is therefore important that the foundation 

design report be clear, concise and accurate. The foundation report is a very 

important document and should be prepared and reviewed accordingly.
145

Preliminary design recommendations based on, and/or transmitted with initial 

subsurface data does not constitute a foundation design report. A foundation design 

report must address each design issue such as axial compression resistance, uplift 

resistance, settlement, lateral resistance and response to lateral loading, seismic, 

scour, etc., in accordance with applicable limit state design methodologies. Only 

with this information can a foundation design report be prepared with appropriate 

content and quality.

The parts of a foundation design report are described below, and are modified from

Cheney and Chassie (2000).

I. Table of Contents.

II. Introduction.

a. Summary of proposed construction, including factored foundation loads.

b. Summary of applicable extreme event limit states and loads.

c. Foundation performance criteria (total and differential settlements, lateral 

deformation).

d. Hydraulic information (if applicable).

e. Summary of site constraints including accessibility, environmental 

restrictions (noise, vibrations, contaminated soil or groundwater, marine 

mammals, fish, wildlife), utility conflicts, as well as any limitations on 

headroom or equipment. 

III. Scope of Explorations.

a. Field explorations (summary of dates and methods, appended results).

b. Laboratory testing (summary of types of tests, appended results).

IV. Interpretation of Subsurface Conditions.

a. Description of formations.

b. Soil types. 

c. Rock types. 

d. Dip and strike of rock.

i.Regional.

ii.Local.

e. Water table data.

i.Perched.

ii.Regional.

iii. Artesian.
146

V. Geomaterial Design Parameters. (Narrative to describe procedure for 

evaluating factual data to establish design values for all soil and rock layers).

a. Shear strength.

b. Compressibility.

c. Design analysis. 

VI. Description of Design Procedures.

a. Summary of results.

b. Explanation of interpretation.

VII. Geotechnical Conclusions and Recommendations. 

VIII. Pile foundations.

a. Type of pile geotechnical resistance: shaft resistance, toe resistance, 

or both.

b. Delineation of unsuitable layers due to compressibility, scour, or 

liquefaction.

c. Suitable pile types: reasons for choice and/or exclusion of types and 

optimization of the recommended section.

d. Estimated pile toe elevation, (average estimated values with probable 

variation potential).

e. Estimated pile lengths.

f. Minimum pile penetration (AASHTO Article 10.7.7 summarizes reasons 

why a minimum penetration depth could be specified to satisfy all 

applicable limit states).

g. Nominal axial compression and axial tension resistances. Lateral load 

resistance and deformation.

h. Location of the neutral plane and drag force.

i. Estimated pile group settlement. Very important for pile groups in 

cohesive soils and large groups in a cohesionless soil deposit 

underlain by compressible soils.

j. Number of test piles and specific test pile locations for maximum utility.

k. Static pile load tests. If used, specify test locations for maximum 

utility.

i. Axial compression.

ii. Axial tension.

iii. Lateral. 

l. Dynamic pile load tests. If used, specify test locations and restrike 

time and frequency.

m. Rapid load tests. If used, specify test locations.
147

n. Resistance determination method (driving criteria). Nominal driving 

resistance depends on the resistance determination method per 

AASHTO Table 10.5.5.2.3-1.

o. Nominal driving resistance.

p. Pre-boring, pile toe reinforcement, or other requirements to reach pile 

penetration requirements or handle potential obstructions.

q. Pile driving requirements: hammer size, alignment and location 

tolerances, sequence of driving, etc.

r.Cofferdams and seals; seal design should consider potential conflicts 

between batter piles driven at alignment tolerance limits and depth of 

sheeting. Group densification inside sheeting for displacement piles 

in sands, or heave for displacement piles in clays should be 

considered.

s. Corrosion effects or chemical attack; particular concern in marine 

environments, old dumps, areas with soil or groundwater 

contaminants.

t. Effects of pile driving on adjacent construction; settlements from 

vibrations and development of excess pore water pressures in soil.

u. Other pile foundation construction issues.

1. Conflicts with existing foundations, foundation remnants, or other 

obstructions.

2. Cobbles and boulders.

3. Groundwater control and/or pile cap excavation stability.

IX. Appendix: Graphic Presentations.

a. Map showing project location.

b. Detailed plan of the site showing proposed structure(s) borehole 

locations and existing structures.

c. Laboratory test data.

d. Finished boring logs and interpreted soil profile.

X. Report Distribution.

Copies of the completed Foundation Report should be transmitted in accordance 

with agency protocol.

The foundation report should be widely distributed to design, construction and 

maintenance engineers involved in the project. The foundation report should also 

furnish information regarding anticipated construction problems and solutions. This 

will provide a basis for the contractor's cost estimates.
148

On conventional design-bid-build contracts, the foundation design report should be 

completed and available to the designer prior to final design. The foundation 

drawings, special provisions, and foundation design report should all be crosschecked for compliance upon completion of final design documents. Conflicts 

between any of these documents greatly increase the potential for construction 

problems. For design-build contracts, factual data is provided by the owner to the 

design-build team that then performs the final geotechnical exploration and 

foundation design report.

5.6.3 Geotechnical Baseline Reports

Uncertainties in subsurface construction and resulting claims resulted in the concept 

of a Geotechnical Baseline Report (GBR). This concept originated in the tunneling 

industry where due to the linear nature of construction activities an unexpected 

condition disrupting construction quickly had significant cost and schedule 

implications. Geotechnical Baseline Reports establish a single source document 

where the geotechnical conditions to be expected (or assumed) are established on a 

contractual basis. This allows the Owner and Contractor to allocate geotechnical 

construction risk according to the established baseline of expected subsurface 

conditions. Risks associated with subsurface conditions equivalent to or better than 

the baseline conditions are borne by the contractor and risk associated with more 

severe subsurface conditions are allocated to the owner.

A GBR represents what is assumed that will be encountered for contract purposes. 

Baselines stated in the GBR should be well defined, reasonable and realistic. These 

should be derived from the site exploration performed, and/or local experience, and 

provide rational for specific requirements, designs or methods to be used. A GBR is 

not a warranty. It does provide a legal foundation for differing site condition claims. 

An example of a GBR statement relative to pile driving would be “Obstructions will 

be encountered between EL 610 and EL 605. For baseline purposes, these 

obstructions are reinforced concrete footing rubble, up to 3 feet thick, covering 15% 

of the substructure footprint at Pier 4.”

A GDR also provides a summary of the site characterization and expected 

subsurface conditions. For projects where both a GDR and GBR exist, a GBR 

should take precedence over a GDR in the hierarchy of contract documents. More 

information on the GBR concept may be found in Essex (2007).
149

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155

CHAPTER 6

PILE TYPES FOR FURTHER EVALUATION

The economic selection of a pile foundation type and section for a structure should 

be based on the specific subsurface conditions as well as the foundation loading 

requirements, performance criteria, construction limitations and schedule, as well as 

the foundation support cost. Piles can be broadly categorized in two main types: 

foundation piles for support of structural loads and piles for earth retention systems. 

The use of piles for earth retention systems is outside the scope of this manual. 

This chapter focuses on the characteristics of driven pile types typically used for 

highway structure foundations. Additional details on pile splices and toe protection 

devices are presented in Chapter 16.

6.1 OVERVIEW OF TYPICAL PILE TYPES

There are numerous types of piles used for foundation support. Figure 6-1 shows a 

pile classification system based on type of material, configuration, installation 

technique and equipment used for installation. Foundation piles can also be 

classified on the basis of their method of load transfer from the pile to the 

surrounding geomaterial. Load transfer can be by shaft resistance, toe resistance or 

a combination of both.

Tables 6-1 to 6-7, modified from NAVFAC (1982), summarize characteristics and 

uses of common driven pile types. The tables are for preliminary guidance only, and 

should be confirmed by local practice. The typical factored resistance may be 

controlled by the structural or available geotechnical resistance which may in turn be 

limited by installation conditions or performance requirements. Although drilled and 

bored piles are included in the pile classification chart shown in Figure 6-1, these 

foundation types are outside the scope of the FHWA driven pile foundation manual. 

Reference should be made to the FHWA reference manuals for drilled shafts (Brown 

et al. 2010), continuous flight auger (CFA) piles (Brown et al. 2007) and micropiles 

(Sabatini et al. 2005) for detailed information on these deep foundation types.
156

Figure 6-1 Pile classification chart.
157

Table 6-1 Timber Piles Technical Summary

PILE TYPE TIMBER PILES 

TYPICAL 

LENGTHS 

15 to 75 feet for Southern Pine. 

15 to 120 feet for Douglas Fir. 

MATERIAL

SPECIFICATIONS 

ASTM D25. 

AWPA UC4A, UC4B, UC4C, UC5A, UC5B and UC5C.

TYPICAL 

FACTORED 

RESISTANCE 

50 to 120 kips. 

MAXIMUM

DRIVING

STRESS 

σdr = ɸda (2.6 Fco). 

ɸda = 1.15. 

Fco = 1.25 ksi for Douglas Fir, 1.20 ksi for Southern Pine.

ADVANTAGES

• Comparatively low in initial cost.

• Permanently submerged piles are resistant to decay.

• Easy to handle.

DISADVANTAGES 

• Difficult to splice.

• Vulnerable to damage in hard driving; both pile head and 

pile toe may need protection.

• Intermittently submerged piles are vulnerable to decay 

unless treated.

REMARKS 

• Best suited for friction piles in granular material.

• Suitable for friction piles with lower factored resistances 

in cohesive soils. 

Figure 6-2 Timber pile typical illustration. 
158

Table 6-2 Steel H-Piles Technical Summary

PILE TYPE STEEL – H-PILES 

TYPICAL 

LENGTHS 15 to 200 feet.

MATERIAL

SPECIFICATIONS 

ASTM - A572, A588, or A690 Grade 50, 60.

(A572 Grade 50 is standard). 

TYPICAL 

FACTORED 

RESISTANCE 

260 to 1,600 kips. 

MAXIMUM

DRIVING

STRESS 

σdr = 0.9 ɸda Fy. 

ɸda = 1.00.

Fy = Yield strength of steel (ksi).

ADVANTAGES

• Available in various lengths and sizes.

• High factored resistance. 

• Small soil displacement.

• Easy to splice. 

• Pile toe protection will assist penetration through harder 

layers and some small obstructions. 

DISADVANTAGES 

• Vulnerable to corrosion where exposed and in corrosive 

soil conditions.

• HP section may be damaged or deflected by major 

obstructions.

REMARKS 

• Best suited for toe bearing on rock.

• Factored resistance reduced in corrosive environments.

• Length and cost overruns often occur when used as a 

friction pile in granular materials.

Figure 6-3 Steel H-pile typical illustration. 
159

Table 6-3 Steel Pipe Piles Technical Summary

PILE TYPE STEEL – PIPE PILES

TYPICAL 

LENGTHS 15 to 200 feet.

MATERIAL

SPECIFICATIONS 

ASTM A252 Grade 2 or 3, API 5L, or API 2B - for pipe.

ACI 318 - for concrete (if filled).

ASTM A572 - for core (if used).

TYPICAL 

FACTORED 

RESISTANCE 

100 to 1,250 kips (closed end, D ≤ 30 in.) with concrete fill. 

660 to 6,500 kips (open end, 16 in. ≤ D ≤ 72 in.) no 

concrete.

MAXIMUM

DRIVING

STRESS 

σdr = 0.9 ɸda Fy. 

ɸda = 1.00 for non-composite during driving.

Fy = Yield strength of steel (ksi). 

ADVANTAGES 

• Closed end pipe can be internally inspected after driving. 

• Low soil displacement for open end installation.

• High factored resistances depending on section. 

• Open end pipe with shoe can be used for obstructions. 

• Open end pipe can be cleaned out and driven further.

• Easy to splice. 

DISADVANTAGES 

• Vulnerable to corrosion where exposed and in corrosive 

soil conditions.

• Potential soil displacement from larger closed end pipe.

REMARKS 

• Provides high bending resistance where unsupported 

length is loaded laterally.

• Open end not recommended as a friction pile in granular 

material due to tendency for length and cost overruns.

 

Figure 6-4 Steel pipe piles typical illustration. 
160

Table 6-4 Precast, Prestressed Concrete Technical Summary

PILE TYPE PRECAST PRESTRESSED CONCRETE PILES 

TYPICAL 

LENGTHS 30 to 150 feet. 

MATERIAL

SPECIFICATIONS 

ACI 318 - for concrete. 

ASTM - A82, A615, A722, and A884 - for reinforcing steel. 

ASTM - A416, A421, and A882 - for prestressing. 

TYPICAL 

FACTORED 

RESISTANCE 

350 to 2,200 kips on solid square piles.

1,500 to 3,000 kips on spun cast cylinder piles.

MAXIMUM

DRIVING

STRESS 

σdr = ɸda (0.85 f’c - fpe ), In compression.

σdr = ɸda (0.095 +fpe ), In tension (normal conditions).

σdr = ɸda (fpe ), In tension (severe conditions).

ɸda = 1.00.

f’c = Concrete compressive strength (ksi).

fpe = Effective prestress (ksi).

ADVANTAGES

• High factored resistances. 

• Corrosion resistance obtainable.

• Hard driving possible.

DISADVANTAGES 

• Vulnerable to handling damage.

• Can have relatively high breakage rate.

• Potential soil displacement effects from large sections. 

• Difficult to splice when insufficient length ordered.

REMARKS • Cylinder piles are well suited for bending resistance. 

Figure 6-5 Precast, prestressed concrete typical illustration. 
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Table 6-5 Monotube Pile Technical Summary

PILE TYPE MONOTUBE PILES

TYPICAL 

LENGTHS 15 to 100 feet.

MATERIAL

SPECIFICATIONS 

ACI 318 - for concrete. 

ASTM A252 - for steel pipe. 

TYPICAL 

FACTORED 

RESISTANCE 

100 to 450 kips.

MAXIMUM

DRIVING

STRESS 

σdr = 0.9 ɸda Fy. 

ɸda = 1.00 for non-composite during driving.

Fy = Yield strength of steel (ksi).

ADVANTAGES

• High factored resistance for relatively shorter lengths.

• Increased shaft resistance from tapered section.

• Fluted shell not easily damaged. 

DISADVANTAGES 

• Potential soil displacement effects. 

• Vulnerable to corrosion where exposed and in corrosive 

soil conditions.

REMARKS • Best suited as a friction pile in granular soils. 

Figure 6-6 Monotube pile typical illustration. 
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Table 6-6 Tapertube Piles Technical Summary

PILE TYPE TAPERTUBE PILES

TYPICAL 

LENGTHS 50 to 150 feet.

MATERIAL

SPECIFICATIONS 

ACI 318 - for concrete. 

ASTM A252 - for steel pipe.

TYPICAL 

FACTORED 

RESISTANCE 

200 to 850 kips. 

MAXIMUM

DRIVING

STRESS 

σdr = 0.9 ɸda Fy. 

ɸda = 1.00 for non-composite during driving.

Fy = Yield strength of steel (ksi).

ADVANTAGES

• High factored resistance for relatively shorter lengths.

• Standard pipe piles may be spliced to tapered sections. 

• Increased shaft resistance from tapered section.

• Reduced concrete fill volume in tapered section.

DISADVANTAGES 

• Potential soil displacement effects. 

• Vulnerable to corrosion where exposed and in corrosive 

soil conditions.

REMARKS • Best suited as a friction pile in granular soils.

Figure 6-7 Tapertube pile typical illustration. 
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Table 6-7 Composite Piles Technical Summary

PILE TYPE COMPOSITE PILES

TYPICAL 

LENGTHS 50 to 200 feet.

MATERIAL

SPECIFICATIONS 

ASTM A572 - for HP section. 

ASTM A252 - for steel pipe.

ASTM D25 - for timber. 

ACI 318 - for concrete. 

TYPICAL 

FACTORED 

RESISTANCE 

100 to 1,250 kips. 

MAXIMUM 

DRIVING

STRESS 

Varies depending on pile materials.

ADVANTAGES

• Composite section can be designed to address loading 

conditions and/or specific site requirements.

• Considerable length can be provided at comparatively 

low cost for wood composite piles.

• High factored resistance for some composite piles. 

• Internal inspection for pipe composite piles.

DISADVANTAGES 

• More complex pile fabrication.

• Difficult to attain good joints between two materials 

except for concrete and H or pipe composite piles.

REMARKS • The weakest of any material used will govern the 

structural design and factored resistance. 

Figure 6-8 Composite piles typical illustration. 
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6.2 TIMBER PILES

Timber piles are usually of round, tapered cross section made from tree trunks of 

Southern Pine or Douglas Fir driven with the smaller end used as the pile toe. 

Southern Pine timber piles can be found to lengths up to 75 feet, and some west 

coast Douglas fir may be up to 120 feet in length. Oak and other timber types have 

also been used for piles, but their use is infrequent today. ASTM D25, Standard 

Specification for Round Timber Piles, presents guidelines on minimum timber pile 

dimensions, straightness, knot sizes, etc. AWPA C3, Piles - Preservative Treatment 

by Pressure Process, contains penetration and retention values for the various 

preservatives. Figure 6-9 presents a photograph of timber piles.

Figure 6-9 Timber piles.

Timber piles are best suited to support modest loads as friction piles in sands, silts 

and clays. The taper of timber piles is effective in increasing the shaft resistance, 

particularly in loose sands. They are not recommended as piles to be driven through 

dense gravel, boulders, or till, or for toe bearing piles on rock since they are 

vulnerable to damage at the pile head and toe in hard driving. Overdriving of timber 

piles can result in the crushing of fibers or brooming at the pile head. This can be 

controlled by using a helmet with cushion material and/or metal strapping around the 

head of the pile. In hard driving situations, a metal shoe should be attached to the 

pile toe. Timber piles are favored for the construction of bridge fender systems and 

small jetties due to the good energy absorption properties of wood.
165

Timber pile splices are difficult and undesirable. AASHTO (2010) LRFD Bridge 

Construction Specifications state that timber piles should not be spliced unless 

specified in the contract documents and approved by the engineer.

Durability is generally not a design consideration if a timber pile is below the 

permanent water table. However, when a timber pile is subjected to alternate 

wetting and drying cycles or located above the water table, damage and decay by 

insects may result. Bacteria and fungi attack can also result in pile damage and 

decay. Such damage reduces the service life of timber piles significantly unless the 

pile is treated with a wood preservative. The most common treatments for timber 

piling are Chromated Copper Arsenate (CCA) for Southern Pine, and Ammoniacal 

Copper Zinc Arsenate (ACZA) for Douglas Fir. Creosote remains a common 

treatment option in some areas of the United States but prohibited in others. 

Creosote cannot be used alone in southern waters due to attack by limnoria 

tripunctata, but should be used as part of a dual treatment with CCA or ACZA. If 

cracking of the pile shaft or head occurs and extends below the prescribed pile cutoff level, the initial preservative treatment will not be effective, and the trimmed end 

of the pile should be treated a second time.

According to Graham (1995), the durability of round timber piling is a function of sitespecific conditions:

1. Foundation piles permanently submerged in ground water will typically last 

indefinitely.

2. Fully embedded, treated foundation piles partially above the ground water 

with a concrete cap will typically last on the order of 100 years or longer.

3. Treated trestle piles over land will generally last as long as utility poles in the 

area, i.e., about 75 years in northern areas and about 40 years in southern 

areas of the United States.

4. Treated piles in fresh water will typically last about five to ten years less than 

land trestle piles in the same area.

5. For treated piles in brackish water, the longevity should be determined by the 

experience in the area.

6. Treated marine piles will typically last about 50 years in northern climates and 

25 years in southern climates of the United States.
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6.3 STEEL H-PILES

Steel H-piles consist of rolled wide flange sections that have flange widths 

approximately equal to the section depth. In most H-piles sections, the flange and 

web thicknesses are the same. They are manufactured in standard sizes ranging 

from 8 to 18 inches. In some cases, W-sections have also been used for piles. 

However, this is generally customized on a project to project basis. Figure 6-10

contains a photograph of H-piles with driving shoes.

Figure 6-10 H-piles with driving shoes. 

H-piles produced today meet the requirements of ASTM A572, Grade 50 steel, as 

ASTM A36 steel H-piles are no longer readily available. Steel sections meeting the 

requirements of ASTM A588 and ASTM A690 are also available. These are high 

strength, low alloy steels developed for improved corrosion resistance in 

atmospheric (ASTM A588) and marine (ASTM A690) environments. However, 

ASTM A588 and A690 steels are typically hard to obtain, and long lead times may 

be necessary if they are specified. ASTM A572, A588 and A690 are all Grade 50 

steels. Some H-piles can also be obtained in A572 Grade 60 steel. Therefore, it is 

possible to use the higher strength of Grade 50 or 60 steel if the pile can be installed 

to maximize the geotechnical resistance in the project soil conditions. Steel H-piles 
167

are very effective when driven into soft rock. They can be driven very hard with 

modern high impact velocity hammers with little likelihood of pile toe damage.

H-piles can develop their nominal resistance through shaft resistance, toe resistance 

or a combination of both. Since H-piles generally displace a minimum amount of 

soil, they can be driven more easily through dense granular layers and very stiff 

clays than displacement piles. In addition, problems associated with soil heave 

during foundation installation are often reduced by using H-piles. However, 

sometimes H-piles will "plug". That is, the soil being penetrated will adhere to the 

web and the inside flange surfaces creating a closed end, solid section. The pile will 

then drive as if it were a displacement pile below the depth of plug formation. 

Plugging can have a substantial effect on both the soil resistance during driving and 

the nominal geotechnical resistance. H-piles can be problematic when used as 

friction piles in some granular deposits. In these conditions they often don’t plug 

during driving, have low dynamic resistances during installation, and result in 

excessive pile lengths.

Experience indicates that corrosion is not a practical problem for steel piles driven in 

natural soil, due primarily to the absence of oxygen in the soil. However, in fill 

materials at or above the water table, moderate corrosion may occur and protection 

may be needed. In 2013, an H-pile supported pier at the I-43 Leo Frigo Bridge 

suddenly settled due severe corrosion. While the concurrent factors that lead to the 

corrosion were considered highly unusual, this occurrence emphasizes the need for 

identifying corrosive environments during the design stage. As noted previously, 

high strength, low allow steels are available for improved corrosion resistance. 

Another common protection method requires the application of pile coatings before 

and after driving. Coal-tar epoxies, fusion bonded epoxies, metallized zinc, 

metallized aluminum and phenolic mastics are some of the pile coatings available. 

Encasement by cast-in-place concrete, precast concrete jackets, or cathodic 

protection can also provide protection for piles extending above the water table. 

Another design option for piles subject to corrosion is to select a heavier section 

than that required by the design loads, anticipating the loss of material caused by 

corrosion. Corrosion losses can be estimated using the information provided in 

Section 6.14.1. However, even with the corrosion protection options or allowances 

mentioned above, certain aggressive soil conditions will preclude the use of steel 

piles altogether.

One of the key advantages of H-piles is the ease of extension or reduction in pile 

length. This makes them suitable for nonhomogeneous soils with layers of hard 

strata or natural obstructions. Splices are commonly made by full penetration 
168

groove welds so that the splice is as strong as the pile in both compression and 

bending. The welding should always be done by qualified welders in accordance 

with approved procedures. Proprietary splices are also commonly used for splicing 

H-piles and Chapter 16 presents additional information on typical splices. A steel 

load transfer cap is not required by AASHTO if the pile head is embedded 12 inches 

into the concrete pile cap. Pile toe reinforcement using commercially manufactured 

pile shoes is recommended for H-piles driven through or into very dense soil or soil 

containing boulders or other obstructions. Pile shoes are also used for H-piles 

driven to rock, particularly for penetration into sloping rock surfaces. Chapter 16 

provides details on available driving shoes.

The disadvantages of H-piles include a tendency to deviate when natural 

obstructions are encountered. Nominal resistance verification of H-piles used as 

friction piles in granular soils based on the observed blow count can also be 

problematic, and can result in significant length overruns. An H-pile in a granular 

profile will often not plug during the dynamic loading of pile installation but may plug 

under the slower static loading condition. Length for length, steel piles tend to be 

more expensive than concrete piles. On the other hand, steel's high factored 

resistance for a given weight can reduce pile driving costs.

6.4 STEEL PIPE PILES

6.4.1 Closed End Steel Pipe

Closed end steel pipe piles consist of seamless, welded or spiral welded steel pipes 

in diameters typically ranging from 8 to 30 inches. Larger pipe pile sizes are 

available, but larger diameters are more commonly driven open ended. Typical wall 

thicknesses for closed end pipe piles range from 0.188 to 1 inch. Pipe piles should 

be specified by grade with reference to ASTM A252, API-5L, or API-2B. In some 

situations, a contractor may propose to supply used pipe not produced under ASTM 

standards. Used pipe piles not meeting ASTM standards must be evaluated by an 

engineer for general condition, drivability, and weldability prior to approval.

A closed end pipe pile is generally formed by welding a 0.75 to 2 inch thick flat steel 

plate to the pile toe. The toe plate thickness generally increases with pile diameter 

and/or with anticipated harder driving conditions. When pipe piles are driven to 

weathered rock or through boulders, a cruciform-reinforced end plate or a conical 

point with rounded nose is often used to minimize pile toe distortion. Figure 6-11

presents a picture of a typical closed end pipe pile with a flat closure plate.
169

Figure 6-11 Typical 16 inch closed end pipe pile. 

Closed end steel pipe piles can be used as friction piles, toe bearing piles, a 

combination of both, or as rock socketed piles. They are commonly used where 

variable pile lengths are required since splicing is relatively easy. With the increased 

ductility requirements for earthquake resistant design, pipe piles are being used 

extensively in seismic areas.

Pipe piles may be left open or filled with concrete. If concrete filled, the piles can 

also have a reinforcing steel cage or structural shape such as an H-section inserted 

into the concrete. Reinforcing steel is required only when the concrete in the pile 

may be under tension from such conditions as uplift, high lateral loads, or for 

unsupported pile lengths.

Most often, pipe piles are driven from the pile head. However, thin wall, closed end 

pipe piles can also be bottom driven using a mandrel. A mandrel is usually a heavy 

tubular steel section inserted into the pile that greatly improves pile drivability. After 

driving, the mandrel is removed and the pile is inspected internally before concrete is 

placed. Typically, pipe piles are spliced using full penetration groove welds. 

Proprietary splicing sleeves are available and should be used only if the splice can 

provide full strength in bending and tension (unless the splice will be located at a 

distance below ground where bending moments and tension loads are small). 

Typical pile splices are described in Chapter 16. The corrosion discussion on Hpiles is also applicable to steel pipe piles.
170

6.4.2 Open End Steel Pipe

Open end pipe piles are frequently installed when hard driving conditions, significant 

penetration depths, or debris is expected. Open end pipe piles are available in 

diameters that range from 8 to 160 inches. When pipe piles are driven open ended, 

wall thicknesses of 0.5 inches or greater are commonly used. Similar to closed end

pipe piles, the open end piles can be seamless, rolled and welded, or spiral welded 

steel pipes. Open end pipe piles should be specified by grade with reference to 

ASTM A252, API-5L, or API-2B. In some situations, a contractor may propose to 

supply used pipe not produced under ASTM standards. Used pipe piles not meeting 

ASTM standards must be evaluated by an engineer for general condition, drivability, 

and weldability prior to approval.

Large diameter open end pipe piles are defined as greater than 36 inches in 

diameter (Brown and Thompson 2015). These are often selected due to their ability 

to resist significant vessel impacts and overturning moments, and/or due to scour, 

ice, or seismic design considerations. Other common applications of large diameter 

open end pipe include fender systems, mooring dolphins, and offshore facilities. 

Large diameter open end pipe piles are frequently used to support main piers on 

major river bridges or other structures with large lateral resistance demands. Large 

diameter open end pipes are typically driven to bear in dense sand or soft rock, or 

are installed as long friction piles. Figure 6-12 shows a group of 42 inch diameter, 

spiral weld, open end pipe piles for a main bridge pier support. A photo of a 12 foot 

diameter open end pipe is presented in Figure 6-13. 

Open end pipe piles can be socketed into bedrock (rock socketed piles), or when 

driving through dense materials, may form a soil plug. The plug makes the pile act 

like a closed end pile and can significantly increase the pile toe resistance. Plugging 

is discussed in greater detail in Section 7.10.7 and is a complex phenomenon. The 

soil plug should not be removed unless the pile is to be internally cleaned out and 

filled with concrete or unless the soil plug is preventing the pile from achieving the 

required penetration depth.

Large diameter open end pipe typically core through the soil during driving due to 

mass soil inertia effects. However, under static loading conditions, these piles 

generally plug providing significant toe bearing resistance. Some recent large 

diameter open end pipe pile projects have incorporated a constrictor plate inside the 

pile to force plugged behavior during driving. A diagram of a constrictor plate design

and a companion photograph are presented in Figures 6-14 and 6-15, respectively. 

Constrictor plates are also discussed in Section 16.3.3.
171

Figure 6-12 42 inch diameter, spiral weld, open end, pipe for main river pier. 

 

Figure 6-13 Large diameter open end pipe piles. 

Open end pipe piles are seldom cleaned out full length unless a rock socket is 

planned or short pile lengths are used. Before concrete placement, steel 

reinforcement and uplift resisting dowels can be added, as necessary.
172

Open ended piles can also be equipped with internal or external steel cutting shoes 

to reduce the potential for toe damage. When hard driving conditions are expected 

such as sloping rock, cobbles and boulders, etc., large diameter open end pipes are 

frequently designed with a thicker wall pipe section over the bottom two diameters to 

lessen the risk of toe damage from high localized stresses.

Figure 6-14 Constrictor plate. 

Figure 6-15 Constrictor plate installed inside pipe pile.
173

6.5 MONOTUBE PILES

The Monotube pile, shown in Figure 6-16, is a proprietary pile driven without a 

mandrel. Monotubes are longitudinally fluted and are tapered over the lower pile 

section. Depending upon the pile section selected nominal pile toe diameters of 8 or 

8.5 inches taper to a butt diameter of 12, 14, 16, or 18 inches over tapered section 

lengths of 10 to 75 feet. Non-tapered extensions are available in 20 and 40 feet for 

splicing to the lower tapered section. Monotube piles are available in 9 to 3 gage 

shell thicknesses or roughly 0.15 to 0.24 inches. Monotube sections are spliced by 

inserting the slightly smaller bottom section of the extension into the top of the 

previously driven section. A fillet weld is then used along the interface between the 

extension and the lower pile section into which it is inserted. Four V shaped notches 

cut into the lower pile section at 90 degree locations are often added to increase 

weld length and splice strength. After driving, Monotube piles are filled with 

concrete.

The fluted and tapered design of Monotube piles has several functional advantages. 

The flutes add stiffness necessary for handling and driving lightweight piles. These 

also increase the surface area while the tapered section improves the soil resistance 

per unit length in compression loading. The flutes are formed by cold working when 

the pile is manufactured as this increases the yield point of the steel to more than 50 

ksi, further improving the pile drivability.

Figure 6-16 Tapered Monotube section (right) with add-on sections (left).
174

6.6 TAPERTUBE PILES

This pile consists of a tapered, 12 sided polygon over the lower section with 

conventional steel pipe pile material as the upper add-on sections. The 15 to 30 feet 

long tapered section steel is available with pile toe diameters ranging from 8 to 

14 inches and pile head diameters of 12 to 24 inches. Wall thickness of the tapered 

sections ranged from 0.1875 to 0.625 inches. The tapered tube bottom section has 

a yield strength of 50 ksi, and the upper pipe pile sections conform to ASTM A252 

Grade 3 steel with a yield strength of 45 ksi. The tapered and pipe sections are 

connected using a full penetration weld.

Tapertube piles are driven from the top and filled with concrete after driving. 

Specialty sleeve splices may be used as an alternative to welding sections together. 

Please see Chapter 16 for more information. A photo of Tapertube piles is 

presented in Figure 6-17. 

Figure 6-17 Tapertube piles (courtesy DFP Foundation Products, LLC).

6.7 SPIN FIN PILES

The Spin Fin pile is a variation of a pipe pile introduced on the west coast in 1983. It 

is a pipe pile with an outside “thread” made of fins that gradually wind around a 

bottom portion of the pile. These fins cause the pile to rotate into the ground during 

driving. Following driving, the pile is incorporated into a pile cap. Pile rotation is 
175

then restrained by the cap preventing the pile from twisting. This results in a 

plugging effect that increases the pile's resistance to tension loads as depicted in 

Figure 6-18. The Spin Fin pile is particularly attractive on projects needing 

increased uplift resistance such as for seismic events or in soils with limited 

overburden materials overlying a hard bearing layer such as a glacial till over 

bedrock. The fins also increase the compression resistance by increasing the 

bearing area in the pile section having fins. A photograph of a 30 inch diameter spin 

fin pile is presented in Figure 6-19. 

Figure 6-18 Spin Fin pile load transfer illustration. 
176

Figure 6-19 30 inch diameter Spin Fin pile. 

6.8 PRESTRESSED CONCRETE PILES

Prestressed concrete piles vary from the most common solid square section to a 

solid octagonal section. As sections increase in size, they are often cast with an 

internal void to reduce handling weight. Prestressed piles can either be pretensioned or post-tensioned. Pre-tensioned piles are usually cast to their full length 

in permanent casting beds. Post-tensioned piles are usually manufactured in 

sections, most commonly cylindrical, and assembled and prestressed to the required 

pile lengths at the manufacturing plant or on the job site. Figure 6-20 depicts 

common prestressed concrete pile sections and Figure 6-21 presents a photograph 

of prestressed piles stored on a barge prior to driving.

The prestressing steel may be in the form of strands or wires which are enclosed in 

a conventional steel spiral and placed in tension. Prestressing steel must conform to 

ASTM A416, A421, and A882. Due to the effects of prestressing, these piles can 

usually be made lighter and longer than reinforced concrete piles of the same size. 

In cases of extreme environmental conditions, an epoxy coating has been used on 

prestressing strands. If this coating is used, it should be dusted with sand before the 

epoxy sets. Then the strand will have sufficient bond strength to carry the prestress 
177

development bond stresses. If an epoxy coating has been used on the strand, it 

should also be used on the tie or spiral reinforcement. However, epoxy coating is 

generally not necessary for prestressed piles since the prestressing force will keep 

the concrete in compression making deterioration less likely.

Figure 6-20 Typical prestressed concrete piles. 

Figure 6-21 Square prestressed concrete piles. 

The primary advantage of prestressed concrete piles compared to conventional 

reinforced concrete piles is durability. Since the concrete is under continuous 

compression, hairline cracks are kept tightly closed and thus prestressed piles are 

usually more resistant to weathering and corrosion than conventionally reinforced 

piles. This characteristic of prestressed concrete removes the need for special steel 

coatings since corrosion is not as serious a problem as for reinforced concrete. 
178

Another advantage of prestressing is that the tensile stresses which can develop in 

the concrete under certain driving and handling conditions are less critical.

Prestressed concrete piles are more vulnerable to damage from striking hard layers 

of soil or obstructions during driving than reinforced concrete piles. This is due to 

the decrease in axial compression strength due to the application of the prestressing 

force. When driven in soft soils, care must also be used since large tension stresses 

can be generated in easy driving.

Prestressed concrete piles cutoff and splicing problems are considered much more 

serious by contractors that drive them infrequently than by those that drive only this 

pile type. Special reinforcement required at the pile head in seismic areas can pose 

problems if actual lengths vary significantly from the planned length. In these cases, 

a splice detail must be included so that the seismic reinforcement is extended into 

the pile cap.

Concrete used for most prestressed concrete piles typically has a 28 day 

compressive strength between 5 and 6 ksi. Recent developments in prestressed 

concrete piles include studies on the use of high performance concrete (e.g., Moser 

2011) to improve pile durability, or ultra-high performance concrete (e.g., Vande 

Voort et al. 2008) for improved durability and load support relative to steel H-piles. 

Belk (2013) studied the use of lightweight aggregate on prestressed concrete piles 

relative to drivability, load support and weight. At the present time, these 

developments have not progressed into mainstream practice.

6.9 CONCRETE CYLINDER PILES

Concrete cylinder piles include spun cast, non-spun cast, and Industrial Concrete 

Products pile types. Each of these cylinder piles types is discussed further below. 

Cylinder piles are sometimes difficult to drive. However, they usually extend directly 

to the superstructure support level avoiding the need for a pile cap, which can result 

in substantial cost savings. Jetting is often used to install cylinder piles to the near 

the desired depth followed by impact driving to attain the required nominal 

resistance. When jetting is used, it must be controlled to minimize degradation of 

the lateral soil resistance and disturbance and removal of soils below the specified 

jetting elevation. The design of the pile wall often includes periodic vent holes that 

are used to reduce the build-up of internal water pressure during driving as well as to 

later avoid build-up of internal gas pressure from organic soil decomposition during 

service life.
179

6.9.1 Spun-Cast Cylinder Piles

Spun-cast concrete cylinder piles are hollow concrete piles which are cast in 8, 12, 

or 16 foot sections. Depending on the manufacturer, cylinder piles are generally 

available in diameters of 36, 42, 54 and 66 inches. Wall thicknesses of 5, 6, or 6.5 

inches are available depending on the pile diameter. Cylinder pile sections are spun 

centrifugally during the casting process to obtain a high density, durable concrete 

that is virtually impervious to moisture. The concrete quality from the spun-casting 

process is unique to spun-cast cylinder piles and concrete compressive strengths of 

8 ksi can routinely be achieved. Once cured, sections are assembled to form a pile 

with the desired pile length. An adhesive joint compound is used between sections. 

The assembled pile is post tensioned and then the post tensioning ducts grouted.

Results of chloride ion penetration and permeability tests on prestressed cylinder 

piles indicate that the spun-cast cylinder piles have excellent resistance to chloride 

intrusion. The post-tensioning process results in a typical prestress level of 1.5 ksi. 

Figure 6-22 shows the typical configuration of this cylinder pile type. A photograph 

of concrete cylinder piles is presented in Figure 6-23. 

Figure 6-22 Typical spun-cast concrete cylinder pile section. 

Generally spun-cast cylinder piles are used for marine structures or land trestles and 

have high resistance to corrosion and high lateral resistance. To prevent freezethaw degradation, air entrainment and adequate spiral reinforcement are important 
180

considerations. The piles typically extend above ground and are designed to resist a 

combination of axial loads and bending moments. Additional design and installation 

considerations associated with concrete cylinder piles are summarized in Hartman et 

al. (2007).

Figure 6-23 Concrete cylinder pile. 

6.9.2 Bed-Cast Cylinder Piles

Cylinder piles can also be cast in a bed with forms rather than centrifugally spun 

cast. These piles are produced to the required length in a single piece and are 

pretensioned instead of being post tensioned like the spun-cast piles. Due to the 

differences in the casting process, these piles do not have the high density, low 

porosity concrete that is characteristic of spun-cast cylinder piles and will therefore 

not have the same resistance to chloride intrusion. Bed-cast cylinder piles have 

typical specified concrete compressive strengths of 5.5 ksi, a diameter of 60 inches, 

and a wall thickness of 7.5 inches.

6.9.3 Industrial Concrete Products (ICP) Piles

These cylindrical piles are pretensioned and spun to compact concrete while curing. 

The spinning process creates a high density mix by forcibly removing water during 

the hydration process while pretension adds tensile strength. Typically, ICP piles 

are cast in sections with steel end plates. Field splices are made by welding the 

section end plates. Sizes range from 10 inch diameter with a 2.25 inch thick wall to 

a 48 inch diameter with a 5.9 inch wall. ICP section lengths vary with the pile 

diameter from 20 to 39 feet on the smallest diameter section up to 33 to 138 feet on 

the largest diameter section. Piles are supplied open end with a flat shoe or with an 

X-pointed shoe.
181

6.10 COMPOSITE PILES

In general, a composite pile is made up of two or more sections of different materials 

or different pile types. Depending upon the soil conditions, various composite 

sections may be used. The upper pile section is often precast concrete, steel pipe, 

or corrugated shell. The lower pile section may consist of steel H, steel pipe, or 

timber pile. Composite piles have limited application and are generally used only 

under special conditions. Some of the more common composite piles are discussed 

below.

6.10.1 Precast Concrete - Steel H-pile Composite Piles

One of the more commonly used composite piles consists of a lower section of steel 

H-pile or pipe pile embedded in an upper pile section of precast concrete. These 

concrete-steel composite piles are often used when scour or uplift requirements 

dictate pile penetration depths that a displacement pile cannot achieve, in 

subsurface conditions where surficial soil layers have high corrosion potential, or in 

other conditions that dictate their use. A photograph of composite square concrete 

piles each with H-pile stinger is presented in Figure 6-24. 

Figure 6-24 Precast concrete piles each with H-pile stinger.
182

6.10.2 Steel Pipe - H-pile Composite Piles

Steel pipe - H-pile composite piles are comprised of a steel pipe upper section with a 

transition to an H-pile lower section. This composite selection may be applicable 

where lateral load demands require the bending resistance of a pipe pile in the 

upper pile length and dense soil conditions necessitate a low displacement pile over 

the lower pile length to satisfy minimum pile penetration depth or bearing layer 

requirements. A steel pipe - H-pile composite piles are shown in Figure 6-25. 

Figure 6-25 Steel pipe - H-pile composite piles.

6.10.3 Corrugated Shell - Timber Composite Piles

Corrugated shell - timber composite piles are sometimes used as foundation piles. 

For this composite pile the timber pile section is permanently located below the 

groundwater level. A concrete filled shell pile is used above the treated or untreated 

timber pile. In the case of the composite corrugated shell - timber pile, the timber 

pile is driven below the water table. A corrugated steel shell is connected to the pile 
183

head of the timber section with a wedge ring driven into the wood. The shell is filled 

with concrete to the cutoff elevation and the pile is complete.

6.10.4 Corrugated Shell - Pipe Composite Piles

This composite pile consists of a pipe pile for the lower section and a corrugated 

shell for the upper portion of the pile. A variety of pipe and shell diameters can be 

used to accommodate a range of loading conditions. The corrugated shell - pile 

composite pile is mandrel driven. The mandrel provides a guide for alignment of the 

two pile sections provided it extends to the pipe pile head or partially into the pipe 

pile. Possible pile joints include: a sleeve joint, a welded joint, and a drive-sleeve 

joint. Once the pipe and shell are driven and connected, they are filled with concrete 

to cutoff grade and any excess shell is removed.

6.11 PILE TYPES INFREQUENTLY USED ON TRANSPORTATION PROJECTS

Additional pile types exist that are sometimes used in specialty applications. Others 

were used more frequently in the past and could be encountered if existing 

foundations are re-used on reconstruction projects. These include Fundex piles, 

Tubex piles, mandrel driven piles, reinforced concrete piles, and pressure injected 

footings. Experience and field testing (load and integrity tests) should be used in 

new designs using these systems or in re-use of existing foundations.

6.11.1 Fundex Piles

The Fundex pile is a unique form of a pipe-cased, cast-in-place concrete 

displacement pile. Instead of the pile being driven into the ground with a hammer, it 

is screwed into the ground with a special 18 inch diameter boring tip. The boring tip 

is fitted to a 14 inch diameter drilling mandrel with a chuck assembly. A drill table 

forces the drilling mandrel into the ground utilizing a constant vertical load and 

torque. When the bearing layer or required penetration depth is reached, a 

reinforcing cage is suspended within the mandrel and concrete is placed. The 

drilling mandrel is then oscillated out of the ground leaving the drill point behind. 

This process results in a cast-in place pile typically installed for unfactored design 

loads in the 40 to 125 ton range.

Some of the advantages of the Fundex piles include minimal vibrations and very low 

noise during installation, drilling equipment that can be used in confined places, and 

a removable mast that allows installation with only 20 feet of overhead clearance.
184

6.11.2 Tubex Piles

Tubex piles are installed similarly to Fundex piles. A pipe-cased hole is created for 

the Tubex pile as the drill point is advanced into the soil. Pipe sections may be 

spliced if additional length is required or if required by headroom limitations. Since 

the steel casing is in place after drilling to the final installation depth, an internal 

inspection can also be performed prior to concrete and reinforcement placement. 

However, unlike the Fundex pile, this casing is then left in place along with the drill 

point. A grout injected version of the Tubex pile is also available. Typically Tubex 

pile diameters are 14 inches or greater. If grouted, this pile is typically installed for 

unfactored design loads of up to 230 tons.

Some of the advantages of the Tubex piles include minimal vibrations and very low 

noise during installation, drilling equipment that can be used in confined places, and 

a removable mast that allows installation with only 20 feet of overhead clearance. In 

addition, the grout injected Tubex pile can be used to insulate the steel casing from 

corrosive environments. 

6.11.3 Pressure Injected Footings (PIF)

This type of driven, cast-in-place pile is often referred to as a Franki pile or pressure 

injected footing. The best site conditions for these piles are loose to medium dense 

granular soils.

This pile type is installed by bottom driving a temporary steel casing into the ground 

using a drop weight driving on a zero slump concrete plug at the bottom of the 

casing. When the required depth has been reached, the steel casing is restrained 

from above and the concrete plug is driven out the bottom of the tube. An enlarged 

base is formed by adding and driving out small batches of zero slump concrete.

Steel reinforcing is then installed prior to adding more concrete to the shaft. It is 

suggested that widely spaced bars be used to allow the low workability mix to 

penetrate to the exterior of the piles. After the base is formed and reinforcement is 

placed, concrete continues to be added and the uncased shaft is formed by 

compacting the concrete with a drop weight in short lifts as the casing is being 

withdrawn. Alternatively, if a high workability mix is used to complete the pile, a 

vibrator can be clamped to the top of the tube and used to compact the concrete into 

place as the casing is withdrawn.
185

6.11.4 Mandrel Driven Piles

In the past, mandrel driven piles were used in many soil conditions except where 

obstacles such as cobbles and boulders were present that could damage the thin 

shells during driving. The thin shells were susceptible to collapse under hydrostatic 

pressure prior to concrete placement. They were best suited for friction piles in 

granular material. Thin shell mandrel driven piles are rarely used today for new 

construction. They may however be encountered when widening or rehabilitating 

existing structures so a general overview of their characteristics remains useful.

The pile shells for mandrel driven piles were generally produced from sections of

corrugated steel and were either of constant diameter, steadily decreasing in 

diameter from the pile head to the pile toe, or diameter decreasing in discrete steps 

over the pile length. Typical tapers were on the order of 1 inch per 8 foot length. It 

was also possible to have different lengths for each section. Separate shell sections 

were usually screw-connected and waterproofed with an O-ring gasket. The 

Raymond Step Taper, Armco Hel-Cor, Republic Corwel, and Guild pile were among 

the pile types previously driven with mandrels. However most of these corrugated 

shell type piles are no longer manufactured. Thin wall pipe piles have also been 

mandrel driven with the mandrel driving on a reinforced section at the pile toe.

The properties of the reusable mandrels dictated the drivability of these shell or thin 

wall pipe pile sections. This resulted in a significant cost advantage for a mandrel 

driven pile since the mandrels result in improved pile drivability and soil resistance at 

low material costs. Construction control of mandrel driven piles should include a 

wave equation analysis that accounts for the improved pile drivability from the 

mandrel. A dynamic formula should not be used for construction control of mandrel 

driven piles. Mandrel driven piles may be costly if it is necessary to drive piles to an 

unanticipated depth that exceeds the mandrel length available at the job site.

6.11.5 Reinforced Concrete Piles

Prestressed concrete piles have replaced reinforced concrete piles in the U.S. 

market. Reinforced concrete piles were manufactured from concrete and had 

reinforcement consisting of a steel cage made up of several longitudinal bars and 

lateral or tie steel in the form of individual hoops or a spiral. While they are no 

longer used in the U.S., a limited discussion on them is presented in case they are 

encountered on a rehabilitation or widening project. Steel reinforcing for reinforced 

concrete piles was governed by ASTM A82, A615, and A884. High yield strength 

steel reinforcement to resist uplift loads had to conform to ASTM A722.
186

Reinforced concrete piles were more susceptible to damage during handling and 

driving because of tensile stresses compared to prestressed piles. Advantages of 

reinforced concrete piles included their lower net compressive stress during driving 

and under foundation loads, and a reduced danger of pile head cracking. In 

addition, these piles were easier to splice than prestressed piles. To reduce 

corrosion of the reinforced concrete joints, splices were located below the ground 

surface, or if under water, the mudline. Segmental pile sections were used to 

produce piles with varied lengths to accommodate variable soil conditions, and were 

easily transported to job sites.

The most common type of jointed pile was a square cross section made of high 

density concrete with each successive unit of shorter length. Typical pile cross 

sections ranged from 10 to 16 inches, but sizes above and below this range were 

produced. Joints between these pile sections were of the mechanical type, including 

bayonet fittings or wedges. The joints had to be well aligned or energy was lost 

during driving and bending stresses would be introduced due to an eccentric 

connection. These piles were best suited for friction piles in sand, gravel and clay.

Another jointed reinforced concrete pile type utilized a hexagonal section. The 

advantages of this cross sectional shape were an improved stress distribution over 

the pile section and an improved resistance to torsional loading.

Special precautions had to be taken when placing piles during cold weather. If piles 

were driven through ice and water before reaching soil, the air and concrete may 

have been at low temperatures relative to the soil and water. Such temperature 

gradients could cause concrete to crack due to non-uniform shrinkage and 

expansion. Although most reinforced concrete piles were jointed, there were 

occasions when non-jointed piles were more economical due to the cost of pile 

segments. Often for a very large project when thousands of piles were used, the 

piles were economically cast on site. Most non-jointed piles had a square cross 

section and were difficult to change in length. Only a few splicing procedures 

existed if a situation arose where a reinforced concrete pile needed to be 

lengthened. The first method of pile lengthening involved the breakdown of the 

projecting pile head to provide a suitable lap for reinforcing steel. Concrete was cast 

to form a joint. A second option was to butt the two piles together within a steel 

sleeve, and use an epoxy cement to join the two piles. The last lengthening method 

involved the use of dowel bars to be inserted into drilled holes with epoxy cement to 

form the joint. If piles were lengthened, the connecting pile sections had to be 

carefully aligned, since excessive bending stresses would result if any eccentricity 
187

existed. Splicing problems tended to become less severe or even non-existent 

when contractors developed experience and techniques. Special reinforcement 

required at the pile head in seismic areas posed problems if actual lengths varied 

significantly from the planned length. In these cases, a splice detail had to be 

included so that the seismic reinforcement was extended into the pile cap.

Reinforced concrete piles are no longer used in the United States. However, they 

are routinely used in Europe, Australia, and many Asian countries for economic 

reasons.

6.12 DESIGN CONSIDERATIONS IN AGGRESSIVE SUBSURFACE 

ENVIRONMENTS

For every design, consideration should be given to the possible deterioration of the 

pile over its design life due to the surrounding environment. This section will 

address design considerations in aggressive subsurface environments where 

corrosion, chemical attack, abrasion, and other factors can adversely affect pile 

durability after installation. An assessment of the in-situ soil conditions, fill materials, 

and groundwater properties is necessary to completely categorize an aggressive 

subsurface condition.

An aggressive environment can generally be identified by soil resistivity and pH 

tests. If either the pH or soil resistivity tests indicate the subsurface conditions are 

aggressive, then the pile selection and foundation design should be based on an 

aggressive subsurface environment. The design of pile foundations in an 

aggressive environment is a developing field. Therefore, a corrosion/degradation 

specialist experienced in underground corrosion should be retained for major 

projects with pile foundations in aggressive environments.

Whenever the pH value of the soil or water is less than 4.5, the foundation design 

should be based on an aggressive subsurface environment. Alternatively, if the 

resistivity is less than 2000 ohms-cm the site should also be treated as aggressive. 

When the soil resistivity test results are between 2000 and 5000 ohms-cm then 

chloride ion content and sulfate ion content tests should be performed. If these test 

results indicate a chloride ion content greater than 100 parts per million (ppm) or a 

sulfate ion content greater than 200 ppm, then the foundation design should be 

based on an aggressive subsurface environment. Resistivity values greater than 

5000 ohms-cm are considered non-aggressive environments. Electro chemical 

classification tests for aggressive environments are described in Chapter 5.
188

Contaminated soil and groundwater can cause significant damage to foundation 

piles in direct contact with the aggressive chemicals. Acidic groundwater is common 

at sites with organic soils, industrial contamination, or mine runoff. The subsurface 

exploration program should indicate if the soil or groundwater is contaminated. If 

industrial contamination is found, the maximum likely concentrations should be 

determined as well as an estimate of the lateral and vertical extent of the 

contamination.

6.12.1 Corrosion of Steel Piles

Steel piles driven through contaminated soil and groundwater conditions may be 

subject to high corrosion rates and should be designed appropriately. Corrosion of 

steel or steel reinforced piles may also occur if piles are driven into disturbed 

ground, landfills or cinder fills, or low pH soils. Corrosion should also be evaluated 

for piles located in a marine environment, or if piles are subject to alternate wetting 

and drying from tidal action. Corrosion rates are a function of the ambient 

temperature, pH, access to oxygen, and chemistry of the aqueous environment 

surrounding the steel member.

6.12.1.1Corrosion in Non-Marine Environments

AASHTO Standard R 27-01 (2010) provides a recommended assessment procedure 

for evaluating corrosion of steel piling in non-marine applications. This 

recommended procedure consists of a Phase I and Phase II assessment. In the 

Phase I assessment, information on the location of the pile cap relative to the 

groundwater table, the soil characteristics, and soil contaminants is obtained. This 

information is used to determine if a Phase II assessment is required.

If the pile cap is at or above the water table, a Phase II assessment is performed to 

evaluate the corrosivity of the site. The Phase II assessment consists of collecting 

continuous soil samples to a depth of 3 feet below the water table and conducting 

laboratory tests on the recovered samples. The site sampling and testing protocol is 

outlined in Figure 6-26. After collecting the necessary information, the possibility of 

uniform or macro cell corrosion is evaluated using the flow chart presented in Figure 

6-27. The final step in the evaluation process includes determining the necessity for 

electrochemical testing, corrosion monitoring, and mitigation techniques. A flow 

chart of this process is presented in Figure 6-28. 
189

Figure 6-26 Soil sampling and testing protocol for corrosion assessment of steel 

piles in non-marine applications. 
190

Figure 6-27 Procedure for uniform or macrocell corrosion assessment of steel piles 

in non-marine applications. 
191

Figure 6-28 Procedure for determination of electrochemical testing, corrosion 

monitoring and corrosion mitigation techniques. 
192

It should be noted that the flow charts do not cover all possibilities for corrosion of 

steel piling at a site. Factors not addressed include chemical contamination, stray 

DC currents, and the presence of high concentrations of microbes. When these 

conditions are present on a project, an underground corrosion specialist should be 

consulted.

For steel piles buried in fill or disturbed natural soils, a conservative estimate of the 

corrosion rate is 0.003 inches per year. Morley (1979) reported corrosion rates of 

0.002 inches per year for steel piles immersed in fresh water, except at the waterline 

in canals where the rate was as high as 0.013 inches per year. The high rate of 

corrosion at the water line was attributed to debris abrasion and/or cell action 

between other parts of the structure.

AASHTO Standard R 27-01 (2010) should be consulted for a detailed step by step 

procedure of corrosion evaluation process and estimation of remaining service life. 

Additional insight into the corrosion of steel piles in non-marine environments is also 

presented in NCHRP Report 408 by Beavers and Dunn (1998).

6.12.1.2Corrosion in Marine Environments

For steel piles in marine environments (salt water), separate zones, each with a 

different corrosion rate, are present along the length of the pile. Tomlinson (1994) 

identifies these zones as follows: 

1. Atmospheric zone: exposed to the damp atmospheric conditions above the 

highest water level or subject to airborne spray.

2. Splash zone: above the mean high tide, but exposed to waves, spray, and 

wash from passing ships.

3. Intertidal zone: between mean high and low tides.

4. Continuous immersion zone: below lowest low tide.

5. Underground zone: below the mudline.

Figure 6-29, after Morley and Bruce (1983), summarizes average and maximum 

probable marine corrosion rates in these zones as well as in the low water zone. 

In corrosive environments, the designer should apply one of the design options 

for piles in corrosive environments discussed in Section 6.12.4.
193

Figure 6-29 Loss of thickness by corrosion for steel piles in seawater 

(after Morley and Bruce 1983). 

6.12.2 Sulfate and Chloride Attack on Concrete Piles

Attack on precast and cast-in-place concrete occurs in soils with high sulfate or 

chloride concentrations. Factors influencing the rate of deterioration on concrete 

piles include the pH of the soil, the solubility of the sulfate or chloride, the movement 

of the groundwater relative to the piles, and the density of the pile concrete.

The reaction between concrete and sulfate begins with sulfate ions in solution. 

Once the sulfate ions in the groundwater come in contact with Portland cement, an 

expansive chemical reaction takes place. Expansion of concrete often leads to 

cracking and spalling which can significantly reduce the available structural 

resistance of a pile foundation.

One method of reducing sulfate attack is to use a dense concrete which is less 

permeable to sulfate ions. Other possible deterrents include using sulfate-resisting 
194

cement, using cement with 25% pozzolanic material, or creating a physical barrier 

between the concrete and the groundwater with some form of pile sleeve.

Chlorides are commonly found in soils, groundwater, or industrial wastes. Instead of 

attacking concrete, chlorides cause corrosion of reinforcement steel with 

consequential expansion and bursting of concrete as the products of steel corrosion 

are formed. Once corrosion begins, it continues at an accelerated rate. This can 

lead to a loss of bond between steel and concrete and extreme reduction of pile 

structural resistance. Protective measures which can reduce corrosion include 

increased concrete cover around the reinforcing steel, and the use of galvanized, or 

epoxy coated reinforcement.

6.12.3 Bacteria, Fungi, Insect, and Marine Borer Attacks on Timber Piles 

Timber piles are subject to attack on land by bacteria, fungi, termites, and beetles, or 

in water by bacteria, fungi, and marine borers. Incidences of marine borer attack on 

timber piles have re-emerged in some areas as previously polluted water has 

improved. As mentioned in Section 6.2, arsenate and creosote pressure treatments 

are the most effective means of protecting timber piles from premature deterioration. 

In southern waters, creosote must be combined with other preservative treatments 

because of attack by limnoria tripuncata. Table 6-8 provides a summary of AWPI 

recommended preservative treatments depending upon foundation use, 

preservative, and wood species. Any environmental restrictions or regulations on a 

preservative treatment in a given situation must be considered.

When designing with timber piles, the wood species is usually not specified unless a 

specific species of wood is more suitable for design loads and/or environmental 

conditions. Certain species are not suitable for preservative treatment, while others 

may provide increased durability. As expected, ASTM standards for timber piles 

vary with geologic region, as land and fresh water piles have less stringent 

preservative treatment requirements than piles used in marine environments.

If timber piles are installed in other aggressive environments such as environments 

containing chemical wastes, a timber pile specialist should be consulted in 

determining the appropriate preservative treatment.
195

Table 6-8 Preservative Retention Requirements (after Collin 2002)

Pile Use

Category

Southern 

Pine

Creosote

(pcf)

Douglas

Fir

Creosote

(pcf)

Southern 

Pine

CCA

(pcf)

Douglas

Fir

ACZA

(pcf)

Foundation 12 17 0.8 1.0

Land & Fresh Water 12 17 0.8 1.0

Marine (Saltwater)

N. of Delaware1 or

San Francisco1 

16 16 1.5 1.5

Marine (Saltwater)

S. of New Jersey2 or 

San Francisco2 

20 20 2.5 2.5

Marine (Saltwater)

Dual Treatment3 20 20 1.0 1.0

1 Where Teredo is expected and Limnoria tripunctata is not expected, creosote or 

creosote solutions provide adequate protection.

2 Where Teredo and Limnoria tripunctata are expected and where pholad attack

is not expected, either dual treatment, or high retentions of CCA for Southern 

Pine or ACZA for Douglas Fir provide maximum protection.

3 In those areas where Limnoria tripunctata and pholad attack is expected or

known, dual treatment provides the maximum protection.

6.12.4 Design Options for Piles Subject to Degradation or Abrasion

When a pile must be installed in an aggressive or abrasive environment, several 

design options can be considered. These design options include:

1. A heavier steel section than required can be used to provide extra thickness 

(H and pipe sections). This method is not effective in running water with 

active bedload to scour the corroded surface.
196

2. Cathodic protection of steel piles in soil below the water table or in marine 

environments. Note that this method of protection tends to be a costly 

solution and requires periodic anode replacement.

3. Concrete encasement of steel piles above the mud line. This method may 

alter the impact absorbing properties of the pile.

4. Use of copper-bearing steel is effective against atmospheric corrosion but the 

cost is greater than conventional steel.

5. Sleeving or encapsulating of reinforced, cast-in-place piles by using metal 

casings, polymer or fiberglass jackets isolates contaminants from concrete.

6. Use of a low water/cement ratio, resistant aggregate, and minimum air 

content consistent with the environment to improve abrasion resistance of 

precast concrete piles.

7. Use of a protective metallic or epoxy paint (isocyanate-cured) or fusion 

bonded epoxy coating on exposed sections of the pile. These options may 

not as effective in running water with active bedload.

8. Use of coal-tar epoxies for corrosion protection in marine environments. This 

method can also lose effectiveness in running water with active bedload.

Protective coatings cannot be replaced after a pile is driven. Therefore, if a 

protective coating is used, the coating should be designed to be durable enough to 

remain undamaged during pile transportation, handling, and placement in the leads 

for driving as well as resistant to the abrasion resulting from pile driving. The 

designer should also note that the shaft resistance on a coated pile may be 

significantly different than on an uncoated pile, depending on the coating.

6.13 SELECTION OF PILE TYPE AND SIZE FOR FURTHER EVALUATION

The selection of appropriate pile types for any project involves the consideration of 

several design and installation factors including pile characteristics, subsurface 

conditions and performance criteria. This selection or elimination process should 

consider the factors listed in Tables 6-1 to 6-7, 6-9, and 6-10. Tables 6-1 to 6-7

summarize typical pile characteristics and uses. Table 6-9 provides pile type 

recommendations for various subsurface conditions. Table 6-10 presents the 

placement effects of pile shape characteristics.
197

In addition to the considerations provided in the tables, the problems posed by the 

specific project location and topography must be considered in any pile selection 

process. Following are some of the usually encountered issues:

1. Vibrations from driven pile installation may affect pile type selection, dictate 

installation equipment selection and/or use of special installation techniques 

such as predrilling, and/or necessitate vibration monitoring of adjacent 

structures. 

2. Urban areas, remote areas, or other locations with limited access may restrict 

driving equipment size and, therefore, pile type and size.

3. Local availability of certain materials and capability of contractors may have 

decisive effects on pile selection.

4. Waterborne operations may dictate use of shorter pile sections due to pile 

handling limitations.

5. Steep terrain may make the use of certain pile equipment costly or 

impossible.

Often several different pile types meet all the requirements for a particular structure. 

In such cases, the final choice should be made on the basis of a cost analysis that 

assesses the over-all cost of the foundation alternatives. This requires that 

candidate pile types be carried forward in the design process for determination of the 

pile section requirements for design loads and constructability. The cost analysis 

should also include time and space limitations, time delays, cost of load testing 

programs, as well as the differences in the cost of pile caps and other elements of 

the structure that may differ among alternatives. For major projects, alternate 

foundation designs should be considered for inclusion in the contract documents if 

there is a potential for cost savings.
198

Table 6-9 Pile Type Selection Based on Subsurface and Hydraulic Conditions.

TYPICAL PROBLEM RECOMMENDATIONS

Boulders overlying bearing

stratum

Use heavy non-displacement pile with a cast 

steel driving shoe and include contingent 

predrilling item in contract.

Loose cohesionless soil

 

Use a tapered pile to develop the maximum unit 

shaft resistance. 

Downdrag and Drag Force

Avoid use of battered piles due to ground 

settlement. A pile-soil bond breaker such as a 

bitumen coating or plastic wrap (if feasible) can 

be used to reduce large drag forces.

Deep soft clay Use rough concrete piles to increase adhesion 

and rate of pore water dissipation.

Artesian pressure

Use solid prestressed concrete pile, tapered 

piles with sufficient collapse strength, or thick

wall closed end pipe with flush boot plate 

depending upon local practice. H-piles without 

driving shoes may also be viable selection. Do 

not use mandrel driven thin-wall shells, as 

generated hydrostatic pressure may cause shell 

collapse. Pile heave also common to closed 

end pile.

Scour

Use uniform section pile with sufficient 

structural resistance to act as a column through 

scour zone. Do not use tapered piles unless a 

large part of the taper extends well below scour 

depth.

Coarse gravel deposits

Use prestressed concrete piles where hard 

driving is expected. In coarse soils, use of Hpiles and open end pipe piles often results in 

excessive pile lengths and cost overruns due to 

pile running during installation. 

* Table modified and reproduced (Cheney and Chassie 1993).
199

Table 6-10 Pile Type Selection Based on Pile Shape Effects.

6.14 HISTORICAL PRICE INFORMATION

Many state transportation agencies post project bid and award information. This 

media may be downloaded for use by designers, contractors, and project managers 

to aid in cost estimating. The corresponding agency standard specifications are also 

typically available to download and should be reviewed to determine the description 

of the work as well as the method of measurement and payment. Tables 6-11 to 

6-17 present recent historical statewide price information for several state agencies.

The description of the work requirements vary from state to state. Prices for 

furnishing piles and driving piles are listed separately for some state agencies, while 

a combined bid price is used by others. Items such as pile shoes, splices, sleeves, 

load tests and mobilization may or may not be incidental, and vary by agency.
200

6.14.1 California

According to the CALTRANS specification, furnished cost includes any length of pile 

installed in the ground, including splices and materials to splice. Driving costs 

included installing and cutting off piles, providing pile tips or shoes and any 

predrilling involved. Table 6-11 presents costs tabulated from 2014 for various pile 

types, including the quantity, weighted unit price per linear foot and number of 

projects for each. Both steel and concrete piles are driven in California and are 

presented in this table. Pile class 90, 140, or 200 identifies the axial compression 

resistance in kips for the prestressed concrete piles in the service limit state.

Table 6-11 CALTRANS, 2014 Contract Cost Data

Description Quantity1,2 Weighted 

Unit Price

No. of 

Projects

FURNISH STEEL PILING (HP 10 X 57) 12,882 $30.53 3 

FURNISH STEEL PILING (HP 14 X 89) 12,100 $45.55 1 

FURNISH STEEL PILING (HP 14 X 117) 22,593 $63.69 2 

FURNISH 24” CIS SHELL PIPE PILING 8,497 $135.31 3 

FURNISH 30” CIS SHELL PIPE PILING 1,262 $145.00 1 

FURNISH 48” CIS SHELL PIPE PILING 1,242 $448.19 2 

FURNISH PILING (CLASS 90) 3,729 $50.00 1 

FURNISH PILING (CLASS 140) 25,583 $25.00 1 

FURNISH PILING (CLASS 200) 19,256 $62.87 2 

DRIVE STEEL PILE (HP 10 X 57) 280 $2,034.57 3 

DRIVE STEEL PILE (HP 14 X 89) 310 $1,050.00 1 

DRIVE STEEL PILE (HP 14 X 117) 424 $1,283.02 2 

DRIVE 24" STEEL PIPE PILE 132 $5,863.63 3 

DRIVE 30" STEEL PIPE PILE 18 $20,200.00 1 

DRIVE 48" STEEL PIPE PILE 23 $23,130.43 2 

DRIVE PILE (CLASS 90) 94 $1,550.00 1 

DRIVE PILE (CLASS 140) 336 $3,100.00 1 

DRIVE PILE (CLASS 200) 261 $2513.22 2 

1Furnished Units in Linear Feet. 

2Driven Units in Individually Driven Sections.

http://www.dot.ca.gov/hq/esc/oe/awards/2014CCDB/2014ccdb.pdf
201

6.14.2 Florida

Although prestressed concrete piles are primarily driven in Florida, steel piles are 

also included by the Florida Department of Transportation in their 2015 Historical 

Cost Summary. Costs are summarized on a statewide basis on a moving 12 month 

average. The data presented below is for the 12 month period ending January 31, 

2016. Prices include furnishing and driving piles, per linear foot, with splices 

included in the per foot price. Pile points or driving shoes are paid per each. Table 

6-12 presents the piling related costs.

Table 6-12 FDOT 2015 Historical Cost Information

Description Quantity1,2

Weighted 

Average 

Price 

No. of 

Projects

PRESTRESSED CONCRETE PILING, 14" SQ 460 $230.00 1 

PRESTRESSED CONCRETE PILING, 18" SQ 21,755 $93.73 7 

PRESTRESSED CONCRETE PILING, 24" SQ 105,468 $90.12 10

STEEL PILING, HP 12 X 53 28,113 $49.11 1 

STEEL PILING, HP 14 X 89 5,185 $90.57 2 

STEEL PILING, 24" DIA. PIPE 560 $224.28 1 

POINT PROTECTION , HP 12 X 53 297 $240.24 1 

POINT PROTECTION , HP 14 X 89 61 $262.55 1 

POINT PROTECTION , 24” PIPE 20 $340.69 1 

STEEL PILING, 24" DIA. PIPE 7,936 $121.27 5 

TEST PILES - PREST CONCRETE,18" SQ 4,335 $198.65 6 

TEST PILES - PREST CONCRETE,24" SQ 9,650 $181.58 9 

TEST PILES - STEEL, HP 12 X 53 1,783 $85.63 1 

TEST PILES - STEEL, HP 14 x 89 535 $128.69 1 

TEST PILES - STEEL, 24" DIA. PIPE 140 $521.59 1 

1Units in Linear Feet for Piles. 

2

Units in Per Each for Point Protection. 

http://www.fdot.gov/programmanagement/Estimates/HistoricalCostInformation/Histor

icalCost.shtm 
202

6.14.3 Indiana

Steel piles are used almost exclusively across the Midwest. Table 6-13 presents the 

weighted average unit bid price for Indiana Department of Transportation pile driving 

items in 2015. This information is tabulated annually and is available on their 

website. Pile costs are per linear foot and include furnishing and driving the piles 

including splices. For closed end steel pipe piles, the price includes the flat end 

plate and concrete fill. Conical tip protection or inside cutting shoes for pipe piles, 

driving shoes for H-piles, and mobilization of the driving system are separate pay 

items. Piles are priced per linear foot installed.

Table 6-13 INDOT 2015 Unit Price Summaries

Description Quantity1,2

Weighted 

Average 

Price

PILE, STEEL PIPE, 14 IN O.D. X 0.250 IN WALL 5,370 $48.45

PILE, STEEL PIPE, 14 IN O.D. X 0.312 IN WALL 13,352 $42.10

PILE, STEEL PIPE, 14 IN O.D. X 0.375 IN WALL 3,318 $55.39

PILE, STEEL PIPE, 14 IN O.D. X 0.50 IN WALL 2,861 $58.51

PILE, STEEL PIPE, 18 IN O.D. X 0.50 IN WALL 1,254 $95.00

PILE, STEEL PIPE, 24 IN O.D. X 0.50 IN WALL 838 $65.00

PILE, STEEL PIPE, 24 IN O.D. X 0.75 IN WALL 7,700 $75.00

PILE, STEEL H, HP 12 X 53 17,314 $44.21

PILE, STEEL H, HP 12 X 74 8,362 $67.82

PILE, STEEL H, HP 12 X 84 3,548 $64.39

PILE, STEEL H, HP 14 X 73 59 $325.00

PILE, STEEL H, HP 14 X 89 669 $109.50

CONICAL PILE TIP, 14 IN 327 $277.81

CONICAL PILE TIP, 16 IN 15 $1000.00

PILE SHOE, 24 IN INSIDE CUTTING SHOE 102 $400.00

PILE SHOE, HP 12 X 53 286 $123.43

PILE SHOE, HP 12 X 74 152 $113.85

PILE SHOE, HP 12 X 84 22 $118.17

PILE SHOE, HP 14 X 73 4 $125.00

PILE SHOE, HP 14 X 89 12 $140.00

1 Units in Linear Feet. 

2 Tip Reinforcement Unit Price for Each. 

http://www.in.gov/dot/div/contracts/pay/ 
203

6.14.4 Maryland

The Maryland State Highway administration maintains a price index that is updated 

twice per year. The piling items, quantities and associated unit costs summarized in 

the July 2015 price index are presented in Table 6-14. In Maryland, furnishing and 

driving pile are included in the same pay item and are priced per linear foot. 

Mobilization of the pile driving system and pile splices are incidental items included 

in the per foot price. However H-pile shoes (pile points) are a separate bid item, 

unless specified and are priced per item. Projects for the Maryland State Highway 

Administration primarily utilize steel piles.

Table 6-14 MDSHA Price Index Published July 2015

Description Quantity1,2

Weighted 

Average 

Price 

Projects

UNTREATED TIMBER PILE 25,855 $21.19 6 

STEEL HP 8 X 36 BEARING PILE 188 $65.00 1 

STEEL HP 12 X 53 BEARING PILE 17,703 $47.74 4 

STEEL HP 12 X 63 BEARING PILE 2,460 $85.00 1 

STEEL HP 12 X 74 BEARING PILE 7,472 $55.00 1 

STEEL HP 14 X 73 BEARING PILE 4,328 $55.32 1 

STEEL HP 14 X 89 BEARING PILE 66,092 $60.25 9 

STEEL HP 14 X 102 BEARING PILE 1,020 $106.51 1 

STEEL HP 8 X 36 BEARING TEST PILE 13 $180.00 1 

STEEL HP 12 X 53 BEARING TEST PILE 568 $110.30 4 

STEEL HP 12 X 63 BEARING TEST PILE 260 $85.00 1 

STEEL HP 12 X 74 BEARING TEST PILE 285 $85.00 1 

STEEL HP 14 X 73 BEARING TEST PILE 183 $100.00 1 

STEEL HP 14 X 89 BEARING TEST PILE 2,250 $87.13 9 

STEEL HP 14 X 102 BEARING TEST PILE 255 $196.19 1 

PILE POINT FOR 12 INCH HP BEARING PILE 144 $119.63 2 

PILE POINT FOR 14 INCH HP BEARING PILE 48 $126.00 1 

1 Pile Units in Linear Feet. 

2 Pile Point Unit Price for Each. 

http://www.roads.maryland.gov/ohd2/MDSHA_PriceIndex_Jul2015.pdf 
204

6.14.5 North Carolina

Multiple geologic conditions dictate that both concrete and steel piles are driven in 

North Carolina. Table 6-15 presents the 2015 average bid prices compiled by the 

North Carolina Department of Transportation. Piles are priced per linear foot, while 

pile points and pile plates are separate items. More detailed information on the 

scope of measurement and payment may be found in the most recent agency 

standard specification.

Table 6-15 2015 NCDOT Bid Quantities and Averages

Description Quantity1,2

Weighted 

Average 

Price 

12" PRESTRESSED CONCRETE PILE 18,220 $58.94

16" PRESTRESSED CONCRETE PILE 28,315 $73.39

24" PRESTRESSED CONCRETE PILE 6,076 $144.80

HP12X53 PILES 43,983 $49.45

HP14X53 GALVANIZED PILES 680 $85.00

HP14X73 PILES 10,016 $62.42

HP14X73 GALVANIZED PILES 1,660 $71.52

HP14X89 PILES 1,195 $75.00

14 IN O.D. X 0.50 STEEL PIPE 3,000 $67.37

36 IN O.D. X 0.625 STEEL PIPE 95 $230.00

14 IN O.D. X 0.50 GALVANIZED STEEL PIPE 480 $145.87

16 IN O.D. X 0.50 GALVANIZED STEEL PIPE 400 $124.53

18 IN O.D. X 0.50 GALVANIZED STEEL PIPE 3,570 $101.47

24 IN O.D. X 0.50 GALVANIZED STEEL PIPE 7,680 $152.39

30 IN O.D. X 0.50 GALVANIZED STEEL PIPE 8,560 $193.34

30 IN O.D. X 0.625 GALVANIZED STEEL PIPE 4,850 $212.30

36 IN O.D. X 0.625 GALVANIZED STEEL PIPE 770 $215.00

PREDRILLING FOR PILES 13,169 $20.38

STEEL PILE POINTS 673 $386.19

PIPE PILE PLATES 262 $168.74

1 Pile Units in Linear Feet. 

2

Pile Point & Plate Unit Price for Each. 

https://connect.ncdot.gov/letting/pages/central-letting-resources.aspx 
205

6.14.6 Pennsylvania

Price history for the Pennsylvania Department of Transportation is presented in 

Table 6-16 . The latest PennDOT price history covers projects let between April 

2013 and April 2015. The majority of PennDot projects involve steel piles, many of 

which are H-piles driven to rock due to local geology. For bearing piles, the unit 

price per linear foot includes furnished and driven pile including splices and cutoff. 

Driving shoes (tips) are a separate pay item. Shoes for timber piles are included in 

the unit price.

Table 6-16 2015 PennDOT Bid Quantities and Averages

Description Quantity1,2

Weighted 

Average

Price

No. of 

Projects

TREATED TIMBER BEARING PILES 14,800 $40.00 1 

CAST-IN-PLACE CONCRETE PILES 4,907 $74.37 7 

CAST-IN-PLACE CONCRETE PILE TIP 22 $804.23 3 

STEEL BEAM BEARING PILES, HP10X42 150 $144.50 1 

STEEL BEAM BEARING PILES, HP10X57 10,376 $63.01 8 

STEEL BEAM BEARING PILES, HP12X53 673 $60.49 3 

STEEL BEAM BEARING PILES, HP12X63 3,350 $54.79 3 

STEEL BEAM BEARING PILES, HP12X74 63,600 $70.31 47

STEEL BEAM BEARING PILES, HP 12X84 5,278 $96.46 5 

STEEL BEAM BEARING PILES, HP14X73 4,497 $71.95 3 

STEEL BEAM BEARING PILES, HP14X89 134,174 $85.00 19

STEEL BEAM BEARING PILES, HP14X102 126,960 $68.90 15 

STEEL BEAM BEARING PILES, HP14X117 26,458 $91.33 3 

NORMAL DUTY PILE TIP - HP 10X57 56 $118.58 2 

NORMAL DUTY PILE TIP - HP 12X63 44 $151.59 2 

NORMAL DUTY PILE TIP - HP 12X74 410 $151.17 13

NORMAL DUTY PILE TIP - HP 14X73 50 $149.04 2 

NORMAL DUTY PILE TIP - HP 14X89 26 $129.00 1 

HEAVY DUTY PILE TIP - HP 12X74 405 $141.92 11

HEAVY DUTY PILE TIP - HP 12X84 49 $104.08 2 

HEAVY DUTY PILE TIP - HP 14X89 1,388 $134.28 18 

HEAVY DUTY PILE TIP - HP 14X117 254 $135.00 1 

1 Pile Units in Linear Feet. 

2Tip Reinforcement Unit Price for Each. 

https://www.dot.state.pa.us/public/Bureaus/design/Pub287/Pub%20287.pdf
206

6.14.7 Texas

A 12-month statewide moving average of bid prices and quantities from the Texas 

Department of Transportation is presented below in Table 6-17. The weighted 

average prices cover the period from March 1, 2015 to February 28, 2016. The unit 

bid price for piling is per linear foot in place. However, splices and pile tips, when 

necessary, are a separate bid item. The majority of driven piles in Texas are 

concrete piles. Please consult the most recent Texas Department of Transportation 

Standard Specification for further information on measurement and payment items, 

as well as the average project bid awards as these values are updated monthly.

Table 6-17 TxDOT 12-Month Project Bid Averages

Description 12 Month 

Quantity1

12 Month 

AVG BID

No. of 

Projects

STEEL H-PILING (HP 12 X 53) 640 $150.00 1 

STEEL H-PILING (HP 14 X 117) 201 $20.00 1 

PRESTR CONCRETE PILING (16 IN SQ) 48,679 $113.38 17 

PRESTR CONCRETE PILING (18 IN SQ) 7,133 $100.20 8 

PRESTR CONCRETE PILING (20 IN SQ) 8,151 $118.80 3 

PRESTR CONCRETE PILING (24 IN SQ) 3,304 $135.00 1 

1 Units in Linear feet. 

http://www.txdot.gov/business/letting-bids/average-low-bid-unit-prices.html 

A brief summary of pay items was presented for selected state agencies covering a 

range of pile types and installation conditions. The designer should consult the most 

current price information from the appropriate agency as price histories and 

specifications change over time.
207

REFERENCES

American Concrete Institute (ACI), (2012). Guide to Design, Manufacture and 

Installation of Concrete Piles. ACI-543R-12, 64 p.

American Association of State Highway and Transportation Officials (AASHTO). 

(2001). Standard Recommended Practice for Assessment of Corrosion of 

Steel Piling for Non-Marine Applications. AASHTO Standard Specifications 

for Transportation Materials and Methods of Sampling and Testing, Part 1B: 

Specifications, 24th Edition, 13 p.

American Association of State Highway and Transportation Officials (AASHTO). 

(2014). AASHTO LRFD Bridge Design Specifications, US Customary Units, 

Seventh Edition, with 2015 Interim Revisions. American Association of State 

Highway and Transportation Officials, Washington, D.C., 1960 p.

Beavers, J.A. and Durr, C.L. (1998). Corrosion of Steel Piling in Non-Marine 

Applications. NCHRP Report 408, National Cooperative Highway Research 

Program, Transportation Research Board, Washington, D.C., 35 p.

Belk, N.A. (2013). Evaluation of Lightweight Aggregate Concrete for Precast, 

Prestressed Driven Piles. PhD Dissertation, Auburn University, 215 p.

Brown, D.A., Dapp, S.D., Thompson, W.R., and Lazarte, C.A. (2007). Design and 

Construction of Continuous Flight Auger (CFA) Piles. FHWA-HIF-07-03, 

Geotechnical Engineering Circular (GEC) No. 8. U.S. Dept. of 

Transportation, Federal Highway Administration, 289 p.

Brown, D. A., Turner, J.P. and Castelli R.J. (2010). Drilled Shafts: Construction 

Procedures and LRFD Design Methods, FHWA-NHI-10-016, Geotechnical 

Engineering Circular (GEC) No. 10. U.S. Dept. of Transportation, Federal 

Highway Administration, 970 p.

Brown, D.A., and Thompson III, W.R. (2015). Current Practices for Design and 

Load Testing of Large Diameter Open-End Driven Pipe Piles. Final Report. 

NCHRP Report 20-05, Topic 45-05, National Cooperative Highway Research 

Program, Washington, D.C., 175 p.
208

Chellis, R.D. (1961). Pile Foundations. Second Edition, McGraw-Hill Book 

Company, New York, NY, 704 p.

Cheney, R.S. and Chassie, R.G. (2000). Soils and Foundations Workshop 

Reference Manual. FHWA HI-00-045, U.S. Department of Transportation, 

National Highway Institute, Federal Highway Administration, Washington, 

D.C., 358 p. 

Collin, J.G. (2002). Timber Pile Design and Construction Manual. American Wood 

Preservers Institute (AWPI), 122 p.

Department of the Navy, (1982). Foundations and Earth Structures Design Manual. 

DM 7.2. Naval Facilities Engineering Command, (NAVFAC), Alexandria, VA, 

279 p. 

Fleming, W.G.K., Weltman, A.J., Randolph, M.F. Elson, W.K. (2008). Piling 

Engineering, Third Edition, Taylor and Francis, New York, New York, NY, 398

p.

Fuller, F.M. (1983). Engineering of Pile Installations. McGraw-Hill, New York, NY, 

286 p.

Graham, J. (1995). Personal Communication.

Hartman, J.J., Castelli, R.J., and Malhotra, S. (2007). Design and Installation of 

Concrete Cylinder Piles. Proceeding of GeoDenver 2007, GSP 158 

Contemporary Issues in Deep Foundations, ASCE, pp. 1-14.

Moser, R., Holland, B. Kahn, L., Singh, P., and Kurtis, K. (2011). Durability of 

Precast Prestressed Concrete Piles in Marine Environment: Reinforcement 

Corrosion and Mitigation Part 1. GDOT Research Project No. 07-30. Office of 

Materials and Research Georgia Department of Transportation, 243 p.

Morley, J. (1979). The Corrosion and Protection of Steel Piling, British Steel 

Corporation, Teesside Laboratories, Report No. T/CS/906/4/78/C.

Morley, J. and Bruce, D.W. (1983). Survey of Steel Piling Performance in Marine 

Environments, Final Report, Commission of the European Communities, 

Document EUR 8492 EN, 25 p.
209

Portland Cement Association (PCA). (1951). Concrete Piles: Design, Manufacture 

and Driving, 80 p.

Prakash, S. and Sharma, H. (1990). Pile Foundations in Engineering Practice. John 

Wiley and Sons, Inc., New York, NY, 768 p.

Rausche, F. (1994). Design, Installation and Testing of Nearshore Piles. 

Proceedings of the 8th Annual Symposium on Deep Foundations, Vancouver.

Sabatini, P.J., Tanyu, B., Armour., P., Groneck, P., and Keeley, J. (2005). Micropile 

Design and Construction, FHWA-NHI-05-039. National Highway Institute, 

U.S. Dept. of Transportation, Federal Highway Administration, Washington, 

D.C., 436 p.

Tomlinson, M.J. (1994). Pile Design and Construction Practice, Fourth Edition, E & 

FN Spon, London, 432 p.

Transportation Research Board. (1977). Design of Pile Foundations. NCHRP 

Synthesis of Highway Practice No. 42, 68 p.

Vande Voort, T., Suleiman, M. T., and Sritharan, S. (2008). Design and 

Performance Verification of Ultra-High Performance Concrete Piles for Deep 

Foundations. Final Report, Iowa DOT, IHRB Project TR-558, CTRE Project 

06-264, Iowa Dept. of Transportation, Ames, IA, 206 p.
210
211

CHAPTER 7

GEOTECHNICAL ASPECTS AND LIMIT STATE DESIGN

7.1 INTRODUCTION

Static analysis methods can be categorized as analytical methods that use 

geomaterial strength and compressibility properties to determine nominal 

geotechnical resistance and deformation. This chapter will focus on analysis 

methods for determining the nominal geotechnical resistance of single piles and pile 

groups in axial compression, uplift, and lateral loading as well as the resulting 

vertical and lateral deformations. Important considerations are as follows:

1. Static analysis methods are an integral part of the design process. Static 

analysis methods are necessary to determine the most cost effective pile type 

and to estimate the number of piles and the required pile lengths for the 

design of substructure elements. The foundation designer must have 

knowledge of the design loads and the project performance criteria in order to 

perform the appropriate static analyses.

2. Many static analysis methods are available. The methods presented in this 

chapter are relatively simple methods that have proven to provide reasonable 

agreement with full scale field results. Many of these methods are also

included in the AASHTO (2014) design specifications. Other more 

sophisticated analysis methods may be used and in some cases may provide 

better results. Regardless of the method used, it is important to continually 

apply experience gained from past field performance of the analysis method. 

3. Designers should fully understand the basis for, the limitations of, and the 

applicability of a chosen method. This is particularly true of computer 

solutions where some underlying assumptions may not be readily apparent. 

A selected method should also have a proven agreement with full scale field 

results. 

4. Evaluation and confirmation of pile drivability is an integral part of limit state 

design. Foundation designs with fewer, larger, pile sections having greater 

nominal resistances are more frequently being used. The ability of these 

piles to achieve the required pile penetration depths necessary to satisfy all 

nominal resistance and deformation requirements is a critical design check. 

A pile drivability analysis requires combining the geotechnical aspects 

detailed in this chapter with the wave equation analysis procedures described 
212

in Chapter 12. A drivability analysis should be performed by the engineer 

during the design stage to assess the constructability of the pile design.

Construction procedures can have a significant influence on the behavior of pile 

foundations. The analysis methods described in this chapter lead to successful 

designs of deep foundations only if appropriate construction techniques and 

construction monitoring methods are used. Construction monitoring should be an 

integral part of the design and construction of any foundation. Static load tests, 

wave equation analysis or dynamic monitoring for construction control should, 

whenever possible, be used to confirm the results of a static design method. These 

topics are discussed in detail in subsequent chapters.

The first few sections of this chapter will briefly cover background information. Static 

analysis procedures for piles subject to compression, uplift, and lateral loads will be 

covered, as well as pile group settlement. The influence of special design events on 

driven pile designs will also be discussed. Limited guidance on design in 

liquefaction susceptible soils is provided. Detailed seismic design is beyond the 

scope of this manual and is covered in GEC-3 by Kavazanjian et.al (2011). Finally, 

this chapter will address construction issues pertinent to static analysis methods and 

foundation design. In all cases, existing character notation for soil resistances as 

well as load and resistance factors utilized by AASHTO (2014) will be presented.

7.1.1 Static Analysis Methods in Limit State Design

There are four general types of static analyses addressed in this chapter. Static 

analyses are performed to determine:

1. Nominal resistance in axial compression of a pile or pile group. These 

calculations are performed to determine the long term resistance of a 

foundation as well as to determine the soil resistance provided from soil 

layers subject to scour, liquefaction, downdrag, or that are otherwise 

unsuitable for long term load support. Static analyses are used to establish 

minimum pile penetration requirements, pile lengths for bid quantities, as well 

as to estimate the soil resistance at the time of driving (SRD) and the required 

nominal driving resistance, Rndr. 

2. Nominal resistance in axial tension of a pile of pile group. These calculations 

are performed to determine the soil resistance to uplift or tension loading 

which, in some cases, may also determine the minimum pile penetration 

requirements.
213

3. Nominal lateral resistance and lateral deformation of a pile or pile group. 

These soil-structure interaction analysis methods consider the soil strength 

and deformation behavior as well as the pile structural properties and are 

used in pile section selection.

4. Settlement of a pile group. These calculations are performed to estimate the 

vertical foundation deformation under the structure’s service loads.

The factored geotechnical resistance of a pile can be defined as the sum of 

geomaterial resistances along the pile shaft and at the pile toe available to support 

the factored loads on the pile. As noted above, static analyses are performed to 

determine the factored resistance of an individual pile and of a pile group as well as 

the deformation response of a pile group to the factored loads. The factored

resistance of an individual pile and of a pile group is the smaller of: (1) the factored 

geotechnical resistance of surrounding soil/rock medium to support the loads 

transferred from the pile(s) or, (2) the factored structural resistance of the pile(s). 

Soil-structure interaction analysis methods are used to determine the deformation 

response of a pile and pile groups to lateral loads. The results from these analyses 

in conjunction with pile group settlement analysis are compared to the performance 

criteria established for the structure. Both vertical (settlement) and lateral 

deformation analyses of pile groups are computed using factored loads at the 

service limit state.

The static pile resistance from the sum of the soil/rock resistances along the pile 

shaft and at the pile toe can be estimated from geotechnical engineering analysis 

using: 

1. Laboratory determined shear strength parameters of the soil and rock 

surrounding the pile.

2. In-situ test data (i.e., SPT, CPT).

3. Back analysis of geomaterial design parameters based on performance data.

Prior to discussing static design methods for estimating pile resistance in detail, it is 

desirable to review events that occur in the pile soil system during and after pile 

driving as well as basic load transfer mechanisms.
214

7.1.2 Events During and After Pile Driving

The soil and some weaker rocks in which a pile foundation is installed are almost 

always disturbed. Several factors influence the degree of disturbance. These 

include the soil and rock type and density, the pile type (displacement, lowdisplacement), and the method of pile installation (driven, vibrated, drilled, jetted). 

For driven piles, substantial disturbance and remolding is unavoidable in soils and 

weaker rocks where the in-situ stresses and rock structure are changed during pile 

installation. 

Over the long term, scour, settlement, pore pressure fluctuations, seismic and other 

extreme events may occur. These situations should be addressed in the applicable 

strength, extreme, and service limit states such that sufficient embedment and 

foundation stiffness is achieved. These topics, among others, will be discussed 

throughout this chapter.

7.1.2.1 Cohesionless Soils

The resistance of piles driven into cohesionless soil depends primarily on the relative 

density of the soil. During driving, the relative density of loose to medium dense 

cohesionless soil is increased close to the pile due to vibrations and lateral 

displacement of soil. This effect is most pronounced in the immediate vicinity of 

displacement piles. Broms (1966) and more recent studies found the zone of 

densification extends as far as 3 to 5.5 diameters away from the pile shaft and 3 to 5 

diameters below the pile toe as depicted in Figure 7-1. 

The increase in relative density increases the resistance of single piles and pile 

groups. The pile type selection also affects the amount of change in relative density. 

Piles with large displacement characteristics such as closed end pipe and precast 

concrete piles increase the relative density of cohesionless material more than lowdisplacement open end pipe or steel H-piles. Similarly, vibratory installation may 

increase the relative density of loose to medium dense cohesionless material more 

than driving with an impact hammer.

In dense cohesionless soil, the relative density may actually decrease since dense 

soils will dilate during shear and displacement. As a result, negative pore pressures 
215

Figure 7-1 Compaction of cohesionless soils during pile driving (after Broms 

1966).

may temporarily be generated during driving which could lead to a temporarily 

increased resistance. The increase in stress, which can occur adjacent to the pile

shaft and/or below the pile toe during the driving process, can be lost by relaxation in 

dense sand and gravels as the negative pore pressures generated during driving are 

dissipated. The phenomena can be explained by the shear strength equation

presented in Equation 7-1.

 Eq. 7-1 

Where: 

τ = shear strength of soil.

 c = cohesion.

σ = total normal stress (pressure) on plane of failure. 

 u = pore water pressure.

φ = angle of internal friction.

Negative pore pressures temporarily increase the soil shear strength, and therefore

pile resistance, by changing the (σ - u) tan φ component of shear strength to (σ + u) 

tan φ. As negative pore pressures dissipate, the shear strength and pile resistance 

decrease. 
216

The pile driving process can also generate high positive pore water pressures in 

saturated cohesionless silts and loose to medium dense fine sands. Positive pore 

pressures temporarily reduce the soil shear strength and the pile resistance. This 

phenomena is identical to the one described below for cohesive soils. The gain in 

resistance with time or soil setup is generally quicker for sands and silts than for 

clays because the pore pressures dissipate more rapidly in cohesionless soils than 

in cohesive soils.

7.1.2.2 Cohesive Soils

When piles are driven into saturated cohesive materials, the soil near the piles is 

disturbed and radially compressed. For soft or normally consolidated clays, the 

zone of disturbance is generally within one pile diameter around the pile. For piles 

driven into saturated stiff clays, there are also significant changes in secondary soil 

structure (closing of fissures) with remolding and loss of previous stress history 

effects in the immediate vicinity of pile. Figure 7-2 illustrates the disturbance zone 

for piles driven in cohesive soils as observed by Broms (1966). This figure also 

notes the heave that can occur when driving displacement piles in cohesive soils. 

Figure 7-2 Disturbance of cohesive soils during pile driving (after Broms 1966).
217

The disturbance and radial compression generate high pore pressures (excess 

positive pore pressures) which temporarily reduce soil shear strength, the nominal 

geotechnical resistance, and therefore the pile penetration resistance or blow count. 

As reconsolidation of cohesive soils around the pile occurs, the high pore pressures 

are diminished, which leads to an increase in shear strength and pile resistance 

(setup). This phenomenon is opposite to "relaxation" described for cohesionless 

soils. The zone and magnitude of soil disturbance are dependent on the soil 

properties of soil sensitivity, driving method, and the pile foundation geometry. 

Limited data available for partially saturated cohesive soils indicates that pile driving 

does not generate high pore pressures and hence significant soil setup does not 

occur. 

7.1.2.3 Additional Soil Resistance Considerations

On many driven pile projects, multiple static analyses are required for a design. 

First, a static analysis is necessary to determine the number and length of piles 

necessary to support the structure loads. A second static analysis may also be 

required to determine the nominal resistance the pile will encounter during 

installation. This second analysis enables the design engineer to determine the 

necessary capability of the driving equipment and the minimum pile section 

requirements based on drivability. Figures 7-3 and 7-4 illustrate situations that 

require two static analyses.

Figure 7-3 shows a situation where piles are to be driven for a bridge pier. In this 

case, the first static analysis performed should neglect the soil resistance in the soil 

zone subject to scour, since this resistance may not be available for long term 

support. The number of piles and pile lengths determined from this analysis will then 

be representative of the long term conditions in the event of scour. At the time of 

pile driving however, the scour zone soil will provide resistance to pile penetration. 

Therefore, a second static analysis is required to estimate the nominal resistance 

encountered by the pile during driving to the embedment depth determined in the 

first analysis. The second static analysis includes the soil resistance in the materials 

above the scour depth as well as the underlying strata.

Figure 7-4 shows another frequently encountered situation in which piles are driven 

through loose uncompacted fill material into the natural ground. The loose fill 

material offers unreliable resistance and is usually neglected in determining the 

number of piles and the pile lengths required. A second static analysis is then 

performed to determine nominal resistance encountered by the pile during driving, 

which includes the resistance in the fill material.
218

Figure 7-3 Situation where two static analyses are necessary due to scour.

Figure 7-4 Situation where two static analyses are necessary due to fill materials. 
219

In both examples, the soil resistance to be overcome during driving will be 

substantially greater than the required nominal resistance,Rn. The geotechnical 

resistance computed in the scourable or unsuitable layer, determined from static 

analysis, is then added to the required nominal resistance when determining the 

required nominal driving resistance,Rdnr. 

The results of multiple static analyses should be considered in the development of 

project plans and specifications. For example, consider a case where scour, uplift 

loading, or some other design consideration (e.g., seismic event) dictates that a 

greater pile penetration depth be achieved than that required for support of the axial 

compressive loads. The static analyses indicate that a soil resistance of 420 kips 

must be overcome to obtain the minimum penetration depth needed for a 300 kip 

nominal resistance in axial compression. This information is germane to the 

drivability assessment for pile section selection. It should also be conveyed in the 

construction documents so that the driving equipment can be properly sized and so 

that the intent of the design is clearly and correctly interpreted by the contractor and 

construction personnel. Specifying only the 300 kip nominal resistance or providing 

only the factored load on the plans could easily be misinterpreted and can lead to 

construction claims. In the above example, the controlling factored load, the nominal 

resistance, the required nominal driving resistance, and the minimum penetration 

depth should all be provided. A more in depth discussion on driving criteria is 

presented in Chapter 17.

7.1.3 Load Transfer

The nominal resistance, Rn, of a pile in homogeneous soil may be expressed by the 

sum of the shaft resistance Rs and toe resistance Rp, or

+= RRR psn Eq. 7-2 

The above equation for nominal resistance assumes that both the pile toe and the 

pile shaft have moved sufficiently with respect to the adjacent soil to simultaneously 

develop the nominal shaft and toe resistances. Generally, the displacement needed 

to mobilize the shaft resistance is smaller than that required to mobilize the toe 

resistance. This simple rational approach has been commonly used for all piles 

except very large piles greater than 36 inches in diameter or width. 

Figure 7-5 illustrates typical load transfer profiles for a single pile. The load transfer 

distribution can be obtained from a static load test where strain gages or telltale rods 
220

Figure 7-5 Typical load transfer profiles. 

are attached to a pile at different depths along the pile shaft. Figure 7-5 shows the 

nominal resistance, Rn, in the pile plotted against depth. The shaft resistance 

transferred to the soil is represented by Rs, while Rp represents the resistance at the 

pile toe. In Figure 7-5(a), the load transfer distribution for a pile with no shaft 

resistance is illustrated. In this case the full axial load at the pile head is transferred 

to the pile toe. In Figure 7-5(b), the axial load versus depth for a uniform shaft 

resistance distribution typical of a cohesive soil is illustrated. Figure 7-5(c) presents 

the axial load in the pile versus depth for a triangular shaft resistance distribution 

and is typical of cohesionless soils.
221

7.1.4 Effective Stress

The vertical effective stress at a given depth below ground surface is the vertical 

stress at that depth due to the weight of the overlying soils. An Effective Stress 

Diagram plots the vertical effective stress versus depth, and is used in many static 

resistance and settlement calculations. An understanding of how to construct and 

use an Effective Stress Diagram is therefore important.

Information needed to construct an Effective Stress Diagram includes the total unit 

weight and thickness of each soil layer as well as the depth of the water table. The 

soil layer thickness and depth of the water table should be available from the project 

boring logs. The total unit weight of each soil layer may be obtained from density 

tests on undisturbed cohesive samples or estimated from Standard Penetration Test 

(SPT) N values in conjunction with the soil visual classification.

The first step in constructing an Effective Stress Diagram is to calculate the total 

vertical stress, σvo, versus depth. This is done by summing the product of the total 

unit weight times the layer thickness versus depth which was demonstrated in 

Section 5.2.2. Similarly, the pore water pressure, u, is summed versus depth by 

multiplying the unit weight of water, γw = 62.4 lb/ft3, times the water height. The 

vertical effective stress, σ’vo, at any depth is then the total vertical stress minus the 

pore water pressure at that depth.

The vertical effective stress at any depth is determined by summing the weights of 

all layers above that depth as follows:

1. For soil deposits above the static water table:

σ'vo= (total soil unit weight, γ)(thickness of soil layer above the desired depth).

2. For soil deposits below the static water table:

σ’vo = (total soil unit weight, γ)(depth) - (unit weight of water, γw )(height of 

water, hw).

The design water table elevation should be carefully selected. It should compared to 

the water level determined in the subsurface exploration and consider seasonal 

variations. 

Figures 7-6 and 7-7 present examples of Effective Stress Diagrams for cases where 

the water table is above and below the ground surface level.
222

Figure 7-6 Effective stress diagram – water table below ground surface. 

Figure 7-7 Effective stress diagram – water table above ground surface. 
223

7.1.5 Resistance Factors

As discussed in Chapter 2, limit state design for highway structures in the U.S. has 

been mandated since 2007, and is specified to achieve objectives of safety, 

serviceability, constructability, and economy. Resistance factors applicable to driven 

pile design are contained in AASHTO (2014) Table 10.6.5.2.3-1. Section 7.2 of this 

chapter reviews nominal resistance methods for the strength limit state, while 

Section 7.3 describes lateral deflection and group settlement for the service limit 

state. Section 7.4 discusses extreme limit state design which must address strength 

and serviceability requirements for the respective factored loads and performance 

requirements.

Resistance factors should not be viewed as a direct replacement for the factor of 

safety previously used in the allowable stress design (ASD) platform. As discussed 

in Chapter 2, a resistance factor is a statistically based multiplier applied to the 

nominal resistance determined from a specific analysis method or analysis 

procedure. It may be a value determined by national practice (i.e. AASHTO) or it 

may be a locally calibrated value determined from past practice, databases, and 

correlation studies. Similarly, load factors are applied to address the uncertainty on 

the load side, with the load factor also acting as a statistically based multiplier on the 

force effect. Both the load and resistance factors are tied to a target reliability index, 

β, which quantifies the probability of non-performance or failure.

Resistance factors to determine the factored resistance via static analysis 

calculations are summarized in Table 7-1. Several static analysis methods are 

documented in the literature. However, only those most commonly used in the US 

transportation industry will be presented in this chapter. Table 7-1 includes several 

static analysis methods that are presented in this manual that do not have calibrated 

resistance factors in AASHTO. The presented resistance factors are the best 

guidance available at the present time. Further research and calibration studies are 

needed in this area. Brown and Thompson (2015) noted that the AASHTO

resistance factor calibration study considered few large diameter pipe piles. 

Therefore the designer should consider this factor if performing analyses for piles 

larger than 24 inches in diameter regardless of pile type. 

Where a static analysis method is used to determine nominal resistance, the 

factored resistance, RRrr, shall be less than or equal to the sum of the shaft resistance 

multiplied by the resistance factor associated with the shaft resistance computation 

method plus the toe resistance multiplied by the resistance factor for the toe 

resistance computation method as presented in Equation 7-3. 
224

 Eq. 7-3 

Where: 

Rr = factored resistance (kips).

φstat = resistance factor (based on the static analysis method).

Rn = nominal resistance (kips).

Table 7-1 Resistance Factors for Static Analysis Methods Presented in this 

Manual (modified from AASHTO 2014)

The pile shaft resistance, Rs, and pile toe resistance, Rp, can further be expressed in 

terms of unit resistance values as shown in Equations 7-4 and 7-5. 

 Eq. 7-4 
225

 Eq. 7-5 

Where: 

fs = unit shaft resistance over the pile surface area (ksf). 

As = pile shaft surface area (ft2).

qp = unit toe resistance over the pile toe area (ksf).

Ap = pile toe area (ft2

).

The range in the resistance factors has primarily depended upon the reliability of the 

particular analysis method with consideration of the following items.

1. The level of confidence in the input parameters. This is a function of the type 

and extent of the subsurface exploration and laboratory testing program. It is 

assumed that AASHTO Section 10.4 on soil and rock properties is followed.

2. Variability of the soil and rock deposits. 

3. Method of static analysis. Resistance factors for some static analysis methods 

are tied to specific correlation procedures for soil properties.

4. Effects of and consistency of the proposed pile installation method.

AASHTO (2014) Article C10.5.5.2.3 notes that where the nominal resistance is 

determined by static load test, dynamic testing, wave equation analysis or dynamic 

formulas, the uncertainty in the nominal resistance is solely due to the reliability of 

the field resistance determination method. Therefore, the resistance factor for the 

field method, φdyn, should be used to determine the number of piles of a given 

factored resistance needed to resist the factored loads in the strength limit state. 

Table 7-2 summarizes resistance factors for field resistance determination methods. 

Individual chapters provide expanded details on the field test methods as well as 

associated performance, analysis, and interpretation information. Chapter 9 covers 

static load tests, Chapter 10 dynamic pile testing, Chapter 11 rapid load tests, 

Chapter 12 wave equation analysis, and Chapter 13 dynamic formulas.

If the pile foundation design consists of a small group (i.e. less than 3 piles per 

substructure unit in this manual) AASHTO (2014) recommends the resistance 

factors in Tables 7-1 and 7-2 be reduced by 20% to reflect the limited ability of the 

small group to accommodate the overstressing of one pile. The definition of a small 

group ranges from 2 or 3 according to Isenhower and Long (1997) to 5 according to 

Paikowsky et al. (2004).
226

Table 7-2 Resistance Factors for Field Determination Methods 

(after AASHTO 2014)
227

7.1.6 Interdiscipline Communication and Coordination

The design and construction process requires clear and concise communication 

among the project professionals practicing in the structural, geotechnical, geologic, 

hydraulic, and construction fields. Figure 2-3 in Chapter 2 presented a flow chart of 

the driven pile design and construction process that highlighted the major 

corroboration areas required in LRFD design and construction of surface 

transportation projects. The content of the interdiscipline communication was further 

detailed in Tables 2-4 and 2-5. Effective communication is essential to achieve a 

cost effective foundation meeting all of the strength, extreme, and service limit state 

design requirements.

7.2 STRENGTH LIMIT STATES

The strength limit state ensures local and global strength and stability against 

statistically significant load combinations occurring during the structure design life. 

Strength limit state design includes an evaluation of the nominal geotechnical and 

structural resistances as well as the loss of lateral and vertical support in the design 

flood event due to scour. Geotechnical aspects of strength limit state design 

includes:

• axial compression resistance of single piles, 

• axial compression resistance of pile groups,

• uplift resistance of single piles, 

• uplift resistance of pile groups,

• bearing stratum punching failure, and

• constructability including drivability.

7.2.1 Determination of Nominal Resistance for Single Piles

7.2.1.1 General

Numerous static analysis methods are available for calculating the nominal 

geotechnical resistance of a single pile. Section 7.2.1.3 of this chapter will detail 

analysis methods for piles in cohesionless, cohesive, and layered soil profiles using 

readily available SPT or laboratory test information. Additional methods based on 

cone penetration test results are also presented. As noted earlier, designers should 

fully understand the basis for, the limitations of, and the applicability of a chosen 

method. The selected method should also have a proven agreement with full scale 
228

field results in soil conditions similar to the project being designed, with the pile type 

being evaluated, and the pile installation conditions (impact driving, vibratory driving, 

etc.) to be used. The AASHTO resistance factors for some static analysis methods 

require specific procedures for determining soil strength properties.

To perform a static analysis, a pile depth is iteratively assumed based on 

geomaterial design parameters and selected design methods. This manual presents 

several design methods with AASHTO (2014) specified resistance factors. 

However, regional geologic settings or construction control techniques may offer 

unique conditions not accounted for in these provided methods, therefore reliability 

calibrations for design methods and resistance factors are encouraged, and may 

supersede the presented guidelines herein if justified. Table 7-3 compares the static 

analysis methods presented in AASHTO (2014) design specifications with those 

contained in this manual. Methods shaded in gray are presented in both documents 

and have a static analysis resistance factor, φstat. Methods without a static analysis 

resistance factor require local calibration or must rely on the resistance factor for the 

field verification method.

Table 7-3 Summary of Static Analysis Methods in GEC-12 and AASHTO (2014)

for Determination of Nominal Resistance 

Analysis Method Soil

Type

Soil 

Information

Required

Presented 

in 

GEC-12

Presented 

in 2014 

AASHTO

Code

AASHTO

φstat

Meyerhof (1976) Cohesionless SPT N No Yes 0.30

Nordlund (1963) Cohesionless φ’ Yes Yes 0.45

α-method (1980) Cohesive su Yes Yes 0.35

β-method (1951)(1979)* Cohesive su No Yes 0.25

λ-method (1972) Cohesive su No Yes 0.40

API RP2A (1993) Mixed su, φ’ Yes No - - - 

β-method (1991)** Mixed φ’ Yes No Differs1

Brown (2001) Mixed SPT N Yes No - - - 

Elsami & Fellenius (1997) Mixed CPTu Yes No - - - 

Schmertmann (1975) Mixed CPT Yes Yes 0.50
229

Notes: φ’ = effective stress friction angle

 su = undrained shear strength

SPT = standard penetration test

CPT = cone penetration test

1

 = β-method in AASHTO uses Skempton (1951), Ersig and Kirby (1979)*;

β-method in GEC-12 based on Fellenius (1991)**

7.2.1.2 Static Analysis Overview

Agencies often specify the use of select design methodologies based on experience. 

The design methods presented in this chapter should be used based on applicability 

to soil conditions, pile type and laboratory testing or subsurface exploration in-situ 

data. Tables 7-4 and 7-5 present an overview of design methods commonly used 

for calculating the nominal resistance of piles in cohesionless and cohesive soils, 

respectively. In layered profiles, the nominal resistance can be calculated by using

the applicable cohesionless and cohesive static analysis methods in appropriate soil 

layers. Table 7-3 identified additional methods that may be used to estimate the 

nominal resistance in mixed profiles as well as commonly used CPT methods.

Static analysis methods are typically used to estimate the required pile length for a 

given nominal resistance to establish pile length quantities in the contract 

documents. The estimated pile lengths and nominal resistance are then confirmed 

during construction using a field resistance determination method. A driving criterion 

as discussed in Chapter 17 is established from these results and used to install the 

remaining production piles. In some cases, static analyses alone are used as the 

resistance determination method and piles are driven to a predetermined toe 

elevation. In this latter case, site variability must be addressed in the design pile 

lengths. Regardless of the static analysis method used, the designer should clearly 

understand the applications and limitations of the selected method, the soil strength 

parameters needed to properly use the method, and the procedures and/or 

correlations used to determine those soil parameters for the method.

The FHWA recommended static analysis methods are the Nordlund method in 

cohesionless soils (Section 7.2.1.3.1), the α-method in cohesive soils (Section 

7.2.1.3.2), and the API method for large diameter pipe piles (Section 7.2.1.3.3). The 

Nordlund and α-method are recommended methods based on FHWA experience 

with their reliability in estimating pile length and the associated nominal resistance 

for conventional pile types in most subsurface conditions. The API method was 

specifically developed for large diameter pipe piles, and thus is the recommended 

static analysis method for evaluating highway structure foundations of this pile type.
230

Table 7-4 Methods of Static Analysis for Piles in Cohesionless Soils

Method Approach Method of 

Obtaining 

Design 

Parameters

Advantages Disadvantages Remarks 

Nordlund 

Method.

SemiEmpirical. Charts 

provided by 

Nordlund.

Estimate of 

soil friction 

angle is 

needed.

Allows for 

increased shaft 

resistance of 

tapered piles 

and includes 

effects of pilesoil friction 

coefficient for 

different pile 

materials. 

No limiting 

value on unit 

shaft

resistance is 

recommended. 

Soil friction 

angle often 

estimated from 

SPT data. 

Limit on pile 

sizes. 

Good approach 

to design that is 

widely used. 

Method is 

based on field 

observations. 

Details 

provided in 

Section 

7.2.1.3.1.

API RP2A. Empirical, 

effective 

stress 

analysis. 

Nq selected 

from Table 7-8

based on soil 

type. 

Developed 

specifically for 

large diameter 

open end pipe.

Application to 

non-LDOEPs 

is limited.

Used almost 

exclusively for 

offshore pile 

design. 

Effective 

Stress 

Method.

Semiempirical. β and Nt

selected 

based on soil 

classification 

and estimated 

friction angle.

β value 

considers pilesoil friction 

coefficient for 

different pile 

materials. Soil 

resistance 

related to 

effective vertical 

stress. 

Results 

affected by 

range in β 

values and in 

particular by 

range in Nt

chosen. 

Good approach 

for design. 

Details 

provided in 

Section 

7.2.1.3.3.

Brown 

Method. 

Empirical. Results of SPT 

tests based of 

N60 values.

Widespread use 

of SPT test and 

input data 

availability. 

Simple method

to use.

Relies solely 

on N60 values, 

which may not 

always be 

available.

Simple method 

based on 

correlations 

with 71 static 

load test 

results. Details 

provided in 

Section 

7.2.1.3.5.

Methods 

based on 

Cone 

Penetration 

Test (CPT) 

data.

Empirical. Results of

CPT tests.

Testing analogy 

between CPT 

and pile. 

Reliable 

correlations and 

reproducible 

test data.

Limitations on 

pushing cone 

into dense 

strata. 

Good approach 

for design. 

Details 

provided in 

Sections 

7.2.1.3.6 and 

7.2.1.3.7.
231

Table 7-5 Methods of Static Analysis for Piles in Cohesive Soils

Method Approach Method of 

Obtaining

Design 

Parameters

Advantages Disadvantages Remarks 

α-Method

(Tomlinson 

Method).

Empirical, 

total stress 

analysis. 

Undrained 

shear strength 

estimate of 

soil is needed. 

Adhesion 

calculated 

from Figures 

7-17 and 7-18. 

Simple 

calculation 

from 

laboratory 

undrained 

shear strength 

values to 

adhesion.

Wide scatter in 

adhesion 

versus 

undrained 

shear strengths 

in literature. 

Limits on su 

strengths in soft 

and medium 

cohesive soils. 

Widely used 

method 

described in 

Section 

7.2.1.3.2. 

API RP2A. Empirical, 

effective 

stress 

analysis. 

Undrained 

shear strength 

estimate of 

soil is needed. 

Developed 

specifically for 

large diameter 

open end pipe 

piles. 

Application to 

non-LDOEPs is 

limited. 

Used almost 

exclusively for 

offshore pile 

design. 

Effective 

Stress 

Method.

SemiEmpirical, 

based on 

effective 

stress at 

failure. 

β and Nt

values are 

selected from 

Table 7-9

based on 

drained soil 

strength 

estimates. 

Ranges in β 

and Nt values 

for most 

cohesive soils 

are relatively 

small. 

Range in Nt

values for hard 

cohesive soils 

such as glacial 

tills can be 

large. 

Good design 

approach 

theoretically 

better than 

undrained 

analysis. 

Details in 

Section 

7.2.1.3.3.

Methods 

based on 

Cone 

Penetration 

Test data.

Empirical. Results of 

CPT tests.

Testing 

analogy 

between CPT 

and pile. 

Reproducible 

test data.

Cone can be 

difficult to 

advance in very 

hard cohesive 

soils such as 

glacial tills.

Good 

approach for 

design. 

Details in 

Section 

7.2.1.3.6 and 

Section 

7.2.1.3.7.
232

7.2.1.3 Nominal Resistance of Single Piles in Soils

The nominal resistance of a single pile is taken as the sum of shaft and toe 

resistances (Rn = Rs + Rp). The calculation assumes that the shaft resistance and 

toe resistance can be determined separately and that these two factors do not affect 

each other. Many analytical and empirical methods have been developed for 

estimating the nominal resistance of piles using this approach. For typical pile sizes 

18 inches and smaller in diameter or width, the nominal resistance is often 

calculated using the Nordlund Method in cohesionless soils and the α-method in 

cohesive materials. The Nordlund Method is described in Section 7.2.1.3.1 and the 

α-method is detailed in Section 7.2.1.3.2. It should be noted that piles larger than 18 

inches were not in the correlation database for either the Nordlund Method or the αmethod. In layered soil profiles, the nominal resistance can be calculated by 

combining these methods in cohesionless and cohesive layers as applicable. Table 

7-3 identified additional methods that may be used to estimate the nominal

resistance in mixed profiles as well as commonly used CPTu and CPT methods.

7.2.1.3.1 Nordlund Method – Cohesionless Soils

The Nordlund Method (1963) is based on field observations and considers the shape 

of pile taper and its soil displacement in calculating the shaft resistance. This 

method also accounts for the differences in soil-pile coefficient of friction for different 

pile materials, and is based on the results of load test programs in cohesionless 

soils. Several pile types were used in these test programs including timber, H, 

closed end pipe, Monotube, and Raymond step taper piles. These piles, which were 

used to develop the method's design curves, had pile widths generally in the range 

of 10 to 20 inches. The larger pile types used today including large diameter open 

end pipe piles, concrete cylinder piles, 24 inch and greater square prestressed 

concrete piles as well as16 and 18 inch H-pile sections are not in the calibration 

database. The Nordlund Method tends to over predict the nominal resistance for 

piles widths larger than 24 inches. Alternative static analysis methods should be 

evaluated for these other pile types and larger pile sizes.

According to the Nordlund Method, the shaft resistance is a function of the following 

variables: 

1. The friction angle of the soil.

2. The friction angle on the sliding surface.

3. The taper of the pile.

4. The effective unit weight of the soil.
233

5. The pile length.

6. The minimum pile perimeter.

7. The volume of soil displaced.

These factors are considered in the Nordlund equation as illustrated in Figure 7-8. 

The Nordlund Method equation for computing the nominal resistance of a pile is as

follows:

 Eq. 7-6 

Where: 

d = depth (feet).

D = embedded pile length (feet).

Kδ = coefficient of lateral earth pressure at depth d.

CF = correction factor for Kd when δ≠φ. 

σ’d = vertical effective stress at the center of depth increment d.

δ = friction angle between pile and soil.

ω = angle of pile taper from vertical.

Cd = pile perimeter at depth d (feet).

Δd = length of pile segment.

αt = dimensionless factor (dependent on pile depth width relationship).

N’q = bearing capacity factor.

Ap = pile toe area (feet).

σ’p = vertical effective stress at the pile toe (ksf).

For a pile of uniform cross section (ω=0) and embedded length D, driven in soil 

layers of the same effective unit weight and friction angle, the Nordlund equation 

becomes:

 Eq. 7-7 

The soil angle of internal friction, φ, influences most of the calculations in the 

Nordlund method. In the absence of laboratory test data, φ can be estimated from 

corrected SPT N values. Section 5.2.3 provides several means of estimating soil 

friction angle based on these corrected SPT N values. Figures 7-8 to 7-14 should 

be used to include associate parameters for design. In addition, Table 7-6 and 7-7

provide factors to evaluate the coefficient of lateral earth pressure with depth.
234

Nordlund developed this method in 1963 with updates in 1979 and did not place a 

limiting value on the shaft resistance. However, Nordlund recommended that the 

vertical effective stress at the pile toe, σ’p, used for computing the pile toe resistance 

be limited to 3 ksf.

Figure 7-8 Nordlund’s general equation diagram for nominal resistance.
235

Figure 7-9 Relationship of δ/φ and pile soil displacement, V, for various pile types 

(after Nordlund 1979). 

Figure 7-10 Design curve for evaluating Kδ for piles when φ = 25 ̊ 

(after Nordlund 1979).
236

Figure 7-11 Design curve for evaluating Kδ for piles when φ = 30 ̊ 

(after Nordlund 1979).
237

Figure 7-12 Design curve for evaluating Kδ for piles when φ = 35 ̊ 

(after Nordlund 1979).

Figure 7-13 Design curve for evaluating Kδ for piles when φ = 40 ̊ 

(after Nordlund 1979).
238

Figure 7-14 Correction factor for Kδ when δ ≠ φ (after Nordlund 1979). 

Figure 7-15 Relationship between maximum unit toe resistance and friction angle 

for cohesionless soils (after Meyerhof 1976).
239

Figure 7-16 Chart for estimating αt coefficient and bearing capacity factor N’q 

(after Bowles 1977).
240

Table 7-6 Design Table for Evaluating Kδ for Piles when ω = 0° and 

Displaced Volume -V = 0.10 to 1.00 ft3/ft

Φ 0.10 0.20 0.30 0.40 0.50 0.60 0.70 0.80 0.90 1.00

25 0.70 0.75 0.77 0.79 0.80 0.82 0.83 0.84 0.84 0.85

26 0.73 0.78 0.82 0.84 0.86 0.87 0.88 0.89 0.90 0.91

27 0.76 0.82 0.86 0.89 0.91 0.92 0.94 0.95 0.96 0.97

28 0.79 0.86 0.90 0.93 0.96 0.98 0.99 1.01 1.02 1.03

29 0.82 0.90 0.95 0.98 1.01 1.03 1.05 1.06 1.08 1.09

30 0.85 0.94 0.99 1.03 1.06 1.08 1.10 1.12 1.14 1.15

31 0.91 1.02 1.08 1.13 1.16 1.19 1.21 1.24 1.25 1.27

32 0.97 1.10 1.17 1.22 1.26 1.30 1.32 1.35 1.37 1.39

33 1.03 1.17 1.26 1.32 1.37 1.40 1.44 1.46 1.49 1.51

34 1.09 1.25 1.35 1.42 1.47 1.51 1.55 1.58 1.61 1.63

35 1.15 1.33 1.44 1.51 1.57 1.62 1.66 1.69 1.72 1.75

36 1.26 1.48 1.61 1.71 1.78 1.84 1.89 1.93 1.97 2.00

37 1.37 1.63 1.79 1.90 1.99 2.05 2.11 2.16 2.21 2.25

38 1.48 1.79 1.97 2.09 2.19 2.27 2.34 2.40 2.45 2.50

39 1.59 1.94 2.14 2.29 2.40 2.49 2.57 2.64 2.70 2.75

40 1.70 2.09 2.32 2.48 2.61 2.71 2.80 2.87 2.94 3.0
241

Table 7-7 Design Table for Evaluating Kδ for Piles when ω = 0° 

and Displaced Volume -V = 1.0 to 10 ft3/ft

φ 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0 9.0 10.0

25 0.85 0.90 0.92 0.94 0.95 0.97 0.98 0.99 0.99 1.00

26 0.91 0.96 1.00 1.02 1.04 1.05 1.06 1.07 1.08 1.09

27 0.97 1.03 1.07 1.10 1.12 1.13 1.15 1.16 1.17 1.18

28 1.03 1.10 1.14 1.17 1.20 1.22 1.23 1.25 1.26 1.27

29 1.09 1.17 1.22 1.25 1.28 1.30 1.32 1.33 1.35 1.36

30 1.15 1.24 1.29 1.33 1.36 1.38 1.40 1.42 1.44 1.45

31 1.27 1.38 1.44 1.49 1.52 1.55 1.57 1.60 1.61 1.63

32 1.39 1.52 1.59 1.64 1.68 1.72 1.74 1.77 1.79 1.81

33 1.51 1.65 1.74 1.80 1.85 1.88 1.92 1.94 1.97 1.99

34 1.63 1.79 1.89 1.96 2.01 2.05 2.09 2.12 2.15 2.17

35 1.75 1.93 2.04 2.11 2.17 2.22 2.26 2.29 2.32 2.35

36 2.00 2.22 2.35 2.45 2.52 2.58 2.63 2.67 2.71 2.74

37 2.25 2.51 2.67 2.78 2.87 2.93 2.99 3.04 3.09 3.13

38 2.50 2.81 2.99 3.11 3.21 3.29 3.36 3.42 3.47 3.52

39 2.75 3.10 3.30 3.45 3.56 3.65 3.73 3.80 3.86 3.91

40 3.00 3.39 3.62 3.78 3.91 4.01 4.10 4.17 4.24 4.30
242

STEP BY STEP PROCEDURE FOR: “NORDLUND METHOD”

Steps 1 through 6 are for computing the shaft resistance and Steps 7 through 9 are 

for computing the pile toe resistance.

STEP 1 Delineate the soil profile into layers and determine the φ angle for each 

layer.

a. Construct Effective Stress Diagram using procedure described in 

Section 7.1.4.

b. Correct SPT field N values for vertical stress using methods from 

Section 5.1.1 to obtain corrected SPT N values, (N1)60. Delineate soil 

profile into layers based on corrected SPT N values.

c. Determine φ angle for each layer from laboratory tests or in-situ data.

d. In the absence of laboratory or in-situ test data, determine the 

average corrected SPT N value, (N1)60, for each soil layer and 

estimate φ angle from Table 5-5 in Chapter 5.

STEP 2 Determine δ, the friction angle between pile and soil based on displaced 

soil volume, V, and the soil friction angle, φ. 

a. Compute volume of soil displaced per unit length of pile, V.

b. Use Figure 7-9 with V and determine δ/φ ratio for pile type.

c. Calculate δ from δ/φ ratio.

STEP 3 Determine the coefficient of lateral earth pressure, Kδ, for each φ angle.

a. Determine Kδ for φ angle based on displaced volume, V, and pile 

taper angle, ω, using either Figure 7-10, 7-11, 7-12, or 7-13 and the 

appropriate procedure described in Step 3b, 3c, 3d, or 3e.

b. If the displaced volume is 0.1, 1.0 or 10.0 ft3

/ft, which correspond to 

one of the curves provided in Figures 7-10 through 7-13 and the φ
243

angle is one of those provided, Kδ can be determined directly from 

the appropriate figure.

c. If the displaced volume is 0.1, 1.0 or 10.0 ft3/ft which correspond to 

one of the curves provided Figures 7-10 through 7-13 but the φ angle 

is different from those provided, use linear interpolation to determine 

Kδ for the required φ angle. Tables 7-6 and 7-7 also provide 

interpolated Kδ values at selected displaced volumes versus φ angle 

for uniform piles (ω = 0).

d. If the displaced volume is other than 0.1, 1.0 or 10.0 ft3/ft which 

corresponds to one of the curves provided in Figures 7-10 through 

7-13 but the φ angle corresponds to one of those provided, use log 

linear interpolation to determine Kδ for the required displaced volume. 

Tables 7-6 and 7-7 also provide interpolated Kδ values at selected 

displaced volumes versus φ angle for uniform piles (ω = 0).

e. If the displaced volume is other than 0.1, 1.0 or 10.0 ft3/ft which 

correspond to one of the curves provided in Figures 7-10 through 

7-13 and the φ angle does not correspond to one of those provided, 

first use linear interpolation to determine Kδ for the required φ angle 

at the displaced volume curves provided for 0.1, 1.0 or 10.0 ft3/ft. 

Then use log linear interpolation to determine Kδ for the required 

displaced volume. Tables 7-6 and 7-7 also provide interpolated Kδ

values at selected displaced volumes versus friction angle for 

uniform piles (ω = 0).

STEP 4 Determine the correction factor, CF, to be applied to Kδ if δ ≠ φ. 

Use Figure 7-14 to determine the correction factor for each Kδ. Enter 

figure with φ angle and δ/φ value to determine CF. 

STEP 5 Compute the average vertical effective stress at the midpoint of each soil 

layer, σ’d (psf).

Note: A limiting value is not applied to σ’d. 
244

STEP 6 Compute the shaft resistance in each soil layer. Sum the shaft 

resistance from each soil layer to obtain the nominal shaft resistance, RRss

(kips). 

 Eq. 7-8 

(for uniform pile cross section)

For H-piles in cohesionless soils, the "box" area should generally be 

used for shaft resistance calculations. An additional discussion on the 

behavior of open pile sections is presented in Section 7.10.7.

STEP 7 Determine the αt coefficient and the bearing capacity factor, N’q, from the 

φ angle near the pile toe.

a. Enter Figure 7-16(a) with φ angle near the pile toe to determine αt

coefficient based on pile length to diameter ratio.

b. Enter Figure 7-16(b) with φ angle near the pile toe to determine

Nq. 

c. If φ angle is estimated from SPT data, compute the average 

corrected SPT N value, (N1)60, over the zone from the pile toe to 3 

diameters below the pile toe. Use this average corrected SPT N 

value to estimate φ angle near pile toe.

STEP 8 Compute the vertical effective stress at the pile toe, σ’p (ksf). 

 Note: The limiting value of σ’p is 3 ksf.

STEP 9 Compute the nominal toe resistance, RRpp (kips). 

 Eq. 7-9 

a. While limiting Rp = qL Ap where qL value is obtained from:

1. Entering Figure 7-15 with φ angle near pile toe determined from 

laboratory or in-situ test data.
245

2. Entering Figure 7-15 with φ angle near the pile toe estimated and 

the average corrected SPT N value, (N1)60, near toe as described 

in Step 7.

b. Use lesser of the two Rp values obtained in steps a and b.

For steel H and open end pipe piles, the selection of the pile toe area 

(steel area only or full cross sectional area of steel and enclosed soil) for 

toe resistance calculations should be based on past experience and 

local correlations with static load test results.

STEP 10 Compute the nominal resistance, Rn, from the sum of the shaft and toe 

resistances, Rn = Rs + Rp. 

STEP 11 Compute the factored resistance, Rr (kips), using select resistance 

factors provided in Section 7.1.5 and Equation 7-3. 

AASHTO (2014) provides a recommended resistance factor for this method in Table 

7-1 for the nominal resistance determined by static analysis. The nominal resistance 

at the strength limit state could also be determined using the resistance factor 

associated with the field verification method, φdyn, as recommended in Table 7-2. 

7.2.1.3.2 α-Method - Cohesive Soils

For piles in clay, a total stress analysis is often used where nominal resistance is 

calculated from the undrained shear strength of the soil. This approach assumes 

that the shaft resistance is independent of the vertical effective stress and that the 

unit shaft resistance can be expressed in terms of an empirical adhesion factor times 

the undrained shear strength.

The unit shaft resistance, fs, is equal to the adhesion which is the shear stress 

between the pile surface and the soil at failure. This may be expressed in equation 

form as:

uas == αsCf Eq. 7-10

Where: 

Ca = adhesion (ksf). 

su = undrained shear strength (ksf).

α = adhesion factor.
246

The adhesion factor α depends on the nature and strength of the clay, pile 

dimension, method of pile installation, and time effects. The values of α vary within 

wide limits and decreases rapidly with increasing shear strength.

It is recommended that Figure 7-17 generally be used for adhesion calculations, 

unless one of the special soil stratigraphy cases identified in Figure 7-18 is present 

at a site. In cases where either Figures 7-17 or 7-18 could be used, the 

inexperienced user should select and use the smaller value obtained from either 

figure. The undrained shear strengths addressed in the design charts ranges from 

approximately 0.5 ksf on the low end to 3.5 to 5.0 ksf on the high end, depending on 

the design chart. All users should confirm the applicability of a selected design chart 

in a given soil condition with local correlations between static resistance calculations 

and static load tests results. This is particularly for cases where the shear strength 

approaches either the upper or lower limit of the design chart.

In Figure 7-17, the pile adhesion, Ca, is expressed as a function of the undrained 

shear strength, su, with consideration of both the pile type and the embedded pile 

length, D, to pile diameter, b, ratio. The embedded pile length used in Figure 7-17

should be the minimum value of the length from the ground surface to the bottom of 

the clay layer, or the length from the ground surface to the pile toe.

Figure 7-18 presents the adhesion factor, α, versus the undrained shear strength of 

the soil as a function of unique soil stratigraphy and pile embedment. The adhesion 

factor from these soil stratigraphy cases should be used only for determining the 

adhesion in a stiff clay layer in that specific condition. For a soil profile consisting of 

clay layers of significantly different consistencies such as soft clays over stiff clays, 

adhesion factors should be determined for each individual clay layer.

The top graph in Figure 7-18 may be used to select the adhesion factor when piles 

are driven through a sand or sandy gravel layer and into an underlying stiff clay 

stratum. This case results in the highest adhesion factors as granular material is 

dragged into the underlying clays. The greater the pile penetration into the clay 

stratum, the less influence the overlying granular stratum has on the adhesion factor. 

Therefore, for the same undrained shear strength, the adhesion factor decreases 

with increased pile penetration into the clay stratum.

The middle graph in Figure 7-18 should be used to select the adhesion factor when 

piles are driven through a soft clay layer overlying a stiff clay layer. In this case, the 

soft clay is dragged into the underlying stiff clay stratum thereby reducing the 

adhesion factor of the underlying stiff clay soils. The greater the pile penetration into 
247

the underlying stiff clay soils, the less the influence the overlying soft clays have on 

the stiff clay adhesion factor. Therefore, the stiff clay adhesion factor increases with 

increasing pile penetration into the stiff clay soils.

Last, the bottom graph in Figure 7-18 may be used to select the adhesion factor for 

piles driven in stiff clays without any different overlying strata. In stiff clays, a gap 

often forms between the pile and the soil along the upper portion of the pile shaft. In 

this case, the shallower the pile penetration into a stiff clay stratum the greater the 

effect the gap has on the shaft resistance that develops. Hence, the adhesion factor 

for a given shear strength is reduced at shallow pile penetration depths and 

increased at deeper pile penetration depths. 

In highly overconsolidated clays, undrained shear strengths may exceed the upper 

limits of Figures 7-17 and 7-18. In these cases, it is recommended that adhesion 

factor, α, be calculated according to API Recommended Practice 2A (1993). Further 

information for this is provided in Section 7.2.1.3.3 of this chapter.

In the case of H-piles in cohesive soils, the shaft resistance should not be calculated 

from the surface area of the pile, but rather from the "box" area of the four sides. 

The shaft resistance for H-piles in cohesive soils consists of the sum of the 

adhesion, Ca, times the flange surface area along the exterior of the two flanges, 

plus the undrained shear strength of the soil, su, times the area of the two remaining 

sides of the "box", due to soil-to-soil shear along these faces. This computation can 

be approximated by determining the adhesion using the appropriate corrugated pile 

curve in Figure 7-17 and multiplying the adhesion by the H-pile "box" area. 

Additional information on the behavior of open pile sections is presented in Section 

7.10.7.

In clays with large shrink-swell potential, static resistance calculations should ignore 

the shaft resistance from the adhesion in the shrink-swell zone. During dry times, 

shrinkage will create a gap between the clay and the pile in this zone and therefore 

the shaft resistance should not be relied upon for long term support.
248

Figure 7-17 Adhesion values for piles in cohesive soils 

(after Tomlinson 1979).

The unit toe resistance, qp, in a total stress analysis for homogeneous cohesive soil 

can be expressed as:

 Eq. 7-11

Where:

Nc = bearing capacity factor. 

su = undrained shear strength (ksf).

The term Nc is a dimensionless bearing capacity factor which depends on the pile 

diameter and the depth of embedment. The bearing capacity factor, Nc, is usually 

taken as 9 for deep foundations. 
249

Figure 7-18 Adhesion factors for driven piles in clay (Tomlinson 1980).

On smaller piles in cohesive soils, the toe resistance contribution to the nominal 

resistance is a low percentage of the overall resistance and is therefore sometimes 

ignored. On larger piles, the movement required to mobilize the toe resistance is 
250

several times greater than that required to mobilize the shaft resistance. At the 

movement required to fully mobilize the toe resistance, the shaft resistance may 

have decreased to a residual value. These factors should be considered when 

performing nominal resistance assessments of various pile sections.

STEP BY STEP PROCEDURE FOR: "α-METHOD"

STEP 1 Delineate the soil profile into layers and determine the pile adhesion, Ca, 

from Figure 7-17 or adhesion factor, α, from Figure 7-18 for each layer.

Enter appropriate figure with the undrained shear strength of the soil, su, 

and determine pile adhesion or adhesion factor based on the embedded 

pile length in clay, D, and pile diameter, b. Use the curve for the 

appropriate soil and embedment condition. For highly overconsolidated 

clays where su exceeds the chart values, see Section 7.2.1.3.3 for 

guidance on α. 

STEP 2 For each soil layer, compute the unit shaft resistance, fs (ksf), using 

Equation 7-10. 

STEP 3 Compute the shaft resistance in each soil layer and the nominal shaft 

resistance, Rs (kips), from the sum of the layer shaft resistances. 

STEP 4 Compute the unit toe resistance, qp (ksf) using Equation 7-11. 

STEP 5 Compute the nominal toe resistance, Rp (kips). 

For open pile sections, refer to the discussion of pile plugging presented 

in Section 7.10.7.

STEP 6 Compute the nominal pile resistance, Rp, from the sum of the shaft and 

toe resistances, Rn = Rs + Rp. 
251

STEP 7 Compute the factored resistance, Rr (kips), using the resistance factors 

provided in Section 7.1.5 and Equation 7-3.

AASHTO (2014) provides a recommended resistance factor for this method in Table 

7-1 for the nominal resistance determined by static analysis. The nominal resistance 

in the strength limit state could also be determined using the resistance factor 

associated with the field verification method, φdyn, as recommended in Table 7-2. 

7.2.1.3.3 P2A Method – Mixed Soil Profiles

The American Petroleum Institute (API) provides a static analysis procedure design 

developed for offshore construction. These projects almost exclusively use large 

diameter, open-end, steel pipe piles which are driven by impact hammer to final 

penetration (API 1993). In NCHRP Report 20-05 on large diameter pipe piles by 

Brown and Thompson (2015), large diameter open end pipe piles are defined as 

having a diameter of 36 inches or greater. Large diameter open end pipe piles can 

be either steel pipe piles or concrete cylinder piles. Recently, large diameter open 

end pipe pile usage has increased significantly on transportation projects. This has 

heightened the need for more accurate nominal resistance estimates on these larger 

piles.

Similar to other design methodologies, the API approach gives consideration to soil 

type. For cohesive soil, shaft resistance, fs, can be determined from Equation 7-12.

 Eq. 7-12

Where: 

α = dimensionless adhesion factor.

su = undrained shear strength at the location in question (ksf).

The factor αff varies based on effective stress and may be calculated using 

Equations 7-13 through 7-15 below.

 when Eq. 7-13

 when Eq. 7-14

 

Eq. 7-15
252

Where:

α = dimensionless factor.

su = undrained shear strength at the sample depth (ksf). 

σ’vo = vertical effective stress at the sample depth (ksf). 

An α value of 1.0 is recommended for unconsolidated clays. Reductions in 

resistance may be practical for very long piles where residual soil strength values 

are approached due to extended driving and subsequent soil displacement. For 

these cases, API (1993) recommends the use of engineering judgment.

The unit toe resistance, qp, for piles in cohesive soil can be determined using 

Equation 7-16. 

up q = 9s Eq. 7-16

Because of the dynamic events during driving, the pile may be installed in an 

unplugged condition, however when static loads are applied, a plugged condition 

may exist. Consideration should be given to these cases when performing a static 

analysis and drivability study.

When installing piles in cohesionless soils, the unit shaft resistance may be 

determined with Equation 7-17.

s = Kf δ 'vo tanδσ Eq. 7-17

Where: 

Kδ = coefficient of lateral earth pressure.

σ’vo = vertical effective stress (ksf).

δ = friction angle between the soil and pile wall.

API (1993) notes that assuming Kδ =0.8 for both tension and compression loading of 

unplugged, open ended pipe pile is appropriate. In addition, for the plugged or 

closed end case the assumption of Kδ =1.0 is recommended.

The unit toe resistance, qp for piles in cohesionless soils may be determined using 

the following relationship:

p vo Nq q = σ ' Eq. 7-18
253

Where:

σ’vo = vertical effective stress (ksf). 

Nq = dimensionless bearing capacity factor. 

Table 7-8 is modified from API (1993) and presents guidelines for design parameters 

using steel pipe piles installed in cohesionless soils. Subsurface exploration and 

testing data for sites in question should be considered where necessary, in lieu of 

these guidelines.

Table 7-8 Design Parameter Guidelines for Cohesionless Siliceous Soil 

(after API 1993)

Density Soil

Soil-Pile 

Friction 

Angle, δ

Limiting 

Unit Shaft 

Resistance, 

(ksf) 

Nq

Limiting 

Unit Toe 

Resistance, 

(ksf) 

Very loose

Loose

Medium

Sand

Sand-Silt*

Silt

15 1.0 8 40

Loose

Medium

Dense

Sand

Sand-Silt*

Silt

20 1.4 12 60

Medium

Dense

Sand

Sand-Silt*

25 1.7 20 100

Dense

Very Dense

Sand

Sand-Silt*

30 2.0 40 200

Dense

Very Dense

Gravel

Sand

35 2.4 50 250

* - In sand-silt soils (soils with significant fractions of both sand and silt), the strength 

values generally increase with increasing sand fractions and decrease with 

increasing silt fractions.
254

STEP BY STEP PROCEDURE FOR: “API METHOD”

STEP 1 Delineate the soil profile into cohesive and cohesionless layers.

 For cohesive layers, determine the average undrained shear strength, 

su, for the layer and the vertical effective stress σ’vo, at the mid-point of 

the layer. Determine the adhesion factor, α, based on the su/ σ’vo ratio 

and Equation 7-13 or 7-14.

 For cohesionless layers, determine the soil-pile friction angle, δ, for the

 layer and the vertical effective stress σ’vo, at the mid-point of the layer. 

 Determine the coefficient of lateral earth pressure, Kδ, based on the pile 

 type and loading direction.

STEP 2 For each soil layer, compute the unit shaft resistance, fs (ksf) using 

Equation 7-12 in cohesive soil layers and Equation 7-17 in cohesionless 

soil layers.

 Note: A limiting unit shaft resistance is applied in cohesionless soils.

STEP 3 Using Equation 7-4, compute the shaft resistance in each soil layer and 

the nominal shaft resistance, Rs, from the sum of the shaft resistance

from each layer.

STEP 4 Compute the unit toe resistance, qp (ksf). 

 For piles terminated in a cohesive layer, use the average undrained 

shear strength, su, in Equation 7-16.

 For piles terminated in a cohesionless layer, determine bearing capacity 

factor, Nq, from Table 7-8 and Equation 7-18.

 Note: A limiting unit toe resistance is applied in cohesionless soils.
255

STEP 5 Compute the nominal toe resistance, Rp (kips) using Equation 7-5.

 For open pile sections, refer to the discussion of pile plugging presented 

in Section 7.10.7.

STEP 6 Compute the nominal pile resistance, Rn from the sum of the shaft and 

toe resistances, Rn = Rs + Rp. 

STEP 7 Compute the factored resistance, Rr (kips), using the applicable 

resistance factor for the field verification method.

In NCHRP Report 507, Paikowsky (2004) evaluated the reliability of the API method 

in cohesive soils but not in sands or mixed profiles. AASHTO (2014) also does not 

provide a recommended resistance factor for the nominal resistance determined by 

the API method. Therefore, the nominal resistance at the strength limit state should 

be determined using the resistance factor associated with the field verification 

method, φdyn, as recommended in Table 7-2. 

7.2.1.3.4 Effective Stress β-Method – Mixed Soil Profiles

Static resistance calculations in cohesionless, cohesive, and layered soils can also 

be performed using an effective stress method. Effective stress methods were 

developed to model the long term drained shear strength conditions. Therefore, the 

effective stress friction angle, φ', should be used in parameter selection.

In an effective stress analysis, the unit shaft resistance is calculated from the 

following expression:

 Eq. 7-19

Where:

β = Bjerrum-Burland beta coefficient = Kδδ tan δ. 

σ’vo = average vertical effective stress along the pile shaft (ksf). 

Kδ = coefficient of lateral earth pressure. 

δ = friction angle between pile and soil.

The unit toe resistance is calculated from:

 Eq. 7-20
256

Where:

Nt = toe resistance coefficient.

σ’p = vertical effective stress at the pile toe (ksf).

Recommended ranges of β and Nt coefficients as a function of soil type and φ' angle 

from Fellenius (2014) are presented in Table 7-9. Fellenius notes that factors 

affecting the β and Nt coefficients consist of the soil composition including the grain 

size distribution, angularity and mineralogical origin of the soil grains, the original soil 

density and density due to the pile installation technique, the soil strength, as well as 

other factors. While Table 7-9 provides the approximate range of β coefficients for 

various soil types, Fellenius noted that values can deviate significantly from those in 

the table. β coefficients seldom exceed 1.0.

For sedimentary cohesionless deposits, Fellenius states Nt ranges from about 30 to 

a high of 120. In very dense deposits such as glacial tills, Nt can be much higher, 

but can also approach the lower bound value of 30. In clays, Fellenius notes that 

the toe resistance calculated using an Nt of 3 is similar to the toe resistance 

calculated from a traditional analysis using undrained shear strength. Therefore, the 

use of a relatively low Nt coefficient in clays is recommended unless local 

correlations suggest higher values are appropriate.

As with any design method, the user should also confirm the appropriateness of 

selected coefficients in a given soil condition with local correlations between static 

resistance calculations and load tests results.

It should be noted that the effective stress method places no limiting values on either 

the shaft or toe resistance.

Table 7-9 Approximate Range of β and Nt Coefficients 

(after Fellenius 2014) 

Soil Type φ' β Nt

Clay 25-30 0.15-0.35 3-30

Silt 28-34 0.25-0.50 20-40

Sand 32-40 0.30-0.90 30-150

Gravel 35-45 0.35-0.80 60-300
257

STEP BY STEP PROCEDURE FOR: “EFFECTIVE STRESS β-METHOD”

STEP 1 Delineate the soil profile into layers and determine φ' angle for each 

layer.

a. Construct Effective Stress Diagram using previously described 

procedure in Section 7.1.4.

b. Divide soil profile throughout the pile penetration depth into layers 

and determine the vertical effective stress, σ’vo, (ksf) at the midpoint 

of each layer.

c. Determine the φ' angle for each soil layer from laboratory or in-situ 

test data.

d. In the absence of laboratory or in-situ data for cohesionless layers, 

determine the average corrected SPT N value, (N1)60, for each layer 

and estimate φ' angle from a select method in Section 5.2.3 of 

Chapter 5. 

STEP 2 Select the β coefficient for each soil layer. 

a. Use local experience to select β coefficient for each layer.

b. In the absence of local experience, use Table 7-9 to estimate β

coefficient from φ' angle for each layer.

STEP 3 For each soil layer compute the unit shaft resistance, fs (ksf) using 

Equation 7-19. 

STEP 4 Compute the shaft resistance in each soil layer and the nominal shaft 

resistance, Rs (kips), from the sum of the shaft resistance from each soil 

layer.

Refer to Section 7.10.7 for additional information on the behavior of open 

pile sections.
258

STEP 5 Compute the unit toe resistance, qp(ksf) using Equation 7-20. 

a. Use local experience to select Nt coefficient.

b. In the absence of local experience, estimate Nt from Table 7-9 based 

on φ' angle.

c. Calculate the vertical effective stress at the pile toe, σ’p (ksf). 

STEP 6 Compute the nominal toe resistance, Rp (kips) using Equation 7-5.

 For open pile sections, refer to the discussion of pile plugging presented 

in Section 7.10.7.

STEP 7 Compute the nominal pile resistance, Rn from the sum of the shaft and 

toe resistances, Rn = Rs + Rp. 

STEP 8 Compute the factored resistance, Rr (kips), using resistance factors 

provided in Section 7.1.5 and Equation 7-3.

AASHTO (2014) does not provide a recommended resistance factor for the nominal 

resistance determined by the β-Method by Fellenius. Therefore, the nominal 

resistance at the strength limit state should be determined using the resistance 

factor associated with the field verification method, φdyn, as recommended in Table

7-2. 

7.2.1.3.5 Brown Method – Mixed Soil Profiles – SPT Data

The Brown Method (2001) is a simple empirical method that uses SPT N60 values 

for calculating unit shaft resistance and unit toe resistance values. The Brown 

Method was based on resistance correlations with 71 static load tests from Caltrans 

projects in a wide variety of soil types. The pile types included closed end pipe, 

open end pipe, H-piles, and precast concrete piles. The method considers 

compression and uplift loading as well as pile installation method (impact driving and 

partial vibratory installation).

Brown reported that the average unit shaft resistance, fs, (ksf) is:

 Eq. 7-21 
259

Where:

Fvs = factor for pile driving method (1.0 for impact or 0.68 for vibratory). 

Ab = Brown’s regression analysis factor based on soil type in (ksf). 

Bb = Brown’s regression analysis factor based on soil type in (ksf/bpf). 

N60 = SPT N-Value corrected for 60% energy transfer.

Limits on the value of N60 were also recommended. If N60 is greater than 50, a value 

of 50 should be used and if N60 is less than 3, use 3. Brown recommended that the 

shaft resistance, Rs, be calculated by multiplying the unit shaft resistance times the 

pile perimeter with the “box” perimeter used for H-piles rather than the actual pile/soil 

contact area and for open end pipe piles only the external surface area. Regression 

factors Ab and Bb, as well as the pile installation factor, Fvvss, are given in Table 7-10. 

Brown (2001) recommended that for impact driven piles, the unit toe resistance, qqpp

(ksf) is calculated: 

 Eq. 7-22

For vibratory installed piles this unit toe resistance should then be multiplied by 0.56. 

The pile toe resistance, Rp, in kips is then calculated as follows:

 Eq. 7-23

Where:

Ap = cross sectional area of the pile material at pile toe (ft2). 

App = cross sectional area of soil plug (ft2) for open end pipe piles or Hpiles at pile toe.

Fp = plug mobilization factor, 0.42 for pipe piles or 0.67 for H-piles. 

Brown recommended the actual steel area at the pile toe be used for Ap on H-piles 

and open end pipe piles.

While the simplicity of Brown’s method is attractive, it is recommended that the 

method be used only for preliminary length estimates until a greater experience base 

is obtained with the method results. Caltrans continues to study and expand on 

Brown’s work as reported by Olson and Shantz (2004). AASHTO (2014) does not 

provide a recommended resistance factor for the nominal resistance determined by 

this method. Therefore, the resistance factor for the applicable field verification 

method, φdyn, should be used if this method is used.
260

Table 7-10 Input Factors for Brown’s Method

Loading 

Condition

Installation 

Method

Soil Type Fvs Ab

(ksf) 

Bb 

(ksf/bpf)

Compression Impact Clay to Sand 1.0 0.555 0.040

“ “ Gravelly Sand to 

Boulders 1.0 0.888 0.888

“ “ Rock 1.0 2.89 2.89

Tension Impact Clay to Sand 1.0 0.522 0.0376

“ “ Gravelly Sand to 

Boulders 1.0 0.835 0.0

“ “ Rock 1.0 2.71 0.0

“ Vibratory Clay to Sand 0.68 0.522 0.0376

“ “ Gravelly Sand to 

Boulders 0.68 0.835 0.0

“ “ Rock 0.68 2.71 0.0

7.2.1.3.6 Eslami and Fellenius Method – CPT Data

In the Eslami and Fellenius method, the unit shaft resistance is correlated to the 

average effective cone tip resistance with a shaft correlation coefficient applied 

based on the soil profile. The unit shaft resistance, fs, (ksf) is calculated from:

 Eq. 7-24

In Which:

 Eq. 7-25

 Eq. 7-26

Where:

Cs = shaft correlation coefficient from Table 7-11. 

qE = Eslami cone stress.

U2 = pore water pressure measured at cone shoulder.

qc = measured cone tip stress.

a = ratio between shoulder area (cone base) unaffected by the pore 

water pressure to total area.
261

Table 7-11 Cs Values for Eslami and Fellenius Method (after Fellenius 2014)

Soil Type Cs (%)

Soft sensitive soil 8.0

Clay 5.0

Silty clay, stiff clay and silt 2.5

Sandy silt and silt 1.5

Fine sand or silty sand 1.0

Sand 0.4

The unit toe resistance is computed using geometric averaging of the cone tip 

resistance over the influence zone at the pile toe, after the cone tip resistances have 

been corrected for pore pressure on the cone shoulder and effective stress. The 

zone of influence at the pile toe is based on the pile diameter, b, and ranges from 4b 

below the pile toe to 8b above the pile toe when the pile is installed through a weak 

zone overlying a dense zone. When a pile is driven through a dense soil into a weak 

soil, the zone of influence is from 4b below the pile toe to 2b above the pile toe. The 

unit toe resistance is calculated from:

 Eq. 7-27

Where:

Cp = toe correction coefficient equal to 1.0 in most cases.

 qEg = geometric average of the cone tip resistance in ksf over the influence 

zone after correction for pore pressure on shoulder and adjustment to 

effective stress. 

The toe correction coefficient is a function of the pile size since larger piles require 

greater movement to mobilize the pile toe resistance. For pile diameters greater 

than 16 inches, the toe correction coefficient should be calculated as follows:

 

Eq. 7-28

Where:

b = pile width or diameter (inches). 
262

STEP BY STEP PROCEDURE FOR: “ESLAMI AND FELLENIUS METHOD”

STEP 1 Obtain CPT test results with outputs of cone stress and pore pressure 

measurements.

STEP 2 For each depth increment, compute the unit shaft resistance, fs (ksf) 

using Equation 7-24 and coefficients provided in Table 7-11. 

STEP 3 Compute the shaft resistance in each soil layer and the nominal shaft 

resistance, Rs (kips), from the sum of the shaft resistance from each 

layer using Equation 7-4.

STEP 4 Compute the unit toe resistance, qp(ksf), using Equation 7-27. 

STEP 5 Compute the nominal toe resistance, Rp (kips) using Equation 7-5.

STEP 6 Compute the nominal pile resistance, Rn from the sum of the shaft and 

toe resistances, Rn = Rs + Rp. 

STEP 7 Compute the factored resistance, Rr (kips), using select resistance 

factors provided in Section 7.1.5 and Equation 7-3.

AASHTO (2014) does not provide a recommended resistance factor for this method. 

Therefore, the nominal resistance in the strength limit state should be determined 

using the resistance factor associated with the field verification method, φdyn, as 

recommended in Table 7-2. 

7.2.1.3.7 Nottingham and Schmertmann Method – CPT Data

One empirical procedure used in U.S. practice was derived from work originally 

published by Nottingham and Schmertmann (1975), and summarized in publication 

FHWA-TS-78-209, "Guidelines for Cone Penetration Test, Performance and Design" 

by Schmertmann (1978).
263

The nominal shaft resistance, Rs, in cohesionless soils may be derived from unit 

sleeve friction of the CPT using the following expression:

 Eq. 7-29

Where: 

Ks = ratio of unit pile shaft resistance to unit cone sleeve friction from 

Figure 7-20 as a function of the full penetration depth, D.

fs = average unit sleeve friction over the depth interval indicated by 

subscript. 

As = pile shaft surface area over fs depth interval.

b = pile width or diameter.

D = embedded pile length.

 0 to 8b = range of depths for segment from ground surface to a depth of 8b.

8b to D = range of depths for segment from a depth equal to 8b to the pile toe.

The transfer function, Ks, relating pile shaft resistance to CPT sleeve friction, varies 

as a function of total pile penetration (depth of embedment/pile diameter), pile 

material type, and type of cone penetrometer used. No limit was imposed on sleeve 

friction values in the procedure originally proposed by Nottingham and Schmertmann 

(1975).

If cone sleeve friction data is not available, Rs can be determined from the cone tip 

resistance as follows:

 Eq. 7-30

Where:

Cf = correction obtained from Table 7-12. 

qc = average cone tip resistance along the pile length.

As = pile shaft surface area (length).

For shaft resistance in cohesive soils, the nominal shaft resistance, Rs, is obtained 

from the sleeve friction values using the following expression:

 Eq. 7-31
264

Where:

α' = ratio of pile shaft resistance to cone sleeve friction, patterned after 

Tomlinson's α-method.

fs = unit sleeve friction. 

As = pile shaft surface area (length).

The value of α' varies as a function of sleeve friction, fs, as shown in Figure 7-20. It 

is expected that this method of calculating pile shaft resistance is less appropriate in 

sensitive soils as the friction sleeve of the cone encounters severely disturbed soils 

behind the cone tip.

The estimation of pile toe resistance is described in Figure 7-21. In essence, an 

elaborate averaging scheme is used to weigh the cone tip resistance values, from 8 

pile diameters above the pile toe, to as much as 3.75 pile diameters below the pile 

toe, favoring the lower cone tip resistance, qc, values within the depth range. The 

authors make reference to a "limit" value of qc between 105 to 315 ksf, that should 

be applied to the nominal unit pile toe resistance, qp, unless local experience 

warrants use of higher values. In the case of mechanical cone soundings in 

cohesive soils, the qp value is reduced by 40 percent to account for toe resistance 

effects on the base of the friction sleeve. As discussed in Section 7.10.7, careful 

consideration of soil plugging phenomena is needed in choosing the cross-sectional 

area over which qp is applied for low displacement open ended pipe and H-piles.

Table 7-12 CPT Cf VALUES

Type of Piles Cf 

Precast Concrete 0.012

Timber 0.018

Steel Displacement 0.012

Open End Steel Pipe 0.008
265

Figure 7-19 Penetrometer design curves for pile shaft friction in sand 

(after FHWA Implementation Package, FHWA-TS-78-209).

Figure 7-20 Design curve for pile shaft friction in clay

(after Schmertmann 1978).
266

Figure 7-21 Illustration of Nottingham and Schmertmann Procedure for estimating 

pile toe resistance (FHWA-TS-78-209).
267

STEP BY STEP PROCEDURE FOR: “NOTTINGHAM AND SCHMERTMANN 

METHOD”

STEP 1 Delineate the soil profile into layers using the cone tip resistance, qc, and 

sleeve friction, fs, values.

STEP 2 Compute the shaft resistance for each soil layer, Rs (kips). 

a. For piles in cohesionless soils, compute shaft resistance, Rs, using 

the average sleeve friction value for the layer, fs, and the K value. 

Note that K should be determined using the full pile penetration depth 

to diameter ratio from Figure 7-19, and not the penetration depth for 

the layer. Conversely, the depth d corresponds to the pile toe depth, 

or the depth to the bottom of the layer, whichever is less. For H-piles 

in cohesionless soils, the pile-soil surface area As, should be the 

"box" area.

For cohesionless layers below a depth of 8b, Equation 7-29 for shaft resistance in a 

layer reduces to:

 Eq. 7-32

Where:

Ks = ratio of unit pile shaft resistance to unit cone sleeve friction from 

Figure 7-19 as a function of the full penetration depth, D. 

fs = average unit sleeve friction over the depth interval indicated by 

subscript. 

As = pile shaft surface area over fs depth interval.

 

For piles in cohesionless soils without sleeve friction data, compute the shaft 

resistance from Equation 7-30. 

b. For piles in cohesive soils, compute the shaft resistance using the 

average sleeve friction value for the layer and Equation 7-31. 
268

STEP 3 Compute the shaft resistance in each soil layer and the nominal shaft 

resistance, Rs (kips), from the sum of the shaft resistance from each 

layer.

STEP 4 Compute the unit pile toe resistance, qp (ksf) using Equation 7-33. 

 Eq. 7-33

Where:

qc1 = average qc over a distance of xb below the pile toe (path 1-2-3 from 

Figure 7-21). Sum qc values in both the downward (path 1-2) and 

upward (path 2-3 direction. Use actual qc values along path 1-2 and 

the minimum path rule along path 2-3. Compute qc1 for x-values from 

0.7 to 3.75 below the pile toe and use the minimum qc1 value 

obtained.

qc2 = average qc over a distance of 8b above the pile toe (path 3-4 from 

Figure 7-21). Use the minimum path rule as for path 2-3 in the qc1

computations.

b = pile width or diameter.

D = embedded pile length.

STEP 5 Compute the nominal toe resistance, Rp (kips). 

For steel H and unfilled open ended pipe piles, use only the steel cross 

section area at the pile toe unless there is reasonable assurance and 

previous experience that a soil plug would form. For a plugged condition 

use the "box" area of the H-pile and the full cross section area for pipe 

pile. Additional information on the plugging of open pile sections is 

presented in Section 7.10.7.

STEP 6 Compute the nominal resistance, Rn from the sum of the shaft and toe 

resistances, Rn = Rs.+ Rp.

STEP 7 Compute the factored resistance, Rr (kips), using select resistance 

factors provided in Section 7.1.5 and Equation 7-3.

AASHTO (2014) provides a recommended resistance factor for this method in Table 

7-1 for the nominal resistance determined by static analysis. The nominal resistance 

in the strength limit state could also be determined using the resistance factor 

associated with the field verification method, φdyn, as recommended in Table 7-2. 
269

7.2.1.4 Nominal Resistance of Single Piles to Rock

Pile foundations on rock are normally designed to carry large loads. For pile 

foundations driven to rock, which include steel H-piles, pipe piles or precast concrete 

piles, the exact area in contact with the rock, the depth of penetration into rock, as 

well as the quality of rock in the immediate pile contact area are largely unknown. 

Pile installation can also alter the characteristics of the rock formation, further 

complicating design procedures.

The distinction between soft rock and hard rock is not well defined. The AASHTO 

definition of soft and hard rock for driven pile foundation design is limited. In 

general, soft rock is defined as rock that can be penetrated by pile driving and hard 

rock is defined as rock that cannot be penetrated. Hence, piles are driven into soft 

rock and driven to hard rock. In many intact rock classification systems, the 

transition between soft rock and hard rock occurs between an unconfined 

compression strength of 200 and 1000 ksf. Kulhawy (1991) proposed 400 ksf as the 

an unconfined compression strength denoting the transition between weak and 

moderate rock. AASHTO notes that a definition of hard rock based on measureable 

rock characteristics has not been widely accepted.

Likins and Goble (1978) reported on the results of an Ohio DOT and FHWA 

sponsored research study on HP 10x42 H-piles driven to shallow bedrock with 

multiple, similarly sized pile hammers. Static and dynamic load tests were 

conducted at the two sites, one with hard rock and one with soft rock. The H-piles 

were driven to a penetration resistance of 20 blows per inch on rock at both sites. At 

the hard rock site, hard limestone with an unconfined compression strength of up to 

1094 ksf was encountered around 22 feet. At the soft rock site, weathered shale 

with unconfined compressive strength of 172 to 316 ksf was encountered around 15 

feet. The A-36 steel H-piles at the hard limestone site were statically load tested to 

as high as 0.93Fy without reaching the Davisson load test failure criteria which is 

described in Chapter 9. At the weathered shale site, significantly lower nominal 

resistances were achieved with failure occurring between 0.28 to 0.71Fy. Relaxation 

occurred on the H-piles driven at the weathered shale site and higher nominal 

resistances were observed on piles that penetrated deeper into the shale. This 

study, even though it was performed with smaller hammers and piles then commonly 

used today, demonstrates that geotechnical aspects control soft rock design and 

structural aspects control hard rock design.

Tomlinson and Woodward (2015) state that piles driven to rock can create very high 

concentrated loads on the rock beneath the pile toe. The ability of the rock to 
270

support the concentrated load depends in part on the compressive strength of the 

rock, the frequency of fissures and joints in the rock mass, and whether the fissures 

and joints are tightly closed or are open and filled with weathered material. They 

further note that very high loads can be supported if the rock is strong and has 

closed joints or joints on a shallow angle to horizontal. Conversely, steeply inclined 

and open joints may provide little resistance as the rock beneath the pile toe slides 

until the joints are closed or the rock mass becomes locked together. The driving of 

piles through weak or broken rock to hard rock can shatter weak rock such that shaft 

resistance is significantly reduced or eliminated. The resulting concentrated load 

may be acceptable for strong intact rock but it may be excessive for a strong but 

closely jointed rock mass.

Published results as well as empirical values for nominal shaft resistances are 

presented in Table 7-13. These results may be used for preliminary estimating 

purposes or as a check of values obtained from field tests. However, they are not 

intended to be used as final design values without the user determining the 

applicability of the underlying method or the suitability of a reported nominal 

resistance value to a given site or geologic formation.

Table 7-13 Published Nominal Shaft Resistance Values in Weak Rock Materials

Rock Description Pile Type Nominal Unit Shaft 

Resistance (ksf)

Source

Moderately strong, slightly weathered, 

Slaty Mudstone H-pile 0.6 1.

Moderately strong, slightly weathered, 

Slaty Mudstone H-pile 3.3 1.

Faintly to moderately weathered , 

moderately strong to strong, Mudstone Pipe Pile 2.6 2.

Very weak, closely fissured argillaceous 

Siltstone (Mercia Mudstone)

Precast 

Concrete 2.7 3.

Very weak, coral detrital Limestone

(carbonate sandstone/siltstone) Pipe pile 0.9 2.

Limestone H-pile 24.0 4.

Weak calcareous Sandstone Pipe Pile 0.9 2.

Sandstone H-pile 20.0 4.

Shale H-pile 12.0 4.

Information sources: 1) George et al. (1976)

2) Tomlinson and Woodward (2015)

3) Leach and Mallard (1980)

4) Illinois DOT Geotechnical Pile Design Guide, AGMU 10.2 (2011)
271

Published results and recommended empirical values for nominal toe resistances 

are presented in Table 7-14. Once again these values may be used for preliminary 

estimating purposes or as a check of values obtained from field tests. However, 

they are not intended to be used as final design values without the user determining 

the applicability of the underlying method or the suitability of a reported nominal 

resistance value to a given site or geologic formation.

Table 7-14 Published Nominal Toe Resistance Values in Weak Rock Materials

Rock Description Pile Type Nominal Unit Toe

Resistance (ksf)

Source

Weak carbonate Siltstone/Sandstone 

(coral detrital limestone) N.A. 106.7 1.

Limestone H-pile 240.0 2.

Weak calcareous Sandstone Pipe Pile 62.6 1.

Sandstone H-pile 200.0 2.

Shale H-pile 120.0 2.

Information sources: 1) Tomlinson and Woodward (2015)

2) Illinois DOT Geotechnical Pile Design Guide, AGMU 10.2 (2011)

Based on the above discussion and information, the determination of nominal shaft 

and toe resistances of piles driven into or on rock is best made on the basis of static 

or dynamic load tests in conjunction with pile driving observations and local

experience. In general, the design of small diameter piles supported on fair to 

excellent quality rock will be controlled by their nominal structural resistance as 

described in Chapter 8. A pile on hard intact rock will fail structurally before hard 

intact rock failure. Piles supported in soft, weakly laminated or weathered rock

should be designed based on the results of static or dynamic load tests.

7.2.1.4.1 Piles Driven into Soft and Weak Rock

Tomlinson and Woodward (2015) note that the shaft resistance on piles driven into 

weak weathered rocks cannot always be calculated from the results of laboratory 

tests on rock cores. Factors such as degradation of the weak rock, reduction in 

shaft resistance due to shattering of the rock structure from driving adjacent piles, 

and formation of an enlarged hole around the pile hamper analytical methods.

The design of axial loaded piles in soft rock should be performed using the same 

procedures and methods used for piles in soil as presented in Section 7.2.2. Static 

and dynamic tests are recommended for design verification until local correlations 
272

and design methods are available. The empirical values provided above for H-piles 

in shale, sandstone, and limestone were developed by Illinois DOT for their practice 

and have been found to yield reasonable estimates of pile length and nominal 

resistance in the regional geology.

The nominal toe resistance of piles in soft and hard rocks can also be computed 

according to Equation 7-34 in Section 7.2.1.4.2 below. For this calculation, the rock 

friction angle, undrained shear resistance, and effective density are needed from 

laboratory tests.

It is recommended that unit resistance values calculated from this equation be 

compared to values obtained from static or dynamic tests and that these 

comparisons be compiled in a local or regional database. This will assist and 

improve foundation design procedures in soft and weak rocks. Piles driven into soft 

rock may not require pile toe protection (i.e. driving shoe).

7.2.1.4.2 Piles Driven to Hard Rock

For piles driven to hard rock, the nominal resistance is generally controlled by the 

structural limit state. In hard rock designs, the nominal structural resistance will 

generally be less than the nominal geotechnical resistance of hard rock.

Where laboratory tests can be made on undisturbed samples of weak or hard rock 

and the undrained shear strength, su, and the rock friction angle, φ, can be obtained, 

the nominal unit toe resistance, qp, in ksf can be calculated using Equation 7-34.

















++=

2

γ γγ

bN

NsPq cusp DN Ptq Eq. 7-34

Where:

su = undrained shear resistance of the rock (ksf). 

γ = effective density of the rock mass (kcf). 

D = pile penetration below the rock surface (ft).

b = pile width or diameter (ft).

Ps = pile toe shape factor of 1.25 for square pile or 1.2 for a circular pile.

 Pt = pile base factor of 0.8 for a square pile or 0.7 for a circular pile.

Nc, Nq, and Nγ are bearing capacity factors from Figure 7-22. 

Equation 7-34 represents wedge failure below a foundation modified with shape 

factors and it should not be confused with Terzaghi’s equation for a spread footing 
273

foundation. Tomlinson and Woodward (2015) added the recommended correction 

factors Ps and Pt for driven pile shapes.

Figure 7-22 Bearing capacity factors for foundations on rock 

(modified from Pells and Turner 1980).

Data from Kulhawy and Goodman (1980; 1987) showed that the unit toe resistance, 

qp, can also be estimated from the Rock Quality Designation (RQD) value of the 

intact rock mass and the unconfined compression strength of the rock, qu. RQD was 

described previously in Chapter 4. For RQD values of 0 to 70%, the nominal unit toe 

resistance can be estimated as 0.33qu. For RQD values of 70 to 100%, the nominal 

unit toe resistance can be linearly interpolated from 0.33qu at an RQD value of 70% 

to 0.80qu at an RQD value of 100%. In hard rock cases, pile toe protection should 

be considered in the design.

A static analysis resistance factor, φstat, for any of the above nominal resistance 

calculation methods for rock is not available. Therefore, the calculated nominal 

resistance should be verified during construction using a nominal resistance field 

verification method and its associated resistance factor, φdyn. 
274

7.2.1.5 Software for Single Pile Nominal Resistance Computations

The FHWA previously sponsored the development and use of the DRIVEN computer 

program for static analysis computations. This program used the FHWA 

recommended Nordlund Method and α-method as presented in this manual for 

calculation of the nominal resistance. The public domain version of the DRIVEN 

program is no longer available from the FHWA. However, a commercial version of 

the program, DrivenPiles, was released in 2015. Numerous other static analysis 

programs are also commercially available. Table 7-15 summarizes the static 

analysis methods incorporated in commercially available programs. It should be 

emphasized that the FHWA does not endorse any particular analysis program and 

the programs are listed alphabetically in Table 7-15. 
275

Table 7-15 Summary of Computer Analysis Software for Axial Single Pile Analysis

Computer 

Program

Static Analysis Methods

in Program

Method Presented 

in GEC-12 

(2016)

Method 

Presented in 

AASHTO 7th

Edition (2014)

AllPile Navfac DM-7 No No 

A-Pile API-RP2A Yes No 

A-Pile US Army COE No No 

A-Pile FHWA (alpha / Nordlund) Yes Yes

A-Pile Lambda Method No Yes

A-Pile NGI (CPT) No No 

A-Pile ICP (CPT) No No 

DrivenPiles Alpha Method Yes Yes

DrivenPiles Nordlund Method Yes Yes

FB-Deep FDOT SPT Method No No 

FB-Deep Schmertmann (CPT) Yes Yes

FB-Deep UF (CPT) No No 

FB-Deep LCPC (CPT) No No 

Unipile Alpha Method Yes Yes

Unipile Beta Method Yes Yes, but differs 

Unipile Elsami and Fellenius (CPT) Yes No 

Unipile Schmertmann (CPT) Yes Yes

Unipile LCPC (CPT) No No 

Unipile Meyerhof (SPT) No Yes

An example soil profile is presented in Figure 7-23 along with the necessary soil 

input parameters for the selected static analysis software programs. Figure 7-24

presents DRIVEN results of nominal resistance, shaft resistance, and toe resistance 

versus depth. For comparison purposes, results for the same soil profile are 

presented in Figures 7-25 and 7-26 for the APILE using the Nordlund and alpha 

Methods and UNIPILE using the Beta Method, respectively.
276

Figure 7-23 Soil profile for computer software comparison. 

700 to 1600 kips

Figure 7-24 Nominal resistance from DRIVEN program using FHWA method.
277

Figure 7-25 Nominal resistance from APILE program using FHWA Method. 

Figure 7-26 Nominal resistance from UNIPILE program using Beta Method. 
278

Of particular note is the difference between the DRIVEN results and the APILE 

FHWA method results. One could reasonably expect both programs to yield the 

same results. However, subtle differences in software coding results in about a 10% 

variation in the toe resistance and nominal resistance at the final 125 foot 

penetration depth. The programs also have differences in the toe resistance and 

nominal resistance at the transition depths between soil layers. The APILE program

uses an averaging of the toe resistance around layer transitions whereas the 

DRIVEN program does not. The results from the UNIPILE program which calculates 

the resistance using effective stresses in all layers yields a lower resistance solution.

Figure 7-27 presents nominal resistance results generated by the APILE - FHWA 

method for three different pile types in the same soil profile. The pile types include 

an HP 18 x 204 H-pile, a 18 inch O.D. closed end pipe pile, an a 24 inch square 

prestressed concrete pile. Note the displaced volume of a 24 inch concrete pile 

exceeds the 2.0 pcf volume limit in Figure 7-9. 

Figure 7-27 Nominal resistance versus depth for three pile types calculated by the 

APILE program with the FHWA Method. 
279

Both of these examples emphasize the importance of the program user to fully 

understand the methodology behind a given program, its accuracy, and its 

limitations in a given application.

In Figure 7-28, a design chart for an 18 inch diameter closed-end pipe pile is 

presented. The nominal resistance results versus depth were calculated using the 

APILE - FHWA method. In the same figure, the factored resistance versus depth is 

presented for different field verification methods. For a factored resistance of 1000 

kips in the soil conditions depicted, a relatively small difference exists between the 

estimated pile penetration depth of 106 feet based on a resistance factor of 0.80 

(static load test and dynamic tests on 2% of the production piles) and the estimated 

pile penetration depth of 108 feet based on a resistance factor of 0.65 (dynamic 

tests on 2% of the production piles). Hence, a static load test may not be justifiable 

based on cost in this situation. Conversely, a significant difference exists between 

the estimated pile penetration depth of 108 feet based on a resistance factor of 0.65 

(dynamic tests on 2% of the production piles) and the estimated pile penetration 

depth of 117 feet based on a resistance factor of 0.50 (wave equation analysis) or 

127 feet based on a resistance factor of 0.40 (Modified Gates formula). An 

evaluation of the nominal and factored resistances versus depth considering the cost 

of the pile length as well as the field verification method allows the designer to 

determine the most appropriate method of field verification for the design.

The design chart in Figure 7-28 allows an evaluation of the nominal and factored 

resistances versus depth considering the cost of the pile length as well as the field 

verification method. This type of analysis allows the designer to determine the most 

appropriate method of field verification for the design.
280

 

Figure 7-28 Design chart of nominal resistance, Rn, and factored resistance, Rr, 

versus depth for one pile type with different field control methods.

7.2.2 Resistance of Pile Groups in Axial Compression

The previous sections dealt with design procedures for single piles. However piles 

for almost all highway structures are installed in groups, due to the heavy foundation 

loads. The following sections of this chapter will address foundation design 

procedures for evaluating the nominal resistance in axial compression of pile groups. 

The nominal resistance in axial compression and settlement of pile groups are 

interrelated; settlement will be further discussed in Section 7.3.2 of this chapter. Pile 

group design computations are primarily performed using computer programs 

designed for this purpose such as FB-Pier or Group 6.0 rather than the simple hand 

calculations presented in this manual.
281

The efficiency of a pile group in supporting the foundation load is defined as the ratio 

of the nominal resistance of the group to the sum of the nominal resistance of the 

individual piles comprising the group. This may be expressed in equation form as:

 

Eq. 7-35

Where: 

ηg = pile group efficiency.

Rng = nominal resistance of pile group.

n = number of piles in pile group.

Rn = nominal resistance of each individual pile in pile group.

If piles are driven into relatively weak cohesive soil or in dense cohesionless material 

underlain by weaker soil, then the nominal resistance in axial compression of a pile 

group may be less than that of the sum of the nominal resistance in axial 

compression of the individual piles. In this case, the pile group has a group 

efficiency of less than 1. In cohesionless soils, the nominal resistance in axial 

compression of a pile group is generally greater than the sum of the nominal 

resistance in axial compression of the individual piles comprising the group. In this 

case, the pile group has group efficiency greater than 1.

Guidance on the group efficiency for piles driven to or into rock is not contained in 

AASHTO. Provided the pile group is driven to or into rock with equally strong 

material beneath, the pile group efficiency should be 1 or greater. Group efficiency 

of groups driven into highly weathered, karst, or other variable conditions should be 

further evaluated by the designer based on local conditions.

The soil medium supporting a pile group is also subject to overlapping stress zones 

from individual piles in the group. The overlapping effect of stress zones for a pile 

group supported by shaft resistance is illustrated in Figure 7-29. 

Piles are typically driven into mixed soil and rock profiles. For pile group analysis, 

the stratum that provides the majority of the nominal geotechnical resistance may be 

considered the controlling stratum.
282

7.2.2.1 Pile Groups in Cohesionless Soils

In cohesionless soils, the nominal group resistance of driven piles with a center to 

center spacing of less than 3 pile diameters is greater than the sum of the nominal 

resistance of the individual piles. The greater group resistance is due to the overlap 

of individual soil compaction zones around each pile which increases the shaft 

resistance due to soil densification. Piles in groups at center to center spacings 

greater than three times the average pile diameter generally act as individual piles.

Design recommendations for estimating group resistance for driven piles in 

cohesionless soil are as follows:

1. The nominal group resistance for driven piles in cohesionless soils not 

underlain by a weak deposit may be taken as the sum of the individual 

nominal pile resistances, provided jetting or predrilling was not used in the 

pile installation process. Jetting or predrilling can result in group efficiencies 

less than 1. Therefore, jetting or predrilling should be avoided whenever 

possible and controlled by detailed specifications when necessary.

2. If a pile group founded in a firm bearing stratum of limited thickness is 

underlain by a weak deposit, then the nominal group resistance is the smaller 

value of either the sum of the nominal resistances of the individual piles, or 

the group resistance against block failure of an equivalent pier, consisting of 

the pile group and enclosed soil mass punching through the firm stratum into 

the underlying weak soil. From a practical standpoint, block failure in 

cohesionless soils can only occur when the center to center pile spacing is 

less than 2 pile diameters, which is less than the minimum center to center 

spacing of 2.5 diameters allowed by AASHTO (2014). The method shown for 

cohesive soils in the Section 7.2.2.3 may be used to evaluate the possibility of 

a block failure.

3. Piles in groups should not be installed at center to center spacing less than 3 

times the average pile diameter unless dictated otherwise by pile size, site 

constraints, pile cap requirements, and construction costs. The minimum 

center to center spacing of 3 diameters is recommended to optimize nominal 

group resistance and to minimize pile installation problems associated with a 

center to center pile spacing of 2.5 diameters. 
283

Figure 7-29 Overlap of stress zones for friction pile group (after Bowles 1988).
284

7.2.2.2 Pile Groups in Cohesive Soils

The nominal group resistance in cohesive soil is usually governed by the sum of the 

nominal resistance of the individual piles, with some reduction due to overlapping 

zones of shear deformation in the surrounding soil. AASHTO (2014) design 

specifications states that the group resistance is influenced by whether the pile cap 

is in firm contact with the ground. If the pile cap is in firm contact with the ground, 

the soil between the piles and the pile group act as a unit.

The following design recommendations are for estimating nominal group resistance 

in cohesive soils. The lesser value of the nominal group resistance, calculated from 

Steps 1 to 4, should be used.

1. For pile groups driven in clays with undrained shear strengths of less 

than 2 ksf and the pile cap not in firm contact with the ground, a group 

efficiency of 0.65 should be used for center to center pile spacing of 

2.5 times the average pile diameter. If the center to center pile spacing 

is greater than 6 times the average pile diameter, then a group 

efficiency of 1.0 may be used. Linear interpolation should be used for 

intermediate center to center pile spacing.

2. For piles in clays with undrained shear strengths less than 2 ksf, and the 

pile cap in firm contact with the ground, a group efficiency of 1.0 may 

be used.

3. For pile groups in clays with undrained shear strength in excess of 2 ksf, 

a group efficiency of 1.0 may be used regardless of the pile cap - 

ground contact.

4. Calculate the nominal pile group resistance against block failure using 

the procedure described in Section 7.2.2.3.

5. Piles in cohesive soils should not be installed at center to center pile 

spacing less than 3 times the average pile diameter unless dictated 

otherwise by pile size, site constraints, pile cap requirements, and 

construction costs. The center to center pile spacing should also not 

be less than 3 feet.

It is important to note that the driving of pile groups in cohesive soils can generate 

large excess pore water pressures. This can result in short term (1 to 2 months after 
285

installation) group efficiencies on the order of 0.4 to 0.8. As these excess pore 

pressures dissipate, the pile group efficiency will increase. Figure 7-30 presents 

observations on the dissipation of excess pore water pressure versus time for pile 

groups driven in cohesive soils. Depending upon the group size, the excess 

pressures typically dissipate within 1 to 2 months after driving. However, in very 

large groups, full pore pressure dissipation may take up to a year.

If a pile group will experience the full group load shortly after construction, the 

foundation designer must evaluate the reduced group resistance that may be 

available for load support. In these cases, piezometers should be installed to 

monitor pore pressure dissipation with time. Effective stress resistance calculations 

can then be used to determine if the increase in pile group resistance versus time 

during construction meets the load support requirements.

Figure 7-30 Measured dissipation of excess pore water pressure in soil 

surrounding full scale pile groups (after O’Neill 1983).
286

7.2.2.3 Block Failure of Pile Groups

Block failure of pile groups is generally only a design consideration for pile groups in 

soft cohesive soils or in cohesionless soils underlain by a weak cohesive layer. For 

a pile group in cohesive soil as shown in Figure 7-31 the nominal resistance of the 

pile group against a block failure is provided by the following expression:

 Eq. 7-36

Where:

Rng = nominal resistance of the pile group.

D = pile embedded length.

B = width of pile group.

Z = length of pile group.

su1 = weighted average of the undrained shear strength over the depth of 

pile embedment for the cohesive soils along the pile group perimeter.

 su2 = average undrained shear strength of the cohesive soils at the base of 

the pile group to a depth of 2B below pile toe level.

Nc = dimensionless bearing capacity factor.

If a pile group will experience the full group load shortly after construction, the 

nominal group resistance against block failure should be calculated using the 

remolded or reduced shear strength rather than the average undrained shear 

strength for su1. 

The bearing capacity factor, Nc, for a rectangular pile group is generally 9. However, 

for pile groups with small pile embedment depths and/or large widths, Nc should be 

calculated from the following equation:

 Eq. 7-37

Where:

Nc = dimensionless bearing capacity factor. 

D = pile embedded length. 

B = width of pile group.

Z = length of pile group.
287

When evaluating possible block failure of pile groups in cohesionless soils underlain 

by a weak cohesive deposit, the weighted average unit shaft resistance for the 

cohesionless soils should be substituted for su1 in calculating the nominal group 

resistance. The pile group base strength determined from the second part of the 

nominal group resistance equation should be calculated using the strength of the 

underlying weaker layer.

Additional guidance can be found in AASHTO (2014) Article 10.6.3.1.2d on the 

punching failure of a footing situated in a strong layer overlying a weak layer. The 

equivalent footing concept for the pile group can be used in conjunction with the 

spread footing guidance.

Figure 7-31 Three dimensional pile group configuration (after Tomlinson 1994).
288

7.2.3 Design for Axial Tension Resistance

The design of piles for axial tension or uplift loading conditions has become 

increasingly important for structures subject to seismic loading. In some cases, the 

uplift resistance determines the minimum pile penetration requirements. In those 

situations, axial tension load tests may be very cost effective as the higher 

resistance factor as a result of testing may reduce minimum pile penetration 

requirements and overall pile lengths. Tension load test procedures are described 

by Kyfor et al. (1992) in FHWA-SA-91-042 as well as in Chapter 9.

7.2.3.1 Axial Tension Resistance of Single Piles

De Nicola and Randolph (1993) note that in fine grained cohesive soils, where 

loading is assumed to occur under undrained conditions, the shaft resistance is 

generally considered equal in compression and in uplift.

In non-cohesive or free draining soils, the uplift resistance of a pile has been more 

controversial. De Nicola and Randolph (1993) state that it has been customary to 

assume that the shaft resistance in uplift is approximately 70% of the shaft 

resistance in compression. Based upon a finite difference parametric study, they 

concluded that a reduction in shaft resistance for uplift in free draining soils should 

be used, and that piles have lower uplift resistance than their compression shaft 

resistance. Conversely, the American Petroleum Institute's (1993) recommended 

design practice considers the pile shaft resistance to be equal in uplift and 

compression loading. Likewise, Altaee et al. (1992) presented a case of an 

instrumented pile in sand where the shaft resistance was approximately equal in 

compression and uplift when residual stresses were considered.

Tomlinson (1994) notes that the shaft resistance under cyclic loading is influenced 

by the rate of application of load as well as the degree of degradation of soil particles 

at the soil-pile interface. Under cyclic or sustained uplift loading in clays, the uplift 

resistance can decrease from the peak value to a residual value. In sands, particle 

degradation or reorientation can also result in decrease in uplift resistance under 

cyclic or sustained uplift loading. Therefore, the designer should consider what 

effect, if any, sustained or cyclic uplift loading will have on soil strength degradation.

Based on the above issues, the factored uplift resistance of a single pile should be 

taken as that of the nominal shaft resistance calculated from any of the AASHTO 

static analysis methods with a resistance factor for uplift, φup, from Table 7-1. From 

AASHTO (2014) Article C10.7.3.10, the uplift resistance factor from static analysis 
289

methods is reduced to 80 percent of the resistance factor for static shaft resistance

in compression. Therefore, the shaft resistance does not need to be further reduced 

for the uplift case.

Equation 7-38 is modified from the general static analysis equation to account for 

uplift. Selection of the factored uplift resistance should also consider the potential 

for soil strength degradation due to the duration or frequency of uplift loading, which 

may not be apparent in the short term static or dynamic test results.

== φφ RRR supnr Eq. 7-38

Where:

Rr = factored resistance (Equation 7-3).

Rs = nominal shaft resistance (Equation 7-4).

φup = resistance factor for uplift analysis method per Table 7-1. 

If a tension static load test is performed for design confirmation, a design uplift 

resistance factor of up to 0.60 may be used. Dynamic tests with signal matching can 

also be used to determine uplift resistance per AASHTO (2014) Article 10.7.3.10. 

Based on a dynamic test, an uplift resistance factor of up to 0.50 is recommended.

Figure 7-32 illustrates a design chart for uplift loading. The nominal resistance and 

nominal shaft resistance in the figure were calculated using the FHWA method

(Nordlund / α-method) in APILE. The nominal shaft resistance has also been 

factored by 0.60 to yield the factored resistance in axial tension expected based on a 

static load test, 0.50 for a dynamic test with signal matching, and by 0.25 in the clay 

layers (α-method) and 0.35 (Nordlund Method) in the sand layers to yield the 

factored resistance in axial tension based solely on static analysis.

For a factored resistance in axial tension of 100 kips, the uplift design chart indicates 

estimated pile lengths of 72, 74, and 78 feet for a tension load test, dynamic test with 

signal matching, and static analysis. Hence, a static analysis may be the most cost 

effective approach for this factored load for a project with a limited number of piles. 

On the other hand, the uplift design chart indicates estimated pile lengths of 81, 84, 

and 112 feet for a factored uplift resistance of 200 kips based on a tension load test, 

dynamic test with signal matching, and static analysis. In this case, the pile length 

savings from either static load testing or dynamic testing may be economically 

justified.
290

Figure 7-32 Design chart for nominal and factored resistance in axial tension.

7.2.3.2 Axial Tension Resistance of Pile Groups

Axial tension or uplift resistance of a pile group is often a significant factor in 

determining the minimum pile penetration requirements and in some cases can 

control the foundation design. A few common conditions where group uplift 

resistance may significantly influence the foundation design include cofferdam seals 

that create large buoyancy forces, cantilever segmental bridge construction, and 

seismic, vessel impact, or debris loading. When piles with uplift loads are driven to a 

relatively shallow bearing stratum, uplift resistance may control the foundation 

design. AASHTO (2014) specifications for group uplift resistance are considered 

relatively conservative, particularly in cohesionless soils.
291

AASHTO specifications (2014) limit uplift resistance of a pile group to the lesser of:

1. The sum of individual pile resistance in axial tension.

2. The resistance in axial tension of the pile group considered as a block. 

Equation 7-39 is modified for factored uplift which considers the total resistance of a 

pile group. Meanwhile, the group uplift resistance is determined differently for 

cohesive and cohesionless soils, which is described in the following two sections.

 Eq. 7-39

Where: 

Rr = factored resistance in axial tension (Eq. 7-4).

Rug = nominal resistance in axial tension of the pile group.

φug = resistance factor for axial tension, 0.50 per Table 7-1. 

7.2.3.2.2 Axial Tension Resistance of Groups in Cohesionless Soils

In cohesionless soils, AASHTO (2014) Article 10.7.3.11 limits the nominal group 

resistance in axial tension to the lesser of two resistances. The first resistance is the 

sum of the individual pile nominal shaft resistance times the analysis method 

resistance factor for uplift, times the number of piles in the group. This is expressed 

in Equation 7-40. 

 Eq. 7-40

Where: 

Rug = nominal resistance in axial tension of pile group.

Rs = nominal shaft resistance.

φup = resistance factor for uplift analysis method per Table 7-1. 

n = number of piles in group. 

The second resistance is defined by the weight of soil contained with the pile group 

block depicted by Figure 7-33. This approach outlined by Tomlinson (1994), uses 

the effective weight of the block of soil extending upward from the pile toe level at a 

slope of 1H:4V. For simplicity in performing the calculation, the weights of the piles 

within the soil block are considered equal to the weight of the soil. The group uplift 

resistance is the lesser resistance determined from Equation 7-40 or Figure 7-33. 
292

Figure 7-33 Uplift of pile group in cohesionless soil (after Tomlinson 1994). 

7.2.3.2.3 Axial Tension Resistance of Groups in Cohesive Soils

Similarly in cohesive soils, AASHTO (2014) Article 10.7.3.11 limits the nominal 

group uplift resistance to the lesser of the resistance computed by Equation 7-40 or 

Equation 7-41. Figure 7-34 depicts the nominal group resistance provided by the 

soil weight contained within the pile group block and the soil shear resistance along 

the group perimeter, as should be used for Equation 7-41.

 Eq. 7-41

Where: 

Rn = nominal resistance (kips).

Rug = nominal resistance in axial tension of the pile group (kips).

D = pile embedded length (feet).

B = width of pile group (feet).

Z = length of pile group (feet).

su = weighted average of the undrained shear strength over the depth of 

pile embedment along the pile group perimeter (ksf).

Wg = effective weight of the pile/soil block including pile cap weight (kips).
293

Figure 7-34 Uplift of pile group in cohesive soils (after Tomlinson 1994). 

7.2.4 Nominal Axial Resistance Changes after Pile Driving

The surrounding soil is greatly disturbed when a pile is driven into it. As the soil 

surrounding the pile regains strength following the installation disturbance, a time 

dependent change in geotechnical resistance occurs. Frequently piles driven in 

saturated clays, and loose to medium dense silts or fine sands gain significant 

resistance after driving has been completed. This phenomenon is called soil setup. 

Occasionally piles driven into dense saturated fine sands, dense silts, or weak 

laminated rocks such as shale, will exhibit a decrease in resistance after the driving 

has been completed. This phenomenon is called relaxation. Case history 

discussions on soil setup and relaxation may be found in Fellenius et al. (1989), and 

Thompson and Thompson (1985), respectively.

7.2.4.1 Relaxation

The nominal resistance of driven piles can decrease with time following driving. This 

decrease in nominal resistance is known as relaxation. Relaxation has been 

observed in dense, saturated, fine grained soils such as non-cohesive silts and fine 

sands. In these cases, the driving process is believed to cause the dense soil near 

the pile toe to dilate (tendency for volume increase), thereby generating negative 
294

pore pressures (suction). The negative pore pressures temporarily increase the 

effective stresses acting on the pile, resulting in a temporarily higher soil strength 

and driving resistance. When these pore pressures dissipate, the effective stresses 

acting on the pile decrease, as does the nominal resistance. Several cases of 

relaxation in silts and fine sands have been reported in the literature including Yang 

(1970), York et al. (1994), Morgano and White (2004), and Richardson (2011).

Relaxation in weak laminated rocks has also been reported. Thompson and 

Thompson (1985) attributed this to a release of locked in horizontal stresses. Rock 

formations reported in the literature to exhibit relaxation include shales; Thompson 

and Thompson (1985), Hussein et al. (1993), Morgano and White (2004); and 

phyllite, Seidel et al. (1992). However, relaxation is possible in other weak 

laminated rock formations such as mudstone, claystone, and siltstone among others.

Because the nominal resistance may decrease (relaxation) after the end of driving, 

nominal resistance assessments from static load testing or dynamic test restrikes 

should be made after equilibrium conditions in the soil have been re-established. In 

the absence of site specific pore pressure data from piezometers, it is suggested 

that static load testing or restriking of piles in dense silts and fine sands be delayed 

for a few days to a week after driving, or longer if possible. In relaxation prone 

shales, it is suggested that static load testing or restrike testing be delayed a 

minimum of ten days to two weeks after driving.

Published cases of the relaxation magnitude of various soil types are quite limited. 

However, data from Thompson and Thompson (1985) as well as Hussein et al. 

(1993) suggest relaxation factors for piles founded in some shales can range from 

0.5 to 0.9. The relaxation factor is defined as the static load test failure load divided 

by the nominal resistance at the end of initial driving. Relaxation factors of 0.5 and 

0.8 have also been observed in two cases where piles were founded in dense sands 

and extremely dense silts, respectively. The importance of evaluating time 

dependent decreases in nominal resistance for piles founded in these materials 

cannot be over emphasized. AASHTO (2014) Article 10.7.3.4.2 states that a restrike 

test should be done after a sufficient time for relaxation to develop if relaxation is 

possible in the soils at the site.

7.2.4.2 Soil Setup

When saturated cohesive soils are compressed and disturbed due to pile driving, 

large excess pore pressures develop. These excess pore pressures are generated 

partly from the shearing and remolding of the soil and partly from radial compression 
295

as the pile displaces the soil. The excess pore pressures cause a reduction in the 

effective stresses acting on the pile, and a reduction in the soil shear strength. This 

results in a reduced nominal resistance during, and for a period of time after, driving.

After driving, the excess pore pressures will dissipate primarily through radial flow of 

the pore water away from the pile (Randolph et al. 1979). With the dissipation of 

pore pressures, the soil reconsolidates and increases in shear strength. This 

increase in soil shear strength results in an increase in the nominal resistance and is 

called soil setup. A similar decrease in resistance to pile penetration with 

subsequent soil setup may occur in loose to medium dense, saturated, fine grained 

sands or silts. The magnitude of the gain in nominal resistance depends on soil 

characteristics, pile material and pile dimensions.

Because the nominal resistance may increase after the end of driving, nominal 

resistance assessments should be made from static load tests, dynamic testing, or 

hammer restrike events performed after equilibrium conditions in the soil have been 

re-established. The time for the return of equilibrium conditions is highly variable 

and depends on soil type and degree of soil disturbance.

In research studies, piezometers installed within three diameters of the pile can be 

used to monitor pore pressure dissipation with time. Effective stress static nominal

resistance calculations can be used to evaluate the increase in resistance with time 

once pore pressures are quantified.

To quantify the largest magnitude of soil setup, static load testing or restrike dynamic 

testing of piles in fine grained soils should be delayed until pore pressures dissipate 

and return to equilibrium. In the absence of site specific pore pressure data or local 

experience, it is preferable that static load testing or dynamic restrike tests of piles in 

clays and other predominantly fine grained soils be delayed for two weeks after 

driving or longer if possible. Unfortunately, delays of this magnitude to quantify or 

confirm the soil setup magnitude are not frequently feasible on most routine projects. 

Therefore, it is best to quantify the soil setup magnitude and time rate of setup in a 

design stage test program in high setup locales. The design stage program should 

include multiple restrike events on dynamic test piles and static load test(s) so that 

both the time rate of setup as well as the setup magnitude can be determined and 

later used for production driving where only shortened restrike intervals can be 

accommodated due to the construction schedule. In sandy silts and fine sands, pore 

pressures generally dissipate more rapidly. In these more granular deposits, three 

to seven days is often a sufficient time delay.
296

Hannigan et al. (2012) presented a case history multiple dynamic monitoring events 

on test piles were used to quantify the time dependent soil setup magnitude and its 

spatial variation for a four-level interchange project. Static load tests were also used 

to confirm the nearly five-fold increase in shaft resistance over time.

Several methods have been proposed to predict soil setup effects. A study by Skov 

and Denver (1988) looked at predicting the nominal resistance in cohesive and 

cohesionless soils sometime after initial driving based on initial driving 

measurements using Equation 7-42.

















+=

o

non t

t ARR log1 1 Eq. 7-42

Where: 

Rn = nominal resistance after time “t” of driving (kips).

Rno = initial nominal resistance at “to” of driving (kips).

A1 = constant based on soil type and subsurface condition. 

(0.2 for sand and 0.6 for clay).

t = time after driving (days).

to = time after driving from which the increase in resistance is linear in 

logarithmic time (days) (typically 0.5 for sand, 1.0 for clay).

Komurka et al. (2003) summarized the state of the practice in estimating and 

measuring soil setup in a report to the Wisconsin Highway Research Program. This 

report summarizes the mechanisms associated with soil setup development and 

reviews several empirical relationships for estimating setup. 

A study by Bullock et al. (2005) assessed five square prestressed concrete piles 

driven in various soils throughout Florida. Resistance was measured through initial 

drive and restrike dynamic testing followed by subsequent O-cell tests. For this, 

they expanded upon the Skov and Denver approach.





 





+





























 = 1log

so o

s

s so t

t

R

m

RR Eq. 7-43

Where: 

Rs = shaft resistance after time “t” of driving (kips).

Rs = initial shaft resistance at “to” of driving (kips).

ms = semilog-linear slope of Rs vs t from multiple restrike tests. 

t = time after driving (days).

to = time of driving (days) (1 day for all tests).
297

Ng (2011) investigated soil setup on steel HP 10x42 piles driven into cohesive soils 

throughout Iowa. Initial drive and restrike dynamic testing was performed to assess 

time dependent resistance changes. This dissertation was presented in SI units, 

therefore the pore water dissipation factor, Ch, should be determined first using 

Equation 7-44, and then converted to English units for inclusion in Equation 7-45.

 (cm2

/min) Eq. 7-44

In which: 

 

Eq. 7-45

Where: 

Rs = shaft resistance after time “t” of driving (kips).

Rso = initial shaft resistance at “to” of driving (kips). 

fc = consolidation factor (non-dimensional regression factor).

Ch = pore water pressure dissipation factor (in2/min for Equation 7-45).

Na = average SPT N-Value over pile length. 

N = SPT N-Value. 

rp = equivalent pile radius (inches).

fr = remolding recovery rate (non-dimensional regression factor).

Lt = embedded pile length at time “t” after initial driving (feet).

Lo = embedded pile length at the time of initial driving (feet).

t = time after initial driving (days).

to = time of driving (days) (0.00694 days for all tests).

Another recent study comparing the nominal resistance from initial drive and restrike 

dynamic testing was performed by Reddy and Stuedlein (2014). They reviewed data 

from 76 piles in the Puget Sound Lowlands of Oregon, with restrikes varying from 5 

to 310 hours. Both closed (CEP) and open end (OEP) steel pipe piles, as well as 

prestressed concrete piles (PSC) were utilized. The setup prediction model 

developed followed a hyperbolic curve and is presented as Equation 7-46. 

 

Eq. 7-46
298

Where:

Rs = shaft resistance after time “t” of driving (kips).

Rso = initial shaft resistance at “to” of driving (kips).

A2 = constant based on pile type (1.72 for PSC piles, 0.70 for CEP, and 

0.77 for OEP). 

t = time after driving (hours).

to = time of driving (hours) (1 hour for all tests).

k1 = regression factor (0.17 for PSC, 0.12 for CEP and 0.15 for OEP).

k2 = regression factor (0.00044 for PSC piles, 0.00078 for CEP piles, 

0.00060 for OEP piles).

For the studies presented above, specific contributing factors were explored (i.e. 

regional soil, pile type, etc.). Therefore, these predictive methods may not translate 

to other areas with different conditions unless modified by local correlation 

experience. However, many of these studies provide a suggested approach for 

agencies to generate a database of tested piles with initial driving, restrike testing 

and static load tests are available to better assess the time dependent changes.

Limited information is available regarding soil setup in unsaturated soils. However, it 

is anticipated that setup would occur in these deposits. 

Rausche et al. (1996) calculated general soil setup factors based on the 

predominant soil type along the pile shaft. The soil setup factor was defined as the 

static load test failure load divided by the end-of-drive wave equation resistance. 

These results are presented in Table 7-16. The database for this study was 

comprised of 99 test piles from 46 sites. The number of sites and the percentage of 

the database in a given soil condition is included in the table. While these soil setup

factors may be useful for preliminary estimates, soil setup is better estimated based 

on site specific data gathered from pile restriking, dynamic measurements, static 

load testing, and local experience. 
299

Table 7-16 Soil Setup Factors (after Rausche et al. 1996)

Predominant Soil 

Type Along Pile 

Shaft

Range in

Soil Setup

Factor

Recommended

Soil Setup

Factors*

Number of Sites

and (Percentage

of Database) 

Clay 1.2-5.5 2.0 7 (15%)

Silt - Clay 1.0-2.0 1.0 10 (22%)

Silt 1.5-5.0 1.5 2 (4%)

Sand - Clay 1.0-6.0 1.5 13 (28%)

Sand - Silt 1.2-2.0 1.2 8 (18%)

Fine Sand 1.2-2.0 1.2 2 (4%)

Sand 0.8-2.0 1.0 3 (7%)

Sand - Gravel 1.2-2.0 1.0 1 (2%)

* Confirmation with Local Experience Recommended. 

7.2.4.2.1 Estimation of Pore Pressures During Driving

According to Lo and Stermac (1965), the maximum pore pressure induced from pile 

driving may be estimated from the following equation.

 Eq. 7-47

Where:

Δum = maximum excess pore pressure (ksf). 

Ko = at rest earth pressure coefficient.

σ’vi = initial vertical effective stress prior to pile driving (ksf).

 (Δu/σ’v)m = maximum value of the pore pressure ratio, Δu/σ'vo, 

measured in a CU triaxial test with pore pressure 

measurements.

Ismael and Klym (1979) presented a case history where the above procedure was 

used. They reported good agreement between measured excess pore pressures 

with estimates from the Lo and Stermac procedure.

Poulos and Davis (1980) summarized measurements of excess pore pressures due 

to pile driving from several case histories. In this compilation, the reported excess 

pore pressure measurements divided by the vertical effective stress were plotted 

versus the radial distance from the pile surface divided by the pile radius. These 
300

results are presented in Figure 7-35 and indicate that the excess pore pressure at 

the pile-soil interface can approach 1.4 to 1.9 times the vertical effective stress, 

depending upon the clay sensitivity.

The foundation designer should evaluate the potential change in nominal resistance 

with time. Once pore pressures are measured or estimated, effective stress static 

resistance calculation methods can be used to quantify the probable change in 

nominal resistance with time.

Figure 7-35 Excess pore water pressure due to pile driving 

(after Poulos and Davis 1980).

7.2.4.3 Implementation of Time Effects During Construction

As discussed above in the relaxation and soil setup sections, changes in nominal 

resistance occur after driving. These are permanent physical changes to the 

nominal resistance. Significant cost savings can be achieved when soil setup is 

effectively used. Many state transportation agencies are aware of soil setup effects, 

and as mentioned in Komurka et al. (2003), if setup effects could be predicted in the 

design phase, it may be possible to reduce pile lengths, reduce pile sections (use 

smaller-diameter or thinner-wall pipe piles, or smaller-section H-piles), or reduce the 

size of driving equipment (use smaller hammers and/or cranes).
301

Conversely, relaxation effects have been studied to a lesser degree, and in areas 

with known weathered shale or weak laminated bedrock, initial drive and restrike 

dynamic testing can prove useful to evaluate the extent of resistance loss (Long and 

Anderson 2014). Potentially problematic pile installations can be avoided when 

relaxation is identified early on and modified construction procedures adopted.

7.2.5 Nominal Lateral Resistance

In addition to axial compression and uplift loads, piles are routinely subjected to

lateral loads. Potential sources of lateral loads on bridge structures include vehicle 

acceleration and braking forces, wind loads, wave and current forces, debris loading, 

ice forces, vessel impact loads, construction procedures, thermal expansion and 

contraction, earth pressures on the back of abutment walls, slope movements, and 

seismic events. These lateral loads can be of the same magnitude as axial 

compression loads and therefore warrant careful consideration during design.

AASHTO (2014) Article 10.7.3.12 addresses the nominal resistance of piles 

subjected to lateral loads. For a long pile, the nominal lateral resistance is controlled 

by the strength of the pile material since the pile will fail structurally before soil failure 

occurs along the entire pile length. The nominal lateral resistance of a short pile, 

defined as a pile with insufficient embedment to prevent toe rotation, is controlled by 

the soil. While short and rigid piles with very large lateral loads are relatively 

uncommon in practice, when encountered, they should be further evaluated in a p-y 

analysis considering both the properties of the geomaterials and the pile’s structural 

resistance. The p-y method is described in detail in Section 7.3.7.3.

AASHTO (2014) notes that the strength limit state for nominal lateral resistance is, in 

most cases, only structural for the reasons described above. Chapter 8 provides 

guidance on structural limit states. The foundation deformation under lateral loading 

must also be within the established performance criterion for the structure. Section 

7.3.7 discusses horizontal pile deflection under the service limit state.

7.2.6 Pile Length Estimates for Contract Documents

Pile length estimates and contract pile quantities should be carefully established 

using a combination of the site subsurface information, static analysis methods, and 

preconstruction test pile programs. Sometimes, pile length estimates can be 

determined from an obvious bearing layer. In other cases, construction control or 

static analysis methods may be relied upon, but the potential bias in either method 
302

should be considered. Local pile driving experience should also be reviewed when 

determining the pile length estimates.

Minimum pile penetration requirements for effects of uplift, settlement, scour, lateral 

deflection, or downdrag should also factor into the contract pile length estimates. In 

cases where the depth of penetration required to achieve the nominal resistance in 

axial compression is less than the depth to satisfy these strength, service or extreme 

limit state requirements, the minimum pile penetration required to satisfy all limit 

state requirements should be used for contract pile quantities. The contract 

documents should also explicitly note the minimum pile penetration depth and the 

required nominal resistance below that depth during driving, Rndr, to avoid 

construction control problems if the required nominal resistance is achieved above 

the minimum pile penetration depth. It is important to note that a minimum pile 

penetration depth is not always required and that the estimated length may be all 

that is necessary for contract documents.

7.2.7 Groundwater Effects and Buoyancy

The nominal resistance in axial compression and tension as well as the lateral 

resistance can be influenced by groundwater elevation changes and the resulting

change in vertical effective stress. Groundwater elevation changes can be 

influenced by droughts, regional construction events such as construction of tunnels, 

or by local site events such as excavation dewatering. Events which cause 

temporary lowering of the groundwater elevation increases vertical effective stresses 

and thus the effective stress acting on the pile. Hence, a pile driven for the required 

nominal resistance in a temporarily lowered groundwater condition will have a 

nominal resistance less than required when the groundwater returns to its normal 

elevation.

Buoyant forces should also be considered in driven pile foundation design. A 

closed-end pipe pile driven through soft soils to rock or a hard bearing layer may 

need a thicker wall section or other measures to counter-act buoyancy effects. In 

this situation, buoyancy effects can raise a sealed closed-end pile from the bearing 

layer if the shaft resistance alone is insufficient to resist buoyancy forces.

7.2.8 Site Dewatering

When a site is dewatered during construction, a temporary increase in effective 

stresses will occur. This causes a corresponding temporary increase in soil shear 

strength that will result in piles driven in a dewatered site to develop a greater 
303

resistance at a shallower pile penetration depth as compared to the non-dewatered 

condition. The soil resistance to be overcome to reach a specified penetration depth 

will also be greater than in the non-dewatered condition. If not considered in the 

design stage, the selected pile type may not be drivable to the required penetration 

depth in the dewatered construction condition. When dewatering is terminated, the

effective stresses acting on the pile will decrease as the water table rises. This will 

result in a decrease in the soil shear strength and a decrease in long term nominal

resistance. Hence piles driven to the nominal resistance in the dewatered condition 

would have less than the required nominal resistance once dewatering is terminated.

For projects where significant dewatering is required, the effects of the dewatering 

on nominal resistance and pile drivability should be evaluated. In these cases, 

multiple static analyses should be performed to determine the nominal resistance 

and drivability requirements under the short term dewatered condition, as well as the 

long term nominal resistance after dewatering has been terminated.

Dewatering can also have negative impacts on nearby structures supported on deep 

and shallow foundations. The increase in effective stress can cause or increase 

drag force on deep foundations or cause consolidation settlements that affect the 

performance of deep and shallow foundations systems. The potential dewatering 

effects on adjacent structures should be evaluated during the design phase.

7.2.8.1 Artesian Conditions

Artesian conditions develop where a permeable soil layer or aquifer is confined by 

two impermeable layers and the permeable soil layer outcrops at an elevation higher 

than where it is encountered in-situ. In this situation the effective stresses acting on 

the pile are directly related to the head of water. Artesian conditions can result in 

lower than expected nominal resistance in the soil layer as well as potential vertical 

groundwater migration issues along the pile-soil interface. Harris et al. (2003) 

describe a driven pile case history where the back calculated effective stress design 

parameters from a static load test program where significantly lower than expected 

due to artesian conditions. A detailed subsurface exploration program can aid in 

identifying artesian conditions. Nominal resistances that require pile lengths that 

punch through confining layers should be avoided in design. Artesian conditions 

should be noted on foundation plans to reduce the risk of construction problems.
304

7.2.9 Scour

Scour is defined as the erosion of soil and rock materials from the streambed and/or 

stream banks due to flowing water. Foundations subject to scour require input from 

all disciplines; hydraulics, structures, geotechnical, and construction. Multidiscipline 

input and communications are required to achieve the most cost effective design.

Though often considered as being localized, scour may consist of multiple 

components including long term aggradation and degradation, local scour, and 

contraction scour. Aggradation and degradation involve the long term streambed 

elevation changes due to an abundance or deficit, respectively, in upstream 

sediment supply. Local scour involves the removal of material from the immediate 

vicinity of a substructure unit and can be either clear-water (free of disturbed 

upstream sediment), or live-bed scour, complicated by the transport of upstream 

sediment into the scour hole following the storm event. In contrast, contraction 

scour occurs due to a constriction of the flow and involves erosion across all or most 

of the channel width and relates directly with the stream stratigraphy at the scour 

location. Contraction scour may be cyclic and or relate to the passing of a flood.

Different materials, subject to any of the above mentioned types of scour, erode at 

different rates. In a flood event, loose granular soils can be eroded away in a few 

hours. Cohesive or cemented soils typically erode more gradually and over several 

cycles of flooding but can experience the same nominal scour depths as those of 

cohesionless deposits. As noted earlier in this chapter, the nominal resistance of a 

driven pile is due to soil resistance along the pile shaft and at the pile toe. 

Therefore, the erosion of the soil materials providing pile support can have 

significant detrimental effects on nominal resistance and must clearly be evaluated 

during the design stage.

Depending on the type of scour and the scour susceptibility of the streambed soils, 

multiple static resistance calculations may be required to evaluate the nominal 

resistance of a pile and to establish minimum pile penetration requirements. In the 

case of local scour, the soil in the scour zone provides resistance at the time of 

driving that cannot be counted on for long term support. Hence, shaft resistance in 

the scour zone, although included for drivability considerations, is ignored for design 

purposes. However, because the erosion is localized, nominal resistance 

calculations should assume that the vertical effective stress is unchanged. The 

effects of non-localized scour on long term pile resistance are more severe. In all of 

degradation, contraction scour, and general scour, a reduction in both the scour 

zone soil resistance and the vertical effective stress is applied to long term 
305

resistance calculations, due to the widespread removal of the streambed materials. 

This added reduction in effective stresses can have a significant effect on the 

calculated shaft and toe resistances. Figure 7-36 provides an illustration of localized 

and non-localized scour. Table 7-17 presents HEC-18 recommended minimum 

scour design flood frequencies and scour design check flood frequencies as a 

function of the hydraulic design flood frequencies.

Figure 7-36 Local and channel degradation scour.

Table 7-17 Hydraulic Design, Scour Design, and Scour Design Check Flood 

Frequencies per HEC-18 (after Arneson et al. 2012)

Hydraulic Design Flood 

Frequency, QD

Scour Design Flood 

Frequency, QS

Scour Design Check Flood 

Frequency, QC

Q10 Q25 Q50

Q25 Q50 Q100

Q50 Q100 Q200

Q100 Q200 Q500
306

The FHWA publication FHWA HIF-12-003, “Evaluating Scour at Bridges” by Arneson 

et al. (2012) more commonly known as HEC-18, recommends that the following pile 

design issues also be considered at bridge sites subject to scour.

1. For pile supported substructures subjected to scour, a reevaluation of the 

foundation design may require a change in the pile length, number, crosssectional dimension and type based on the loading and performance 

requirements and site-specific conditions.

2. Piling should be designed for additional lateral restraint and column action 

because of the increase in unsupported pile length after scour. The 

unsupported pile length is discussed in Chapter 8.

3. Local scour holes at piers and abutments may overlap one another in some 

instances. If local scour holes do overlap, the scour is indeterminate and may 

be deeper. The top width of a local scour hole on each side of the pier 

ranges from 1.0 to 2.8 times the depth of local scour. A top width value of 2.0 

times the depth of local scour on each side of a pier is recommended. 

4. Perform the bridge foundation analysis on the basis that all streambed 

material in the scour prism above the total scour line has been removed and 

is not available for nominal resistance or lateral support. In areas where the 

local scour is confined to the proximity of the footing, the lateral ground 

stresses on the pile length which remains embedded may not be significantly 

reduced from the pre-local scour conditions.

5. Placing the top of the footing or pile cap below the streambed a depth equal 

to the estimated long term degradation and contraction scour depth will 

minimize obstruction to flood flows and resulting local scour. Even lower 

footing elevations may be desirable for pile supported footings when the piles 

could be damaged by erosion and corrosion from exposure to river or tidal 

currents. However, in deep water situations, it may be more cost effective to 

situate the pile cap above the mudline and design the foundation accordingly.

6. Stub abutments positioned in the embankment should be founded on piling 

driven below the elevation of the thalweg including long term degradation and 

contraction scour in the bridge waterway to assure structural integrity in the 

event the thalweg shifts and the bed material around the piling scours to the 

thalweg elevation.
307

AASHTO (2014) Article 10.7.3.6 notes that scour effects must be considered in 

determining the minimum pile embedment depth and the required nominal driving 

resistance,Rndr. The pile penetration depth after the design scour event must satisfy 

the required nominal resistance for axial and lateral loading. Therefore, the piles 

need to be driven to the required nominal resistance plus the shaft resistance lost 

due to scour. The shaft resistance in the scour zone should be calculated using the 

same static analysis procedure used for design except the scour zone resistance 

should not be included in the long term nominal resistance. The bias of the static 

analysis method should also be considered. 

If restrike dynamic measurements with signal matching are performed during 

construction, the shaft resistance in the scour zone materials can be determined and 

that resistance subtracted from the nominal resistance. The material below the 

scour elevation must provide the required nominal resistance after the scour event.

7.2.10 Downdrag

Downdrag is ground settlement, or downward soil movement, relative to a pile. 

Current AASHTO (2014) guidance treats downdrag and the resulting drag force as 

an additional load that must be supported by the pile foundation. Therefore, 

downdrag is addressed in AASHTO under the geotechnical strength limit state. The 

AASHTO approach has generally been conservative in practice as there have been 

relatively few transportation structures where drag forces have been problematic and 

corrective action required. However, a design procedure that addresses the effects 

of downdrag in the geotechnical service limit state and in the pile structural strength 

limit state is preferred to the current AASHTO geotechnical strength limit state 

approach. The recommended design approach for downdrag is presented under the 

service limit state in Section 7.3.6.1.
308

7.3 SERVICE LIMIT STATES

Service limit states for foundations include settlement, lateral movement, overall 

stability and scour at the design flood. Foundation movement should be limited to 

structural tolerances for both the total and differential deformations. As mentioned in 

AASHTO (2014), the economics of limiting settlement, rotation and horizontal 

movements should be compared with the cost of designing the respective 

superstructure to tolerate such movement. This includes maintenance of either 

structural or roadway elements. Deformations should be computed assuming scour

will occur at the design flood. 

The service limit state provides limits on stress, deformation, and cracking under 

regular service conditions. Service limit state geotechnical considerations in driven 

pile foundation designs include:

• vertical deformation – settlement,

• horizontal movements,

• rotation,

• overall stability, and

• deformations due to scour at the design flood (100 year event).

Settlements should be determined using the Service I Load combination which 

includes normal operational factors only. If the pile foundation is supported in or on 

cohesive deposits subject to time dependent consolidation settlement, transient 

loads may be excluded from the settlement analysis. Settlement methods are 

covered in Section 7.3.5.

Load combinations for all applicable service limit states should be evaluated for 

horizontal movement and foundation rotation. Horizontal movement is measured at 

the foundation top, and should be within structural limits based on column length and 

stiffness. Sufficient geotechnical resistance and structural stiffness to resist shear 

and overturning loads should be checked during design. Lateral loading from the 

structure or earthen elements is covered in Section 7.3.7. Rotational movement is 

likewise measured at the foundation top and should be restricted.

Service limit state considerations are further discussed in AASHTO (2014) Articles 

10.5.2.1 through 10.5.2.4 and associated commentary. All applicable service limit 

state load combinations must be evaluated.
309

7.3.1 Tolerable Vertical Deformations and Angular Distortion 

Tolerable deformations are limited by the structure type and function, design service 

life, and anticipated performance at respective displacement levels. Vertical, 

horizontal, and rotational displacements should be considered during design, where 

tolerable movement criteria shall be established by empirical procedures, structural 

analysis, or both (AASHTO 2014).

Agencies often limit tolerable vertical deformations to restrictive values such as 1 

inch or less without a rational basis. While there are no technical reasons to set 

arbitrary deformation limits, there are practical reasons for limits such as the 

deformation tolerances of attached structures and utilities. Drainage, ride quality, 

and safety are additional considerations in evaluating the deformation magnitude 

that can be tolerated by the design.

Moulton (1985) performed a detailed study of 314 bridges to develop a tolerable 

movement criterion for highway bridges. He recommended limiting values of 

angular distortion for multi-span and simple span structures. Angular distortion is 

illustrated in Figure 7-37, and is defined by the differential settlement of the 

foundation, Sd, relative to the span length, Ls. Both the differential settlement and 

span length should be expressed in the same units. The tolerable movement criteria 

recommend by Moulton and illustrated in SHRP 2 Project R19B is presented in 

Table 7-18. Moulton also recommended that lateral deformations be limited to 1.5 

inches for all structures.

Figure 7-37 Angular distortion due to non-uniform bridge settlement. 
310

Table 7-18 Tolerable Movement Criteria for Bridges

Type of Bridge Limiting Angular Distortion, 

Sd/Ls (radians)

Multi-span bridges 0.004

Simple-span bridges 0.005

Note the limiting angular distortion criterion was developed for tolerable bridge 

movements in the longitudinal direction and should not be used for tolerable bridge 

movements in the transverse direction.

7.3.1.1 Load Factor for Vertical Deformations 

AASHTO (2014) specifications include a load factor for vertical deformations or 

settlement, γSE, which is to be considered on a project-specific basis. AASHTO 

specifications currently recommend a value of 1.0 for this load factor in lieu of 

project-specific information to the contrary. This load factor is used to assess force 

effects from settlement on the structural design such as the generation of secondary 

moments within a given span due to settlement of a substructure support. For 

example, a settlement load factor of 1.25 does not indicate the computed 

settlements from a given method should be multiplied by 25% to limit the probability 

of exceeding tolerable settlements. Rather, the settlement load factor is applied to 

the limit state load combinations to determine the settlement effects on service and 

strength limit states. A research effort is in progress to develop additional guidance 

on application and use of the γSE load factor.

7.3.2 S-0 Concept for Vertical Deformations

Due to the variability of subsurface conditions, the vertical deformation that occurs at 

each substructure location will generally vary. In addition, the predicted deformation 

and actual deformation will differ. Therefore, a bridge structure should be designed 

to accommodate a realistic value of differential settlement and angular distortion in 

the longitudinal direction using the S-0 concept proposed by Duncan and Tan 

(1991). The key points of this method are as follows:

1. The actual settlement of a substructure unit can be as large as the value 

calculated by a given method.
311

2. The actual settlement of an adjacent substructure unit can be zero instead of 

the value calculated using the same method.

The maximum differential settlement between two adjacent foundations using this 

approach is therefore the maximum total settlement calculated for either foundation 

unit supporting the span. The maximum angular distortion is then the maximum total 

settlement divided by the span length. These values represent the design 

differential settlement, DSd, and the design angular distortion, DA, respectively.

Figure 7-38 illustrates the computed settlement and associated computed angular 

distortion for a four span bridge supported by two abutments and three piers. The 

design differential settlement and design angular distortion using the S-0 concept 

are illustrated in Figure 7-39. 

Figure 7-38 Example settlement and angular distortion profile 

(after Modjeski and Masters 2015).
312

Figure 7-39 Settlement profile with design differential settlement and design 

angular distortion from S-0 concept (after Modjeski and Masters 2015).

7.3.3 Construction Point Concept

In the example presented in Figure 7-39, the computed settlements and resultant 

angular distortion are based on the foundation loads applied instantaneously at the 

same time. In reality, the loads on the foundation are gradually applied. Similarly, 

settlements from those incrementally applied loads occur as construction 

progresses. Therefore, settlement and angular distortion relative to several critical 

points during construction process should be evaluated separately by the designer.

Figure 7-40(a) illustrates the typical sequences in the construction of a bridge pier, 

while Figure 7-40(b) shows the associated applied load from each sequence and 

resulting vertical deformation. Plotting the load and deformation results as shown in 

this figure allow settlement and resulting distortion to be assessed relative to a given 
313

Figure 7-40 Construction point concept: (a) identification of key construction points, 

(b) estimated load-deformation behavior (after Modjeski and Masters 2015).

point in the construction process. In a typical bridge structure project, settlement 

that occurs after pier column and superstructure construction but prior to wearing 

surface construction is often the most relevant. For example, settlement occurring 

prior to superstructure construction may not be significant to the superstructure 

design.

Vertical deformation and angular distortion should be determined and evaluated 

using the construction point concept in limit state designs. The four span bridge 

example from Figure 7-38 is revisited to illustrate the construction point concept. In 

Figure 7-41, the calculated settlements (solid line), and hypothetical maximum 

angular distortion (dashed line) from the S-0 concept are once again plotted. Using 

construction point load and deformation data similar to Figure 7-40, the range in 

anticipated settlement and the resulting range in angular distortion are then 

computed. The results are then depicted by the hatched zone pattern in Figure 

7-41. 
314

Figure 7-41 Settlement profile with angular distortion from construction point 

concept (after Modjeski and Masters 2015).

7.3.4 Recommended Procedure for Vertical Deformation Analysis

Modjeski and Masters, Inc., et al. (2015) in SHRP2 Report S2-R19B-RW-1 

presented a recommended design procedure for determining vertical deformations 

or settlement. Their recommended procedure can be summarized as follows:

1. The total foundation settlement should be computed at each substructure 

location using an appropriate analysis method. Several appropriate analysis 

methods are described in detail in Section 7.3.5.

a. Calculate the total foundation settlement, Sta, using all applicable loads in 

the Service I load combination.

b. Calculate the total foundation settlement, Stp, before construction of the 

bridge superstructure using all applicable substructure loads in the 

Service I load combination.

c. Calculate, Str, the relevant total settlement: Str = Sta – Stp. 
315

2. 

a. Assume that the relevant settlement at each substructure support location 

can be as large as the magnitude calculated from an appropriate analysis 

method and that the settlement at the adjacent substructure support is zero. 

Hence, the differential settlement, Sd, for a bridge span is the maximum 

relevant settlement value of either support location.

b. As illustrated in Figure 7-37, calculate the angular distortion, A, in radians 

from the ratio of the differential settlement, Sd, divided by the span length, Ls, 

or: Ad = Sd / Ls. 

3. Compute the modified angular distortion, Am, by multiplying A from step 2b by 

the load factor for settlement, γSE. 

4. Check the modified angular distortion with the owner-specified criteria. 

Alternatively, angular distortion should be limited to 0.004 radians for multispan bridges and 0.005 radians for single span bridges if no owner-specified 

criteria are provided. Other angular distortion limits may be applicable or 

appropriate due to:

• vertical clearance,

• cost of mitigation via larger foundations, realignment, ground 

improvement, or surcharge,

• rideability,

• tolerable deformation limits of other associated structures,

• roadway drainage,

• aesthetics, and

• safety.

5. Evaluate the ramifications of the computed angular distortions on the 

structure. Modify the foundation design if necessary based on the structural 

ramifications.

6. The above procedure is also recommended for structures where foundations 

have been designed for equal total settlement due to the uncertainty in the 

settlement prediction from any given method.
316

7.3.5 Pile Group Settlement

Pile groups supported in and underlain by cohesionless soils will produce only 

immediate settlements. This means the settlements will occur immediately as the 

pile group is loaded. Pile groups supported in and underlain by cohesive soils may 

produce both immediate settlements and primary consolidation and secondary 

compression settlements that occur over a period of time. In highly overconsolidated clays, the majority of the foundation settlement will occur immediately. 

Primary consolidation settlements will generally be the major source of foundation 

settlement in normally consolidated clays.

The settlement of a pile group is likely to be many times greater than the settlement 

of an individual pile carrying the same load per pile as each pile in the pile group. 

Figure 7-42 illustrates that for a single pile, only a small zone of soil around and 

below the pile toe is subjected to vertical stress. This also illustrates that for a pile 

group, a considerable depth of soil around and below the pile group is stressed. The 

settlement of the pile group may be large, depending on the compressibility of the 

soils within the stressed zone.

Figure 7-42 Stress zone from single pile and pile group (after Tomlinson 1994). 
317

Methods for estimating settlement of pile groups are provided in the following 

sections. Methods for estimating single pile settlements are not provided because 

piles are usually installed in groups.

7.3.5.1 Elastic Compression of Piles

The pile group settlement methods discussed in the following sections only consider 

soil settlements and do not include the settlement caused by elastic compression of 

pile material due to the imposed axial load. Therefore, the elastic compression 

should also be computed and this settlement added to the group settlement 

estimates of soil settlement. The elastic compression, Δ, can be computed by the 

following expression:

AE

QL =∆ Eq. 7-48

Where: 

Q = unfactored axial load (kips). 

L = total pile length (inches).

A = pile cross sectional area (in2

). 

E = elastic modulus of pile material (ksi).

The modulus of elasticity for steel piles is 29,000 ksi. For concrete piles, the 

modulus of elasticity varies with concrete compression strength and is generally on 

the order of 4,000 ksi. The elastic compression of short piles is usually quite small 

and can often be neglected in design. For pile with significant shaft resistance, 

much of the load will be transferred to the soil over the length of the pile. In those 

cases, the elastic compression will be overestimated by Equation 7-48.

7.3.5.2 Group Settlement in Cohesionless Soils

Settlement in cohesionless soils is generally considered elastic for deformation 

calculations. Meyerhof (1976) provided elastic settlement correlations based on 

SPT and CPT test data as described in 7.3.5.2.1 and 7.3.5.2.2, respectively. Elastic 

settlement may also be estimated by the Hough (1959) method shown in Equation 

7-49. The stress increase in a soil layer due to the foundation loads should be 

determined using the equivalent footing method with consideration of the soil 

stratigraphy.





 



 ∆+ =

vo

vo

o C

HS '

' log 1

' σ

σσ Eq. 7-49
318

Where:

S = total layer settlement (feet).

Ho = initial soil layer thickness (feet).

C’ = dimensionless bearing capacity index from Figure 7-43, 

determined from the average corrected SPT N value for the layer

with consideration of the SPT hammer type.

σ’vo = vertical effective stress at midpoint of layer prior to 

stress increase (ksf).

Δσ = average change in vertical stress in the layer in (ksf).

Cheney and Chassie (2002) report that FHWA experience with this method indicates 

the method is usually conservative and can overestimate settlements by a factor of 

2. This conservatism is attributed to the use of the original bearing capacity index 

chart from Hough (1959) which was based upon SPT donut hammer data. Based 

upon average energy variations between SPT donut, safety, and automatic

hammers reported in technical literature, Figure 7-43 now includes a correlation 

between SPT N values from safety and automatic hammers and bearing capacity 

index. The safety hammer values are considered N60 values. This modification 

should improve the accuracy of settlement estimates with this method.

Figure 7-43 Values of the bearing capacity index, C', for granular soil 

(data from Hough 1959).
319

7.3.5.2.1 Method Based on SPT Test Data

Meyerhof (1976) recommended that the settlement of a pile group in a 

homogeneous sand deposit not underlain by a more compressible soil at a greater 

depth may be conservatively estimated by the following expression:

)60(1

4

N

BIp S ff = Eq. 7-50

For silty sand, use:

)60(1

8

N

BIp S ff = Eq. 7-51 

In which

5.0

8

' 1 ≥−=

B

D I f Eq. 7-52

Where: 

S = estimated total settlement (inches). 

pf = design foundation pressure (ksf).

B = width of pile group (feet).

N1(60) = average corrected SPT N value within a depth B below pile toe.

D’ = 2/3 pile embedded length (feet).

If = influence factor for group embedment. 

For piles in cohesionless soils underlain by cohesive deposits, the method presented 

in Sections 7.3.5.3 should be used.

7.3.5.2.2 Method Based on CPT Test Data

Meyerhof (1976) recommended the following relationship to estimate maximum 

settlements using cone penetration test results for saturated cohesionless soils.

ca

ff

q

p BI

S

2 = Eq. 7-53

In which

5.0

8

' 1 ≥−=

B

D I f Eq. 7-54

Where: 

S = estimated total settlement (inches).

pf = design foundation pressure (ksf).

B = width of pile group (feet).
320

If = influence factor for group embedment.

qca = average cone tip resistance within depth of B below the pile toe (ksf).

D’ = 2/3 pile embedded length (feet).

7.3.5.3 Group Settlement in Cohesive Soils

Terzaghi and Peck (1967) proposed that pile group settlements could be evaluated 

using an equivalent footing situated at a depth of 1/3 D above the pile toe. This 

concept is illustrated in Figure 7-44. For a pile group consisting of only vertical piles, 

the equivalent footing has a plan area (B)(Z) that corresponds to the perimeter 

dimensions of the pile group as depicted previously in Figure 7-31. The pile group 

load over this plan area is then the stress transferred to the soil through the 

equivalent footing. The load is assumed to spread within the frustum of a pyramid 

with side slopes at 30 ̊ and to cause uniform additional vertical stress at lower levels. 

Figure 7-44 Equivalent footing concept. 

The stress at any level is equal to the load carried by the group divided by the plan 

area of the base of the frustum at that level. AASHTO (2014) Article 10.7.2.3.1 

states that the load used in calculating group settlement is the permanently applied 

load. Equation 7-55 should be used to calculate the change in stress for a given 

depth, z, below the equivalent footing.

 Eq. 7-55

Where: 

Δσ’d = change in stress below equivalent footing (inches).
321

Q = unfactored permanent load (kips). 

B = width of pile group (feet).

Z = length of pile group (feet).

z = depth below equivalent footing. 

Rather than fixing the equivalent footing at a depth of a D above the pile toe for all 

soil conditions, the depth of the equivalent footing should be adjusted based upon 

soil stratigraphy and load transfer mechanism to the soil. Figure 7-45 presents the 

recommended equivalent footing location and stress distribution proposed by 

Duncan and Buchignani, (1976) for a pile group driven through a soft clay layer and 

into a medium or firm layer. 

Figure 7-45 Stress distribution below equivalent footing for a pile group in firm clay

(after Duncan and Buchignani, 1976).

Figure 7-46 presents other recommended locations of the equivalent footing for the 

following load transfer and soil resistance conditions:

a. toe resistance piles in hard clay or sand underlain by soft clay, 

b. piles supported by shaft resistance in clay, 

c. piles supported by shaft resistance in sand underlain by clay, and

d. piles supported by shaft and toe resistance in layered soil profile. 
322

Figure 7-46 Stress distribution below equivalent footing for a pile group 

(modified from Cheney and Chassie 2000). 
323

Consolidation settlements are calculated based on the stress increase in the 

underlying layers. Equations 7-56 through 7-58 provide primary consolidation 

settlement estimations for cohesive soil. Parameters determined from a 

consolidation test are shown in the plot of void ratio versus vertical effective stress in 

Figure 7-47. 

Figure 7-47 Plot of void ratio vs. vertical effective stress from consolidation test. 

For normally consolidated (σ’vo = σp) cohesive soil layers:

 Eq. 7-56

For overconsolidated soil with (σ’vo + Δσ ≤ σp):

 Eq. 7-57
324

For overconsolidated soil with (σ’vo + Δσ > σp):

 Eq. 7-58

Where: 

Sc = settlement estimate from primary consolidation (feet).

eo = initial soil layer void ratio.

Ho = initial soil layer height (feet).

σp = preconsolidation stress (psf).

σ’vo = vertical effective stress at midpoint of each layer prior to loading (psf).

Δσ = additional pressure from structural loading (psf).

Cr = recompression index from consolidation test.

Cc = compression index from consolidation test.

STEP BY STEP PROCEDURE FOR GROUP SETTLEMENT IN COHESIVE SOILS

STEP 1 Determine the new load imposed on soil by the pile group.

a. Determine the location of the equivalent footing. For pile groups 

supported primarily by toe resistance, the equivalent footing is placed at 

the pile toe as illustrated in Figure 7-46 (a). For pile groups supported 

primarily by shaft resistance, the equivalent footing is placed at a depth 

of 2⁄3 D as shown in Figure 7-46 (b, c, and d).

b. Determine the dimensions of the equivalent footing. For pile groups 

consisting only of vertical piles, the equivalent footing (unless modified 

for load transfer as in Figure 7-46) has the same dimensions as the 

length and width of the pile group from Figure 7-31. For pile groups 

supported primarily by shaft resistance that include batter piles, the plan 

area of the footing should be calculated from the dimensions of the pile 

group at depth 2⁄3 D, including the plan area increase due to the pile 

batter. For toe resistance groups with batter piles, the equivalent footing 

area should be the dimensions of the pile group at depth D, including the 

area increase due to pile batter.

c. Determine the stress distribution to soil layers below the equivalent 

footing up to the depth at which the stress increase from the equivalent 

footing is less than 10% of existing vertical effective stress at that depth. 

Remember that the equivalent footing size may be increased and the 
325

footing stress correspondingly reduced as a result of load transfer above 

the footing location or in groups with batter piles. The depth at which the 

stress increase is less than 10% will provide the total thickness of 

cohesive soil layer or layers to be used in performing settlement 

computations. AASHTO (2014) Article 10.7.2.3.1 states that the load 

used in calculating settlement is the permanently applied load.

d. Divide the cohesive soil layers in the affected stress increase zone into 

several thinner layers with thickness of 5 to 10 feet. The thickness of 

each layer is the thickness H for the settlement computation for that 

layer.

e. Determine the existing vertical effective stress, σ’vo, at the midpoint of 

each layer.

f. Determine the imposed stress increase, Δσ, at the midpoint of each 

affected soil layer based on the appropriate stress distribution.

STEP 2 Determine consolidation test parameters.

Plot the results of consolidation test(s) as shown in Figure 7-47 followed by 

determination of settlement parameters.

STEP 3 Compute settlements.

Using the appropriate settlement equation, Equation 7-56, 7-57 or 7-58, 

compute the settlement of each affected soil layer. Sum the settlements of all 

layers to obtain the total estimated soil settlement from the pile group. Add 

the elastic compression of the pile under the design load to obtain the total 

estimated pile group settlement.

7.3.5.4 Time Rate of Settlement in Cohesive Soils

Settlement analyses in cohesive soils should also evaluate the time required for the 

anticipated settlement to occur. In time rate computations, the time for 90% 

consolidation to occur is typically used to determine the total time required for 

primary settlement. The time rate of settlement of a cohesive soil deposit can be 

calculated from:
326

 

Eq. 7-59

Where:

t = time for settlement to occur (days).

 T = theoretical time factor for percentage of primary consolidation to 

occur (Table 7-19).

Hv = maximum vertical drainage path in the cohesive layer (feet). 

Cv = coefficient of consolidation (ft2/day).

The term Hv should not be confused with the term Ho used in the settlement 

equations for cohesive soils. Hv is the maximum distance water must travel from the 

compressible cohesive deposit to reach a more permeable layer. In the case of a 

cohesive layer overlain and underlain by a permeable granular layer, Hv would be 1⁄2 

the cohesive layer thickness. However, if the cohesive layer were overlain by a 

permeable granular layer and underlain by a non-permeable rock layer, Hv would be 

the full thickness of the cohesive deposit. Additional discussion on time rate of 

consolidation can be found in Samtani and Nowatzki (2006).

Table 7-19 Time Factors for Settlement

Primary Settlement (%) Time Factor (T)

10 0.008

20 0.031

30 0.071

40 0.126

50 0.197

60 0.287

70 0.403

80 0.567

90 0.848

7.3.5.5 Group Settlement in Layered Soils

Piles are often installed in a layered soil profile consisting of cohesionless and 

cohesive soils or in soil profiles where an underlying soil stratum of different 

consistency is affected by the pile group loading. In these cases, group settlement 

will be influenced by the stress increase in and compressibility of the affected layers. 
327

Figures 7-46(a), 7-46(c) and 7-46(d) may be used to determine the location of the 

equivalent footing and to evaluate the resulting stress increase in a soil layer. The 

settlement of each layer is then calculated using the appropriate settlement equation 

presented in Section 7.3.5.2 or Section 7.3.5.3. 

STEP BY STEP PROCEDURE FOR GROUP SETTLEMENT IN LAYERED SOIL 

PROFILES

STEP 1 Determine the new load imposed on soil by the pile group.

a. Determine the location of the equivalent footing. For pile groups 

supported primarily by toe resistance, the equivalent footing is placed 

at the pile toe as illustrated in Figure 7-46 (a). For pile groups 

supported primarily by shaft resistance in sands underlain by 

cohesive soils, the equivalent footing is placed at a depth of 8/9 D as 

shown in Figure 7-46 (c). For pile groups in layered soils supported 

by a combination of shaft and toe resistance, the equivalent footing is 

placed at 2⁄3 D as shown in Figure 7-46 (d).

b. Determine the dimensions of the equivalent footing. For pile groups 

consisting only of vertical piles, the equivalent footing (unless 

modified for load transfer as in Figures 7-46 (c) and 7-46 (d)) has the 

same dimensions as the length and width of the pile group from 

Figure 7-31. For pile groups supported primarily by shaft resistance 

that include batter piles, the plan area of the footing should be 

calculated from the dimensions of the pile group at the equivalent 

footing depth that includes the plan area increase due to the pile 

batter. For toe resistance groups with batter piles, the equivalent 

footing area should be calculated from the dimensions of the pile 

group at depth D, including the plan area increase due to the pile 

batter.

c. Determine the stress distribution to soil layers below the equivalent 

footing up to the depth at which the stress increase from the 

equivalent footing is less than 10% of existing vertical effective stress 

at that depth. Remember that the equivalent footing size may be 

increased and the footing stress correspondingly reduced as a result 

of load transfer above the footing location or in groups with batter 

piles. The depth at which the stress increase is less than 10% will 

provide the total thickness of soil to be evaluated in the settlement 
328

computations. AASHTO (2014) Article 10.7.2.3.1 states that the load 

used in calculating settlement is the permanently applied load.

d. Divide the soil layers in the affected stress increase zone into several 

thinner layers of 5 to 10 feet in thickness. The thickness of each 

layer is the thickness H for the settlement computation for that layer.

e. Determine the existing vertical effective stress, σ’vo, at the midpoint of 

each soil layer.

f. Determine the imposed stress increase, Δσ, at the midpoint of each 

affected soil layer based on the appropriate stress distribution.

STEP 2 Determine consolidation test parameters for each cohesive layer.

 Plot results of consolidation test(s) as shown in Figure 7-47. 

 Determine σp, eo, Cr and Cc values from the consolidation test data.

STEP 3 Determine bearing capacity index for each cohesionless layer.

 Determine the average corrected SPT N value, for each cohesionless 

 layer. Use N60 or the appropriate SPT hammer type in Figure 7-43 to 

 obtain the bearing capacity index for each layer. The safety hammer N 

 values in Figure 7-43 are considered representative N60 values.

STEP 4 Compute settlements.

 Using the appropriate settlement equation, compute the settlement of 

each affected soil layer. Sum the settlements of all layers to obtain the 

total estimated soil settlement from the pile group. Add the elastic 

compression of the pile under the design load to obtain the total 

estimated pile group settlement.

7.3.5.6 Group Settlement Using the Janbu Tangent Modulus Approach

The previous methods of group settlement analyses assume a linear relationship 

between induced stress and soil strain. However except at very small strains, a nonlinear relationship exists between stress and strain. Figure 7-48 illustrates that a 

stress increase at a small original stress will result in a larger strain than the same 

stress increase applied at a greater original stress.
329

Janbu (1963, 1965) proposed a tangent modulus approach that is referenced in the 

Canadian Foundation Engineering Manual (1985). In this method, the stress strain 

relationship of soils is expressed in terms of a dimensionless modulus number, m, 

and a stress exponent, j. Values of the modulus number can be determined from 

conventional laboratory triaxial or oedometer tests. The stress exponent, j, can 

generally be taken as 0.5 for cohesionless soils and 0 for cohesive soils.

Figure 7-48 Non-linear relation between stress and strain in soil 

(after Fellenius 1990).

The following seven equations are used to calculate the strain for normally and over 

consolidated, cohesionless and cohesive soils. The terms used in these four 

equations are as follows:

ε = strain from the increase in effective stress.

Es = elastic modulus of soil (ksf).

mn = dimensionless modulus number.

mnr = dimensionless recompression modulus number.

j = stress exponent in Table 7-20. 

σ’1 = new effective stress after stress increase (ksf). 

σ’o = effective stress prior to stress increase (ksf). 

σ’p = preconsolidation stress (ksf). 

σr = constant reference stress = 2 ksf. 
330

For cohesionless soils (j > 0), the strain induced by an increase in effective stress 

may be expressed as shown in Equation 7-60: 































 − 













 =

j

r

o

j

n r jm σ

σ

σ

σ

ε

1 ' ' 1 Eq. 7-60

For dense coarse grained soils (j = 1), the following equation should be used to 

calculate the strain induced by an increase in effective stress:

[ ] o Es

'' 1 1−= σσε Eq. 7-61

For sandy or silty soil (j = 0.5), the Equation 7-62 should be used to calculate the 

strain induced by an increase in effective stress:

[ ] p o mn

'' 2 = − σσε Eq. 7-62

For overconsolidated sandy or silty soil, where the final stress will exceed the 

preconsolidation stress, Equation 7-63 should be used to calculate the induced 

strain:

[ ] [ ] p

n

o mnr m '' 2 '' 2 = 1 σσε +− 1 − σσ Eq. 7-63

For cohesive soils, the stress exponent is zero, (j = 0). The strain induced by an 

increase in effective stress in a normally consolidated cohesive soil is then as 

follows:































 = mn o

'

' ln 1 1

σ

σ ε Eq. 7-64

For overconsolidated cohesive soils, if the applied foundation stress exceeds the 

preconsolidation stress, the following equation should be used to calculate the 

strain:

































+































 =

o n p

p

mnr m '

' ln 1

'

' ln 1 1

σ

σ

σ

σ

ε Eq. 7-65

For overconsolidated cohesive soils, if the applied foundation stress does not 

exceed the preconsolidation stress, the following equation should be used to 

calculate strain:
331

 

Eq. 7-66

In cohesionless soils, the modulus number can be calculated from the soil modulus 

of elasticity, Es (ksf), and the previously described terms using Equation 7-67: 

 Eq. 7-67

In cohesive soils, the modulus number, mn, or recompression modulus number, mnr, 

can be calculated from the initial void ratio, eo, and the compression index, Cc, or 

recompression index, Cr. The modulus number is calculated from:

 Eq. 7-68

The recompression modulus number, mmnnrr, is calculated by substituting the 

recompression index, Cr, for the compression index, Cc, as follows: 

 Eq. 7-69

The Janbu tangent modulus approach is quite adaptable to calculating pile group 

settlements in any soil profile. For reference purposes, typical and normally 

conservative modulus number and stress exponent values from the Canadian 

Foundation Engineering Manual (1985) are presented in Table 7-20. These values 

may be useful for preliminary settlement estimates. A step by step procedure for

this method follows.

STEP BY STEP PROCEDURE FOR PILE GROUP SETTLEMENT BY JANBU 

METHOD

STEP 1 Determine the new load imposed on soil by the pile group.

a. Determine the location of the equivalent footing. For pile groups 

supported primarily by toe resistance, the equivalent footing is placed at 

the pile toe as illustrated in Figure 7-46 (a). For pile groups supported 

primarily by shaft resistance in sands underlain by cohesive soils, the 

equivalent footing is placed at a depth of 8/9 D as shown in Figure 

7-46(c). For pile groups in layered soils supported by a combination of 

shaft and toe resistance, the equivalent footing is placed at 2⁄3 D as 

shown in Figure 7-46 (d).
332

b. Determine the dimensions of the equivalent footing. For pile groups 

consisting only of vertical piles, the equivalent footing (unless 

modified for load transfer as in Figures 7-46 (c) and 7-46 (d) has the 

same dimensions as the length and width of the pile group from 

Figure 7-31. For pile groups supported primarily by shaft resistance 

that include batter piles, the plan area of the footing should be 

calculated from the dimensions of the pile group at the equivalent 

footing depth that includes the plan area increase due to the pile 

batter. For toe resistance groups with batter piles, the equivalent 

footing area should be calculated from the dimensions of the pile 

group at depth D, including the plan area increase due to the pile 

batter.

c. Determine the stress distribution to soil layers below the equivalent 

footing up to the depth at which the stress increase from the 

equivalent footing is less than 10% of existing vertical effective stress 

at that depth. Remember that the equivalent footing size may be 

increased, and the footing stress correspondingly reduced, as a 

result of load transfer above the footing location, or in groups with 

batter piles. The depth at which the stress increase is less than 10% 

will provide the total thickness of the soil to be analyzed and the 

number of soil layers for settlement calculations. AASHTO (2014) 

Article 10.7.2.3.1 states that the load used in calculating settlement is 

the permanently applied load.

d. Divide the soil layers in the affected stress increase zone into several 

thinner layers of 5 to 10 feet in thickness. The thickness of each 

layer is the thickness H for the settlement computation for that layer.

e. Determine the existing effective stress, σ’vo, at the midpoint of each 

soil layer.

f. Determine the preconsolidation stress, σp, at the midpoint of each soil 

layer and whether the soil layer is overconsolidated or normally 

consolidated.

g. Determine the new effective stress, σ’1, at the midpoint of each 

affected soil layer based on the equivalent footing stress distribution.
333

Table 7-20 Typical Modulus and Stress Exponent Values 

(after Canadian Geotechnical Society 1985)

Soil Type Consistency 

Range in 

Modulus 

Number, mn

Stress 

Exponent, j

Glacial Till Very Dense to Dense 1000-300 1.0

Gravel --- 400-40 0.5

Sand Dense 400-250 0.5

Sand Medium Dense 250-150 0.5

Sand Loose 150-100 0.5

Silt Dense 200-80 0.5

Silt Medium Dense 80-60 0.5

Silt Loose 60-40 0.5

Silty Clay & Clayey Silt Hard-Stiff 60-20 0 

Silty Clay & Clayey Silt Stiff-Firm 20-10 0 

Silty Clay & Clayey Silt Soft 10-5 0 

Marine Clay Soft 20-5 0 

Organic Clay Soft 20-5 0 

Peat --- 5-1 0 

STEP 2 Determine modulus number and stress exponent for each soil layer.

 Use laboratory test data to compute modulus number for each layer.

 Preliminary settlement estimates can be made by using assumed

modulus numbers based on soil type as indicated in Table 7-20. 

STEP 3 Select the appropriate strain computation equation for each layer.

 Select the strain equation applicable to each layer depending upon

whether the soil layer is cohesive or cohesionless, and overconsolidated 

or normally consolidated.

STEP 4 Compute settlements.

Using the appropriate strain computation equation, compute the 

settlement, S, of each affected soil layer of thickness, Ho. Sum the 

settlements of all layers until the stress increase is less than 10% of the 
334

existing vertical effective stress at that depth obtain the total estimated 

soil settlement from the pile group. Add the elastic compression of the 

pile under the design load to obtain the total estimated pile group 

settlement.

7.3.5.7 Group Settlement Using the Neutral Plane Method

As the previous sections demonstrate, most of the group settlement methods select 

the depth of the equivalent footing based upon the assumed load transfer behavior. 

A preferred solution is to determine the depth of the neutral plane, and place the 

equivalent footing at or below the neutral plane location. The neutral plane occurs at 

the depth where the unfactored permanent load plus the load from negative shaft 

resistance is equal to the positive shaft resistance plus the toe resistance. The 

design should aim to locate the neutral plane in competent soils. When this is done, 

group settlements are usually well within acceptable limits.

The position of the neutral plane and the resulting negative shaft resistance can be 

determined from a static calculation. As previously stated, the neutral plane is the 

depth at which the sum of the unfactored permanent load plus the negative shaft 

resistance is equal to the positive shaft resistance plus the toe resistance. Above 

the neutral plane, the settlement of the soil is greater than the settlement of the pile. 

Any shaft resistance above the neutral plane is negative shaft resistance, since by 

definition the soil settlement is greater than the pile settlement. Therefore, the soil 

settlement transfers load to the pile. Below the neutral plane, the settlement of the 

soil is less than the settlement of the pile and load is transferred from the pile to the 

soil. Accordingly, pile settlement equals soil settlement at the neutral plane. 

Therefore, pile settlement is controlled by the soil compressibility below the neutral 

plane.

The following step by step procedure adapted from Goudreault and Fellenius (1994) 

is recommended for determination of the neutral plane.

STEP BY STEP PROCEDURE FOR DETERMINING THE NEUTRAL PLANE 

DEPTH

STEP 1 Perform a static resistance calculation.

a. Determine the nominal resistance, RRnn, from a static resistance 

calculation. 
335

b. Plot the load transfer versus depth by subtracting the shaft resistance 

at a given depth from the nominal resistance. This computation is 

identified as Curve A in Figure 7-49. 

STEP 2 Determine the load transfer to the pile above the neutral plane.

a. Determine the unfactored permanent load, Qd. 

b. Plot the load transfer to the pile versus depth by adding the shaft 

resistance at a given depth to the sustained load. This computation is 

labeled as Curve B in Figure 7-49. 

Figure 7-49 Neutral plane (after Goudrealt and Fellenius 1994).

STEP 3 Determine the depth of the neutral plane.

a. The neutral plane is at the depth of where Curves A and B intersect.

b. The location of the neutral plane will move if the permanent load is 

changed or the soil resistance versus depth is altered. Hence, 

design or construction decisions altering the permanent load, or soil 
336

resistance versus depth, will require reevaluation of the neutral plane 

location under the changed conditions. Preaugering, jetting, use of 

bitumen coatings, etc. are but a few of the factors that can change 

the soil resistance versus depth and thus the neutral plane location.

Goudreault and Fellenius (1994), note that the magnitude of group settlement 

between the neutral plane and the pile toe level is generally small. This is because 

the piles below the neutral plane act as reinforcing elements and the compression of 

the pile-reinforced soil is small. Therefore, for most cases they recommend 

calculating the pile group settlements based on locating the neutral plane at the pile 

toe.

The group load is distributed below the neutral plane at a slope of 1H:2V. As in the 

previous methods, the soil materials below the neutral plane should be evaluated for 

settlement until the stress increase is less than 10%. Group settlements are 

generally calculated based upon the stress increase and the resulting strain as 

presented for the Janbu method in Section 7.3.5.6. However, the methods 

presented for layered soils in Section 7.3.5.5 could also be used.

7.3.6 Settlement Due to Downdrag

When soil moves downward relative to the pile, it creates a drag force on, and 

therefore within, the pile. The downward soil movement creates the potential for 

downward pile movement. This downward pile movement is referred to as 

downdrag. The subsurface conditions, pile installation methods, pile loading 

sequences, as well as the pile and ground surface configuration determine the 

magnitude of the drag force and the downdrag movement. 

The design approach for downdrag specified by AASHTO treats drag force as an 

additional load to be resisted in a geotechnical strength limit state analysis. 

However, drag force does not affect geotechnical strength. As the pile head axial 

compression load approaches the nominal geotechnical resistance, all shaft 

resistance is positive or acting upward, hence no drag force exists. This section 

provides further explanation of these concepts and a recommended approach for 

calculating drag force and downdrag movement is presented in Section 7.3.6.1. The 

recommended approach is based on the neutral plane method developed by 

Fellenius (1989), and modified by Siegel et al. (2013). The recommended method 

should be used in evaluating the structural strength and geotechnical service limit 

states, respectively.
337

The drag force and/or the downdrag movement may be large or small, but for 

practical consideration they always exist primarily due to the contrasting stiffness 

between the pile and the surrounding geomaterials as well as due to soil disturbance 

and soil stress changes caused by pile installation. This is an important realization. 

It means that drag force and downdrag are not design considerations in only special

circumstances. Rather, drag force and downdrag should be evaluated in all driven 

pile designs. The determination of drag force and its effect on the structural 

resistance should be considered in the structural strength limit state analysis. 

Determination of the downdrag movement, since it contributes to pile head 

settlement, should be part of a geotechnical service limit state analysis.

Negative shaft resistance occurs as the soil moves downward relative to the pile. 

The accumulation of negative shaft resistance with depth produces the drag force on 

the pile. When piles are installed through a soil deposit undergoing consolidation, a 

large drag force can develop. Battered piles should be avoided in soil conditions 

where large soil settlements are expected because of the additional bending forces 

imposed on the piles which can cause pile deformation and damage.

There exists a depth along the pile where the sum of the permanent load on the pile 

plus the negative shaft resistance is equal to the positive shaft resistance on the pile 

plus the toe resistance. This depth is the location of the neutral plane. The 

maximum drag force and the maximum axial compression stress in the pile occur at 

the neutral plane.

Figure 7-50 presents plots of the calculated load in a pile versus depth as a function 

of time. This data set was collected by the Minnesota Department of Transportation 

(MnDOT) using sister bar mounted vibrating wire strain gages embedded in a 12.75 

inch O.D. concrete filled pipe pile. A large abutment footing and wall were 

constructed atop the pile concurrent with abutment backfilling. As the sustained pile 

head load increased, the development and location of the neutral plane 30 feet 

above the pile toe is readily apparent. The maximum compression force in the pile 

and the drag force at the neutral plane are also easily identified. The effects of any 

residual driving stresses are not included in the load versus depth profiles due to 

installation of the embedded instrumentation following pile driving.
338

Figure 7-50 Calculated load versus depth in 12.75 inch O.D. concrete filled pipe 

pile as a function of time (after MnDOT Geotechnical Manual).

Figure 7-51 illustrates the changes that occur as the pile approaches the 

geotechnical strength limit state. Note that the location of the neutral plane moves 

up the pile toward the ground surface. This results in a reduction in the magnitude 

of the negative shaft resistance as well as the drag force in the pile. At the 

geotechnical strength limit state, geotechnical failure, all shaft resistance is positive 

as the entire pile is moving downward relative to the soil during plunging failure. 

Hence, drag force does not alter the geotechnical strength limit state.
339

Figure 7-51 Change in neutral plane, negative shaft resistance, and drag force 

during transition to geotechnical strength limit (after MnDOT Geotechnical Manual).

The recommended design procedure for downdrag is presented in Section 7.3.6.1. 

The recommended approach addresses settlement considerations due to downdrag 

in the geotechnical service limit state and pile structural considerations in the 

structural limit state. Figure 7-52 presents a conceptual illustration of soil and pile 

movement and resulting pile forces.

Figure 7-52 Conceptual illustration of soil and pile movement (left) and resulting 

neutral plane and pile forces (right) (adapted from Siegel et al. 2013). 

Figure 7-53 illustrates the most common situation where large negative shaft 

resistance develops when fill is placed over a compressible layer immediately prior 

to, or after, piles are driven. Effective stress changes due to dewatering can also 

cause negative shaft resistance to develop such as shown in Figure 7-54. 
340

Figure 7-53 Common downdrag situation due to fill weight.

Figure 7-54 Common downdrag situation due to ground water lowering.

Briaud and Tucker (1993) presented the following criteria to assist in identifying 

situations where drag force and downdrag may be significant and should be carefully 
341

evaluated in the design. This list is not meant to be all inclusive and other situations

can also cause significant drag force and downdrag. Ultimately the designer must 

determine when the magnitude of the drag force and/or downdrag movement must 

be addressed in the design. Briaud and Tucker listed the following events for where 

significant drag force and downdrag merits careful consideration in the design: 

1. The total settlement of the ground surface will be larger than 4 inches.

2. The settlement of the ground surface after the piles are driven will be larger than 

0.4 inches.

3. The height of the embankment placed on the ground surface exceeds 5 feet.

4. The thickness of the soft compressible layer is larger than 30 feet.

5. The water table will be lowered by more than 10 feet.

6. The piles will be longer than 80 feet.

7.3.6.1 Recommended Approach for Downdrag

Siegel et al. (2013) proposed a downdrag design approach using the neutral plane 

method within the LRFD framework. This approach is the FHWA recommended 

design method for downdrag. It does not treat the drag force as an additional load 

that must be supported. Rather, drag force is a settlement consideration in the 

geotechnical service limit state and is a structural consideration in the pile structural 

limit state. The approach recognizes that drag force develops on all piles, 

regardless of soil and loading conditions.

The information necessary to implement the downdrag analysis procedure includes:

• Unfactored structural loads to determine the permanent load.

• Defined subsurface stratigraphy with appropriate parameters for all layers.

• Soil behavior models to characterize load-deformation response, 

(instrumented load tests, or t-z and q-z models).

• Information on fill placement including amount, lateral extent, and timeline.

In the recommended approach, all loads and resistances should be unfactored. The 

use of factored loads or factored resistances will distort load-transfer relationships 

and lead to erroneous predictions of settlement and drag force.
342

STEP BY STEP PROCEDURE FOR DOWNDRAG ANALYSIS 

STEP 1 Assume soil consolidation and ground settlement will occur. 

STEP 2 Using an appropriate static analysis method for the pile type and 

subsurface conditions, determine the nominal shaft, mobilized toe, and 

total mobilized resistance as a function of pile penetration depth.

 Shaft resistance is typically fully mobilized at relatively small pile 

movements of 0.10 inches or less. The full toe resistance however may 

require a toe movement as much as 4 to 5% of the pile diameter

depending on the geomaterial at the pile toe. An assessment of the 

mobilized toe resistance magnitude can be made using engineering 

judgment along with t-z and q-z behavior in static analysis software, or 

from t-z and q-z values derived from instrumented static load tests.

 Example output of the required static analysis results versus pile 

penetration depth is included in Figure 7-55. An illustration of the 

percentage toe resistance mobilized relative to the toe movement 

normalized by the pile diameter is presented in Figure 7-56. 

Figure 7-55 Plot of static analysis results (after Siegel et al. 2013).
343

Figure 7-56 Plot of normalized toe resistance versus toe movement 

(after Siegel et al. 2013).

STEP 3 Select the pile toe elevation for analysis.

STEP 4 Develop the axial load and resistance versus depth diagram at the 

selected pile toe elevation. 

Plot the accumulated shaft resistance versus depth (∑ Rs) from Step 2 

on the load and resistance diagram. This is depicted by the solid line in 

Figure 7-57. 

Determine the unfactored permanent load on the pile. Add the 

unfactored permanent load to the accumulated shaft resistance versus 

depth. This is indicated by the solid line in Figure 7-58. 

Subtract the sum of the accumulated shaft resistance at a given depth 

from the nominal resistance at the selected the pile toe elevation (Rn - ∑

Rs). Plot this resistance on the load and resistance diagram. This is 

denoted by dashed lines in Figure 7-59. Graphs a, b, and c, represent 

the total mobilized resistance based on the shaft resistance plus a 
344

mobilized toe resistance of 100%, 50%, and 0% of the nominal toe 

resistance, respectively.

Figure 7-57 Axial load and resistance plot of cumulative shaft resistance vs. depth 

(after Siegel et al. 2013).
345

Figure 7-58 Axial load and resistance plot of unfactored permanent load plus 

cumulative shaft resistance vs depth (after Siegel et al. 2013).

Figure 7-59 Axial load and resistance plot including mobilized resistances 

(after Siegel et al. 2013).
346

STEP 5 Determined the location of the neutral plane, the magnitude of the 

maximum axial compression load in the pile, and the magnitude of the 

drag force.

 For the example given in Figure 7-60, the neutral plane occurs at depth 

A, B, or C, depending on the magnitude of the mobilized toe resistance. 

If 100% of the toe resistance is mobilized, the neutral plane occurs at a 

depth of 30 feet (point A), the maximum axial load in the pile is 80 tons, 

and the drag force is 40 tons (80 tons – 40 ton permanent pile head 

load). Conversely, if no toe resistance is mobilized, the neutral plane 

occurs at a depth of 20 feet (point C), the maximum axial load in the pile 

is 60 tons, and the drag force is 20 tons (60 tons – 40 ton permanent pile 

head load).

Figure 7-60 Axial Load and resistance plot including neutral plane location based 

on mobilized toe resistance (after Siegel et al. 2013).

STEP 6 Check the structural strength limit state due to loading conditions

including drag force. The factored structural resistance of the pile in the 

strength limit state in axial compression, Pr, must exceed the factored 

permanent load and factored drag force per Equation 7-70. 

 Eq. 7-70
347

Where: 

 Qd = permanent load on pile (kips).

 DF = drag force on pile (kips). 

 γp = load factor for drag force in neutral plane analysis.

 Pr = factored axial compression resistance of pile (kips).

 An AASHTO load factor for drag force, γp, determined by static analysis 

methods using the neutral plane procedure is not yet available. 

Therefore, local calibration is required for implementation of this 

approach. The Minnesota DOT has adopted a load factor of 1.1 for drag 

force with the neutral plane downdrag procedure while a local calibration 

effort is in progress. This load factor was based on an equivalent 

minimum factor of safety 1.5 for material strength evaluation.

STEP 7 Calculate the settlement due to downdrag with respect to the neutral 

plane.

 a. Calculate the thickness of compressible soil, tsoil, beneath the neutral 

plane.

soil il −= DDt np Eq. 7-71

Where: 

tsoil = thickness of compressible soil beneath neutral plane (feet).

Dil = depth from reference to top of incompressible layer (feet).

Dnp = depth from reference to neutral plane (feet).

 b. Determine settlement due to downdrag from stress increase.















 ∆ =

s

p

dd soil E

tS σγ Eq. 7-72

Where: 

Sdd = settlement due to downdrag (feet).

tsoil = thickness of compressible soil beneath neutral plane (feet).

γp = load factor for downdrag.

∆σ = increase in vertical stress (ksf).

Es = elastic modulus of in-situ soil (ksf). 
348

It should be noted that effective stress changes such as approach fill 

placement after pile driving will alter the plot of the permanent load plus 

cumulative shaft resistance. This will in turn alter the location of the 

neutral plane, the maximum compression force in the pile, the magnitude 

of the drag force, and calculated settlement due to downdrag. Similarly, 

changes in the permanent load to be applied to the pile or changes in 

the pile toe elevation will also alter the analysis results.

7.3.6.2 Methods for Reducing Downdrag and Drag Force

In situations where the permanent load and drag force exceed the structural strength 

limit state, or where pile settlement exceeds the geotechnical service limit state, 

methods for mitigating the drag force should be evaluated. The following techniques

have been used in cases with large drag forces: 

a. Increase the structural resistance

In cases where the factored structural resistance is insufficient, the pile structural 

resistance may be increased by using a pile section with greater structural 

resistance. The use of a thicker wall section in a steel pipe pile, a heavier H-pile 

section of the same size, or use of higher strength pile materials are possible 

solutions. A larger pile size can also be evaluated although this will also increase 

the pile surface area and therefore drag force.

b. Increase the number of piles

An increase in the number of piles will result in a decrease in the permanent load 

carried per pile. This will reduce the maximum axial compression force carried by 

the pile section.

c. Reduce soil settlement by preloading

Preconsolidation of compressible soils can be achieved by preloading and 

consolidating the soils prior to pile installation. This approach is often used for 

bridge foundations in fill sections. Prefabricated vertical drains are often used in 

conjunction with preloading to shorten the time required for consolidation. Additional 

information on prefabricated vertical drains is available in “Prefabricated Vertical 

Drains,” FHWA RD 86/168 by Rixner et al. (1986) and in "Ground Improvement 

Methods" manual by Elias et al. (2004).
349

d. Use lightweight fill material

Construct structural fills using lightweight fill material to reduce the drag forces. 

Lightweight fill materials often used, depending upon regional availability, include 

geofoam, foamed concrete, wood chips, blast furnace slag, and expanded shales. 

Additional information on lightweight fills is available in Elias et al. (2004). Geofoam 

blocks being placed for embankment construction are shown in Figure 7-61. 

e. Use a friction reducer

Bitumen coating and plastic wrap are two methods commonly used to reduce the 

friction at the pile-soil interface. Bitumen coatings should only be applied to the 

portion of the pile which will be embedded in the negative shaft resistance zone. 

Case histories on bitumen coatings have reported reductions in negative shaft 

resistance from as little as 47% to as much as 90%. Goudreault and Fellenius

(1994) suggest that the reduction effect of bitumen may be analyzed by using an 

upper limit of 200 psf as the pile-soil shear resistance or adhesion in the bitumen 

coated zone.

Figure 7-61 Geofoam block approach embankment (courtesy MnDOT).
350

One of the major problems with bitumen coatings is protecting the coating during pile 

installation, especially when driving through coarse soils. An inexpensive solution to 

this problem is to weld an oversized collar around the pile where the bitumen ends. 

The collar opens an adequate size hole to permit passage of the bitumen for 

moderate pile lengths in fine grained soils. Figure 7-62 presents a photograph on an 

over-sized collar between the uncoated lower pile section and white washed bitumen 

coating on the upper pile section.

Bitumen coatings can also present additional construction problems associated with 

field coating and handling. The bitumen coating used must have relatively low 

viscosity to permit slippage during soil consolidation, yet high enough viscosity and 

adherence to insure the coating will stick to the pile surface during storage and 

driving. The bitumen must also have sufficient ductility to prevent cracking and

spalling of the bitumen during handling and driving. Therefore, the climate at the 

time of pile installation should be considered in selection of the proper bitumen 

coating. The use of bitumen coatings can be quite successful provided proper 

construction control methods are followed. However, bitumen coatings should not 

be casually specified as the solution to drag forces. 

Figure 7-62 Over-sized collar for bitumen coating protection. 
351

Plastic wrap has proven to be an economically attractive friction reducer, particularly 

for abutment piles driven behind and before construction of MSE walls. Tawfig 

(1994) performed laboratory tests on 0.006 inch thick polyethylene sheets used as a 

friction reducer. The laboratory test results indicated plastic wraps reduced the pilesoil shear resistance from between 78% for a one wrap layer to 98% for a two layer 

wrap with mineral oil lubricant. The laboratory test data indicated the pile-soil shear 

resistance of a one wrap layer was about 200 psf and only 20 psf for the lubricated 

two wrap system.

f. Prevent direct contact between soil and pile

Pile sleeves are sometimes used to eliminate direct contact between pile and 

soil. Bentonite slurry has been used in the past to achieve the same purpose. 

However these methods are generally more expensive and less effective than 

other solutions.

7.3.7 Horizontal Pile Foundation Deflection

Historically, designers often used prescriptive values for the lateral load resistance of 

vertical piles, or have added batter piles to increase a pile group's lateral resistance 

when it was believed that vertical piles could not provide the needed lateral 

resistance. However, vertical piles can be designed to withstand significant lateral 

loads. Modern analysis methods should be employed in the selection of the pile 

type and pile section.

Coduto (1994) notes that a foundation system consisting of only vertical piles 

designed to resist both axial and lateral loads is more flexible, and thus more 

effective at resisting dynamic loads, as well as less expensive to build. Bollman 

(1993) reported that the Florida Department of Transportation often uses only 

vertical piles to resist lateral loads, including ship impact loads because vertical piles 

are often less expensive than batter piles. In areas where seismic lateral loading is 

a serious concern, batter piles can deliver excessively large horizontal forces to the 

structure during the earthquake event. This phenomenon was observed during the 

Loma Prieta earthquake of 1989 in California and discussed in greater detail by 

Hadjian et al. (1992). In earthquake areas, lateral loads should be resisted by 

ductile vertical piles, and batter piles should be avoided whenever possible.

Sophisticated analysis methods are now readily available that allow the lateral loaddeflection behavior of piles to be rationally evaluated. Lateral loads and moments 

on a vertical pile are resisted by the flexural stiffness of the pile and mobilization of 

resistance in the surrounding soil as the pile deflects. The flexural stiffness of a pile 
352

is defined by the pile's modulus of elasticity, E, and moment of inertia, I. The 

geomaterial resistance to an applied lateral load is a combination of geomaterial 

compression and shear resistance, as shown in Figure 7-63. 

The design of laterally loaded piles must evaluate both the pile structural response 

and geomaterial deformation to lateral loads. The nominal structural resistance 

must be determined. In addition, the pile deformation under the service loading 

conditions must be calculated and compared to foundation performance criteria.

Figure 7-63 Soil resistance to a lateral pile load (after Smith 1989).

The design of laterally loaded piles requires the combined skills of the geotechnical 

and structural engineer. It is inappropriate for the geotechnical engineer to analyze 

a laterally loaded pile without a full understanding of pile-structure interaction. 

Likewise it is inappropriate for the structural engineer to complete a laterally loaded 

pile design without a full understanding of how pile section or spacing changes may 

alter the geotechnical response. Because of the interaction of pile structural and 

geotechnical considerations, the economical solution of lateral pile loading problems 

requires interdisciplinary coordination between the structural and geotechnical 

engineer.
353

Soil, pile, and load parameters have significant effects on the lateral resistance of 

piles. The factors influencing these parameters are as follows:

1. Geomaterial Parameters

a. Soil or rock type and physical properties such as shear strength, friction, 

density, groundwater level, and moisture content.

b. Coefficient of horizontal subgrade reaction in (pcf). This coefficient is defined 

as the ratio between a horizontal stress per unit area of vertical surface (psf) 

and the corresponding horizontal displacement (inches). For a given 

deformation, the greater the coefficient, the greater the lateral load resistance.

2. Pile Parameters

a. Physical properties such as shape, material, and dimensions.

b. Pile head conditions (rotational constraint, if any).

c. Method of pile placement such as driving, jetting, etc.

d. Group action.

3. Lateral Load Parameters

a. Static (monotonic or cyclic) or dynamic.

b. Eccentricity (moment coupled with shear force).

7.3.7.1 Pile Head Fixity

The pile to pile cap connection can behave as a pinned connection, fixed 

connection, or somewhere in between depending on the design detail. Wilson et al. 

(2006) summarized the design effect of pile to pile cap connection. If the pile 

extends only a nominal distance into the pile cap, it will behave as a pinned 

connection. A pinned connection provides restraint against translational movements 

but does not restrain rotation of the pile head relative to the cap. Conversely, a fixed 

connection requires the pile to be embedded two to three diameters into pile cap or 
354

to be fitted with a specially designed connection. A fixed connection provides 

restraint against rotation and all directions of movement of the pile head relative to 

the pile cap.

At the service limit state, a pinned connection will typically have more horizontal 

movement than a fixed connection as illustrated in Figure 7-64. At the strength limit 

state, Figure 7-65 illustrates that a fixed head condition will generally result in a 

larger bending moment at the pile head compared to a pinned connection, but may 

or may not have larger bending moments below the pile head.

Figure 7-64 Effect of pile head fixity on translation at service limit state 

(after Wilson et al. 2006).

Figure 7-65 Effect of pile head fixity on moment in piles at strength limit state 

(after Wilson et al. 2006).
355

7.3.7.2 Lateral Design Methods

The basic design approaches for lateral soil resistance analysis of vertical piles 

consist of lateral load tests or analytical methods. Both of these approaches are 

described in greater detail in the following sections.

1. Lateral Load Tests

Full scale lateral load tests can be conducted at a site during either the design or 

construction stage. The load-deformation data obtained is used to finalize or confirm 

the design for the particular site. Factors such as loading rate, cyclic (single or multidirectional) versus monotonic application of design forces, and magnitude of axial 

load should be considered in developing appropriate field testing procedures. 

Lateral load tests may not be economically justifiable on many projects but are 

essential on projects controlled by lateral load demand. Chapter 9 provides 

additional details on lateral load test procedures and interpretation.

2. Analytical Methods

The analytical methods are based on theory and empirical data and permit the 

rational consideration of various site parameters. Two common approaches are 

Broms' (1964a, 1964b) hand calculation method, and Reese's (1984) computer 

solution. Both approaches consider the pile to be analogous to a beam on an elastic 

foundation. FHWA IP-84-11 by Reese (1984) presents details of both methods.

Broms' method provides a relatively easy hand calculation procedure to determine 

lateral loads and pile deflections at the ground surface. Broms' method also ignores 

the axial load on the pile. As lateral load demand has increased along with the need 

for improved deformation estimates more detailed load-deformation computer

analyses have become the norm.

Reese's p-y method is a more rigorous computer analysis that was originally 

available in the 1993 DOS based COM624 computer program which was developed 

under an FHWA research grant. That method is now incorporated in that program’s 

proprietary successor, the LPILE program (Isenhower and Wang 2014). The p-y 

method permits the inclusion of more complete modeling parameters of a specific 

problem. The program output provides distributions versus depth of moment, shear, 

soil and pile moduli, and soil resistance for the entire length of pile, including 

moments and shears in above ground sections.
356

For the design of all major pile foundation projects, the p-y method should be used. 

Some of the more common software programs that perform lateral loading analysis 

using the p-y method include LPILE, FBPIER, and ALLPILE. It should be 

emphasized that the FHWA does not endorse the use of any particular software 

program. However the p-y method, described further in the following section, is the 

FHWA recommended analysis method for lateral load design.

7.3.7.3 p-y Method

The interaction of a pile-soil system subjected to lateral load has long been 

recognized as a complex function of nonlinear response characteristics of both pile 

and soil. The most widely used nonlinear analysis method is the p-y method, where 

p is the soil resistance per unit pile length and y is the lateral soil or pile deflection. 

This method, illustrated in Figure 7-66, models the soil resistance to lateral load as a 

series of nonlinear springs. As noted previously, some of the more common 

software programs that perform lateral loading analysis using the p-y method include 

LPILE, FBPIER, and ALLPILE.

Reese (1984, 1986) has presented procedures for describing the soil response 

surrounding a laterally loaded pile for various soil conditions by using a family of p-y 

curves. The procedures for constructing these curves are based on experiments 

using full sized, instrumented piles and theories for the behavior of soil and rock 

under stress.

The geomaterial modulus subgrade reaction is defined as follows:

y

p k −= Eq. 7-73

Where: 

k = modulus of subgrade reaction (psi).

p = soil resistance per unit pile length (lbs/inch).

y = lateral deflection (inch). 
357

Figure 7-66 Typical lateral analysis pile-soil model.

The negative sign indicates that the ground resistance opposes pile deflection. The 

ground’s modulus is the secant modulus of the p-y curve and is not constant except 

over a small range of deflections. Typical p-y curves are shown in Figure 7-67. 

Ductile p-y curves, such as curve A, are typical of the response of soft clays under 

static loading and sands. Brittle p-y curves, such as curve B, can be found in some 

stiff clays under dynamic loading conditions. 
358

Figure 7-67 Typical p-y curves for ductile and brittle soil (after Coduto 1994).

The factor most influencing the shape of the p-y curve is the soil properties. 

However, the p-y curves also depend upon depth, soil stress strain relationships, 

pile width, water table location, and loading conditions (static or cyclic). 

Representative values for the clay ε50 strain parameter are provided in Table 7-21

while modulus of subgrade reaction values for clays and sand are shown in Table 

7-22. Procedures for constructing p-y curves for various soil and water table 

conditions as well as static or cyclic loading conditions are typically provided in the 

p-y program documentation.

Table 7-21 Representative Values of ε50 for Clays

Clay Consistency Average Undrained Shear Strength, su, 

(ksf) ε50

Soft Clay 0.25-0.50 0.02

Medium Clay 0.50-1.0 0.01

Stiff Clay 1.0-2.0 0.007

Very Stiff Clay 2.0-4.0 0.005

Hard Clay 4.0-8.0 0.004
359

Table 7-22 Representative Modulus of Subgrade Reaction Values, ks and kc, for 

Clays and Sands

Soil

Type

Avg. 

Undrained 

Shear 

Strength, su, 

(ksf) 

Soil 

Condition 

Relative to 

Water Table

ks-Static

Loading

(pci)

kc-Cyclic

Loading

(pci)

Soft Clay 0.25-0.50 --- 30 ---

Medium Clay 0.50-1.0 --- 100 ---

Stiff Clay 1.0-2.0 --- 500 200

Very Stiff Clay 2.0-4.0 --- 1000 400

Hard Clay 4.0-8.0 --- 2000 1000

Loose Sand --- Submerged 20 20

Loose Sand --- Above 25 25

Med Dense Sand --- Submerged 60 60

Med Dense Sand --- Above 90 90

Dense Sand --- Submerged 125 125

Dense Sand --- Above 225 225

The p-y programs solve the nonlinear differential equations representing the 

behavior of the pile-soil system to lateral (shear and moment) loading conditions in a 

finite difference formulation using Reese's p-y method of analysis. The strongly 

nonlinear reaction of the surrounding soil to pile-soil deflection is represented by the 

p-y curve prescribed to act on each discrete element of the embedded pile. For 

each set of applied boundary (static) loads the program performs an iterative 

solution which satisfies static equilibrium and achieves an acceptable compatibility 

between force and deflection (p and y) in every element.

The shape and discrete parameters defining each individual p-y curve may be input 

by the user or generated by the program. Layered soil systems are characterized by 

conventional geotechnical data including soil type, shear strength, density, depth, 

and stiffness parameters, and whether the loading conditions are monotonic or cyclic 

in nature.

For batter piles, Awoshika and Reese (1971) proposed a modifying constant ranging 

from 0 to 2 be applied to the pult value based on the direction of the applied lateral 

load relative to the orientation of the pile batter. This modifying constant 

proportionally modifies the p-values. If the pile head is inclined away from the 
360

direction of the horizontal load, the modifying constant ranges from 1 to 2. If the pile 

head is inclined toward the applied horizontal load the modifying constant ranges 

from 1 to 0. With this approach, predicted behavior has been reported to reasonable 

agree with lateral test results on outwardly battered piles but less successful for 

inwardly battered piles. Hence, full scale load tests to better evaluate batter pile 

response should be considered on important projects.

The influence of applied loads (axial, lateral and moment) at each element can be 

modeled with flexural rigidity varying as a function of applied moment. In this 

manner, progressive flexural damage such as cracking in a reinforced concrete pile 

can be treated more rigorously. Programs typically include a subroutine which 

calculates the value of flexural rigidity at each element under the boundary 

conditions and resultant pile-soil interaction conditions.

Typical p-y analysis output summarizes the input information and the analysis 

results. The input data summarized includes the pile geometry and properties, and 

soil strength data. Output includes the generated p-y curves at various depths 

below the pile head and the computed pile deflections, bending moments, stresses 

and soil moduli as functions of depth below the pile head. This information allows an 

analysis of the pile’s structural resistance. Internally generated (or input) values of 

flexural rigidity for cracked or damaged pile sections are also output. Graphical 

output presentations versus depth include the computed deflection, slope, moment, 

and shear in the pile, and soil reaction forces similar to those illustrated in Figure 

7-68. 

The p-y analyses characterize the behavior of a single pile under lateral loading 

conditions. A detailed view is obtained of the load transfer and structural response 

mechanisms to design conditions. Considerable care is required in extrapolating the 

results to the behavior of pile groups (pile-soil-pile interaction, etc.), and accounting 

for the effects of different construction processes such as predrilling or jetting.

In any lateral analysis case, the analyst should verify that the intent of the modeling 

assumptions, all elastic behavior for example, is borne out in the analysis results. 

When a lateral load test is performed, the measured load-deflection results versus 

depth should be plotted and compared with the analysis predicted behavior so that 

an evaluation of the validity of the p-y curves used for design can be made. Figure 

7-69 illustrates a comparison between the measured load–deflection curve and one 

predicted by COM624P.
361

Figure 7-68 Graphical presentation of p-y analysis results (after Reese 1986). 

Figure 7-69 Comparison of measured and COM624P predicted load-deflection 

behavior versus depth (after Kyfor et al. 1992).
362

7.3.7.4 Strain Wedge Method

The p-y methods are generally applicable to piles that have the ability to bend and 

deflect. For short, stiff, piles, the strain wedge (SW) method may be a more realistic 

modeling approach as a short stiff pile tends to rotate rather than bend. Background 

details on the strain wedge method are provided in Norris (1986) and Ashour et al. 

(1998a).

In the strain wedge method, a passive soil wedge is modelled to resist lateral pile 

loads. A three dimensional wedge is incorporated into the beam on elastic 

foundation problem and can accommodate variations in shape, depth, loading and 

pile deflection due to changes in strain. Additional research with this model has 

added the use of multiple soil layers and the effect of pile head conditions (Ashour et 

al. 2002). Within the wedge, the mobilized strains form the relationship between 

passive resistance and horizontal displacement as shown in Figure 7-70. 

Figure 7-70 Strain wedge developed in soil (after Ashour et al. 1998a).
363

The soil stress-strain behavior can be estimated or directly incorporated from 

laboratory tests and included in the wedge profile. For the so called multi-sublayer 

technique, individual soil parameters are assigned for each respective soil layer in 

the wedge. The wedge shape changes with variations in soil properties as depicted 

in Figure 7-71. This analysis method therefore offers a potential benefit over 

traditional p-y modeling; the elastic springs are not independently modeled along the 

continuous pile length as shown in Figure 7-72. In addition, pile group effects are 

addressed from the overlapping wedges of each pile in the group.

Implementation of the SW method may be performed with software such as the 

Deep Foundation System Analysis Program (DFSAP), and has shown positive 

correlation with measured pile response from field tests (Ashour et al. 1998a). 

However widespread use of DFSAP has yet to occur, and many transportation 

agencies continue to use p-y based lateral pile analyses.

Table 

Figure 7-71 Proposed geometry of compound passive wedge 

(after Ashour et al. 1998a).
364

Figure 7-72 Soil-pile interaction with multiple soil layers (after Ashour et al. 1998a).

7.3.7.5 Single Piles

Lateral analysis software readily performs lateral analysis for a single pile. The 

applied loads to be analyzed should be factored loads (at the respective strength, 

service, and extreme event limit states) while the geotechnical resistance has a 

resistance factor of 1.0 per AASHTO (2014). Pile bents frequently consist of a 

single row of piles. Therefore, these are modeled as a single pile along the 

alignment of a bridge, and as a pile group in the perpendicular axis. More 

frequently, piles are installed in groups with two or more rows of piles. The following 

section addresses lateral pile analysis in greater detail with content applicable to 

both a single pile (lead row) and pile groups.

7.3.7.6 Pile Groups

The ability of a pile group to resist lateral loads from vessel impact, debris, wind, or 

wave loading, seismic events, and other sources is a significant design issue. The 

deflection of a pile group under a lateral load is typically 2 to 3 times larger than the 

deflection of a single pile loaded to the same intensity. Holloway et al. (1981) and 

Brown et al. (1988) reported that piles in trailing rows of pile groups (nearest the 

point of load application) have significantly less resistance to a lateral load than piles 

in the lead row, and therefore exhibit greater deflections. This is due to the pile-soilpile interaction that takes place in a pile group. The pile-soil-pile interaction results 

in the lateral resistance of a pile group being less than the sum of the lateral 

resistance of the individual piles comprising the group. Hence, laterally loaded pile 

groups have a group efficiency of less than 1.
365

The lateral resistance of an individual pile in a pile group is a function of its position 

in the group and the center to center pile spacing. Brown et al. (1988) proposed a 

p-multiplier, Pm, be used to modify the p-y curve of an individual pile based upon the 

piles row position. An illustration of the p-multiplier concept is presented in Figure 

7-73. For piles in a given row, the same Pm value is applied to all p-y curves along 

the length of the pile. In a lateral load test of a 3 by 3 pile group in very dense sand 

with a center to center pile spacing of 3b, Brown found the leading row of piles had a 

Pm of 0.8 times that of an individual pile. The Pm values for the middle and back row 

of the group were 0.4 and 0.3, respectively.

McVay et al. (1995) performed centrifuge model tests on a 3 by 3 pile group having 

center to center pile spacings of 3b and 5b. A dense and loose sand condition was

simulated in the centrifuge model tests. For the dense sand case at a center to 

center spacing of 3b, the centrifuge model test results were similar to Brown's field 

results. However, McVay also found that the Pm values were influenced by soil 

density and the center to center spacing. The Pm results from McVay's centrifuge 

tests as well as other recent results for vertical piles in 3 x 3 pile groups are 

summarized in Table 7-23. McVay's centrifuge tests indicated lateral load group 

efficiencies in sands on the order of 0.74 for a center to center pile of 3b and 0.93 for 

a center to center spacing of 5b. Field studies in cohesive soils have also shown 

that pile-soil-pile interaction occurs. Brown et al. (1987) reported Pm values of 0.7, 

0.5, and 0.4 for the lead, second, and third row of a laterally loaded pile group in stiff 

clays. 

Additional work on this topic has included full scale lateral load testing of a 16 pile 

group in loose sand by Ruesta and Townsend (1997), and a 9 pile group in clayey 

silt by Rollins et al. (1998). A scaled model study of a cyclically laterally loaded pile 

group in medium clay has also been reported by Moss (1997). The center to center 

pile spacing, Pm results, and pile head deflections reported in these studies are 

included in Table 7-23. NCHRP Project 24-09 entitled “Static and Dynamic Lateral 

Loading of Pile Groups” was also completed by Brown et al. (2001). The objective 

of this study was to develop and validate an improved design method for pile groups 

subjected to static and dynamic lateral loads. The information summarized in Table 

7-23 has been averaged and incorporated into AASHTO (2014). However, the 

distinction between soil type and test type is not made. The designer should 

consider if other p-multiplier, Pm, values may be more applicable in situations similar 

to those described in Table 7-22 which have comparatively lower Pm values than 

AASHTO.
366

Figure 7-73 Illustration of p-multiplier concept for lateral group analysis.
367

Table 7-23 Laterally Loaded Pile Group Studies

Brown and Bollman (1993) proposed a p-multiplier procedure for the design of 

laterally loaded pile groups. It is recommended that this approach, outlined in the 

step by step procedure that follows, be used for the design of laterally loaded pile 

groups. For a center to center pile spacing of 3b, AASHTO (2014) design 

specifications recommend p-multiplier, Pm, values of 0.8 for the lead row, 0.4 for the 

second row, and 0.3 for the third and subsequent rows. For a center to center pile 

spacing of 5b, the AASHTO specified Pm values are 1.0 for the lead row, 0.85 for the 

second row, and 0.7 for the third and subsequent rows. For center to center pile

spacing between 3B and 5B, interpolation is recommended to determine the 

Soil 

Type

Test

Type

Center to 

Center 

Pile 

Spacing

Calculated 

p-Multipliers, Pm

For Rows

1, 2, & 3+

Deflection, 

inches Reference

Stiff Clay Field Study 3b 0.70, 0.50, 0.40 2 

Brown et al. 

(1987)

Stiff Clay Field Study 3b 0.70, 0.60, 0.50 1.2 Brown et al. 

(1987)

Stiff Clay Field Study 3.3b 0.82, 0.61, 0.45 3.5

Rollins et al.

(2006)

Stiff Clay Field Study 4.4b 0.90, 0.80, 0.69 1.6

Rollins et al. 

(2006)

Stiff Clay Field Study 5.65b 0.95, 0.88, 0.77 2.6

Rollins et al. 

(2006)

Medium

Clay 

Scale ModelCyclic Load 3b 0.60, 0.45, 0.402.4 at

50 cycles 

Moss 

(1997)

Clayey Silt Field Study 3b 0.60, 0.40, 0.40 1.0-2.4

Rollins et al. 

(1998)

V. Dense

Sand Field Study 3b 0.80, 0.40, 0.30 1 

Brown et al. 

(1988)

M. Dense 

Sand

Centrifuge 

Model 3b 0.80, 0.40, 0.30 3 McVay et al. (1995)

M. Dense 

Sand

Centrifuge 

Model 5b 1.0, 0.85, 0.70 3 McVay et al. (1995)

Loose

M. Sand 

Centrifuge 

Model 3b 0.65, 0.45, 0.35 3 McVay et al. (1995)

Loose

M. Sand 

Centrifuge 

Model 5b 1.0, 0.85, 0.70 3 McVay et al. (1995)

Loose

F. Sand Field Study 3b 0.80, 0.70, 0.30 1-3 

Ruesta et al. 

(1997)
368

appropriate Pm value. Figure 7-74 shows typical load and moment versus deflection 

plots from this procedure which can be performed using multiple individual analyses 

with a p-y software program. The analyses can also be performed with less effort 

using pile group software such as the FB-Pier, FB-MultiPier or GROUP computer 

programs.

The computer program FB-Pier was developed with FHWA support as the primary 

design tool for analysis of pile groups under axial and lateral loads. This program, 

which is a successor of the LPGSTAN program by Hoit and McVay (1994) is a nonlinear, finite element analysis, soil structure interaction program. FB-Pier uses a pmultiplier approach in evaluation of laterally loaded pile groups under axial, lateral, 

and combined axial and lateral loads. The program is capable of analyzing driven 

pile and drilled shaft foundation supported sound walls, retaining walls, signs and 

high mast lighting. FB-MultiPier replaced FB-Pier and functions similarly. FBMultipier contains additional features and can apply loading variations from multiple 

piers connected by bridge spans.

Fayyazi et al. (2012) studies the effects of the center to center pile spacing between 

rows as well as the spacing between piles within the same row in two model pile 

groups in sand. They observed that the spacing within a row can have a significant 

effect of the lateral pile group resistance due to the edge effects from overlapping 

zones of influence between two piles in the same row. They found that edge effects 

can result in overestimating the lateral load resistance by as much as 30%. Their 

research is ongoing with the goal of providing a procedure of p-multiplier selection 

that considers both row and inter row pile spacing within a group.

STEP BY STEP DESIGN PROCEDURE FOR LATERALLY LOADED PILE 

GROUPS USING LATERAL SINGLE PILE ANALYSIS

STEP 1 Obtain factored lateral loads.

STEP 2 Develop p-y curves for single pile.

a. Obtain site specific single pile p-y curves from instrumented lateral 

pile load test at site.

b. Use p-y curves based on published correlations with soil properties. 

c. Develop site specific p-y curves based on in-situ test data.
369

Figure 7-74 Typical plots of (a) Load versus deflection and (b) Bending moment 

versus deflection for pile group analysis (adapted from Brown and Bollman 1993).
370

STEP 3 Run software analyses.

a. Perform p-y analyses using the Pm value for each row position to 

develop load-deflection and load-moment data.

b. Use AASHTO (2014) design specification recommendations for Pm

values. For a center to center pile spacing of 3b, recommended Pm

values are 0.8 for the lead row, 0.4 for the second row, and 0.3 for 

the third and subsequent rows. For a center to center pile spacing 

of 5b, recommended Pm values are 1.0 for the lead row, 0.85 for the 

second row, and 0.7 for the third and subsequent rows. Interpolate 

for center to center pile spacing between 3b and 5b to determine for 

the appropriate Pm value.

c. Determine shear load versus deflection behavior for piles in each 

row. Plot load versus pile head deflection results similar to as 

shown in Figure 7-74(a).

STEP 4 Estimate group deflection under lateral load.

Average the load for a given deflection from all piles in the group (i.e. each of the 

four rows) to determine the average group response to a lateral load as shown in 

Figure 7-74(a).

Divide the lateral load to be resisted by the pile group by the number of piles in the 

group to determine the average lateral load resisted per pile.

Enter load-deflection graph similar to Figure 7-74(a) with the average load per pile to 

estimate group deflection using the group average load deflection curve.

STEP 5 Evaluate pile structural acceptability.

a. Plot the maximum bending moment determined from p-y software 

analyses versus deflection for each row of piles as illustrated in Figure 

7-74(b).

b. Check the pile structural adequacy for each p. Use the estimated 

group deflection under the lateral load per pile to determine the 

maximum bending moment for an individual pile in each row.
371

c. Determine maximum pile stress from p-y software output associated 

with the maximum bending moment.

d. Compare maximum pile stress with pile yield stress.

STEP 6 Perform refined pile group evaluation that considers superstructuresubstructure interaction.

7.3.7.6.1 Lateral Resistance Increases Through Ground Improvement

Methods to improve the lateral resistance of pile groups in weak near surface soils 

were evaluated by Rollins and Brown (2011) in NCHRP Report 698, Design 

Guidelines for Increasing the Lateral Resistance of Highway-Bridge Pile Foundations 

by Improving Weak Soils. This study concluded that significant increases in lateral 

resistance in soft clays and loose sands can be achieved through soil replacement 

or ground improvement techniques. Figure 7-75 illustrates the appropriate treatment 

areas to improve lateral resistance around new foundations in weak soils as well as 

the treatment areas around existing foundations to improve their lateral resistance.

Tables 7-24 and 7-25 summarize the treatment options studied, the improvement in 

lateral resistance achieved, the treatment costs, and the cost savings. The standard 

alternative to ground improvement is to add more piles in order to resist the same 

lateral load. This option would also require a larger pile cap. The approximate cost 

of the additional piles and larger cap necessary to resist the same lateral load was 

estimated for each ground improvement technique and is reported in the third 

column of Table 7-25. The difference between this cost and the ground 

improvement cost in column 2 is the reported cost savings in column 4. For more 

detailed information, please refer to the NCHRP study.
372

Figure 7-75 Ground improvement treatment areas for increased lateral resistance 

of pile groups in weak soils (Rollins and Brown 2011).
373

Table 7-24 Summary of Ground Improvement Method and Increase in Lateral 

Resistance (after Rollins and Brown 2011)

Treatment Method

Treatment

Dimensions

(LxWxD)

Treatment 

Volume

(yd3) 

Untreated 

Lateral 

Resistance 

(kips)

Increase in 

Lateral 

Resistance 

(kips)

Percent 

Increase in 

Resistance

Jet grouting below cap. 15’ x 10.5’ x 10’ 58.3 282 500 160

Jet grouting adjacent to cap. 6,6’ x 13’ x 12’ 38.1 214 398 185

Soil mixing adjacent to cap. 4’ x 11’ x 10’ 16.3 282 170 60

Weal flowable fill below cap. 13.5’ x 8.8’ x 6’ 26.4 232 24 10

Flowable fill adjacent to cap. 6’ x 12’ x 6’ 16.0 265 145 55

Compacted fill to edge of cap. 9.6’ x 8.75’ x 3.5’ 10.9 232 23 10

Compacted fill 5 feet 

beyond edge of cap. 14.6’ x 8.75’ x 3.5’ 16.6 232 40 18

Rammed aggregate piers 

adjacent to cap top. 

13, 2.5 dia x 13’ 

deep 29.5 285 40 14

Rammed aggregate piers 

adjacent to cap top. 

13, 2.5’ dia. X 10.5 

deep 23.6 50 35 70

Table 7-25 Summary of Ground Improvement Method and Associated Costs for 

Increase in Lateral Resistance after (Rollins and Brown 2011)

Treatment Method

Ground 

Improvement 

Cost

Pile and Pile 

Cap Cost for 

Same Lateral 

Load 

Increase

Savings 

Relative to 

Adding 

Additional 

Piles

Ground 

Improvement 

Lateral 

Support Cost 

($ / kip)

Jet grouting below cap. $28,500 $84,200 $55,700 57

Jet grouting adjacent to cap. $38,000 $69,360 $31,360 95

Soil mixing adjacent to cap. $10,000 $30,345 $20,345 59

Weak flowable fill below cap. $3,180 $4,335 $1,155 133

Flowable fill adjacent to cap. $3,600 $26,010 $22,410 25

Compacted fill to edge of cap. $544 $4,335 $3,791 24

Compacted fill 5 feet beyond edge of cap. $828 $8,670 $7,842 21

Rammed aggregate piers adjacent to cap top. $4,225 $8,670 $4,445 106

Rammed aggregate piers adjacent to cap top. $4,225 $8,670 $4,445 121

 
374

7.3.8 Lateral Squeeze of Foundation Soil and Solutions

Bridge abutments supported on piles driven through soft compressible cohesive 

soils may tilt forward or backward depending on the geometry of the backfill and the 

abutment. This problem is illustrated in Figure 7-76. Large horizontal movements 

may cause damage to the structure. The unbalanced fill loads displace the soil 

laterally. This lateral displacement may bend the piles, causing the abutment to tilt 

toward or away from the fill. Lateral squeeze can similarly adversely affect pier 

locations if stockpiles are placed adjacent to a pier location.

The following rules of thumb are recommended for determining whether tilting will 

occur, as well as estimating the magnitude of horizontal movement.

1. Lateral squeeze and abutment tilting can occur if:

 Eq. 7-74

Where: 

γf = unit weight of fill (pcf).

hf = height of fill (feet).

su = undrained sear strength of soft cohesive soil (psf).

2. If abutment tilting can occur, the magnitude of the horizontal movement can 

be estimated by the following formula:

 Eq. 7-75

Where: 

Sh = horizontal abutment movement (inches).

Sv = vertical fill settlement (inches).

Mitigation of lateral squeeze may be provided by several means. The four solutions 

below represent primary methods to prevent lateral squeeze.

a. Delay installation of abutment piling until after fill settlement has stabilized 

(best solution).

b. Provide expansion shoes large enough to accommodate the movement.

c. Use steel H-piles to provide high tensile strength in flexure.

d. Use lightweight fill materials to reduce driving forces.
375

Figure 7-76 Examples of abutment tilting due to lateral squeeze.

7.3.9 Overall Stability

Abutments are often supported by a deep foundation system and are also subject to 

the potential for overall stability failure. Although the stiffness of deep foundations 

may provide resistance against this shear loading, if the slip surface is below the pile 

tip elevation, or if large earthen loads are generated, global stability failure should be 

evaluated. To a lesser extent, this situation may apply to walls or other cut slopes 

which are also founded on structural elements. Per AASHTO (2014), the Service I 

Load Combination and load factor should be used to assess global stability. When 

the stabilizing system includes a structural element (deep foundation), a resistance 

factor of 0.65 for sliding is recommended for use in stability calculations. As 

provided in the AASHTO (2014) commentary, the 0.65 resistance factor is taken as 

the inverse of the typical 1.5 safety factor used in the ASD approach. A detailed 

discussion on the mechanisms of shear failure and slope stability is beyond the 

scope of this manual. References can be made to Sabatini et al. (1997), Duncan 

and Wright (2005) and Tanyu et al. (2008). 
376

7.4 EXTREME EVENT LIMIT STATES

Extreme event limit states involve events with a low occurrence probability. 

However, these events can have a detrimental impact on the foundation and 

structure if not considered in design. As noted in AASHTO (2014), the extreme 

event limit state exists to protect against structural collapse and preserve life. 

Furthermore, AASHTO (2014) provides guidance on extreme event loading cases, 

including ice loading and vehicle and vessel impacts. Arneson et al. (2012) details 

scour development near bridge piers and abutments, while Kavazanjian et al. (2011) 

provides an in depth discussion on seismic events. Background information on ice 

loads can be found in in Montgomery et al. (1984). In depth coverage for vessel 

impact design is available in AASHTO (2009) Guide Specifications and Commentary 

for Vessel Collision Design of Highway Bridges. The following sections provide a 

discussion on these event types, while the documents referenced above address 

these topics in greater detail. 

Extreme event limit state evaluations ensure structural survival of a bridge under 

unique major occurrences such as earthquakes, floods, and vehicle or vessel 

collisions having return periods significantly greater than the bridge design life. 

Extreme event limit states for driven pile foundation design include:

• the check flood for scour, 

• vessel collision, 

• vehicle collision, 

• seismic loading, 

• ice and debris loading, and 

• other site-specific situations determined by the design engineer.

7.4.1 Extreme Event Scour During Check Flood 

Many bridges are constructed over rivers, bodies of water, and tidal areas where 

scour is a consideration in the foundation design. These structures encounter 

repeated aggradation or degradation of material in the scour zone which affects the 

foundation resistance. The check flood is an extreme event that is unlikely to 

happen during the design life of the structure. Arneson et al. (2012) details design 

requirements for the check flood where soil materials within the scour zone are 

assumed to be removed and provide no geotechnical resistance for lateral, axial

compression, or tension loads.
377

For the applicable factored loads, the pile foundation must have adequate factored 

axial and lateral resistance. In the extreme event limit state, the check flood for 

scour differs from the design flood as follows:

1. The 500-year check flood has a different and typically deeper scour prism.

2. The loads used for structural modeling in the extreme event limit state are 

unfactored. 

3. Resistance factors of 1.0 are used for the geotechnical resistances on lateral 

and axial compression loads, and 0.8 or less for uplift loads. 

Overall stability of the bridge should be evaluated during the check flood, where 

storm surges, tides or other floods can detrimentally impact the foundation and 

overall structure. Debris loads during the check flood should likewise be included in 

this analysis. Piles should be driven to penetration depths to overcome the loss of 

shaft resistance within the scour prism.

Loading combinations with the check flood are further explained in Section 7.4.4, as 

well as in HEC-18 (Arneson et al. 2012). Section 7.2.9 also provides additional 

background on scour design considerations.

7.4.2 Seismic and Seismic Induced Downdrag

The design issues associated with pile foundation design for seismic events are 

significant and include liquefaction effects on soil resistance, ground movements, 

seismic induced foundation loads, and seismic induced drag forces. As noted in 

AASHTO (2014) design specifications, bridges shall be designed for a low 

probability of collapse but may suffer significant damage or disruption of service as a 

result of earthquake loading. A pseudo-static analysis approach is commonly 

utilized for many bridges and will be covered in this section. In addition, subsequent 

sections of this manual will discuss seismic action effects with respect to the 

extreme event limit states. However, more detailed references should be consulted 

on the seismic design procedure including Geotechnical Engineering Circular No. 3 

by Kavazanjian et al. (2011) and FHWA-NHI-15-004 LRFD Seismic Analysis and 

Design of Bridges by Marsh et al. (2014).
378

7.4.2.1 AASHTO Recommendations for the Equivalent Static Seismic Force

For earthquake loads, the horizontal ground displacements are transmitted from the 

foundation to superstructure where accelerations occur. As a result, inertial forces 

are generated and applied back to the foundation which must then resist these 

loads. The superstructure must be designed to resist brittle failure as a result of 

seismic loading (Marsh et al. 2014) while the foundation must resist effects of 

liquefaction, ground movement, drag forces and the increased superstructure loads. 

Before defining the required nominal geotechnical resistance, the increased loads 

resulting from seismic action must first be determined.

STEP BY STEP PROCEDURE FOR: “EVALUATION OF EARTHQUAKE LOAD”

The seismic hazard of a site contributes to expected loads that can develop for a 

1000 year design earthquake (AASHTO 2014). AASHTO Article 3.10 outlines steps 

to determine the elastic seismic coefficient, Csm. However the end goal is to define 

the equivalent static horizontal seismic force, Pe(x). This relationship is shown in 

Equation 7-76, with the following steps providing a discussion of this process.

 Eq. 7-76

Where: 

Pe(x)= equivalent static horizontal seismic force acting on superstructure.

Csm = elastic seismic response coefficient (dimensionless).

W = equivalent weight of the superstructure.

STEP 1 Define the Site Ground Coefficient and Spectral Coefficients.

The site peak ground acceleration coefficient, PGA, short period spectral coefficient, 

Ss and long period spectral coefficient, S1, are determined by inspecting contour 

seismic maps developed for such a purpose. These maps were developed by the 

U.S. Geological Survey for AASHTO and may be found in the Chapter 3 of AASHTO 

(2014) design specifications. In addition, the U.S. Geological Survey website 

includes an application (http://earthquake.usgs.gov/designmaps/us/application.php) 

which performs a search and presents these output values, after the user enters the 

site location and site classification.
379

STEP 2 Determine the Site Classification.

Using results of the subsurface investigation, the site is classified as A to F. The 

upper 100 feet of the subsurface profile is to be averaged to define the shear wave 

velocity, SPT N-Value, and undrained shear strength. Table 7-26 should be used for 

this determination.

Table 7-26 Site Class Definition

Site Class Soil Type and Profile

A Hard rock with measured shear wave velocity, > 5,000 ft/s. 

B Rock with 2,500 ft/sec < < 5,000 ft/s. 

C Very dense soil and soil rock with 1,200 ft/sec < < 2,500 ft/s, or with 

either > 50 blows/ft, or > 2.0 ksf. 

D Stiff soil with 600 ft/s < < 1,200 ft/s, or with either 15 < < 50 

blows/ft, or 1.0 < < 2.0 ksf. 

E Soil profile with < 600 ft/s or with either < 15 blows/ft or < 1.0 

ksf, or any profile with more than 10 ft of soft clay defined as soil with PI 

> 20, w > 40 percent and < 0.5 ksf. 

F Soils requiring site-specific evaluations, such as:

• Peats or highly organic clays (H > 10 ft of peat or highly organic clay 

where H = thickness of soil). 

• Very high plasticity clays (H > 25 ft with PI > 75). 

• Very thick soft/medium stiff clays (H >120 ft). 

 = average shear wave velocity

 = averageundrained shear strength

PI = Plasticity Index

 = average SPT N- value

w = moisture content 

STEP 3 Determine the Site Factors.

Site factors corresponding to the zero, short and long periods of acceleration should 

be determined using Table 7-27 to Table 7-29. Straight line interpolation should be 

used for intermediate values in any of the three tables. If the site is classified as 

Class F, a site-specific geotechnical investigation with dynamic response analysis 

should be performed to determine these acceleration period parameters.
380

Table 7-27 Site Factor Values, Fpga, at Zero Period Acceleration

Site Class PGA < 0.1 PGA = 0.2 PGA = 0.3 PGA = 0.4 PGA > 0.5

A 0.8 0.8 0.8 0.8 0.8

B 1.0 1.0 1.0 1.0 1.0

C 1.2 1.2 1.1 1.0 1.0

D 1.6 1.4 1.2 1.1 1.0

E 2.5 1.7 1.2 0.9 0.9

F * * * * * 

Table 7-28 Site Factor Values, Fa, for Short Period Coefficient Ss

Site Class Ss < 0.25 Ss = 0.5 Ss = 0.75 Ss = 1.0 Ss > 1.25

A 0.8 0.8 0.8 0.8 0.8

B 1.0 1.0 1.0 1.0 1.0

C 1.2 1.2 1.1 1.0 1.0

D 1.6 1.4 1.2 1.1 1.0

E 2.5 1.7 1.2 0.9 0.9

F * * * * * 

Table 7-29 Site Factor Values, Fv, for Long Period Coefficient, S1

Site Class S1 < 0.1 S1 = 0.2 S1 = 0.3 S1 = 0.4 S1 > 0.5

A 0.8 0.8 0.8 0.8 0.8

B 1.0 1.0 1.0 1.0 1.0

C 1.7 1.6 1.5 1.4 1.3

D 2.4 2.0 1.8 1.6 1.5

E 3.5 3.2 2.8 2.4 2.4

F * * * * * 

STEP 4 Characterize the Design Response Spectrum and Determine the Elastic 

Seismic Response Coefficient, Csm

The Design Response Spectrum as shown in Figure 7-77 is created using the zero, 

short, and long period acceleration values as well as the site factor values 

determined from the steps above.
381

Figure 7-77 Design response spectrum (after AASHTO 2014).

The elastic seismic response coefficient is a function of the Design Response 

Spectrum and a hand calculation may be used to determine this exact value.

Equation 7-77 to Equation 7-79 should be employed for this purpose.

 Eq. 7-77

In which

 Eq. 7-78

 Eq. 7-79

Where: 

Csm = elastic seismic response coefficient (dimensionless).

As = peak seismic ground acceleration coefficient modified by shortperiod site factor (dimensionless).

SDS = Csm value with a period of 0.2 second = FaSs

Tm = period of vibration of mth mode(s).

To = reference period to define spectral shape = 0.2Ts. 

Ts = corner period when spectrum changes from independent to inversely 

proportional = SD1 /SDS

FPGA= zero period site factor.
382 

PGA = peak ground acceleration coefficient.

Fa = short period site factor.

Ss = short period spectral coefficient. 

SD1 = Csm value with a period of 1.0 seconds = FvS1. 

The elastic seismic response coefficient, Csm, is then substituted into Equation 7-76 

as noted to determine the equivalent static force. The structural engineer applies 

this force to the superstructure using a method outlined in Article 4.7.4.3 of the 

AASHTO (2014) design specifications. The minimum analysis requirement depends 

on the seismic zone in which the bridge exists, as well as the bridge regularity and 

operational classification. Table 7-30 classifies the site as Seismic Zone 1 through 4 

depending on the acceleration coefficient, SD1. If the site is classified as Seismic 

Zone 2 through 4, a liquefaction assessment is required for multispan bridges.

Table 7-30 Seismic Zones

Acceleration Coefficient, SD1 Seismic Zone

SD1 ≤ 0.15 1 

0.15 < SD1 ≤ 0.30 2 

0.30 < SD1 ≤ 0.50 3 

0.50 < SD1 4 

For the Extreme Event I check, the factored loads resulting from seismic action are 

then applied to the foundation for an axial, lateral and overturning analysis. A 

resistance factor of 1.0 is used on the resistances for axial compression loads and 

0.8 or less for lateral and uplift loads as recommended by AASHTO (2014).

7.4.2.2 Liquefaction 

Liquefaction is defined as a loss of shear strength and stiffness which results from 

built-up pore water pressure during cyclic loading. Soil types most susceptible to 

liquefaction are saturated, very loose to medium dense, fine to medium grained 

sands and non-plastic silts. However, liquefaction has also occurred in saturated, 

very loose to medium dense gravels. 

In seismically active areas where peak earthquake acceleration will be greater than 

0.1g, the soil susceptibility to liquefaction should be evaluated. A commonly used 

procedure for identification of liquefaction susceptible soils was proposed by Seed et 

al. (1983). This liquefaction evaluation approach is detailed in Lam and Martin 
383

(1986), the Commentary for Article 10.5.4 of AASHTO (2014), and Chapter 6 of 

GEC-3 by Kavazanjian et al. (2011). If the soils are found to be subject to 

liquefaction during the design event, the pile foundation must be designed to 

accommodate the following resulting behavior: the loss of resistance in the liquefied 

zone, the seismic induced loads, as well as the anticipated vertical and horizontal 

displacements, and the resulting drag force. Alternatively, the liquefaction potential 

may be mitigated through ground improvement techniques.

Pile foundations in liquefiable soils must penetrate through the zone of liquefaction 

and develop adequate resistance in the underlying deposits. Evaluation of 

compression and uplift resistances during the seismic event can be made by 

assigning residual strength properties to the liquefiable layers. Appropriate residual 

strengths should be used to analyze axial and lateral loading. Residual strengths of 

sands and silty sands can be approximated from SPT resistance values using a 

correlation proposed by Seed (1987) and updated by Seed and Harder (1990). 

Additional correlations on the residual strengths of sands and silty sands have been 

reported by Olson and Stark (2002), as well as Idriss and Boulanger (2007). Figure 

7-78 presents the correlation developed by Idriss and Boulanger between the 

equivalent clean sand SPT corrected blow count and the ratio of the residual shear 

strength, sr, divided by the vertical effective stress, σ’vo. 

Figure 7-78 Correlation between the Residual Undrained Strength Ratio, sr / σ’vo

and equivalent clean sand SPT blow count, (N1)60-e (Idriss and Boulanger 2007). 
384

The SPT N-value used for the equivalent clean sand in Figure 7-78 should be 

corrected for both energy and vertical effective stress as well as for the percent 

fines, as described by Seed and Harder (1990). The correction for the percent fines 

is provided in Table 7-31. 

Table 7-31 Blow Count Correction, Ncorr, for the Equivalent Clean Sand Blow 

Count, (N1)60-e 

Percent Passing #200 Sieve Ncorr

0-9 0 

10-24 1 

25-49 2 

50-74 4 

> 75 5 

The equivalent clean sand blow count is then determined using Equation 7-80

below.

 Eq. 7-80

Where: 

(N1)60-e = equivalent clean sand blow count.

(N1)60 = SPT N value corrected for energy and overburden stress. 

Ncorr = correction for percent fines from Table 7-31. 

In a Washington State research report, Kramer (2008) proposed an alternate 

approach for estimating the residual strength of liquefied soil. This approach 

estimated the residual strength based on a weighted value determined by four 

methods. This residual strength was based on 20% of the value determined 

following Idriss (1998), 20% of the value determined in accordance with from Olson 

and Stark (2002), 20% of the value determined following Idriss and Boulanger 

(2007), and 40% of the value determined using the Kramer-Wang hybrid model, 

Kramer and Wang (2007).

Following a seismic event that induces soil liquefaction, the liquefied layer will 

generally consolidate as pore water pressure dissipates. Settlement magnitude can 

be estimated using procedures in Kavazanjian et al. (2011). Fellenius and Siegel 

(2008) noted that the location of the zone of liquefaction relative to the neutral plane 

is very important. If liquefaction occurs in soil layers above the location of the 

neutral plane before liquefaction, the liquefaction event will have limited effect on the 
385

pile. If liquefaction occurs in soil layers located below the pre-liquefaction location of 

the neutral plane, it will increase the axial compression load in the pile as well as 

result in additional pile settlement. The structural design of the pile section and the 

settlement resulting from liquefaction occurring below the neutral plane should be 

evaluated in the design. The pile foundation must be structurally capable of 

supporting the increased drag force and foundation settlement occurring after 

liquefaction must be within the structure's performance criteria.

Liquefaction induced lateral spread can impose significant bending moments in piles 

driven through liquefiable soils. Therefore, piles in liquefiable soils should be flexible 

and ductile in order to accommodate lateral loads. The maximum bending moment 

of piles in liquefiable soils is often evaluated in a p-y analysis by assigning Reese's 

soft clay p-y curve with low residual shear strengths and high ε50 values to the 

liquefiable layer. Brandenberg et al. (2007) performed centrifuge model tests and 

found p-multiplier values of 0.05 in loose sand and 0.30 in dense sand could be 

used to model pile p-y response in fully liquefied granular layers.

7.4.3 Ice and Collisions

Vehicle and vessel collision as well as ice loads are included in the Extreme Event II 

load combination provided in AASHTO (2014). Each is treated as an independent 

extreme event. However they often occur in tandem with other loading cases such 

as large wind loads and vessel impact. Since these loads are typically applied to the 

superstructure, they are analyzed for structural loading cases, and therefore do not 

directly alter the geotechnical resistance as with scour and earthquakes. The 

Extreme Event II limit state equation can be expressed as Equation 7-81 below.

Q DL LL WA FR ( CVorCTorIC ) pii ∑ γγ ++++= 0.10.10.15.0 Eq. 7-81

Where: 

γi = load factor.

Qi = force effect.

γp = load factor for permanent loads (from Table 2-3 of Chapter 2).

DL = dead loads. 

LL = live loads. 

WA = water load. 

FR = friction load. 

IC = ice load. 

CT = vehicular collision force. 

CV = vessel collision force. 
386

From the equation above, ice loads as well as vehicle and vessel collisions are 

treated as separate events. Therefore they will be independently discussed.

7.4.3.1 Ice Loads

AASHTO (2014) provides consideration for ice loading that occurs in freshwater 

bodies such as lakes and rivers. Specific mention of saltwater ice loading is left to

specialists in that area, and are thus beyond the scope of this manual. The 

expected ice forces are assumed to act directly on piers. Four loading conditions 

may occur including; 1) dynamic impact from ice sheet collision with the bridge pier, 

2) static load resulting from thermal expansion of ice sheets around piers, 3) load

from hanging dams or ice jams and 4) static uplift or vertical loads that result from 

ice adhesion when water levels fluctuate.

Section 3.9 of AASHTO (2014) provides in depth detail on the magnitude and 

application of ice loads resulting from the above mentioned factors. A discussion of 

dynamic forces resulting from ice collisions is mainly provided by Montgomery et al. 

(1984), where forces depend on the flow size, the ice strength and thickness, and 

pier geometry. Meanwhile, presumed values are noted in AASHTO (2014) as to ice 

crushing strength.

Static loads applied from thermal expansion and hanging dams or ice jams are 

accounted for in the structural design of superstructure members. Increased forces 

are therefore applied during a static structural analysis, where they generate 

alternative loading conditions for the foundations. To account for vertical forces from 

ice adhesion in rapid water level fluctuations, AASHTO (2014) recommends 

Equation 7-82 to determine the vertical force surrounding circular piers. Equation 

7-83 should be used for oblong piers.















 = + 75.0

2 03.0 35.080

i

i

t

R tF Eq. 7-82















 = + + 75.0

25.1 2 03.0 35.0802.0

i

i i

t

R tCtF Eq. 7-83

Where: 

F = vertical force.

ti = ice thickness (feet).

C = perimeter of pier excluding half circles at ends of oblong pier (feet).

R = radius of pier (feet).
387

7.4.3.2 Vehicle Collison

When bridge piers are within 30 feet of the roadway edge or within 50 feet of the 

railway centerline, vehicular collision should be evaluated. In the absence of crash 

protection, an equivalent 600 kip static force is to be applied 5 feet above grade and 

at an angle of 0 to 15 feet from the edge of pavement. This loading condition is 

based on full scale tractor trailer collision tests. A distributed load of no more than 5 

feet wide by 2 feet high may be alternatively applied for wall piers, if considered a 

suitable replacement for an anticipated vehicle.

The designer may also utilize a barrier to redirect or absorb the vehicular collision. 

For this, an embankment, guardrail or other crash protection system may be used. 

AASHTO (2014) requires this barrier to be structurally independent of the pier and 

ground mounted. If within 10 feet of the pier, the barrier must be 54 inches high, 

while a height of 42 inches must be reached if outside of this distance.

7.4.3.3 Vessel Collision

Bridges that span navigable bodies of water should be designed for vessel impact. 

Bow, deck house, or mast impacts are typically applied to superstructure elements. 

A head on collision with piers may occur and should consider the full vessel mass. 

Based on an assessment of vessels passing the bridge, a design vessel is selected 

and given a deadweight tonnage (DWT) value of representative size and weight. 

The head on collision force to the pier may be calculated as follows. 

s = 15.8 VP DWT Eq. 7-84

Where: 

Ps = equivalent static vessel impact force (kips).

DWT = deadweight tonnage (tonne).

V = vessel impact velocity (ft/s).

After determination of the equivalent static impact force, the load is to be applied as 

either 100 percent in the direction parallel to the channel centerline or 50 percent in 

the direction normal to the channel centerline. Both loading cases should be 

evaluated. Overall stability is determined by applying a point load at the mean high 

water level as shown in Figure 7-79. AASHTO (2014) notes that all substructure 

components exposed to direct impacts shall be designed to resist the applied loads.
388

Figure 7-79 Application of equivalent impact force 

(after AASHTO 2014).

7.4.4 Combined Extreme Events

Two extreme event combinations with scour are given in AASHTO (2014). Extreme 

Event I is in combination with earthquake loading, while the Extreme Event II 

combination includes ice, vessel and vehicle collisions and hydraulic loads. As 

noted in Arneson et al. (2012), the Extreme Event I combination has a low 

occurrence probability for both the check flood and earthquake loading. Therefore 

scour for the mean discharge or normal non-flood flow may be applied to this event 

combination. For the Extreme Event II combination, research is ongoing to assess 

the probability of joint loading conditions during the check flood, and judgement 

should be used based on site-specific factors. If ice or debris jams near the 

structure dictates the use of a more extreme flood event than the check flood, this 

may be used to assess the extreme event limit state. 

7.5 DETERMINATION OF MINIMUM PILE PENETRATION

Foundation settlement and resultant structure deformation should be kept within 

tolerable limits as described in Section 7.3.1. Once all limit state analyses have 

been performed, minimum pile penetration depths should be specified, if necessary. 

Minimum pile penetration depths may be required to limit structure vertical or lateral 

deformations under applied loads. 
389

A minimum pile penetration depth should be specified only if needed. In many 

cases, the inclusion of only an estimated length is satisfactory with recognition that 

once piles achieve their required nominal driving resistance, driving may be 

terminated shorter than the estimated length. The ability of an appropriately sized 

pile hammer to drive piles to the specified minimum penetration depth should be 

evaluated during the design phase in a wave equation drivability analysis. The 

drivability analysis is used to check constructability and viability of achieving the 

minimum pile penetration depth. An alternative pile type or use of pile installation 

aids may need to be specified if the drivability analysis indicates the minimum pile 

penetration depth is not obtainable using conventional driving procedures.

AASHTO (2014) design specifications state that the minimum pile penetration depth 

shall be the pile toe elevation needed to satisfy the following requirements, as 

applicable, to the site and loading conditions. Additional guidance on each topic 

affecting the minimum pile penetration requirement is provided in the noted sections.

a. Single pile and pile group settlement (service limit state – Section 7.3.5). 

b. Lateral deflection (service limit state – Section 7.3.7). 

c. Uplift (strength limit state – Section 7.2.3).

d. Penetration depth into soils to accommodate drag forces from static 

settlement stresses (service limit state – Section 7.3.6). 

e. Penetration depth into soils due to liquefaction (strength and extreme 

event limit states – Section 7.4.2). 

f. Penetration depth into soils needed to provide adequate pile axial 

(compression and uplift) and lateral resistance after scour (strength limit 

state – Section 7.2.9, and extreme event limit states – Section 7.4.1).

g. Penetration depth into soils necessary to achieve fixity for resisting the

applied lateral loads (strength limit state – Section 7.2.5, and service limit 

state – Section 7.3.7). It should be noted that AASHTO is silent on the 

definition of fixity, which can vary if defined by the deflection profile, the 

bending moment profile, or the pile head deflection.

h. Axial uplift and lateral resistance to resist extreme event limit states loads

(extreme event limit state – Section 7.4). 
390

7.6 DETERMINATION OF Rndr TO ESTABLISH CONTRACT DRIVING CRITERIA

The required nominal driving resistance, RRnnnnrr, is used to establish the driving 

criteria. This resistance along with the field method for resistance verification should 

be specified in contract documents. The nominal driving resistance includes many 

factors such as the axial compression loads, resistance from scourable or liquefiable 

layers, as well as time dependent soil strength changes. Some of these conditions 

result in an increased soil resistance at the time of installation. In these cases, the 

pile should be driven to a higher nominal resistance to accommodate these future 

resistance losses. Inclusion of the RRnnnnrr in conjunction with the factored load and the 

field verification method can reduce the risk of contractor claims. A minimum 

penetration depth may also be required in addition to the RRnnnnrr to also satisfy uplift or 

lateral loading requirements and/or serviceability.

An example calculation for determining RRnnnnrr is illustrated in Equation 7-85 based on 

the project information provided in Table 7-32. The factored axial load, Q, is 200 

kips in compression. Given the site conditions, some resistance will be lost in a soil 

layer subject to scour. The resistance lost in the scour zone should be determined 

by an appropriate static analysis method and should not be factored per AASHTO 

(2014) Article 10.7.3.6. For this example, it is estimated that 50 kips of resistance 

will be lost in the scour zone. The pile will also be driven into a weathered shale 

formation that historically has exhibited a loss in toe resistance following initial 

driving. The resistance lost to relaxation should be factored based the resistance 

verification method. For this example, it is estimated that 100 kips of toe resistance 

will be lost. Based on field verification by dynamic testing with signal matching, the 

resistance loss from relaxation is then 154 kips.

Equations 7-85 and 7-86 are simplified from AASHTO (2014) to provide the 

necessary calculation steps. 

Table 7-32 Summary of Load and Resistance Information

Q (kips) 200

φdyn 0.65

Rn (kips) 308

Rscour (kips) 50

Rrelax (kips) 154 

Rndr (kips) 512
391

Example computation of Rn

dyn

n

Q R

φ = Eq. 7-85

R kips n 308

65.0

200 ==

Example computation of Rndr

ndr n scour ++= RRRR relax Eq. 7-86

R kips ndr 512

65.0

100 50308  =



 



 ++=

Table As shown in the above example, Rndr is considerably higher than the nominal 

resistance due to the site specific considerations of scour and relaxation. The load 

and resistance factors used in this determination may be found respectively in Table 

2-3 in Chapter 2 and Table 7-2 of this chapter.

7.7 DRIVABILITY ANALYSIS

Greater pile penetration depths are increasingly being required to satisfy minimum 

penetration depth requirements due to extreme events such as scour, vessel impact, 

ice and debris loading, as well as seismic events. Therefore, the ability of a pile to 

be driven to the required penetration depth has become increasingly more important. 

AASHTO (2014) Article 10.7.8 states that a drivability analysis should be performed 

by the engineer and that it be conducted during the design stage. Pile drivability 

refers to the ability of a pile to be driven to the required pile penetration depth and/or 

to the required resistance, and within specified material strength limits. All of the 

previously described static analysis methods are meaningless if the pile cannot be 

driven to the required depth and nominal resistance without sustaining damage. The 

limit of pile drivability is the maximum soil resistance a pile can overcome at a 

reasonable blow count without being damaged while being driven by an 

appropriately sized and properly operating driving system.

Primary factors controlling the nominal geotechnical resistance of a pile are the pile 

type and length, the soil conditions, and the method of installation. Since the pile 

type, length and method of installation can be specified, it is often erroneously 

assumed that the pile can be installed as designed to the estimated penetration 

depth. However, the pile must have sufficient drivability to overcome the soil 
392

resistance encountered during driving to reach the estimated or specified pile 

penetration depth. If a pile section does not have a drivability limit in excess of the 

soil resistance to be overcome during driving, it will not be drivable to the desired 

pile penetration depth. The failure to adequately evaluate pile drivability is one of 

the most common deficiencies in driven pile design practice.

In evaluating the drivability of a pile, the soil disturbance during installation and the 

time dependent soil strength changes should be considered. Both soil setup and 

relaxation have been described earlier in this chapter. For economical pile design, 

the foundation designer must match the soil resistance to be overcome at the time of 

driving with the pile impedance (discussed below), the pile material strength, and the 

pile driving equipment.

7.7.1 Factors Affecting Drivability

A pile must satisfy two aspects of drivability. First, the pile must have sufficient 

stiffness to transmit driving forces large enough to overcome soil resistance. 

Second, the pile must have sufficient structural strength to withstand the driving 

forces without damage.

The primary controlling factor on pile drivability is the pile impedance, EA/C. Once 

the pile material is selected, and thus the pile modulus of elasticity, E, and the pile 

wave speed, C, only increasing the pile cross sectional area, A, will improve the pile 

drivability. For steel H-piles, the designer can improve pile drivability by increasing 

the H-pile section without increasing the H-pile size. This will increase the area and 

impedance without significantly changing the soil resistance on the section. The 

drivability of steel pipe piles can be improved by increasing the pipe wall thickness. 

This again increases the pile impedance without increasing the soil resistance. For 

open ended pipe piles, an inside-fitting cutting shoe can improve drivability by 

delaying the formation of a soil plug and thereby reducing the soil resistance to be 

overcome. Most concrete piles are solid cross sections. Therefore, increasing the 

pile area to improve drivability is usually accompanied by an increase in the soil 

resistance to driving.

A lesser factor influencing pile drivability is the pile material strength. The influence 

of pile material strength on drivability is limited, since strength does not alter the pile 

impedance. However, a pile with a higher pile material strength can tolerate higher 

driving stresses that may allow a larger pile hammer to be used. This may allow a 

slightly higher nominal resistance to be obtained before refusal driving conditions or 

pile damage occurs.
393

Other factors that may affect pile drivability include the driving system characteristics 

such as ram weight, stroke, and speed, as well as the actual system performance in 

the field. The dynamic soil response can also affect pile drivability. Soils may have 

higher damping characteristics or elasticity than assumed, both of which can reduce 

pile drivability. Dynamic soil response is discussed in greater detail in Chapters 10, 

11, and 12. Pile installation aids such as predrilling, jetting and spudding, can assist 

in meeting project pile penetration requirements. However, these installation aids do 

not change the drivability of the pile section.

Even if the geotechnical and structural resistance both indicate a large geotechnical 

resistance could be used, this large geotechnical resistance may still not be 

obtainable because driving stresses may exceed material driving stress limits. A pile 

cannot be driven to a static nominal resistance that is as high as the structural 

resistance of the pile because of the additional dynamic resistance or damping 

forces generated during pile driving. Pile structural resistance and driving stresses 

are presented in Chapter 8.

7.7.2 Methods for Determining Pile Drivability

There are three available methods for predicting and/or checking pile drivability. As 

design tools, all of the methods have advantages and disadvantages and are 

therefore presented in order of increasing cost and reliability.

1. Wave Equation Analysis 

This computer program accounts for pile impedance and calculates estimated 

driving stresses as well as the relationship of pile penetration resistance (blow 

count) versus nominal resistance (Goble and Rausche 1986). Wave equation 

analyses performed in the design stage require assumptions on the hammer 

type and performance level, the drive system components, as well as the soil 

response during driving. These shortcomings are reflected in variations 

between predicted and actual field behavior. Even with these shortcomings, 

the wave equation is a powerful design tool that can and should be used to 

check drivability in the design stage, to design an appropriate pile section, or 

to specify driving equipment characteristics. As noted previously, AASHTO 

(2014) design specification state that a wave equation drivability analysis 

should be performed during design in the strength limit state. Additional 

information on the wave equation, including its use as a construction control 

tool, is presented in Chapter 12.
394

2. Dynamic Testing and Analysis

Dynamic measurements can be made during pile installation to calculate 

driving stresses and to estimate nominal resistance at the time of driving. 

Time dependent changes in nominal resistance can be evaluated during 

restrike tests. Signal matching analysis of the dynamic test data can provide 

soil parameters for a refined wave equation analysis. A shortcoming of this 

method as a design tool is that it must be performed during pile driving. 

Therefore, in order to use dynamic testing information to confirm drivability or 

to refine a design, a test program is required during the design stage. 

Additional details on dynamic testing and analysis, including its use as a 

construction control tool, is presented in Chapter 10.

3. Static Load Tests 

Static load tests are useful for confirming pile drivability and the nominal 

resistance prior to production pile driving (Kyfor et al. 1992). Test piles are 

normally driven to estimated lengths and load tested. The confirmation of pile 

drivability through static load test programs is the most accurate method of 

confirming drivability and nominal resistance since a pile is actually driven 

and statically load tested. However, this advantage also illustrates one of its 

shortcomings as a design tool, in that a test program is required during the 

design stage. Other shortcomings associated with static load tests for 

determining drivability include:

a. cost and time delay that limit their suitability to certain projects.

b. assessment of driving stresses and the extent of pile damage, if any, is

not provided by the test.

c. can be misleading on projects where soil and/or rock conditions are 

highly variable.

Additional details on static load tests, including its use as a construction 

control tool, are presented in Chapter 9. Rapid load tests can also be used to 

evaluate nominal resistance and are discussed in Chapter 11. 

As design and construction control tools, methods 1 and 2 offer additional 

information and complement static load tests. Used properly, methods 1 and 2 can 

yield significant savings in material costs or reduction of construction delays and 
395

risks of claims. These methods can be used to reduce the number of static load 

tests and increase the usefulness and applicability of the static load test performed. 

A determination of the increase (soil setup) or decrease (relaxation) in geotechnical 

resistance with time can also be made if piles are tested under restrike conditions 

after initial driving.

7.7.3 Drivability versus Pile Type

Drivability should be checked during the design stage of all driven piles. It is 

particularly important for closed end steel pipe piles where the impedance of the 

steel casing may limit pile drivability. Although the designer may attempt to specify a 

thin wall pipe in order to save material cost, a thin wall pile may lack the drivability to 

develop the required nominal resistance or to achieve the necessary pile penetration 

depth. This concept is illustrated in the wave equation example problem presented 

in Section 12.5.8 of Chapter 12. Wave equation drivability analyses should be 

performed in the design stage to select the pile section and wall thickness per 

AASHTO (2014) design specifications. An example of a wave equation drivability 

assessment is presented in Section 12.5.3.

Steel H-piles and open pipe piles, prestressed concrete piles, and timber piles are 

also subject to drivability limitations. This is particularly true as extreme events and 

the trend to use larger nominal resistances require increased pile penetration 

depths. The drivability of long prestressed concrete piles can be limited by the pile's 

tensile strength.

7.8 CONSIDERATIONS FOR BATTER PILE DESIGN AND CONSTRUCTION

Battered or inclined piles are piles that are driven on an inclination from vertical and 

are often considered when large static lateral loads are expected or where structural 

rigidity is required. A battered pile can typically resist larger horizontal loads and 

experience less deflection than their vertical counterparts at the same loading levels. 

When static loads are applied, battered piles generally perform well. However under 

dynamic horizontal loads or for seismic events, the increased structural stiffness 

benefit decreases. In seismic design particularly, earthquake induced 

displacements generate load, and as force is a function of stiffness, the increased 

stiffness from battered piles results in larger applied horizontal forces to the 

structure. Where significant drag forces are expected, battered piles should be 

avoided due to the increased bending moment loads along the pile length.
396

The response of a pile group to applied loads will differ if the pile group consists 

solely of vertical piles, or if the pile group consists of a combination of vertical and 

batter piles. Figure 7-80 illustrates the response of a foundation unit subjected to a 

horizontal load and an overturning moment. In Case A, the foundation unit is 

supported by a single pile that rotates about a point below the ground surface in 

response to the applied load and moment. In Case B, the foundation unit is

supported by two vertical piles and it translates horizontally and displaces vertically 

downward in response to the applied load and overturning moment. In Case C, the 

foundation unit is supported by an outwardly battered lead pile and a trailing vertical 

pile. This foundation unit translates horizontally and displaces vertically upward in 

response to the applied load and overturning moment. Hence, the influence of 

batter piles on the foundation response must be considered in the foundation design.

Figure 7-80 Variation of foundation response depending on group configuration 

and batter (after Wilson et al. 2006).

For batter piles, the horizontal component of the axial load contributes to, or 

subtracts from the horizontal resistance of the group depending on the batter pile 

direction relative to the applied load. Wilson et al. (2006) noted that the analysis for 

a pile group containing batter piles is similar to the procedures described earlier in 

Section 7.3.7.5 with the following exceptions:
397

1. Using a rigid cap assumption, the vertical load at the head of each pile is 

resolved into the axial pile load based on the batter angle. The horizontal 

component of the axial pile load is then computed. 

2. The sum of all horizontal components is computed, and the total compared to 

the applied horizontal load. If the sum of the horizontal components is less 

than the applied horizontal load, the remainder of the horizontal load must be 

resisted through pile bending computed in a p-y analysis. If the sum of the 

horizontal components are greater and opposite of the applied horizontal 

load, the piles will bend in the opposite direction of the applied load.

3. If no vertical load is applied to a batter pile by the pile cap, the batter pile will 

have no axial load. Without an axial load, it will have no horizontal 

component to resist an applied horizontal load. A battered pile in this 

scenario behaves as a vertical pile and the horizontal resistance is provided 

by pile bending.

Careful assessment of the direction of the applied horizontal loads and 

corresponding need, location, and orientation of batter piles is required when batter 

piles are incorporated into a foundation design. As noted above, adding batter piles 

to resist horizontal loads is inefficient and could be more economically handled by 

vertical piles if the batter pile receives no vertical load from the pile cap. 

Batter pile interference with the surrounding environment should be carefully 

reviewed during the design. The selected batter angle as well as pile deviation from 

the selected batter angle should be evaluated for potential conflicts with adjacent inuse foundations, cofferdams and excavation support systems, utilities, abandoned 

deep foundations from previous construction, tunnels, and other constraint posed by 

the surrounding environment. In situations where an offshore lead and template 

system are used with batter piles, additional bending stresses from the weight of the 

pile hammer and the unsupported portion of the pile above the template can cause 

additional stresses that must be considered in addition to the axially oriented driving 

stresses. From a constructability standpoint, batter angles greater than 1H:3V 

should be avoided.

7.9 CORROSION AND DETERIORATON

Corrosion and deterioration of all pile types must be considered for the respective 

environmental conditions. Section 6.12 of Chapter 6 provides extensive detail on 
398

design considerations for aggressive subsurface environments. AASHTO (2014) 

recommends at minimum, an evaluation for steel corrosion particularly in fill soils, 

low pH soils, and marine environments. For concrete piles, sulfate, chloride and 

acid attack should be evaluated. Timber piles are subject to decay from wetting and 

drying cycles as well as attack from marine bores and insects. Additional details can 

be found in AASHTO (2014) Article 10.7.5 and associated commentary.

7.10 ADDITIONAL DESIGN AND CONSTRUCTION CONSIDERATIONS

The previous sections of this chapter addressed typical geotechnical design and 

analysis procedures for pile foundation design. However, the designer should be 

aware of additional design and construction considerations that can influence the 

reliability of static analysis procedures in estimating nominal resistance. These 

issues include the influence of time, predrilling or jetting, driving piles though 

embankments, soil densification, and the plugging of open pile sections on soil 

resistance. Pile driving induced vibrations can also influence the final design and 

resistance calculation results if potential vibration levels dictate changes in pile type 

or installation procedures. These final sections serve to fill the gaps between design 

and construction events that may influence long term resistance and/or construction 

procedures.

7.10.1 Minimum Pile Spacing, Clearance, and Cap Embedment

AASHTO (2014) design specifications recommend a minimum center-to-center pile 

spacing of no less than 30 inches or 2.5 pile diameters. In addition, the pile cap 

edge should be greater than 9 inches from the side of the nearest pile. As 

discussed previously in this chapter, center-to-center spacing less than 3.0 

diameters can create construction difficulties and lower pile group axial and lateral 

resistances. Therefore, the benefit of a smaller cap size should be weighed against 

the reduced resistances and constructability issues that can arise with the smaller 

spacing.

The pile head should extend a minimum of 12 inches into the pile cap. Any 

damaged portions of the pile head should be trimmed and the damaged portion 

removed down to undamaged material before pouring the cap. If piles are attached 

to the cap by embedded bars or strands, the piles should extend a minimum of 6 

inches into the pile cap.
399

When a reinforced concrete bent cap is cast in place atop piles, the concrete cover 

should be a minimum of 6 inches on all sides of the piles. Pile misalignment may 

occur. Therefore an additional allowance should be given for cover in conjunction 

with specified maximum permissible pile misalignment and its remedy. Where pile 

reinforcement is anchored into the cap satisfying AASHTO (2014) Article 5.13.4.1 

requirements, the projection may be less than 6.0 inches.

7.10.1.1Special Considerations for Large Pile Sizes 

AASHTO design specifications do not address minimum pile spacing requirements 

or pile cap embedment for large diameter open end pipe piles. The synthesis study 

on large diameter pipe piles by Brown and Thompson (2015) presents limited details 

on this subject and most state agencies do not currently have standard plan details 

for these large pile sizes.

One detail used to connect large diameter pipe piles into the pile cap consists of 

internally cleaning out the pipe pile to the desired depth, if necessary, and then 

inserting a reinforcing cage into the pile that also extends up into the pile cap. The 

pile is then filled with a concrete plug. In some cases, design details require the 

internal cleanout extended to the scour depth or mudline, and in other instances, the 

concrete plug extends only for the length needed for load transfer.

Another design approach is use an internal reinforcing cage in conjunction with 

welded shear studs on the exterior surface of the pile that extends into the pile cap. 

At least two rows of shear studs are frequently used to develop sufficient load 

transfer from the pile to cap. In addition to the use of rebar cages or cages and 

shear studs, large diameter piles heads typically extend on the order of 3 feet into 

the pile cap to develop the required structural connection.

Little information was provided with respect to center to center pile spacing. 

However because of the large diameter, comparison with drilled shaft minimum 

spacing guidelines may be appropriate. The presented pile layout in Brown and 

Thompson (2015) shows the minimum spacing to be on the order of 3 pile diameters 

center-to- center. Since large diameter open end piles are often selected for a 

foundation design due to their high lateral load resistance, greater center-to-center 

pile spacing may be applicable for the lateral load demand.
400

7.10.2 Identification of High Rebound Soils

During pile driving, the hammer impact causes temporary compression of the soil 

and pile system. For certain soil types, the deformed soil system can spring back 

close to its original position. When this occurs, it proves difficult to continue pile 

driving operations and to achieve the nominal resistance (Hussein et al. 2006). This 

effect is known as high rebound, or elastic rebound, and is generally due to 

increased pore water pressure from driving occurring near the pile toe. As driving 

continues, pore water pressure increases and causes a larger rebound, leading to 

refusal like driving conditions. If pile driving is paused, the pore water pressure 

dissipates and subsequently smaller rebound occurs.

Studies have shown that for dense silty sand, hard silty clay, glacial tills, silt and 

other fine grained saturated soils, high elastic rebound can occur when using 

displacement piles (Likins 1983; Hussein et al. 2006; Cosentini et al. 2010). 

Research is ongoing to determine what specific soil properties or in-situ tests may 

be used to locate and quantify high rebound soils beyond a basic assessment of soil 

types and saturation. Local experience may be of great value in areas where these 

soil conditions exist, as recommendations on pile drivability may provide a significant 

cost savings in design and construction. Pile driving equipment with heavy rams 

operating at short hammer strokes is helpful in these installation conditions.

7.10.3 Soil and Pile Heave

As noted by Hagerty and Peck (1971), whenever piles are driven, soil is displaced. 

This can result in both upward movement (pile heave) and lateral movements of 

previously driven piles. These soil movements can be detrimental to the resistance 

of previously driven piles as well as to adjacent facilities. Obviously, the greater the 

volume of soil displaced by pile driving, the greater the potential for undesirable 

movements of previously driven piles, or damage to adjacent structures. Heave of 

piles primarily supported by toe resistance is particularly troublesome since the pile 

may be lifted from the bearing stratum, thereby greatly reducing the soil resistance 

and increasing the foundation settlement when loaded. Haggerty and Peck noted 

that saturated, insensitive clays display incompressible behavior during pile driving 

and have the greatest heave potential.

When piles are to be installed in cohesive soils, it is recommended that the potential 

magnitude of vertical and lateral soil movements be considered in the design stage. 

If calculations indicate that movements may be significant, use of an alternate low 

displacement pile, or specifying a modified installation procedure (such as predrilling 
401

to reduce the volume of displaced soil) should be evaluated. A step by step 

procedure adapted from Haggerty and Peck for estimating soil and pile heave in 

saturated insensitive clay follows. The procedure assumes a regular pile driving 

sequence and a level foundation surface. The paper by Haggerty and Peck should 

be consulted for modifications to the recommended procedure for conditions other 

than those stated.

STEP BY STEP PROCEDURE FOR ESTIMATING SOIL AND PILE HEAVE

STEP 1 Calculate the estimated soil heave at the ground surface.

a. Divide the volume of inserted piles by the volume of soil enclosed by 

the pile foundation to obtain the volumetric displacement ratio.

b. Estimate the normalized soil heave (soil heave / pile length) from 1⁄2 

the volumetric displacement ratio calculated in Step 1a.

c. Calculate the soil heave at the ground surface by multiplying the 

normalized soil heave in Step 1b by the average length of piles.

STEP 2 Determine the depth of no pile-soil movement.

a. Figure 7-81 illustrates that an equilibrium depth, dd, exists where the 

potential upward pushing and downward resisting forces on the pile 

shaft are equal.

b. Calculate the pile-soil adhesion along the entire pile shaft using the αmethod described in Section 7.2.1.3.2. 

c. Through multiple iterations determine the equilibrium depth, dd, where 

the adhesion from the upward pushing force equals the adhesion from 

the downward resisting force. Note that only shaft resistance is 

considered in calculating the downward resisting force.

STEP 3 Calculate the estimated pile heave.

a. Calculate the percentage of pile length resisting heaving soil. 

 

Eq. 7-87
402

Where:

L% = percentage of pile length resisting heaving soil (%).

D = pile embedded length (feet).

d = equilibrium depth (feet).

b. Calculate the estimated pile heave by multiplying the estimated soil 

heave from Step 1c by the percentage of pile length resisting heave 

from Step 3a.

Figure 7-81 Balance of forces on pile subject to heave 

(after Haggerty and Peck 1971).

More recent work on this topic was present in Sagaseta and Whittle (2001) where 

the shallow strain path method (SSPM) was used to predict ground movements due 

to pile driving in clay. Results from this method typically yield a reasonable first 

estimate of pile heave.

7.10.4 Piles Driven Through Embankment Fills

Approach embankments are frequently constructed before the commencement of 

pile driving operations. Piles driven through embankment fills should penetrate a 

minimum of 10 feet below original grade unless refusal driving conditions are 
403

encountered on bedrock or a competent bearing layer. An alternative foundation 

type may be more effective in cases where refusal driving conditions occur at a pile 

penetration depth less than 10 feet below the embankment base. Ideally, bedrock or 

a bearing layer that may cause refusal driving conditions is identified by the 

subsurface exploration program performed before foundation construction begins. 

To minimize drag forces, piles should not be driven through embankments or 

constructed over compressible materials until after 90% primary settlement is 

complete.

7.10.5 Effect of Predrilling, Jetting and Vibratory Installation on Nominal 

Resistance

Piles are sometimes predrilled or jetted to a prescribed depth in order to attain the 

pile penetration depths required, as well as to reduce other foundation installation 

concerns, such as ground vibrations. Jetting is usually performed in cohesionless 

soils that can be freely eroded by water jets. Jetting, which can be very effective in 

sands, is usually ineffective in cohesive soils. For clays, and other drillable 

materials, such as thin layers of rock, predrilling the pile locations is more effective. 

The predrilled hole can be slightly smaller, equal to, or slightly larger than the pile 

diameter.

The use of predrilling or jetting will result in greater soil disturbance than considered 

in standard static pile resistance calculations. Therefore, when predrilling or jetting 

is contemplated, the effect of either of these construction procedures on calculated 

compression, uplift, and lateral soil resistance should be considered. Poulos and 

Davis (1980) reported that the shaft resistance should be reduced by 50% of the 

originally calculated resistance in the jetted zone, if the pile is jetted and then driven 

to the final penetration. McClelland et al. (1969) reported that a decrease in shaft 

resistance over a predrilled depth can range from 50 to 85% of that calculated 

without predrilling, depending upon the size of the predrilled hole. Hence, the 

probable reduction in compression, uplift, and lateral resistance from jetting or 

predrilling should be evaluated whenever predrilling or jetting is being considered.

Agencies are often requested to allow pile installation with a vibratory pile hammer 

instead of an impact hammer. Mosher (1987) summarized the results from five sites 

where piles were installed by both impact and vibratory hammers. This study 

concluded that for a majority of the cases, piles installed in sand with a vibratory 

hammer had a lower nominal resistance than impact driven piles at the same site. 

Mosher also concluded that time dependent soil strength changes occurred equally 

for both installation methods. Hence, even with soil setup, the resistance of the 
404

vibratory installed piles did achieve the resistance of the impact driven piles. 

However, it was also observed that impact driving a vibratory installed pile would 

increase the resistance of the vibratory installed pile to that of an impact driven pile.

O’Neill and Vipulanandan (1989) performed a laboratory evaluation of piles installed 

with vibratory hammers. This laboratory study found impact driven piles had a 25% 

greater unit shaft resistance and a 15 to 20% higher unit toe resistance than 

vibratory installed piles in medium dense to dense, uniform, fine sand. However, in 

very dense, uniform, fine sand, the impact driven pile had a 20 to 30% lower unit 

shaft resistance and approximately a 30% lower unit toe resistance than the 

vibratory installed pile.

Ghose-Hajra et al. (2015) compared the effects of pile installation using an impact 

and a vibratory hammer on the long term soil setup on two H-piles and two open-end 

pipe piles driven in a soft clay formation in southeast Louisiana. This study found 

that the impact driven piles always had a greater soil resistance than the companion 

vibratory installed pile at comparable time intervals.

The above studies indicate use of vibratory pile installation rather than impact driving 

will affect the nominal pile resistance that can be achieved at a given pile penetration 

depth. Therefore, communication between design and construction personnel 

should occur, and the influence of vibratory pile installation be evaluated when it is 

proposed. Impact driving a specific final depth of vibratory installed piles may 

provide a foundation that meets the engineer’s performance requirements at 

reduced installation cost.

7.10.6 Densification Effects on Nominal Resistance and Installation Conditions

As illustrated in the previously presented Figure 7-1, driving a pile in cohesionless 

soil influences the surrounding soils to a distance of about 3 to 5 pile diameters 

away from the pile. The soil displacement and vibrations resulting from driving pile 

groups in cohesionless soils can further densify cohesionless materials. The use of 

displacement piles also intensifies group densification effects in cohesionless soils.

Densification can result in the soil resistance as well as the resistance during pile 

driving being significantly higher than that calculated for a single pile in static 

analysis calculations. The added confinement provided by cofferdams or the 

sequence of pile installation can further aggravate a group densification issue. Piles 

should be installed from the center of the group outward in order to reduce group 

densification effects due to installation sequence. Densification can cause 
405

significant construction problems if scour, seismic, or other considerations require 

achieving pile penetration depths that cannot be achieved.

Potential densification effects should be considered in the design stage. Studies by 

Meyerhof (1959) and Kishida (1967) indicate that an increase in the soil friction 

angle of up to 4 degrees would not be uncommon for piles in loose to medium dense 

sands. It is expected that the increase in soil friction angle would be less for dense 

sands or cohesionless soils with a significant fine content. Densification affects the 

nominal resistance to be overcome during driving. As a constructability check, static 

analyses should be performed using higher soil strength parameters than used for 

design to assess nominal resistance due to densification. Results from these static 

analyses may indicate that a low displacement pile should be used, the pile spacing 

should be increased, or that a pile installation aid should be specified in order to 

obtain the required pile penetration depth.

7.10.7 Plugging of Open Pile Sections

Open pile sections include open end pipe piles and H-piles. The use of open pile 

sections has increased, particularly where special design events dictate large pile 

penetration depths. When open pile sections are driven, they may behave as low 

displacement piles and "core" through the soil, or act as displacement piles if a soil 

plug forms near the pile toe. It is generally desired that open sections remain 

unplugged during driving and plugged under static loading conditions.

Stevens (1988) reported that plugging of pipe piles in clays does not occur during 

driving if pile accelerations (along the plug zone) are greater than 22g's. Holloway 

and Beddard (1995) reported that hammer blow size (impact force and energy) 

influenced the dynamic response of the soil plug. With a large hammer blow, the 

plug "slipped" under the dynamic event, whereas under a lesser hammer blow the 

pile encountered toe resistance typically of a plugged condition. From a design 

perspective, these cases indicate that pile penetration of open sections can be 

facilitated if the pile section is designed to accommodate a large pile hammer. Wave 

equation analyses can provide calculated accelerations at selected pile segments.

Static soil resistance calculations must determine whether an open pile section will 

exhibit plugged or unplugged behavior. Studies by O'Neill and Raines (1991), 

Raines et al. (1992), as well as Paikowsky and Whitman (1990) suggest that 

plugging of open pipe piles in medium dense to dense sands generally begins at a 

pile penetration to pile diameter ratio of 20, but can be as high as 35. For pipe piles 
406

in soft to stiff clays, Paikowsky and Whitman (1990) reported plugging can occur at 

penetration-to-pile diameter ratios of 10 to 20.

The above studies suggest that plugging in any soil material may occur under static 

loading conditions once the penetration to pile diameter ratio exceeds 20 in dense 

sands and clays, or 20 to 30 in medium sands. However, plugging is difficult to 

predict with certainty in all cases. For large diameter open ended pipe piles, 

plugging during driving is significantly less likely to occur particularly for piles larger 

than 36 inches in diameter. Plugging of large diameter pipe piles under static 

loading conditions is more likely than in the dynamic conditions. However, it is still 

subject to uncertainty. Forced plugging of large diameter pipe piles using an internal 

constrictor plate as illustrated in Figures 6-14 and 6-15 is sometimes used to reduce 

this uncertainty.

NCHRP Report 20-05 by Brown and Thompson (2015) provides a summary of the 

current industry practices with large diameter open end piles and provides an in 

depth discussion on plugging. This report postulates that large inertial effects 

prevent plug development during driving. Under static loading conditions, plug or 

unplugged behavior for these large diameter piles is uncertain. Further research is 

ongoing regarding design methods on large diameter pipe piles.

An illustration of the difference in the soil resistance mechanism that develops on a 

pipe pile with an open and plugged toe condition is presented in Figure 7-83. 

Paikowsky and Whitman (1990) recommend that the static resistance of an open 

end pipe pile be calculated from the lesser of the following equations:

Plugged Condition:

n soso += AqAfR ppp Eq. 7-88

Unplugged Condition:

n sisisoso −++= WAqAfAfR ppp Eq. 7-89

Where:

Rn = nominal resistance (kips).

fso = exterior unit shaft resistance (ksf).

Aso = pile exterior surface area (ft2).

fsi = interior unit shaft resistance (ksf).

Asi = pile interior surface area (ft2).
407

qp = unit toe resistance (ksf).

App = cross sectional area of pile and soil plug at pile toe (ft2).

Ap = cross sectional area of pile material at pile toe (ft2).

Wp = weight of soil plug (kips).

The soil stresses and displacements induced by driving an open pile section and a 

displacement pile section are not the same. Hence, a lower unit toe resistance, qqpp, 

should be used for calculating the toe resistance of open end pipe piles compared to 

a typical closed end condition. The value of the interior unit shaft resistance in an 

open end pipe pile is typically on the order of 1/3 to 1/2 the exterior unit shaft 

resistance, and is influenced by soil type, pile diameter, and pile shoe configuration. 

These factors will also influence the length of soil plug that may develop.

For open end pipe piles in cohesionless soils, Tomlinson (1994) recommends that 

the static soil resistance be calculated using a limiting value of 105 ksf for the unit 

toe resistance, regardless of the pile size or soil density. Tomlinson states that 

higher unit toe resistances do not develop, because yielding of the soil plug rather 

than geotechnical failure of the soil below the plug governs the resistance.

For open end pipe piles driven in stiff clays, Tomlinson (1994) recommends that the 

static soil resistance be calculated as follows when field measurements confirm a 

plug is formed and carried down with the pile:

 Eq. 7-90

Where:

Rn = nominal resistance (kips).

Ca = pile adhesion from Figure 7-17 (ksf). 

As = pile shaft surface area (ft2).

su = average undrained shear strength at the pile toe (ksf).

Ap = toe area of a plugged pile (ft2).

Static soil resistance calculations for open end pipe piles in cohesionless soils 

should be performed using the Paikowsky and Whitman equations. Toe resistance 

should be calculated using the Tomlinson limiting unit toe resistance of 105 ksf, once 

Meyerhof's limiting unit toe resistance, determined from Figure 7-15, exceeds 105 

ksf. For open end pipe piles less than 24 inches in diameter, and in predominantly 

cohesive soils, the Tomlinson equation should be used. Conversely, for moderate 

(24 to 36 inch O.D.) and large (36+ inch) diameter open ended pipe piles, plugging is 

less certain under static conditions and should be carefully analyzed.
408

The API method incorporated into the APILE program calculates the plugged and 

unplugged resistance of a open end pipe pile. The program also contains an 

automated routine that calculates whether a plugged or unplugged condition will 

develop. The results for a 30 inch open end pipe along with the analyzed soil profile 

are presented in Figure 7-82. 

Figure 7-82 Variation in expected static resistance from APILE API method based 

on plugged, unplugged, and automatic calculated plugging. 

The plugging phenomenon in H-piles can be equally difficult to analyze. However, 

the distance between flanges of an H-pile is smaller than the inside diameter of most 

open end pipe piles. Therefore, an H-pile is more likely to be plugged under static 

loading conditions where the “box” area of the pile toe is be used for static 

calculation of the toe resistance in cohesionless and cohesive soils. The toe 

resistance for H-piles driven to rock is usually governed by the pile structural 

strength, calculated based on the steel cross sectional area, and should not include 

the area of a soil plug, if any.
409

Figure 7-83 Plugging of open end pipe piles 

(after Paikowsky and Whitman 1990).

Figure 7-84 Plugging of H-piles (after Tomlinson 1994).
410

For H-piles in cohesionless soils, arching between the flanges may occur, and the 

"box" perimeter can be used for shaft resistance calculations. In most cohesive 

soils, the shaft resistance is calculated from the sum of the adhesion, Ca, along the 

exterior of the two flanges plus the undrained shear strength of the soil, su, times the 

surface area of the two remaining sides of the "box" due to soil-to-soil shear along 

these two faces. Figure 7-84 illustrates that the calculation of shaft resistance in Hpiles in stiff clays can still be problematic. Sheared clay lumps can develop above 

the plug zone, in which case the shaft resistance may only develop along the 

exterior surfaces of the flanges in the sheared lump zone.

The above discussions highlight the point that a higher degree of uncertainty often 

exists for static soil resistance calculations of open pile sections than for 

displacement piles. Soil plug formation and plug response is often different under 

static and dynamic loading. This can complicate nominal resistance evaluations of 

open pile sections with all dynamic methods (wave equation, dynamic testing, and 

dynamic formulas). Therefore, for large diameter open end pipe piles, greater than 

18 inches, or for H-piles designed to carry their load primarily in shaft resistance, a 

static load test is recommended for resistance verification.
411

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427

CHAPTER 8

STRUCTURAL ASPECTS AND LIMIT STATES

Driven piles resist a variety of vertical loads, moments and lateral loads. The 

previous chapter discussed a pile’s and pile group’s geotechnical resistance to these 

loads. This chapter considers the pile’s structural limit states which can govern the 

design in situations where both lateral and axial loads are applied. The analysis of 

the effects of vessel impact, wind, scour, and earthquake events in structure design 

often generates factored loads that can tax the pile’s structural resistance.

8.1 INTRODUCTION

This chapter addresses structural limit states, but is not meant to cover structural 

design in a comprehensive way. Please review AASHTO (2014) for specific 

structural design equations and requirements. A driven pile must satisfy stress and 

buckling checks under static loading conditions during its design life as well as under 

dynamic, driving induced loads. Therefore, the material strength limits are 

compared to:

1. The factored driving stresses.

2. The factored design loads during the pile’s design life.

In almost all cases, the highest stress levels occur in a pile during driving. High 

driving stresses are necessary to cause pile penetration. The pile must be stressed 

to overcome the nominal geotechnical resistance, plus any dynamic resistance 

forces, in order to be driven to support the pile design load. The high strain rate and 

temporary nature of the loading during pile driving allow a substantially higher driving 

stress limitation than for the static design case. Wave equation analyses can be 

used to predict driving stresses prior to installation. During installation, dynamic 

testing can be used to monitor driving stresses.

Factored loads are briefly considered in Chapter 2. The factored structural 

resistance is summarized in this chapter, in conformance with the AASHTO (2014) 

LRFD Bridge Design Specification. AASHTO (2014) provides a discussion on 
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concrete piles in Article 5, steel piles in Article 6, and timber piles in Article 8. 

Composite concrete filled steel pipe piles are also discussed.

The factored structural resistances for piles given in AASHTO (2014) are a function 

of the following variables:

1. Average section yield strength, such as:

a. Fy, the yield strength for steel piles or steel reinforcement.

b. f’c, the ultimate compression strength for concrete, typically at 28 days.

c. Wood crushing strengths.

2. Factors for incision, loading rate, submersion, cross sectional shape and analysis 

method for timber piles.

3. Resistance factor, φi , which allows for variations in loading condition, materials, 

construction dimensions, and method reliability.

4. Load factor, γi , to account for uncertainty in the factored service loads.

Driving stress limits, group layout, preliminary cap design, in-service stress limits, 

and buckling of piles are addressed in this chapter.

8.2 BASIC STRUCTURAL PROPERTIES OF DRIVEN PILES

8.2.1 Material Properties

Primary pile materials include steel, concrete and timber. The properties of these 

materials can significantly affect static nominal structural resistance and nominal 

driving resistance. Steel piles are generally produced to meet a minimum design 

yield stress which is used during load evaluations. Common steel Pipe pile and Hpile designations are given in Table 8-1 and Table 8-2 respectively, along with the 

minimum required yield stress. H-pile sections were traditionally available as Grade 

A-36 steel; however ASTM A572 requires new sections to be produced with a 

minimum yield stress of 50 ksi. H-piles of grade A-36 are therefore no longer 

manufactured. For all steel piles, the elastic modulus for static calculations, Es, is 

assumed to be 29,000 ksi, per AASHTO (2014) specifications.
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Table 8-1 Common Steel Pipe Pile Grades and Yield Stress 

Designation/Grade Yield Stress, Fy, ksi

ASTM A-252 Grade 2 35

ASTM A-252 Grade 3 45

ASTM A-252 Grade 3 (Mod) 50-80

Table 8-2 Common Steel H-pile Grades and Yield Stress

Designation/Grade Yield Stress, Fy, ksi 

A-36 36

ASTM A-572-50 50

ASTM A-572-60 60

For concrete piles, the elastic modulus/concrete strength is likewise a limiting 

variable. However, reinforcing steel can be added to increase tensile strength. The 

elastic modulus of concrete is typically estimated based on the compression 

strength. Without confirming test data, the AASHTO (2014) recommended method 

to determine the elastic modulus of concrete is presented in Equation 8-1. This 

approach assumes normal weight concrete with a unit weight of 145 pcf.

 Eq. 8-1 

Where: 

Ec = elastic modulus of concrete (ksi).

f’c = concrete compression strength at 28 days, unless otherwise 

specified (ksi).

Timber pile properties vary by tree species. Douglass-Fir and Southern Pine are the 

predominant timber pile species used across the United States. However, Red Oak 

and Red Pine are also used. Reference values for compression stress parallel to 

grain, Fco, and elastic modulus, Eo, are provided in AASHTO (2014) Article 8.4.1.4 

and are shown in Table 8-3. These values are typically modified by shape and size 

factors which are further discussed in Section 8.4.
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Table 8-3 Timber Pile Compression Stress and Elastic Modulus Reference 

Values (after AASHTO 2014)

Species Fco

ksi 

Eo 

ksi 

Pacific Coast Douglas-Fir 1.25 1500

Red Oak 1.10 1250

Red Pine 0.90 1280

Southern Pine 1.20 1500

8.2.2 Pile Section Definitions

Beyond material properties, the general shape and element dimensions affect 

modes of structural resistance. The relative thickness of steel generally limits the 

magnitude of force that can be reasonably transferred to the pile either during 

driving, or for structural considerations. Axial compression and flexural resistance is 

also derived from the pile shape, and as relatively large loads are transferred to steel 

and concrete piles, the element dimensions should be accounted for. Figure 8-1

and Figure 8-3 provide general shape dimensions for H-piles and pipe piles, 

respectively.

Figure 8-1 H-pile section dimensions.
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Figure 8-2 Table of H-pile section properties (modified from Skyline Steel 2015).

Figure 8-3 Pipe pile section dimensions.
432

The largest stresses in piles occur at the greatest distance from the neutral axis. 

This is at the outer edge(s) of a single pile and at the outer edge of the outermost 

piles of a group. Under combined axial and flexural loading, the pile area, the 

stiffness or moment of inertia, and distance from neutral axis to the edge of the 

section, must be defined to check maximum stress.

The moment of inertia, II (in4), for various common shapes may be computed with 

Equation 8-2 to 8-4. A reference should be made to the AISC (2011) shape tables 

or Figure 8-2 for H-piles, where tabulated structural property values about the strong 

and weak axis are provided. Sections should be selected from those commonly 

available from local suppliers. Otherwise, scheduling delays may result as 

manufacturers do not continuously hold all pile sizes in stock.

For solid, circular sections:

 Eq. 8-2 

For voided circular sections:

 Eq. 8-3 

For solid, square sections: 

 

Eq. 8-4 

Where: 

D = outside diameter (inches).

D1 = inner diameter, 2 r1= D-2 t, (inches).

t = pipe pile wall thickness (inches).

b = width/height of square (inches).

To compute moment induced stresses, it is also convenient to compute the elastic 

section modulus, S, as in Equation 8-5.

 

Eq. 8-5 
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Where: 

S = elastic section modulus (in3).

I = moment of inertia (in4). 

c = distance from centroid to outer edge (inches).

8.2.3 Effective Length and Buckling

From a structural view, piles act as columns and therefore under axial or moment 

loads, an effective length could be considered for simplified frame analyses. Pile 

end conditions are used to approximate an effective length factor, K, as shown in 

Figure 8-4, where the pile toe is generally assumed fixed for both translation and 

rotation. In the absence of sufficient bracing, (e.g. very soft soils, piles extended 

though water, large scour) the pile head may be affected by sidesway, and therefore 

a fixed rotation and translation condition may not apply. For these conditions, the 

design value of K should be applied. For example, if a pile extends into a pier cap, 

which is subject to severe scour, rotation is generally prevented by the pier cap 

mass and stiffness, while lateral movement may result from reduced soil confining 

pressure (reduced lateral bracing) in combination with existing loads. In this case, a 

fixed rotation and free translation condition may exist. In pile bents, depending on 

the foundation’s connection to the superstructure, the bent cap could allow rotation 

and translation perpendicular to the long axis of the bent cap, but free translation 

with fixed rotation along the long axis of the bent cap. To have a rotationally fixed 

pile top condition, Rollins and Stenlund (2010) observed that rather than defining a 

rule-of-thumb for minimum pile embedment length into the pile cap, the moment 

capacity of the pile cap to pile connection should be designed with pile embedment 

and cap reinforcement details such that the moment capacity of the connection 

exceeds the moment capacity of the pile.

Buckling is generally of concern when piles extend through water or air, or for 

liquefaction, where an absence of confining stress is clearly recognizable. Concern 

for buckling has also been expressed for piles extending through very soft soils or 

peat. However, pile buckling is a function of the confining stress (bracing) and 

applied load (Davisson 1963).

To characterize buckling in soft soils, a load test program was performed by the 

Bethlehem Steel Corporation which suggested that even soft soils provide adequate 

support (Bethlehem Steel Corporation 1970). One such H-pile in this study 

extended though 31 feet of water and 29 feet of organic silt where the pile sank 

under its own weight. A load of 200 tons produced a gross settlement of 0.63 inches 

while net settlement, after unloading, was measured at 0.02 inches. No pile buckling 
434

occurred. In addition, Coduto et al. (2016) suggests that, “even the softest soils 

provide enough lateral support to prevent underground buckling in piles subject only 

to axial loads, especially when a cap is present and provides rotational fixity to the 

pile top.” 

Figure 8-4 Effective length factors, K (after AASHTO 2014).

A more conservative approach to this issue is presented in Chance (2003) where the 

critical buckling load is determined through the use of computer software such as 

LPILE. For this method, a pile-soil model is generated and incremental loads are 

applied to evaluate the resulting deflection. This method may provide the design 

engineer with a deflected pile shape to assess buckling for a given factored load in 

lieu of using prescriptive minimum soil strength values to characterize an unbraced 

length.

8.3 STRUCTURAL CONSIDERATIONS AND RESISTANCE FACTORS

The limitations on nominal static design stresses for driven piles in various codes 

generally represent the static resistance which can be consistently developed with 

traditional driving equipment and methods.
435

The factored resistance must be greater than factored loads applied to the pile. The 

recommended AASHTO limits for factored pile design stresses will generally keep 

the driving stresses within recommended limits. Factored loads are covered in 

Article 3 of the AASHTO Specification (2014) while driving stress limits are 

presented in the respective pile material sections for concrete (Article 5), steel 

(Article 6), and timber (Article 8).

8.3.1 Depth to Fixity

The unbraced length, ll, or laterally unsupported length is defined by AASHTO 

(2014) as the distance between two braced points that resist buckling or distortion 

modes. For embedded piles, the unbraced length is considered for scour and pile 

stickup through air and/or water. For preliminary analysis, when lateral loads are 

applied, the effective length, K, for flexural or torsional resistance calculations is 

taken as the total unsupported length, plus an embedded depth to “fixity.” If a lateral 

pile analysis with p-y curves for soil-structure interaction has been performed as 

discussed in Chapter 7, the depth to fixity concept is unnecessary. Most software 

with lateral analysis also includes additional features to determine a pile’s buckling 

capacity given the soil model and a pile model with the expected stick-up above the 

ground level.

For preliminary calculations, however, the depth to fixity below the ground may be 

evaluated based on soil type and soil strength parameters as shown in Equation 8-6 

for clays and Equation 8-8 for sands. Table 8-4 contains the rate of increase in soil 

modulus for sands, nh, and should be used as applicable in the following depth to 

fixity estimates. 

For clays:

 

Eq. 8-6 

 Eq. 8-7 

For sands: 

Eq. 8-8 

Where:

df = depth to fixity below the ground (ft).

E = elastic modulus of pile material (ksi). 
436

Es = elastic modulus of clay soil (ksi). 

su = undrained shear strength of clay (ksf).

Iw = weak axis moment of inertia of pile (ft4).

nh = rate of increase of soil modulus with depth (Table 8-4) (ksi/ft).

Table 8-4 Rate of Increase of Soil Modulus with Depth for Sands (ksi/ft) 

(after AASHTO 2014)

Consistency Dry or Moist Submerged

Loose 0.417 0.208

Medium 1.110 0.556

Dense 2.780 1.390

8.3.2 Limiting Slenderness Ratio

Piles extending through air or water are unbraced over some length and therefore, 

for axial compression, the slenderness ratio should be checked during design. For 

non-composite steel piles, which are not fully embedded, slenderness ratio limits 

should be satisfied as follows: 

 Eq. 8-9 

Where:

K = effective length factor (Figure 8-4) (dimensionless).

l = unbraced length, or laterally unsupported length plus df (inches).

rs = minimum radius of gyration, (inches).

8.3.3 Resistance Factors

A discussion and step by step determination of the nominal structural resistance for 

timber, steel, and concrete piles is provided in the following sections. The AASHTO 

(2014) specifications form the basis of these respective sections. Following the 

Load and Resistance Factor Design (LRFD) approach, a resistance factor is applied 

to the calculated nominal structural resistance.

In practical terms, the imposed factored load must be less than or equal to the 

factored resistance. Chapter 2 provides a discussion on load combinations in which 

load factors are applied to respective load effects. The critical load combination is 
437

applied as axial, shear and moment loads on the structural member and will be 

described herein as follows:

Factored axial load:

( ) ∑ iii →= PQQn u γ Eq. 8-10

Factored moment load:

( ) ∑ iii →= MQQn u γ Eq. 8-11

Factored shear load:

( ) ∑ iii →= VQQn u γ Eq. 8-12

The nominal structural resistance is specific to pile properties such as material, size, 

and shape. A separate discussion based on pile material is provided in Section 8.4 

through Section 8.7. On the resistance side, the factored structural resistance will 

be described as shown below.

Factored resistance in axial compression:

∑ φφ == PPR rncii Eq. 8-13

Factored resistance in flexure:

∑ φφ == MMR rnfii Eq. 8-14

Factored resistance in shear:

∑ φφ == VVR rnvii Eq. 8-15

AASHTO (2014) recommended resistance factors for driving and structural 

resistance are provided in Table 8-5 and Table 8-6, respectively. As an example, for 

a steel H-pile with good driving conditions and with the use of a pile tip (e.g. pile 

point or shoe), φda = 1.0 would be used in Equation 8-33 (Section 8.5.1) along with 

the material yield stress to determine the nominal driving stress limits. For nominal 

structural resistance, φc = 0.60, would be entered into Equation 8-34 (Section 8.5.2). 

A similar procedure is therefore used for alternative pile types, driving conditions, 

and loading cases.
438

Table 8-5 Resistance Factors During Pile Driving

Resistance Determination Method Pile Drivability Resistance Factor

Steel Piles (AASHTO 6.5.4.2) φda = 1.0

Concrete Piles (AASHTO 5.5.4.2) φda = 1.0

Timber Piles (AASHTO 8.5.4.2) φda = 1.15

Table 8-6 Structural Limit Resistance Factors for Piles in Compression

 
439

8.4 TIMBER PILES

Timber piles are typically round, tapered, and used for modest loading levels. 

Several species of timber are suitable for use as piles, however Douglas Fir and 

Southern Pine are by far the most common types used in the United States. 

Equations to assess maximum driving stresses (Section 8.4.1) or axial and flexural 

stresses (Section 8.4.2) will be subsequently described, while reference design 

values shown in Table 8-7 are factored into these calculations. For the load cases 

described herein, the reference value is modified with an adjustment factor specific 

to the loading case and pile properties.

Table 8-7 Reference Design Values for Timber Piles, ksi (after AASHTO 2014)

Species Fco Fbo Fvo Eo

Pacific Coast Douglas-Fir1 1.25 2.45 0.115 1500

Red Oak2 1.10 2.45 0.135 1250

Red Pine3 0.90 1.90 0.085 1280

Southern Pine4 1.20 2.40 0.110 1500 1 Use Douglas Fir-Latch reference values for connection design. 

 2 Northern and Southern Red Oak. 

 3 United States grown Red Pine. Use Northern Pine reference values for connection design.

 4 Reference strengths apply to Lobololly, Longleaf, Shortleaf and Slash Pine. 

Where:

Fco = reference value for compression stress parallel to grain (ksi).

Fbo = reference value for strength in flexure (ksi).

Fvo = reference value for strength in shear (ksi).

Eo = reference value elastic modulus (ksi).

The time effect factor, Cλ, shown in Table 8-8 is utilized in consideration of specific 

limit states. AASHTO (2014) commentary provides that the Strength I condition be 

used when the cumulative duration of a bridge live load is two months, while the 

Strength II condition is used for cumulative live loading less than two months. 

Furthermore for permanent loads, the Strength IV limit state should be used.

Table 8-8 Time Effect Factors (after AASHTO 2014)

Limit State Cλ

Strength I 0.8

Strength II 1.0

Strength III 1.0

Strength IV 0.6

Extreme Event I 1.0
440

A size reduction factor, CF, is applied to timber piles in flexure; however this factor is 

only applied if the critical section in bending has a diameter greater than 12 inches 

(i.e. area of 144 in2). This would be the sawn lumber beam equivalent of a depth of 

12 inches (i.e. area of 144 in2). If the critical section area is less than 144 in2, the 

shape reduction factor is set equal to 1.0. Conversely, if the critical section area is 

greater than 144 in2, Equation 8-16 should be used to determine the size reduction 

factor. Considering the AASHTO (2014) criteria, “the equation should be entered 

with a b equal to the depth of a square beam having the same cross-sectional area 

as that of a round member” (Showalter 2012).

 Eq. 8-16

Where:

CF = size factor.

b = depth of beam or width of dimension lumber.

The wet service factor, CM, provides adjustment based upon moisture content and 

although not all timber piles are submerged, and assumption of wet use conditions 

may be made. A wet service factor equal to 1.0 is therefore recommended. 

Several adjustment factors provided in the AASHTO (2014) specifications do not 

apply to timber piles such as the volume factor Cv, incising factor Ci, deck factor Cd, 

flat-use factor Cfu. Since AASHTO (2014) design specifications do not provide 

specific guidance on these values for timber piles, the values are assumed as 1.0 in 

this manual to prevent unwarranted increases or reductions in the structural 

resistance limits.

8.4.1 Driving Stresses 

Timber piles should be installed to limit excessive pile stress development that may 

result in shearing or brooming near the pile top. Additional stresses may develop 

along the pile as well, especially as the tapered section decreases near the pile toe. 

AASHTO (2014) specifications limit maximum compression and tension driving 

stresses as Equation 8-17.

 Eq. 8-17
441

Where:

σdr = driving stress (ksi).

φda = resistance factor during driving (1.15 for timber piles, Table 8-5).

Fco = reference value for compression stress parallel to grain (Table 8-7).

8.4.2 Structural Resistance

AASHTO (2014) summarizes axial and flexural loading resistance limits in Article 

8.6, 8.8 and 8.10. The following sections provide design checks to determine 

structural strength limits, where Table 8-7 -provides reference values, Table 8-8

contains time effect factors, and Equation 8-16 may be used to determine a size 

factor. The equations below apply to all timber elements, and therefore several 

adjustment factors introduced in the AASHTO (2014) specification do not specifically 

pertain to piles and are assumed to equal 1.0.

8.4.2.1 Axial Compression Parallel to Grain 

For axial compression loading, the factored structural limit state is taken as follows:

= φPP nr Eq. 8-18 

Where:

Pr = factored resistance in axial compression (kips).

φ = resistance factor (0.9 for timber pile compression parallel to grain).

Pn = nominal resistance in axial compression (kips).

STEP BY STEP PROCEDURE FOR: “NOMINAL COMPRESSION RESISTANCE”

STEP 1 Determine the adjusted resistance in compression,Fc. 

The nominal resistance for axial compression should be evaluated using Equation 

8--19 and 8-20 which are provided in AASHTO (2014) Article 8.4.

C = KFCO iFM

CCCCCFF λ Eq. 8-19 

In which:

φ

5.2 CKF = Eq. 8-20 
442

Where:

Fc = adjusted resistance in compression (ksi).

Fco = reference resistance in compression (Table 8-7) (ksi).

CKF = format conversion factor where φc = 0.9.

CM = wet service factor (1.0 for piles).

CF = size factor (1.0 for or Equation 8-16 for piles).

Ci = incising factor (Not Applicable for piles-1.0).

Cλ = time effect factor (Table 8-8).

STEP 2 Determine the nominal resistance in axial compression, Pn

Following adjustment of the reference resistance in compression, the nominal 

resistance in axial compression is calculated from Equation 8-21. If the piles are 

sufficiently braced, the column stability factor, Cp, may be taken as 1.0. However, 

Equation 8-22 should be used to determine the column stability factor, Cp, if 

sufficient bracing does not exist.

= CAFP pgcn Eq. 8-21 

In which:

0.1

2

1

2

1 2

 ≤− 

 



 + − + = c

B

c

B

c

B Cp Eq. 8-22 

≤= 0.1

c

cE

F

F

B Eq. 8-23 

2

2

e

cE

cE L

K Eb

F = Eq. 8-24 

= CEE to Eq. 8-25 

Where:

Fc = adjusted resistance in compression parallel to grain (ksi).

Pn = nominal resistance in axial compression (kips).

Ag = gross cross-sectional area (in2). 

Cp = column stability factor.

CM = wet service factor (1.0 for piles).

Ci = incising factor (Not Applicable for piles-1.0).

c = round timber pile factor (0.85).
443

 KcE = round timber pile factor (0.76).

b = depth or width (inches).

Le = effective length taken as KL (Figure 8-4, inches).

Eo = reference modulus of elasticity (ksi) (Table 8-7).

8.4.2.2 Flexure

For flexure, the factored structural limit state is taken as:

 Eq. 8-26 

Where:

Mr = factored flexural resistance (kip-in).

φ = resistance factor (0.85 for timber pile flexure).

Mn = nominal flexural resistance (kip-in).

STEP BY STEP PROCEDURE FOR: “NOMINAL FLEXURAL RESISTANCE”

STEP 1 Determine the adjusted resistance in flexure, FFbb. 

A round timber pile is assumed in this manual for flexure. If alternative pile cross 

sections are used, consideration should be given to the new shape. To determine 

the adjusted resistance in flexure, the following equations should be applied.

 Eq. 8-27 

In which:

 

Eq. 8-28 

Where:

FFbbvv = reference resistance in flexure (Table 8-7) (kips).

CCKKKK = format conversion factor where φf = 0.85.

CCMM = wet service factor (1.0 for piles).

CCKK = size factor (1.0 or Equation 8-16 for piles).

CCffuu = flat use factor (Not Applicable for piles-1.0).

CCii = incising factor (Not Applicable for piles-1.0).

CCnn = deck factor (Not Applicable for piles-1.0).

CCλλ = time effect factor (Table 8-8).
444

STEP 2 Determine the nominal flexural resistance, Mn. 

Following adjustment of the reference resistance in flexure, the nominal flexural 

resistance is calculated as follows.

= bn SFM Eq. 8-29

Where:

Mn = nominal flexural resistance (kip-in).

S = elastic section modulus (in3).

Fb = adjusted resistance in flexure (Table 8-7) (ksi).

8.4.2.3 Combined Flexure and Axial Compression

For combined flexure and compression, the factored structural limit state must 

satisfy checks for beam rupture and lateral buckling as shown in Equation 8-30. If 

the applied loads are not sufficiently resisted by the pile section, an alternative pile 

design is required.

0.1

1

2

≤















 −

+ 















cE g

u

r

u

r

u

AF

P

M

M

P

P Eq. 8-30 

In which:

2

2

e

cE

cE L

K Eb

F = Eq. 8-31 

= CCEE iMo Eq. 8-32 

Where:

Pr = factored resistance in axial compression (kips) (Equation 8-18).

Pu = factored axial load (kips) (Equation 8-10). 

Mr = factored resistance in flexure (kip-in) (Equation 8-26).

Mu = factored moment load (kip-in) (Equation 8-11).

Ag = gross cross-sectional area (in2).

CM = wet service factor (N/A -1.0 for piles).

Ci = incising factor (N/A -1.0 for piles).

 KcE = round timber pile factor (0.76). 

b = pile width or diameter (inches).

Le = effective length taken as KL (Figure 8-4, inches).

Eo = reference modulus of elasticity (ksi) (Table 8-7).
445

8.5 STEEL PILES

8.5.1 Driving Stresses 

Steel piles can handle higher stresses than concrete or timber during driving, and 

the limit is governed the steel yield strength. Several grades of steel are routinely 

produced and higher grades may be specified if warranted. Table 8-1 provides an 

overview of typical steel grades and their respective yield strength. Pile driving does 

not typically generate sufficiently high tensile stresses to yield steel piles, therefore a 

stress limit is only required in compression. For steel piles, AASHTO (2014) 

specifications limit nominal compression driving stresses as follows:

 Eq. 8-33 

Where:

σdr = driving stress limit (ksi).

φda = resistance factor during driving (1.0 for steel piles, Table 8-5).

Fy = yield stress of steel (ksi) (Table 8-1).

8.5.2 Structural Resistance

8.5.2.1 Axial Compression

For axial compression loads, the factored structural limit state is taken as:

 Eq. 8-34 

Where:

Pr = factored resistance in axial compression (kips).

φc = resistance factor (Table 8-6).

Pn = nominal resistance in axial compression (kips).

To determine the nominal resistance in axial compression, pile strength and buckling 

failure should be considered. The following step by step procedure should be used 

to calculate the nominal resistance.
446

STEP BY STEP PROCEDURE FOR: “NOMINAL COMPRESSION RESISTANCE”

STEP 1 Determine the equivalent nominal yield resistance, Po. 

The equivalent nominal yield resistance, Po, is a function of the material yield stress, 

cross sectional area and slenderness reduction factor, if applicable. For non-slender 

piles in compression, the slenderness reduction factor, Q, is taken as 1.0. However 

for slender piles, the full nominal yield strength under uniform axial compression is 

limited by local buckling. This reduction factor is governed by section buildup, pile 

dimensions and material properties, therefore, a further discussion of slender

members and direction for calculating Q may be found in AASHTO (2014) Article 

6.9.4.2.2.

Po = QF Agy Eq. 8-35

Where:

Ag = gross cross-sectional area (in2

).

Po = equivalent nominal axial yield resistance (kips).

 Fy = yield stress of steel (Table 8-1) (ksi).

Q = slender element reduction factor (dimensionless).

To satisfy the slender element requirement for local buckling, Equation 8-36 is used 

for H-piles and Equation 8-38 is used for unfilled pipe piles. 

y

c st

f

f

F

Ek

t

b

64.0

2

≤ Eq. 8-36

And:

≤≤ 76.035.0 c k

In which:

ww

c

td

k 4 = Eq. 8-37

Where:

bf = flange width (inches).

tf = flange thickness (inches).

Fy = yield stress of steel (Table 8-1) (ksi).

Est = elastic modulus of steel (ksi).

dw = web depth (inches).

tw = web thickness (inches).
447

y

st

F

E

t

D

≤ 11.0 Eq. 8-38

Where:

D = diameter of pipe (inches).

t = wall thickness (inches).

Fy = yield stress of steel (Table 8-1) (ksi).

Est = elastic modulus of steel (ksi).

STEP 2 Determine the elastic critical buckling resistance, Pe. 

In determination of the nominal resistance in axial compression, buckling may occur 

with a lack of sufficient bracing. This topic is discussed further in Section 8.2.3. 

AASHTO (2014) requires both flexural and torsional modes of buckling be checked if 

applicable. For fully embedded piles, the flexural buckling mode will be used. 

However, when the pile extends through water or air, doubly symmetric open section 

members (e.g., H-piles) must be evaluated for torsional buckling as well. The critical 

failure mode is the lesser buckling resistance, and is employed to define the nominal 

resistance in axial compression. 

Flexural buckling:

2

2





 





=

s

st g

e

r

Kl

AE

P

π Eq. 8-39 

Where:

Pe = elastic critical buckling resistance (kips).

Est = elastic modulus of steel (ksi).

Ag = gross cross-sectional area (in2

).

K = effective length in the plane of buckling (Figure 8-4) (dimensionless).

l = unbraced length in the plane of buckling (Section 8.3) (inches).

rs = radius of gyration about axis normal to plane of buckling (inches).

Torsional buckling:

( ) yx

g

zz

st w

e II

A

GJ

lK

CE P

+





 



 = + 2

2 π Eq. 8-40
448

In which:

4

2 hI

C y

w = Eq. 8-41 

EG st = 385.0 Eq. 8-42 

Where:

Pe = elastic critical buckling resistance (kips).

Est = elastic modulus of steel (ksi).

Cw = warping torsional constant (doubly symmetric open sections) (in6

).

Kz = effective length for torsional buckling (Figure 8-4) (dimensionless). 

lz = unbraced length for torsional buckling (inches).

G = shear modulus (ksi). 

J = St. Venant torsional constant (in4).

Ag = gross cross-sectional area (in2

).

 Ix, Iy = moments of inertia about the major and minor principal axes of cross 

section, respectively (in4

).

h = distance between flange and centroids (inches).

STEP 3 Determine the nominal resistance in axial compression, Pn. 

With the above resistances defined, the nominal resistance for axial compression 

may be evaluated using the following equations, which are provided in AASHTO 

(2014) Article 6.9.4.1.

If ≥ 44.0

o

e

P

P

: 

e

o

P

P

= PP on 658.0 Eq. 8-43

If < 44.0

o

e

P

P

: 

n PP e = 877.0 Eq. 8-44

Where:

Pn = nominal resistance in axial compression (kips).

Po = equivalent nominal yield resistance (Equation 8-35) (kips).

Pe = elastic critical buckling resistance (Equation 8-39 or 8-40) (kips).
449

8.5.2.2 Flexure

The factored resistance in flexure is computed as follows

= φ MM nfr Eq. 8-45

Where:

Mr = factored resistance in flexure (kip-in).

φf = resistance factor (Table 8-6).

Mn = nominal resistance in flexure (kip-in).

The nominal flexural resistance is a function of pile shape as well as general pile 

properties. Steel piles are primarily H-piles or pipe piles. Therefore the step by step 

procedure that follows will consider only these two steel pile types. If alternative 

sections are used, the engineer is referred to Article 6.12.2.2 of the AASHTO (2014) 

specifications. Steel H-piles and I-sections are treated equally for flexural 

resistance. Hence, part A of this procedure applies to both steel H-piles and 

miscellaneous I sections.

STEP BY STEP PROCEDURE FOR: “NOMINAL FLEXURAL RESISTANCE”

A. Steel H-Sections.

STEP 1 Check flange slenderness ratio and limiting slenderness.

f

f

f t

b

2

λ = Eq. 8-46 

yf

st

pf F

E λ = 38.0 Eq. 8-47 

yf

st

rf F

E λ = 83.0 Eq. 8-48 

Where:

λf = slenderness ratio for flange.

λpf = limiting slenderness ratio for a compact flange.

λrf = limiting slenderness ratio for a non-compact flange.
450

bf = flange width (inch).

tf = flange thickness (inch).

Est = elastic modulus of steel (ksi).

Fyf = minimum yield strength of lower strength flange (ksi).

STEP 2 Determine the nominal flexural resistance.

To determine the nominal flexural resistance, the above slenderness definitions 

should first be resolved. These functions serve to define the limiting flexural 

resistance. In the case where the limiting slenderness ratio of a compact flange is 

greater than the slenderness ratio, the plastic moment about the weak axis will limit 

resistance. For H-piles, Eq. 8-49 can be used. Conversely, Eq. 8-51 should be 

used when the slenderness ratio is greater than the limiting slenderness ratio of a 

compact flange.

If : 

 Eq. 8-49 

In which, for HP-sections about weak axis:

 Eq. 8-50 

If , the nominal flexural resistance about the weak axis is:

 

Eq. 8-51 

Where:

Mn = nominal resistance in flexure (kip-in).

Mp = plastic moment about the weak axis (kip-in).

Sy = elastic section modulus about weak axis (in3

).

Zy = plastic section modulus about weak axis (in3).

λf = slenderness ratio for flange (Equation 8-46, dimensionless).

λpf = limiting slenderness ratio for a compact flange (Equation 8-47, 

dimensionless).
451

Est = elastic modulus of steel (ksi).

Fy = yield stress of steel (ksi).

Fyf = minimum yield strength of lower strength flange (ksi).

B. Steel Pipe Piles.

STEP 1 Check diameter to thickness ratio. 

If the diameter to thickness ratio is sufficiently large, local buckling limits flexural 

resistance. To determine whether the plastic moment or local buckling will govern 

flexural resistance, Equation 8-52 should be applied. If Equation 8-52 is satisfied, 

the plastic moment will yield the steel pile and Step 2a should follow. Conversely, 

local buckling will limit flexural resistance if Equation 8-2 is not satisfied, and 

therefore Step 2b should follow.

 

Eq. 8-52

Where:

D = outside diameter of pipe (inch).

t = pipe thickness (inch).

Est = elastic modulus of steel (ksi).

Fy = yield strength of steel (Table 8-1) (ksi).

STEP 2a Determine nominal flexural resistance by plastic moment. 

The nominal flexural resistance can be taken as follows: 

 Eq. 8-53

Where:

Mn = nominal resistance in flexure (kip-in).

Mp = plastic moment (kip-in).

Zy = plastic section modulus about weak axis (in3

).

Fy = yield strength of steel (Table 8-1) (ksi).
452

STEP 2b Determine nominal flexural resistance by local buckling. 

Where local buckling will limit the nominal resistance in flexure, the following checks 

apply.

If 

y

st

F

E

t

D

≤ 31.0 : 

yy

st

n SF

t

D

E

M

























= +

021.0 Eq. 8-54

If 

y

st

F

E

t

D

> 31.0 : 

n cr y = SfM Eq. 8-55

In which:

t

D

E f st

cr

33.0 = Eq. 8-56

Where:

D = outside diameter of pipe (inch).

t = pipe thickness (inch).

Es = elastic modulus of steel (ksi).

Fy = yield strength of steel (Table 8-1) (ksi).

Mn = nominal flexural resistance (kip-in).

Sy = elastic section modulus about weak axis (in3

).

fcr = elastic local buckling stress (ksi).

8.5.2.3 Combined Axial Compression and Flexure

Combined axial compression and flexure checks are only applied to pile groups with 

vertical piles. At this time, AASHTO (2014) does not have a recommendation to 

include battered piles. For combined compression and flexure of vertical piles, 

AASHTO (2014) requires the factored structural limit state to satisfy the following 

limit state checks. 

If < 2.0

r

u

P

P : 
453

0.1

0.2

≤















++

ry

uy

rx

ux

r

u

M

M

M

M

P

P Eq. 8-57 

If ≥ 2.0

r

u

P

P : 

0.1

0.9

0.8 ≤















+ +

ry

uy

rx

ux

r

u

M

M

M

M

P

P Eq. 8-58 

Where:

Pu = factored load in axial compression (kips).

Pr = factored resistance in axial compression (kips) (Section 8.5.2.1). 

Mux = factored moment about x-axis (kip-ft).

Mrx = factored resistance in flexure about x-axis (kip-ft) (Section 8.5.2.2).

Muy = factored moment about y-axis (kip-ft).

Mry = factored resistance in flexure about y-axis (kip-ft) (Section 8.5.2.2).

8.5.2.4 Shear

Piles used for bridge foundations are generally not also used to resist high shear 

loads as significant lateral pile deflections may negatively impact bridge 

serviceability. For shear loads, the factored structural limit state is taken as: 

= φ VV nvr Eq. 8-59 

Where:

Vr = factored resistance in shear (kips).

φv = resistance factor (Table 8-6).

Vn = nominal resistance in shear (kips).

A straightforward calculation of the nominal structural resistance in shear is shown in 

Equation 8-60 and Equation 8-61 for an H-pile section. Reference should be made 

to the AASHTO (2014) specifications for additional design requirements concerning 

piles subject to significant shear loads or for alternative non-composite pile sections.

n = CVV p Eq. 8-60 

p yw ww = 58.0 tdFV Eq. 8-61 
454

Where:

Vn = nominal resistance in shear (kips).

C = ratio of shear-buckling resistance to shear yield strength, typically 1.0 

for H-piles.

Vp = plastic shear force (kips).

Fyw = yield strength of web (Table 8-1) (ksi).

dw = web depth of pile section (inch).

tw = web thickness of pile section (inch).

8.5.3 Example Calculations for H Pile Structural Resistance.

The following example provides calculations for an HP 14x117 which will support an 

integral abutment. This H-pile section is produced with a yield stress, Fy, of 50 ksi, 

an elastic modulus, Est, of 29,000 ksi while the radius of gyration in the weak 

direction, rs, is 3.59 inches (Figure 8-2). Based upon project specific conditions, an 

unbraced length, l, of 120 inches is assumed for scour. In addition, rotation is fixed 

while translation is free at the pile head, whereas both rotation and translation are 

fixed for at the pile toe, thus an effective length factor of 1.2 results (Figure 8-4).

Under the given conditions, the pile section easily satisfies main member 

slenderness limits (kl/r<120). However a further inspection of local buckling under 

compression loads is required. First the buckling coefficient is determined.

952.0

805.02.14

4 4 = = = td inches inches

k

ww

c [Eq. 8-37] 

But, kc must satisfy:

kc ≤≤ 76.035.0

Therefore:

kc = 0.76

After calculating the buckling coefficient, the following condition is checked.

y

c st

f

f

F

Ek

t

b

64.0

2

≤ [Eq. 8-36]

ksi

ksi

inches

inches

50

000,29*76.0 64.0

805.0*2

9.14 ≤
455

 

Based upon Equation 8-36, the pile section is a non-slender element, and therefore 

the slenderness reduction factor Q is set equal to 1.0. The equivalent nominal yield 

resistance is determined using Equation 8-35.

 [Eq. 8-35]

 

Po = 1720 kips

Next, elastic critical buckling resistance in the section is determined.

 

[Eq. 8-39]

 

Pe = 6120 kips

From Step 3 of Section 8.5.2.1, the nominal resistance in axial compression, Pn, is 

determined by applying either Equation 8-43 or 8-44. For this example, the ratio of 

elastic critical buckling resistance, Pe, to yield resistance, Po, satisfies criteria for 

Equation 8-43 and is therefore shown in the following calculations.

If : 

 

Therefore:

 [Eq. 8-43]

 

Pn = 1529 kips
456

After calculation of the nominal resistance in axial compression, a resistance factor 

can be applied to determine the factored structural resistance in axial compression. 

Because good driving conditions were encountered and pile toe protection was not 

necessary, a resistance factor, φc = 0.60, is used (Table 8-6). This factor is applied 

when only axial compression is considered. When using the combined axial and 

flexure interaction equations as shown in Section 8.5.2.3, φc = 0.70 is used.

= φ PP ncr [Eq. 8-34] 

P kips r = 529,1*60.0

Pr = 917 kips 

Continuing with the example, the factored resistance in flexure is determined. To 

begin, the flange slenderness is inspected for the HP 14x117 pile section. In 

addition, compact and non-compact flange slenderness ratios are calculated.

Flange slenderness ratio:

f

f

f t

b

2

λ = [Eq. 8-46]

inches

inches

f 805.0*2

9.14

λ =

λf = 9.25

Limiting slenderness ratio for a compact flange: 

yf

st

pf F

E λ = 38.0 [Eq. 8-47]

ksi

ksi

pf 50

000,29 λ = 38.0

λpf = 9.25

Limiting slenderness ratio for a non-compact flange:

yf

st

rf F

E λ = 83.0 [Eq. 8-48]
457

ksi

ksi

rf 50

000,29 λ = 83.0

λrf = 19.98

After determining the slenderness ratio functions, a comparison is then made. For 

this particular pile, Equation 8-48 will be used to define the nominal flexural 

resistance. Reference should be made to Figure 8-2 for H-pile section properties 

such as the elastic and plastic section moduli.

If fpf ≤< λλλ rf : 

≤< 98.1925.915.9

Therefore:

yf y

yf

st

pff

y

y

n Zf

F

Z E

S

M

























































−















 −−=

45.0

11

λλ [Eq. 8-51] 

3

3

3

4.91*50

50

000,29 45.0

15.925.9

4.91

5.59 11 ksi in

ksi

in ksi

in M n

































































−















 −−=

Mn = 4555 kip-in

Following the calculation of nominal resistance in flexure, a resistance factor can be 

applied to determine the factored structural resistance in flexure. The AASHTO 

recommended resistance factor for flexure is, φf = 1.0, is used (Table 8-6).

= φ MM nfr [Eq. 8-45] 

M r = 4555*0.1 kip − in

Mr = 4555 kip-in
458

Finally, the nominal resistance in shear of the pile section is determined using 

Equations 8-60 and 8-61. For the HP 14x117 pile section, the web depth is 14.2 

inches and the web thickness is 0.805 inches (Figure 8-2). The plastic shear force is 

calculated using Equation 8-61.

 [Eq. 8-61]

 

Vp = 332 kips

The nominal structural resistance in shear is calculated using Equation 8-60.

 [Eq. 8-60]

 

Vn = 332 kips

By multiplying the nominal structural resistance in shear by the AASHTO (2014) 

recommended resistance factor, the factored structural resistance in shear is 

determined.

 [Eq. 8-59] 

 

Vr = 332 kips
459

8.6 CONCRETE PILES

8.6.1 Driving Stresses 

Concrete by nature is strong in compression while tensile strength is mainly derived 

from reinforcing steel and prestressed longitudinal rebar. The additional tensile 

strength from prestressing allows for somewhat larger tensile stresses to develop 

and is therefore featured in nearly all concrete piles manufactured in the United 

States today. Unlike timber and steel piles, concrete piles are susceptible to 

damage in both compression and tension under normal driving conditions. The use 

of a pile cushion (Chapter 15) slows the impact velocity and thus, reduces dynamic 

forces during driving. Pile cushions are therefore employed when driving piles to 

lessen the probability of damage.

Driving stress limits are applied to the gross concrete area. In severe corrosive 

environments, the nominal tension stress is limited to the effective prestress after 

losses. AASHTO (2014) specifications limit the nominal driving stresses in 

compression and tension for concrete piles as follows.

For compression:

( ) dr da pec = φσ '85.0 − ff Eq. 8-62 

For tension:

( ) dr da pec = φσ '095.0 + ff Eq. 8-63

Where:

σdr = driving stress (ksi).

φda = resistance factor during driving (1.0 for concrete piles, Table 8-5).

f’c = concrete compression strength at 28 days, unless otherwise 

specified (ksi).

fpe = effective prestressing stress in concrete (ksi). 

In Chapter 10 of the AASHTO (2014) specifications, the effective prestressing stress 

in the concrete is designated as fpe, as shown in Equation 8-62 and 8-63. This value 

should typically be on the order of 0.5 to 1.0 ksi for most piles on highway projects, 

and should certainly be less than f’c. 
460

8.6.2 Structural Resistance

8.6.2.1 Axial Compression

For axial compression loading, the factored structural limit state is taken as:

= φ PP ncr Eq. 8-64

Where:

Pr = factored resistance in axial compression (kips).

φc = resistance factor (Table 8-6).

Pn = nominal resistance in axial compression (kips).

To determine the nominal resistance in axial compression, a straightforward 

calculation is performed considering either spiral or tie reinforcement. As mentioned 

in the ASSHTO (2014) commentary, reduction factors are placed on the respective 

equations to account for unintended eccentricity. Further details on axial resistance 

of concrete piles can be found in Article 5.7.4.4 of AASHTO (2014).

For members with spiral reinforcement:

[ ( ) ( )] Pn gc str ps yr str peps EfAAFAAAf st cu = '85.085.0 −−+−− ε Eq. 8-65 

For members with tie reinforcement: 

[ ( ) ( )] Pn gc str ps yr str peps EfAAFAAAf st cu = '85.080.0 −−+−− ε Eq. 8-66 

Where:

Pn = nominal resistance in axial compression (kips).

f’c = concrete compression strength at 28 days, unless otherwise 

specified (ksi).

fpe = effective stress in the prestressing steel after losses (ksi). 

Fyr = yield stress of reinforcing steel (ksi). 

Ag = gross cross-sectional area (in2).

Astr = cross sectional area of longitudinal reinforcement (in2).

Aps = cross sectional area of prestressing steel (in2).

Est = elastic modulus of prestressing steel (in2

).

εcu = failure strain of concrete in compression (in/in).
461

8.6.2.2 Biaxial Flexure

For biaxial flexure, the factored structural limit state is taken as:

= φ MM nfr Eq. 8-67 

Where:

Mr = factored resistance in flexure (kip-ft).

φf = resistance factor (Table 8-6).

Mn = nominal resistance in flexure (kip-ft).

Biaxial flexural resistance must satisfy the following checks. Additional information 

may be found in Article 5.7.4.5 of the AASHTO (2014) specifications.

If u AfP gc ≥ φ '10.0 : 

0.1 1111≤−+=

rxy ryrx φPPPP o

Eq. 8-68 

In which:

( ) ( ) n gc str ps yr str peps EfAAFAAAfP st cu = '85.0 −−+−− ε Eq. 8-69 

If u AfP gc < φ '10.0 : 

≤ 0.1 















+

ry

uy

rx

ux

M

M

M

M Eq. 8-70 

Where:

Pu = factored axial load.

Prx = factored resistance in axial compression determined on the basis that 

only eccentricity, ey, is present (kips).

Pry = factored resistance in axial compression determined on the basis that 

only eccentricity, ex, is present (kips).

Prxy = factored resistance axial compression with biaxial flexure (kips).

Mux = factored moment about x-axis (kip-in).

Mrx = factored resistance in flexure about x-axis (kip-in) (Section 8.5.2.3).

Muy = factored moment about y-axis (kip-in).

Mry = factored resistance in flexure about y-axis (kip-in) (Section 8.5.2.3).

φc = resistance factor for axial compression (Table 8-6).
462

f’c = concrete compression strength at 28 days, unless otherwise 

specified (ksi).

fpe = effective stress in the prestressing steel after losses (ksi). 

Fyr = yield stress of reinforcing steel (ksi). 

Ag = gross cross-sectional area (in2).

Astr = cross sectional area of longitudinal reinforcing steel (in2).

Aps = cross sectional area of prestressing steel (in2).

Ep = elastic modulus of prestressing steel (in2

).

εcu = failure strain of concrete in compression (in2).

The analysis of the prestressed concrete section response to a combination of an 

axial load and two orthogonal moments is complex. A successful and practical 

approach to the analysis of the pile cross section is offered by the FB-Pier Program. 

The concrete and the prestressing steel stress strain relationships are assumed. 

For concrete, the FB-Pier program assumes a maximum concrete strength of 

0.85ff′cc to include loading time effects on the concrete strength and all points on the 

stress strain curve are reduced to 85 percent of the short time values.

Bi-axial interaction diagrams are determined for each of an increasing set of axial 

loads up to the maximum axial strength condition. An illustration of one of these 

interaction diagrams for a particular axial load is shown in Figure 8-5. These 

diagrams are determined for the entire range of axial loads up to the axial failure 

case. With increasing axial load the maximum moment strength becomes smaller. 

A three dimensional interaction diagram can then be constructed with the axial load 

on the vertical axis and a particular interaction diagram at each level of axial load. 

Imagine a stack of these interaction diagrams. Thus, a three dimensional failure 

surface is defined. The equation of the failure surface can be generated by fitting a 

surface through the interaction diagrams at each level.
463

Figure 8-5 Moment interaction diagram.

When the necessary failure surfaces are available, the analysis at a particular load 

level can be checked by examining whether the vector of the forces on the section 

(axial, Mx and My) falls within or outside the failure envelope. The deformations 

associated with the three applied forces make it possible to determine the 

displacements associated with the various load levels. This elegant and powerful 
464

analysis algorithm produces excellent results. Well-designed graphics make it 

possible for the foundation specialist to easily evaluate the results.

The analysis has been discussed for prestressed concrete piles and they are 

probably the most challenging to deal with. FB-Pier can also analyze steel piles and 

concrete filled pipes using the same concepts described above.

8.7 COMPOSITE PILES

Composite piles for structural applications shall be defined as concrete filled steel 

pipe piles. Guidance for other composite pile types such as a concrete pile with an 

H-pile stinger is not provided for in AASHTO (2014) specifications. Therefore 

structural resistances over the length of respective pile materials should be 

evaluated considering the primary section material.

8.7.1 Driving Stress

Concrete filled steel pipe piles are driven unfilled. Therefore, driving stresses for the 

non-composite steel section should be used and reference should be made to 

Section 8.5.1, Driving Stresses: Steel Piles.

8.7.2 Structural Resistance

8.7.2.1 Axial Compression

The behavior of composite sections is somewhat different than non-composite 

members and is therefore evaluated by alternate means. AASHTO (2014) provides 

evaluation methods to assess the strength limit state of composite sections, 

provided the following criteria are met.

1. The cross-sectional area of steel is at least 4 percent of the total crosssectional area. If the cross sectional area of steel is less than this limit, the pile 

is considered non-composite and should be evaluated following procedures in 

Section 8.6 Concrete Piles.

2. The concrete compression strength is between 3.0 ksi and 8.0 ksi. 

Commentary provided in AASHTO (2014) notes the lower limit is imposed to 

encourage use of good quality concrete.
465

3. The yield strength of longitudinal reinforcement to determine the compression 

resistance cannot exceed 60.0 ksi.

For axial compression loading of concrete filled steel pipes, the factored structural 

limit state is taken as:

 Eq. 8-71 

Where:

Pr = factored resistance in axial compression (kips).

φ = resistance factor (Table 8-6).

Pn = nominal resistance in axial compression (kips).

STEP BY STEP PROCEDURE FOR: “NOMINAL COMPRESSION RESISTANCE”

STEP 1 Determine the normalized column slenderness factor, λ. 

The normalized column slenderness factor is be evaluated using the following 

equation. However, if the pile is fully embedded, λ may be taken as 0 (AASHTO 

2014). 

 Eq. 8-72 

In which: 

 Eq. 8-73 

 Eq. 8-74 

 

Eq. 8-75 

Where: 

λ = normalized column slenderness factor.

Ast = cross sectional area of steel (in2

). 

Ac = cross sectional area of concrete (in2). 

Astr = cross sectional area of longitudinal reinforcing steel (in2). 

K = effective length factor (Figure 8-4). 

l = unbraced length in the plane of buckling (inches).

rs = radius of gyration about axis normal to plane of buckling (inches).
466

Fe = nominal compression resistance of composite section (ksi).

Fy = yield stress of steel (Table 8-1) (ksi).

Fyr = yield stress of reinforcing steel (ksi). 

f’c = concrete compression strength at 28 days, unless otherwise 

specified (ksi).

Ee = modified elastic modulus of steel for composite column (ksi).

Est = elastic modulus of steel (ksi).

Ec = elastic modulus of concrete (Equation 8-1) (ksi).

C1 = composite column constant 1 (1.00 for concrete filled pipes). 

C2 = composite column constant 2 (0.85 for concrete filled pipes). 

C3 = composite column constant 3 (0.40 for concrete filled pipes). 

STEP 2 Determine the nominal resistance in axial compression, Pn. 

After determining the normalized column slenderness ratio, a relatively 

straightforward calculation of the nominal resistance in axial compression is made 

using either Equation 8-73 or Equation 8-74.

If : 

 Eq. 8-76

If : 

 Eq. 8-77 

Where: 

λ = normalized column slenderness factor. 

Pn = nominal resistance in axial compression (kips).

Ast = cross sectional area of steel (in2

). 

Fe = nominal resistance in axial compression of composite section (ksi).
467

8.8 LAYOUT OF PILE GROUPS

A group of piles is typically required to support large structural loads. An initial group 

layout must be determined to perform in-service stress checks. The possible loads 

are illustrated in Figure 8-6. The various load combinations, as given in Article 3.4 of 

the 2014 AASHTO Standard Specification, should be investigated to determine the 

pile stress conditions. Pile group layouts can be determined through the use of 

computer software such as GROUP or FB-Pier, which utilize soil structure 

interaction (t-z and p-y curves) to distribute loads and deflections throughout the pile 

group. Alternatively, a hand calculation for the Rigid Cap Method may be performed 

as follows.

STEP BY STEP PROCEDURE FOR: “LAYOUT OF PILE GROUPS”

STEP 1 Estimate the number of piles required to resist structural loads.

An initial, or trial, group layout may be computed by dividing the factored axial load 

acting on the pile cap/group by the factored geotechnical resistance of a single pile, 

and then rounding the number up (say, by 15% or more, depending upon magnitude 

of the moments and lateral loads) to a constructible pile layout. Therefore, the 

number of piles is estimated as:

 (number of piles, round up) Eq. 8-78

Where:

Pu = largest factored, axial load of the superstructure (kips).

Rr = factored resistance of a single pile (geotechnical) (kips).

n = number of piles in pile group.

Please note however that for many highway structures, lateral or tension loading 

may govern the design. Although the Equation 8-78 provides a preliminary guide to 

estimate the number of piles for axial compression loads, additional piles may be 

required to resist other loads and limit deflection.

STEP 2 Generate trial pile group configuration.

A trial configuration for the group of piles should be developed with n-piles. 

Minimum center-to-center pile spacing requirements by AASHTO (2014) 

specifications is the lesser of 2.5 feet or 2.5 pile diameters. However, as 
468

recommended in the previous edition of this manual, experience with pile group 

loading and efficiency has shown that 3.0 pile diameters is a more practical pile 

spacing minimum. Example pile group layouts can be found in CRSI Design 

Handbook (2002) and other sources.

STEP 3 Determine the maximum single pile axial load.

The trial group configuration consisting of n-piles should be checked for adequate 

single pile axial resistance under the combined superstructure/substructure axial 

loads and moments. The various factored load combinations, and not just the 

combination with the largest axial load, should be checked to determine the critical 

loading case. If the stiffness of all piles in the group are equal, then the maximum 

single pile factored axial load, Pui, may be computed with the rigid cap method as:

 Eq. 8-79

Where:

Pui = maximum single pile factored axial load (kips).

Puz = factored axial load from superstructure/substructure acting upon pile 

cap (kips).

Wc = estimated weight of pile cap (kips).

Ws = estimated weight of soil above pile cap, if applicable (kips).

n = number of piles in pile group.

Mux = factored moment about the x axis acting on the pile cap (kip-ft).

 Muy = factored moment about the y axis acting on the pile cap (kip-ft).

x = distance along x-axis from the center of the column to each pile

center (feet).

y = distance along y-axis from the center of the column to each pile 

center (feet).

If the maximum single pile factored axial load exceeds the pile factored structural 

resistance, add one, or more, piles to the group or increase the pile spacing if Pui > 

Puz and recompute the maximum single pile factored axial load (Pui). If the moments 

in one direction are substantially larger than the other, it may be desirable to make 

the cap unsymmetrical.
469

Figure 8-6 Pile group layout.

(Note: Column supported on pile cap not shown for clarity)

 
470

8.9 PRELIMINARY DESIGN OF PILE BENT AND GROUP CAPS

The purpose of this section is to provide guidelines to develop a preliminary size of a 

pile cap for the purposes of cost estimating. Information for complete, 

comprehensive structural design, including design for seismic and other extreme 

events, is beyond the scope of this manual, and not presented herein.

A state-of-the-practice manual on pile cap design was recently published by the 

Concrete Reinforcing Steel Institute (CRSI). This manual, “Design Guide for Pile 

Caps” (Mays 2015), was developed to provide the practicing engineer with a detailed 

overview of pile cap design, analysis methods, and detailing. It follows the American 

Concrete Institute ACI-14: Building Code Requirements for Structural Concrete and 

Commentary (ACI 2014) and covers piles with nominal loads up to 400 tons. Design 

procedure for vertical, lateral/overturning, and seismic loading is presented. 

Spreadsheets for pile cap design, following ACI-14, are available for download at 

http://www.crsi.org/pilecaps.cfm. This guide and spreadsheets are useful tools to 

develop a preliminary cap design for detailed analysis following AASHTO (2014).

In bridge design, the LRFD structural design of a concrete pile bent cap is similar to 

that of structural design for a concrete pier cap (Wilson et al. 2006). However, 

several design considerations are unique to cap design including: pile embedment 

into the cap, structural design depth, pile concrete cover requirements, pile 

misalignment tolerance, concrete pile anchorage, and minimum pile spacing 

requirements.

The design and size of the pile cap is dependent on the pile group layout, pile loads, 

and superstructure loads. Thus, an iterative design is required to optimize overall 

economics. The horizontal dimensions of the pile cap for the trial pile group 

configuration may be estimated, per AASHTO (2014) Article 10.7.1.2, by using: (i) 

the minimum center-to-center pile spacing of not less than 30.0 inches or 2.5 pile 

diameters; and (ii) a minimum edge of cap to side of pile distance of 9 inches. 

Maximum width and/or length of pile cap may be dictated by project constraints.

The thickness of the pile cap is a sum of the pile embedment into the cap, clear 

space between the cap reinforcing steel and the top of (embedded) piles, and 

thickness required for structural support. Per AASHTO (2014) Article 10.7.1.2, the 

piles shall project at least 12.0 inches into the cap after damaged pile material has 

been removed. If the piles and cap are attached by embedded bars or strands, this 

thickness may be reduced to 6.0 inches. The reinforced concrete must be designed 

with consideration of flexure and shear, for the factored loads. Potential shear 
471

failures include punching about a single pile, punching about a pair of piles, a single 

pile across the corner of the cap, and punching of the (superstructure) column 

across the width(s) of the cap.

STEP BY STEP PROCEDURE FOR: “LAYOUT OF PILE GROUPS”

STEP 1 Estimate the total thickness of cap. 

An initial, trial total (including pile embedment and 3 inches clear space between top 

of piles and reinforcing steel) thickness of the pile cap may be estimated from 

experience, agency guidelines or standards, or with the following equation: 

 Eq. 8-80

Where:

tcap = thickness of cap (inches).

Pui = maximum single pile factored axial load (kips).

This initial, trial thickness should be refined by examining punching shear in the 

reinforced concrete pile cap. Equations for these preliminary calculation steps are 

shown below.

STEP 2 Determine dimensions for computations. 

1. Determine effective shear depth to concrete reinforcement, dv. The 

effective shear depth, dv, is taken as the distance between the resultants 

of the tensile and compression forces due to flexure, as illustrated in 

Figure 8-7, where:

 Eq. 8-81
472

Where:

dv = effective depth to reinforcement (inches).

tcap = thickness of cap (inches).

dc = clear space (inches).

ds = distance to center of steel (inches).

df = pile embedment into cap (inches). 

h = structural depth, thickness of cap less pile embedment (inches).

The effective shear depth for a regularly reinforced section, per AASHTO (2014) 

Article 5.8.2.9, need not be taken to be less than the greater of 0.9ds or 0.72h

(inches). Thus, an assumed value of the effective shear depth, dv, for preliminary 

sizing may be taken as the maximum of 0.9ds or 0.72h (inches).

Figure 8-7 Cap section and notation.
473

2. Determine critical punching shear perimeter, bo, around the column 

supported on the pile cap. The shear force applied the shear perimeter 

is the load acting outside that perimeter.

For square columns:

 Eq. 8-82 

For rectangular columns:

 Eq. 8-83 

For circular columns: 

 Eq. 8-84

Where:

bo = critical punching shear perimeter (inches).

dv = effective shear depth (inches).

bc = column side for square columns (inches).

c1 = small column side for rectangular columns (inches).

c2 = large column side for rectangular columns (inches).

co = column diameter (inches).

STEP 3 Check punching (two-way) shear at: (i) at a distance dv/2 from face of 

column and at the face of the column, as illustrated in Figure 8-8; and (ii) around 

individual and pair of piles as illustrated in Figures 8-9 and 8-10. 

1. Compute total applied two-way, punching shear stress, Vu, at critical section.

For shear around column:

 Eq. 8-85 

Where:

Vu = factored shear force (kips). 

Pu = factored axial load of the superstructure (kips).
474

For shear around individual or pair of piles: 

 Eq. 8-86 

Where:

Vu = factored shear force (kips). 

Rr = factored (geotechnical) axial resistance of a single pile (kips).

ni = number of piles whose centers lie inside the two-way shear critical 

section.

Figure 8-8 Critical punching (two-way) shear sections around column.
475

Figure 8-9 Critical section from overlapping critical perimeters 

(after AASHTO 2014).

2. Determine the nominal shear strength of concrete, per AASHTO Article 

5.13.3.6.3, for two way action. Without transverse reinforcement, the nominal 

resistance is calculated using Equation 8-84. Sections with transverse 

reinforcement resist shear by means presented in Equation 8-85. 

Without Transverse Reinforcement (AASHTO 5.13.3.6.3-1):

 Eq. 8-87 

With Transverse Reinforcement (AASHTO 5.13.3.6.3-2):

 Eq. 8-88 

In which (AASHTO 5.13.3.6.3-3,-4):

 Eq. 8-89 

 

Eq. 8-90 
476

Where:

Vn = nominal shear resistance (kips).

Vs = nominal shear resistance provided by steel reinforcement (kips).

Vc = nominal shear resistance provided by concrete tensile strength (kips).

βc = ratio of the long side to the short side of the load. 

f’c = concrete compression strength at 28 days, unless otherwise 

specified (ksi).

Fy = specified yield stress of reinforcement (ksi). 

bo = critical punching (two-way) shear perimeter (inches).

dv = effective shear depth (inches).

Astv = area of transverse reinforcement within distance, s (in2

).

s = spacing of the transverse reinforcement (inches).

Figure 8-10 Pile punching shear for: (a) circular pile; (b) rectangular H pile; (c) 

circular pile near corner of cap; and (d) two nearby circular piles.
477

3. Compute the factored shear resistance from:

= φ VV nvr Eq. 8-91 

Where:

Vr = factored shear resistance (kips).

φv = resistance factor (0.9 for normal weight concrete shear).

Vn = nominal shear resistance (kips).

4. Check that the factored shear resistance is greater than the factored shear 

load.

Vu < Vr Eq. 8-92 

Where:

Vu = factored shear load (kips).

Vr = factored shear resistance (kips).

If the above condition is not satisfied, increasing shear reinforcement or the effective 

shear depth should be considered. 

STEP 4 Check One-Way Beam Shear, per AASHTO Article 5.13.3.6.2.

1. Compute total applied beam (one-way) shear stress, Vu, at critical sections, 

per AASHTO (2014) Article 5.8.3.2. Determine the beam shear distance from 

column to critical section as illustrated in Figure 8-11. Determine beam shear 

distance from the pile to the critical section(s), as illustrated in Figure 8-12. 

2. For each direction of pile cap, check number of piles that lie outside of the 

critical section. Compute total applied one-way, beam shear stress, Vu, at 

critical section. 

== RnQV roiiu γ Eq. 8-93 

Where:

Vu = factored shear stress (kips). 

Rr = factored nominal resistance of a single pile (geotechnical) (kips).

no = number of piles which lie outside of the one-way critical shear 

section.
478

Figure 8-11 One-way beam shear critical sections from column. 

Figure 8-12 One-way beam shear critical sections from pile(s). 
479

3. Determine the nominal shear strength of concrete, per AASHTO Article 

5.8.3.3. The nominal shear resistance, Vn, without prestressed reinforcement, 

is the lesser of Equation 8-91 or 8-92 per AASHTO Article 5.8.3.3-2:

 Eq. 8-94

 Eq. 8-95 

Where:

Vn = nominal shear resistance (kips).

Vs = nominal shear resistance provided by steel reinforcement (kips).

Vc = nominal shear resistance provided by concrete tensile strength (kips).

f’c = concrete compression strength at 28 days, unless otherwise 

specified (ksi).

bv = width of interface (inches).

dv = effective depth to reinforcement (inches).

4. Compute the factored shear resistance from:

 

Where:

Vr = factored shear resistance (kips).

φv = resistance factor (0.9 for normal weight concrete shear per AASHTO 

Article 5.5.4.2.1).

Vn = nominal shear resistance (kips).

5. Check that the factored shear resistance is greater than the factored shear 

load.

Vu < Vr Eq. 8-96

Where:

Vu = factored shear force (kips).

Vr = factored shear resistance (kips)
480

STEP 5 Check Bending in Cap. 

1. Compute factored bending moment, for each direction of pile cap, from piles 

to edge of column.

( ) u = ∑ r ≤ φMyxRM n , Eq. 8-97 

Where: 

Mu = factored moment load (kip-in).

Rr = factored (geotechnical) axial resistance of a single pile (kip). 

x,y = distance from edge of column to each pile (in both x and y directions) 

(inches).

Mn = nominal flexural resistance (kip-in).

φ = resistance factor (0.9 for normal weight concrete in shear or torsion).

The moment resistance of a reinforced concrete beam is determined based on the 

assumption of a rectangular distribution of the compression stress in the concrete at 

failure. The design must fail by yielding the tension steel to assure ductility. The 

nominal flexural resistance, Mn, for a non-prestressed rectangular section can be 

computed with the following equation (after AASHTO 5.7.3.2.2-1):





 



  − −  



 = − 2 2

a dfA a n dfAM stststt scscstc Eq. 8-98

In which:

bf

fAfA

bf

fAfA

ca

c

ssss

c

ssss

'85.0

''

'85.0

''

1

1

1

− = − == β

β β Eq. 8-99

Where:

Mn = nominal flexural resistance (kip-in).

Astt = area of tension reinforcement (in2).

fst = stress in the mild steel tension reinforcement at nominal flexural 

resistance (ksi), as specified in AASHTO Article 5.7.2.1.

dst = distance from extreme compression fiber to the centroid of the tensile 

reinforcement (in).

a = c β1, depth of the equivalent stress block (in).

b = width of the compression face.

c = distance between the neutral axis and the compression face (in). 

β1 = stress block factor, which cancels in Equation 8-99. 

Astc = area of compression reinforcement (in2).
481

fsc = stress in the mild steel compression reinforcement at nominal flexural 

resistance (ksi), as specified in AASHTO Article 5.7.2.1.

dsc = distance from extreme compression fiber to the centroid of the 

compression reinforcement (in).

The equation for moment strength, Mn, can be solved for the area of steel, Ast. 

However, a/2 is small compared with d so an approximate value can be assumed for 

a/2, Ast can be calculated and the steel area adjusted to arrive at a satisfactory Ast. 

If d-a/2 is assumed to be 0.9d, Ast will be quite close and it will be satisfactory for 

preliminary design. If an improved Ast is desired it can be obtained made by 

determining an improved aa with knowledge of the preliminary Ast and with the new aa

the next cycle of Ast can be determined and it will probably be final. 

Per AASHTO Article 5.7.3.3.2, the tensile reinforcement should be adequate to 

develop a factored flexural resistance, Mr of at least 1.33 times the factored moment. 

See Article 5.7.3.3.2 for more details. 

The final, structural design of pile caps is beyond the scope of this manual. See 

AASHTO (2014), ACI (2014), etc. for guidance on detailed structural design.

8.9.1 Cap Considerations for Large Pile Sizes 

For Large Diameter Open-Ended Piles (LDOEPs), general analysis steps in the 

above section should be followed. No strict guidance can be offered at this time 

regarding minimum spacing and embedment into the pier cap, specifically for large 

pile sizes. Case studies presented in Brown and Thompson (2014) present few 

details on this subject as most state agencies do not currently have standard plans 

for these pile sizes. Pile tops were however embedded further into the cap, than 

with smaller pile sections. This was on the order of 3 feet into the pier cap for 

LDOEPs.

Pile spacing considerations should reflect guidelines proposed for drilled shafts due 

to the comparable element diameter. The presented pile plan layout in Brown and 

Thompson (2014) shows a minimum pile spacing on the order of 3 pile diameters 

(center to center), which is consistent with typical drilled shaft group layouts (Brown 

et al. 2010).
482

REFERENCES

American Association of State Highway and Transportation Officials (AASHTO). 

(2014). AASHTO LRFD Bridge Design Specifications, US Customary Units, 

Seventh Edition, with 2015 Interim Revisions. American Association of State 

Highway and Transportation Officials, Washington, D.C., 1960 p.

American Institute of Steel Construction (AISC). (2011). Steel Construction Manual, 

Fourteen Edition, (Third Printing). American Institute of Steel Construction 

Chicago, IL, 2192 p.

American Concrete Institute (ACI) (2014). Building Code Requirements for 

Structural Concrete and Commentary, ACI 318-14. Farmington Hills, MI, 520 

p.

Bethlehem Steel Corporation. (1970). Steel H Piles, Handbook 2196-B. Bethlehem 

Steel Corporation, Bethlehem, PA, 64 p.

Brown, D. A., Turner, J.P. and Castelli R.J. (2010). Drilled Shafts: Construction 

Procedures and LRFD Design Methods, FHWA-NHI-10-016. Geotechnical 

Engineering Circular (GEC) No. 10, U.S. Dept. of Transportation, Federal 

Highway Administration (FHWA), Washington D.C., 970 p.

Brown, D.A., and Thompson III, W.R. (2015). Current Practices for Design and 

Load Testing of Large Diameter Open-End Driven Pipe Piles. Final Report. 

NCHRP Report 20-05, Topic 45-05, National Cooperative Highway Research 

Program, (NCHRP) Washington, D.C., 175 p. 

Coduto, D P., and Kitch, W.A., and Yeung, M.R. (2016). Foundation Design: 

Principles and Practices, Third Edition. Pearson Education, Upper Saddle 

River, NJ:, 686 p.

Davisson, M.T. (1963). Estimating Buckling Loads for Piles. Proceedings of the 

Second Pan-American Conference on Soil Mechanics and Foundation 

Engineering, Brazil, Vol. 1, pp. 351-371.
483

Davisson, M.T. and K.E. Robinson. (1965). Bending and Buckling of Partially 

Embedded Piles. Proceedings Sixth International Conference on Soil 

Mechanics and Foundation Engineering, University of Toronto Press. 

Montreal, Canada, pp. 243-246.

Mays, T. (2015). Design Guide for Pile Caps, First Edition. Concrete Reinforcing 

Steel Institute (CSRI), Schaumburg, IL, 152 p.

Reese, L.C. (1984). Handbook on Design of Piles and Drilled Shafts Under Lateral 

Load, FHWA-IP-84-11. U.S. Department of Transportation, Federal Highway 

Administration, Office of Implementation, Washington, D.C., 392 p.

Rollins, K. and Stenlund, T. (2010). Laterally Loaded Pile Cap Connections. Utah 

Department of Transportation, Research Division, Report UT-10.16, 133 p. 

Showalter, J. (2012). Round Timber Poles and Piles. Wood Design Focus, Forest 

Products Society, Vol. 22, No. 4, pp. 14-18.

Skyline Steel. (2015). Steel H-pile Technical Datasheets. Skyline Steel, 

http://www.skylinesteel.com. 

White, C.D., and Castrodale, R.W. (2004). Precast/Prestresed Concrete Institute 

Bridge Design Manual – Chapter 20 Precast Prestressed Concrete Piles, 

MNL 133. Precast/Prestressed Concrete Institute, Chicago, IL, 171 p.

Wilson, K.E., Kimmerling, R.E., Goble, G.G., Sabatini, P.J., Zang, S.D., Zhou, J.Y. 

Amrhein, W.A., Bouscher, J.W. and Danaovich, L.J. (2006). LRFD for 

Highway Bridge Substructures and Earth Retaining Structures Reference 

Manual, FHWA NHI-05-094. U.S. Dept. of Transportation, Federal Highway 

Administration (FHWA), Washington, D.C., 1730 p.
484
485

APPENDIX A

LIST OF FHWA/ NHI RESOURCES RELEVANT TO DEEP FOUNDATIONS

Abu-Hejleh, N., DiMaggio, J.A., Kramer, W.M., Anderson, S., and Nichols, S. (2010). 

Implementation of LRFD Geotechnical Design for Bridge Foundations: 

Reference Manual. FHWA-NHI-10-039. National Highway Institute, Federal 

Highway Administration, Washington, D.C.

Abu-Hejleh, N., Kramer, W.M., Mohamed, K., Long, J.H., and Zaheer, M.A. (2013). 

Implementation of AASHTO LRFD Design Specifications for Driven Piles, 

FHWA-RC-13-001. U.S. Dept. of Transportation, Federal Highway 

Administration, 71 p.

Arneson, L.A., Zevenbergen, L.W., Lagasse, P.F., and Clopper, P.E. (2012). 

Evaluating Scour at Bridges, Fifth Edition, FHWA-HIF-12-003, Hydraulic 

Engineering Circular (HEC) No. 18. U.S. Dept. of Transportation, Federal 

Highway Administration, 340 p.

Azizinamini, A., Power, E.H., Myers, G.F., and Ozyildirim, H.C. (2014). Bridges for 

Service Life Beyond 100 Years, Innovative Systems Subsystems, and 

Components, S2-R19A-RW-1 Strategic Highway Research Program 2 

(SHRP), Transportation Research Board, Washington, D.C., 248 p.

Briaud J-L. and Miran, J. (1992). The Cone Penetration Test, FHWA-SA-91-043. 

U.S. Department of Transportation, Federal Highway Administration, Office of 

Technology Applications, Washington, D.C., 161 p.

Brown, D.A., and Thompson, W.R. (2011). Developing Production Pile Criteria from 

Test Pile Data. National Cooperative Highway Research Program (NCHRP) 

Synthesis 418, Washington, D.C,. 54 p.

Brown, D. A., Turner, J.P. and Castelli R.J. (2010). Drilled Shafts: Construction 

Procedures and LRFD Design Methods, FHWA-NHI-10-016, Geotechnical

Engineering Circular (GEC) No. 10. U.S. Dept. of Transportation, Federal 

Highway Administration, 970 p.
486

Brown, D.A., Dapp, S.D., Thompson, W.R., and Lazarte, C.A. (2007). Design and 

Construction of Continuous Flight Auger (CFA) Piles. FHWA-HIF-07-03, 

Geotechnical Engineering Circular (GEC) No.08. U.S. Dept. of 

Transportation, Federal Highway Administration, 289 p.

Cheney, R.S. and Chassie, R.G. (2000). Soils and Foundations Workshop 

Reference Manual. FHWA HI-00-045, U.S. Department of Transportation, 

National Highway Institute, Federal Highway Administration, Washington, 

D.C., 358 p.

Elias, V., Welsh, J.P., Warren, J., Lukas, R.G., Collin J.G., and Berg, R.R. (2006). 

Ground Improvement Methods Volumes I and II, FHWA-NHI-06-019 and 

FHWA NHI-06-020. National Highway Institute, Federal Highway 

Administration, U.S. Department of Transportation, Washington D.C.

Federal Highway Administration (FHWA). (1996). Geotechnical Engineering 

Notebook DT-15. Differing Site Conditions. U.S. Dept. of Transportation, 

Federal Highway Administration, Washington, D.C. 36 p.

Federal Highway Administration (FHWA). (2002a). Life-Cycle Cost Analysis Primer, 

IF-02-047. Federal Highway Administration, U.S. Department of 

Transportation. Washington, D.C., 24 p.

Geotechnical Guideline 13 (1985). Geotechnical Engineering Notebook, U.S. 

Department of Transportation. Federal Highway Administration, Washington, 

D.C., 37 p.

Ghosn, M., Moses, F., and Wang, J. (2003). Design of Highway Bridges for Extreme 

Events, NCHRP Report 489. National Cooperative Highway Research 

Program. Transportation Research Board, Washington, D.C., 174 p.

Goble, G.G. and Rausche, F. (1976). Wave Equation Analysis of Pile Driving – 

WEAP Program, FHWA IP-76-14.3., U.S. Department of Transportation, 

Federal Highway Administration, Office of Research and Development, 

Washington, D.C., Volumes I-IV.

Goble, G.G. and Rausche, F. (1986). Wave Equation Analysis of Pile Driving - 

WEAP86 Program. U.S. Department of Transportation, Federal Highway 

Administration, Implementation Division, McLean, Volumes I-IV.
487

Hawk, H. (2003). Bridge Life-Cycle Cost Analysis, NCHRP483. Transportation 

Research Board of the National Academies, Washington, D.C. 96 p.

Kavazanjian, E., Wan, J-N. J., Martin, G.R., Shamsabadi, A., Lam, I., Dickenson, 

S.E., and Hung, C.J. (2011). LRFD Seismic Analysis and Design of 

Transportation Geotechnical Features and Structural Foundations, FHWANHI-11-032, Geotechnical Engineering Circular (GEC) No. 3. U.S. Dept. of 

Transportation, Federal Highway Administration, Washington, D.C., 592 p.

Kimmerling, R.E. (2002). Shallow Foundations, FHWA-IF-02-054, Geotechnical 

Engineering Circular (GEC) No. 6. U.S. Dept. of Transportation, Federal 

Highway Administration, Washington, D.C., 310 p.

Kyfor, Z.G., Schnore, A.R., Carlo, T.A., and Bailey, P.F. (1992). Static Testing of 

Deep Foundations, FHWA-SA-91-042. U.S. Department of Transportation, 

Federal Highway Administration, Office of Technology Applications, 

Washington, D.C., 174 p.

Lam, I.P. and Martin, G.R. (1986). Seismic Design of Highway Bridge Foundations. 

Volume II - Design Procedures and Guidelines, FHWA-RD-86-102. U.S. 

Department of Transportation, Federal Highway Administration, Office of 

Engineering and Highway Operations, McLean, VA, 167p.

Long, J., and Anderson, A. (2014). Improved of Driven Pile Installation and Design in 

Illinois: Phase 2, FHWA-ICT-14-019. Illinois Department of Transportation, 

Bureau of Material and Physical Research, Springfield, IL, 84 p.

Marsh, M.L., Buckle, I.G., and Kavazanjian Jr, E. (2014). LRFD Seismic Analysis 

and Design of Bridges, FHWA-NHI-15-004. National Highway Institute, U.S. 

Dept. of Transportation, Federal Highway Administration, Washington, D.C., 

608 p.

Mayne, P.W., Christopher, B., Berg, R., and DeJong, J. (2001). Manual on 

Subsurface Investigations, FHWA NHI-01-031, U.S. Dept. of Transportation, 

National Highway Institute, Federal Highway Administration, Washington, 

D.C., 301 p.

Munfakh, G., Arman, A., Collin, J.G., Hung, J.C.-J., and Brouillette, R.P. (2001). 

Shallow Foundations Reference Manual, FHWA-NHI-01-023. National 

Highway Institute, Federal Highway Administration, Washington, D.C., 222 p.
488

Rausche, F., Likins, G.E., Goble, G.G., Hussien, M. (1986). The Performance of Pile 

Driving Systems; Inspector's Manual. U.S. Department of Transportation, 

Federal Highway Administration, Washington, D.C., 92 p.

Rausche, F., Thendean, G., Abou-matar, H., Likins, G.E. and Goble, G.G. (1996). 

Determination of Pile Drivability and Capacity from Penetration Tests, 

DTFH61-91-C-00047, Final Report. U.S. Department of Transportation, 

Federal Highway Administration, McLean, VA, 432 p.

Reese, L.C. (1984). Handbook on Design of Piles and Drilled Shafts Under Lateral 

Load. Report No. FHWA-IP-84-11, U.S. Department of Transportation, 

Federal Highway Administration, Office of Implementation, McLean, VA, 386 

p.

Reese, L.C. (1986). Behavior of Piles and Pile Groups Under Lateral Load, FHWARD-85-106. U.S. Department of Transportation, Federal Highway 

Administration, Office of Engineering and Highway Operations Research and 

Development, Washington, D.C., 311 p.

Rixner, J.J., Kraemer, S.R. and Smith, A.D. (1986). Prefabricated Vertical Drains 

Volume I, Engineering Guidelines, FHWA-RD-86-168. U.S. Department of 

Transportation, Federal Highway Administration, Office of Engineering and 

Highway Operations Research and Development, McLean, VA, 117 p.

Sabatini, P. J., Elias, V., Schmertmann, G. R., and Bonaparte, R. (1997). Earth 

Retaining Systems FHWA-SA-96-038, Geotechnical Engineering Circular 

(GEC) No. 2. U.S. Department of Transportation, Federal Highway 

Administration, Washington, D.C., 161 p.

Sabatini, P.J., Tanyu, B., Armour., P., Groneck, P., and Keeley, J. (2005). Micropile 

Design and Construction, FHWA-NHI-05-039. National Highway Institute, 

U.S. Dept. of Transportation, Federal Highway Administration, Washington, 

D.C., 436 p.

Samtani, N.C. and Nowatzki, E.A. (2006). Soils and Foundations: Reference 

Manual, Vol. 1, FHWA-NHI-06-088, U.S. Dept. of Transportation, National 

Highway Institute, Federal Highway Administration, Washington, D.C., 462 p.
489

Samtani, N.C., Nowatzki, E.A., and Mertz, D.R. (2010). Selection of Spread Footings 

on Soils to Support Highway Bridge Structures, FHWA-RC TD-10-001. U.S. 

Department of Transportation, Federal Highway Administration, Washington, 

D.C., 98 p.

Schmertmann, J.H. (1978). Guidelines For Cone Penetration Test, Performance, 

and Design, FHWA-TS-78-209. U.S. Department of Transportation, Federal 

Highway Administration, Washington, D.C., 145 p.

Tanyu B.F., Sabatini, P. J., and Berg, R.R. (2008). Earth Retaining Structures, 

FHWA-NHI-07-07. U.S. Department of Transportation, Federal Highway 

Administration, Washington, D.C., 792 p.

Wightman, W., Jalinoos, F., Sirles., and Hanna, K. (2003), Applications of 

Geophysical Methods to Related Highway Problems. U.S. Dept. of 

Transportation, Federal Highway Administration (FHWA), Washington D.C., 

716 p.

Wilson, K.E., Kimmerling, R.E., Goble, G.C., Sabatini, P.J., Zang, S.D., Zhou, J.Y. 

Amrhein, W.A., Bouscher, J.W. and Danaovich, L.J. (2006). LRFD for 

Highway Bridge Substructures and Earth Retaining Structures Reference 

Manual, FHWA NHI-05-094. U.S. Dept. of Transportation, Federal Highway 

Administration, Washington, D.C., 1730 p.
490
491

Appendix B

LIST OF ASTM AND AASHTO PILE DESIGN AND TESTING SPECIFICATIONS

American Association of State Highway and Transportation Offices (AASHTO). 

Standard Method of Test for High Strain Dynamic Testing of Piles, AASHTO 

Designation T-298-33.

American Association of State Highway and Transportation Officials (AASHTO). 

(1978). Manual on Foundation Investigations Second Edition.. AASHTO 

Highway Subcommittee on Bridges and Structures, Washington, D.C., 196 p.

American Association of State Highway and Transportation Officials (AASHTO). 

(2001). Standard Recommended Practice for Assessment of Corrosion of 

Steel Piling for Non-Marine Applications. AASHTO Standard Specifications 

for Transportation Materials and Methods of Sampling and Testing, Part 1B: 

Specifications, 24th Edition, 13 p.

American Association of State Highway and Transportation Officials (AASHTO). 

(2011) Guide Specifications for LRFD Seismic Bridge Design, 2nd Edition, 

with 2012, 2014, and 2015 Interim Revisions. American Association of State 

Highway and Transportation Officials, Washington, D.C., 331 p.

American Association of State Highway and Transportation Officials (AASHTO). 

(2014). AASHTO LRFD Bridge Design Specifications, US Customary Units, 

Seventh Edition, with 2015 Interim Revisions. American Association of State 

Highway and Transportation Officials, Washington, D.C., 1960 p.

ASTM A27-13. (2014). Standard Specification for Steel Castings, Carbon, for 

General Application. Book of ASTM Standards, Vol. 1.02, ASTM International, 

West Conshohocken, PA, 4 p. 

ASTM A572-15. (2015). Standard Specification for High-Strength Low-Alloy 

Columbium-Vanadium Structural Steel. Book of ASTM Standards, Vol. 1.04, 

ASTM International, West Conshohocken, PA, 4 p.
492

ASTM D1143-07. (2014). Standard Test Methods for Deep Foundations Under 

Static Axial Compressive Load. Book of ASTM Standards, Vol. 4.08, ASTM 

International, West Conshohocken, PA, 15 p.

ASTM D1452-09. (2014). Standard Practice for Soil Exploration and Sampling by 

Auger Borings. Annual Book of ASTM Standards, Vol. 4.08, ASTM 

International, West Conshohocken, PA, 6 p. 

ASTM D1586-11. (2014). Standard Test Method for Standard Penetration Test 

(SPT) and Split-Barrel Sampling of Soils. Annual Book of ASTM Standards, 

Vol. 4.08, ASTM International, West Conshohocken, PA, 9 p.

ASTM D1587-12. (2014). Standard Practice for Thin-Walled Tube Sampling of Soils 

for Geotechnical Purposes. Annual Book of ASTM Standards, Vol. 4.08, 

ASTM International, West Conshohocken, PA, 4 p.

ASTM D2113-14. (2014). Standard Practice for Rock Core Drilling and Sampling of 

Rock for Site Investigation. Annual Book of ASTM Standards, Vol. 4.08, 

ASTM International, West Conshohocken, PA, 20 p.

ASTM D2573-08. (2012). Standard Test Method for Field Vane Shear Test in 

Cohesive Soil. Annual Book of ASTM Standards, Vol. 4.08, ASTM 

International, West Conshohocken, PA, 8 p.

ASTM D3689-07. (2014). Standard Test Methods for Deep Foundations Under 

Static Axial Tensile Load. Book of ASTM Standards, Vol. 4.08, ASTM 

International, West Conshohocken, PA,13 p.

ASTM D4633-10. (2014). Standard Test Method for Energy Measurement for 

Dynamic Penetrometer. Book of ASTM Standards, Vol. 4.08, ASTM 

International, West Conshohocken, PA, 7 p.

ASTM D4719-07. (2014). Standard Test Methods for Prebored Pressuremeter 

Testing in Soils Annual. Book of ASTM Standards, Vol. 4.08, ASTM 

International, West Conshohocken, PA, 10 p.

ASTM D4945-12. (2014). Standard Test Method for High-Strain Dynamic Testing of 

Piles. Book of ASTM Standards, Vol. 4.08, ASTM International, West 

Conshohocken, PA, 9 p.
493

ASTM D4971-02. (2014). Standard Test Method for Determining the In Situ Modulus 

of Deformation of Rock Using the Diametrically Loaded 76-mm (3-in.) 

Borehole Jack. Book of ASTM Standards, Vol. 4.08, ASTM International, 

West Conshohocken, PA, 7 p.

ASTM D5778-12. (2014). Standard Test Method for Electronic Friction Cone and 

Piezocone Penetration Testing of Soils. Book of ASTM Standards, Vol. 4.08, 

ASTM International, West Conshohocken, PA, 20 p.

ASTM D5882-07. (2013). Standard Test Method for Low-Strain Dynamic Testing of 

Piles Annual Book of ASTM Standards, Vol. 4.09, ASTM International, West 

Conshohocken, PA, 6 p.

ASTM D6032-08. (2014). Standard Test Method for Determining Rock Quality 

Designation (RQD) of Rock Core. Annual Book of ASTM Standards, Vol. 

4.09, ASTM International, West Conshohocken, PA, 5 p.

ASTM D6635-07. (2014). Standard Test Method for Performing the Flat Dilatometer. 

Book of ASTM Standards, Vol. 4.09, ASTM International, West 

Conshohocken, PA, 16 p.

ASTM D7012-14. (2014). Standard Tests Method for Compressive Strength and 

Elastic Moduli of Intact Rock Core Specimens under Varying States of Stress 

and Temperatures. Annual Book of ASTM Standards, Vol. 4.09, ASTM 

International, West Conshohocken, PA, 9 p.

ASTM D7383-10 (2010). Standard Test Methods for Axial Compressive Force Pulse 

(Rapid) Testing of Deep Foundations. Annual Book of ASTM Standards, Vol. 

4.08, ASTM International, West Conshohocken, PA, 9 p. ASTM

ASTM Vol 4.08. (2014). Soil and Rock I, Vol. 4.08, ASTM International, West 

Conshohocken, PA, 1826 p.

ASTM Vol 4.09. (2014). Soil and Rock II, Vol. 4.09, ASTM International, West 

Conshohocken, PA, 1754 p.
494
495

Appendix C

PILE HAMMER INFORMATION

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

1 DELMAG D 5 OED 10.51 1.10 9.55

2 DELMAG D 8-22 OED 20.10 1.76 11.42

3 DELMAG D 12 OED 22.61 2.75 8.22

4 DELMAG D 15 OED 27.09 3.30 8.21

5 DELMAG D 16-32 OED 40.20 3.52 11.42

6 DELMAG D 22 OED 40.61 4.91 8.27

7 DELMAG D 22-02 OED 48.50 4.85 10.00

8 DELMAG D 22-13 OED 48.50 4.85 10.00

9 DELMAG D 22-23 OED 51.22 4.85 10.56

10 DELMAG D 25-32 OED 66.34 5.51 12.04

11 DELMAG D 30 OED 59.73 6.60 9.05

12 DELMAG D 30-02 OED 66.20 6.60 10.03

13 DELMAG D 30-13 OED 66.20 6.60 10.03

14 DELMAG D 30-23 OED 73.79 6.60 11.18

15 DELMAG D 30-32 OED 75.44 6.60 11.43

16 DELMAG D 36 OED 83.82 7.93 10.57

17 DELMAG D 36-02 OED 83.82 7.93 10.57

18 DELMAG D 36-13 OED 83.82 7.93 10.57

19 DELMAG D 36-23 OED 88.50 7.93 11.16

20 DELMAG D 36-32 OED 90.56 7.93 11.42

21 DELMAG D 44 OED 90.16 9.50 9.49

22 DELMAG D 46 OED 107.08 10.14 10.56

23 DELMAG D 46-02 OED 107.08 10.14 10.56

24 DELMAG D 46-13 OED 96.53 10.14 9.52

25 DELMAG D 46-23 OED 107.08 10.14 10.56

26 DELMAG D 46-32 OED 122.19 10.14 12.05

27 DELMAG D 55 OED 125.00 11.86 10.54

28 DELMAG D 62-02 OED 152.45 13.66 11.16

29 DELMAG D 62-12 OED 152.45 13.66 11.16

30 DELMAG D 62-22 OED 164.60 13.66 12.05

31 DELMAG D 80-12 OED 186.24 17.62 10.57

32 DELMAG D 80-23 OED 212.50 17.62 12.06

33 DELMAG D100-13 OED 265.68 22.07 12.04

35 DELMAG D 19-52 OED 43.20 4.00 10.80

36 DELMAG D 6-32 OED 13.52 1.32 10.23
496

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

37 DELMAG D 12-32 OED 31.33 2.82 11.11

38 DELMAG D 12-42 OED 33.30 2.82 11.81

39 DELMAG D 14-42 OED 34.50 3.09 11.18

40 DELMAG D 19-32 OED 42.44 4.00 10.61

41 DELMAG D 19-42 OED 43.24 4.00 10.81

42 DELMAG D200-42 OED 492.04 44.09 11.16

43 DELMAG D120-42 OED 301.79 26.45 11.41

44 DELMAG D150-42 OED 377.33 33.07 11.41

45 DELMAG D125-42 OED 313.63 27.56 11.38

46 DELMAG D 21-42 OED 55.75 4.63 12.04

47 DELMAG D 5-42 OED 10.56 1.10 9.60

48 DELMAG D160-32 OED 393.45 35.27 11.16

49 DELMAG D260-32 OED 639.36 57.32 11.16

50 FEC 1200 OED 22.50 2.75 8.18

51 FEC 1500 OED 27.09 3.30 8.21

52 FEC 2500 OED 50.00 5.50 9.09

53 FEC 2800 OED 55.99 6.16 9.09

54 FEC 3000 OED 63.03 6.60 9.55

55 FEC 3400 OED 73.01 7.48 9.76

56 FEC D-18 OED 39.70 3.97 10.00

61 MITSUBIS M 14 OED 25.25 2.97 8.50

62 MITSUBIS MH 15 OED 28.14 3.31 8.50

63 MITSUBIS M 23 OED 43.01 5.06 8.50

64 MITSUBIS MH 25 OED 46.84 5.51 8.50

65 MITSUBIS M 33 OED 61.71 7.26 8.50

66 MITSUBIS MH 35 OED 65.62 7.72 8.50

67 MITSUBIS M 43 OED 80.41 9.46 8.50

68 MITSUBIS MH 45 OED 85.43 10.05 8.50

70 MITSUBIS MH 72B OED 135.15 15.90 8.50

71 MITSUBIS MH 80B OED 149.60 17.60 8.50

81 LINKBELT LB 180 CED 8.10 1.73 4.68

82 LINKBELT LB 312 CED 15.02 3.86 3.89

83 LINKBELT LB 440 CED 18.20 4.00 4.55

84 LINKBELT LB 520 CED 26.31 5.07 5.19

85 LINKBELT LB 660 CED 51.63 7.57 6.82

90 HITACHI HNC65 ECH 56.42 14.33 3.94

91 HITACHI HNC80 ECH 69.43 17.64 3.94

92 HITACHI HNC100 ECH 86.79 22.05 3.94

93 HITACHI HNC125 ECH 108.49 27.56 3.94

101 KOBE K 13 OED 25.43 2.87 8.86

103 KOBE K22-Est OED 45.35 4.85 9.35

104 KOBE K 25 OED 51.52 5.51 9.35

107 KOBE K 35 OED 72.18 7.72 9.35
497

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

110 KOBE K 45 OED 92.75 9.92 9.35

112 KOBE KB 60 OED 130.18 13.23 9.84

113 KOBE KB 80 OED 173.58 17.64 9.84

120 ICE 180 CED 8.13 1.73 4.70

121 ICE 422 CED 23.12 4.00 5.78

122 ICE 440 CED 18.56 4.00 4.64

123 ICE 520 CED 30.37 5.07 5.99

124 ICE 640 CED 40.62 6.00 6.77

125 ICE 660 CED 51.63 7.57 6.82

126 ICE 1070 CED 72.60 10.00 7.26

127 ICE 30-S OED 22.50 3.00 7.50

128 ICE 40-S OED 40.00 4.00 10.00

129 ICE 42-S OED 42.00 4.09 10.27

130 ICE 60-S OED 59.99 7.00 8.57

131 ICE 70-S OED 70.00 7.00 10.00

132 ICE 80-S OED 80.00 8.00 10.00

133 ICE 90-S OED 90.00 9.00 10.00

134 ICE 100-S OED 100.00 10.00 10.00

135 ICE 120-S OED 120.00 12.00 10.00

136 ICE 200-S OED 100.00 20.00 5.00

137 ICE 205-S OED 170.00 20.00 8.50

139 ICE 32-S OED 26.01 3.00 8.67

140 ICE 120S-15 OED 132.45 15.00 8.83

142 MKT DE-20C OED 20.00 2.00 10.00

143 MKT DE-30C OED 28.00 2.80 10.00

144 MKT DE-33C OED 33.00 3.30 10.00

145 MKT DE333020 OED 40.00 4.00 10.00

146 MKT DE 10 OED 8.80 1.10 8.00

147 MKT DE 20 OED 16.00 2.00 8.00

148 MKT DE 30 OED 22.40 2.80 8.00

149 MKT DA35B SA OED 23.80 2.80 8.50

150 MKT DE 30B OED 23.80 2.80 8.50

151 MKT DA 35B CED 21.00 2.80 7.50

152 MKT DA 45 CED 30.72 4.00 7.68

153 MKT DE 40 OED 32.00 4.00 8.00

154 MKT DE 35 OED 35.00 3.50 10.00

155 MKT DE 42 OED 42.00 4.20 10.00

157 MKT DE 50C OED 50.00 5.00 10.00

158 MKT DE 70C OED 70.00 7.00 10.00

159 MKT DE 50B OED 42.50 5.00 8.50

160 MKT DA55B SA OED 40.00 5.00 8.00

161 MKT DA 55B CED 38.20 5.00 7.64

162 MKT DE 70B OED 59.50 7.00 8.50
498

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

163 MKT DE-50B OED 50.00 5.00 10.00

164 MKT DE-70B OED 70.00 7.00 10.00

165 MKT DE-110C OED 110.00 11.00 10.00

166 MKT DE-150C OED 150.00 15.00 10.00

167 MKT DA 35C CED 21.00 2.80 7.50

168 MKT DA 55C CED 38.20 5.00 7.64

171 CONMACO C 50 ECH 15.00 5.00 3.00

172 CONMACO C 65 ECH 19.50 6.50 3.00

173 CONMACO C 550 ECH 25.00 5.00 5.00

174 CONMACO C 565 ECH 32.50 6.50 5.00

175 CONMACO C 80 ECH 26.00 8.00 3.25

176 CONMACO C 100 ECH 32.50 10.00 3.25

177 CONMACO C 115 ECH 37.38 11.50 3.25

178 CONMACO C 80E5 ECH 40.00 8.00 5.00

179 CONMACO C 100E5 ECH 50.00 10.00 5.00

180 CONMACO C 115E5 ECH 57.50 11.50 5.00

181 CONMACO C 125E5 ECH 62.50 12.50 5.00

182 CONMACO C 140 ECH 42.00 14.00 3.00

183 CONMACO C 160 ECH 48.75 16.25 3.00

184 CONMACO C 200 ECH 60.00 20.00 3.00

185 CONMACO C 300 ECH 90.00 30.00 3.00

186 CONMACO C 5200 ECH 100.00 20.00 5.00

187 CONMACO C 5300 ECH 150.00 30.00 5.00

188 CONMACO C 5450 ECH 225.00 45.00 5.00

189 CONMACO C 5700 ECH 350.00 70.00 5.00

190 CONMACO C 6850 ECH 510.00 85.00 6.00

191 CONMACO C 160 ** ECH 51.78 17.26 3.00

192 CONMACO C 50E5 ECH 25.00 5.00 5.00

193 CONMACO C 65E5 ECH 32.50 6.50 5.00

194 CONMACO C 200E5 ECH 100.00 20.00 5.00

195 CONMACO C 300E5 ECH 150.00 30.00 5.00

196 CONMACO C 1750 ECH 1050.00 175.00 6.00

204 VULCAN VUL 01 ECH 15.00 5.00 3.00

205 VULCAN VUL 02 ECH 7.26 3.00 2.42

206 VULCAN VUL 06 ECH 19.50 6.50 3.00

207 VULCAN VUL 08 ECH 26.00 8.00 3.25

208 VULCAN VUL 010 ECH 32.50 10.00 3.25

209 VULCAN VUL 012 ECH 39.00 12.00 3.25

210 VULCAN VUL 014 ECH 42.00 14.00 3.00

211 VULCAN VUL 016 ECH 48.75 16.25 3.00

212 VULCAN VUL 020 ECH 60.00 20.00 3.00

213 VULCAN VUL 030 ECH 90.00 30.00 3.00

214 VULCAN VUL 040 ECH 120.00 40.00 3.00
499

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

215 VULCAN VUL 060 ECH 180.00 60.00 3.00

220 VULCAN VUL 30C ECH 7.26 3.00 2.42

221 VULCAN VUL 50C ECH 15.10 5.00 3.02

222 VULCAN VUL 65C ECH 19.18 6.50 2.95

223 VULCAN VUL 65CA ECH 19.57 6.50 3.01

224 VULCAN VUL 80C ECH 24.48 8.00 3.06

225 VULCAN VUL 85C ECH 25.99 8.52 3.05

226 VULCAN VUL 100C ECH 32.90 10.00 3.29

227 VULCAN VUL 140C ECH 35.98 14.00 2.57

228 VULCAN VUL 200C ECH 50.20 20.00 2.51

229 VULCAN VUL 400C ECH 113.60 40.00 2.84

230 VULCAN VUL 600C ECH 179.16 60.00 2.99

231 VULCAN VUL 320 ECH 60.00 20.00 3.00

232 VULCAN VUL 330 ECH 90.00 30.00 3.00

233 VULCAN VUL 340 ECH 120.00 40.00 3.00

234 VULCAN VUL 360 ECH 180.00 60.00 3.00

235 VULCAN VUL 505 ECH 25.00 5.00 5.00

236 VULCAN VUL 506 ECH 32.50 6.50 5.00

237 VULCAN VUL 508 ECH 40.00 8.00 5.00

238 VULCAN VUL 510 ECH 50.00 10.00 5.00

239 VULCAN VUL 512 ECH 60.00 12.00 5.00

240 VULCAN VUL 520 ECH 100.00 20.00 5.00

241 VULCAN VUL 530 ECH 150.00 30.00 5.00

242 VULCAN VUL 540 ECH 200.00 40.90 4.89

243 VULCAN VUL 560 ECH 300.00 62.50 4.80

245 VULCAN VUL 3100 ECH 300.00 100.00 3.00

246 VULCAN VUL 5100 ECH 500.00 100.00 5.00

247 VULCAN VUL 5150 ECH 750.00 150.00 5.00

248 VULCAN VUL 6300 ECH 1800.00 300.00 6.00

251 RAYMOND R 1 ECH 15.00 5.00 3.00

252 RAYMOND R 1S ECH 19.50 6.50 3.00

253 RAYMOND R 65C ECH 19.50 6.50 3.00

254 RAYMOND R 65CH ECH 19.50 6.50 3.00

255 RAYMOND R 0 ECH 24.38 7.50 3.25

256 RAYMOND R 80C ECH 24.48 8.00 3.06

257 RAYMOND R 80CH ECH 24.48 8.00 3.06

258 RAYMOND R 2/0 ECH 32.50 10.00 3.25

259 RAYMOND R 3/0 ECH 40.63 12.50 3.25

260 RAYMOND R 150C ECH 48.75 15.00 3.25

261 RAYMOND R 4/0 ECH 48.75 15.00 3.25

262 RAYMOND R 5/0 ECH 56.88 17.50 3.25

263 RAYMOND R 30X ECH 75.00 30.00 2.50

264 RAYMOND R 8/0 ECH 81.25 25.00 3.25
500

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

265 RAYMOND R 40X ECH 100.00 40.00 2.50

266 RAYMOND R 60X ECH 150.00 60.00 2.50

270 MENCK MHU 100C ECH 73.71 11.10 6.64

271 MENCK MH 68 ECH 49.18 7.72 6.37

272 MENCK MH 96 ECH 69.43 11.02 6.30

273 MENCK MH 145 ECH 104.80 16.53 6.34

274 MENCK MHU 195 ECH 143.74 21.36 6.73

275 MENCK MHU 220 ECH 162.17 24.84 6.53

276 MENCK MHU 400 ECH 294.82 51.09 5.77

277 MENCK MHU 600 ECH 442.28 75.52 5.86

278 MENCK MHU 1000 ECH 737.38 126.98 5.81

279 MENCK MHU 1700 ECH 1253.24 207.15 6.05

280 MENCK MHU 2100 ECH 1548.29 257.18 6.02

281 MENCK MHU 3000 ECH 2211.90 370.23 5.97

282 MENCK MRBS 500 ECH 45.07 11.02 4.09

283 MENCK MRBS 750 ECH 67.77 16.53 4.10

285 MENCK MRBS 850 ECH 93.28 18.96 4.92

286 MENCK MRBS1100 ECH 123.43 24.25 5.09

287 MENCK MRBS1502 ECH 135.59 33.07 4.10

288 MENCK MRBS1800 ECH 189.81 38.58 4.92

289 MENCK MRBS2500 ECH 262.11 63.93 4.10

290 MENCK MRBS2502 ECH 225.95 55.11 4.10

291 MENCK MRBS2504 ECH 225.95 55.11 4.10

292 MENCK MRBS3000 ECH 325.36 66.13 4.92

293 MENCK MRBS3900 ECH 513.34 86.86 5.91

294 MENCK MRBS4600 ECH 498.94 101.41 4.92

295 MENCK MRBS5000 ECH 542.33 110.23 4.92

296 MENCK MRBS6000 ECH 759.23 132.27 5.74

297 MENCK MRBS7000 ECH 631.40 154.00 4.10

298 MENCK MRBS8000 ECH 867.74 176.37 4.92

299 MENCK MRBS8800 ECH 954.53 194.01 4.92

300 MENCK MBS12500 ECH 1581.83 275.58 5.74

301 MKT No. 5 ECH 1.00 0.20 5.00

302 MKT No. 6 ECH 2.50 0.40 6.25

303 MKT No. 7 ECH 4.15 0.80 5.19

304 MKT 9B3 ECH 8.75 1.60 5.47

305 MKT 10B3 ECH 13.11 3.00 4.37

306 MKT C5-Air ECH 14.20 5.00 2.84

307 MKT C5-Steam ECH 16.20 5.00 3.24

308 MKT S-5 ECH 16.25 5.00 3.25

309 MKT 11B3 ECH 19.15 5.00 3.83

310 MKT C826 Stm ECH 24.40 8.00 3.05

311 MKT C826 Air ECH 21.20 8.00 2.65
501

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

312 MKT S-8 ECH 26.00 8.00 3.25

313 MKT MS-350 ECH 30.80 7.72 3.99

314 MKT S 10 ECH 32.50 10.00 3.25

315 MKT S 14 ECH 37.52 14.00 2.68

316 MKT MS 500 ECH 44.00 11.00 4.00

317 MKT S 20 ECH 60.00 20.00 3.00

318 IHC S-30 ECH 21.70 3.53 6.15

319 IHC S-40 ECH 28.93 4.85 5.97

320 IHC S-35 ECH 25.53 6.63 3.85

321 IHC S-70 ECH 51.25 7.73 6.63

322 IHC S-90 ECH 65.90 9.94 6.63

323 IHC S-120 ECH 89.37 13.48 6.63

324 IHC S-150 ECH 110.06 16.60 6.63

325 IHC S-200 ECH 145.64 22.00 6.62

326 IHC S-280 ECH 205.31 30.06 6.83

327 IHC S-400 ECH 292.60 44.20 6.62

328 IHC S-500 ECH 366.09 55.30 6.62

329 IHC S-600 ECH 443.54 67.00 6.62

330 IHC S-900 ECH 658.36 99.45 6.62

331 IHC S-1200 ECH 891.05 134.60 6.62

332 IHC S-1800-L ECH 1170.39 166.00 7.05

333 IHC S-2300 ECH 1681.48 254.00 6.62

334 IHC S-2000 ECH 1473.97 222.65 6.62

335 IHC SC-30 ECH 21.81 3.76 5.80

336 IHC SC-40 ECH 29.86 5.51 5.42

337 IHC SC-50 ECH 36.82 7.29 5.05

338 IHC SC-60 ECH 44.95 13.30 3.38

339 IHC SC-75 ECH 54.80 12.15 4.51

340 IHC SC-110 ECH 81.89 17.46 4.69

341 IHC SC-150 ECH 109.35 24.30 4.50

342 IHC SC-200 ECH 152.51 30.20 5.05

343 IHC SC-250 ECH 179.80 37.26 4.83

344 IHC S-750 ECH 550.79 83.11 6.63

345 IHC S-800 ECH 589.97 88.15 6.69

346 IHC S-1400 ECH 1033.84 147.94 6.99

347 IHC S-1800 ECH 1340.21 195.64 6.85

348 IHC S-2500 ECH 1843.16 275.80 6.68

349 HERA 1900 OED 44.41 4.19 10.60

350 HERA 1250 OED 24.85 2.76 9.02

351 HERA 1500 OED 29.81 3.31 9.02

352 HERA 2500 OED 49.70 5.51 9.02

353 HERA 2800 OED 55.70 6.18 9.02

354 HERA 3500 OED 69.59 7.72 9.02
502

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

355 HERA 5000 OED 99.45 11.03 9.02

356 HERA 5700 OED 113.38 12.57 9.02

357 HERA 6200 OED 123.30 13.67 9.02

358 HERA 7500 OED 149.19 16.54 9.02

359 HERA 8800 OED 174.99 19.40 9.02

360 ICE I-12obs OED 30.21 2.82 10.71

361 ICE I-19obs OED 43.24 4.02 10.77

362 ICE I-30obs OED 71.45 6.62 10.80

363 ICE I-36obs OED 90.68 7.94 11.42

364 ICE I-46obs OED 107.74 10.15 10.62

365 ICE I-62obs OED 164.98 14.60 11.30

366 ICE I-80obs OED 212.40 17.70 12.00

367 ICE I-8v2obs OED 17.60 1.76 10.00

368 ICE I-100obs OED 264.45 23.61 11.20

369 BSP SL20 ECH 14.11 3.31 4.27

370 BSP SL30 ECH 21.69 5.51 3.94

371 FAIRCHLD F-45 ECH 45.00 15.00 3.00

372 FAIRCHLD F-32 ECH 32.55 10.85 3.00

374 BSP CX40 ECH 28.21 6.61 4.27

375 BSP CX50 ECH 37.61 8.82 4.27

376 BSP CX60 ECH 47.01 11.02 4.27

377 BSP CX75 ECH 52.08 13.23 3.94

378 BSP CX85 ECH 60.75 15.43 3.94

379 BSP CX110 ECH 78.11 19.84 3.94

381 BSP HH3 ECH 26.02 6.61 3.94

382 BSP HH5 ECH 43.38 11.02 3.94

383 BSP HH7 ECH 60.78 15.43 3.94

384 BSP HH8 ECH 69.50 17.64 3.94

385 BSP HH9 ECH 78.17 19.84 3.94

386 BSP HH11-1.2 ECH 95.55 24.25 3.94

387 BSP HH14-1.2 ECH 121.59 30.86 3.94

388 BSP HH16-1.2 ECH 138.87 35.27 3.94

391 BSP HA30 ECH 260.37 66.14 3.94

392 BSP HA40 ECH 347.16 88.18 3.94

393 BSP HH11-1.5 ECH 119.31 24.25 4.92

394 BSP HH14-1.5 ECH 151.83 30.86 4.92

395 BSP HH16-1.5 ECH 173.54 35.27 4.92

396 BSP CG180 ECH 131.92 26.45 4.99

397 BSP CG210 ECH 153.91 30.86 4.99

398 BSP CG240 ECH 175.90 35.27 4.99

399 BSP CG270 ECH 197.88 39.68 4.99

400 BSP CG300 ECH 219.87 44.09 4.99

401 BERMINGH B23 CED 22.99 2.80 8.21
503

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

402 BERMINGH B200 OED 18.00 2.00 9.00

403 BERMINGH B225 OED 29.25 3.00 9.75

404 BERMINGH B300 OED 40.31 3.75 10.75

405 BERMINGH B400 OED 53.75 5.00 10.75

406 BERMINGH B-21 OED 53.25 4.63 11.50

410 BERMINGH B300 M OED 40.31 3.75 10.75

411 BERMINGH B400 M OED 53.75 5.00 10.75

412 BERMINGH B400 4.8 OED 43.20 4.80 9.00

413 BERMINGH B400 5.0 OED 45.00 5.00 9.00

414 BERMINGH B23 5 CED 22.99 2.80 8.21

415 BERMINGH B250 5 OED 26.25 2.50 10.50

416 BERMINGH B3505 OED 47.20 4.00 11.80

417 BERMINGH B4005 OED 59.00 5.00 11.80

418 BERMINGH B4505 OED 77.88 6.60 11.80

419 BERMINGH B5005 OED 92.04 7.80 11.80

420 BERMINGH B5505 OED 108.56 9.20 11.80

421 BERMINGH B550 C OED 88.00 11.00 8.00

422 BERMINGH B2005 OED 18.00 2.00 9.00

424 BERMINGH B2505 OED 35.40 3.00 11.80

425 BERMINGH B3005 OED 35.40 3.00 11.80

431 BERMINGH B6005 OED 160.95 13.64 11.80

432 BERMINGH B6505 C OED 253.00 22.00 11.50

433 BERMINGH B6505 OED 202.86 17.64 11.50

434 BERMINGH B-9 OED 21.00 2.00 10.50

435 BERMINGH B-32 OED 81.08 7.05 11.50

436 BERMINGH B-64 OED 166.50 14.11 11.80

437 BERMINGH B-6505HD OED 220.50 22.05 10.00

441 MENCK MHF5-5 ECH 38.69 11.02 3.51

442 MENCK MHF5-6 ECH 46.43 13.23 3.51

443 MENCK MHF5-7 ECH 54.17 15.43 3.51

444 MENCK MHF5-8 ECH 61.91 17.64 3.51

445 MENCK MHF5-9 ECH 69.65 19.84 3.51

446 MENCK MHF5-10 ECH 77.39 22.05 3.51

447 MENCK MHF5-11 ECH 85.13 24.25 3.51

448 MENCK MHF5-12 ECH 92.87 26.45 3.51

449 MENCK MHF3-3 ECH 24.76 7.05 3.51

450 MENCK MHF3-4 ECH 30.96 8.82 3.51

451 MENCK MHF3-5 ECH 38.69 11.02 3.51

452 MENCK MHF3-6 ECH 46.43 13.23 3.51

453 MENCK MHF3-7 ECH 54.17 15.43 3.51

454 MENCK MHF10-15 ECH 124.73 33.06 3.77

455 MENCK MHF10-20 ECH 166.28 44.07 3.77

456 MENCK MHF 5-14 ECH 108.34 30.86 3.51
504

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

457 MENCK MHU135T* ECH 110.59 17.99 6.15

458 MENCK MHU500T* ECH 368.74 65.96 5.59

459 MENCK MHU 300S ECH 221.20 35.73 6.19

460 MENCK MHU 270T ECH 221.20 35.73 6.19

461 MENCK MHU 200T ECH 162.24 26.75 6.07

462 MENCK MHU 400T ECH 324.37 52.45 6.18

463 MENCK MHU 500T ECH 405.53 65.96 6.15

464 MENCK MHU 700T ECH 567.72 92.88 6.11

465 MENCK MHU 840S ECH 619.22 92.88 6.67

466 MENCK MHU 600B ECH 457.03 65.96 6.93

467 MENCK MHU 600T ECH 486.63 80.39 6.05

468 MENCK MHU 800S ECH 604.57 99.93 6.05

469 MENCK MHU1200S ECH 884.84 145.71 6.07

470 MENCK MHU1500S ECH 1106.07 178.94 6.18

471 MENCK MHU1700T ECH 1400.86 227.36 6.16

472 MENCK MHU1900S ECH 1400.86 227.36 6.16

473 MENCK MHU 150S ECH 110.59 17.99 6.15

474 MENCK MHU2700S ECH 1990.19 318.77 6.24

475 MENCK MHU 135T ECH 110.59 17.99 6.15

476 MENCK MHU 750T ECH 604.57 99.93 6.05

477 MENCK MHU1100T ECH 899.66 145.71 6.18

478 MENCK MHU150S* ECH 110.59 17.99 6.15

479 MENCK MHU600B* ECH 457.03 65.96 6.93

481 JUNTTAN HHK3A ECH 26.05 6.62 3.94

482 JUNTTAN HHK4A ECH 34.73 8.82 3.94

483 JUNTTAN HHK5A ECH 43.41 11.03 3.94

484 JUNTTAN HHK6A ECH 52.10 13.23 3.94

485 JUNTTAN HHK7A ECH 60.75 15.43 3.94

486 JUNTTAN HHK10A ECH 86.83 22.05 3.94

487 JUNTTAN HHK12A ECH 104.19 26.47 3.94

488 JUNTTAN HHK14A ECH 121.56 30.88 3.94

491 JUNTTAN HHK9A ECH 78.14 19.85 3.94

494 JUNTTAN HHK16A ECH 138.92 35.29 3.94

495 JUNTTAN HHK18A ECH 156.29 39.70 3.94

496 JUNTTAN HHK20A ECH 173.65 44.11 3.94

497 JUNTTAN HHK4SL ECH 43.40 8.82 4.92

498 JUNTTAN HHK3AL ECH 17.37 6.62 2.63

499 JUNTTAN HHK4AL ECH 23.15 8.82 2.63

500 JUNTTAN HHK5AL ECH 28.94 11.03 2.63

501 HPSI 110 ECH 44.00 11.00 4.00

502 HPSI 150 ECH 60.00 15.00 4.00

503 HPSI 154 ECH 61.60 15.40 4.00

504 HPSI 200 ECH 80.00 20.00 4.00
505

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

505 HPSI 225 ECH 90.00 22.50 4.00

506 HPSI 650 ECH 32.50 6.50 5.00

507 HPSI 1000 ECH 50.00 10.00 5.00

508 HPSI 1605 ECH 83.00 16.60 5.00

509 HPSI 2005 ECH 95.10 19.02 5.00

510 HPSI 3005 ECH 154.33 30.87 5.00

511 HPSI 3505 ECH 176.33 35.27 5.00

512 HPSI 2000 ECH 80.00 20.00 4.00

514 UDDCOMB H2H ECH 16.62 4.40 3.77

515 UDDCOMB H3H ECH 24.88 6.60 3.77

516 UDDCOMB H4H ECH 33.18 8.80 3.77

517 UDDCOMB H5H ECH 41.47 11.00 3.77

518 UDDCOMB H6H ECH 49.76 13.20 3.77

519 UDDCOMB H8H ECH 82.19 17.60 4.67

520 UDDCOMB H10H ECH 86.88 22.05 3.94

521 DAWSON HPH1200 ECH 8.72 2.30 3.79

522 DAWSON HPH1800 ECH 13.72 3.30 4.16

523 DAWSON HPH2400 ECH 17.32 4.19 4.13

524 DAWSON HPH6500 ECH 46.98 10.25 4.58

525 DAWSON HPH4500 ECH 32.56 7.72 4.22

526 DAWSON HPH9000 ECH 66.30 10.47 6.33

530 BRUCE SGH-0312 ECH 26.00 6.60 3.94

531 BRUCE SGH-0512 ECH 43.34 11.00 3.94

532 BRUCE SGH-0712 ECH 60.68 15.40 3.94

533 BRUCE SGH-1012 ECH 86.77 22.05 3.94

534 BRUCE SGH-0412 ECH 34.67 8.80 3.94

535 BANUT S3000 ECH 26.04 6.62 3.94

536 BANUT S4000 ECH 34.72 8.82 3.94

537 BANUT S5000 ECH 43.41 11.03 3.94

538 BANUT S6000 ECH 52.09 13.23 3.94

539 BANUT S8000 ECH 69.45 17.64 3.94

540 BANUT S10000 ECH 86.81 22.05 3.94

541 BANUT 3 Tonnes ECH 17.35 6.61 2.62

542 BANUT 4 Tonnes ECH 23.14 8.82 2.62

543 BANUT 5 Tonnes ECH 28.92 11.02 2.62

544 BANUT 6 Tonnes ECH 34.72 13.23 2.62

545 BANUT 7 Tonnes ECH 40.49 15.43 2.62

550 ICE 70 ECH 21.00 7.00 3.00

551 ICE 75 ECH 30.00 7.50 4.00

552 ICE 110-SH ECH 37.72 11.50 3.28

553 ICE 115-SH ECH 37.95 11.50 3.30

554 ICE 115 ECH 46.00 11.50 4.00

555 ICE 160-SH ECH 64.00 16.00 4.00
506

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

556 ICE 160 ECH 64.00 16.00 4.00

557 ICE 220 ECH 88.00 22.00 4.00

558 ICE 275 ECH 110.00 27.50 4.00

559 ICE DKH-3U ECH 26.00 6.60 3.94

560 HMC 28A ECH 28.00 7.00 4.00

561 HMC 28B ECH 21.00 7.00 3.00

562 HMC 62 ECH 46.00 11.50 4.00

563 HMC 86 ECH 64.00 16.00 4.00

564 HMC 119 ECH 88.00 22.00 4.00

565 HMC 149 ECH 110.00 27.50 4.00

566 HMC 187 ECH 138.00 34.50 4.00

567 HMC 19D ECH 14.00 3.50 4.00

568 HMC 38D ECH 28.00 7.00 4.00

569 APE D 8-42 OED 19.80 1.76 11.25

570 APE D 1-42 OED 1.32 0.21 6.33

571 APE D 19-42 OED 47.13 4.19 11.25

572 APE D 30-42 OED 74.42 6.62 11.25

573 APE D 36-42 OED 89.30 7.94 11.25

574 APE D 46-42 OED 114.11 10.14 11.25

575 APE D 62-42 OED 153.80 13.67 11.25

576 APE D 80-42 OED 198.45 17.64 11.25

577 APE D 100-42 OED 248.06 22.05 11.25

579 APE D 16-42 OED 39.69 3.53 11.25

580 APE D 16-52 OED 39.69 3.53 11.25

581 APE D 25-42 OED 62.01 5.51 11.25

582 APE D 125-42 OED 310.08 27.56 11.25

583 APE D 50-42 OED 124.03 11.03 11.25

584 APE D 12-42 OED 29.77 2.65 11.25

585 APE D 36-26 OED 89.30 7.94 11.25

586 APE D 128-42 OED 317.25 28.20 11.25

587 APE D 138-42 OED 342.00 30.40 11.25

588 APE D 160-42 OED 396.90 35.28 11.25

589 APE D 180-42 OED 446.51 39.69 11.25

590 APE D 225-42 OED 558.00 49.60 11.25

591 APE 5.4mT ECH 26.00 12.00 2.17

592 APE 7.2mT ECH 51.30 16.20 3.17

593 APE D 220-42 OED 540.81 48.46 11.16

594 APE 15-60 ECH 150.00 30.00 5.00

595 APE 10-60 ECH 100.00 20.00 5.00

596 APE 400U ECH 400.00 80.00 5.00

598 APE 750U ECH 750.00 120.00 6.25

599 APE D 100-13 OED 300.04 23.70 12.66

600 BSP DX20 ECH 14.11 3.31 4.27
507

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

601 BSP DX25 ECH 18.09 4.41 4.10

602 BSP DX30 ECH 21.71 5.51 3.94

603 BSP LX2.5-SA ECH 14.47 5.51 2.63

604 BSP LX4-SA ECH 23.15 8.82 2.63

605 BSP LX5-SA ECH 28.94 11.03 2.63

606 BSP CGL370 ECH 271.22 55.11 4.92

607 BSP CGL440 ECH 325.47 66.14 4.92

608 BSP CGL520 ECH 379.71 77.16 4.92

609 BSP CGL590 ECH 433.96 88.18 4.92

610 BSP LX7-SA ECH 40.52 15.44 2.63

625 BRUCE SGH-1212 ECH 104.13 26.46 3.94

626 BRUCE SGH-1312 ECH 112.81 28.66 3.94

627 BRUCE SGH-1315 ECH 141.01 28.66 4.92

628 BRUCE SGH-1412 ECH 121.48 30.87 3.94

629 BRUCE SGH-1415 ECH 151.85 30.87 4.92

630 BRUCE SGH-1612 ECH 138.84 35.27 3.94

631 BRUCE SGH-1615 ECH 173.55 35.27 4.92

632 BRUCE SGH-1618 ECH 208.26 35.27 5.90

633 BRUCE SGH-1619 ECH 219.83 35.27 6.23

634 BRUCE SGH-1812 ECH 156.19 39.68 3.94

635 BRUCE SGH-1815 ECH 195.24 39.68 4.92

636 BRUCE SGH-2012 ECH 173.55 44.09 3.94

637 BRUCE SGH-2015 ECH 216.94 44.09 4.92

638 BRUCE SGH-2312 ECH 199.58 50.71 3.94

639 BRUCE SGH-2315 ECH 249.48 50.71 4.92

640 BRUCE SGH-3012 ECH 260.32 66.14 3.94

641 BRUCE SGH-3013 ECH 282.02 66.14 4.26

642 BRUCE SGH-3015 ECH 325.40 66.14 4.92

643 BRUCE SGH-4012 ECH 347.10 88.19 3.94

644 BRUCE SGH-4212 ECH 364.45 92.59 3.94

645 BRUCE SGH-5012 ECH 433.87 110.23 3.94

650 Twinwood V20B ECH 35.58 9.04 3.94

651 Twinwood V100D ECH 87.66 22.27 3.94

652 Twinwood V160B ECH 140.58 35.71 3.94

653 Twinwood V400A ECH 263.84 67.02 3.94

656 Pilemast 24-750 ECH 1.50 0.75 2.00

657 Pilemast 24-900 ECH 1.80 0.90 2.00

658 Pilemast 24-2000 ECH 4.00 2.00 2.00

659 Pilemast 24-2500 ECH 5.00 2.50 2.00

660 Pilemast 36-3000 ECH 9.00 3.00 3.00

661 Pilemast 36-5000 ECH 15.00 5.00 3.00

669 MVE M-12 OED 30.21 2.82 10.71

670 MVE M-19 OED 49.38 4.02 12.30
508

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

671 MVE M-30 OED 83.35 6.62 12.60

801 DKH PH-5 ECH 43.40 11.02 3.94

802 DKH PH-7 ECH 60.75 15.43 3.94

803 DKH PH-7S ECH 60.75 15.43 3.94

804 DKH PH-10 ECH 86.79 22.05 3.94

805 DKH PH-13 ECH 112.83 28.66 3.94

806 DKH PH-20 ECH 216.98 44.09 4.92

807 DKH PH-30 ECH 325.47 66.14 4.92

808 DKH PH-40 ECH 433.96 88.18 4.92

809 DKH DKH-713 ECH 112.92 28.66 3.94

850 PILECO D8-22 OED 18.66 1.76 10.60

851 PILECO D12-42 OED 29.89 2.82 10.60

852 PILECO D19-42 OED 42.51 4.01 10.60

853 PILECO D25-32 OED 58.41 5.51 10.60

854 PILECO D30-32 OED 70.07 6.61 10.60

855 PILECO D36-32 OED 84.16 7.94 10.60

856 PILECO D46-32 OED 107.48 10.14 10.60

857 PILECO D62-22 OED 161.31 13.67 11.80

858 PILECO D80-23 OED 197.57 17.64 11.20

859 PILECO D100-13 OED 246.85 22.04 11.20

860 PILECO D125-32 OED 308.67 27.56 11.20

861 PILECO D225-22 OED 555.34 49.58 11.20

862 PILECO D250-22 OED 617.06 55.09 11.20

863 PILECO D138-32 OED 340.61 30.41 11.20

864 PILECO D180-32 OED 444.27 39.67 11.20

865 PILECO D280-22 OED 688.55 61.73 11.16

866 PILECO D160-32 OED 395.08 35.28 11.20

867 PILECO D400-12 OED 810.10 88.15 9.19

868 PILECO D600-12 OED 1215.10 132.22 9.19

869 PILECO D800-22 OED 1620.20 176.30 9.19

921 BRUCE SGH-0212 ECH 17.34 4.40 3.94

922 BRUCE SGH-0715 ECH 75.77 15.40 4.92

923 BRUCE SGH-1015 ECH 108.47 22.05 4.92

924 BRUCE SGH-1215 ECH 130.16 26.46 4.92

925 BRUCE SGH-2512 ECH 216.94 55.12 3.94

926 BRUCE SGH-2515 ECH 271.17 55.12 4.92

927 BRUCE SGH-3512 ECH 303.71 77.16 3.94

928 BRUCE SGH-3515 ECH 379.64 77.16 4.92

929 BRUCE SGH-4015 ECH 433.87 88.19 4.92

930 BRUCE SGH-4215 ECH 455.56 92.59 4.92

931 BRUCE SGH-4512 ECH 390.48 99.21 3.94

932 BRUCE SGH-4515 ECH 488.10 99.21 4.92

933 BRUCE SGH-4712 ECH 407.84 103.62 3.94
509

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

934 BRUCE SGH-4715 ECH 509.80 103.62 4.92

935 BRUCE SGH-4719 ECH 645.74 103.62 6.23

936 BRUCE SGH-5015 ECH 542.34 110.23 4.92

937 BRUCE SGH-5715 ECH 618.26 125.66 4.92

938 BRUCE SGH-6015 ECH 650.80 132.28 4.92

939 BRUCE SGH-7015 ECH 759.27 154.32 4.92

940 BRUCE SGH-8015 ECH 867.74 176.37 4.92

949 JUNTTAN HHK10S ECH 108.49 22.05 4.92

950 JUNTTAN HHK28S ECH 303.78 61.73 4.92

951 JUNTTAN HHK5S ECH 54.27 11.03 4.92

952 JUNTTAN HHK7S ECH 75.97 15.44 4.92

953 JUNTTAN HHK9S ECH 97.68 19.85 4.92

954 JUNTTAN HHK12S ECH 130.24 26.47 4.92

955 JUNTTAN HHK14S ECH 151.95 30.88 4.92

956 JUNTTAN HHK16S ECH 173.65 35.29 4.92

957 JUNTTAN HHK18S ECH 195.36 39.70 4.92

958 JUNTTAN HHK20S ECH 217.07 44.11 4.92

959 JUNTTAN HHK25S ECH 271.22 55.11 4.92

960 JUNTTAN HHK36S ECH 390.56 79.36 4.92

961 JUNTTAN HHU5A ECH 54.27 11.03 4.92

962 JUNTTAN HHU7A ECH 75.94 15.43 4.92

963 JUNTTAN HHU9A ECH 97.64 19.84 4.92

964 JUNTTAN HHU12A ECH 130.19 26.45 4.92

965 JUNTTAN HHU14A ECH 151.88 30.86 4.92

966 JUNTTAN HHU16A ECH 173.58 35.27 4.92

968 JUNTTAN SHK100-3 ECH 26.91 6.61 4.07

969 JUNTTAN SHK100-3 ECH 35.89 8.82 4.07

970 JUNTTAN SHK100-3 ECH 44.84 11.02 4.07

971 JUNTTAN SHK100-3 ECH 53.82 13.23 4.07

972 JUNTTAN SHK110-5 ECH 44.98 11.02 4.08

973 JUNTTAN SHK110-5 ECH 53.82 13.23 4.07

974 JUNTTAN SHK110-5 ECH 65.62 15.43 4.25

975 JUNTTAN SHK110-5 ECH 77.42 17.64 4.39

976 JUNTTAN SHK110-5 ECH 87.74 19.84 4.42

977 JUNTTAN SHK100-5 ECH 44.84 11.02 4.07

978 JUNTTAN SHK100-5 ECH 53.82 13.23 4.07

979 JUNTTAN SHK110-7 ECH 65.63 15.43 4.25

980 JUNTTAN SHK110-7 ECH 77.43 17.64 4.39

981 JUNTTAN SHK110-7 ECH 87.74 19.84 4.42

998 HYPOTHET EX 4 OED 23.38 2.75 8.50

999 SELF Drop/10t ECH 300.00 20.00 15.00

1001 DFI-Corp HHA250-4 ECH 25.18 5.51 4.57

1002 DFI-Corp HHA300-4 ECH 28.75 6.61 4.35
510

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

1003 DFI-Corp HHA325-4 ECH 30.36 7.16 4.24

1004 DFI-Corp HHA350-4 ECH 31.80 7.70 4.13

1005 DFI-Corp HHA400-6 ECH 51.92 8.80 5.90

1006 DFI-Corp HHA450-6 ECH 57.04 9.92 5.75

1007 DFI-Corp HHB500-6 ECH 66.66 11.00 6.06

1008 DFI-Corp HHB600-6 ECH 77.88 13.20 5.90

1020 J&M 70B HIH ECH 21.00 7.00 3.00

1021 J&M 82 HIH ECH 32.80 8.20 4.00

1022 J&M 115 HIH ECH 46.00 11.50 4.00

1023 J&M 160 HIH ECH 64.00 16.00 4.00

1024 J&M 220 HIH ECH 88.00 22.00 4.00

1025 J&M 275 HIH ECH 110.00 27.50 4.00

1026 J&M 345 HIH ECH 138.00 34.50 4.00

1134 Pilemer DKH-3U ECH 26.00 6.60 3.94

1135 Pilemer DKH 10L ECH 86.79 22.05 3.94

1201 Liebherr H 50/3 ECH 28.97 6.60 4.39

1202 Liebherr H 50/4 ECH 35.02 8.80 3.98

1203 Liebherr H 85/5 ECH 43.34 11.00 3.94

1204 Liebherr H 85/7 ECH 60.16 15.43 3.90

1205 Liebherr H 110/7 ECH 60.16 15.43 3.90

1206 Liebherr H 110/9 ECH 78.01 19.85 3.93

1251 ICE I-30 V2 OED 71.71 6.62 10.84

1261 APE D 19-52 OED 47.13 4.19 11.25

1262 APE D 16-32 OED 39.69 3.53 11.25

1263 APE D 19-32 OED 47.13 4.19 11.25

1264 APE D 25-32 OED 62.01 5.51 11.25

1265 APE D 30-32 OED 74.42 6.62 11.25

1266 APE D 36-32 OED 89.30 7.94 11.25

1267 APE D 46-32 OED 114.11 10.14 11.25

1268 APE D 62-22 OED 153.80 13.67 11.25

1269 APE D 80-23 OED 198.45 17.64 11.25

1270 APE D 100-32 OED 248.06 22.05 11.25

1271 APE D 120-32 OED 349.69 27.60 12.67

1272 APE D 70-42 OED 173.64 15.44 11.25

1273 APE D 25-52 OED 62.01 5.51 11.25

1274 APE D 30-52 OED 74.42 6.62 11.25

1275 APE D 36-52 OED 89.30 7.94 11.25

1276 APE D 46-52 OED 114.11 10.14 11.25

1277 APE D 50-52 OED 124.03 11.03 11.25

1278 APE D 62-52 OED 153.80 13.67 11.25

1279 APE D 70-52 OED 173.64 15.44 11.25

1280 APE 7.5a ECH 24.00 12.00 2.00

1281 APE 7.5b ECH 20.40 10.20 2.00
511

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

1282 APE 7.5c ECH 15.20 7.60 2.00

1283 APE 9.5a ECH 50.66 16.00 3.17

1284 APE 9.5b ECH 44.32 14.00 3.17

1285 APE 7-3 ECH 42.00 14.00 3.00

1286 APE 8-3 ECH 48.00 16.00 3.00

1287 APE 8 ECH 16.00 8.00 2.00

1288 APE 8a ECH 24.00 12.00 2.00

1289 APE 10-4 ECH 80.00 20.00 4.00

1321 MENCK MHU 240U ECH 221.20 35.73 6.19

1322 MENCK MHU 440S ECH 324.37 52.45 6.18

1323 MENCK MHU 360U ECH 324.37 52.45 6.18

1324 MENCK MHU 550S ECH 404.06 65.96 6.13

1325 MENCK MHU 450U ECH 404.06 65.96 6.13

1326 MENCK MHU 660S ECH 485.17 80.39 6.04

1327 MENCK MHU 540U ECH 485.17 80.39 6.04

1328 MENCK MHU 720T ECH 588.19 99.93 5.89

1329 MENCK MHU 650U ECH 588.19 99.93 5.89

1330 MENCK MHU1000S ECH 736.48 126.98 5.80

1331 MENCK MHU 900T ECH 736.48 126.98 5.80

1332 MENCK MHU 810U ECH 736.48 126.98 5.80

1333 MENCK MHU1700S ECH 1272.95 207.15 6.15

1334 MENCK MHU2100S ECH 1573.92 257.18 6.12

1335 MENCK MHU3000S ECH 2216.56 370.23 5.99

1336 MENCK MHU1400B ECH 1032.08 145.71 7.08

1337 MENCK MHU3500S ECH 2582.43 385.85 6.69

1371 IHC S-3000 ECH 2211.93 332.44 6.65

1372 IHC S-4000 ECH 2948.91 444.30 6.64

1401 FAMBO HR250 ECH 1.81 0.55 3.28

1402 FAMBO HR500akk ECH 3.62 1.10 3.28

1403 FAMBO HR500 ECH 4.34 1.10 3.94

1404 FAMBO HR1000 ECH 8.68 2.20 3.94

1405 FAMBO HR1500 ECH 13.02 3.31 3.94

1406 FAMBO HR2000 ECH 17.36 4.41 3.94

1407 FAMBO HR2750 ECH 23.87 6.06 3.94

1408 FAMBO HR3000 ECH 26.04 6.61 3.94

1409 FAMBO HR4000 ECH 34.72 8.82 3.94

1410 FAMBO HR5000 ECH 43.40 11.02 3.94

1411 FAMBO HR7000 ECH 60.75 15.43 3.94

1412 FAMBO HR8000 ECH 69.45 17.64 3.94

1413 FAMBO HR10000 ECH 86.79 22.05 3.94

1501 ICE I-12v2 OED 29.63 2.82 10.50

1502 ICE I-8v2 OED 18.69 1.76 10.60

1503 ICE I-19v2 OED 46.14 4.01 11.50
512

Type Hammer 

Make

Hammer 

Model

Hammer 

Type

Rated Energy 

kip-feet

Ram Weight 

kips

Stroke 

feet

1504 ICE I-30v2 OED 76.05 6.61 11.50

1505 ICE I-36v2 OED 93.73 7.94 11.81

1506 ICE I-46v2 OED 119.77 10.14 11.81

1507 ICE I-62v2 OED 172.37 14.59 11.81

1508 ICE I-80v2 OED 208.30 17.64 11.81

1509 ICE I-100v2 OED 260.37 22.05 11.81

1510 ICE I-125v2 OED 310.10 27.56 11.25

1511 ICE I-160v2 OED 393.45 35.27 11.16

1512 ICE IP-2 ECH 17.36 4.41 3.94

1513 ICE IP-3 ECH 26.04 6.61 3.94

1514 ICE IP-5 ECH 43.40 11.02 3.94

1515 ICE IP-7 ECH 60.75 15.43 3.94

1516 ICE IP-10 ECH 86.78 22.04 3.94

1517 ICE IP-13 ECH 112.83 28.66 3.94

1520 ICE I-138v2 OED 328.62 30.40 10.81

1531 SPI D 19-42 OED 42.61 4.02 10.60

1532 SPI D 30-32 OED 72.08 6.80 10.60

1601 DELMAG D 2 OED 1.78 0.49 3.61

1602 DELMAG D 4 OED 3.60 0.84 4.30

1603 DELMAG D 8-12 OED 20.10 1.76 11.42

1604 DELMAG D 12-52 OED 33.98 2.82 12.05

1605 DELMAG D 16-52 OED 40.20 3.52 11.42

1606 DELMAG D 25-52 OED 66.34 5.51 12.04

1607 DELMAG D 30-52 OED 75.44 6.60 11.43

1611 DELMAG D138-32 OED 339.51 30.44 11.16

1612 DELMAG D180-32 OED 442.64 39.68 11.16

1613 DELMAG D300-32 OED 737.73 66.14 11.16

1614 DELMAG D400-32 OED 983.64 88.18 11.16

1620 HMC TD19 OED 46.09 4.01 11.49

1621 HMC TD30 OED 69.87 6.61 10.57
513

Type Hammer 

Make 

Hammer 

Model

Hammer 

Type

Rated Power 

kW 

Ecc. Mass

kips 

Frequency

Hz 

620 MAIT 34 VIB 227.00 1.23 33.30

621 MAIT 42 VIB 309.00 1.52 33.30

622 MAIT 54 VIB 450.00 0.98 33.30

623 MAIT 68 VIB 531.00 1.23 33.30

624 MAIT 120 VIB 674.00 1.74 30.00

698 ICE 50B VIB 432.00 10.42 26.70

699 ICE 3117 VIB 235.00 1.12 28.30

700 ICE 23-28 VIB 21.00 0.10 26.70

701 ICE 216 VIB 130.00 0.46 26.70

702 ICE 216E VIB 130.00 0.46 26.70

703 ICE 11-23 VIB 164.00 0.46 31.70

704 ICE 223 VIB 242.00 0.46 38.30

705 ICE 416L VIB 242.00 0.92 26.70

706 ICE 812 VIB 375.00 1.82 26.70

707 ICE 815 VIB 375.00 1.84 26.70

708 ICE 44-30 VIB 242.00 1.30 20.00

709 ICE 44-50 VIB 377.00 1.30 26.70

710 ICE 44-65 VIB 485.00 1.30 27.50

711 ICE 66-65 VIB 485.00 1.95 21.70

712 ICE 66-80 VIB 597.00 1.95 26.70

713 ICE 1412B VIB 597.00 2.04 21.00

714 ICE 1412C VIB 470.00 2.02 23.00

715 ICE V125 VIB 984.00 1.04 25.80

716 ICE 14RF VIB 242.00 1.01 38.30

717 ICE 14-23 VIB 164.00 1.17 35.00

718 ICE 22-23V VIB 164.00 0.92 26.90

719 ICE 22-30 VIB 250.00 0.92 26.90

720 HMC 3+28 VIB 21.00 0.11 26.80

721 HMC 3+75 VIB 56.00 0.11 36.10

722 HMC 13+200 VIB 149.00 0.35 26.70

723 HMC 13S+200 VIB 149.00 0.35 26.70

724 HMC 13H+200 VIB 164.00 0.35 29.80

725 HMC 25+220 VIB 164.00 0.61 20.90

726 HMC 26+335 VIB 242.00 0.71 25.60

727 HMC 26S+335 VIB 242.00 0.71 25.60

728 HMC 51+335 VIB 242.00 1.21 19.50

729 HMC 51+535 VIB 377.00 1.21 26.40

730 HMC 51S+535 VIB 377.00 1.21 26.40

731 HMC 51+740 VIB 485.00 1.21 27.50

732 HMC 76+740 VIB 485.00 1.82 21.70

733 HMC 76+800 VIB 597.00 1.82 26.10

734 HMC 115+800 VIB 597.00 1.35 20.40

735 HMC 230+1600 VIB 1193.00 2.69 20.40
514

Type Hammer 

Make 

Hammer 

Model

Hammer 

Type

Rated Power 

kW 

Ecc. Mass

kips 

Frequency

Hz 

750 MKT V-2B VIB 52.00 0.15 30.00

751 MKT V-5C VIB 138.00 0.43 28.33

752 MKT V-20B VIB 242.00 0.20 28.33

753 MKT V-30 VIB 448.00 1.47 28.33

754 MKT V-35 VIB 485.00 1.60 28.33

755 MKT V-140 VIB 1341.00 1.17 23.33

770 APE 3 VIB 10.58 0.00 38.30

771 APE 6 VIB 10.58 0.01 38.30

772 APE 15 VIB 59.67 0.11 30.00

773 APE 20 VIB 59.67 0.15 38.30

774 APE 20E VIB 59.67 0.15 38.30

775 APE 50 VIB 194.00 0.23 30.00

776 APE 50E VIB 194.00 0.23 30.00

777 APE 100 VIB 194.00 0.32 30.00

778 APE 100E VIB 194.00 0.14 30.00

779 APE 100HF VIB 260.00 0.14 43.00

780 APE 150 VIB 260.00 0.14 30.00

781 APE 150T VIB 260.00 0.17 30.00

782 APE 150HF VIB 466.00 0.32 43.00

783 APE 200 VIB 466.00 0.29 30.00

784 APE 200T VIB 466.00 0.34 30.83

785 APE 200T HF VIB 738.00 0.34 43.00

786 APE 300 VIB 738.00 0.34 25.00

787 APE 400B VIB 738.00 0.78 23.33

788 APE 600 VIB 800.00 1.05 23.30

789 APE Tan 400 VIB 1476.00 1.37 23.33

790 APE Tan 600 VIB 1800.00 2.11 23.30

791 APE 200-6 VIB 470.00 0.43 30.00

811 MGF RBH 80 VIB 50.00 0.60 30.00

812 MGF RBH 140 VIB 85.00 1.04 26.67

813 MGF RBH 200 VIB 125.00 0.74 26.67

814 MGF RBH 320 VIB 200.00 0.79 26.67

815 MGF RBH 460 VIB 255.00 1.13 26.67

816 MGF RBH 1050 VIB 460.00 1.55 22.50

817 MGF RBH 1575 VIB 700.00 1.16 22.50

818 MGF RBH 2400 VIB 975.00 1.77 23.50

880 ICE 23RF VIB 384.00 0.83 38.30

881 ICE 1412BT VIB 1193.00 1.67 21.70

882 ICE 23-40 VIB 30.00 0.19 31.80

883 ICE 28-35 VIB 261.00 1.16 27.30

884 ICE 28RF-35 VIB 261.00 1.16 27.30

885 ICE V360 VIB 783.00 0.94 25.00

886 ICE V360 T VIB 1566.00 1.88 25.00
515

Type Hammer 

Make 

Hammer 

Model

Hammer 

Type

Rated Power 

kW 

Ecc. Mass

kips 

Frequency

Hz 

887 ICE 44-30V VIB 250.00 0.92 26.00

888 ICE 44-70 VIB 585.00 0.92 28.10

889 ICE 66-70 VIB 585.00 0.92 23.00

890 ICE 7RF VIB 154.00 0.51 38.30

891 ICE 66-70HS VIB 585.00 0.92 26.70

892 ICE 66-80HS VIB 597.00 0.92 29.20

893 ICE 100c-Tdm VIB 1774.00 1.83 26.67

894 ICE 423 VIB 377.00 0.92 38.30

895 ICE 32RF VIB 391.00 1.16 33.30

896 ICE 36RF VIB 431.00 1.30 33.30

897 ICE 46RF VIB 678.00 1.66 38.30

898 ICE 64RF VIB 663.00 1.16 32.50

899 ICE 44B VIB 595.00 1.30 30.00

900 Mueller MS16HF VIB 219.00 1.16 39.20

901 Mueller MS25H2 VIB 218.00 0.90 28.00

902 Mueller MS25H3 VIB 218.00 0.90 28.00

903 Mueller MS50H2 VIB 419.00 1.20 27.00

904 Mueller MS50H3 VIB 419.00 1.20 27.00

905 Mueller MS25HHF VIB 274.00 0.58 27.30

906 Mueller MS50HHF VIB 562.00 1.17 27.30

907 Mueller MS100HHF VIB 750.00 2.33 24.90

908 Mueller MS120HHF VIB 895.00 2.30 25.60

909 Mueller MS200HHF VIB 837.00 4.25 22.90

910 Mueller MS-10HFV VIB 203.00 0.39 39.30

911 Mueller MS-16HFV VIB 294.00 0.53 39.20

912 Mueller MS-24HFV VIB 720.00 0.85 39.20

913 Mueller MS-32HFV VIB 551.00 1.05 39.60

914 Mueller MS-48HFV VIB 823.00 1.69 39.20

915 Mueller MS-62HFV VIB 735.00 1.82 35.00

1039 J&M 11-23 VIB 164.00 0.92 31.70

1040 J&M 1412 VIB 559.00 1.67 21.70

1041 J&M 1412T VIB 1119.00 1.67 21.70

1042 J&M 216 VIB 149.00 0.92 26.70

1044 J&M 22-23 VIB 164.00 0.92 20.80

1045 J&M 22-30 VIB 261.00 0.92 27.50

1050 J&M 28-35 VIB 261.00 1.17 27.50

1051 J&M 360 VIB 783.00 0.94 21.70

1052 J&M 416 VIB 250.00 0.92 26.70

1053 J&M 416B VIB 261.00 0.92 26.70

1054 J&M 416S VIB 250.00 0.92 26.70

1055 J&M 815 VIB 429.00 0.92 26.70

1056 J&M 44-30 VIB 250.00 0.92 20.00

1057 J&M 44-50 VIB 399.00 0.92 26.70
516

Type Hammer 

Make 

Hammer 

Model

Hammer 

Type

Rated Power 

kW 

Ecc. Mass

kips 

Frequency

Hz 

1058 J&M 44-65 VIB 552.00 0.92 27.50

1060 J&M 66-65 VIB 552.00 0.92 21.70

1061 J&M 66-80 VIB 559.00 0.92 26.70

1100 PVE 14M VIB 190.00 1.01 28.30

1101 PVE 23M VIB 234.00 1.66 27.50

1102 PVE 25M VIB 294.00 0.98 28.30

1103 PVE 27M VIB 294.00 0.98 28.30

1104 PVE 38M VIB 392.00 0.92 28.30

1105 PVE 50M VIB 440.00 1.20 28.30

1106 PVE 52M VIB 564.00 0.75 28.30

1107 PVE 105M VIB 784.00 1.52 22.50

1108 PVE 110M VIB 784.00 0.80 22.50

1109 PVE 200M VIB 1130.00 1.45 23.30

1110 PVE 2307 VIB 190.00 0.47 38.30

1111 PVE 1420 VIB 190.00 1.01 33.30

1112 PVE 2315 VIB 234.00 1.09 38.30

1113 PVE 2520 VIB 294.00 1.81 33.30

1114 PVE 2310VM VIB 190.00 0.72 38.30

1115 PVE 2315VM VIB 234.00 1.09 38.30

1116 PVE 2316VM VIB 294.00 1.16 38.30

1117 PVE 2319VM VIB 392.00 1.37 38.30

1118 PVE 2323VM VIB 392.00 0.83 38.30

1119 PVE 2332VM VIB 564.00 1.16 38.30

1120 PVE 2335VM VIB 784.00 1.27 38.30

1121 PVE 40VM VIB 564.00 1.45 33.30

1122 PVE 50VM VIB 564.00 1.20 30.00

1123 PVE 55M VIB 403.00 1.17 28.33

1124 PVE 82M VIB 565.00 1.76 28.33

1125 PVE 300M VIB 1796.00 6.21 23.33

1126 PVE 16VM VIB 335.00 0.35 38.33

1127 PVE 20VM VIB 395.00 0.41 38.33

1128 PVE 24VM VIB 395.00 0.52 38.33

1129 PVE 28VM VIB 403.00 0.61 38.33

1130 PVE 2070VM VIB 1130.00 1.52 33.33

1131 PVE 2312VM VIB 252.00 0.26 38.33

1132 PVE 2350VM VIB 790.00 1.09 38.33

1142 PTC 30HP VIB 196.00 0.87 27.00

1143 PTC 40HD VIB 269.00 0.87 28.00

1144 PTC 50HD1 VIB 255.00 0.87 25.00

1145 PTC 50HD2 VIB 290.00 0.87 25.00

1146 PTC 65HD VIB 305.00 0.87 26.00

1147 PTC 60HD VIB 305.00 0.87 28.00

1148 PTC 75HD VIB 410.00 0.87 25.00
517

Type Hammer 

Make 

Hammer 

Model

Hammer 

Type

Rated Power 

kW 

Ecc. Mass

kips 

Frequency

Hz 

1149 PTC 100HD VIB 451.00 0.87 23.00

1150 PTC 100HDS VIB 564.00 0.87 23.00

1151 PTC 175HD VIB 611.00 0.87 23.00

1152 PTC 240HD VIB 988.00 0.87 23.00

1153 PTC 240HDS VIB 988.00 0.87 30.00

1154 PTC 120HD VIB 410.00 0.87 23.00

1155 PTC 130HD VIB 564.00 0.87 23.00

1156 PTC 200HD VIB 710.00 0.87 23.00

1157 PTC 265HD VIB 1080.00 0.87 24.00

1340 H&M H-150 VIB 94.00 0.11 28.30

1341 H&M H-1700 VIB 165.00 0.20 20.00

1431 BRUCE SGV-80 VIB 112.20 0.13 33.33

1432 BRUCE SGV-100 VIB 142.60 0.18 30.00

1433 BRUCE SGV-200 VIB 184.80 0.31 28.83

1434 BRUCE SGV-300 VIB 211.20 0.35 27.50

1435 BRUCE SGV-400 VIB 286.00 0.44 26.67

1436 BRUCE SGV-450 VIB 323.40 0.48 26.67

1437 BRUCE SGV-600 VIB 451.50 0.72 26.67

1438 BRUCE SGV-1000 VIB 569.10 1.03 25.00

1630 LBFoster 4150 VIB 335.00 0.53 25.00