Wind and Seismic Effects

Proceedings of the 28th Joint Meeting

NIST SP 904

PUBLICATIONS

A1110S 0M3053

lAT'L INST OF STAND & TECH R.I.C.
The National Institute of Standards and Technology was established in 1988 by Congress

to "assist industry in the development of technology . . . needed to improve product quality,

to modernize manufacturing processes, to ensure product reliability . . . and to facilitate rapid

commercialization ... of products based on new scientific discoveries."

NIST, originally founded as the National Bureau of Standards in 1901, works to strengthen

U.S. industry's competitiveness; advance science and engineering; and improve public health,

safety, and the environment. One of the agency's basic functions is to develop, maintain, and retain

custody of the national standards of measurement, and provide the means and methods for

comparing standards used in science, engineering, manufacturing, commerce, industry, and

education with the standards adopted or recognized by the Federal Government.

As an agency of the U.S. Commerce Department's Technology Administration, NIST

conducts basic and applied research in the physical sciences and engineering, and develops

measurement techniques, test methods, standards, and related services. The Institute does generic

and precompetitive work on new and advanced technologies. NIST's research facilities are located

at Gaithersburg, MD 20899, and at Boulder, CO 80303. For more information contact the Public

Inquiries Desk, 301-975-3058.
Wind and

Seismic

NIST SP 904

PROCEEDINGS OF

THE 28TH JOINT

MEETING OF

THE U.S.-JAPAN

COOPERATIVE PROGRAM

IN NATURAL RESOURCES

PANEL ON WIND AND

SEISMIC EFFECTS

Building and Fire Research Laboaratory

National Institute of Standards and Technology

Gaithersburg, MD 20899-0001

U.S. DEPARTMENT OF COMMERCE

Michael Kantor, Secretary

TECHNOLOGY ADMINISTRATION

Mary L. Good, Under Secretary for Technology

National Institute of Standards and Technology

Arati Prabhakar, Director

Effects

Issued August 1996

Noel J. Raufaste,

EDITOR
National Institute of Standards and Technology Special Publication 904

Natl. Inst. Stand. Technol. Spec. Publ. 904, 613 pages (Aug. 1996)

CODEN: NSPUE2

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1996

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, DC 20402-9325
PREFACE

This publication contains the Proceedings of the 28th Joint Meeting of the U.S.-Japan Panel on

Wind and Seismic Effects. The meeting was held at the National Institute of Standards and

Technology in Gaithersburg, Maryland during 14-17 May 1996. Forty-six papers were written,

21 by U.S. members and 25 by Japanese members. Thirty-four papers were presented orally;

16 by the U.S.-side and 18 by the Japan-side. The papers were organized into five themes:

Wind Engineering; Earthquake Engineering; Storm Surge and Tsunamis; Summary of Joint

Cooperative Research Programs; and Report of Task Committee Workshops conducted during

the past year. Also, eight papers were presented at two mini-symposia during the technical site

visit segment of the Joint Panel Meeting, four from each side. The papers were presented at the

University of Minnesota, Minneapolis and at the Oregon Department of Geology and Mineral

Industries, Portland. The Japan-side papers are included in these proceedings in the section,

Japan-side Papers Presented at Mini- Symposia.

BACKGROUND

Responding to the need for improved engineering and scientific practices through exchange of

technical data and information, research personnel, and research equipment, the United States

and Japan in 1961 created the U.S.-Japan Cooperative Science Program. Three collateral

programs comprise the Cooperative Science Program. The U.S.-Japan Cooperative Program in

Natural Resources (UJNR), one of the three, was created in January 1964. The objective of

UJNR is to exchange information on research results and exchange scientists and engineers in

the area of natural resources for the benefit of both countries. UJNR is composed of 16 Panels

each responsible for specific technical subjects.

The Panel on Wind and Seismic Effects was established in 1969. Nineteen U.S. and seven

Japanese agencies participate with representatives of private sector organizations to develop

and exchange technologies aimed at reducing damages from high winds, earthquakes, storm

surge, and tsunamis. This work is produced through collaboration between U.S. and Japanese

member researchers working in 1 1 task committees. Each committee focuses on specific

technical issues, e.g., earthquake strong motion data. The Panel provides the vehicle to

exchange technical data and information on design and construction of civil engineering

lifelines, buildings, and water front structures, and to exchange high wind and seismic

measurement records. Annual meetings alternate between Japan and the United States (odd

numbered years in Japan; even numbered years in the United States). These one-week

technical meetings provide the forum to discuss ongoing research and research results; oneweek technical study tours follow the meetings.

The National Institute of Standards and Technology (NIST) provides the U.S.-side chair and

secretariat. The Public Works Research Institute (PWRI), Japan, provides the Japan Side chair

and secretariat.

iii
Cooperative research is done through formal Panel Programs. In 1981, cooperative research in

Large-Scale Testing was started under the auspices of the Panel. Also, in 1981, joint research

on Reinforced Concrete Structures was initiated. Full-scale testing was performed at the

Building Research Institute (BRI), one of the six Japanese members' organizations, with

supporting tests in Japan and in the United States Two years later, a joint research program on

Steel Structures was initiated. Full-scale testing again was led by BRI with supporting tests in

the United States and Japan. The U.S. -Japan coordinated program for Masonry Building

Research was started in 1985. A U.S.-Japan coordinated program on Precast Seismic Structural

Systems was initiated in 1991. A joint program on Seismic Performance of Composite and

Hybrid Structures was initiated in 1993. In 1994, a joint program was started on Physical and

Numerical Simulation of Structural Damages Due to Liquefaction and Development of

Countermeasure Techniques.

Task Committee meetings, exchanges of data and information through technical presentations

at annual Panel meetings, exchanges of guest researchers, visits to respective research

laboratories and informal interactions between Panel meetings, joint workshops and seminars,

and joint cooperative research programs all contribute to the development and effective

delivery of knowledge that has influenced design and construction practices in both countries.

Guest research exchanges have advanced the state of technology in areas of steel, concrete, and

masonry structures under seismic forces; developed techniques to analyze risks from

liquefaction; modeled water seepage in dam foundations; performed comparative analyses of

seismic design of U.S. and Japanese bridges.

Direct communication between counterpart country organizations is the cornerstone of the

Panel. Effective information exchanges and exchanges of personnel and equipment have

strengthened domestic programs of both countries. There are opportunities for experts in

various technical fields to get to know their foreign counterparts, conduct informal exchanges,

bring their respective views to the frontiers of knowledge, and advance knowledge of their

specialties.

The Panel's activities resulted in improved building and bridge standards and codes and design

and construction practices in hydraulic structures in both countries, for example:

created and exchanged digitized earthquake records used as the basis of design and

research for Japan and the United States;

transferred earthquake engineering information and strong-motion measurement

techniques for use by seismically active countries, e.g., Australia, Canada, Italy, Mexico,

Peru, Taiwan, Turkey, and North Africa;

produced data that advanced retrofit techniques for bridge structures;

developed field test data for use in aerodynamic retrofit of bridge structures;

produced full-scale test data that advanced seismic design standards for buildings;

advanced technology for repairing and strengthening reinforced concrete, steel, and

masonry structures;

improved in-situ measurement methods for soil liquefaction and stability under seismic

iv
loads;

created a database comparing Japanese and U.S. standard penetration tests to improve

prediction of soil liquefaction;

created a database on storm surge and tsunamis and verified mathematical models of

tsunami and storm surge warning systems;

established a library resource of current research on wind and earthquake engineering

and on storm surge and tsunamis;

published proceedings of Panel meetings, Task Committee Workshops, and special

publications such as List of Panel Publications and translated two-volume series on

earthquake resistant construction using base isolation systems;

gained better knowledge of both countries' research, design, and construction

capabilities from in-depth visits to the host country's laboratories and building and

public work projects. Results of such visits contribute to creation of new Task

Committees, agendas for Joint Panel meetings and task committee workshops, special

visits of U.S. -Japan researchers, and joint collaborative research.

HIGHLIGHTS OF THE TECHNICAL SITE VISITS

Eight technical sites were visited at four geographic locations during the 28th Joint Panel

Meeting.

Washington, DC

1. Ronald Reagan Federal Triangle Building

This 298 000 m2 Federal office building, on 4.5 hectares of land in central Washington, DC,

completes Congress' 1972 plan to redevelop the Federal Triangle; and the periphery of

Pennsylvania Avenue, Constitution Avenue, and 14th Street. The building's main function is

to serve as a trade center.

Congress created the Pennsylvania Avenue Development Corporation (PADC) in 1972, along

with the General Services Administration (GSA), to provide oversight to all construction within

the Federal Triangle. In the Winter of 1995, U.S. budget cuts forced elimination of PADC's

role one-year before it scheduled close-down. Congress assigned GSA's National Capital

Region with full responsibility to manage the completion of this $720 million project through

1997 when GSA will take ownership of the building and begin repaying the construction loan

from rent collected by GSA tenants. While this facility is a Federal building, it was not funded

by appropriations. Loans were secured from the Federal Financing Bank.

Site construction started in 1991; more than 840 000 m

3 of earth was excavated. Since the

water level is about 4 m below street level, aim thick perimeter slurry wall was constructed

down to bed rock about 27 m below street level. The wall is anchored into rock straturns. The

water level could not be drawn down because neighboring buildings are on timber piles. Over

191 000 m

3 concrete was used for this cast-in-place trade center. About 33 700 metric tons

V
of epoxy coated reinforcing rods were used and more than 1 730 metric tons of structural steel.

Below grade concrete girders, some 4 m deep and 24 m long, transfer loads onto the 9 m by 18

m grids for the 33 400 m

2 parking. The building's facade is faced with 6 650 m3 of Indiana

limestone.

The height of the building conforms to Washington, DC Fine Arts Commission's 33.5 m height

limitation (not to exceed the height of the U.S. Capitol) and design (conforms to the

neoclassical style of its 1920's and 1930's-era neighbors). However, its interior is contemporary with a half-cone-shaped atrium, curved wall mirrors, large auditorium, and

37 000 m

2 space for exhibitions, restaurants, and private-sector trade-related offices.

2. Smithsonian Quadrangle Museums

The delegation visited the underground 34 000 m

2 museum designed for the Smithsonian

Institution to house its vast art collection from Japan and Africa. Opened in 1987, the

Quadrangle Museums comprise the National Museum of African Art and the Arthur M. Sackler

Gallery and an education center, the S. Dillion Rippley Center. To satisfy Washington, DC

Commission of Fine Art's concerns about constructing a new museum on this site, the

Smithsonian Institution proposed and received approval to construct the African and Asian art

museums below ground. Only five percent of its volume would be above ground. This design

permitted unimpeded visibility of the original Smithsonian Institution Building, known as the

Renwick "Castle" (1855), from Independence Avenue. The layout, to hold this vast collection,

involved three levels of viewing space below grade. It was designed to local seismic code. A 20

m slurry wall was constructed along with 27 dewatering wells. Since a high water table was

present, a 2.6 m double slab was constructed as ballast to prevent upheaving and a double wall

construction to reduce moisture migration into the museum spaces further. Soil lm to 2 m

thick was placed on the roofs for construction of a public garden. An underground passage

connects to an adjourning art museum — the Freer Gallery.

Miami, FL

3. The National Hurricane Center, National Oceanic and Atmospheric

Administration

The National Hurricane Center (NHC), maintains a continuous watch on tropical storms over

the Atlantic, Caribbean, and Gulf of Mexico from 15 May through 30 November. The Center

prepares and distributes hurricane watches and warnings for the public, and prepares and

distributes marine and military advisories for other users. NHC coordinates with Hurricane

Warning Offices, Weather Service Forecast Offices, and Weather Service Offices when a

tropical storm threatens the United States. The Center prepares and distributes probabilities of

hurricane/tropical storm conditions for 45 locations along the East and Gulf coasts.

NHC conducts R&D to improve tropical storm forecasting and concentrates on transferring

theoretical advances to operational forecasting. Within NHC is the Aerial Reconnaissance

Coordination All Hurricanes Unit that schedules aerial reconnaissance requirements and

assigns flying units to meet these requirements. Data from the reconnaissance aircraft is downvi
linked from satellites and provided to the hurricane specialist for forecasting and warning.

Tropical storm position estimates are issued between 2-hourly intermediate advisories

whenever a storm with a well-defined radar center is within 200 nautical miles of land-based

radar in the United States. These estimates give the center location in map coordinates and

distance and direction from a well-known point.

NHC is improving its ability to: forecast a storm strike at a landfall because evacuation of

people costs about $ 1 million per 1.6 Ion (not including costs of closing down businesses) and

to develop better prediction and 3-D models. Forecasting predictions have improved at the rate

of about 1 percent a year over the last 20-years (from 120 to 100 nautical miles out). Since

more than 50 million people live along the east U.S. coast to Texas, NHC places high priority

toward improving forecasting accuracy.

HRC's Hurricane Research Division (HRD) is concerned with tropical meteorology using radar

and instrumented aircraft. HRD produces real-time surface wind analysis for major storms

giving forecasters better data. HRD is improving future real-time surface wind analysis and

storm surge modeling and developing products for emergency management officials for post

storm search and rescue efforts, damages assessments, and recovery efforts. HRD's staff total

30 persons who are operating with $4 million budget.

Homestead, FL

4. Habitat for Humanity

Following the 1992 Hurricane Andrew, a 18 hectare site known as Jordan Commons,

Homestead, Florida was selected as the site for 187 dwelling units that replace housing

destroyed by the Hurricane. Jordan Commons is one of many self-help, sweat-equity housing

projects established throughout the United States under the auspices of Habitat for Humanity.

Here, qualified families are required to donate 400 hours of their time toward the construction

of their homes. Forty units will be constructed using all steel frames. The other homes are constructed of reinforced masonry blocks -- the traditional building material in southern

Florida. All units will be fully insulated and equipped with solar water heating systems. Each

unit cost $75,000 paid over 20 years or 30 years, interest free. This community features a sustainable environment. All constniction materials are recycled.

The project site was selected by HUD which has teamed with the National Association of

Home Builders Research Center (NAHBRC) to provide the technical assistance in construction

code approval and builder education. Assembly of cold formed steel members is performed onsite. The American Iron and Steel Institute (AISI) donated the steel for framing and produced

a training manual for owners and contractors who are constructing the houses. NAHBRC

provides quality control methods through use of their research facility and seminars and oneweek, hands-on, steel framing construction training for carpenters, builders, and new owners of

these residences.

vii
Building codes have been revised as a result of damages sustained from the 1992 Hurricane.

Building codes now require low rise buildings to resist lateral wind velocity of 52 m/s and wind

borne missiles traveling at 15 m/s.

Since the use of cold formed steel for frame construction is not explicit in the building code,

prototype housing will be evaluated for performance and cost. NAHBRC developed

performance criteria of the steel framing and recommended lateral bracing to meet local

building codes. The residences' exterior walls are 16 mm, 4-ply plywood; all windows require

functional shutters. The steel exterior C-shaped wall studs are 1.1 mm (43 mils) thick.

Headers and beams are 1.4 mm (58 mils) thick. The exterior wall studs are about 30 percent

thicker than minimum code requirements.

Minneapolis, Minnesota

5. St. Anthony Falls Hydraulic Laboratory, University of Minnesota

The Laboratory is part of the Department of Civil and Mineral Engineering, University of

Minnesota. Its focus is on fluid mechanics, waters resources engineering, and aerodynamics

and wind engineering. Their principal facilities include a glass walled flume to study pressure

fluctuations in hydraulic-jump spillways and turbidity currents; 76 m multipurpose river

channel used to study navigation conditions and develop geometries for water facilities and

structures; and the 7.5: 1 contraction boundary Layer Wind Tunnel high-speed test section

measures 1.7 mby 1.8 m in cross section by 1 6 m long, with velocities up to 45 m/s. Its lowspeed test section measures 2.4 m by 2.4 m by 18 m long, with velocities up to 19 m/s, in the

return leg of the tunnel. The wind tunnel is available under contract. Recent past studies

include the flow of wind around buildings and determination of the peak and fluctuation

pressures on buildings.

The Hydraulic Laboratory's budget supports seven staff and graduate students. The budget

varies between $1-2 million annually with a typical project costing about $150,000. The

budget for the wind tunnel is about $250,000.

During this visit, the delegation conducted a mini-symposium featuring four papers, two by

wind tunnel graduate students:

o Review of On-going Research on Effect of Turbulence on Mean and Fluctuation Pressures of

Buildings and

o Wind Tunnel Simulation ofAtmospheric Boundary Layers

and two from the Japan-side:

o New Evaluation Technique of Wind Loads on External Building Elements by Hisashi Okada, of

BRI and

o Wind Tunnel Study on Long-Span Bridges Conducted at the Large Boundary Layer Wind

Tunnel by Hiroshi Sato of PWRI.

Discussions followed each paper. Professor Farell, our host, discussed his research on wind

loading around circular cylinders including cooling towers and distributed related papers.

viii
Eden Prairie, Minnesota

6. MTS Systems Corporation

MTS was founded in 1962 as the Mechanical Testing and Simulation Company to perform

aerospace stress simulations and material performance testing for transport and experimental

aircraft. Today MTS is: o developing vehicle testing and software control technology to replicate test track

conditions in the laboratory,

o performing civil engineering testing including simulations of material testing systems for

rock, soils, and concrete, and

o developing actuators and controls for large scale test facilities.

Eight percent of its revenue from the approximate $150 million sales is put back in to research.

MTS's engineering and management data are available through its internal web site.

Following a tour of the facilities, the delegation learned about MTS's future direction of

dynamic testing using large-scale structural test facilities. Currently MTS is developing Japan's

Public Works Research Institute's shake table to replicate the loading from the Hyogoken

Nanbu (Kobe) Earthquake. Also, they provided the major components for Japan's Science and

Technology Agency's National Research Institute for Earth Science and Disaster Prevention large

shake table. MTS said its future is in producing modular multi-span portable shake tables such

as the 2-45 ton 14 m by 14 m tables being designed for the University of Nevada under FEMA

sponsorship. MTS, in collaboration with MIT, is studying a large scale seismic simulator for

the Idaho National Engineering Laboratory. Work is underway to perform in-situ, artificial

earthquake tests for highway bridges using SR1/MTS' RESCUE Technique (Repeatable

Earthquake Shaking by Controlled Underground Explosives).

Portland Oregon

7. CORPS Portland Earthquake Engineering District

The delegation visited the CORPS Portland District Offices and on another day the CORPS'

Bonneville Dam, located about 60 1cm east of Portland that spans the Columbia River

separating Oregon and Washington States. Staff from the Structural and Architecture Design

Section presented a comprehensive and lively review of the Districts' responsibilities about

seismic design and construction and provided a set of the latest CORPS earthquake related

design manuals:

o Dam Safety. The District manages 2 1 multipurpose hydroelectric power dams; provides

dam safety and evaluation, emergency action plans, and dam safety training programs.

The District enacted seismic studies of dam safety programs in 1990 and in 1994

established more stringent requirements based on new findings of Oregon's earthquake

ground motion. Seismic evaluations must be done for all projects in Zone 2 and greater.

Dynamic response stress analysis is required in seismic Zone 2 or greater when the sitespecific peak ground motion is 0. 15 g or greater. During the early 1990's, seismic risk of

northern Oregon has been changed from Zone 2 to Zone 3. The District is reevaluating

the maximum credible earthquake values considering new earthquake sources and

ix
evaluating critical structures to assure that they meet project safety requirements.

There is consideration of using GPS to monitor dams' performance.

o Ground Motion Studies . District staff is mapping fault zones and estimating

earthquake damage along non-known faults where earthquakes are envisioned to occur. Based on new knowledge, the District staffs are revising project design criteria guidelines

to ensure uniform revisions are made to highways, bridges, dams, and buildings.

o Seismic Analysis of Embankment Dams. Based on recent data, the District Office has

revised its dam safety evaluation guidelines for liquification potential and embankment

deformations. The design procedures used for embankment dams are in the engineering

manual, Stability ofEarth and Rockfill Dams that uses a static analysis with an additional

horizontal force (calculated based on the seismic coefficient for the region). These

analyses are based on Simplified Proceduresfor Evaluating Soib Liquification Potential and

Rationalizing the Seismic Coefficient Method

.

o Seismic Technical Center. The Center conducts seminars on lifeline safety and training,

determines the vulnerability of Army facilities against earthquakes, develops databases

on civil works' facilities and cost estimates to remedy the damaged facilities, develops

recommendations to revise the CORPS' Emergency Manual on Evaluating Lifelines , and

develops seismic information bulletins and the CORPS' Engineering Manual. The Center

is collaborating with the Air Force in its development of an Air Force Lifelines

Assessment Manual. The staff participates in the Interagency Committee on Seismic

safety in Construction and in civil and military seismic oversight groups.

o Bonneville Seismic Rehabilitation . The original dam, powerhouse, and lock were

constructed during 1933 to 1937; a second powerhouse was constructed between 1974

and 1981. Result from the District's latest dam safety assurance evaluation shows the

dam is adequate for current stability requirements. Now, there is a reassessment

underway of the reinforced downstream piles for the original powerhouse. Initial

analysis showed the up stream end and intermediate pilasters would be stressed beyond

yield under the more stringent earthquake requirements. The main intake piers at the

intake roof and the intake roof itself were determined to be areas of vulnerability under

the more stringent requirements. The District will perform further strength evaluation

based on consideration on nonelastic energy dissipation behavior, response modification

factors and ductility of concrete structures. In addition, a 2-D and 3-D localized models

of up stream and down stream wall-roof system will be done.

Portland Oregon

8. Oregon Department of Geology and Mineral Industries (DGMI)

The Department's mission concerning natural disaster reduction includes: providing a costeffective source of geological information for State citizens and using that information to create

partnerships for reducing life loss and property damage from geologic hazards; serving as experts on the state's geologic processes; and serving as the State's entry point to recommend

geologic hazards safety legislation. Before 1975, there were no seismic requirements for

constructing buildings in Oregon. Based on findings from various seismic vulnerability and

zonation studies of the early and mid-1980's greater attention has been given to better

X
designing against seismic loads. Seismic records show the Portland Greater Metropolitan Area

experienced 5-10 reasonably large earthquakes during the past 100 years. As instrumentation

improved, greater number of seismic records were detected. Before 1993, Oregon was

designated a Zone 2 level of risk; now North Oregon is designated, Zone 3. Recent studies

revealed a large earthquake could result in $ 127 billion of damages to buildings in Portland.

About 50 percent of the State's buildings were constructed before seismic design criteria were

available. The costs to rehabilitate these buildings are estimated at $16 billion; too costly for

the State to fund. DGMI has identified the highest risk unreinforced masonry buildings. The

estimated cost to retrofit these buildings is $2 billion. A recent Department report

recommends: develop a State wide seismic standard; create an inventory of buildings; increase

State seismic engineering expertise; identify financial incentives that will speed seismic retrofit

alterations; and stress education to achieve public support for raising the needed money to

realize seismic retrofit. A Department Task Force is performing a triage of the most important

actions based on available funding. Their schedule of actions includes: a 5-year plan to perform

an inventory of State buildings; a 15-year plan to complete the necessary rehabilitation of

triaged buildings; and a 75-100 year plan where all buildings will either have been retrofitted or were constructed during the 21st century.

A mini-symposium was held featuring four papers; two from the Japan-side include:

o A Seismic Retrofitting Methodfor Reinforced Concrete Bridge Piers by Hisanori Otsuka,

Jun-ichi Hoshikuma and Kazuhiro Nagaya of PWRI and

o Experimental Study on the Bearing Capacity of Cast-In-Place Concrete Piles Damaged by the

Hanshin Earthquake by Hiroyuki Nakajima, Hidenao Hayashi and Kenji Kosa of the

Hanshin Expressway Public Corporation.

DGMI presented two papers:

o Earthquake Risk in Oregon by Matthew Mabey from DGMI and

o Discussion of the Draft Report Prepared by DGMI's Seismic Rehabilitation Task Force by David

BugniofDGMI.

Following discussion, the delegation visited three unreinforced masonry buildings undergoing

seismic retrofit:

Pioneer Federal Courthouse . This historic building is being proposed for retrofit with base

isolation. Three types are under consideration: high damping rubber; lead rubber; and friction

pendulum. Shear wave velocities tests were performed in a 100 m hole to characterize the soil

conditions around the building and to develop base isolation specifications. The basement

floor will be replaced to permit installation of 50-70 base isolators. The estimated cost for this

job is $ 13 million.

Multnomah County Library. This partially reinforced 1912 building was constructed of clay

tile infills (inside) and its exterior is of masonry infills. The building experienced high uplift

loads in the basement causing walls to buckle. The building has a stiff diaphragm pans in

place and reinforced concrete beams. A conventional seismic retrofit program commenced in

xi
1994 consisting of retrofitting the foundation using pin piles to tie it to the earth. The

estimated retrofit cost is expected at $25 million.

Multnomah Hotel . This 1910 abandoned building was one of Portland's finest hotels. The Eshape building, in cross-section, is constructed of lightly reinforced concrete frames and infill

masonry walls. Embassy Suites Hotel Corporation purchased the building and is retrofitting it

for seismic safety. Thirty- 12 m pin piles were installed to tie the foundation to the ground.

Shear walls were installed from its basement to the top of the building. They are tied into

existing columns; three in the E-W direction and four in the N-S direction. Gunite was used to

face the walls. Drag bars were installed on the third floor to transfer the tension loads to the

shear walls. Struts were installed between the wings and the gunite faced shear walls for

structural stability. The cost of this retrofit is $20 million; $1.5 million for the structural

retrofit.

Noel J. Raufaste, Secretary-General

U.S. -side Panel on Wind and Seismic Effects

xii
ABSTRACT

This publication is the Proceedings of the 28th Joint Meeting of the U.S. -Japan Panel on Wind

and Seismic Effects. The meeting was held at the National Institute of Standards and

Technology, Gaithersburg, Maryland, during 14-17 May 1996. The Proceedings include the

program, list of members, panel resolutions, task committee reports, and the 46 technical

papers written for this joint meeting.

The papers were presented within five themes: 1. Storm Surge and Tsunamis, 2. Earthquake

Engineering, 3. Joint Cooperative Research Program, 4. Wind Engineering, and 5. Summaries

of Task Committee Workshop Reports (oral presentations only).

KEYWORDS: Bridges; building technology; concrete; design criteria; disaster reduction;

earthquakes; geotechnical engineering; ground failures; lifelines; liquefaction; masonry; repair

and retrofit; risk assessment; seismic; standards; storm surge; structural engineering; tsunamis;

and wind loads.

xiii

CONTENTS

PREFACE iii

ABSTRACT xiii

AGENDA FOR 28TH JOINT PANEL MEETING xix

LIST OF PANEL MEMBERS xxvii

LIST OF TASK COMMITTEE MEMBERS Ivii

RESOLUTIONS Ixiii

THEME I: Storm Surge and Tsunamis

U.S. Operational Storm Surge Forecast Models 3

Wilson A. Shaffer and Jye Chen

Long-term Shoreline Change in Japan 7

Shigenobu Tanaka

Assessment ofHurricane Opal Impact on East Pass Florida with Airborne Lidar 23

William E. Roper

THEME II: Earthquake Engineering

Mitigation ofLateral Flow ofLiquefied Ground 39

Dcuo Towhata, Yoshihiro Kogai, Kei Amimoto, and Hendri Gusti Putra

Evaluation ofLiquefaction Potential Using the CPT 53

Richard S. Olsen, Joseph P. Koester, and Mary E. Hynes

Estimation ofEffects ofLiquefaction-Induced Ground Flow on Bridge Foundations 69

Keiichi Tamura, Yoshio Ninomiya, and Tadashi Hamada

Preliminary Guidelinesfor Liquefaction Assessment Using Shear Wave Velocity 77

Ronald D. Andrus and Kenneth H. Stokoe, II

Damage Investigation ofSteel Buildings in Specific Areas Observedfrom the 1995

Hyogoken-Nanbu Earthquake 85

Midorikawa Mitsumasa, Hasegawa Takashi, Mukai Akiyoshi, Nishiyama Isao,

Fukuta Toshibumi, and Yamanouchi Hiroyuki

An Integrated Program to Improve the Performance ofSteel Moment Resisting

Frame Buildings 91

Stephen A. Mahin and Michael G. Mahoney

xv
Recent Developments in Strong-Motion Attenuation Relationships 101

William B. Joyner and David M. Boore

Topographical Effect in the Evaluation ofDesign GroundMotion 117

I. Okawa, T. Kashima, and S. Koyama

Preliminary Estimates of Ground-Motion Amplificationfor the Northridge and

Hyogo-ken Nanbu Earthquakes ofJanuary 1 7, 1994 and 1995 125

Roger D. Borcherdt

Damage to Communication Tunnels Caused by Liquefaction and Other Phenomenon

During the Hanshin-Awaji Earthquake and Future Countermeasures 135

Yuzo Yamaguchi and Masayuki Tsuchiya

Effect ofLoad Path on Damage to Concrete Bridge Piers 149

Andrew W. Taylor, Ashraf El-Bahy, William Stone, and Sashi Kunnath

Guide Specificationsfor Reconstruction andRepair ofHighway Bridges Which

Suffered Damage Due to the Hyogo-ken Nanbu Earthquake 159

Kazuhiko Kawashima, Kazuhiro Nishikawa, Masanori Nakano, Shigeki Unjoh,

Yoshitomi Kimura, and Jun-ichi Hoshikuma

Comparison of U.S. and Japanese Highway Bridge Column Seismic Design 1 75

W. Phillip Yen, James D. Cooper, Hisanori Otsuka, and Torn Terayama

Residual Displacement Response Spectrum and its Application to Reinforced

Concrete Bridge Piers 185

Kazuhiko Kawashima, Gregory A. MacRae, Jyun-ichi Hoshikuma, and Kazuhiro Nagaya

Analysis ofDamaged Foundations ofHighway Bridges in the 1995 Hyogo-ken

Nanbu Earthquake 205

Michio Okahara, Masanori Nakano, and Yoshitomi Kimura

Evaluation ofFault Rupture Geometry on Design Ground Motions 223

D.R.H. O'Connell, and J.P. Ake

Damage to Port Facilities ofKOBE Port by the 1995 Hyogo-ken Nanbu Earthquake

and Their Reconstruction 241

Tatsuo Uwabe, Ken Oikawa, Tatsuo Wako, and Takahiro Sugano

Major Factors ofDamages ofReinforced Concrete Buildings due to the 1995

Hyogo-ken Nanbu Earthquake 249

H. Hiraishi, T. Kaminosono, M. Teshigawara, and S. Nakata

Push Over Laboratory Testing of Unreinforced Masonry Infills 257

Ghassan Al-Chaar, Steve Sweeney, and Parmalee Brady

Designs and Techniquesfor the Restoration ofthe Earthquake-Damaged

Hanshin Expressway Kobe Route 271

Hisamori Ohtsuka, Hiroyuki Nakajima, Hidenao Hayashi, and Kenji Kosa

xvi
Action and Concept on Performance-Based Design and Engineering System

in Japan 285

Hiroyuki Yamanouchi, Hisashi Okada, and Yuji Ohashi

Compositesfor Construction 293

John B. Scalzi

Arrangement ofKyoshin Net 295

Keiichi Ohtani, Shigeo Kinoshita, and Hiroyuki Fujiwara

Seismic Stabilityfor Sliding ofRockfill Dams 303

Tadahiko Sakamoto, Tadahiko Fujisawa, Akira Nakamura, and Tomoya Iwashita

THEME III: Summary of Joint Cooperative Research Programs

U.S.-Japan Cooperative Earthquake Research Program on Composite and Hybrid

Structures - Research Progress and Current Status 319

Subhash C. Goel and Isao Nishiyama

Design ofLarge-Scale Liquefaction Experiment System 327

K. Ishihara, T. Kagawa, N. Ogawa, C. Minowa, K. Sakai, and A. Abe

THEME IV: Wind Engineering

Influence of Vent Configurations on Bridge Response to Wind 333

Harold R. Bosch

Aerodynamic Characteristics ofSlotted Box Girders 343

Hiroshi Sato and Katsuya Ogihara

Wind Engineering Research in the United States 357

Ahsan Kareem and Eleanore Sabadell

Estimation ofRelevant Internal Pressure and Wind Loads on Cladding 373

H. Okada and L. Kong

Descriptive Epidemiology of Wind-Related Mortality 381

Josephine Malilay, Annemarie Wasley, and Leslie A. Fiedler

JAPAN-side PAPERS PRESENTED AT MINI-SYMPOSIA

New Evaluation Technique of Wind Loads on Building External Elements 391

H. Okada and L. Kong

Wind Tunnel Studies on Long-Span Bridges Conducted at the Large Boundary

Layer Wind Tunnel in Tsukuba 399

Hiroshi Sato

xvii
A Seismic Retrofitting Methodfor Reinforced Concrete Bridge Piers and Their

Analysis 427

Hisanori Otuska, Jun-ichi Hoshikuma, and Kazuhiro Nagaya

Experimental Study on the Bearing Capacity ofCast-in-Place Concrete Piles

Damaged by the Hanshin Earthquake 441

Hiroyuki Nakajima, Kidenao Hayashi, and Kenji Kosa

MANUSCRIPTS AUTHORED FOR PANEL MEETING BUT NOT PRESENTED ORALLY

Analyses ofLiquefaction in the 1995 Hyogo-ken Nanbu Earthquake 453

Osamu Matsuo, Takao Shimazu, and Tatsuya Tsutsumi

Photogrammetric Measurement of Three Dimensional Ground Surface

Displacement Caused by the Kobe Earthquake 465

Hiroshi Murakami and Minoru Hoshino

A Real Time Systemfor Forecasting Storm Surges Caused by the Typhoons 475

Tatsuo Konishi

Windborne Debris in Hurricanes and Typhoons 485

Joseph E. Minor

Liquefaction Potential ofCoarse Foundation Materials: Lost Creek Dam 495

Ronald A. Oaks

U.S. Guidelinesfor the Seismic Rehabilitation ofBuildings 517

Christopher Rojahn and Ugo Morelli

Northridge Earthquake Responses ofBase-Isolated (USC) and Conventionally

Designed (Olive View) Hospital Buildings 527

Mehmet K. Celebi

Recent U.S. Experience in Design and Testing ofBeam-to-Column Connections

for Moment-Resisting Steel Frames 539

Ferdinand LoRay Schwartz

APPENDIX -Summaries of Task Committee Reports 549

xviii
AGENDA

28th TECHNICAL PANEL MEETING on

WIND AND SEISMIC EFFECTS

14-17 May 1996

Tuesday 14 May

1000 OPENING CEREMONIES

(Lecture Room B, Administration Building)

Call to order by Noel RAUFASTE, Secretary-General US-side Panel

Opening remarks by Dr. Robert HEBNER, Deputy Director, National Institute

of Standards and Technology

Remarks by Akihiro FUJITA, Counsellor for Science and Technology, Embassy

of Japan

Remarks by Richard N. WRIGHT, Chairman US-Side, Panel on Wind and

Seismic Effects, Director, Building and Fire Research Laboratory

Remarks by Tadahiko SAKAMOTO, Chairman Japan- Side, Panel on Wind and

Seismic Effects, Director- General, Public Works Research Institute

Introduction of U.S. Members by U.S. Panel Chairman

Introduction of Japan Members by Japan Panel Chairman

Elect Joint Meeting Chairman

Adopt Agenda

Adjourn

1115 Group Photograph

1130 Lunch: Hosted by Dr. Robert HEBNER, Deputy Director, National Institute

of Standards and Technology

xix
THEME - STORM SURGE AND TSUNAMIS

1245-1405 Technical Session - STORM SURGE AND TSUNAMIS

Chairman: Mr. Tadahiko SAKAMOTO

U.S. Operational Storm Surge Forecast Models, Wilson A. SHAFFER and Jye

CHEN, NOAA

Long-term Shoreline Change in Japan, Shigenobu TANAKA, PWRI

Assessment of Hurricane Opal Impact on East Pass Florida with Airborne Lidar,

William ROPER, CORPS

Discussion

Break

TASK COMMITTEE MEETINGS

1425-1700 Task Committee Meetings

T/C B Testing and Evaluation Procedures for Building Systems Mr. K. Ohtani (Japanside Chair) and H.S. Lew (U.S. -side Chair)

T/C C Design Evaluation and Improvement of Structures Dr. Hirashi (Japan-side

Chair) and Dr. Ken Chong (U.S. -side Chair)

T/C D Dam Earthquake Engineering Dr. Okahara (Japan-side Acting Chair) and Dr.

William Roper (U.S. -side Chair)

T/C E Design for Wind and Wind Hazard Mitigation Drs. Okada and Sato (Japanside Co-Chairs) Drs. Nora Sabadel and Andrew Taylor (U.S. -side Co-Chairs)

T/C F Disaster Prevention Methods for Lifeline Systems Dr. Otsuka (Japan-side Acting

Chair) and Dr. Riley Chung (U.S. -side Chair)

T/C K Wind and Earthquake Engineering for Offshore and Coastal Facilities Dr.

Sugano (Japan-side Acting Chair) and Dr. Charles Smith (U.S. -side Chair)

1700 Conclusion of Day 1

XX
Wednesday 15 May

THEME - EARTHQUAKE ENGINEERING

0830-1130 Technical Session - Earthquake Engineering Part 1

Chairman: Dr. Richard WRIGHT

0830 Mitigation of Lateral Flow of Liquefied Ground, Ikuo TOWHATA, Univ. of

Tokyo, Yoshihiro KOGAI, Fudo Construction Co., Kei AMIMOTO, Univ. of

Tokyo, and Hendri GUSTI-PUTRA, Andalas Univ., Indonesia

0850 Evaluation of Liquefaction Potential Using the CPT, Richard OLSEN, Joseph

KOESTER, and Mary Ellen HYNES, WES

0910 Estimation of Effects of Liquefaction-induced Lateral Spreading on Bridge

Foundations, Keiichi TAMURA, Yoshio NINOMIYA, Tadashi HAMADA,

PWRI

0930 Preliminary Guidelines for Liquefaction Assessment Using Shear Wave Velocity,

Ronald ANDRUS, NIST and Kenneth Stokoe, University of Texas

0950 Discussion

1010 Break

1030 Damage Investigation of Steel Buildings in Specific Areas Observed from the

1995 Hyogoken Nanbu Earthquake, Mitsumasa MIDORIKAWA, Takashi

HASAGAWA, Yoshiaki MUKAI, Isao NISHIYAMA, Toshifumi FUKUTA,

Hiroyuki YAMANOUCHI, BRI

1050 An Integrated Program to Improve the Performance of Steel Moment Resisting

Frame Buildings, Stephen MAHIN, Univ of Calif Berkeley and Michael

MAHONEY, FEMA

1110 Discussion

1 130 Adjourn for Lunch

1 135 Lunch: Hosted by Dr. William ANDERSON, Director, Hazard Mitigation

Section, Civil and Mechanical Systems Division, National Science Foundation

xxi
1245-1425 Technical Session - Earthquake Engineering (Continued)

Chairman: Dr. Richard WRIGHT

1245 Recent Developments in Strong-Motion Attenuation Relationships, William B.

JOYNER and David M. BOORE, USGS

1305 Topographical Effect in the Evaluation of Design Earthquake Motion, Izuru

OKAWA, Toshihide KASHIMA, Shin KOYAMA, BRI

1325 Preliminary Estimates of Ground-Motion Amplification for the Northridge and

Hyogo-ken Nambu Earthquakes of January 17, 1994 and 1995, Roger

BORCHERDT, USGS

1345 Damage to Communication Tunnels Caused by Liquefaction and Other Phenomenon during the Hanshin-Awaji Earthquake and future Countermeasures,

yuzo YAMAGUCHI and Masayuki TSUCHIYA, NTT

1405 Discussion

1425 Break

TASK COMMITTEE MEETINGS

1440-1700 Task Committee Meetings

T/C A Strong-motion Data and Applications Dr. Sugano (Japan-side Acting Chair and

Dr. Roger Borcherdt (U.S. -side Chair)

T/C G Structural Control and Intelligent Material Systems Dr. Otsuka (Japan-side

Chair) and Dr. Shi-Chi Liu (U.S. -side Chair)

T/C H Soil Behavior and Stability During Earthquakes Dr. Tamura (Japan-side Chair)

and Dr. A.G. Franklin, U.S.-side Chair)

T/C I Storm Surge and Tsunamis Dr. Okahara (Japan-side Acting Chair) and Dr.

Blackford (U.S.-side Chair)

T/C J Wind and Earthquake Engineering for Transportation Systems Dr. Sato (Japanside Acting Chair) and Mr. JamesCooper (U.S.-side Chair)

1700 Conclusion of Day 2

xxii
Thursday 16 May

THEME - EARTHQUAKE ENGINEERING - PART II

0815-1135 Technical Session - Earthquake Engineering-Part II

Chairman: Mr. Tadahiko SAKAMOTO

08 15 Effect of Load Path on Damage to Concrete Bridge Piers, Andrew TAYLOR,

NIST, Ashraf EL-BAHY, Univ. of Central Florida, William STONE, NIST, and

Safshi KUNNATH, Univ. of Central Florida

0835 Guide Specifications for Reconstruction and Repair of Highway Bridges Which

Suffered Damage due to the Hyogo-ken Nanbu Earthquake, Kazuhiko

KAWASHIMA, Tokyo Institute of Technology, Kazuhiro NISHIKAWA,

Masanori NAKANO, Shigaki UNJOH, Yoshitomi KIMURA, and Jun-ichi

HOSHIKUMA, PWRI

0855 Comparison of U.S. and Japanese Highway Bridge Column Seismic Design,

Philip YEN and James Cooper, FHWA, and Hisanori OTSUKA and Toru

Terayama, PWRI

09 15 Residual Displacement Response Spectrum and its Application to Reinforced

Concrete Bridge Piers, Kazuhiko KAWASHIMA, Tokyo Institute of Technology,

Gregory A. MACRAE, Univ. of Wash., Jun-ichi HOSHIKUMA and Kazuhiro

NAGAYA, PWRI

0935 Discussion

0955 Break

1015 Analysis of Damaged Foundations of Highway Bridge in the Hyogoken Nanbu

Earthquake, Michio OKAHARA, Masanori NAKANO, Yoshitomi KIMURA,

PWRI

1035 Evaluation of Fault Geometry on Design Ground Motions, Jon P. AKE and

D.R.H. O'CONNELL, BUREC

1055 Damage to Port Facilities of Kobe Port by the 1995 Hyogo-ken Nanbu Earthquake and their Reconstruction, Tatsuo UWABE, PHRI, Ken OIKAWA, Port

Construction Bureau, and Tatsuo WAICO, and Takahiro SUGANO, PHRI

1115 Discussion

1 135 Adjourn for lunch

1135 Lunch: Hosted by Richard KRIMM, Acting Associate Director, Mitigation

Directorate, Federal Emergency Management Agency

xxiii
1235-1620 Technical Session - Earthquake Engineering (Continued)

Chairman: Mr. Tadahiko SAKAMOTO

1300 Major Factors of Damages of Reinforced Concrete Building due to the 1995

Hyogo-ken Nanbu Earthquake, Hisahiro HIRAISHI, Takashi KAMINOSONO,

Masaomi TESHIGAWARA and S. NAKATA, BRI

1320 Pushover laboratory testing of unreinforced Masonry Infills, Steven Sweeney,

Pamalee Brady, and Ghassan Al-Chaar, CERL

1340 Design and Techniques for the Restoration of the Earthquake Damaged Hanshin

Expressway Kobe Route, Hisanori OHTSUKA, PWRI, and Hiroyuki

NAKAJIMA, Hidenao HAYASHI, and Kenji KOSA, Hanshin Expressway Public

Corporation

1400 Action and Concept on Performance-Based Design and Engineering System in

Japan, Hiroyuki YAMANOUCHI, Hisashi OKADA, Yuji OHASHI, BRI

1420 Discussion

1440 Break

1500 Composites for Construction, Jack SCALZI, NSF

1520 Arrangement for Kyoshin-net, Keiichi OHTANI, Shigeo KINOSHITA, Hiroyuki

HUJIWARA, NRIESDP

1540 Seismic Stability for Sliding of Rockfill Dams, Tadahiko SAKAMOTO, Tadahiko

FUJISAWA, Akira NAKAMURA, Tomoya IWASHITA, PWRI

1600 Discussion

1620 Break

THEME - SUMMARY JOINT COOPERATIVE RESEARCH PROGRAMS

Chairman: Mr. Tadahiko SAKAMOTO

1640-1740 Technical Session - Joint Cooperative Research Programs

1640 U.S. -Japan Cooperative Earthquake Research Program on Composite and

Hybrid Structures - Research Progress and Current Status, Subhash C. GOEL,

University of Michigan and Isao NISHIYAMA, BRI

1700 Design of Large-scale Liquefaction Experiment System, Kenji ISHIHARA,

Science Univ. of Tokyo, Takaaki KAGAWA, Wayne Univ., Nobuyuki OGAWA,

NIED, Chikahiro MINOWA, NIED, K. SAKAI, Kiso-Jiban Consultants Co., and

A. ABE, Tokyo Soil Research Co.

1720 Discussions

1740 Conclusion of Day 3

xxiv
Friday 1 7 May

THEME - WIND ENGINEERING

Chairman: Dr. Richard WRIGHT

0815-1030 Technical Session - Wind Engineering

0815 Influence of Vent Configurations on Bridge Response to Wind, Harold BOSCH,

FHWA

0835 Aerodyanic Characteristics of Slotted Box Girders, Hiroshi SATO, Katsuya

OGIHARA, PWRI

0855 Discussion

0910 Wind Engineering Research in the United States, Ahsan KAREEM, Notre Dame

University and Eleanore SABADELL, NSF

0930 Estimation of Relevant Internal Pressure and Wind Loads on Cladding, Hisashi

OKADA, Linze KONG, BRI

0950 Descriptive Epidemiology of Wind-related Mortality, Josephine MALILAY,

Annemarie WASLEY, and Leslie A. FIEDLER, CDC

1010 Discussion

1030 Break

THEME - SUMMARY RECENT PANEL WORKSHOPS

Chairman: Dr. Richard WRIGHT

1045-1140 Workshop Reports

1045 T/C "A" Strong Motion Data, Site Response to Strong Earthquake Motions, 16-17 January 1996, Yokosuka.

1055 T/C "F" Disaster Prevention Methods for Lifeline Systems, 6th Workshop on Earthquake Disaster Preventionfor Lifeline Systems, 18-

19 July 1995, Osaka.

1 105 T/C "I" Storm Surge and Tsunamis, 4th Workshop on Storm Surge and

Tsunamis, Hawaii, 1 April 1996.

1115 T/C "J" Wind and Earthquake Engineering for Transportation Systems,

10th Workshop on Bridge Engineering, 30-31 May 1995,

Tsukuba.

1125 Discussion

1140 Adjourn for Lunch

1 140 Lunch: Hosted by Dr. Walter HAYS, Deputy for Research Applications, United

States Geological Survey (Courtyard)

XXV
TASK COMMITTEE REPORTS AND RESOLUTIONS

Report of Task Committees

Chairman: Dr. Richard WRIGHT

Strong-Motion Data and Applications

Testing and Evaluation Procedures for Building Systems

Evaluation and Improvement of Structures

Dam Earthquake Engineering

Design for Wind and Wind Hazard Mitigation

Disaster Prevention Methods for Lifeline Systems

Structural Control and Intelligent Materials Systems

Soil Behavior and Stability During Earthquakes

Storm Surge and Tsunami

Wind and Earthquake Engineering for Transportation Systems

Wind and Earthquake Engineering for Offshore and Coastal Facilities

Break

Adoption of Final Resolutions

Break

CLOSING CEREMONIES

1610 Call to Order by Noel J. RAUFASTE, Secretary- General, U.S.-Side Panel

Closing Remarks by Tadahiko SAKAMOTO, Chairman Japan- Side Panel

Closing Remarks by Richard N. WRIGHT, Chairman U.S.-Side Panel

1630 Conclusion of 28th Joint Panel Technical Sessions

1245-1445

T/CA

T/CB

T/CC

T/CD

T/CE

T/CF

T/CG

T/CH

T/CI

T/CJ

T/CK

1445-1500

1500-1600

1600-1610

xxvi
LIST OF PANEL MEMBERS

UNITED STATES-SIDE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1996

Dr. Richard N. Wright

Chairman

Director, Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301-975-5900 FAX: 301-975-4032

e-mail:rwright@nist.gov

Mr. Noel J. Raufaste

Secretary-General

Head, Cooperative Research Programs

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301-975-5905 FAX: 301-975-4032

e-mail: nraufaste@nist.gov

Mr. Kharaiti L. Abrol

Principal Structural Engineering Consultant (088)

Room 475

Department of Veterans Affairs

810 Vermont Avenue, NW

Washington, DC 20420

202-565-5579 FAX: 202-565-5478

Dr. John Ake

Geophysicist

Seismotectonics and Geophysics Section

CodeD-8330

P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-236-4195 x276 FAX: 303-236-9127

e-mail:jake@seismo.usbr.gov

xxix
Mr. Charles E. Anderson

Seismic Safety Coordinator

P.O. Box 25007 (D-8110)

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-236-3999 x534 FAX: 303-236-9099

e-mail:canderson@do.usbr.gov

Dr. Celso S. Barrientos

Supervisory Physical Scientist

National Environmental Satelite Data

Information Service - Code E/RA28

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

5200 Auth Road

Camp Springs, MD 20746

301-763-8102 FAX: 301-763-8020

e-mail:cbarrientos@nesdis.noaa.gov

Dr. Eddie N. Bernard

Director, Pacific Marine Environmental Laboratory

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

7600 Sand Point Way, NE

BINC15700/Building3

Seattle, WA 981 15-0070

206-526-6800 FAX: 206-526-6815

e-mail:bernard@pmel.noaa.gov

Mr. Mike Blackford

Director, Pacific Tsunami Warning Center

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

91270 Fort Weaver Road

Ewa Beach, HI 96706-2928

808-689-8207 FAX: 808-689-4543

e-mail:mike@ptwc.noaa.gov

Dr. David M. Boore

U.S. Geological Survey

345 Middlefield Road MS 977

Menlo Park, CA 94025

415-329-5616 FAX: 415-329-5163

e-mail:boore@samoa.wr.usgs.gov

xxx
Dr. Roger D. Borcherdt

Branch of Seismology

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415-329-5619 FAX: 415-329-5163

e-mail:borcherdt@samoa.wr.usgs.gov

Mr. H. Lee Butler

Chief, Research Division

Coastal Engineering Research Center

U.S. Army Engineer Waterways Experiment Station

Office CEWES-CR

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2405 FAX: 601-634-4314

Mr. Harish Chander

U.S. Dept. Of Energy (EH-31),

CXXI, Room 2016 (GTN)

19901 Germantown Road

Germantown, MD 20874-1290

301-903-6681 FAX: 903-8693

e-mail:harish.chander@hq.doe.gov

Dr. Ken P. Chong

Program Director, Mechanical & Structural Systems

National Science Foundation

4201 Wilson Boulevard - Room 545

Arlington, VA 22230

703-306-1361 FAX: 703-306-0291

e-mail: Kchong@NSF.gov

Dr. Riley Chung

Group Leader, Earthquake Engineering

Structures Division

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301-975-6056 FAX: 301-869-6275

e-mail:rchung@enh.nist.gov

xxxi
Mr. James D. Cooper

Chief, Structures Division, HNR-10

Federal Highway Administration

U.S. Department of Transportation

6300 Georgetown Pike

McLean, VA 22101

703-285-2060 FAX: 703-285-2766

e-mail:jcooper@cais.com

Dr. A.G. Franklin

Chief, Earthquake Engineering &. Geophysics Division

CEWES-GH Geotechnical Laboratory

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2658 FAX: 601-634-4134

e-mail:FRANKLA@EX1 .wes.army.mil

Mr. William Freeborn

Housing and Urban Development

Room 8132

451 Seventh Street, SW

Washington, DC 20410

202-708-4370 FAX: 202-619-8360

E-mail:WiUiam_E._Freeborne@HUD.GOV

Mr. James H. Gates

Caltrans - Structures

P. O. Box 94287

Sacramento, CA 94274

916-227-8773 FAX: 916-227-8174

Mr. Peter E. Gurvin

Director, Building Design and Engineering Division

Foreign BuUding Operations

Building SA-6, Room 335

U.S. Department of State

Washington, DC 20520

703-875-6117 FAX: 703-875-6204

xxxii
Dr. Walter W. Hays

Deputy for Research Applications

Office of Earthquake, Volcanoes and Engineering

905 National Center

U.S. Geological Survey

U.S. Department of the Interior

Reston, VA 22092

703-648-6711 FAX: 703-648-6717

e-mail:whays@usgs.gov

Mr. Larry C. Hultengren

Senior Structural Engineer

Office of Civil/Structural Engineering

Foreign Building Operations

Building SA-6, Room 346

U.S. Department of State

Washington, DC 20520

703-875-6194 FAX: 703-875-6204

Dr. William B. Joyner

Geophysicist

Branch of Seismology

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415-329-5640 FAX: 415-329-5163

e-mail:j oyner@samoa. wr. usgs . gov

Mr. Roger M. Kenneally

Structural Engineer

Structural and Seismic Engineering Branch

Mail Stop T10-L1

U.S. Nuclear Regulatory Commission

Washington, DC 20555

301-415-6303 FAX: 301-415-5074

e-mail:rmk@nrc.gov

Dr. George Lee

Director

National Center for Earthquake Engineering Research

State University of New York at Buffalo

Red Jacket Quadrangle

Buffalo, NY 14261

716-645-3391 FAX: 716-645-3399

xxxiii
Dr. H.S. Lew

Chief, Structures Division

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301-975-6061 FAX: 301-869-6275

e-mail:hsl@nist.gov

Dr. Shih-Chi Liu

Program Director, Structural Systems

Division of Biological and Critical Systems

National Science Foundation

4201 Wilson Boulevard - Room 545

Arlington, VA 22230

703-306-1362 FAX: 703-306-0291

e-mail:cliu@nsf.gov

Dr. Francis G. McLean

Value Engineering Program Manager

Code D-8170

P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-236-9120x236 FAX: 303-236-6737

e-mail:fmclean@do.usbr.gov

Dr. Josephine Malilay

Diaster Assessment and Epidemiology Section

National Center for Environmental Health

F-46 Center for Disease Control and Prevention

4770 Buford Highway, NE

Atlanta, GA 30341

770-488-7350 FAX: 770-488-7335

Dr. Raymond E. Meyer

Senior Program Manager

Office of Foreign Disaster Assistance

SA14/Room 1705

USAIDAVashington, D.C. 20523-1443

703-875-1522 FAX: 703-875-1708

xxxiv
Mr. Ugo Morelli

Policy Manager

Federal Emergency Management Agency

500 C Street, SW

Washington, DC 20472

202-646-2810 FAX: 202-646-2577

Mr. Howard D. Nickerson

Earthquake Engineering and Weapons Specialist

Naval Facilities Engineering Command

Hoffman Building #2, Room 12S63, Code 04B2

200 Stovall Street

Alexandria, VA 22332

202-433-8599 FAX: 202-433-8777

Dr. William E. Roper

Assistant Director

Research and Development (Civil Works)

U.S. Army Corps of Engineers

20 Massachusetts Avenue, N.W.

Washington, DC 20314-1000

202-761-0257 or 1845 FAX: 202-761-0907

email:Bill.Roper@inet.hq.USACE.army.mil

Dr. J. Eleonora Sabadell

Director, Natural and Technological Hazards Mitigation Program

Directorate of Engineering

National Science Foundation

4201 Wilson Boulevard - Room 545-03

Arlington, VA 22230

703-306-1361 FAX: 703-306-0291

e-mail:esabadel@nsf.gov

Dr. Charles E. Smith

Research Program Manager

Offshore Minerals Management

Technology Assessment and Research Branch

Minerals Management Service

U.S. Department of the Interior

381 Elden Street, MS 4800

Herndon, VA 22070-48 1

7

703-787-1559 FAX: 703-787-1555

e-mail:smithc@smtp.mms.gov

xxxv
Mr. Stanley Strickland

Chief, Air Base Systems Branch

Stop 37, Wl/FIVC/OL, Bldg. 1 120

139 Barnes Drive, STE 2

Tyndall AFB, FL 2403-6001

904-283-3709 FAX: 904-283-3722

Dr. Andrew Taylor

Research Structural Engineer

Structures Division

Bldg. 226, Room B- 158

Building and Fire Research Laboratory

National Institute of Standards and Technology

Gaithersburg, MD 20899- 1 000

301-975-6078 FAX: 301-869-6275

e-mail:ataylor@enh.nist.gov

xxxvi
ALTERNATE MEMBERS

Dr. Clifford J. Astill Program Director

Division of Biological and Critical Systems

National Science Foundation

4201 Wilson Boulevard - Room 545

Arlington, VA 22230

703-306-1361 FAX: 703-306-0291

Dr. Mehmet K. Celebi

Research Civil Engineer

Branch of Earthquake and Geomagnetic Information

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415-329-5623 FAX: 415-329-5163

Mr. Michael Changery

Chief, Global Analysis Branch

National Climatic Data Center

National Oceanic and Atmospheric Administration

U. S. Department of Commerce

Federal Building

AshviUe, NC 28801

704-271-4765 FAX: 704-271-4246

Dr. C.Y. Chen

Senior Civil/Geotechnical Engineer

Office of Foreign Buildings

Department of State

Code SA-6, Rm. 347

Washington, DC 20520

703-875-6207 FAX: 703-875-6204

Mr. Vincent P. Chiarito

Research Structural Engineer

Structural Mechanics Division

Structures Laboratory

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2714 FAX: 601-634-3412

xxxvii
Dr. James F. Costello

Senior Structural Engineer

Structural and Geological Engineering Branch

Office of Nuclear Regulatory Research

U.S. Nuclear Regulatory Commission

Mail Stop T10-L1

Washington, DC 20555

301-415-6009 FAX: 301-415-5074

Mr. Lucian G. Guthrie, P.E.

Structures Branch, Eng. Div.

U.S. Army Corps of Engrs.

UQUSACE (CECW-ED)

Washington, D.C. 20314-1000

202-761-8673 FAX: 202-761-4716

e-mail:civil works. eng 1 postguthrie ( Lucian Guthrie

)

Dr. Robert L. Hall

Chief, Structural Analysis Group

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2567 FAX: 601-634-3412

Dr. James R. Houston

Director, Coastal Engineering Research Center

U.S. Army Engineer Waterways Experiment Station/WESCV-Z

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2000 FAX: 601-634-2055

e-mail:houston@coafs 1 .wes.army.mil

Mr. James Lander

Geophysicist

Cooperative Institute for Research in

Environmental Sciences

University of Colorado

Campus Box 449, Room 152 RL3

3100 Marine Street

Boulder, CO 80309

303-497-6446 FAX: 303-497-6513

xxxviii
Mr. Tingley K. Lew

Research Structural Engineer

Structures Division

Naval Facilities Engineering Service Center - Code C62

11 10 23rd Ave.

Port Hueneme, CA 93043-4370

805-982-1234 FAX: 805-982-1418

Mr. Michael Mahoney

Physical Scientist

Federal Emergency Management Agency

500 C Street, SW

Washington, DC 20472

202-646-2794 FAX: 202-646-4387

Dr. Erdal Safak

Research Structural Engineer

U.S. Geological Survey

Box 25046, MS 966

Denver Federal Center

Denver, CO 80225

303-273-8593 FAX: 303-273-8600

Dr. John B. Scalzi

Program Director, Structures and Building Systems

National Science Foundation

4201 Wilson Blvd. - Room 545

Arlington, VA 22230

703-306-1361 FAX: 703-306-0291

E-mail:jscalzi@nsf.gov

xxxix
JAPAN-SIDE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

Steering Committee Membership List

Mr. Tadahiko Sakamoto

Chairman, Japan-side . Panel on Wind and Seismic Effects, UJNR

Director-General

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2821

FAX: 0298-64-2148

e-mail: tsakamo@pwri.go.jp

Dr. Michio Okahara

Secretary-General, Japan-side

Panel on Wind and Seismic Effects, UJNR

Research Coordinator for Advanced Technology

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-3725

FAX: 0298-64-4344

e-mail: okahara@pwri.go.jp

Mr. Naotoshi Baba

Director, Road Disaster Prevention Section

Road Bureau

Ministry of Construction

2-1-3 ,Kasumigaseki,Chiyoda-ku,

Tokyo 100

TEL: 03-5251-1896

FAX: 03-5251-1949

XL
Mr. Kunio Enoshima

Senior Assistant for Disaster Prevention

Planning Division

Administration Department

Japan Meteorological Agency

1 -3-4 ,Ohtemachi,Chiyoda-ku

Tokyo 100

TEL: 03-3212-8341

FAX: 03-3212-2453

Mr. Tadahiko Fujisawa

Director, Dam Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2832

FAX: 0298-64-0164

e-mail: fujisawa@pwri.go.jp

Mr. Takeshi Hashimoto

Director,Earthquake Disaster Countermeasures Division

Prevention Bureau

National Land Agency

1-2-2, Kasumigaseki, Chiyoda-ku,

Tokyo 100

TEL: 03-3593-3311

FAX: 03-3501-5199

Dr. Hisahiro Hiraishi

Structural Engineering Department

Building Research Institute

Ministry of Construction

1 ,Tatehara,Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-6639

FAX: 0298-64-6773

XLi
Dr. Susumu Iai

Chief, Geotechnical Earthquake

Engineering Laboratory

Port and Harbour Research Institute

Ministry of Transport

3-l-l,Nagase,Yokosuka -shi

Kanagawa-ken 239

TEL: 0468-44-5028

FAX: 0468-44-4095

Dr. Takamasa Inatomi

Director, Structural Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1 ,Nagase,Yokosuka-shi,

Kanagawa-ken 239

TEL: 0468-44-5027

FAX: 0468-44-0839

Dr. Toshio Iwasaki

President, Civil Engineering Research Laboratory

Kyodo Building 5F, 3-3-4,Kanda,

Kaji-cho, Chiyoda-ku,

Tokyo 101

TEL: 03-3254-9481

FAX: 03-3254-9448

Mr. Takashi Kaminosono

Head, Construction Techniques Division

Building Research Institute

Ministry of Construction

1 ,Tatehara, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2151

FAX: 0298-64-2989

e-mail: kamino@kenken.go.jp

Dr. Kazuhiko Kawashima

Professor, Department of Civil Engineering Tokyo Institute of Technology

2-12-1 ,0-okayama,Meguro-ku,

Tokyo 152, Japan

TEL: 03-5734-2922

FAX: 03-3729-0728

XLii
Dr. Yasuyuki Koga

Director, Earthquake Disaster

Prevention Department

Public Works Research Institute

Ministry of Construction

1 Asahi,Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2829

FAX: 0298-64-0598

e-mail: koga@pwri.go.jp

Dr. Hiroshi Kofune

Director, Hydraulic Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi

Kanagawa-ken 239

TEL: 0468-44-5009

FAX: 0468-41-3888

Mr. Kiyoshi Mimura

Director,Geographic Department

Geographical Survey Institute

Ministry of Construction

1 , Kitasato, Tsukuba-shi

Ibaraki-ken 305

TEL: 0298-64-1111

FAX: 0298-64-1804

Dr. Chikahiro Minowa

Cooperative Research Officer

National Research Institute for

Earth Science and Disaster Prevention

Science and Technology Agency

3- 1 , Tennodai, Tsukuba-shi,

Ibaraki-ken,305

TEL: 0298-51-1611

FAX: 0298-51-5658

e-mail: minowa@geo.bosai.go.jp

XLiii
Mr. Eishi Mochizuki

Director, Seismology and Volcanology

Research Department

Meteorological Research Institute

Japan Meteorological Agency

1- 1, Nagamine, Tsukuba-shi

Ibaraki-ken 305

TEL: 0298-53-8675

FAX: 0298-51-3730

e-mail: emochizu@mri-jma.go.jp

Mr. Akihito Nakahama

Director, Disaster Prevention Office

Coast Construction and Disaster Prevention Division

Ports and Harbour Bureau, MOT

2- 1-3, Kasumigaseki, Chiyoda-ku,

Tokyo 100

TEL: 03-3580-4452

FAX: 03-5511-8280

Mr. Akira Nakamura

Head, Fill Type Dam Division

Dam Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2413

FAX: 0298-64-0164

e-mail: nakamura@pwri.go.jp

Dr. Tadahiko Nakao

Director, River Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2830

FAX: 0298-64-1168

XLiv
Dr. Shinsuke Nakata

Director, Production Department

Building Research Institute

Ministry of Construction

1 , Tatehara, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-1111

FAX: 0298-64-2989

Dr. Nobuyuki Narita

Professor, Tokyo Metropolitan University

17-15, Yawata-6, Ichikawa-shi,

Chiba-ken 272

TEL: 0473-35-0324

FAX: 0473-35-0324

Dr. Isao Nishiyama

Research Planning and Information

Department Coordination for International Research Cooperation

Building Research Institute

Ministry of Construction

1 , Tatehara,Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2151

FAX: 0298-64-2989

e-mail: isao@kenken.go.jp

Dr. Setsuo Noda

Director- General

Port and Harbour Research Institute

Ministry of Transport

3-1-1 ,Nagase,Yokosuka-shi,

Kanagawa-ken 239

TEL: 0468-44-5000

FAX: 0468-42-9265

Dr. Shigeyuki Noto

Director, Structures Division

Civil Engineering Research Institute

Hokkaido Development Agency

1-3, Hiragishi, Toyohira-ku, Sapporo-shi, Hokkaido 062

TEL: 011-841-1111

FAX: 011-820-2714

e-mail: noto@hdbceri.go.jp

XLV
Dr. Tsuneo Okada

Professor, University of Tokyo

7-22- 1 , Roppongi, Minato-ku,

Tokyo 106

TEL: 03-3402-6231

FAX: 03-3402-1908

e-mail: okada@kappa.iis.u-tokyo.ac.jp

Dr. Masao Okamoto

Director, Disaster Prevention Investigation Section

River Bureau

Ministry of Construction

2- 1-3, Kasumigaseki, chiyoda-ku,

Tokyo 100

TEL: 03-5251-1884

FAX: 03-5251-1946

Mr. Keiichi Ohtani

Director, Disaster Prevention Research Division

National Research Institute for Earth Science and Disaster Prevention

Science and Technology Agency

3- 1, Tennodai, Tsukuba-shi,

Ibaraki-ken,305

TEL: 0298-51-1611

FAX: 0298-51-1622

Dr. Hisanori Ohtsuka

Head, Earthquake Engineering Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2932

FAX: 0298-64-0598

e-mail: otsuka@pwri.go.jp

Mr. Shin Ohtsuka

Director, Typhoon Research Division

Meteorological Research Institute

Japan Meteorological Agency

Ministry of Transport

1-1, Nagamine, Tsukuba-shi

Ibaraki-ken 305

TEL: 0298-53-8552

FAX: 0298-53-8545

XLVi
Mr. Hiroshi Sasaki

Director, Building Guidance Division

Building Disaster Prevention Section

Housing Bureau

Ministry of Construction

2- 1 -3,Kasumigaseki,Chiyoda-ku,

Tokyo 100

TEL: 03-3580-4311

FAX: 03-3580-7050

Dr. Hiroshi Sato

Head, Structure Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1 ,Asahi,Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2874

FAX: 0298-64-0565

e-mail: hsato@pwri.go.jp

Dr. Keiichi Tamura

Head, Ground Vibration Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2926

FAX: 0298-64-0598

Mr. Shigenobu Tanaka

Head, Coastal Engineering Division

River Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2327

FAX: 0298-64-1168

e-mail: stanaka@pwri.go.jp

XLvii
Dr. Tetsushi Uehara

Director, Office of Disaster Prevention

Research Policy Division

Research and Development Bureau

1-2-2, Kasumigaseki Chiyoda-ku

Tokyo 100

TEL: 03-3581-5271

FAX: 03-3595-0980

Dr. Tatsuo Uwabe

Chief, Earthquake Disaster Prevention Laboratory

Structural Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi,

Kanagawa-ken 239

TEL: 0468-44-5030

FAX: 0468-44-0839

Dr. Makoto Watabe

Professor, Faculty of Environmental Information,

Keio University

5322, Endo, Fujisawa-shi,

Kanagawa-ken 252

TEL: 0466-47-5111

FAX: 0466-47-5041

Mr. Asao Yamakawa

Director, International Affairs Division

Economic Affairs Bureau

Ministry of Construction

2- 1-3, Kasumigaseki, Chiyoda-ku,

Tokyo 100

TEL: 03-5251-1836

FAX: 03-3502-3955

Dr. Hiroyuki Yamanouchi

Director, Structural Engineering Department

Building Research Institute

Ministry of Construction

1 .Tatehara, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2151

FAX: 0298-64-6773

e-mail: yamanoch@kenken.go.jp

XLviii
Dr. Koichi Yokoyama

Director, Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2835

FAX: 0298-64-0565

e-mail: yokoyama@pwri.go.jp

Secretaries Committee Membership List

Dr. Michio Okahara

Secretary-GeneralJapan-side

Panel on Wind and Seismic Effects, UJNR

Research Coordinator for Advanced Technology

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-3725

FAX: 0298-64-4344

e-mail: okahara@pwri.go.jp

Mr. Takekazu Akagiri

Head, Planning Division

Geographic Department Geographical Survey Institute

1 , Kitasato, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-1111

FAX: 0298-64-1804

Dr. Hisahiro Hiraishi

Structural Engineering Department

Building Research Institute

Ministry of Construction

1 , Tatehara, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-6639

FAX: 0298-64-6773

XLix
Mr. Koichi Honda

Director, Research Planning and Coordination

Section Administration Division

Civil Engineering Research Institute of

Hokkaido Development Bureau

1-3, Hiragishi, Toyohira-ku, Sapporo-shi,

Hokkaido 062

TEL: 011-841-1111

FAX: 01 1-824-1226

Dr. Susumu Iai

Chief, Geotechnical Earthquake

Engineering Laboratory

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi

Kanagawa-ken 239

TEL: 0468-44-5028

FAX: 0468-44-4095

Dr. Kazuhiko Kawashima

Professor, Department of Civil Engineering

Tokyo Institute of Technology

2- 12-1, O-okayama, Meguro-ku,

Tokyo 152, Japan

TEL: 03-5734-2922

FAX: 03-3729-0728

e-mail: kawashima@cv.titech.ac.jp

Dr. Yasuyuki Koga

Director, Earthquake Disaster

Prevention Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2829

FAX: 0298-64-0598

e-mail: koga@pwri.go.jp

L
Mr. Osamu Matsuo

Head, Soil Dynamics Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305, Japan

TEL: 0298-64-2933

FAX: 0298-64-2576

Dr. Chikahiro Minowa

Cooperative Research Officer

National Research Institute for Earth Science and Disaster Prevention

Science and Technology Agency

3- 1 , Tennodai, Tsukuba-shi,

Ibaraki-ken, 305, Japan

TEL: 0298-51-1611

FAX: 0298-51-5658

e-mail: minowa@geo.bosai.go.jp

Mr. Akira Nakamura

Head, Fill Type Dam Division

Dam Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2413

FAX: 0298-64-0164

e-mail: nakamura@pwri.go.jp

Mr. Masanori Nakano

Head, Foundation Engineering Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305, Japan

TEL: 0298-64-2873

FAX: 0298-64-0565

Li
Mr. Kazuhiro Nishikawa

Head, Bridge Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305, Japan

TEL: 0298-64-2905

FAX: 0298-64-0565

e-mail: knisika@pwri.go.jp

Dr. Nobuyuki Ogawa

Head, Earthquake Engineering Laboratory

National Research Institute for Earth Science and Disaster Prevention

Science and Technology Agency

3- 1 , Tennodai, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-51-1611

FAX: 0298-51-8512

e-mail: ogawa@geo.bosai.go.jp

Dr. Hisashi Okada

Head of Aerodynamic Division

Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-6641

FAX: 0298-64-6773

e-mail: okada@kenken.go.jp

Mr. Masami Okada

Head, The First Research Laboratory,

Seismology and Volcanology Research Department

Meteorological Research Institute

Japan Meteorological Agency,

Ministry of Transport

1-1, Nagamine, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-53-8677

FAX: 0298-51-3730

e-mail: mokada@mri-jma.go.jp

Lii
Dr. Izuru Okawa

Head, Geotechnical Engineering Division

Structural Engineering Department

Building Research Institute

Ministry of Construction

1 ,Tatehara, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-6642

FAX: 0298-64-6773

e-mail: okawa@kenken.go.jp

Mr. Keiichi Ohtani

Director, Disaster Prevention Research Division

National Research Institute for Earth Science and Disaster Prevention

Science and Technology Agency

3- 1 ,Tennodai, Tsukuba-shi,

Ibaraki-ken, 305

TEL: 0298-51-1611

FAX: 0298-51-1622

Dr. Hisanori Ohtsuka

Head, Earthquake Engineering Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1 >Asahi,Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2932

FAX: 0298-64-0598

e-mail: otsuka@pwri.go.jp

Dr. Hiroshi Sato

Head, Structure Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2874

FAX: 0298-64-0565

e-mail: hsato@pwri.go.jp

Liii
Mr. Masashi Satou

Head, Structures Section

Structures Division

Civil Engineering Research Institute

Hokkaido Development Bureau

1-3, Hiragishi, Toyohira-ku

Sapporo-shi, Hokkaido,062

TEL: 011-841-1111

FAX: 011-820-2714

Dr. Keiichi Tamura

Head, Ground Vibration Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1 ,Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2926

FAX: 0298-64-0598

Mr. Shigenobu Tanaka

Head, Coastal Engineering Division

River Department

Public Works Research Institute

Ministry of Construction

1 ^Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2327

FAX: 0298-64-1168

e-mail: stanaka@pwri.go.jp

Dr. Tatsuo Uwabe

Chief, Earthquake Disaster Prevention Laboratory, Structural Engineering

Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi,

Kanagawa-ken 239, Japan

TEL: 0468-44-5030

FAX: 0468-44-0839

Dr. Hiroyuki Yamanouchi

Director, Structural Engineering Department

Building Research Institute

Ministry of Construction

1 , Tatehara, Tsukuba-shi,

Liv
Ibaraki-ken 305

TEL: 0298-64-2151

FAX: 0298-64-6773

e-mail: yamanoch@kenken.go.jo

Dr. Koichi Yokoyama

Director, Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1 , Asahi, Tsukuba-shi,

Ibaraki-ken 305

TEL: 0298-64-2835

FAX: 0298-64-0565

LV

May 1996

LIST OF TASK COMMITTEE MEMBERS

Task Committee US Side

A. Strong-Motion Data and R.D. Borcherdt*

Applications C.J. Astill

M.K. Celebi

A.G. Franklin

W.B. Joyner

F.G. McLean

C.S. Barrientos

T.K. Lew

E. Safak

+J. Kimball

+J. Hunt

W. Freeborn

J. Ake

H. Meyers

B. Testing and Evaluation Procedures H.S. Lew*

for Building Systems V.P. Chiarito

J.E. Sabadell

C.E. Smith

K. Abrol

+ S. Sweeney

+A. Bieniawski

W. Freeborn

C. Evaluation and Improvement K.P. Chong*

of Structures +J.O. Jirsa

H.S. Lew

T.K. Lew

H.D. Nickerson

+ D.H. Oh

+ M.A. Phipps

J.B. Scalzi

+ S. Woodson

+ S.A. Asar

M.K. Celebi

K.L. Abrol

R.L. Hall

+A. Bieniawski

Japanese Side

S. Iai*

K. Tamura

R. Honda

Y. Yamanouchi

I. Okawa

K. Mimura

K. Otani

+M. Miyata

+Y. Ichii

E. Mochizuki

M. Sato

K. Ohtani*

0. Matsuo

S. Nakata

1. Nishiyama

T. Kaminosono

+T. Fukuda

+A. Tanaka

+ S. Morino

+H. Noguchi

+A. Wada

H. Hiraishi*

Y. Fujishiro

H. Okada

I. Nishiyama

T. Kaminosono

+ M. Teshigawara

+A. Mukai

+ H. Fukuyama

+ S. Kotani

+ H. Kuwamura

+ F. Watanabe

+M. Nakajima

+ H. Shiobara

+J. Kanda

Lvii
+K. Mutreja

P. Brady

D. Ellsworth

W. Freeborn

D. Earthquake Engineering for W.E. Roper*

Dams " R.L. Hall

L.G. Guthrie

+R. Davidson

+M.E. Hynes

A.G. Franklin

+R. Navidi

+L. Von Thun

+W. Allerton

F.G. McLean

E. Design for Wind and Wind

Hazard Mitigation

F. Disaster Prevention Methods

for Lifeline Systems

A. Taylor*

E. Sabadell*

R.M. Chung*

C.J. Astill

M.K. Celebi

J.D. Cooper

+G. Al-Chaar

T.K. Lew

J.B. Scalzi

+J.S. Spencer

+S. Wu

+ S. Sommer

+T. Kabeyazawa

+A. Wada

+B. Tsuji

+Y. Matsus

H. Yoshida*

+T. Koga

T. Fujisawa

+Y. Kato

+1. Nagayama

+K. Watanabe

+Y. Niitaka

+T. Sasaki

+M. Toyota

+Y. Yamaguchi

+T. Iwashita

+J. Kashiwai

+Y. Nakamura

+J. Tamura

+Y. One

+H. Watanabe

+T. Omachi

+T. Kawakami

+Y. Iida

+K. Yoshikawa

+H. Nishida

H. Sato*

H. Okada*

K. Yokoyama*

K. Tamura

H. Otsuka

O. Matsuo

+ S. Tanaka

+M. Kaneko

+J. Hoshikuma

+Y. Kumaki

N. Ogawa

T. Uwabe

+Y. Kajiya

+T. Katayama

+ H. Kameda

+ S. Takada

+T. Sato

+J. Toma

Lviii
G. Structural Control and S.C. Liu*

Intelligent Material V.P. Chiarito

Systems K.P. Chong

+J.R. Hayes

L.C. Hultengren

G. Lee

H. Soil Behavior and Stability A.G. Franklin*

During Earthquakes C.J. Astill

R.D. Borcherdt

C.Y. Chen

F.G. McLean

C.E. Smith

R.M. Chung

+K. Honda

+ H. Suzuki

+K. Shimamura

+Y. Taguchi

H. Otsuka*

+ S. Unjo

Y. Yamanouchi

H. Okada

+H. Fukuyama

+H. Fujitani

+N. Inoue

C. Minowa

M. Sato

IC. Kawashima

+Y. Fujino

4-H. Iemura

+Y. Inoue

+T. Ishimaru

+K. Yoshida

+K. Asano

+A. Nishitani

+A. Ogawa

+K. Ymazaki

+H. Nakajima

+T. Moritani

+Y. Goto

+Y. Makiguchi

+S. Okamoto

K. Tamura*

IC. Yokoyama

M. Nakano

H. Yoshida

O. Matsuo

+T. Hamada

+M. Iiba

C. Minowa

S. Iai

+K. Zen

T. Sugano

+J. Nishikawa

+ S. Yasuda

+1. Towata

+J. Koseki

+H. Suzuki

Lix
+Y. Yamaguchi

+J. Toma

I. Storm Surge and Tsunami M. Blackford*

C.J. Astill

C.S. Barrientos

E.N. Bernard

H.L. Butler

J. Lander

W.E. Roper

A.M. Hittelman

J. Wind and Earthquake Engineering J.D. Cooper*

for Transportation Systems A.G. Franklin

J.H. Gates

J.B. Scalzi

H.S. Lew

+P. Yen

+H.R. Bosch

S. Tanaka*

+ S. Sato

+ S. Iwasaki

R. Kojima

+N. Nagai

M. Okada

+T. Konishi

+T. Sasajima

+N. Syudo

+Y. Kawada

+T. Yamashita

+T. Goto

+A. Nakayama

+K. Satake

Y. Fujishiro*

M. Nakano

+Y. Kimura

+M. Ishida

H. Sato

+K. Ogihara

K. Nishikawa

+J. Murakoshi

K. Tamura

+R. Honda

H. Otsuka

+ S. Unjo

K. Yokoyama

T. Inatomi

+ S. Tanaka

M. Sato

+Y. Fujino

K. Kawashima

+A. Ogawa

+K. Yamazaki

+H. Nakajima

+H. Akiyama

+T. Moritani

LX
K. Wind and Earthquake Engineering C.E. Smith* T. Uwabe*

for Offshore and Coastal Facilities C.S. Barrientos M. Nakan

W.E. Roper T. Inatomi

J.E. Sabadell T. Sugano

+M.C. Miller S. Iai

M.K. Celebi +A. Nozu

M. Okada

+H. Shiojiri

+M. Kazama

+H. Tuchida

+Y. Yoshida

* Chairman

+ TC Members (not Panel Members)

Lxi

RESOLUTIONS

RESOLUTIONS OF THE TWENTY-EIGHTH JOINT MEETING

U.S.-JAPAN PANEL ON WIND AND SEISMIC EFFECTS (UJNR)

National Institute of Standards and Technology

May 14-17, 1996

The following resolutions are hereby adopted:

1. The Twenty-Eighth Joint Panel Meeting provided the forum to exchange valuable

technical information that was beneficial to both countries. In view of the importance

of cooperative programs on the subject of wind and seismic effects, the continuation of

Joint Panel Meetings is considered essential.

2. The following activities have been conducted since the Twenty-Seventh Joint Meeting:

a. Technology Exchanges. Technical experts, technical documents, research

reports, and proceedings for workshops have been exchanged. These exchanges

contributed to the development of research programs and enhanced ongoing

research in both countries.

b. Task Committee Workshops. The Panel held four workshops with technical site

visits:

1 . Task Committee (A) Workshop on Site Response to Strong Earthquake

Motions, 16-17 January 1996, Yokosuka.

2. Task Committee (F) 6th Workshop on Earthquake Disaster Prevention

for Lifeline Systems, 18-19 July 1995, Osaka.

3. Task Committee (I) 4th Workshop on Storm Surge and Tsunamis, 1-3

April 1996, Hawaii.

4. Task Committee (J), 1 1th Bridge Engineering Workshop, 30-3 1 May

1995, Tsukuba.

3. The Panel will continue to seek methods to contribute to the IDNDR such as exchanging Proceedings of Joint Panel Meetings and of Task Committee Workshops

with their respective country's National Committees for the IDNDR.

4. Japanese and U.S. members continued analysis of the most probable causes of failures

from the 17 January 1995 Hyogo-ken-Nanbu Earthquake and the 17 January 1994

Northridge Earthquake. The Japan-side discussed its analysis of the Hyogo-ken-Nanbu

Earthquake at the 28th Joint Panel Meeting. These deliberations presented important

data and information about the reasons for building and infrastructure damages. The

Panel recognizes the importance of using these findings in its ongoing work and of the

need to continue performing future joint post-disaster investigations to improve

scientific knowledge, engineering, and construction practices.

5. The Panel recognizes the importance of providing earthquake science and technology

expertise to the U.S.-Japan Earthquake Disaster Mitigation Partnership of the April 17,

1996 Agreement, U.S.-Japan Common Agenda. The Panel will provide technical

LXV
leadership and participation in appropriate research projects aimed at realizing the

objectives of the priority themes for joint collaborations. Appropriate Task Committees

will respond to the Partnership in planning their work.

6. The Panel approved the Task Committee reports presented during the 28th Joint Panel

Meeting. Each report included objectives, scope of work, accomplishments, future

plans, and other information.

7. The Panel endorsed the following six proposed Task Committee Workshops during the

coming year:

a. Task Committee (B), Workshop on Load Path and Test Observation

Requirements for Quasi-Dynamic and Shake Testing, planned before the

29th Joint Panel Meeting

b. Task Committees (C <SlG), 2nd Workshop on Structural Control and

Intelligent Material Systems, planned for November 1996, College Park, MD,

USA

c. Task Committee (D), 1st Workshop on State-of-the-Art Review of U.S.-Japan

Research on Earthquake Engineering for Dams, proposed for 19-22

November 1996, WES, Vicksburg, MS USA

d. Task Committees (G «Sl J), 4th Workshop on Earthquake Protective Systems

for Bridges, December 1996, Japan

e. Task Committee (J), 12th Bridge Engineering Workshop, 26 October - 3

November 1996, USA.

f. Task Committee (J), 3rd Workshop on Seismic Retrofit of Bridges, December

1996, Japan

In addition, three Task Committees plan to conduct Workshops in 1997, following the

29th Joint Panel meeting:

a. Task Committee (E), 1st Workshop on Design for Wind and Wind Hazard

Mitigation, October or November 1997, USA.

b. Task Committee (F), 7th Joint Workshop on Disaster Prevention for Lifeline

Systems, proposed for June or July 1997, Seattle

c. Task Committee (K) 3rd Workshop on Wind and Earthquake Engineering

for Offshore and Coastal Facilities, July 1997, Japan

Scheduling for the Workshops will be performed by the U.S. and Japan Chairmen of the

respective Task Committees with concurrence of the Joint Panel Chairmen. Both sides'

Secretaries- General shall be kept informed of the planning. Results of each activity

conducted before the 29th Joint Meeting shall be presented at the 29th Joint Panel

Meeting.

8. The Panel recognizes the importance of continuing its joint research programs on

Composite and Hybrid Structures and on Soil Liquefaction and Countermeasures.

9. The Panel approved creation of the Task Committee on Design for Wind and Wind

Hazard Mitigation (T/C "E").

Lxvi
10. The Panel published its third issue, Wind and Seismic Effects Newsletter, Winter 1996.

The newsletter featured the Hyogo-ken Nanbu Earthquake and the 27th Joint Panel

Meeting and summarized damage investigations, workshops and review of two U.S.-side

Panel Agencies and one Japan-side Panel Agency. The Panel endorses continuation of

publishing Panel news as reported in the newsletter. The U.S.-side will publish the

fourth newsletter during the fall of 1996. The Panel encourages both sides to use the

electronic media as much as possible in communicating Panel findings and summaries of

activities. Therefore, the Panel has endorsed creation of a Home Page on the Internet.

NIST will create and maintain the Panel's Home Page. In addition, future Panel

newsletters will be made available through the Panel's Home Page on the Internet.

1 1 . The Panel recognizes the importance of continued exchange of personnel, technical

information, research results, and recorded data that lead to mitigating losses from

strong winds and earthquakes. The Panel also recognizes the importance of using

available large-scale testing facilities and other complementary capabilities in both

countries. Thus, these activities should be continued, strengthened, and expanded. The

Panel will provide official endorsement to facilitate these exchanges.

12. The Twenty-Ninth Joint Panel Meeting of the UJNR Panel on Wind and Seismic

Effects will be held at the Public Works Research Institute, Japan in May 1997.

Specific dates, program, and itinerary will be proposed by the Japan-side with

concurrence of the U.S. -side Panel.

Lxvii

STORM SURGE AND TSUNAMIS

U.S. Operational Storm Surge Forecast Models

by

Wilson A. Shaffer* and Jye Chen*

1. INTRODUCTION

Historically, hurricane-related deaths have mostly

been caused by drowning due to coastal flooding.

The 1900 hurricane that struck Galveston, Texas

caused the deaths of 10,000 to 12,000 people.

(This is the greatest natural disaster to befall the

U.S.) Galveston Island was overwashed by a storm surge, with waves riding above the surge.

Virtually every building on the island was

destroyed. But even in those early days of

meteorology, there was a warning issued. Dr.

Isaac Kline headed the weather office in

Galveston. By observing the swell striking the

coast in the morning, he correctly foretold the

approach of the hurricane. He spent much of the

day riding along the coast, warning residents of

the great storm. Many were saved through his

actions.

Weather observations and forecasts matured

considerably since that devastating 1900

hurricane. We now use satellites, doppler radar,

and aircraft reconnaissance to observe storms

from their inception. We run sophisticated

numerical models of the atmosphere which give a computer forecast of where a storm is going, as well as a hint of its future intensity. Forecasters

at our National Hurricane Center combine all of

this information to make forecasts of a storm's

track and intensity out to 72 hours. This

information is disseminated through various

electronic means to inform the emergency

management community and the public.

Unfortunately, hurricane forecasting is not an

exact science. Track errors have improved over

the past two decades, but is currently at 1 82 km

(100 nmi) for the average 24 hour forecast. As a

result or lack of forecast skill, a relatively large area of the coastline must be warned and

evacuated. Typically, 24 hours before landfall is the time when emergency managers must make

an evacuation decision. Implicit in the track

error is timing error. Using typical forward

hurricane speeds, we see that a storm can make

its landfall either 6 hours before or 6 hours after

its projected time. Along our East Coast, this

means that the storm could come at either high

tide or low tide.

Another important forecast parameter is the

strength, or intensity, of a hurricane. Today, we

have only marginal "skill" in forecasting the

intensity of a storm. We know, for example, that

most hurricanes intensify as they go over warmer

water and die out after land falling. However,

there are exceptions to even these generalities.

Hurricane Alicia (Galveston 1983) was

intensifying as it passed over Houston, Texas.

Hurricane Andrew (Miami 1992) seemed to loose

strength over the Gulf Stream, then intensified as

it struck Miami.

2. STORM SURGE MODELING

The NWS's Techniques Development Laboratory

has developed two numerical models to forecast

storm surges: the Special Program to List the

Amplitudes of Surges from Hurricanes

(SPLASH) (Jelesnianski, 1972) and the Sea,

Lake, and Overland Surges from Hurricanes

(SLOSH) (Jelesnianski et. al., 1992) models.

SPLASH treated surges only up to a smoothed

* Techniques Development Laboratory, Office of Systems

Development, National Weather Service, NOAA, Silver Spring, m j

3
coastline. SLOSH considers overland flooding,

flow up rivers, through sub-grid cuts, and along

barriers. The SPLASH model, because of its

limitations, is no longer used by the NWS.

The original goal of TDL's modeling effort was

to produce a model that could be run in real-time

to estimate the surge before a hurricane makes

landfall. If a hurricane such as Camille was

about to make landfall near Pass Christian,

Mississippi, the NWS would run the SLOSH

model to compute the possible flooding the

hurricane would cause. Because of the

uncertainty in forecasting the landfall location,

several runs would be needed to capture the

possible scenarios of the storm making landfall,

all within the error limitations of the forecast.

Such computer runs are time-consuming. Only a

limited number of SLOSH runs can be prepared,

run, and analyzed as a storm approaches. When

the next forecast package is issued containing a change in the hurricane's track or characteristics, a new set of runs would need to be prepared and

run. This would continue for every forecast

package up to the landfall time. Clearly, this

process would be time consuming and difficult.

Can you imagine the number of computer runs

that would be made for a storm like Donna

(1960) as it skirts the U.S.'s East Coast?

3. MODEL INPUT AND VALIDATION

The model was developed and tested extensively

against observations from past hurricanes. Only

parameters that could reasonably be forecast by

NHC were incorporated as input to SLOSH. The

hurricane was described numerically by its track,

intensity, and size. The required track begins

well before landfall and extends well beyond

landfall. This assures that the storm's impact is

correctly forecast numerically. The intensity of

the storm is described in the model as the

pressure drop from the ambient pressure outside

the storm to the lowest pressure at the center of

the storm. The outer ambient pressure can be

obtained as the pressure of the first anticyclonic

isobar from a weather map. In a forecast, the

value of 1012 mb is usually chosen. Aircraft

reconnaissance measurements of central pressure

are used to determine the central pressure of the

storm. The storm's size is parameterized through

the radius of maximum wind, as measured by

reconnaissance aircraft.

The SLOSH model has been tested extensively

(Jarvinen and Lawrence, 1985), with computed

surge levels compared to observed high water

marks and tide gage observations. In general, we

claim that the observed and forecast surge levels

vary by about +/- 20%. However, the reader

should keep in mind that these comparisons are done after the hurricane, when the track, size, and

intensity are known. For such comparisons, the

model is run with the same type of input as it is under true forecast conditions.

Not only are the maximum surge values

compared, but we also compare the tide gage

record, when possible, to the time-history

generated by the model. Wind speed and

direction are also compared whenever we have

corresponding forecasts and observational data.

Some of these comparisons are given in

Jelesnianski, 1992. More recently, we compared

the SLOSH wind field with the analyzed winds

of Powell et. al., 1995. For most "well behaved"

storms, the computed and the analyzed fields

compare quite well. Hurricane Emily (1993)

proved to be quite anomalous; SLOSH

parametric winds and analyzed winds differed

considerably.

4. EVACUATION STUDIES

The time required to complete an evacuation

along the coastline were poorly understood

before the 1980's. Most evacuations took place

when the NWS issued a formal "hurricane

warning", roughly 18 hours before landfall.

Some coastal areas clearly needed more time to

complete an evacuation. People living on the

southern Florida Keys have approximately 100

miles to drive simply to get off the Keys, then

another 30 miles to get to the Miami area. The

city of New Orleans, Louisiana and its

4
surrounding areas have a population of over a

million persons. Much more time would be

needed for an evacuation than the 18 hour

warning period. Clearly, more information

related to evacuations was needed by the NHC.

The NHC undertook several pilot studies to assist

emergency management in assessing their risk

due to storm surge. Instead of waiting for a hurricane to threaten, the NWS could run its SLOSH model during a convenient time to assess the potential flooding a hurricane of a given

category would generate.

This methodology has expanded to simulation

study in which they examine a full range of

hurricane categories and land falling directions.

For a typical stretch of coastline, NHC runs the

SLOSH model for hurricanes making landfall

from 10 different directions. Approximately a dozen runs will be made for each direction, with

the landfall changed by approximately 15 km for

successive model runs. Also, all five of the

Saffir-Simpson hurricane categories are simulated for areas south of Cape Hatteras, NC.

To the north, category 5 storms are considered

extremely unlikely.

No one could be expected to assimilate such a large number of model runs and the

overwhelming amount of flooding information

generated by each run. Composites are formed

which group together all storms of an individual

direction. These Maximum Envelopes of Water

(MEOW's) are formed as the highest surge

forecast at each model gridpoint from the

hurricanes of the chosen track direction.

Further compositing is done to depict each SaffirSimpson category's potential threat. These

composites, called the Maximum Of the

MEOW's, or MOM's, are formed by taking the

highest value at each gridpoint for MEOW's of a given category.

5. FORECASTER/EMERGENCY

MANAGEMENT AIDS

A computer graphics program has been written

by TDL to aid NWS forecasters and emergency

managers in their response to the hurricane

threat. This program allows the user to view

either MEOW's or MOM's for a given basin. The

user can also inquire to find the projected surge

heights at any location, as well as to window or zoom in to a given geographic area. This

program has been distributed to NWS offices and

to emergency managers trained at the

NWS/FEMA workshop on hurricanes and

evacuation planning.

In addition, there are several useful computer

programs available that aid emergency managers

in interpreting NHC products and in making the

appropriate evacuation decisions.

6. FORECASTS DURING HURRICANE

OPAL, 1995

Throughout the lifetime of hurricane Opal, 1995

from its formation over the Yucatan Peninsula,

during its threat to the U.S. Gulf Coast, and as it died over the continental U.S., forecasts were

continually being issued by NHC. Aircraft

reconnaissance was flown during most of this

period to gather information on the storm's

intensity and movement. Satellite estimates of

the storm's intensity and location were also being

made.

Emergency management officials made

preparations at various times to respond to the

storm's increasing threat to their local

jurisdictions. Approximately three days before

the hurricane was projected to make landfall,

emergency management began their preliminary

actions. Personnel were called and alerted of the

storm. Checkouts of operating procedures were

conducted. Fuel tanks of emergency generators were topped off. Police and fire departments were put on alert.

As the hurricane came closer to the U.S., addition

actions had to be taken. At approximately 24

hours before landfall, a go/no go decision on

evacuation needs to be made. Studies have

5
shown (J. Baker, personnel communication) that

it takes 4-6 hours from the time that the

emergency manager makes his decision to gather

the county officials or city management officials

to make the final decision.

Emergency management officials must keep in

mind that the hurricane is an extensive

phenomenon, not simply a point on a map.

When we concentrate only on the eye position,

we all have a tendency to think of that point as the storm and calculate our distance to the storm

accordingly. In reality, gale force winds may

extend outward from this center over 200 km.

Most emergency managers have chosen to have

their evacuation completed by the time gale force

winds reach the coastline, or when water rises to an elevation that would flood evacuation routes. In the vast majority of cases, the gale force winds

will arrive first. If the hurricane is moving with a forward speed of 20 km/sec, the gale force winds

would arrive approximately 10 hours before the

hurricane's eye made landfall. Disaster could

result from not taking such a "pre-landfall"

hazards time into account.

7. REFERENCES

Houston, S. H., W. A. Shaffer, M. D. Powell, and

J. Chen, "Comparisons of SLOSH parametric and

HRD analyzed surface wind fields in recent

hurricanes", Preprints, 21st Conf. on Hurricanes

and Tropical Meteorology, Miami, Florida.

Amer. Meteor. Soc, Boston, Massachusetts, 1995

Jarvinen, B. J. and M. B. Lawrence, "An

evaluation of the SLOSH storm-surge model",

Bull. Amer. Meteor. Soc, 66, 1408-1411 1985.

Jelesnianski, C. P., 1972: SPLASH (Special

program to list the amplitudes of surges from

hurricanes): I. Landfall storms, NOAA Technical

Memorandum, NWS TDL-46, National Oceanic

and Atmospheric Administration, U.S.

Department of Commerce, 33pp.

Jelesnianski, C. P., J. Chen, and W. A. Shaffer,

SLOSH, Sea, lake, and overland surges from

hurricanes, NOAA Technical Report, NWS-48,

National Oceanic and Atmospheric

Administration, U.S. Department of Commerce,

71pp, 1992.

6
Long-term Shoreline Change in Japan

Shigenobu TANAKA*

ABSTRACT

Damages due to beach erosion have increased in recent years in Japan. Changes in

the coastline of whole Japan were investigated

by comparison of topographical maps. It re- sults in that a net loss was 5,059 ha in 70 year

period before 1978 and 2,395 ha in 15 year period after 1978. An annual net loss rate had

been 72 ha per year before 1978 while it grew

160 ha per year after 1978. Erosion was severe along Hokkaido and the Japan Sea coast in

both former period and latter one. A comparison of the erosion rates and crustal movements

shows that the crustal subsidence in Hokkaido

was favorable to the severe erosion while the

crustal slight uplift on the Japan Sea coast did

not support the erosion. Primary cause of

long-term and overall erosion was considered

to be a net offshore sand drift.

KEYWORDS: Beach Erosion, Comparison of

Topographical Maps, Crustal Movement, Offshore Sand Drift

1. INTRODUCTION

Japan is an island country, and its people

have deeply related to the ocean since ancient

times. Geographically speaking, Japan is an

arc-shaped chain of islands located in the

northwestern, monsoon region of the Pacific

Ocean. Compared to other places in the world,

Japan is very susceptible to the harsh effects of

waves. In addition, it has also suffered damage

from countless tidal waves, or tsunami, because it is situated in the circum-Pacific earthquake region.

The land area of Japan is 378,000 km2 and

the length of the coastline is as long as 34,500

km. Compared with other countries, the coast

length in relation to the land area of the country is quite long. Therefore, it is easy to understand the deep relation between the country

and the surrounding ocean. The people of Japan have lived near the sea since the beginning

of agricultural culture, and the use of coastal

regions has grown along with the development

of civilization. However, at the same time,

lives and property have been lost countless

times due to the destructive energy of the

ocean. Then protection from coastal disasters

has increased in importance along with expanded use of the ocean.

In recent years, instances of damages due to

coastal erosion have gained frequent attention

in the mass media. Such treatments of the subject seem to suggest that the cause of the problem lies in the development of river basins

since the period of Japan's high economic

growth. Actually, however, coastal erosion

was a problem in some regions before the beginning of the high economic growth, and the

problems involved are very complex. So far,

the followings have been considered to be the

causes of coastal erosion: the construction of

large-scale structures on the coast, decreases in

the supply of sediment from rivers and seacliffs, ground subsidence, and the influence of

ocean waves and currents. However, it is still

impossible to identify these precisely at

present.

Though there is an accumulation of shoreline evolution of coasts data in which erosion

is severe, data of the same preciseness with re- gard to tendencies and amounts of erosion is

not available for the country as a whole.

Changes in the shoreline between two points

of time can be investigated through the following methods: comparison of topographical

maps, comparison of aerial photographs,

sounding measurements, etc. However, the

easiest method of investigating the entire

* Head, Seacoast Division

Public Works Research Institute

1, Asahi, Tsukuba, Ibaraki, 305, Japan

7
country with the same precision is the use of

topographical maps. Regarding sandy beach

shorelines, Koike (1977) used maps to investigate the 70 year period following 1890, and

Tanaka, et al. (1973,1974) used aerial photographs to investigate major changes in the

shoreline in the 20 year period following the

World War II. Since these reports were of more than a decade ago and damages around the whole country due to coastal erosion have increased re- cently, there is a need to grasp the situation of

coastal erosion for the whole country. For this

purpose, the newest maps as of 1992 were compared with those of 1978 to determine the situation of erosion since 1978. In addition, by comparing these results with the situation prior to

1978, it was also possible to investigate trends

in the erosion situation.

2. NATIONWIDE SURVEY OF THE CURRENT SITUATION OF COASTAL EROSION

2. 1 Method of the Survey

In this study, all of the maps were on a scale

of 50,000:1 of the Geographical Survey Institute. The newest maps as of 1978 (hereafter referred to as second edition maps) were compared with those available in 1992 (hereafter

referred to as current maps). In all, there were

607 maps that covered the whole country.

Also, I compared the earliest, or, first edition

maps with the second edition maps, focusing

mainly on sandy beach shorelines. Although,

in this comparison only 386 maps were used, it was possible to investigate nearly the entire

country and to compare differences between

the erosion from the Meiji Era to 1978 and that

from 1978 to 1992.

In Fig. 1 , the areas covered by Koike, or in

his compilation of the material taken from

Tanaka, et al., are shown along with those of

the present survey. From Fig. 1 , it can be seen

that the present survey covers a larger area than

these earlier studies. It should be noted, however, that the legend for that from 1978 to 1992

is not shown, since the entire country was covered in this study.

The first edition maps were published between 1887 and 1934, and the average date of

publication is 1905. On the other hand, second

edition maps were published between 1957 and

1978, with the average year being 1973. The

current maps were published between 1976 and

1992, with the average of 1986.

Fig. 1 Extention of Investigation Area

8
Table 1 Shoreline Change in Area

(Unit:ha)

1905- 1978 1978- 1992

Erosion Accretion Net Loss Erosion Accretion Net Loss

Total * 12,539 7,480 5,059 4,605 2,210 2,395

nv/ivivciiuu Hnkkairlo 4,534 1,869 2,665 1,921 631 1,290

1Aomori \\J\ I Ivl 1 676 60 616 182 94 88

Akita 447 123 324 153 43 110

1 dl 1 1 Cam CI LCI 84 53 31 65 43 22

1 vv cxvo 21 13 8 8 9 -1 Miv/ani ivi ly oyi 68 292 -224 79 52 27

Niiaata 847 444 403 221 121 100

Fukushima 1 UIXUvl III 1 IU 259 68 191 65 73 -8

1 UCX 1 drxl 300 306 -6 114 176 -62

Chiba 315 658 -343 249 127 122

Tnkx/n 0 0 0 36 79 -43

Kananawa 62 127 -65 37 26 11

Shi7i inka 327 867 -540 21 43 -22

Tnuflmfl I vjy di i la 305 36 269 26 16 10

Ishikawa 384 31 353 38 26 12

1Fukui Ui\UI 56 28 28 100 19 81

Airhi 298 198 100 40 25 15

1Mie V 1 1 w 196 200 -4 51 26 25

Wakfl\/^infl v v dfvci y cii i ici 38 20 18 16 20 -4

Kvntn I \y ulu 49 9 40 10 12 -2

Osaka 20 17 3 1 17 -16

H\/onr» nyuy*j 635 148 487 36 89 -53

Tottori 137 346 -209 106 42 64

.^himanp Ol III 1 BCll lw 51 118 -67 89 19 70

Okavama 165 105 60 31 5 26

Hiroshima 1 III vol III 1 Id 40 30 10 79 3 76

Vpmani i di i idy uuiiphiii 191 37 154 55 10 45

l_l 1 II 1 IC 296 97 199 53 24 29

Kanawa 1 \UUUVV d 61 102 -41 21 20 1 Toki 1 wf\UOI i<;hima III 1 Id 233 157 76 28 11 17

\VJOI II 114 123 -9 78 75 3

F-i iki in IV

a

i ui\uui\d 189 64 125 3 10 -7

Saga 39 1 38 3 1 2

Nagasaki 85 3 82 134 22 112

Kumamoto 17 1 16 7 3 4

Oita 356 97 259 90 8 82

Miyazaki 170 366 -196 95 46 49

Kagosima 474 266 208 264 144 120

Okinawa 341 332 9 1,117 597 520

*Total does not include Okinawa pref.

As to the intervals between publication of years - So> 1 regarded these intervals as a 70

first and second edition maps, its average was year Period UP to 1978 - Similarly, as to the in67.5 years and its maximum interval was 74 tervals between publication of second edition

9
and current maps, its average was 14.6 years

and its maximum was 18 years. So, I regarded

these intervals as a 15 year period from 1978.

The method of measurement was as follows.

First, the old maps and new ones were compared precisely and a drawing for measurement

was made by sketching coastlines that had

changed. Areas of coastal change are roughly

divided into cases of erosion, sediment accu- mulation, reclamation, dredging, and estuaries.

In this study, however, cases of estuaries, or

river mouths, were excluded since there are significant seasonal and temporary fluctuations in

the amount of sediment due to flood discharge

from river. Also, cases of reclamation and

dredging were excluded. A digitizer was used

to measure the areas as well as lengths and

widths of changed parts.

Resolution of the equipment used was 0.1

mm. But there were errors due to such things as the widths of lines in the drawings (0.2-0.3

mm), readings of the maps and orientation,

consequently, overall preciseness of measurements was equivalent to measuring the shoreline changes of about 20 m in cross-shore direction. Also, precision for area was approximately one hectare.

2.2 Results and Considerations

(1) Coastal Areas which Changed due to Erosion and Sedimentation

There are 39 prefectures with seacoast in Japan. Figure 2 shows each location of prefecture. The areas of erosion, sedimentation and

net loss of land are shown in Table 1 for each

prefecture. There are data of former 70 year

period and of last 15 years. Though the results

regarding Okinawa Prefecture are also shown

in the table, these were disregarded in calculating the total since there were regions that diverge greatly from the actual situation. In the

70 year period, the area of erosion for the whole

country was 12,539 ha. In contrast to this, the area of sedimentation was only 7,480 ha. This

indicates a net loss of 5,059 ha. In the 15 year

period, in contrast to the area of erosion for the

entire country that was 4,605 ha, the area of ac- cretion was 2,210 ha. This indicates a net loss

of 2,395 ha. In both periods, erosion exceeded

sedimentation and the tendency toward erosion

Fig. 2 Location of Prefecture

10
Table 2 Shoreline Change Considering Extention of Sandy Coast

OdllUy DudUi I 1905-1978 1978-1992

Erosion Accretion Net Loss Erosion Accretion Net Loss

/I V (km) (m) (m) (m) (m) (m) (m)

T/-\+'i 1 *

i oiai * 9,499.

1

13.2 7.9 5.3 4.8 2.3 2.5

LJ f\ l/l/Q | nOKKdIQO 1 ,99o.1 22.7 9.4 13.4 9.6 3.2 6.5

Momori ODD. / 18.4 1.6 16.8 5.0 2.6 2.4

MKIla 14b./ 30.5 8.4 22.1 10.4 2.9 7.5

y amagaia 49.o 17.0 10.7 6.3 13.2 8.7 4.5

iwaxe lOU.b 2.1 1.3 0.8 0.8 0.9 -0.1

IVIIydy 1 1 <57 C 1 D/ .D 4.1 17.4 -13.4 4.7 3.1 1.6

£9U.l 29.2 15.3 13.9 7.6 4.2 3.4

ruK.U5nima i A A 7 22.6 5.9 16.6 5.7 6.4 -0.7

IUdrdK.1 1 co.o 24.3 24.8 -0.5 9.2 14.3 -5.0

oi nud 1 Q 22.7 47.4 -24.7 17.9 9.1 8.8

1 \Jt\y\J 1 1 ^ Q.I I O.O 0.0 0.0 0.0 3.2 6.9 -3.8

i\dl IdydWd oy.y 6.9 14.1 -7.2 4.1 2.9 1.2

Qhi7iinl<a Ul H£m.\J\Jt\CL CCp?71 A)n 14.4 38.2 -23.8 0.9 1.9 -1.0

1 Uydll Id DO./ 46.4 5.5 40.9 4.0 2.4 1.5

IblllKdWd 1 QQ Q loo.o 20.3 1.6 18.7 2.0 1.4 0.6

iukui loo.o 4.1 2.1 2.1 7.4 1.4 6.0

Aicni 111.7 26.7 17.7 9.0 3.6 2.2 1.3

ivne 4^b.4 4.6 4.7 -0.1 1.2 0.6 0.6

waKayarna 135.0 2.8 1.5 1.3 1.2 1.5 -0.3

i\yoio 81 .8 6.0 1.1 4.9 1.2 1.5 -0.2

Ocal^a £ 1 .0 9.3 7.9 1.4 0.5 7.9 -7.4

nyuyu 31.4 7.3 24.0 1.8 4.4 -2.6

1 ouon 9o.l 14.7 37.2 -22.5 11.4 4.5 6.9

ormi iidiic £09.0 1.9 4.4 -2.5 3.3 0.7 2.6

VJKayaiTla IO/.D 10.5 6.7 3.8 2.0 0.3 1.7

nilUol III lid 1 Q7 O 2.1 1.6 0.5 4.2 0.2 4.1

t drnagucni A -t A C 4.6 0.9 3.7 1.3 0.2 1.1

CI III lit; oUo.o 9.8 3.2 6.6 1.7 0.8 1.0

rxdgdWd 19D.4 3.1 5.2 -2.1 1.1 1.0 0.1

i OKUsnima llo.o 20.2 13.6 6.6 2.4 1.0 1.5

rxuoni loU.U 6.3 6.8 -0.5 4.3 4.2 0.2

r UMJUlvd 1 QQ 7 10.0 3.4 6.6 0.2 0.5 -0.4

Saaa 54 6 7.1 0.2 7.0 0.5 0.2 0.4

Nagasaki 787.1 1.1 0.0 1.0 1.7 0.3 1.4

Kumamoto 141.8 1.2 0.1 1.1 0.5 0.2 0.3

Oita 190.8 18.7 5.1 13.6 4.7 0.4 4.3

Miyazaki 173.0 9.8 21.2 -11.3 5.5 2.7 2.8

Kagosima 754.8 6.3 3.5 2.8 3.5 1.9 1.6

Okinawa 732.9 4.7 4.5 0.1 15.2 8.1 7.1

Total does not include Okinawa pref.

11
Table 3 Shoreline Change in Length for 1978 - 1992

Changed Length Proportion to Sandy Beach

Erosion Accretion Total Erosion Accretion Total

(km) (km) (km) (%) (%) (%)

Total * ! 1,320.4 640.0 1,960.4 13.9 6.7 20.6

Hokkaido 521.7 198.8 720.5 26.2 10.0 36.2

Aomori 64.5 26.6 91.1 17.6 7.3 24.8

Akita 38.9 12.5 51.4 26.5 8.5 35.0

Yamagata 16.0 14.2 30.2 32.4 28.8 61.2

Iwate 3.1 2.9 6.0 3.1 2.9 6.0

Miyagi 27.2 16.3 43.5 16.2 9.7 26.0

Niigata 56.2 37.0 93.2 19.4 12.8 32.1

Fukushima 19.3 18.0 37.3 16.8 15.7 32.5

Ibaraki 25.7 26.1 51.8 20.8 21.2 42.0

Chiba 53.5 32.8 86.3 38.5 23.6 62.1

Tokyo 7.9 14.3 22.2 6.9 12.6 19.5

Kanagawa 14.8 10.8 25.6 16.5 12.0 28.5

Shizuoka 7.7 9.2 16.9 3.4 4.1 7.4

Toyama 8.7 5.6 14.3 13.2 8.5 21.8

Ishikawa 15.2 8.0 23.2 8.1 4.2 12.3

Fukui 31.1 7.3 38.4 23.0 5.4 28.4

Aichi 11.3 8.2 19.5 10.1 7.3 17.5

Mie 14.8 9.7 24.5 3.5 2.3 5.7

Wakayama 4.6 4.9 9.5 3.4 3.6 7.0

Kyoto 4.9 4.3 9.2 6.0 5.3 11.2

Osaka 0.2 5.9 6.1 0.9 27.4 28.3

Hyogo 13.0 25.4 38.4 6.4 12.5 19.0

Tottori 34.6 13.1 47.7 37.2 14.1 51.2

Shimane 26.8 5.4 32.2 9.9 2.0 11.9

Okayama 13.6 2.9 16.5 8.6 1.8 10.5

Hiroshima 25.4 1.0 26.4 13.6 0.5 14.1

Yamaguchi 14.9 4.1 19.0 3.6 1.0 4.6

Ehime 21.3 10.3 31.6 7.0 3.4 10.4

Kagawa 9.1 6.3 15.4 4.6 3.2 7.8

Tokushima 9.8 4.2 14.0 8.5 3.6 12.1

Kochi 26.3 20.4 46.7 14.6 11.3 25.9

Fukuoka 1.2 3.4 4.6 0.6 1.8 2.4

Saga 0.9 0.3 1.2 1.6 0.5 2.2

Nagasaki 43.4 8.9 52.3 5.5 1.1 6.6

Kumamoto 2.6 0.7 3.3 1.8 0.5 2.3

Oita 32.9 3.7 36.6 17.2 1.9 19.2

Miyazaki 26.0 11.3 37.3 15.0 6.5 21.6

Kagosima 71.3 45.2 116.5 9.4 6.0 15.4

Okinawa 250.8 131.7 382.5 34.2 18.0 52.2

*Total does not include Okinawa pref.

continues. The annual average loss was 72 ha/ these figures that the rate of loss of national

year in the 70 year period, while it was 160 ha/ land area is increasing extremely.

year in the 15 year period. It can be seen from Looking at prefectures separately, Hokkaido

12
shows the greatest erosion for both periods,

making up 2/5 of the figure for the entire country. Though Hokkaido also had the most sedimentation, since this was not as much as the

erosion, the amount of net loss area was large.

In fact, its net loss is more than half of that for

the country.

(2) Average Cross-shore Change

Each prefecture has different length of

coast, and there are also differences in the

lengths of sandy coasts. For this reason, comparisons of the extensions of sandy coasts were

attempted. Table 2 shows the results of dividing the changed areas in Table 1 by lengths of

sandy coasts that is based on a survey of Geographical Survey Institute in 1975 to get the av- erage cross-shore change. During the 70 year

period, there was an average net retreat of 5.3

m. During the 15 year period, there was an av- erage net retreat of 2.5 m. In other words,

sandy coasts receded an average of 1 m over a 6

year period in the 15 year period. Also, during

the 70 year period, prefectures with over 10 m

of sedimentation were Miyagi, Chiba,

Shizuoka, Tottori, and Miyazaki ; but during

the 15 year period, with the exception of

Shizuoka Prefecture, all of these switched to

net retreat due to erosion. Even in the case of

Shizuoka Prefecture, the average amount of

sedimentation decreased from 23.8 m to 1 m;

and it is thought that erosion will eclipse sedimentation in the future. During the 15 year period, the prefectures with large average net re- treat included Chiba, Akita, Tottori, Hokkaido,

and Fukui.

(3) Shoreline Length which Changed due to

Erosion and Sedimentation

Table 3 shows shoreline length changed due

to erosion and sedimentation, and the proportions of these two processes in sandy coastal

extension in the 15 year period. The total length

of erosion was 1,320 km while the sedimentation was only 640 km, and an advance to retreat

ratio becomes 0.48. The erosion area of 4,605

ha and sedimentation area of 2,210 ha also

gives almost the same ratio. Calculation with

the areas and the lengths show average widths

to be 34.9 m for erosion areas and 34.5 m for areas of sedimentation. In addition, 13.9% of

all sandy coast in Japan are subject to erosion,

while sedimentation is apparent in 6.7%. In

other words, in the last 15 years, about 20% of

sandy coastal areas have experienced erosion

or sedimentation of 20 m or more. Prefectures

with eroded sandy shoreline exceeding 20% include Hokkaido, Akita, Yamagata, Ibaraki,

Chiba, Fukui and Tottori. Furthermore, in

Yamagata, Chiba and Tottori prefectures, the

length of coastline changes of 20 m or more

caused by erosion or sedimentation exceeded

half of their sandy coast.

(4) Rates of Beach Erosion

According to Koike, the ratio advanced

coastlines to from 1890 to 1965 is 1.24. The

ratio for the 20 years following the World War

II is 1.05. According to the present study, the

ratio and erosion for the 70 year period, which

nearly corresponds with the earlier period just

cited above, is 0.57, while for the 15 year period it is 0.48. Though the results of both studies correspond in showing the tendency towards increased erosion, there is a wide discrepancy in the numbers. The principal reason

for this is that Koike only measured coastline

changes of more than 50 m in width for those

coastlines that had sandy coast more than 500

m wide and more than 5 km long. For instance,

the advance/retreat ratio for Koike's Region

II(see Fig. 1) is 84km/36km=2.33. In the

present study, however, the ratio for the total of

the Shimokita-Hachinohe Coast, the Sendai

Bay Coast and the Kashima-nada Coast is

1 16km/122km=0.95. There are large

defference between the Koike's and present

study. But, when the area west of the

Shiriyazaki Cape, which is north top of the Region II, is excluded and only areas with more

than 50 m of change in width are collected, the

result is 96km/50km=1.92. This latter figure is

closer to Koike's. This discussion leads us to

consider that the shoreline in areas of the type

which Koike investigated where there is a very

long and wide coast, is somewhat in favor of

accretion . As can be gleaned from the information obtained by comparing topographical maps as described above, it can be seen that there are obvious topographical changes in Japan. In particular, erosion greatly exceeds sedimentation, and

13
N

1905-1978

1978-1992

Advance/Retreat ratio

>1.0 Sedimentation

=1.0 No Net Loss

<1.0 Erosion

Japan Sea

Land uplift is taken as positive (shaded area)

Unit : mm/year

(after Kato)

1905-1978 0.84

1978-1992 0.75

1905-1978

1978-1992

Pacific Ocean

1905-1978

1978-1992

.0.9.7.

.0.69. oL

500km

J I I

Fig. 3 Vertical Crustal Movements at the Coast (1951-1982) and Advance/Retreat Ratio

is in fact becoming worse.

3. COMPARISON WITH EROSION AND

CRUSTAL MOVEMENT

Figure 3 shows advance/retreat length ratios

for large blocks of coast. In this figure, upper is

the one for the 70 year period and lower is for

the 15 year period. In all blocks and both periods the ratio is less than 1.0, that is, erosion has

exceeded accretion and there has been a net

loss. Further, in any blocks, the ratios became

lower. This tendency shows that beach erosion

in Japan has become severer. Looking at

blocks separately, erosion is severe in

Hokkaido and the coast along the Japan Sea.

Figure 3 also shows vertical crustal movements in Japan as deduced from tidal records

for the period 1951 - 1982 after Kato, et

al.(1983). Shaded area indicates crustal uplift.

It can be found that there had been rapid subsidence in northeast part of Japan. This fact agrees

well with the severe erosion in north part of Japan. In other part, there is no good agreement

between vertical crustal movements and beach

erosion. Especially, on the Japan Sea coasts,

though the vertical crustal movement was positive, the beach erosion was rather severe. In

addition, in the areas facing the Japan Sea, because the economic growth of them has been

not so high compared to the Pacific Ocean side,

there are few large-scale structures on the coast

nor so much development in river basins. After

all, it is concluded that the vertical crustal

movement might be one of the causes of severe

erosion of Japan, however, it is not the major

one and the mechanism of erosion is still uncertain.

4. TYPICAL EXAMPLES OF EROSION

Let us look at examples of coastal erosion

14
Japan Sea

Fig. 4 Location of Kaike Coast

and its major causes. As stated previously,

coastal erosion is a result of the complex interactions of many factors. Careful examination

is, therefore, required in considering possible

countermeasures

.

4. 1 Erosion from Old Days

(1) Kaike Coast

Kaike coast is located in the midwest part of

the Japan Sea coast. There are Kaike hot spring

and Yonago city behind it. Between Yonago

City and Sakaiminato City in Tottori Prefecture, there is an arc-shaped coastline lined with

white sand and pine trees

by the name of

"Yumigahama", which

means arc-shaped beach in

Japanese. (see Fig. 4) It is

considered that Kaike

beach had been formed

from the discharged sediment from Hino River

whose upstream basin had

been famous for iron production from iron sand. As

the iron production caused

a lot of sediment discharge

into the Hino River(see

Photo 1), the shoreline of

Kaike beach progressed,

for example, 180m in 30

years from 1868. However, as the iron production

Photo 1 Digging Decomposed Granite

for Separation of Iron Sand

declined around 1920 due to modernization,

the beach began to suffer from erosion. In particular, about 300 m of sandy beach has been

eroded away due to wave action at the front

beach of Kaike hot springs, with losses both of

hot spring wells as well as of hotels (see Photo

2). In view of countermeasures against erosion, this coast was one of the first coasts in Japan at which many countermeasures carried

out. First, jetties were constructed in 1947.

Once the beach was restored, but in a few years

it became unsuccessful. Next, seawall was applied. However, erosion could not be stopped.

Photo 2 Hotel is about to Collapse

15
Photo 3 Detached Breakwaters and "Tombolo" at Kaike Beach

Japan Sea

Before moving After moving( 1 977)

Fig. 5 Houses Moved Inland due to Erosion at Shimoniikawa Coast

Finally, detached breakwater was introduced

since 1971, then tombolo formed behind the

detached breakwater, and beach erosion

stopped (see Photo 3).

(2) Shimoniikawa Coast

The Shimoniikawa coast is situated near the

mouth of the Kurobe River which runs through

the eastern part of Toyama Prefecture. In this

region, villages are located along the arc of the

fan-shaped delta of the Kurobe River at which

underground water is plentiful. In this coast,

erosion has been a serious problem since ancient days due to sand and gravel moving in a westerly direction, and there are records that

villages have had to relocate inland several

times(see Fig. 5). Further, swells peculiar to

Toyama Bay are exceptionally huge, and

coastal disasters have occurred repeatedly in

conjunction with erosion. Still more, the slope

of the seabed is very steep, and there is no place

for coastal sand to accumulate on the coast.

Judging by above fact, it is thought that sediment is swept away to the deep ocean. So far,

seawall was completed but detached breakwater has been carried out mainly near villages as

16
1940 1960

Legend

/

o

Sekiya

Diversion Channel C O 00

Niigata West

Harbor

Shinano River

Fig. 6 Erosion and Major Causes in Niigata Coast

an emergent measure against erosion and high

wave.

(3) Niigata Coast

Niigata is located in the central part of the

Japan Sea coast. The Niigata Plain is one of the

biggest alluvial lowlands in Japan and has several sand dune ridges. The Shinano River,

which is the longest river in Japan, and the

Agano River run into the sea at Niigata city. It

is said that Niigata coast has been eroded since

1890's. Figure 6 shows the erosion history of

this coast. A meteorological observatory was

constructed 100m landward from the shoreline

in 1928, but it was collapsed in the water in

1949 (see Photo 4). Major causes of the beach

erosion have been thought to be 1) interruption

of sand drift in a westerly direction due to the

construction of breakwater of Niigata West

Harbor, 2)reduction in the amount of sediment

discharge from the Shinano River due to the

operation of Ookouzu flood channel and

3)ground subsidence due to extraction of

ground water for natural gas. For the countermeasure, seawall was constructed first. The

beach could not be restored enough untill the

detached breakwater was introduced. In stead

of detached breakwaters, however, artificial

reef method has been recently applied because

Niigata city, the major city along the Japan Sea,

is famous for sunset scenery.

4.2 Problem of Coastal Structure Interrupting

Littoral Sand Drift

The Shimokita-Hachinohe Coast is located

in the north Pacific coast of Honshu Island. In

Photo 4 Collapsed Meteorological

Observatory at Niigata Coast

17
Photo 5 Severe Erosion and Sedimentation around Misawa Fishery Port

Predominant Wave

Fig. 7 Beach Changes around the Port Interrupting Littoral Drift

this area, a northerly littoral drift is predominant. In Photo 5, while large advance of the

shoreline can be seen on the southern side,

there is severe erosion on the northern side.

Though seawalls and wave dissipating works

have been built, the area of erosion is gradually

expanding to the north direction. In 1987, sediment stretching 3 km in length and 150 m in

width accumulated on the southern side of the

breakwater. On the other hand, on the northern

side there was erosion for 6 km. Uda (1991)

shows a diagram shown in Fig. 7 for this kind

of situation.

4.3 Erosion near the Lee of Large Offshore

Structure

The Oarai Beach is situated on the northern

end of the Kashima-nada coast, which is located in the central Pacific coast. There is

Oarai harbor next to the beach, and in conjunction with expansion of the harbor, large-scale

breakwaters have been built. As a result, however, erosion problems have developed (see

Photo 6). Uda (1991) also shows a diagram

shown in Fig. 8 for this kind of situation. Because waves behind the breakwater become

smaller, a difference in wave height occurs between in the lee side of the offshore breakwater

18
Photo 6 Seawall Damage by Erosion at Oarai Beach

Littoral Drift

and out of it. This condition develops a circulating current that flows

toward inside of the harbor near

the beach. This current causes lit- toral drift into the harbor. On the

other hand, in the area out of the

lee of offshore breakwater, because littoral drift is induced by incident waves, the direction of lit- toral drift is out of harbor for this

wave condition. Therefore, Erosion occurs in the area where the

tip of the breakwater is projected

upon the shore in the direction of

the predominant waves, and this

area stretches in both directions

along the shore. On the harbor

side, however, as an effect of the

circulating current and calming of wave, sand

accumulates in the harbor and near the breakwater (see Photo 7).

5. CAUSES OF EROSION

As noted previously, though coastal erosion

is a phenomenon that has been observed since

long ago, its causes are complex and have not

yet been clarified. In this regard, let us consider

the movement of sand on the coast.

Predominant Wave

Circulating Current

Littoral Drift

Sedimentation

Sedimentation .Erosion

Fig. 8 Beach Changes in the Lee of Offshore

Breakwater

Figure 9 shows the relations involved in the

movement of sand along the coast. The

amounts of sand entering the rectangle in units

of time are: R of sediment from rivers, R from

rocky coasts, Bj of sand blown in from sur- rounding areas by the wind, Lj of sand carried

in by littoral drift, and Ot of sand carried in

from offshore by onshore drift. Conversely,

amounts of sand carried out of the region in

units of time are: B2 of by the wind, L2 by lit- toral drift, and 02 by offshore drift. When the

19
amounts coming in

and going out are

balanced, there is no

change in the

amount of sand

within the system;

but when the amount

going out exceeds

the amount supplied,

then there is a net

loss in the system

and erosion occurs.

It is supposed that

the effect of the imbalance of Bj and B2 is small in Japan.

From the macro scale point of view the

sand moved due to littoral drift Lt and L2 will remain nearshore. Th erosion and

sedimentation like section 4.2 and 4.3 by

the imbalance of L( and L2 is therefore

considered to be diminished as a whole

while there are local severe erosion and

sedimentation. Further, in spite of the

sediment discharge from river, there were

coasts that suffered from erosion from

early days as Shimoniikawa coast. And

the net loss of national land has been increased quickly. Additionally, the crustal

movement could not give enough explanation of erosion all over Japan. These

facts imply that there has been the imbalance of

Ol and 02

, that is, a net offshore sand drift, and

this has played important role in sand budget of

littoral zone. Therefore it must be noted that the hydraulic

mechanism in shallow water under waves and

currents should be investigated further in order

to understand the mechanism of the net offshore sand drift and to cope with it.

6. CONCLUSION

In Japan, beach erosions have become se- vere in recent years. Investigation by comparison of topographical maps shows that a net loss

was 5,059 ha in 70 year period before 1978 but

2,395 ha in 15 year period after 1978. It indicates that an annual net loss rate had been 72 ha

per year before 1978 while it grew 160 ha per

year after 1978. The length ratio of the sediPhoto 7 Beach Changes in the Lee of Oarai Harbor

River

Fig.

0i v 02

9 Sand Movement along Coast

mentation to erosion for the 70 year period is

0.57, while for the 15 year period it is 0.48.

While Japanese high economic growth began

around 1960, beach erosion was severe here

and there long before that. Further, although

the river basin facing to the Pacific Ocean was

developed higher than that of the Japan Sea, the

Japan Sea coast has suffered from severer ero- sion than that of the Pacific Ocean coast. Erosion was severe in Hokkaido and along the Japan Sea in both former period and latter one. A

comparison of the erosion rates and crustal

movements shows that the crustal movement

could not give a good explanation to the overall

characteristic of erosion. And considerations

from a macro scale point of view imply that the

net offshore drift is suspected of the cause of

net losses found in topographical maps. All

these results suggest the necessity of further investigation on hydraulic mechanism in shallow

20
water.

REFERENCES

Kato, T. & K. Tsumura, 1983. Vertical Crustal

Movements in Japan as Deduced from Tidal

Records (1951-August, 1982), Report of the

Coordinating Committee for Earthquake Prediction, 29, 368-380 (in Japanese).

Koike, K., 1977. The Recent Change of Sandy

Shorelines in Japan, Komazawa Geography,

13, 1-16.

Tanaka, N., H. Ozasa & A. Ogasawara, 1973.

Note of the Investigations on Change of Shoreline, Part 1 — the Investigation on Shoreline

Change on the Basis of Aerial Photographs --, Technical note of Port &Harbor Res. Inst., 163,

l-95(in Japanese).

Tanaka, N., H. Ozasa, 1974. Note of the Investigations on Change of Shoreline, Part 2 — the

Investigation on Shoreline Change on the Basis

of Aerial Photographs --, Technical note of Port

&HarborRes. Inst., 192, l-106(in Japanese).

Uda, T., 1991. Beach Erosion in Japan and its

Classification, Coastal Sediments '91, ASCE,

1856-1870.

ASSESSMENT OF HURRICANE OPAL IMPACT ON EAST PASS.

FLORIDA WITH AIRBORNE LIDAR 1

William E. Roper

Director, Civil Works Research and Development

U.S. Army Corps of Engineers

Washington, DC

ABSTRACT

Advances in remote sensing

technology are providing a number of

practical tools for application to Corps of

Engineer Civil Works field projects and new

directions in research programs.

Hydrographic survey operations using the

jointly developed U.S./Canadian Scanning

Hydrographic Operational Airborne LIDAR

Survey (SHOALS) is faster and less costly

than traditional methods. It is particularly

useful to rapidly assess bottom changes

following major storm or tsumani events. Its operating and system performance

characteristics are summarized. As part of the

rapid assessment response to hurricane Opal's

passage through East Pass, Florida the

SHOALS system was used to describe bottom

changes and navigation channel impacts.

Hurricane Opal's track and characters are presented along with the results of the

SHOALS survey and evaluation of storm

induced bathymetric changes in the area of

East Pass.

1.0 INTRODUCTION

The Corps of Engineers is involved

in virtually all navigation dredging

operations performed in the United States.

The Corps dredging mission entails

maintenance and improvement of 25,000

miles of commercially navigable channels

serving 400 ports, including 130 of the

nation's 150 largest cities. Connecting

waterways and inlets to the nation's ports

and harbors handle about 2 billion tons of

commerce each year. Emergency dredging

is often needed following major storms at

coastal inlets due to sediment relocation

induced by storm surge, wave action, and

ebb flushing currents.

The Corps dredges an average of 250

to 300 million cu yd of sedimentary material

at a current expenditure level of about $500

million per year in order to maintain and

operate the nation's existing navigation

system. Dredging is the single most costly

item in the Corps' Civil Works Operations

and maintenance program.

Coastal and navigation dredging

projects almost always require hydrographic

survey. This information is essential in the

planning, design, construction, operation

and maintenance of projects. During the

1980s research programs were initiated to

'Presented at the United States - Japan Natural Resources Program Workshop, Two Great

Tsunamis, University of Hawaii at Hilo, April 1- 5, 1996.

23
explore opportunities to decrease cost,

improve response time, and enhance product

quality in areas of hydrographic surveying,

and subbottom geology (Roper, 1995). This

paper presents the results of one of those

research programs and its application for

rapidly assessing storm induced damage

from Hurricane Opal to East Pass, Florida.

2.0 AIRBORNE LIDAR BATHYMETRY

In the 1980s, the U.S. Army Corps of

Engineers begin investigating technologies

that could augment existing survey

capabilities at comparable cost and provide

fast, accurate surveys. In March 1988, the

Corps began a cost-shared program with the

Canadian government to design, construct

and field verify an airborne LIDAR (Light

Detection And Ranging) hydrographic

surveying system. The Scanning

Hydrographic Operations Airborne LIDAR

Survey (SHOALS) development program

was implemented through a Memorandum

of Understanding under the U.S./Canadian

Defense Development Sharing Program.

Optech, Inc. of Toronto, Canada, under

contract was responsible for development of

the SHOALS system.

The SHOALS system provides the

capability for high spatial density utilizing

state-of-the-art LIDAR technology

consisting of a scanning laser

transmitter/receiver that produces 200

soundings per second. SHOALS operates

from a helicopter and includes a separate

ground based data processing system. It is a highly mobile system capable of rapidly

covering large areas. This unique capability

produces a more comprehensive

measurement of inlets and the adjacent near shore bathymetry. Output from SHOALS is

an accurate X, Y, and Z (position/depth)

from each laser (Lillycrop, Parson, etal,

1 994) sounding that is easily formatted for

most data analysis software tools such as Geographic Information Systems (GIS),

volumetric computations, or other

contouring and mapping systems.

The development approach began in

1988 with a conceptual design, followed by

detailed design, construction, laboratory

testing, field testing, and operational use.

2.1 LIDAR BATHYMETRY

TECHNOLOGY

The system operates by emitting

laser pulses that travel from an airborne

platform to the water surface where most of

the laser energy is back scattered to the

airborne receiver as illustrated in Figure 1

.

The remaining energy at the water's surface

propagates through the water column and

reflects off the sea bottom back to the

airborne detector. The time difference

between the surface return and the bottom

return corresponds to water depth.

As the light travels through the water

column and reflects off the sea bottom it undergoes scattering, absorption, and

refraction, which attenuate the return energy

and limit the maximum depth of penetration

or depth of bottom detection. The maximum

depth the system is able to detect is related

to a complex interaction of bottom radiance,

incident sun angle and intensity,

and water turbidity. As a rule-of-thumb, the

SHOALS system is capable of sensing

bottom to depths equal to two or three times

24
the Secchi depth. A Secchi depth is

calculated using an oceanographers tool, a black and white Secchi disk, that is lowered

by a line down into the water column. The

point where the Secchi disk is no longer

visible is the Secchi depth. Thus, if a Secchi

depth was measured to be 5 m then the

maximum depth of bottom detection would

be between 1 0 m and 1 5 m depending on the

composition of the suspended material in the

water column.

2.2 THE SHOALS SYSTEM

SHOALS is comprised of two

separate systems: the airborne system and

ground based data processing system. The

airborne system operates from a Bell 212

helicopter and performs the task of primary

data acquisition. The ground based data

processing system provides the data postprocessing to calculate position and depth

for each laser pulse. This design permits

SHOALS to be highly mobile system

capable of producing a 4 meter sounding

grid under normal operating conditions,

however, the scan pattern and survey speed

can be modified to obtain even higher or lower sounding densities. The SHOALS

system performance specifications are presented in Table 1

.

2.2. 1 Airborne System

The airborne system is divided into

three subsystems (Lilllycrop and Banic,

1993); Transceiver (TRS), Airborne

Positioning and Auxiliary Sensors (APASS),

and Acquisition, Control and Display

(ACDS). These combined with the groundbased Data Processing System comprise the

SHOALS system.

The transceiver subsystem consists

of the laser, scanner, and receiver. The

function of the TRS is to transmit laser

pulses in a defined scan pattern and receive

back scattered energy from these pulses to

produce laser depth soundings and aircraft

altitude information. The laser is 200 Hz

Nd:YAG operating in the infrared and green

frequencies. Returned laser energy is

detected using several optic sensors providing the ability to discriminate between

surface, bottom, and land returns.

The Aircraft Positioning and

Auxiliary Sensors functions are to collect

information from the Global Positioning

System (GPS), inertial reference system

(IRS), and video imagery system.

Differential GPS is used for horizontal

positioning and the IRS provides

information about aircraft attitude, including

roll, pitch, heading, and vertical

acceleration. Included as an auxiliary sensor

is a video camera to record a video image of

the areas being surveyed.

Central to the SHOALS system is the

ACDS, which provides and operator

interface and monitors and controls the

airborne system. The ACDS provides five

functions: data collection, operator interface,

pilot guidance, airborne depth processing,

and system integrity. The data collection

function acquires and manages all data as it flows through the system and records it on

high density magnetic tape at a rate of over

300 kbytes per second. The operator

interface

allows human interaction between the

operator and the system with access to all

elements of the airborne system. The pilot

guidance function provides aid to the pilot in

25
navigating to the survey site and along each

survey line. The airborne depth processing

function calculates and displays preliminary

water depth in real time, providing a means

for data quality checking during the survey

mission.

The last and perhaps most important

function is system integrity. This function

serves as the airborne system coordinator

which continually monitors and interrogates

communications between the various system

components. Without precise (billionths of

seconds) component communications timing

the system could not functions as a unit

(Parson, Lillycrop, etal 1995).

2.2.2 Data Processing System

The airborne system acquires a

tremendous volume of raw data during a

single mission. The Data Processing System

(DPS) is the hardware and software required

to post-process the LIDAR data. Its main

functions are to 1) import airborne data

stored on high density data tape;

2) perform quality control checks on initial

depths and horizontal positions; 3) provide

display and edit capabilities; 4) calculate

depth and position (XYZ) values for each

sounding; and 5) output final positions and

depths for each sounding.

The interface between the airborne

system and DPS is via the high density tape

containing the raw data acquired during the

survey mission. All of the data types (GPS,

IRS, and LIDAR returns, etc) are collected

at varying rates and recorded in

asynchronous format. The primary task of

DPS is to transfer the raw data from the

survey and store it in a data base which

requires some degree of preprocessing so

that the information can be synchronized

into a complete data set.

The DPS possesses a fully automated

capability to post-process the data and

update the database with corrected depth and

horizontal positions within the accuracies

presented in

Table 1 . The software accomplishes this by

identifying the surface and bottom returns

from the airborne data. Depths are determined by computing the differences

between the arrival times of the surface and

bottom returns and applying corrections for

depth biases associated with light

propagation, water level fluctuations, and

various inherent system characteristics

(Guenther and Thomas 1984a and Guenther

and Thomas 1984b). Sophisticated

modeling algorithms are used to predict and

apply corrections associated with these

biases (Guenther, 1985). Surface are modeled and removed so that depths can be

referenced to a common mean water surface.

Determining a mean water surface and

isolating the aircraft's vertical fluctuation

allows for an accurate estimation of the

aircraft height above the mean water level.

Applying tides then produces a depth

reference to a known water level datum such

as mean low water.

A manual processing capability

allows hydrographers to evaluate anomalous

data by providing display and edit functions

of sounding data and system parameters.

Video imagery of the survey area permits

visual scrutiny of the area to aid the

hydrographer in deciding whether to exclude

suspect data from further processing.

Output from the DPS is an accurate digital

26
data set of XYZ (positions/depths) for each

laser sounding that is compatible with most

GIS and other contouring and mapping

systems.

2.3 APPLICATIONS OF SHOALS

Since accepting the SHOALS system

in March of 1994 and following successful

field trials in Sarasota, Florida, it has been

used at project sites in the eastern U.S. and

at —project sites on the western U.S. These

project locations and sponsors are shown in

Table 2.

The SHOALS System has

consistently provided bathymetric soundings

at a normal density of four meters and

surveying rates of 9 Km2/hr with digitized

results produced within a day. The system

provides detailed inlet bathymetry which can

reveal features not produced by other survey

methods. It can also be used to evaluate

near shore sediment change behavior as a

result of seasonal change and storm events.

2.4. PLANNED SHOALS

ENHANCEMENTS

Over the last year our field experience with

the SHOALS system has identified four

areas for potential enhancement of its

capability. As a result additional applied

research will be focused on:

1) Three-dimensional positioning of

SHOALS using kinematic Global

Positioning System (KGPS). If SHOALS

can be positioned accurately with KGPS

then depth calculations would no longer

require tide measurements.

2) Improved data procession time. The

original specification was to process one

hour of airborne survey data in seven hours, a ratio of 1 :7. We have been working to

accelerate this performance specification and

currently we are at 1 :2.

3) Shallow water measurement capability.

The original performance specification was

for SHOALS to be able to measure a minimum depth of 1 .7 meters. However,

with enhanced signal processing we can now

measure at zero water depth and profile the

near beach area. We are currently testing

the accuracy of this new capability and will

have it permanently implemented in the

depth calculation algorithm by late 1996.

4) Sensor fusion. We are currently

evaluating other remote sensors that can be

added to the SHOALS platform to collect

useful data for managing coastal projects.

One such sensor is a hyperspectral sensor

that when fused with SHOALS depth data

can be used to accurately classify sea bottom

type such as seagrass, sand, mud, silt, rock,

and coral, in optically clear water.

3.0 HURRICANE OPAL

Hurricane Opal achieved hurricane

status on September 27, 1995 with sustained

winds of 75 mph just off the east coast of the

Yucatan Peninsula of Mexico as shown in

Figure 2. Opal made landfall early on

September 28th and moved in a zig-zag path

across the Yucatan Peninsula for three days

until it reentered the Gulf late on September

30th. It gathered strength over the warm

gulf waters for the next four days. Initially

Opal moved west across the gulf and then

turned sharply northeast and headed directly

toward the western panhandle of Florida.

Opal made landfall at 1700 central daylight

27
time (CDT) on October 4, 1995 on the

western Florida panhandle in the area of

Destin, Florida. Satellite imagery of Opal

just prior to landfall in Figure 3 show a compact classic hurricane form with heavy

rain bands and a well defined eye. The eye

passed west of East Pass making direct

landfall on Santa Rosa island near the town

of Mary Esther. Hurricane Opal generated

winds up to 67 m/sec (150 mph or 130

knots), classifying it as a great hurricane

(Category 4 on the Saffir-Simpson Scale).

The following statistics summarize this

storm:

Minimum Barometric Pressure

916 MB

Maximum Wind Speed

67 m/sec

Maximum Storm Surge

14.6 ft (4.45 m)(MLW)

During the storm, a 14-ft surge

pushed a tremendous amount of water

through the inlet and across Santa Rosa

Island into Choctawhatchee Bay. The

elevated bay, augmented by fresh water

runoff from the Choctawhatchee River

basin, drained out to sea through East Pass.

Considering the height of the surge, it is

likely that there were tremendous currents in

the inlet and between the jetties as the ebb

jet escape out to the Gulf.

Most of the coastal damage from

Opal was caused by storm surge rather than

from wind. Three surge effects can be seen

in an aerial photograph of East Pass taken

just after landfall (Figure 4): First, a narrow

over wash fan projects into East Pass near

the west end of the highway bridge. This

part of Santa Rosa Island was low and

provided little resistance to over wash.

Second, almost half of the west jetty was

over washed, with sand being carried from

Santa Rosa Island into the channel. Third,

Norriego point was breached just to the

northwest of the condominiums. Mobile

District engineers, who conducted a postOpal reconnaissance, estimated that bout 80

percent of the above-water portion of the

point was removed (Mr. L. Robinson, USAE

District, Mobile, unpublished

reconnaissance report).

4.0 SHOALS SURVEY AT EAST PASS

On October 10, 1995, just one week

after Opal, the SHOALS system was used to

survey East Pass. The survey was

completed in just over one hour covering

about one square mile and recording nearly

200,000 individual depth readings (Morang,

Irish, and Pope, 1996). The results are shown in Figure 5. Each grey band

represents a 1000 meter swath width form a SHOALS scan along the flight line.

Converging jetties mark the mouth of the

inlet. The state plane coordinates are in feet

and north is to the top.

The SHOALS generated results were

compared with project channel

requirements to determine changed dredging

requirements. There was also an evaluation

of risk to project structure, particularly the

jetties, as a result storm induced changes

within the inlet system. This analysis

included a comparison of the SHOALS data

with a 1990 survey (Morang, Irish, and Pope

1996). A graphic display of the differences

is shown in Figure 6. This clearly shows a

significant growth in the scour hole at the tip

of the west jetty. If it continues to deepen

28
the stability of the jetty tip maybe at risk.

There was also an accumulation of sediment

north of the spur jetty.

5.0 SUMMARY

SHOALS is a rapid, accurate survey

tool for accessing post storm changes in

bathymetry. This can assist in post storm

recovery planning and better allocation of

recovery assets, and lead to an improved

understanding of sediment dynamics of inlet

and coastal systems. The SHOALS system

is moving toward full deployment for Corps

of Engineer projects. For the last nine

coastal inlet surveys conducted there was a savings of $300K over traditional methods

and a better product produced. When fully

deployed savings in the range of over

$3 million is expected annually at Corps of

Engineers projects. Through cooperative

project applications for NOAA, Navy, Coast

Guard and others the savings to the Federal

government could be significantly greater.

Planned improvement will further expand

SHOALS capability, accuracy, and

timeliness.

5.0 REFERENCES

1 . Guenther, G.C., Airborne Laser

Hydrography, System Design and

Performance Factors, National Ocean

Service, Rockville, MD, p 385, 1985.

2. Guenther, G.C., and Thomas, R.W.L.,

"Propagation-Induced Depth Measurement

Biases and Peak Bottom Return Power

Relationships for Airborne Laser

Hydrography, " NOAA Technical Report

NOS 1062, National Oceanic and

Atmospheric Administration, U.S.

Department of Commerce, Washington, DC,

p. 121, 1984a.

3. Guenther, G.C., and Thomas, R.W.L.,

"Effectives of Propagation-Induced Pulse

Stretching in Airborne Laser Hydrography,"

In Proceedings SPIE Ocean Optics, Vol.

489, pp. 286-297, 1984b.

4. Lillycrop, W.J., and Banic, J.R.,

"Advancements in the U.S. Army Corps of

Engineers Hydrographic Survey

Capabilities: The SHOALS System,"

Marine Geodesy, Volume 15, 1993.

5. Lillycrop, W.J., Parson, L.E., Estep, L.L.,

Laroque, P.E., Guenther, G.C., Reed, M.D.,

and Truitt, C. L., "Field test results of the

U.S. Army Corps of Engineers airborne

LIDAR Hydrographic survey system," In

Proceedings U.S. Army Corps ofEngineers

1994 Training Symposium, Surveying and

Mapping Remote Sensing/GIS, New

Orleans, LA, pp SM:2A 1-5, 1994.

6. Morang, A., Irish, J.L., and Pope J., "Hurricane Opal Morphodynamic Impacts

on East Pass, Florida: Preliminary

Findings," U. S. Army Engineer Waterways

Experiment Station, Vicksburg, MS, 1996.

7. Parson, L.E., Lillycrop, W.J., Klein, C.J.,

Ives, CP., Orlando, S.P., "LIDAR

Bathymetric Applications for Surveying

Florida Bay," Journal ofCoastal Research,

1995.

8. Roper, W.E., "New Applications of

Remote Sensing Technology in Coastal

Engineering," Third Thematic Conference

on Remote Sensing for Marine and Coastal

Environments, Seattle, WA., 1995

29
Figure 1 . LIDAR Operating Principle

30
Table 1. SHOALS System Performance Requirements

Maximum Depth K 'd > 3 day

Kd > 4 night

Minimum Depth 0.9 meters

Vertical Accuracy +/- 30 cm

Horizontal Accuracy +/- 6 meters

Sounding Density 3-15 meters

Operating Altitude 200 - 1000 m

Scan Swath Width 14 aircraft altitu de

Operating Speed 0-50 m/sec

Operating Temperature 5 - 40 C

Data Processing 2 hrs of processing for 1 hr of data

Aircraft Bell 212

System Mob/Demob 8 hrs to install / 6 hrs to demob

V where K is the diffuse

attenuation coefficient of water

and d is depth. In clear water the

maximum depth is 40 m.

31
Table 2. Projects Surveyed by the SHOALS

System through March 1996

Sponsor Project

U.S. Navy Elgin AFB, FL and Gulf of California, CA

National Oceanic and

Atmospheric Administration

(NOAA)

Tampa Harbor, Florida Bay, FL

Corps of Engineers

Waterways Experiment Station

Longboat Key Beach, New Pass, Palm Beach Harbor,

Palm Beach IWW, Vero Beach, Ponce de Leon Inlet,

Canaveral Harbor, St.Augustine Inlet, FL

Jacksonville District Estero Island, Gasparilla Island, Key Biscayne, Miami

Beach, St. Lucie Inlet, Ft. Pierce Inlet, Hamlets Cove,

East Pass, Jupiter Beach, Sarasota Pass, FL

New York Harbor Morches Inlet, Shinnecock Inlet, Long Island Beach,

NY

Philadelphia District Hereford Inlet, Great Egg Harbor Island, Barnegate Inlet,

NJ

New England District Nantucket Beach, Martha's Vineyard Beach, MA

New Orleans District Mississippi River Ship Channel oulets to the Gulf of

Mexico, LA

Mobile District Mobile Bay Navigation Channel, AL

Galveston District Galveston Beach, Gulf Intercoastal Waterway, Houston

Ship Channel, TX

Los Angeles District Oceanside Inlet, Los Angeles/Long Beach Harbor, CA

32
Figure 2: Path ofHurricane Opal

Figure 3. Imagery of Opal just prior to landfall

33
Figure 4. Aerial photograph of East Pass after Hurricane Opal (north is to the top)

34
N513000

m 2 m

CO CO CO CO CO CO ro CO o

o

o o

o

o o o o

+ + +

N512000 +

N5TI000 +

N510000 +

N509000 +

N508000 +

N507000 +

N506000

m

CO

CO

o

8

East Pass, FL

Oct 1995

Post-Hurricane Opal

m

COCO

o

o

o

+ N513000

* + + N512000

..+' N511000

+ N510000

+ N509000

+ N508000

+ N507000

+ N506000

m_lCOCO

->l o

o

o

Figure 5. SHOALS survey of East Pass, Florida

35
03

Figure 6. Sea floor elevation changes between 1990 and 1995 at East Pass

36
EARTHQUAKE ENGINEERING

Mitigation of Lateral Flow of Liquefied Ground

by

Ikuo Towhata*

Yoshihiro Kogai**

Kei Amimoto***

and

Hendri Gusti Putra+

ABSTRACT

The lateral displacement of liquefied subsoil

during earthquakes is one of the causes of

damages to such facilities as river dikes and

lifeline networks. The present study aims at

developing measures which mitigates this lateral ground movement. A wall made of sheet

piles or compacted sand is installed in a model

deposit of sand and is shaken dynamically.

The observed behavior of a wall showed its

capability to prevent the lateral ground displacement. A theoretical study on such a wall

was conducted simultaneously in order to calculate the reduced displacement of liquefied

ground as well as the bending moment and

shear force in walls. Finally, a example analysis

was conducted to make sure the efficiency of

the walls in mitigating the lateral ground displacement. Both experimentally and analytically, a sheet-pile wall and a wall of compacted

sand exhibited a good mitigation of the ground

movement.

KEYWORDS: earthquake, liquefaction, flow,

shaking table tests, sand.

1. INTRODUCTION

The present text is concerned about mitigation

of lateral displacement of liquefied subsoil induced by earthquakes. Although an overall densification is the ultimate measure to prevent

lateral movement caused by liquefaction, there

is a situation in which it is not feasible. For

example, existing facilities in a developed area do not allow a compaction work under their

foundations. A financial restriction could be

another reason why a compaction is not possible. As long as the displacement is not so

significant, a limited extent of liquefaction may

be tolerated. In this respect, alternative measures are investigated in this study by which a

lateral flow of liquefied subsoil can be mitigated.

2. METHOD OF SHAKING TABLE TEST

A series of shaking table tests were carried

out in order to study the effects of embedded

walls which reduce the lateral flow of liquefied

sandy slopes. The studied walls were of two

types; a sheet-pile wall and a wall of a compacted sand.

Loose deposits of Toyoura sand were made in a container which measured 100cm in length

and 20cm in width. The thickness of model

deposits was variable in the horizontal direction, making the surface (Fig.l) gradient 15%.

The stress-strain behavior of liquefied sand

undergoing lateral flow is characterized by a peak strength followed by softening and a re- sidual state. This behavior can be reproduced

under a low confining pressure as occurs in

shaking table tests by making the density of

sand much lower than in situ. Hence, in the

present study, the void ratio of sand was about

1.05, making the relative density around -20%.

This low density was attained by first placing

wet sand in a container, and then submerging

it under water. The base shaking was generated

in the longitudinal direction of the model deposits. The amplitude of acceleration was 200

to 300 gals while the frequency was 4Hz.

* Professor of Civil Engineering, University of Tokyo, Tokyo, Japan

* Research Engineer, Fudo Construction

Company

** Former student, University of Tokyo

+ Andalas University, Sumatera, Indonesia

39
3. OBSERVED DISPLACEMENT

Fig.l illustrates the ultimate configuration of a

control case in which no mitigation was installed. This model deposit liquefied upon shaking and flew laterally until the surface became

level. No further movement occurred after this

configuration because the level surface was as- sociated with the minimum potential energy.

The grid of colored sand in Fig.l demonstrates

the subsurface deformation. The horizontal displacement is zero at the bottom, while taking

the maximum near the surface.

The second test was run on a model with a

sheet-pile wall. A model of a sheet pile was

made of an aluminum plate with a thickness of

0.5, 0.8, or 1.0mm. The ultimate deformation

of a model in Fig.2 was observed when the

thickness of a sheet-pile was 0.8mm. It is seen

in Fig3 that the elevation of the ground surface

is different between the front and the rear sides

ofthe wall. This difference induced a differential

lateral earth pressure along the wall, resulting

in distortion of the wall. It should be noted

that the installedwall could not completely prevent the soil displacement, although the displacement was made remarkably smaller in

Fig.2 than in the control case of Fig.l where

no mitigative measure was taken. The ground

surface on both sides of the wall was level in

the ultimate stage, because the minimum potential energy is attained when the surface was

level.

Fig3 indicates the ultimate displacement when

the central part of the model was compacted to

make the void ratio ranging from 0.75 to 0.79.

This void ratio is equivalent with a relative

density of 50 to 60%. The lateral movement of

sand was prevented by this compacted sand

wall, generating a difference in elevation of the

ground surface on the front and the rear sides

of the wall. The ground surface became level

after flow. It should be noted again that the

displacement in this figure is smaller than what

was seen in Fig.l without a remedial measure.

Nevertheless, there was still some displacement

in both sides of the wall. Finally, it is demonstrated in all figures that the ground surface

subsided near the upper end while uplift was

the case near the bottom of the slope.

4. EFFECTS OF WALL SIZE ON OBSERVEDDISPLACEMENT

Similar to the bending deformation of a cantile- ver beam, the differential earth pressure between

the front and the rear faces of the wall induces

bending of a sheet-pile wall. It is easily speculated that the bending stiffness of a sheet pile

has a substantial influence on the deformation

of the wall. This is demonstrated in Fig.4 where

the time history of the lateral displacement at

the ground surface adjacent to a sheet-pile wall

is compared for a family of the wall thickness.

A thicker wall has a greater stiffness, making

the displacement smaller. It is interesting that

the thicker wall completes its displacement at

earlier times than thinner ones do. Similarly, a

thicker sheet pile reduces the subsidence of the

ground surface near the upper end of a slope; see Fig.5.

The effects of the size of a compacted sandy

wall on the magnitude of ground flow were

similarly studied. The width of the compacted

domain was varied from 0cm, which is the case without mitigation, up to 15cm. Fig.6 demonstrates that the lateral displacement at the

ground surface near the compacted area decreased as the range of compaction increased.

Fig.7 reveals that the subsidence decreased simultaneously.

5. THEORY FOR STRUCTURAL ANALYSIS ON MITIGATIVE MEASURES

It was demonstrated in the preceding sections

that both a sheet-pile wall and a compacted

sandy wall can reduce the lateral displacement

of liquefied ground. It is, however, left unknown how to predict the reduced ground displacement as well as the bending moment, shear

force, shear strain and others in these mitigative

walls. Since the deflection of those walls is

caused by the difference of earth pressure between the front and the rear faces of the walls,

it is essential to calculate this earth pressure

difference in the prediciton of displacement.

Figs.l to 3 indicated that the elevation of the

ground surface became different after the lateral

flow, producing the earth pressure difference.

It is therefore important to calculate the displacement of ground movement induced by liquefaction.

40
No mitigation

« 1 m

Fig.l Ultimate deformation of control case

without mitigation.

Sheet -pile wall: 0.8mm

\

1 Y

« 1 m

Fig.2 Ultimate deformation reduced by a

sheet-pile wall.

Compacted sand

/

^~-—A-Jmm^ fT /

/ koxitfmm. - —/—r—7 7 ^ -J^k^A5cn\iM / J J

/

7

:*•'•'

.f*.i l&j ,,

1 m

Fig.3 Ultimate deformation reduced by

compacted sand wall.

41
£ 15

v

'_ Thickness of sheet -pile model:

Accel. =260-320 gals

ei;1.05

mitigation

•o 0.5mm

0.8mm a

• A> .. JL 0 m

m

\ _}A

2 3

Time, second

Fig.4 Effects of sheet-pile thickness on lateral

displacement of ground surface.

10 — »rm Subsidence

90cm -- jT-->

o2.

u 2

of

o

s

J?

6

8

Time, second

1 2 3

T

**9>

n. Thickness of sheet-pile model :

-

"8.

v. 'x 1 .0mm \«

V*..

- v 0.8mm

Fig.5 Effects of sheet-pile thickness on

subsidence of ground surface (at 10cm from

upper end).

42
Displacement _

Compacted

50cm 50cmE 15 - Width of compacted sand

S

E

o 10

Accel. = 220 -280 gals

e=1.00-1.06

2 3

Time, second

Fig.6 Effects of width of compacted soil on

lateral displacement of ground surface.

Subsidence

Compacted

;ii-.w.>;!: ,

35cm

65cmTime, second

£

a 1 o

c

•D 2

w

n

°-•• ... 15cm

~o~._. "10cm~ c> ^ W\Nocompaction

idth of compact

ed sand

:

Fig.7 Infects of width of compacted soil on

subsidence of ground surface (at 35cm from

upper end).

43
Towhata et al. (1992) developed a closed-form

solution of a flow displacement of ground

caused by liquefaction by using the principle

of the minimum potential energy. Towhata

( 1995) further developed this solution in order

to deal with such an irregular topography as shown in Fig.8. It is seen in this figure that

the analyzed subsoil is composed of a surface

unliquefied crust, a liquefied sandy layer, and

a stable base. The surface crust is modeled as an elastic column which moves together with

the liquefied layer. The liquefied sand is modeled as a viscous and constant-volume liquid.

Previous model tests indicated that this liquefied

layer develops a sinusoidal mode of lateral displacement (Fig.9). It was shown by previous

model tests as well that the lateral displacement

is continuous at the interface between the liquefied layer and the surface crust; i.e., no slip at

this interface.

The solution for an irregular topography is composed of a nodal equation;

ax k

4yH* ~7

ix

=0 (1)

where H and P denote the thickness of the

liquefied layer and the weight of the surface

crust per unit area, while y the unit weight of

the liquefied sand, E the elastic modulus of the

surface crust, T the thickness of the crust. F is

the lateral displacement at the surface and x is

the horizontal coordinate. The variation of H

with x coordinate is expressed as H=H+bx (2)

Finally, x

+ and x" in Eq. 1 stand for the positive

and negative sides of a node. This equation is

associated with the minimum potential energy.

Nodal equation for a sheet-pile wall

A nodal equation for a sheet-pile wall is going

to be developed in what follows by using the

principle of the minimum potential energy. A

sheet-pile wall is modeled by a cantilever elastic

beam in which the bending stiffness is denoted

as E

p

I and the lateral deflection by p. The

bottom of the wall at the unliquefied base is

fixed while the top is a free end (Fig. 10);

p = 0 and dp/dz = 0 at z=B (3)

d

2 p/dz2 =0and d3 p/dz* = 0

at z=B+H (4)

The strain energy stored in the sheet pile is

obtained by

Strain energy=LB+H-^

Vck

dz (5)

This is added to the analytical expression of

the potential energy of a liquefied ground

(Towhata, 1995) and the total energy is made

minimum by searching for a closed-form solution of displacement; F (variational principle of

mathematics). Consequently,

XI

L dxlx~ L n n

it

.+ r

(6)

(7)

\er4Ef

L dx\

+ m^UW^§-

x~ \_ it n\ it) jtr

Eq.5 gives the earth pressure difference which

is shown to be constant in the vertical direction.

The different elevation of the ground surface

between two sides produces a different over- burden after flow which, since the liquefied

sand is liquid, is equal to a net earth pressure.

Eq.6 is the volume compatibility between the

wall deflection and the soil movement.

44
Displacement: U +

Rg.8 Irregular topography.

45
U = F

FIg.9 Sinusoidal distribution of lateral displacement.

46
Mitigation of lateral flow of

liquefied subsoil

(1) Sheetpile wall

«- z=B*H

Fig. 10 Model of a sheet-pile wall.

(2) Wall of elastic unliquefied soil

z=B*H

z=B

Fig.l 1 Model of a elastic wall of compacted

sand.

47
Deflection

H

p(z)

Lateral

earth

pressure : |I

t

H,

I

-Z=B*H

Liquefied

sand

^Z=B

Fig. 12 Bending of shet pile undergoing large

distortion.

O

Q.

0.6

Assessed net earth pressure

vs. different elevation of

ground surface, H1

$> 0.A

Z3

C/i

</> 0)

I—CL

B 0.2 i—o

a> 4—

0)

H = 7.84kN/m3 -H,

o

D

0.0 h^" ° J

a 1 .0mm sheet pile

0.8mm

o 0.8mm

0.8mm

compacted 15cm backfil

2 A 6 8

Difference in elevation, , cm

Fig. 13 Relationship between assesed lateral

pressure and elevation difference.

10

48
Nodal equation for a compacted soil wall pressure in the vertical direction (Eq.6),

A compacted soil column is modeled by an

elastic body which is of a shear modulus of Gs and width of Ls (Fig. 11). The strain energy

was calculated on this wall and, eventually, a

nodal equation was derived as

6. OBSERVED EARTH PRESSURE DIFFERENCE

2(Bending moment)

M

^ ~

(B+H-z)2

+{p(fl+tf)-p(z)}2

Fig. 13 compares the earth pressure assessed

experimentally by using Eq.12 against the elevation difference at the ground surface. The

bending moment at the bottom of a wail (z=B)

was used. It is interesting that the earth pressure

difference is reasonably equal to the elevation

difference times the unit weight of liquefied

sand. Hence, the liquefied sand around the wall

appears to behave as a liquid.

7. EXAMPLEANALYSIS ON MITIGATIVE

MEASURES

Eq.6 suggested theoretically that the earth pres- sure difference exerted on a sheet-pile wall is

constant in the vertical direction and is produced

by the different elevation of the ground surface

around the wall.

To study the earth pressure experimentally,

sheet-pile walls employed in model tests were

equipped with strain gauges which made it possible to monitor the bending moment and eventually the lateral earth pressure exerted on walls.

When this earth pressure was studied, a care was taken of the large deformation of the sheet

pile which increased the bending moment along

the wall. Fig. 12 illustrates the bending of a

sheet pile due to lateral pressure. When the

distortion is significant, the lateral pressure

comes to have a vertical component near the

top of the wall and makes an additional bending

moment near the bottom. The bending moment

is calculated below by considering the horizontal

and vertical components of the normal earth

pressure;

Bending moment=-J jl cos 8(rj -z)dr]

-l?wn9{p(Ti)-ftz)}dn (ID

where |X denotes the lateral earth pressure which

is normal to the face of the pile, 0 is the angle

of rotation, and T) is a local vertical coordinate.

Considering 6 to be reasonably small,

cos 0« land sin 0 * tan 0 « dp\dr\. By further

considering a uniform distribution of the earth

An example calculation was performed on the

mitigative effects of walls installed in a liquefiable subsoil. Fig. 14 illustrates a modeled situation in which an embankment rests on a level

liquefiable deposit. Zero displacement is as- sumed on both lateral ends of the model. The

analysis of Case 1 has no mitigative measure.

Case 2 has a berm as a counter weight over a 200m interval. Case 3 has a sheet pile wall in

addition at the foot of the embankment. This

wall penetrates into the unliquefied base and

its bending stiffness is Ep

Ip= 35300 kNmVm.

Case 4 has one more sheet-pile wall at the end

of the berm.

When the state ofshaking and flow is maintained

for a sufficiently long period of time, the lateral

movement of the subsoil continues until the

minimum potential energy is attained. The ultimate displacementatthisminimum-energy state

is indicated in Fig. 15. Although Case 1 without

mitigation appears to have less significant displacement than others, it actually has the largest

displacement of the embankment (x>300m).

The counter weight in Case 2 reduces the displacement of the embankment to 75%, while

an additional displacement is generated in the

berm itself. The sheet-pile walls in Case 3 and

Case 4 made the displacement negligible.

The ultimate displacement in Fig. 15 did not

clearly indicate the efficiency of a counter

weight. Actually, a berm decelerates the lateral

ground movement by increasing the total mass

of the moving soil. Since liquefied sand be49
Ultimate lateral displacement in embankment sheet pile model

0 100 200

Horizontal coordinate, x, meter

300

Fig. 15 Ultimate displacement in model

subsoil.

Si

4)£

15

10 c

t>Eo

o

~L 5

in

-1 1 1 rTime history of surface displacement at the foot

of an embankment

Case 1

Case 2

Case 3 4

5 10 15 Time, second

20

Fig. 16 Calcualted time history of lateral

displacement

fc4

%

E

a, j c&

2 2

v> n

Time history of subsidence!

qq 1 at the top of an embank- I ment

Case 2

Case 3 Case_4 -

5 10 15 Time, second

20

Fig. 17 Calcualted time history of subsidence

ofembankment.

50
comes unlikely to flow further at the end of

shaking, due to regained shear strength (Towhata, 1995), the ground flow stops then. Therefore, it is important to pay attention not only to

the ultimate displacement as in Fig. 15 but the

development of flow with time.

Figs. 16 and 17 demonstrate the time history of

lateral displacement at the foot of the embankmentas well as the time history ofthe subsidence

of the embankment. It is interesting that Case

2 with a berm has a remarkably smaller displacement than Case 1. Case 3 and Case 4

with sheet-pile walls are of even smaller displacement. Thus, it seems that a berm and sheetpile walls are very efficient measures to mitigate

the lateral displacement of liquefied subsoil.

8. CONCLUSION

An experimental study was made of the mitigation of lateral flow due to liquefaction by

using such subsurface walls as sheet piles and

compacted sand. These measures prevent the

lateral movement of ground when it is subjected

to a differential earth pressure and distorts laterally. A structural analysis on these wall together

with a counter-weight berm was made possible

by using a principle ofminimum potential energy. Example analyses revealed the efficiency

of these measures.

9. ACKNOWLEDGMENT

Model tests on sheet-pile walls and a compacted

soil wall were conducted with a financial support

provided by the Association for the Development of Earthquake Prediction in Tokyo. This

aid is deeply appreciated by the authors.

REFERENCES

Towhata,I., Sasaki,Y., Tokida,K., Matsumoto,H., Tamari,Y., and Yamada,K. (1992):

"Prediction of Permanent Displacement

of Liquefied Ground by Means ofMinimum

Energy Principle," Soils and Foundations,

Vol.32, No.3, pp.97-116.

I.Towhata (1995): "Liquefaction and Associated Phenomenon," Theme lecture, IS-Tokyo

•95, Tokyo.

51

Evaluation of Liquefaction Potential using the CPT

by

Richard S. Olsen* Joseph P. Koester* and Mary E. Hynes*

ABSTRACT

Prediction of liquefaction resistance using in situ

test data has progressed, over the last 15 years,

from techniques based on the Standard Penetration

Test (SPT) to the Cone Penetrometer Test (CPT).

Current CPT-based techniques for prediction of

liquefaction resistance typically only rely on

correlations to the cone tip (or end-bearing)

resistance. This paper presents an improved

technique for prediction of liquefaction resistance

which uses both ofthe primary CPT measurements

(cone and sleeve friction resistances).

KEYWORDS: CPT, Liquefaction, Normalization,

Sands, Silt correction, Silts, SPT

1 INTRODUCTION

This paper will describe an updated technique for

estimating liquefaction resistance that uses both

Cone Penetration Test (CPT) measurements (i.e.,

cone and sleeve friction resistances). This

technique was originally developed in 1982

(Olsen, 1984) and improved in 1988 (Olsen, 1988)

and 1995 (Olsen & Koester, 1995). All other

CPT-based techniques use only the cone resistance

to estimate liquefaction resistance (Seed &

De Alba, 1985, Robertson & Campanella, 1985,

Ishihara, 1985).

CPT sleeve friction resistance is employed in the

current approach because it is an index of high

strain behavior. If a soil has an unstable structure,

the passage of soil around the cone tip will disturb

the initial soil structure; the sleeve will measure the

effect of high strain. The cone resistance is an

index of soil structure strength and the sleeve

friction resistance is an index of high strain

strength.

2 LIQUEFACTION DEFINED

The classic definition of liquefaction involves

breakage of the grain-to-grain contacts in a loose,

saturated honeycomb sand (due to the earthquake

shearing action) and the temporary suspension of

sand grains in water. The consequence of

liquefaction was originally defined as "flow

failure," but over the last 20 years a given strain

level (e.g., 10 percent double amplitude cyclic

strain, or 5 percent strain potential) has been used

as a failure criterion. Loose to medium dense

sands can experience liquefaction, however, loose

sands exhibit more strain potential than do medium

dense sands. Loose sands that liquefy can cause

large slope movements, whereas medium dense

sands might deform only slightly.

In the 1960's, liquefaction resistance was typically

determined using expensive cyclic triaxial

laboratory tests. However, undisturbed sampling

of sand is very difficult (e.g., loose sands density

and dense sands may dilate during sampling).

Also, natural in situ variability is more complex

than can be generalized with a few tests on

undisturbed soil samples. For these reasons, the

late Professor H. Bolton Seed began to emphasize

the SPT for liquefaction resistance determination

in the early 1970's.

3 NORMALIZATION OF PARAMETERS

Normalization ofparameters is important for CPTand SPT-based liquefaction resistance

determination techniques because it corrects for

numerous trend effects. This section describes

stress normalization to determine equivalent CPT

and SPT measurements at a vertical effective stress ofone atmospheric pressure (atm) (=100 kPa or ~ 1 ton/ft

2).

Earthquake Engineering & Seismology Branch,

Geotechnical Laboratory,

USAE Waterways Experiment Station,

Vicksburg,MS 39180

53
ofone atmospheric pressure (atm) (» 100 kPa or =1

ton/ft

2

).

3.1 CPT normalization

Recent research has significantly impro/ed our

understanding ofthe exponent relationship of cone

resistance with vertical effective stress (Olsen, 1994). It is now known that CPT

measurements (actually all geotechnical properties) are exponentially influenced by vertical effective

stress. The difference between an exponentially

based sand strength and linear based sand strength

(i.e., ov tan <b) decreases with increased ov level.

Normalized cone and sleeve friction resistances are described below:

r

sle -A-l , y

(2)

was used for normalization. The CPT stress exponent, c, can be determined using the CPT soil

characterization chart in Figure 1. Alternatively,

for depths between 4 and 10 meters, the stress exponent can be roughly estimated as 0.6 for sand

and 1.0 for clay.

3.2 SPT normalization

The normalized Standard Penetration Test (SPT)

blow count (N,)60 is defined as:

W)« = ^60 " = Nm ^

where

N60 = Measured SPT blowcount (blows per

foot) adjusted to correspond to a

driving energy equivalent to 60% of

theoretical maximum free-fall

hammer energy

(N,)60 = Normalized SPT (N60 at ov =l atm)

Cn = SPT normalization factor n ( = stress exponent for the SPT

ov = vertical effective stress (atm units)

Rf

= 100 (3)

where

Vertical effective stress (atm)

qcie = Normalized cone resistance (using a

variable exponent)

qc = Measured cone resistance (atm)

C = Cone resistance stress exponent

file = Normalized sleeve friction

% = Measured sleeve resistance (atm)

S = Sleeve stress exponent

(approximately equal to c)

Rf = Friction ratio

The normalized cone resistance, qcle , has a "1"

subscript to represent normalization to one

atmosphere vertical effective pressure

(approximately 100 kPa or 1 ton/ft

2 (tsf)); the "e"

subscript signifies that a variable stress exponent

The field-measured SPT blow count (N) must be

converted to an equivalent N60 either based on an

understanding of the hammer type or field SPT

energy measurements. The stress exponent, n, in

Equation 4 was originally established based on

SPT chamber tests in the laboratory at varying

chamber confining stresses (Marcuson &

Bieganousky, 1977,Gibbs&Holtz, 1957). These

tests were performed using a short drill rod and a

constant 2 meters depth of drilling mud. The

constant borehole fluid pressure (due to the

constant 2 meter column of drill mud) reduced the

overall confining stress surrounding the SPT

sampler at high chamber stresses (Olsen, 1994).

This borehole fluid influence results in calculated

stress exponents, n, which are too low compared to

field stress conditions. New evidence (Olsen,

1994) indicates that the SPT stress exponent, n, is closer in value to the CPT determined stress exponent value, c.

54
4 HISTORICAL FIELD-BASED

TECHNIQUES FOR PREDICTION OF

LIQUEFACTION RESISTANCE

The industry standard for field occurrence-based

liquefaction resistance prediction for the last 25+

years has been the SPT technique (Seed and Idriss,

1971, Seed & De Alba, 1986). The SPT-based

liquefaction technique still provides a good

estimate of liquefaction resistance. This section

will describe current SPT and CPT techniques for

estimating liquefaction resistance stress ratio

(Cyclic Stress Ratio, or CSR).

4.1 SPT liquefaction technique

Professor H. B. Seed developed the classic

relationship between the normalized SPT blow

count (N,)60 and the CSR causing liquefaction in

the field as shown in Figure 2. Increasing fines

content (i.e., silts in sand) generally increases

liquefaction resistance (in all but very loose sands).

One means of accounting for fines content

influence on liquefaction resistance is to determine an equivalent clean sand SPT blow count (N^g,,

using the percent passing the #200 sieve as shown

in Figure 3 and using Equation 5.

tfU* = W« + ^

(5)

where

(Nlc)6o = Silt-corrected clean sand

normalized SPT blow count

AA^ = Silt correction for blow count

(Ni)6o = Normalized SPT blow count

4.2 CPT liquefaction techniques

Most historical CPT techniques for estimating

liquefaction resistance only rely on the cone

resistance measurement. The classic cone

resistance-based CPT liquefaction technique is summarized in Figure 4 (Seed & De Alba, 1986).

The only CPT-based liquefaction resistance

prediction technique using both cone and sleeve

friction CPT measurements was developed by

Olsen (1984, 1988, 1995). Contours of

liquefaction resistance CSR shown in Figure 5

(Olsen, 1988) were established on a CPT soil

characterization chart based on: 1) cyclic

laboratory test results and 2) trends of

CPT-predicted SPT equivalent clean sand blow

counts. The unique feature of this CPT-based

technique is that soil index tests are not required to

predict liquefaction resistance. Other techniques

require soil gradation tests to determine percent

passing the #200 sieve or D50 to estimate the

liquefaction resistance.

5 UPDATED CPT TECHNIQUE FOR

PREDICTION OF LIQUEFACTION

RESISTANCE

The proposed CPT-based technique for prediction

of liquefaction resistance is an improvement on the

method depicted in Figure 5. Three principle steps

involved in the development and application ofthe

technique are described in the next three sections.

5.1 Correlating cyclic laboratory results to

normalized CPT data

Cyclic laboratory tri axial and simple shear

strengths from nine projects were correlated to

normalized CPT parameters from nearby CPT

soundings. The cyclic laboratory strengths

(obtained using a failure criterion of 5% strain)

were converted to normalized CSR proportionate

at 15 equivalent uniform load cycles (appropriate

for a magnitude 7.5 earthquake) and adjusted to an

equivalent vertical effective stress of 1 atm.

Normalized CPT parameters (qcle and Rf ) were

determined from nearby CPT soundings for each

corresponding laboratory data value. These qci e and Rf values were located on the CPT soil

characterization chart and the corresponding

normalized laboratory CSR values were reported

at each point as shown in Figure 6. Circles bound

each CSR value in recognition of the fact that no

sand layer is uniform; the laboratory liquefaction

CSR should fall somewhere within the circles.

The trend ofthese laboratory values indicates that

normalized liquefaction resistance increases with

friction ratio, at least within the soil mixture

(i.e., sand mixtures and silt mixtures) and clay

classification zones.

55
5 .2 Equivalent strain-based cyclic strength ofclay

Clays have the potential for straining during

earthquakes. Figure 6 contains results of several

laboratory cyclic tests on clay (three test results

located in the clay zone are marked with "A").

These tests show that normally consolidated clays

may potentially develop 5% cyclic strain if cyclic

shear stress approaches the static strength. A static

undrained strength c/p ratio (undrained strength

divided by consolidation vertical effective stress)

for normally consolidated clay is typically equal to

0.3 (Olsen, 1995). An earthquake induced CSR of

0.26 when compared to the clay c/p ratio of 0.3

represents an induced shear stress of 80% and

normally consolidated clays will deform at these

stress levels. Normally consolidated soft clays

should therefore be expected to have equivalent

liquefaction resistance CSR between 0.25 and 0.30

as is reflected in Figure 6.

5.3 Establishing CPT-predicted equivalent clean

sand SPT blow count

Figure 7 shows the concept used to construct

equivalent clean SPT blow count contours on the

CPT soil characterization chart. Trends of

liquefaction resistance can be estimated if trends of

equivalent clean sand SPT blow counts are known.

The first step required is to plot contours of

CPT-predicted normalized SPT blow count (N^o

(illustrated with solid lines in Figure 7

(Olsen, 1994)). These (N^ contours have

corresponding equivalent clean sand liquefaction

resistance ratios (i.e., cyclic stress ratio, CSR) also

shown in Figure 7. The silt correction for blow

count (AAL ) becomes larger with increased fines

content (and/or decreasing D50). CPT-based soil

classification can be used to estimate the fines

content for the purpose of estimating &Nliq as shown in Figure 7. The silt corrected blow count,

(Nlc)60, contours, which equals N, + ANliq , are

illustrated with dashed lines in Figure 7 and bend

away from the (N,)^ contours within the silty sand

portion ofthe chart and have larger values of CSR

(than underlying (N ,) 60 contours).

The concepts illustrated in Figure 7 to determine

the equivalent clean sand SPT blowcounts (Nlc)60 were implemented in Figure 6. These (Nlc)60 contours (dashed lines) in Figure 6 bend into the

region of the chart where normalized laboratory

CSR's are available. The CSR contours agree with

the laboratory CSR data points in the soil mixture

zone. The CSR contours were established based

only on laboratory CSR within the clayey soil

mixture and clay zone.

5.4 Final CPT correlation for prediction of

liquefaction resistance

The final CPT chart for estimation of liquefaction

resistance CSR is shown in Figure 8, based on the

discussion in sections 5.1 through 5.3. These

SPT-based CSR and laboratory cyclic test-based

CSR trends match within the "Sand mixtures" zone ofthe CPT soil characterization chart. These

CSR contours were extended down into the soil

mixture and clay region of the chart using trends

from the cyclic laboratory test results. These

contours improve the 1988 version (Olsen, 1988)

as a result of improved understanding of stress

effects, availability ofmore laboratory cyclic data,

and better CPT-based soil classification.

6 COMPARISON OF CPT-BASED

APPROACHES

This section will compare the differences between

the "cone resistance only"-based liquefaction

evaluation approach and the updated liquefaction

technique using both CPT measurements. These

two techniques will also be compared using a unique plotting procedure.

6.1 CPT liquefaction resistance prediction

technique comparison

The updated CPT liquefaction technique shows

that as friction ratio increases, the predicted

liquefaction resistance also increases (for a given

qcle value). However, cone resistance-based

techniques (i.e., Seed technique) infer a constant

liquefaction resistance for all friction ratio levels.

The cone resistance-based techniques, therefore,

must use soil indices (such as percent passing #200

sieve or D50) to approximate the soil type effects.

As stated earlier, sleeve friction resistance is an

index ofthe soil structure sensitivity and as such is an important contribution for prediction of

liquefaction resistance.

56
6.2 Plotting the improved CPT based liquefaction

technique onto the Seed CPT liquefaction chart

The CSR contours on the qc, to Rf liquefaction

resistance correlation chart in Figure 8 can be

replotted as Rf contours on a CSR to qcl chart with

the results shown in Figure 9. An example can

illustrate how this chart-to-chart transformation

occurs. The Rf = 0.5% contour (shown at the

bottom of Figure 10) was established using a

vertical projection of the Rf = 0.5% in the chart at the top of Figure 10, upward through the CSR

contours. The intersection of the CSR contours

with Rf at the top chart provides the q cl and CSR

values to establish the Rf line in the lower chart of

Figure 10.

Also shown in Figure 9 is the technique developed

by Seed and De Alba (1986) so that these two

techniques can be compared. Seed's technique

uses contours of soil indices to determine

liquefaction resistance whereas the Olsen

technique uses contours of friction ratio (Rf)

(which is the ratio of the two CPT strength-based

measurements). In general, increasing friction

ratio (for a given cone resistance level) indicates a

higher lateral stress condition or higher large-strain

strength. In either case, the Olsen technique shows

a higher liquefaction resistance with increased

friction ratio for a given normalized cone

resistance. While these two different sets of

contours are not parallel (in Figure 9), the Olsen

technique was designed to predict liquefaction

resistances equal to the SPT liquefaction

technique.

The Olsen technique does not require nearby soil

borings (for soil index determination), but

confirmation SPT blow counts and related soil

indices will always be required for technique

confirmation. However, this technique does have

the potential for reducing the number of required

borings and soil samples for a project.

CPT-based soil classification lines (from Figure 1)

were superimposed onto Figure 9 with the results

shown in Figure 1 1 . Also shown is Seed's contour

for 10% fines content sand. The Olsen and Seed

contours representing 10% fines content cross at two points and are relatively close considering the

technical independence of the approaches.

Examples using the Olsen and Seed techniques are shown in Figure 12. Low fines content sands

(i.e., fines < 5%) generally have friction ratios of

0.3 to 0.5 % and medium dense sands have

normalized cone resistances of approximately 90

arm (900 kPa). This typical medium dense sand

corresponds to a point on the chart representing a CSR of approximately 0.19 using the Olsen

technique. This point on the chart also falls

between the D50 = 0.4 and 0.8 mm contours for the

Seed technique. The D50=0.4 to 0.8 mm contours

represents low fines content (i.e., < 5% fines)

sand.

7 CONCLUSIONS

An improved CPT-based technique for estimating

liquefaction resistance is presented. This

technique uses both cone and sleeve friction

measurements. It allows a more refined prediction

of liquefaction resistance compared to techniques

that use only the cone resistance.

8 ACKNOWLEDGMENT

The data presented herein, unless otherwise noted,

were obtained from various projects at Waterways

Experiment Station of the United Stated Army

Corps of Engineers. Permission was granted by

the Chief ofEngineers to publish this information.

9 REFERENCES

1 Ishihara, K. (1985) Stability of Natural

Deposits During Earthquakes, Proceedings,

11th International Conference on Soil

Mechanics and Foundation Engineering

(ICSMFE), San Francisco, Vol. 1, 321-376

2 Olsen, R. S. and Koester, J. P. (1995).

Prediction of Liquefaction Resistance using

the CPT, Proceedings, International

Symposium on Cone Penetration Testing - CPT '95, Linkoping, Sweden, Oct. 1995

3 Olsen, R. S. and Mitchell, J. K. (1995). CPT

Stress Normalization and Prediction of Soil

Classification, Proceedings, International

Symposium on Cone Penetration Testing - CPT '95, Linkoping, Sweden, Oct. 1995

57
4 Olsen, R. S.(1995). Prediction of Clay

Strength using the combination of Cone and

Sleeve Resistance, Proceedings, International

Symposium on Cone Penetration Testing - CPT '95, Linkoping, Sweden, Oct. 1995

5 Olsen, R. S. (1994). Normalization and

Prediction of Geotechnical Properties using

the Cone Penetrometer Test (CPT), Technical

Report GL-94-29. U.S. Army Engineer

Waterways Experiment Station, Vicksburg,

MS. August 1994

6 Olsen, R. S. (1994). Normalization and

Prediction of Geotechnical Properties using

the Cone Penetrometer Test (CPT),

Ph.D. Dissertation, University of California,

Berkeley, CA.

7 Olsen, R. S. (1988). Using the CPT for

Dynamic Response Characterization,

Proceedings, Earthquake Engineering and

Soil Dynamics II Conference, 1 988, ASCE.

8 Olsen, R. S., (1984). Liquefaction Analysis

using the Cone Penetrometer Test,

Proceedings, Eighth World Conference on

Earthquake Engineering, San Francisco.

9 Robertson, P. K. and Campanella, R. G.

(1985) Liquefaction Potential of Sands using

the CPT, ASCE Journal of Geotechnical

Engineering, Vol. Ill, No. 3, 384-402

10 Seed, H. B. and De Alba, P. (1986) Use of

SPT and CPT Tests for Evaluating the

Liquefaction Resistance of Sands,

Proceedings, In Situ 86 Specialty Conference,

Blacksburg, VA, Geotechnical Special

Publication No. 6

1 1 Seed, H. B., Idriss, I. M. and Arango, T.

(1983). Evaluation of Liquefaction

Potential Using Field Performance Data,

Journal of Geotechnical Engineering

Division, ASCE, Vol. 109, No. 3, March

1983, pp. 458-492.

12 Seed, H. B. and Idriss I. M. (1971) Simplified

Procedure for Evaluating Soil Liquefaction

Potential, ASCE Geotechnical Journal, Vol

97, SM9, 1249-1275

58
iooo4 _i i i i i_

CO

-—

'

'c

E

<DO

c

03

-t—i CO

'COCDi_

<Dc

o

o

"O

<uN

E

s

4-

100

10-

5

-

4

Cone resistance1

, stress exponent

(for cone

resistance

normalization)

100

Use the best possible ( SCN= -1 cone resistance normalization technique; the variable stress exponent technique (Olsen, 1994) (Olsen & Mitchell, 1995)

provides good results for shallow (e.g., 1 meter)

to deep conditions (>60 meters)

10

0.1

4 5 6 7 0.1

10

Friction Ratio (%)

Use the best available technique for cone resistance normalization

Linear normalization

is adequate for hand calculations (Olsen and Douglas, 1981)

(Douglas, Olsen & Martin, 1981)

Constant stress exponent

normalization (for sands)

is adequate for many

situations (Robertson,

1988, 1989, 1994)

Variable stress exponent is the most advanced stress normalization technique,

required for shallow and deep situations and

soil mixtures (Olsen, 1984, 1988, 1994)

(Olsen and Mitchell, 1995)

qc1L = _=~ q c is=

(St)

o. 6

1c1e /= \C (av ) c

Figure 1. CPT soil characterization chart having SCN and stress exponent contours

(Olsen & Mitchell, 1995)

59
Figure 2. Seed technique for estimating liquefaction resistance based on the SPT

(Seed, Idriss & Arango, 1983)

10

Percent Passing the #200 Sieve

Figure 3. SPT silt correction blow count (AN„.) to fines content (Olsen, 1988)

60
t—i—«—i—i—i—'—i

1—i—

r

M>7.5 earthquake*

_L _L

40 80 120 160 200

Modified Con* Penetration Resistance, - Isf

240

Figure 4. Seed technique for CPT-based prediction of liquefaction resistance (Seed & De Alba, 1986)

1000- 800

600-

400-

w

o 200-

COE

a>c 100-

80-

60-

(1 40-

u

cr

ance

20-

V)

'in o

a. 10-

8-

Cone 6-

0) 4- o

©o 2-

CPT procedure tor estimating the

cyclic shear stress ratio to achieve

at least 5% shear strain, ((i^,/*,)

),

,

lor a vertical effective stress of 1 tsf and

15 equivalent shear stress cycles

SILT MIXTURES

0 °:t f(S>

'S ~*» fit! «»,Wi

CtAYS The ((T„/a<

)5x), from this technique lor sandy

classifications have the same representation as from

the SPT liquefaction technique (Seed, et.al. 1985)

0.2 0.4 0.6 0.8 , 2 4 6 8

Corrected Friction Ratio (%) in terms of tsl

1 FR, =1*100=-^- 100 = too -W (of-'

Figure 5. CPT prediction of liquefaction resistance using both CPT measurements (Olsen, 1988)

61
1UUU

100-

8*cBin 1

01c8

u

a

oz

10-

Approximate CPT predicted equivalent clean sand

blow count

(Nio)eo

( Nl)60 = 25

(N1

)60 = 15

(N1 )60 = 8

Measured

normalized CSR

from cyclic laboratory tests

CPT predicted normalized SPT blow counts {N,)B0 (Olsen 1994) -1

1 1 1—i—i—r-r

0.1

1

1.0

(N1 )60 = 4

(^60 = 2

-I 1—I—1111

10.0

Normalized Friction Ratio (%)

Figure 6. The components for determining the CPT liquefaction technique using CPT predicted SPT

blow contours and cyclic laboratory test results

Friction Ratio (%) (Log scale)

Figure 7. Illustration showing how to construct the CPT predicted equivalent clean sand SPT blow count

(Olsen, 1988, 1984)

62
1000 I I I

>*

0-I—

03

E

X

2

CL

CL

03

(D

'c

E-4—

03,

0O

c

CD

-»—

<

to

0

c

o

O

"D0

,N

"CDF

4-

100

2-

10

Olsen & Koester(1995)

CPT predicted normalized liquefaction

resistance Cyclic Stress Ratio (CSR)

0.1

i

1—i—i—i—

r

5 6 7 8 9

i

1 1—i—i—

r

4 5 6 7 8 9

100

-8

-7-6

_5

-4

-3

-2

Q_ -

-9 mlu

^

-8 0

O -7

-6 03

-5

1 * (/>

"(/> -4

-3 0

c

o -2 O

a0KJ

"J

-8 E

-7 o

-6

-5

-4

-3

-2

0.1

1 10

Friction Ratio (%)

Figure 8. Final relationship for prediction of liquefaction resistance cyclic stress ratio (CSR) using both

CPT measurements

63
0.7

if)

O

0.6

a:

u o

CO

CO

CD

55

o 0.4

"o

O

<DO

£ 0.3

"co CO

CD

a:

.2

0 2

-t—> o

CD

f 0.1

0.0

t 1 1 r 1 1 1 r t 1 1 r 1 1 1 r i 1 1 r

The H. Bolton Seed CPT liquefaction technique

(Seed & De Alba, 1986)

(using cone resistance and soil indices)

| % fines >35

D

50 (mm) = 0.1

-15 -10 <5

#• » 0.2 0.25 0.25 0.4 0.8

OA

J I I L J I I I I I I I I I I I I I I I I I I 0 50 100 150 200 250

CPT Normalized Cone Resistance q c1 (atm)

(one atmospheric pressure (atm) = 100 KPA = 1 ton/ft2

)

Figure 9. Determining liquefaction resistance based on the CPT normalized cone resistance and friction

ratio (from Olsen & Koester, 1995) and with comparison to the Seed & De Alba (1986) technique (the latter based on normalized cone resistance and soil indices from soil samples).

64
1000 _l I I I I I I I I _l I I I I I I I

Rf =0.5%'

Olasn& KoesIer(199S) qc1 to RfCPT chart

with contours of liquefaction CSR (Figure 7)

Sets of qc

-j and CSR

are determined along

a vertical line projection

of Rf = 0.5 % through

intersections of the CSR contours

> 0.1 0.15 0.2 0.3 CPT predicted CSR

I /I T I I I I 1 10

Friction Ratio (%)

CPT qc

-| based

Liquefaction CSR chart (Figure 8)

50 100 150 200

Normalied Cone Resistance

250

Figure 10. Illustration of the conversion of CSR contours on the qc1 versus R, chart to the R, =

contour on the CSR versus qc1 chart

65

0.5 %
0.7

CO

o

i

1 1 1

1

1 1 1 1

1

1 1 1 r

Soil classification contours were developed

using the Olsen & Koester (1995) CPT

liquefaction contours and the CPT

based soil classification contours

r 0.6 l-from Olsen & Mitchell (1995)

CPT based

(Olsen, 1988)

sands with low

Normally / \

fines content

consolidated/ \ (fines =*\0%)

soft clay

t—i—

r

t 1 1 r

0.5

0.4

CO

C£

CO

CO

CD

55

o

"o

O

CDO

c

TO

"to CO

CD

Qd

c

o

o

CO 4—

CD

I 0.1

0) sz

S a> QLl_ CO O

c

O d) p

E

E E

in inV

(/)

E

CMd

ii dn "D o T3 o

C in c

TO Q LL (0 Q

0.3

0.2

o

0\se° C?

Seed technique

Olsen technique

Contours

representing

10% fines

0.0 J I I L 1 J I I L 1 J I L J I I L J I I L

0 50 100 150 200 250

CPT Normalized Cone Resistance q c1 (atm)

(one atmospheric pressure (atm) = 100 KPA = 1 ton/ft2

)

Figure 11. Superposition of the CPT soil classification lines on the CSR versus qc1 chart based on contours

of friction ratio. Note the comparison between 10% fine content sands contours for the Olsen and Seed

techniques.

66
0.7

0.6 -

0.5 -

0.4

0.3 -

0.2

n 0.1

0.0

t 1 1 r i r i 1 1 r

The H. Bolton Seed CPT liquefaction technique

(Seed & De Alba, 1986)

(using cone resistance and soil indices)

1 % fines >35

D^mm) = 0.1

-15 -10 <5

r » 0.2 0.25 0.25 0.4 0.8

0.A

Example (The Seed technique)

q c1 = 80 tsf and D50 = 0.4 (fines < 5%)

CSR, =0.18

Exampl Olsen technique)

q c1 = 120 atm (i.e. 120 tsf or 12 MPa)

and Friction Ratio (R f ) = 0.3 %.

Remember R t = (fs / qc

)1 00

using sleeve resistance (fs

) and

cone resistance (qc ) thus CSR 1 = 0.22

J I I L J I I l_ 0 50 100 150 200 250

CPT Normalized Cone Resistance qc1 (atm units)

(one atmospheric pressure (atm) = 100 kPa = 1 ton/ft2

)

Figure 12. Example determination of normalized liquefaction resistance CSR using the Olsen and Seed

CPT-based techniques

67

ESTIMATION OF EFFECTS OF LIQUEFACTION- INDUCED GROUND FLOW

ON BRIDGE FOUNDATIONS

by

Keiichi Tamura n

, Yoshio Ninomiya 2) and Tadashi Hamada 3)

ABSTRACT

The Hyogo- ken Nanbu Earthquake of January 1995

caused liquefaction in wide area. Especially, the

liquefaction induced the ground flow with the

movement of sea walls along the waterfront, and

this ground flow caused large deformation of bridge

foundations. This paper summarizes a feature of the

liquefaction— induced ground flow and the results of

analysis on the ground movement and the ground

flow force acted on a bridge foundation.

KEYWORDS: Bridge Foundation, Ground Flow,

Hyogo- ken Nanbu Earthquake, Liquefaction

1. INTRODUCTION

The Hyogo- ken Nanbu Earthquake of January 1995

caused widespread liquefaction in the reclaimed

land. Furthermore, the liquefaction induced the

ground flow with the movement of sea walls, and

this ground flow caused large deformation of bridge

foundations. In Japan, the liquefaction- induced

ground flow has been recognized since the Niigata

earthquake of 1964 from an engineering point of

view, however it was the first time that the ground

flow caused the extensive damage to the newly

constructed highway bridges.

This paper presents a feature of the

liquefaction- induced ground flow and an estimation

of the ground movement and the force acted on a bridge foundation by the ground flow.

2. OUTLINE OF LIQUEFACTION- INDUCED

GROUND FLOW

Where the soft ground underwent liquefaction, the

resulting decrease of bearing capacity of the ground

and the ground subsidence caused various damage.

Besides that, near the water's edge, the liquefaction

caused the movement of sea walls and ground flow,

which resulted in the damage to the structures

behind sea walls. Photos 1 and 2 show the examples

of damage caused by ground flow.

Photo 1 shows damage to a parking lot on the north

shore of Uozakihama. This photo shows that the sea

wall shifted out into the watercourse followed by

the subsidence and movement of the ground behind

it. Photo 2 was taken on the north shore of Port

Island. In Photo 2, it appears that the quaywall in

the foreground moved to the watercourse and the

quaywall in the background did not move, however,

the quaywall in the background also moved about

30cm, which was clarified by the survey of bridge

location.

3. GROUND MOVEMENT BASED ON AERIAL

PHOTOGRAMMETRY

It may be possible to judge the occurrence of

ground flow by the movement of sea walls near watercourses as seen in Photos 1 and 2, however, it

is difficult to make judgments of this kind for

locations further from watercourses. Even close to a watercourse, it is impossible to quantitatively

estimate the amount of movement of the ground

simply by inspecting the damage to structures.

The aerial photogrammetry was employed to estimate the amount of ground displacement. An

outline of the procedure is as follows:

1) Head, Ground Vibration Division, Public Works

Research Institute, Ministry of Construction

2) Deputy Head, Structures Division, Civil

Engineering Research Institute, Hokkaido

Development Bureau

3) Research Engineer, Ground Vibration Division,

Public Works Research Institute, Ministry of

Construction

69
Step 1. Selection of the Target Points

The target points selected for aerial photogrammetry are points such as lane markers of roads

that have not been dislodged before and after the

earthquake by factors other than the ground flow.

Step 2. Photographic Inspection of Coordinates of

the Target Points Before the Earthquake

Step 3. Photographic Inspection of the

Coordinates of the Target Points After the

Earthquake

Step 4. Calculation of the Movement Vector

The difference between the coordinates of a target

point measured in steps 2 and 3 is defined as the

movement vector.

In order to study the effect of ground flow on bridge

foundations, the target points were set along the

Route 5, Bay Shore Line of the Hanshin

Expressway, and separate survey was performed to

find movement vectors of the expressway bridge

piers. The movement vectors of the bridge piers

were found in the same manner as the case of

ground movement vector. The as— built drawings

and the GPS measurements were used to obtain

bridge pier coordinates before and after the

earthquake, respectively. The difference in the

coordinates of each target point before and after the

earthquake may be composed of three constituents:

displacement caused by ground flow, displacement

caused by diastrophism, and measurement error. The movement vector found in this study excluded

the diastrophism.

Figures 1 and 2 show the results of estimating the

movement vector of the ground and bridge piers.

Figure 1 shows the area near the north shore of

Rokko Island. The quaywall in the east- west

direction (A- A') and the quaywall in the

north- south direction (B-B') moved to the north

and the west, respectively, and the ground behind

the walls moved north— west direction. Directly at the edge of the watercourse, the ground moved

more than 2 m, and a bridge pier moved

approximately 0.9m. The maximum ground

displacement in this area was about 3m. The

amount of ground movement and bridge pier

generally decreases as the distance from the

watercourse increases.

Figure 2 shows the area near the south shore line

of Uozakihama, which lies opposite to the north

shore of Rokko Island. At this area, the southward

movement of the east- west quaywall (C-C) was

accompanied by the southward movement of the

ground and bridge piers behind this quaywall.

Similar phenomena occurred on other artificial

islands, however, no superstructure on the Hanshin

Expressway felt down under the effects of ground

flow.

4. ESTIMATION OF GROUND FLOW FORCE BY

BACK ANALYSIS

As an example of back analysis of ground flow force,

the case of Shin- Syukugawa Bridge on Route 5,

Hanshin Expressway is here presented.

As Figure 3 shows, Shin- Syukugawa Bridge is a 3- span continuous steel slab Gerber box girder

bridge crossing a watercourse. The bridge was

completed in 1993. The pier P-134, which was

analyzed in this study, was a 4— column steel rigid

frame pier and was supported by cast- in- place

concrete piles with diameter of 1.5 m and 34 m

long. The foundation was constructed in the

reclaimed soil and was supported by a diluvial layer.

Extensive liquefaction occurred around the bridge,

and the P-131 and P-134 moved approximately

0.5 m and 1 m toward the waterway with the

movement of sea walls, respectively. The

longitudinal component of the movement of P-134

was 0.7 m. On the piers P- 131 and P- 134, all the

bearing supports were destroyed and the girder felt down from the supports, however, it remained on

the piers.

As shown in Figure 4, the foundation was so

idealized that a rigid footing is supported by piles

that are supported by soils, considering nonlinear

properties of pile bodies and the ground. Figure 5

shows an analytical model of Shin- Syukugawa

Bridge. The second layer from the surface was

considered to be liquefied. At first, we assumed that

the first layer on the liquefied layer acted lateral

force on the foundation with the ground flow, and

70
calculated the relationship between the lateral force

and displacement, which is indicated by a fine line

in Figure 6. Based on this assumption, the

maximum lateral force is limited to the passive

earth pressure of the non- liquefied layer, which

was estimated as approximately 3,200 tf, however, even if this maximum load was considered, the

amount of residual displacement of the bridge pier

could not be accounted.

Then, in addition to the passive earth pressure of

the non- liquefied layer, we assumed that some

portion of overburden pressure of liquefied layer

also acted on the foundation, and the thick line in

Figure 6 was obtained. This overburden pressure

corresponds to the pressure acting on piles when

slurry goes through between piles in the liquefied

layer. As seen from Figure 6, the movement of

P-134 may be accountable by considering that the

passive earth pressure of non- liquefied layer and a portion of overburden pressure of liquefied layer

acted on the foundation. From the five cases of back

analysis, the contribution ratio of overburden

pressure was estimated as 0.2.

5. CONCLUSIONS

An outline of the liquefaction- induced ground flow

was presented. The ground movement caused by

the ground flow was estimated by comparing aerial

photos taken before and after the earthquake, and

the maximum movement was found to be

approximately 3m. The ground flow force acted on a bridge foundation was estimated by back analysis

using a nonlinear foundation model. From the

analysis of this study, the ground flow force may be

estimated as the sum of the passive earth pressure

of the non- liquefied layer and 20% of the

overburden pressure of the liquefied layer.

REFERENCES

1) Committee for Investigation on the Damage of

Highway Bridges by the Hyogo-ken Nanbu

Earthquake, 1995: Report on the Damage of

Highway Bridges by the Hyogo-ken Nanbu

Earthquake, 1995 (in Japanese)

2) Hamada M., Isoyama R. and Wakamatsu K.: The

1995 Hyogoken- Nanbu (Kobe) Earthquake,

Liquefaction, Ground Displacement and Soil

condition in Hanshin Area, Association for

Development of Earthquake Prediction, 1995

3) Japan Road Association: Reference for Applying

the Guide Specifications for Reconstruction and

Repair of Highway Bridges Suffered Damage from

the Hyogo- ken Nanbu Earthquake of 1995, 1995

71
Photo 1 Damage to a Parking Lot at Uozakihama
2ao

JJS

2$/

100m

Note : Numbers at arrows and in circles represent residual displacements

of the ground and bridge piers, respectively.

Figure 1 Residual Displacements of the Ground and Bridge Piers at Rokko Island

73
74
^ 36,000

/ KOBE

q ©

a- «• SO 1. 0^- cT

' bar

as

©

s

ft sr.

©

s

® © @ OSAKA \

I 7 . 700 g- .y? ,,P ft. /s

T &oJLjg.« s.20o^ft_^«

P-134 PIER MOVED 1m \

P- 131 P-130

/ PIER MOVED 0.5m

Figure 3 General View of Shin- Syukugawa Bridge

Vo m-Ho

A/W|

a/w|

A/Wf

AAM

aaM

A/Wi

aaM

A/wf

A/W|

AA/Vp

(a) Analytical Model

AXIAL FORCE

i

k H

AAM

AAM

AAM

AAM

AAM

AAM

kv

Pnu

ULTIMATE PULL-OUT FORCE

ULTIMATE PUSH-IN FORCE

aan-1 kv

VERTICAL DISPLACEMENT

Ptu

PASSIVE EARTH PRESSURE

HU LATERAL DISPLACEMENT

(b) Lateral Force vs. Lateral Displacement

(c) Vertical Force vs. Vertical Displacement

0

4c h 4>v CURVATURE

(d) Bending Moment vs. Curvature

Figure 4 Idealization of Pile Foundation for Nonlinear Analysis

75
3.7m

15.3 m

4.0 m

3.55 m

0.9 m

4.6m

0.8m

2.7m

1.0m

2.2m

Bs Non-Liquefied Layer

1 Liquefied Layer

Bs N=15

As N=50

As N=15

Dg N=10

Dc EES

Ds N=50

Dc N=15

Ds N=28

Passive Earth

Pressure

Overburden

Pressure

i()=1.5m

l=33.9m

n=52

Figure 5 Analytical Model of Shin- Syukugawa Bridge

ft YIELD HORIZONTAL STRENGTH

* ULTIMATE HORIZONTAL STRENGTH

5000

OVERBURDEN

PRESSURE

PASSIVE EARTH

PRESSURE

0.0 0.2 0.4 0.6 0.8 1.0 LATERAL DISPLACEMENT AT PILE TOP (m)

Figure 6 Relationship between Lateral Force and Displacement at Pile Top

76
Preliminary Guidelines for Liquefaction Assessment using Shear Wave Velocity

by

Ronald D. Andrus 1 and Kenneth H. Stokoe, II 2

ABSTRACT

This paper presents preliminary guidelines for

assessing the liquefaction resistance of soils

using small-strain shear wave velocity. The

guidelines are based on field performance data

from 17 earthquakes and in situ shear wave

velocity measurements at over 40 different sites

in soils ranging from sandy gravel with cobbles

to profiles including silty clay layers.

Additional data are needed from denser soil sites

shaken by stronger ground motions to further

validate the proposed liquefaction potential

boundaries.

KEYWORDS: Building technology; earthquake

engineering; in situ measurements; seismic

testing; shear wave velocity; soil liquefaction.

1. INTRODUCTION

Currently there is no widely accepted procedure

for assessing the liquefaction resistance of

granular soils using small-strain shear wave

velocity, Vs- A number of Vs-based procedures

have been proposed during the past decade.

These procedures were developed from

analytical studies (Stokoe et al. 1988; Andrus

1994), laboratory cyclic triaxial test results

(Tokimatsu et al. 1992), or a limited amount of

field performance data (Robertson et al. 1992;

Kayen et al. 1992; Lodge 1994).

The use of Vs as a field index of liquefaction

resistance is justified since both are influenced

by density, confinement, stress history, and

geologic age. The advantages of using Vs

include:

• V s can be accurately measured in situ

using a number of techniques such as the

seismic crosshole test, the Seismic Cone

Penetration Test (SCPT), or the SpectralAnalysis-of-Surface-Wave (SASW) test;

• Measurements are possible in soils that are hard to sample, such as gravelly soils, and

at sites where borings or soundings may

not be permitted, such as many landfills;

• Measurements can be performed in small

laboratory specimens, making direct

comparisons between laboratory and

field behavior possible;

• Vs is directly related to small-strain shear

modulus, G max, a required parameter in

analytical procedures evaluating dynamic

shearing strain in soils; and

• For large earthquake magnitudes and long

durations of shaking the cyclic shear strain

needed for liquefaction decreases and

approaches the threshold strain in sand

(= 0.02%), thus making it possible to conduct more analytical evaluations of

liquefaction using Vs and Gmax as basic

parameters (Dobry et al. 1981 ; Seed et al. 1983).

The two main limitations of using Vs to evaluate

liquefaction resistance are the lack of a sample

for identifying non-liquefiable fine-grained soils,

and its high sensitivity to weak interparticle

bonding, caused by aging or cementation, which

is eliminated at large strains. Thus, a limited

amount of drilling and penetration testing should

be performed to identify weakly cemented soils

and non-liquefiable clayey soils.

This paper presents preliminary guidelines,

currently under development (Andrus and

Stokoe 1996), for assessing liquefaction

resistance using Vs- The guidelines are based

on field performance data from 17 earthquakes

and in situ Vs measurements at over 40 different

sites (112 test arrays), resulting in a total of 177

liquefaction and non-liquefaction case histories.

The 17 earthquakes are listed as follows.

1 Research Civil Engineer, Structures Division,

Building and Fire Research Laboratory, National

Institute of Standards and Technology,

Gaithersburg, Maryland, 20899.

2Professor, Department of Civil Engineering,

University of Texas, Austin, Texas, 78712.

77
Earthquake Magnitude

1906 San Francisco, California 7.7

1964 Niigata, Japan 7.5

1975 Haicheng, PRC 7.3

1979 Imperial Valley, California 6.5

1981 Westmorland, California 5.9

1983 Borah Peak, Idaho 6.9

1 986 Event LSST4, Taiwan 6.5

1 986 Event LSST7, Taiwan 6.5

1 986 Event LSST8, Taiwan 6.2

1986 Event LSST12, Taiwan 6.2

1986 Event LSST13, Taiwan 6.2

1986 Event LSST16, Taiwan 6.2

1 987 Elmore Ranch, California 5.9

1987 Superstition Hills, California 6.5

1 989 Loma Prieta, California 7.0

1993 Hokkaido-nansei-oki, Japan 7.8

1995 Hyogo-ken Nanbu, Japan 6.9

2. ESTIMATING LIQUEFACTION

RESISTANCE

The most widely used procedure for evaluating

the liquefaction resistance of granular soils is the

stress approach by Seed and his colleagues

(1971, 1982, and 1985) based on modified

Standard Penetration Test (SPT) blow count.

Following the general format of this approach,

Robertson et al. (1992) developed a procedure

based on modified Vs.

2.1 Cyclic Stress Ratio

In the stress approach, liquefaction resistance is related to the ratio of cyclic shear stress to initial

vertical effective stress, called cyclic stress ratio.

The cyclic stress ratio, Wo'v, at a particular

depth in a level soil deposit can be expressed as (Seed and Idriss 1971):

T av/o' v = 0.65 (amax/g) (ov/c'v)

r

d (1)

where Tav is average cyclic shear stress caused

by the earthquake, a'v is initial effective vertical

(overburden) stress, ov is total overburden

stress, a max is peak horizontal ground surface

acceleration, g is acceleration of gravity, and rd

is a shear stress reduction factor. The factor rd can be estimated by:

r d = 1 - 0.015z (2)

where z is depth in meters.

2.2 Modified Shear Wave Velocity

Robertson et al. (1992) modified Vs by:

Vsi =Vs (P a/a, v ) 0- 25(3)

where Pa

is a reference stress, typically 100 kPa,

and CJ'y is in kPa. They chose to modify in terms

of tf v to follow the traditional way SPT and

Cone Penetration Test (CPT) data are modified.

The exponent of 0.25 was selected based on

laboratory studies (Hardin and Drnevich 1972).

Equation 3 assumes that the coefficient of earth

pressure at rest equals 1

.

2.3 Development of V s-based Charts

Field performance data for magnitude 7

earthquakes are shown in Figure 1. The

occurrence of liquefaction is based on the

appearance of sand boils, ground cracks and

fissures, or ground settlement. For each case

history, the shear wave velocity shown is the

average of values reported for the most

vulnerable layer, and modified using Equation 3. Since most attenuation relationships are for the

randomly oriented horizontal component of

ground motion rather than the larger of two

horizontal components, the values of amax used

to estimate the cyclic stress ratios are based on

the randomly oriented horizontal component that

would have occurred at the site in the absence of

liquefaction.

The preliminary liquefaction potential boundary

shown in Figure 1 is drawn to pass through the

origin, include nearly all liquefaction case

histories, and become vertical at higher values of

Vsi . A boundary passing through the origin is supported by the following relationship between

cyclic stress ratio and Vsi f°r a l me of constant

average cyclic shear strain, Yav> (Dobry 1996):

Xav/O'v = f(Yav)V S l 2

(4)

A vertical boundary at higher values of V si is justified since dense granular soils exhibit

dilative behavior even at large strains. Based on

engineering judgment, a value of about 280 m/s

is assumed as an upper bound for most

contractive soil types. This potential boundary

is similar to the boundaries proposed by Kayen

et al. (1992) and Lodge (1994) using field

performance data primarily from the 1989 Loma

Prieta, California earthquake.

For earthquakes with magnitude between 5.5

and 8.5, liquefaction potential boundaries can be

formed using all available field performance

data, as shown in Figure 2. The preliminary

magnitude scaling factors used to construct these

boundaries are listed in column 6 of Table 1. These scaling factors compare well with SPTbased factors developed in recent years by other

investigators, also listed in Table 1. The

78
boundaries for magnitude 6, 6.5 and 7.5

earthquakes shown in figure 2 correctly predict

liquefaction at all sites where surface

manifestations of liquefaction were observed.

The few liquefaction cases that lie slightly below

the liquefaction boundary for magnitude 7

earthquakes are sites where the ground is sloping

and/or where liquefaction may have been

marginal.

It is interesting to note that similar liquefaction

potential boundaries can be constructed using

Vs and amax directly, as shown in figure 3. The

number of non-liquefaction case histories

incorrectly classified in figure 3 is more or less

the same as the number of non-liquefaction case

histories incorrectly classified in figure 2.

Figure 3 suggests that liquefaction will never occur in any earthquake if the shear wave

velocity in the upper 12 m of soil exceeds about

240 m/s. A similar but preliminary conclusion

was reached by Seed et al. (1983). They

concluded that liquefaction will never occur if the shear wave velocity in the upper 1 5 m of soil

exceeds about 366 m/s. The boundaries shown

in figure 3 provide a simpler assessment

procedure than the stress-based approach.

However, their application should be limited to

initial screening of sites with characteristics

similar to the database (depth of most vulnerable

layer less than 12 m, and depth of water table

0.5-7.6 m).

3. CONCLUSIONS

The case histories presented represent a wide

range of soil types, ranging from sandy gravel

with cobbles to profiles including silty clay

layers. Thus, for current use, the preliminary

liquefaction potential boundaries presented in

figures 1 and 2 are recommended for all

liquefiable soil types. Non-liquefiable soil types

include fine-grained soils with clay contents

greater than 15%, liquid limits greater than 35%,

or moisture contents less than 90% of the liquid

limit (Seed and Idriss 1982). The potential

boundaries presented in figure 3 are suggested

for initial screening of sites only where

conditions are similar to the database, which can be viewed from the figure. Additional welldocumented case histories of all types of soil

that have and have not liquefied during

earthquakes should be compiled, particularly

denser deposits shaken by stronger ground

motions.

4. ACKNOWLEDGMENTS

The writers gratefully acknowledge the valuable

review of this work by the Workshop (1996).

We also thank Ricardo Dobry, Rensselaer

Polytechnic Institute, and Riley Chung, National

Institute of Standards and Technology, for their

comments, which greatly enhanced the quality

of this paper.

5. REFERENCES

1. Ambraseys, N. N. (1988). "Engineering

Seismology," Earthquake Engrg. and

Structural Dynamics . Vol. 17, p. 1-105.

2. Andrus, R. D. (1994). "In Situ

Characterization of Gravelly Soils That

Liquefied in the 1983 Borah Peak

Earthquake," Ph.D. Dissertation. Univ.

of Texas at Austin, 533 p.

3. Andrus, R. D., and Stokoe, K. H., II (1996). "Guidelines for Evaluation of

Liquefaction Resistance using Shear

Wave Velocity," Proc. NCEER

Workshop on Evaluation of

Liquefaction Resistance , held in Salt

Lake City, UT, January 4-5

.

4. Arango, I. (1996). "Magnitude Scaling

Factors for Soil Liquefaction

Evaluation," Paper presented at the

NCEER Workshop on Evaluation of

Liquefaction Resistance , held in Salt

Lake City, UT, January 4-5.

5. Dobry, R. (1996). Personal

communication.

6. Dobry, R., Stokoe, K. H., II, Ladd, R.

S., and Youd, T. L. (1981).

"Liquefaction Susceptibility from SWave Velocity," Proc. In Situ Testing

to Evaluate Liquefaction Susceptibility . ASCE National Convention, held in St.

Louis, MO.

7. Hardin, B. O. and Drnevich, V. P.

(1972). "Shear Modulus and Damping

of Soils, Measurement and Parameter

Effects," J. of the Soil Mechanics and

Foundation Div. . ASCE, New York,

NY, Vol. 98, SM7, pp. 667-692.

8. Kayen, R. E., Mitchell, J. K., Seed, R.

B., Lodge, A., Nishio, S., and Coutinho,

R. (1992). "Evaluation of SPT-, CPT-,

and Shear Wave-Based Methods for

Liquefaction Potential Assessment

Using Loma Prieta Data," Proc. 4th

79
Japan-U.S. Workshop on Earthquake

Resistant Design of Lifeline Facilities

and Countermeasures for Soil

Liquefaction, Technical Report

NCEERZ92Z0019, held in Honolulu, HI,

M. Hamada and T. D. O'Rourke, Eds.,

Nat. Ctr. for Earthquake Engrg. Res.,

Buffalo, NY, Vol. 1, pp. 177-204.

9. Lodge, A. L. (1994). "Shear Wave

Velocity Measurements for Subsurface

Characterization," Ph.D. Dissertation . Univ. of California at Berkeley.

10. Robertson, P. K., Woeller, D. J., and

Finn, W. D. L. (1992). "Seismic Cone

Penetration Test for Evaluating

Liquefaction Potential Under Cyclic

Loading," Can. Geotech. J. , Vol. 29, pp.

686-695.

11. Seed, H. B., and Idriss, I. M. (1971).

"Simplified Procedure for Evaluating

Soil Liquefaction Potential," J. of Soil

Mechanics and Foundation Div. . ASCE,

New York, NY, Vol. 97, SM9, pp.

1249-1273.

12. Seed, H. B., and Idriss, I. M. (1982).

Ground Motions and Soil Liquefaction

During Earthquakes. Monograph , EERI,

Oakland, CA, 134 p.

13. Seed, H. B., Idriss, I. M., Arango, I. (1983). "Evaluation of Liquefaction

Potential Using Field Performance

Data," J. of Geotech. Engrg. . ASCE,

Vol. 109, No. 3, pp. 458-482.

14. Seed, H. B., Tokimatsu, K., Harder, L.

F., and Chung, R. M. (1985).

"Influence of SPT Procedures in Soil

Liquefaction Resistance," J. of Geotech.

Engrg. . ASCE, Vol. Ill, No. 12, pp.

1425-1445.

15. Stokoe, K. H., II, Roesset, J. M.,

Bierschwale, J. G., and Aouad, M.

(1988). "Liquefaction Potential of

Sands from Shear Wave Velocity,"

Proc. 9th World Conf. on Earthquake

Engrg,., held in Tokyo, Japan, Vol. Ill,

pp. 213-218.

16. Tokimatsu, K., Kuwayama, S., and

Tamura, S. (1991). "Liquefaction

Potential Evaluation Based on Rayleigh

Wave Investigation and Its Comparison

with Field Behavior," Proc. 2nd Int. Conf. on Recent Advances in Geotech.

Earthquake Engrg. and Soil Dynamics , held in St. Louis, MO, S. Prakash, Ed.,

Univ. of Missouri-Rolla, Vol. I, pp.

357-364.

17. Workshop (1996). NCEER Workshop

on Evaluation of Liquefaction

Resistance, held in Salt Lake City, UT,

January 4-5.

18. Youd, T. L. (1996). "Magnitude

Scaling Factors," Proc. NCEER

Workshop on Evaluation of

Liquefaction Resistance , held in Salt

Lake City, UT, January 4-5.

Table 1. Magnitude Scaling Factors Obtained by Various Investigators (modified after Youd 1996)

Moment Magnitude Scaling Factor

Magnitude,

Mw

(1)

Seed and

Idriss (1982)

(2)

Ambraseys

(1988)

(3)

Youd

(1996)

(4)

Arango

(1996)

(5)

This Paper

(6)

5.5 1.43 2.86 3.26 2.25 - 3.00 2.75

6.0 1.32 2.20 2.15 1.75-2.00 2.15

6.5 1.19 1.69 1 1.50 1.50-1.57 1.65

7.0 1.08 1.30 1.20 1.25 1.25

7.5 1.00 1.00 1.00 1.00 1.00

8.0 0.94 0.67 0.90 0.75 0.75

8.5 0.89 0.44 0.77 0.62 0.60

80
0.8

0.6

>1t-

(0 GC

COw

05

o

IO

0.4

0.2

0.0

Data Based on:

Average value of Vsi

amax for the randomly oriented

horizontal component

Solid - liquefaction

Open - no liquefaction

~r 1 Measurement Type:

Crosshole

o • Downhole, SCPT

A a SASW

100 200 300

Modified Shear Wave Velocity, Vsi , m/s

400

Figure 1. Preliminary Liquefaction Potential Boundary Based on Modified Shear Wave Velocity and

Cyclic Stress Ratio with Results from Magnitude 6.9 to 7.0 Earthquakes.

81
0.8

0.6

0.4

o

>> O 0.2

0.0

Data Based on: Mw = 5.9 to 6.2 earthquakes

Average value of Vsi

amax f°r the randomly oriented

horizontal component

Solid - liquefaction

Open - no liquefaction

Liquefaction

Mw

= 6

No

Liquefaction

0 100 200 300 400

Modified Shear Wave Velocity, V§1 , m/s

(a) Mw

= 6

0 100 200 300 400

Modified Shear Wave Velocity, Vsi, m/s

(b) Mw

= 6.5

0.8

0.6

0.4

0.2

0.0

1 1 1TM T T

Data Based on: Mw = 6.9 to 7.0 earthquakes

Average value of Vgi amax for the randomly oriented

horizontal component

Solid - liquefaction

Open - no liquefaction

Liquefaction

Mw = 7

No

Liquefaction

'Soil may be weakly

cemented with carbonate

0.8

0 100 200 300 400

Modified Shear Wave Velocity, Vsi, m/s

(c) Mw

= 7

> 0.6

03 CC

(/> w

4—

»

CO

_o"o

>. O

0.4

0.2

0.0

Data Based on: Mw = 7.3 to 7.8 earthquakes

"Average value of V§1

amax 'or A16 randomly oriented

horizontal component

Solid - liquefaction

Mw = 7.5

No

Liquefaction

0 100 200 300 400

Modified Shear Wave Velocity, Vsi, m/s

(d) Mw

= 7.5

Figure 2. Preliminary Liquefaction Potential Boundaries Based on Modified Shear Wave Velocity and

Cyclic Stress Ratio with Results from 17 Earthquakes.

82
0.8

0.6

s

°- 4 e

CO

0.2

0.0

1 1 r T

Data Based on: Mw = 6.9 to 7.0 earthquakes

Average value of Vs

amax for the randomly oriented

horizontal component

•Depth<7m

Solid - liquefaction

Open - no liquefaction

Liquefaction

Mw = 6

No

Liquefaction

100 200 300 400

Shear Wave Velocity, Vs, m/s

(a) Mw

= 6

0.8 i

1 1 1 1 1

Data Based on: Mw = 6.9 to 7.0 earthquakes

- Average value of Vg

amax f°r fr> e randomly oriented

horizontal component

0.6 h Depth < 7 m

Solid - liquefaction

Open - no liquefaction

4 DO

0.4

E

CD

0.2

0.0

Liquefaction

Mw = 6.5

No

Liquefaction

100 200 300 400

Shear Wave Velocity, Vs, m/s

(b) Mw

= 6.5

0.8

0.6

i

°- 4

£

CO

0.2

0.0

1

1

1

1

"

Data Based on: Mw = 6.9 to 7.0 earthquakes

Average value of Vs

amax f°r the randomly oriented

horizontal component

"Depth < 12 m

Solid - liquefaction

Open - no liquefaction

Liquefaction

Mw = 7

No

Liquefaction

'Soil may be weakly

cemented with carbonate

0.8

100 200 300 400

Shear Wave Velocity, Vs, m/s

(c) Mw

= 7

0.6

£ 0.4

E

CO

0.2

0.0

1

1 1 1 1

Data Based on: Mw = 7.3 to 7.8 earthquakes

•Average value of Vg

amax f° r the randomly oriented

horizontal component

-Depth < 1 2 m

Solid - liquefaction

Open - no liquefaction

I Liquefaction

Mw = 7.5

No

Liquefaction

100 200 300 400

Shear Wave Velocity, Vs, m/s

(d) Mw

= 7.5

Figure 3. Preliminary Liquefaction Potential Boundaries Based on Shear Wave Velocity and Peak

Horizontal Ground Surface Acceleration with Results from 17 Earthquakes.

83

damage Investigation of

Steel buildings in specific Areas Observed from

the 1995 hyogoken-nanbu earthquake

By

MTOORIKAWA Mitsumasa 1 ), HASEGAWA Takashi2), MUKAI Akiyoshi3), NISHIYAMA Isao4), FUKUTA Toshibumi5

), and YAMANOUCHI Hiroyuki6)

ABSTRACT

An investigation team investigated the damage to

steel building structures that suffered from the

1995 Hyogoken-nanbu earthquake. The team

investigated the damage to a total number of 650

steel buildings, located in the three specific areas of the city of Kobe where many structures

suffered significant damage, in order to

determine the damage ratio of each damage level

for all steel buildings in these areas. The

damage to all steel buildings taller than or equal

to three stories was investigated in the three

areas: the parts of Higashinada-ku, Chuo-ku,

and Hyogo-ku. The investigation was made

mainly from the exterior and also from the inside

as much as possible. The scope of the

investigation was to record the extent and types

of damage to steel buildings. The damage

levels were classified into four levels, that is,

No damage, Minor damage, Moderate damage,

and Severe damage including Collapse.

Among the investigated buildings, 112 were

rated as Severe/Collapse, 93 as Moderate, 212

as Minor, and 220 as No damages. The results

showed that the ratio of the number of the

buildings of Severe/Collapse damage which

were designed and constructed according to the

current seismic provisions of the Building

Standard Law of Japan (BSU) is about one

third of that according to the former ones. KEY WORDS: Hyogoken-nanbu earthquake,

damage survey, steel buildings, specific areas,

damage levels, previous and current seismic

codes.

INTRODUCTION

In the early morning of January 17, 1995, an

earthquake, which measured Mjma 7.2 on the

JMA (Japan Meteorological Agency) magnitude

(moment magnitude Mw 6.9), shook the Kobe

and Awaji areas. The Hyogoken-nanbu

earthquake produced some of the highest

accelerations and velocities ever recorded in

Japan, both on the ground and in structures.

The highest JMA Seismic Intensity assigned

was VII, which is the severest intensity, at several locations spreading over the east-to-west

belt in the affected region. Significant damage

was observed in some parts of the following

areas: Suma-ku, Nagata-ku, Hyogo-ku, Chuoku, Nada-ku, Higashinada-ku, Ashiya City,

Nishinomiya City, Takarazuka City, and Awajishima Island. Damage to building structures

varied widely but included numerous cases of

partial and complete collapse. This event has

been the costliest natural disaster in the history

with an estimated 10 trillion yen in property

damage. The official death toll was 6,308

which included 789 death related to the indirect

effects of the earthquake, and there were 43, 177

injured. Damaged houses including

complete/partial destroy and slight damage were

436,416, and damaged buildings were 4,702.

These results mentioned above were investigated

by the Fire Defense Agency and summed up

upon December 27, 1995.

1) Head, Structure Div., Building Research

Institute, Ministry of Construction, 1 Tachihara,

Tsukuba, Ibaraki, 305 Japan.

2) Research Engineer, Structural Dynamics Div.,

BRI, MOC.

3) Senior Research Engineer, Structure Div.,

BRI, MOC.

4) Coordinator for International Research

Cooperation, BRI, MOC.

5) Head, Large Scale Structure Testing Div.,

BRI, MOC.

6) Director, Structural Engineering Dept., BRI,

MOC.

85
Shortly after the Hyogoken-nanbu earthquake,

the Building Research Institute established an

investigation project team of research engineers

to conduct a post-earthquake investigation.

The project team covered the following subjects:

recorded ground and structure motions,

structural engineering including reinforced

concrete buildings, steel buildings, and wooden

houses, geotechnical engineering, emergency

risk assessment, fire-related aspects, and

nonstructural damage. The results of the

activities of the project team have been reported

inRef. 1.

After a month from the Hyogoken-nanbu

earthquake, investigation teams of research

engineers and practitioners were established to

conduct a post-earthquake damage investigation

on buildings that suffered from the earthquake.

One of the investigation teams investigated the

damage to steel building structures. The

members of the investigation team of steel

buildings were from the Building Research

Institute, the Kozai Club, and the Lath Sheet

Kogyo-kai. This team worked in four days of

February, 1995, and investigated the damage to a total number of 655 buildings considered to be

steel structures, located in the three specific areas of the city of Kobe as shown in Fig. 1, in

order to determine the damage ratio of each

damage level for all steel buildings in these

areas. DAMAGE SURVEY

The damage to all steel buildings taller than or equal to three stories was surveyed in the

following three areas of the city of Kobe as shown in Fig. 1: the parts of Higashinada-ku,

Chuo-ku, and Hyogo-ku, where many

structures suffered significant damage, from

February 20 to 23, 1995. In Higashinada-ku,

the surveyed area of 7,589,000m2 included three

railways of Hankyu, JR and Hanshin. The

surveyed area of 989,000m2 in Chuo-ku

covered from the north side of Sannomiya

Railway Station to the Kobe Marine

Meteorological Observatory. In Hyogo-ku,

surveyed was the area of 1,296,000m2 where

Municipal Subway, Kobe Express Railway and

JR are passing through. The total area surveyed was 9,874,000m2

. Old steel

buildings built with light-gauged beams and

columns were not investigated, although many

of them were significantly damaged. The

investigation was made mainly from the exterior

of buildings, and also from the inside as much

as possible. The scope of the investigation

was to record the extent and types of damage to

steel buildings. For each building surveyed,

the investigators filled out the standard data

forms and took some photographs showing

notable damage. The damage levels were

classified into four levels, that is, No damage,

Minor damage, Moderate damage, and Severe

damage including Collapse.

The following general guideline was adopted to

judge the damage level.

Severe/Collapse Damage: Collapse of the

entire building or a story of the building.

Serious damage to structural members such as beams, columns, and connections; and

significant residual lateral deformations to an

extent considered to be beyond repair.

Moderate Damage: Buckling and/or rupture of

bracing members; and plastification of beams

and columns. Small residual lateral

deformations to an extent considered to be

repairable. Any damage not assigned as Minor

or Severe/Collapse.

Minor Damage: No damage to major structural

members such as beams and columns. Minor

cracking and spalling of exterior finishes.

After completing the field investigation, the

structural type of all steel buildings surveyed

was reexamined and the year of construction of

them was identified from the official records.

Some buildings had been demolished or under

demolition during the field investigation. They

were not included in this damage investigation.

RESULTS

A total number of 655 buildings considered to

be steel structures were investigated in the field

investigation. As a result of the reexamination,

it was found out that five of 655 buildings were

not steel structures. Therefore, the investigated

results for 650 steel buildings are presented in

this report.

Table 1 shows the number of the buildings with

respect to the number of stories, though it unexpectedly includes the one and two story

buildings. The number of the buildings with

respect to the number of stories slightly varies in

86
each area. The reason is that the ratio of

residential area is large in the parts of

Higashinada-ku and Hyogo-ku, while that is small in the part of Chuo-ku. In total, almost a

half of the investigated buildings are three

stories tall and about 85% of them are lower

than or equal to five stories.

Table 2 shows the number of the buildings with

respect to occupancy, and the number in

parentheses indicates the ratio of them. In the

part of Chuo-ku, the number of the investigated

buildings with respect to occupancy is the

largest for shops followed by residential houses

and offices. On the contrary, in the parts of

Higashinada-ku and Hyogo-ku, that is the

largest for residential houses followed by shops

and offices. These results are also related to

the ratio of residential area. In total, the ratios

of the buildings with respect to occupancy are 37% for shops, 36% for residential houses, and

16% for offices.

Table 3 shows the number of the buildings with

respect to damage level, and the number in

parentheses indicates the ratio of them. Among

650 steel buildings, 112 were rated as Severe/Collapse, 93 as Moderate, 212 as Minor,

and 220 as No damages. On comparing the

ratio of the buildings of Severe/Collapse damage

among three areas, the largest is 25.0% in the

part of Higashinada-ku followed by 19.3% in

the part of Chuo-ku and 11.5% in the part of

Hyogo-ku. In total, the ratio of the buildings

of Severe/Collapse damage is 17.2%; that of

Moderate damage 14.3%; that of Minor damage

32.6%; and that of No damage 33.8%.

Figure 2 shows the comparison of the number

and ratio of the buildings at each damage level

between the buildings designed and constructed

according to the previous seismic provisions of

BSLJ (before 1981) and the current ones (after

1982). The current seismic code has been

revised in 1981. The number of the buildings

according to the previous code is 315, and that

according to the current code is 241. The

ratios of the buildings, designed according to the

previous and current codes, of Severe/Collapse

damage are 25% and 9% respectively, while

those of No damage is 20% and 51%

respectively. It is noted that the ratio of the

buildings, designed according to the current

code, of Severe/Collapse damage is much

smaller than and the ratio of No damage is much

larger than the ratio of the buildings designed

according to the previous code.

Table 4 shows the number of the buildings

damaged to each structural member and element,

and the number in parentheses indicates the ratio

of them. The number and ratio of the buildings

at each damage level are compared between the

buildings designed according to the previous

and current codes. The ratios of the buildings,

damaged to columns, designed according to the

previous and current codes are 10.2% and

4.6%; those to beams 3.2% and 3.3%; those to

braces 6.7% and 1.7%; those to column bases

16.5% and 5.0%; those to bolted connections

2.9% and 0.4%; and those to welded

connections 3.2% and 4.1%, respectively.

The ratios of the buildings, damaged to columns,

braces, column bases, and bolted connections,

designed according to the current code are smaller than those designed according to the

previous code. It is remarkable that the ratios

of the buildings, damaged to beams and welded

connections, designed according to the current

code are almost the same as or even larger than

those designed according to the previous code.

Figure 3 shows the number and ratio of the

buildings at each damage level by the number of

stories. The ratio of the buildings of No or Minor damages are more than 60%. The ratio

of the buildings of Severe/Collapse damage is

19.4% for three stories, 22.4% for four stories,

10.0% for five stories, and less than 10% for

the buildings taller than five stories. It is

pointed out that the ratio of the three and four

story buildings of Severe/Collapse damage is

about 20% and much higher than that of the

other buildings. Among the buildings

designed according to the previous code, the

ratio of the buildings of Severe/Collapse damage

is 24.5% for three stories, 30.8% for four

stories, and 21.2% for five stories. Among the

buildings designed according to the current code,

the ratio of the buildings of Severe/Collapse

damage is 11.8% for three stories, 10.5% for

four stories, and 3.0% for five stories.

SUMMARY AND CONCLUSIONS

The damage investigation to steel building

structures that suffered from the 1995

Hyogoken-nanbu earthquake was conducted in

87
order to determine the damage ratio of each

damage level for all steel buildings in specific

three areas of the city of Kobe, where many

structures suffered significant damage: the parts

of Higashinada-ku, Chuo-ku, and Hyogo-ku.

A total number of 650 steel buildings, taller than or equal to three stories, were investigated in the

field investigation.

The conclusions from the presented

investigation can be summarized as follows:

(1)Approximately a half of the investigated

buildings are three stories tall and about 85% of

them are lower than or equal to five stories.

(2)The ratios of the buildings with respect to

occupancy are 37% for shops, 36% for

residential houses, and 16% for offices.

(3)Among the investigated buildings, 112 were

rated as Severe/Collapse, 93 as Moderate, 212

as Minor, and 220 as No damages. On

comparing the ratio of the buildings of

Severe/Collapse damage among three areas, the

largest is 25% in the part of Higashinada-ku

followed by 19% in the part of Chuo-ku and

12% in the part of Hyogo-ku. In total, the

ratio of the buildings of Severe/Collapse damage

is 17%; that of Moderate damage 14%; that of

Minor damage 33%; and that of No damage

34%.

(4)The ratios of the buildings, designed

according to the previous and current codes, of

Severe/Collapse damage are 25% and 9%

respectively, while those of No damage is 20%

and 51% respectively. It is noted that the ratio

of the buildings, designed according to the

current code, of Severe/Collapse damage is much smaller than and the ratio of No damage is

much larger than the ratio of the buildings

designed according to the previous code.

(5)The ratios of the buildings, damaged to

beams and welded connections, designed

according to the current code are almost the

same as or even larger than those designed

according to the previous code. On the

contrary, the ratios of the buildings, damaged to

columns, braces, column bases, and bolted

connections, designed according to the current

code are smaller than those designed according

to the previous code.

(6)The ratio of the buildings of Severe/Collapse

damage is 19% for three stories, 22% for four

stories, 10% for five stories, and less than 10%

for the buildings taller than five stories. It is pointed out that the ratio of the three and four

story buildings of Severe/Collapse damage is

much higher man that of the other buildings.

Acknowledgments

The writers gratefully wish to express their

gratitude to the members of the damage

investigation team of steel buildings from the

Kozai Club and the Lath Sheet Kogyo-kai for

their cooperation in the field investigation. The

writers also gratefully acknowledge the support

of the municipal authorities of the city of Kobe.

Reference

1) Building Research Institute, "A Survey

Report for Building Damages due to the 1995

Hyogoken-nanbu Earthquake," March, 1996.

88
Fig. 1 Surveyed Area

Table 1 Number of Buildings with Respect to the Number of Stories

Story Higashinada-ku Chuo-ku Hyogo-ku Total

1 2(1.2%) 0 (0.0%) 0 (0.0%) 2 (0.3%)

2 5 (3.0%) 0 (0.0%) 2 (0.7%) 7(1.1%)

3 88 (52.4%) 57 (30.5%) 154 (52.2%) 299 (46.0%)

4 45 (26.8%) 48 (25.7%) 73 (24.7%) 166 (25.5%)

5 13 (7.7%) 30 (16.0%) 37 (12.5%) 80(12.3%)

6 3(1.8%) 10 (5.3%) 9(3.1%) 22 (3.4%)

7 1 (0.6%) 9 (4.8%) 7 (2.4%) 17 (2.6%)

8 0 (0.0%) 10 (5.3%) 3(1.0%) 13 (2.0%)

9 1 (0.6%) 9 (4.8%) 4(1.4%) 14 (2.2%)

^10 0 (0.0%) 4(2.1%) 5(1.7%) 9(1.4%)

Unidentified 10 (6.0%) 10 (5.3%) 1 (0.3%) 21 (3.2%)

Total 168 (100%) 187(100%) 295 (100%) 650 (100%)

Table 2 Number of Buildings with Respect to Occupancy

Occupancy Higashinada-ku Chuo-ku Hyogo-ku Total

Office 35 (16.0%) 29(12.4%) 73 (16.7%) 137 (15.4%)

Shop 52 (23.7%) 127 (54.3%) 152 (34.8%) 331 (37.2%)

Residence 91 (41.6%) 50(21.4%) 177 (40.5%) 318 (35.7%)

Hospital 5 (2.3%) 3(1.3%) 5(1.1%) 13 (1.5%)

School 0 (0.0%) 1 (0.4%) 0 (0.0%) 1 (0.1%)

Factory 10 (4.6%) 0 (0.0%) 8(1.8%) 18(2.0%)

Warehouse 13 (5.9%) 0 (0.0%) 7(1.6%) 20 (2.2%)

Public office 0 (0.0%) 2 (0.9%) 0 (0.0%) 2 (0.2%)

Unidentified 13 (5.9%) 22 (9.4%) 15 (3.4%) 50 (5.6%)

Total 219 (100%) 234(100%) 437 (100%) 890 (100%)

Note: Some of the buildings have two or more occupancies.

89
Table 3 Number of Buildings at Each Damage Level

lugajllillaUa'^U Hi aa c h i n a f\ a - Vi i riyogo-icu i oiai No Damage 59 ( 35.1 96) 65 ( 34.8 96) 96 ( 32.5 96) 220 ( 33.8 96)

Minor Damage 32 ( 19.1 %) 52 ( 27.8 96) 128 ( 43.4 96) 212 ( 32.6 96) Moderate 29 ( 17.3 96) 28 ( 15.0 %) 36 ( 12.2 96) 93 ( 14.3 96) Severe/Collapse 42 ( 25.0 96) 36 ( 19.3 %) 34 ( 11.5 %) 112 ( 17.2 96)

Unidentified 6 ( 3.6 96) 6 ( 3.2 96) 1 ( 0.3 96) 13 ( 2.0 %)

Total 168 ( 100 96) 187 ( 100 96) 295 ( 100 96) 650 ( 100 96)

Number of Buildings at Each Damage Level

No Minor Severe/

Damage Damage Moderate Collapse Others

Ratio of Buildings at Each Damage Level

Previous Code

315

Current Code

241

Unidentified

94

No

,Damage

Previous Code

Current Code

100 200 300

Unidentified

Number

400

Minor Severe/

Damage Moderate Collapse Others

, __ .

35 •34 I El

20% 40% 60% 80% 100%

Fig. 2 Number and Ratio of Buildings at Each Damage Level

Table 4 Number of Buildings Damaged to Each Structural Member and Element

Member &

Element Column Beam Brace Column Base Bolted Connection Welded Connection

Previous Code

315

32

(32/315=10.2%)

10 (10/315=3.2%)21 (21/315=6.7%)52

(52/315=16.5%)

9 (9/315=2.9%)10 (10/315=3.2%)

Current Code

241

11 (11/241=4.6%)8 (8/241=3.3%) 4

(4/241=1.7%)

12 (12/241=5.0%)1 (1/241=0.4%)

10 (10/241=4.1%)

Unidentified

94

10 (10/94=10.6%)0

(0/94=0%)

7

(7/94=7.4%)

11 (11/94=11.7%)0

(0/94=0%)

0 (0/94=0%)

Note: The number of buildings damaged to columns and beams includes that damaged to bolted and welded connections.

in all areas

iH Ci, fin &h Ci, tt, fa

in io oo cyi o

Story ^

| 50% * 40%

Story

Fig. 3 Number and Ratio of Buildings at Each Damage Level by Story Number

90
AN INTEGRATED PROGRAM TO IMPROVE THE PERFORMANCE

OF STEEL MOMENT RESISTING FRAME BUILDINGS

by

STEPHEN A. MAHIN 1 AND MICHAEL G. MAHONEY2

ABSTRACT

This paper reviews the performance of welded steel

moment resisting frame (WSMF) buildings during the

Northridge earthquake, and examines some of the

studies being undertaken in the U.S. as part of a

project funded by the Federal Emergency Management

Agency (FEMA) to devise improved methods for

designing new steel frame structures and for

inspecting, evaluating and repairing seismic damage to these types of structures.

One of the important lessons of the Northridge

earthquake of January 17, 1994, was the widespread

and unanticipated brittle fractures in welded steel

beam to column connections. No casualties or collapses occurred as a result of these connection

failures, and some welded steel moment frame

buildings were not damaged at all. However, a wide

spectrum of brittle connection damage did occur,

ranging from minor cracking to completely severed

columns. Damage was so severe in some buildings

that all of the moment resisting connections at one or more floors fractured, or significant permanent lateral

displacements occurred. The result has been a loss of

confidence in the procedures used to design and

construct welded connections in steel moment frames,

and a concern these structures may not be sufficiently

safe.

FEMA has initiated a two phase program to develop

reliable and cost-effective methods for the inspection,

evaluation, repair, rehabilitation and construction of

steel frame structures. This program is being managed

and administered by the SAC Joint Venture. This joint

venture consists of three not-for-profit organizations:

the Structural Engineers Association of California

(SEAOC), the Applied Technology Council (ATC)

and the California Universities for Research in

Earthquake Engineering (CUREe).

KEYWORDS

Steel Buildings, Moment-Resisting Frames, Design

Criteria, Inspection, Evaluation, Repair,

Rehabilitation, Experimental Investigations, Dynamic

Analyses

1. INTRODUCTION

One of the important lessons of the Northridge

earthquake of January 17, 1994, was the widespread

and unanticipated brittle fractures in welded steel

beam to column connections. The economy,

versatility and presupposed high plastic deformation

capacity of welded steel moment-resisting frame

(WSMF) buildings has resulted in their common usage

in Los Angeles as well as in other regions of high and

moderate seismicity. No casualties or collapses

occurred as a result of these connection failures, and

some steel moment frame buildings in areas of

moderate shaking were not damaged at all. However,

a wide spectrum of brittle connection damage did

occur, ranging from minor cracking to completely

severed columns. The most commonly observed

damage occurred in or near the welded joint of a girder

bottom flange to the supporting column flange, and

complete brittle fractures occurred in many of these

joints. Damage was so severe in some buildings that

all of the moment resisting connections at one or more

floors fractured, or significant permanent lateral

displacements occurred.

Thus far, almost 200 damaged steel buildings have

been identified. Damage occurred in new as well as old buildings; and in both tall and short structures.

While inadequate workmanship played a role in the

Nishkian Prof, of Structural Engineering, 777 Davis

Hall, University of California, Berkeley, CA 94720

Senior Geophysicist, Federal Emergency

Management Agency, 500 C St. SW, Washington,

DC 20472

91
damage observed in some structures, most damaged

buildings are believed to be constructed consistent

with modern codes and standards of practice.

The effect of these observations has been a loss of

confidence in the procedures used to design and

construct welded connections in steel moment frames,

and a concern that structures incorporating these

connections may not be sufficiently safe. Current

professional judgment is that historic practices used in

the U.S. for the design and construction of WSMF

connections do not provide adequate reliability and

safety. As a consequence, pre-qualified connection

details and design methods contained in the major U.S.

building codes have been rescinded, and emergency

code provisions generally stipulate that new designs

must be substantiated by testing or test-backed

calculations.

To develop effective and economic design procedures

and construction standards, and to restore public and

professional confidence, several fundamental

questions must be answered. These include:

• What happened to WSMF buildings during the

Northridge earthquake?

• What caused the observed damage?

• How to identify WSMF buildings that may

have sustained damage?

• How safe are damaged WSMF buildings and

do they need to be repaired?

• How can damaged buildings be reliably

repaired and/or upgraded?

• How to design and construct new buildings so they will not sustain similar damage in future

earthquakes?

• Can the vulnerability of existing WSMF

buildings to future earthquakes be reliably

determined and mitigated through effective

rehabilitation procedures?

• What are the economic, social and political

impacts of any required changes in design and

construction practices?

Answering these questions involves consideration of

many complex technical, professional and economic

issues including metallurgy, welding, fracture

mechanics, connection behavior, system performance,

and practices related to design, fabrication, erection

and inspection. Unfortunately, current knowledge of

many of these issues is inadequate.

The Federal Emergency Management Agency (FEMA),

as part of its role in the National Earthquake Hazards

Reduction Program (NEHRP), is charged with reducing

j the ever increasing price that earthquakes are costing

this country. Preventing losses before they happen, an

action which we refer to as mitigation, is the only truly

j effective way of reducing this cost. In this light, the role

ofthis nation's building code and standards in mitigating

earthquake losses is critical, and FEMA is committed to supporting this nation's seismic codes and standards to keep them among the best in the world.

Because the damage that resulted from the Northridge

earthquake had called into question the design

assumptions and building code requirements associated

with steel moment frame construction, and because it was not known at that time what repair or retrofitting

techniques would be adequate, and because of the

unknown life safety hazard of the damaged buildings

(including critical facilities), it was crucial that adequate

hazard assessment criteria and repair and/or retrofitting

procedures be identified as soon as possible. This

would then need to be followed by the development of

reliable and cost-effective methods for the inspection,

evaluation, repair, rehabilitation and construction of

steel frame structures.

To resolve this issue in the most effective manner

possible, FEMA elected to following a two phase

approach. Phase 1 was the development of guidance to

provide guidance for the repair and retrofitting of

damaged buildings; Phase 2 is the research and

•development of a final resource document for the

various building codes and standards that address the

design of new construction and the repair and/or

rehabilitation of existing steel moment frame buildings.

Both phases would be part of a coordinated, problemfocused program of research, investigation and

professional development in order to develop and

validate reliable and cost-effective seismic-resistant

design and rehabilitation procedures for WSMF

structures.

Both phases of this program have been contracted to,

and are managed and administered by the SAC Joint

Venture. This joint venture consists of three not-forprofit professional and educational organizations: the

Structural Engineers Association of California

(SEAOC), the Applied Technology Council (ATC)

and the California Universities for Research in

Earthquake Engineering (CUREe). The program

actively involves engineers, researchers, construction

experts and other specialists from throughout the U.S.

92
As of this paper, the Phase 1, which included limited

amounts of laboratory and field testing as well as other

topical investigations, is essentially complete. The

primary product of Phase 1 is the Interim Guidelines:

Evaluation, Repair, Modification and Design of Welded

Steel Moment Frame Structures [1]. This document has

been completed and is now available from FEMA.

Copies of the Interim Guidelines can ordered directly

from FEMA by calling 1-800-480-2520. The document

identification number is FEMA-267.

The goal of the second phase is to identify long-term

solutions and to develop and verify reliable and costeffective seismic design criteria for steel frame

structures.

2. PROJECT DESCRIPTION

The focus of the first phase of the project was on the

development of specific design advisories, guidelines

and other criteria for design, inspection, evaluation,

repair, modification and rehabilitation of WSMF

structures. The Interim Guidelines [1] developed in

Phase 1 were written by a committee of ten experts

from a variety of disciplines. The Guidelines were

subjected to extensive review by engineers,

researchers, building regulators and other public

officials, and representatives from the steel and

construction industries. The scope of the Interim

Guidelines includes welding procedures, quality

assurance, post-earthquake actions, and new

construction. Specific chapters cover: (a) welding and

metallurgy; (b) quality control and assurance; (c)

visual inspection; (d) non-destructive testing; (e)

classification and implications of damage; (f) postearthquake evaluation; (g) post-earthquake inspection;

(h) post-earthquake repair and modification; and (i)

new construction.

The following specific tasks were performed under the

Phase 1 effort to support the development of the

Interim Guidelines and the planning of Phase 2:

1. Inspect and Assess Damaged Buildings;

2. Perform a Detailed Assessment of Select Buildings;

3. Characterize Ground Motions;

4. Develop Design Advisories;

5. Assessment of Current Knowledge;

6. Conduct Laboratory Tests;

7. Develop Draft Interim Guidelines;

8. Conduct a Users Workshop to Review the Guidelines;

9. Finalize and Publish Interim Guidelines, and;

10. Conduct Regional Training Seminars.

The following sections describe some of these studies

as they relate to the inspection, evaluation and repair

of existing WSMF buildings.

2.1 Survey ofNorthridge Building Damages

A systematic effort has been undertaken to assess the

actual performance of steel buildings the Northridge

earthquake. Three levels of survey were conducted.

In the first, a brief questionnaire was sent to more than

two hundred, randomly-selected owners of steel

buildings. This preliminary survey was used to

estimate the overall scope of damage to steel buildings

and to help identify geographic areas where steel

buildings were damaged. Based on this and other

more detailed surveys, the Interim Guidelines

recommended detailed inspection of steel buildings

where ground motions exceeded 0.2g.

A second level of survey was carried out by engineers

on damaged steel frames. Detailed information was

obtained on 89 buildings regarding the types and

locations of damage observed and their structural

configuration, materials and detailing. This second

survey has been supplemented by a detailed survey of

damage in eight buildings selected for dynamic

analyses. Results of these surveys were used to further

identify methods that can be used to select buildings

for inspection, and the joints within a suspect buildings

that should be inspected. For example, it was found

that on average, 70% of the floors of the buildings

surveyed had serious damage to at least one welded

joint. Only 25% of the connections were found with

no damage. About 20% of the building frames

included in the study had more than 40% of their

connections damaged; and in a few instances, all of the

connections at one or more floors were damaged.

Damage in low rise structures appears to be more or

less uniformly distributed over height (with slightly .more severe damage occurring near the bottom of the

structure), whereas tall buildings exhibited greatest

damage in the upper half. Survey results also show

that damage tends to congregate; thus, finding a

severely damaged connection as part of an inspection

process could serve as a trigger for inspection of other

nearby connections.

2.2 Detailed Analyses of Northridge Steel Buildings

Twelve buildings subjected to the Northridge

earthquake were selected for detailed analysis by

consultants using elastic and nonlinear analysis

93
programs. Buildings were selected with heights

ranging from 2 to 1 7 stories, and in locations varying

from Santa Clarita on the north of the epicenter to

Santa Monica on the south. The analyses were

intended to help identify the causes of the damage, as well as the reliability of various analytical methods

and modeling assumptions in predicting the observed

damages. In four of the buildings, recordings of

response during the Northridge earthquake were

available and in three other cases, ambient vibration

tests were performed. For comparison purposes, two

buildings that did not have any apparent damage and

were immediately adjacent to damaged buildings were

included in the study.

The analysis results indicate that the buildings were

very strong when compared with the design forces

incorporated in current U.S. building codes. For many

of the building sites the estimated response spectra

were nearly double those considered in building codes

(assuming elastic response, R = 1). Elastic analysis

results showed that the most heavily damaged

buildings were only stressed 2 to 3 times their

capacities, suggesting that the buildings may be 4 to 8

times stronger than required. The main reason for this

over-strength appears to be the use of large members

to satisfy stringent drift requirements.

Analysis results show poor correlation with observed

damage. The most heavily stressed joints are the most

likely to be damaged; however, the precise location

and severity of damage is not reliably predicted by

analysis. The 60% most highly stressed connections

have roughly equal chance of being damaged.

Similarly, areas of low computed stress were subject to damage. The reasons for these differences include the

effect of initial defects and poor workmanship, and the

limitations of the analytical methods. More detailed

elastic and inelastic analyses shed some light on this

last possibility. First, both types of analyses indicate

that higher mode effects were very important. Thus,

static lateral force methods in the elastic range, and

nonlinear push-over analysis are of limited value for

longer period structures. The predicted distribution of

damage is highly sensitive to small perturbations in the

spectral characteristics of the ground motions

considered.

Another aspect of this difficulty is that traditional

models ignore or only crudely treat deformations in

the beam to column panel zones. More refined

analyses, especially those performed in the inelastic

range, indicated that panel zone deformations were

significant. Review of experimental data and finite

element results suggests that in addition to anticipated

axial deformations, high local bending and shear

distortions are induced in both the beam and column

flanges near the welded joint due to panel zone

yielding and other causes. Composite slab effects

were also found to have an important effect on local

deformations, and overall structural stiffness and

periods. Current methods for modeling these

phenomenon are quite simple and may be inadequate

in many situations.

2.3 Preliminary Test Program

A total of 37 tests of full size beam-to-column

connection subassemblies were done as part of this

preliminary investigation. Twelve test specimens (like

those shown schematically in Fig. 1) were constructed

utilizing pre-Northridge details, half of the specimens

had W36xl50 beams and half had W30x99 beams.

Fourteen-inch wide-flange sections were used as columns in both cases. Dual certified (fy > 50 ksi)

steel was used in all cases. Slabs were not included in

any of the specimens. These specimens exhibited

brittle appearing fractures; some specimens fractured

without any plastic deformation, while a few were able

to deform to a plastic rotation of about 0.02 prior to

fracturing.

These damaged specimens were repaired or upgraded.

Repair consisted of simply rewelding the connections

using high notch toughness FCAW procedures;

backing bars were removed, the root pass of the CJP

weld on the beam flange to column flange connection

was air-arc gouged and repaired with a fillet weld.

This is the prevalent practice in repairs of damaged

buildings in the Los Angeles area. Test results

indicated that the repaired specimens, constructed with

careful quality control, were able to retain their predamage strength and stiffness. Plastic rotation

capacities were not significantly different from those

achieved in the first tests.

Some of the specimens were upgraded in an attempt to

improve their plastic deformation capacity. In the cases investigated, inclined haunches were applied to one or both sides of the beam at its connection to the

column. This detail attempted to move the plastic

hinge in the beam away from the face of the column to the end of the haunch. These tests supplemented

earlier tests by others which utilized trapezoidal and

rectangular shaped cover plates, vertical fins, or side

plates. Test results on the triangular haunches indicate

94
that they are generally able to increase the plastic

deformation capacity of the connection to at least a

plastic rotation of 0.03.

Four beam-column assemblages were removed during

the demolition of a heavily damaged building. These

specimens were tested in their damaged condition as well as following repairs. Four tests have also been

conducted on details believed appropriate for new

construction. Simple weldment specimens were used

to assess effects of various weld procedures, initial

defects, materials, repair methods and loading rates.

2.4 Recommendations for Evaluation, Inspection and

Repair

Based on the results of the studies described above,

additional studies undertaken by others, practical

considerations and expert opinion, a preliminary

methodology was developed as part of this Project for

the inspection, evaluation and repair of WSMF

buildings subject to damaging earthquakes. The

overall process is illustrated in Table 1. This consists

of a preliminary evaluation in which key parameters

are examined to identify the need to conduct a detailed

inspection and evaluation of a particular building. A

building believed to have sustained structural damage

is then subjected to additional analysis and physical

inspection to assess its likely state of damage,

represented by two parameters: Dj and P, where Dj is

the likely damage index at a floor "i" and P which

represent the probability that the damage index for any

floor in the building exceeds 1/3. Recovery actions to

be taken depend on the values of Dj and P, and range

from doing nothing to completely modifying the

structural system. Detailed procedures for repairing or modifying damaged connections are included in the

Interim Guidelines [1].

2.4.1 Selection Of Buildings For Detailed Inspection

The widespread damage to welded and other

connections in WSMF buildings during the Northridge

earthquake supports the need for rigorous postearthquake evaluation of buildings incorporating

vulnerable welded moment-resisting connections. The

Interim Guidelines identify several triggers that would

be appropriate to initiate a detailed evaluation as indicated in Table 1 . The Interim Guidelines indicate

that a structural analysis is not necessary, but may be

performed.

As a class, it was believed that existing undamaged

WSMF buildings appear to have a lower risk of

collapse than many other types of buildings with

known seismic vulnerabilities, the performance of

which is currently implicitly accepted. Consequently,

mandated or emergency programs to upgrade the

performance of these buildings does not appear

necessary to achieve levels of life safety protection

currently tolerated in the U.S. However, the risk of

collapse is definitely greater than previously thought.

Thus, the Interim Guidelines recommended that

individual owners should be made aware of the

increased level of seismic risk and encouraged to

perform modifications to provide more reliable

seismic performance, particularly in buildings housing

many persons, or in critical occupancies.

2.4.2 Evaluation of Damage State, Safety and

Recovery Actions

Once a building is suspected of having sustained

earthquake damage, the damage state and safety of the

building must be assessed. The Interim Guidelines

recommend an eight step procedure, as outlined

below:

1 . Assign connections to groups

2. Determine sample of connections to be initially

inspected in each group

3. Make inspections and find connection damage

factor, dj (0 < dj < 10)

4. Perform additional inspections near badly damaged

connections

5. Estimate average damage index D for connections

in group

6. Determine probable Dmax at any floor and

probability P that one floor has Dmax

> 1/3

7. Determine appropriate actions from Dmax and P

8. Prepare report to owner

The welded steel moment connections in a structure are divided into groups expected to have similar

mechanical characteristics. An initial sampling of

connections from each group is made. Visual and nondestructive methods are used to identify structural

damage in these connections. Three methods have

been recommended in the Interim Guidelines to select

the inspected connections: random sampling; a

deterministic pattern selection process; and an

analytically-based process. Because of the weak

correlation between analytical predictions and

observed damage, it was felt adequate to randomly

95
select the locations of the connections to be inspected.

Typically, 10 to 30% of the connections are inspected

as a minimum, with a larger percentage required when

smaller numbers of connections form the group.

Detection of severe damage triggers the need for

additional inspections of adjacent connections. This

and other requirements (Table 3) can substantially

enlarge the number of inspections.

It is desirable to rationally estimate the effect of any

detected damage has on the stiffness, strength and

deformability of the connections. Given the difficulties

in making this assessment, a simple connection

damage index dj is suggested in the Interim

Guidelines. The connection damage indices vary

between 0 (for no damage) and 10 for total loss of

effective capacity. Various categories of damage are considered, including damage to welds, girders,

columns, shear tabs, and panel zones.

It is also recognized that the structural deformations

causing the damage in the welded beam to column

connections may cause damage in other types of

welded or bolted connections. Thus, the Interim

Guidelines recommend that inspection be extended to

these types of connections under certain

circumstances.

To assess the safety and need for specific recovery

actions, a floor level damage index Dj is computed.

This index ranges from 0 (no damage) to 1 (complete

loss of connection capacity at a floor). First order,

second moment statistical procedures (assuming

normal distributions of damage) are used along with

the dj values determined for the inspected connections

to estimate the maximum likely value of Dj at a floor

and the likelihood (P) that any Dj exceeds 1/3. The

values of P and the maximum D[ value for all floors

(Dmax)are used to determine appropriate recovery

actions. Recommended recovery actions are shown in

Table 2.

When a WSMF experiences damage to a significant

percentage of its moment-resisting connections (on the

order of 25% in any direction of resistance), in

addition to repair, consideration should be given to modifying the configuration of the individual damaged

connections and possibly some or all of the

undamaged connections to provide improved

performance in the future. However, partial

modifications to the structural system should be made

with due consideration of the effect on overall system

behavior. When a WSMF building has had many

seriously damaged connections (on the order of 50%

in any direction of resistance), owners should be

informed that this damage may have highlighted basic

deficiencies in the existing structural system, or a

geologic feature which unusually amplifies site

motion. In such cases the existing system should be

both repaired and modified to provide an acceptably

reliable structural system. Modifications may consist

.either of local reinforcement of individual connections

and/or alteration of the structure's basic lateral-forceresisting system.

2.4.3 Inspection Procedures

The Interim Guidelines contain precise

recommendations for the conduct of the inspections.

These include methods for exposing, visual

inspections and nondestructive testing of connections,

qualifications of inspectors, and a standardized

classification and reporting system. Emphasis is placed on the need for careful visual inspection, as supplemented by the use of magnetic particle testing

and liquid dye penetrant testing. While the use of

ultrasonic testing is encouraged, difficulties exist in

detecting fractures in joints, especially at the beam to

column connections where backing, runout tabs and

other conditions may lead to spurious images

2.4.4 Repair of Earthquake Damages

Currently, the vast majority of connections damaged

by the Northridge earthquake are being simply

repaired to their pre-earthquake condition. As used in

the Interim Guidelines, the term repair means

restoration of the strength, stiffness and deformation

capacity of structural elements that have been

damaged or have construction defects. Modification

means actions taken to enhance the capacity of either

damaged or undamaged elements.

Currently, there is no clear evidence that one welding

procedure can always produce better welds and

structural performance. However, it appears that poor

workmanship and inadequate structural details can

produce poor performance regardless of the welding

procedure. Thus, it is imperative that repair

techniques fully account for proper welding

procedures and incorporate rigorous quality control

and assurance programs. A written Welding

•Procedure Specification is required to provide specific

instruction to the welder and inspector for each weld

condition encountered. In particular, care must be

96
exercised in using proper pre-heat and post-heat,

electrode diameters and deposition rates, and

inspection procedures. Special reference is made in

the Interim Guidelines to the desirability of using

weld filler material with high impact toughness (>20

ft-lbs at 0_ F).

Several examples of repairs are cited in the Interim

Guidelines. In situations where a fracture has occurred

in or near the weld joining the beam and column

flanges, it is recommended that the backing bar be

removed and the root pass be air-arc back gouged and

subsequently repaired with a fillet weld. Any fractures

in the weld and base metal should be air-arc gouged,

ground to achieve proper fit up, rewelded, and

inspected. In some circumstances the fracture may

extend into the column (e.g., a divot of column base

material may be removed by the fracture) and it may

be necessary to rebuild the column section by filling

the divot with weld material, grinding the repair weld

surface smooth, and inspecting the repaired column

flange. In all welding situations, it is recommended

that the engineer should consult with knowledgeable

experts in the fields of welding and inspection.

In some cases it is necessary to replace portions of a member. For instance, it may be necessary to remove

the bottom flange of a beam due to severe damage or

to gain access to make repairs to the column flange. In

such cases, the beam flange may be replaced by a

plate. Care needs to be exercised in orienting the plate

with the rolling direction in the proper direction. In

some cases, the plate can be welded in the same

location as the original flange (welding the plate to the

beam web along the K line as well as to the column

and beam flange). In other cases it may be acceptable

to weld a plate to the surface of the existing flange. In

both cases, care must be taken regarding local details,

welding procedures and avoidance of local or lateral

buckling of the flange and/or web.

It may be desirable to improve behavior of a connection through modification. Currently, most

approaches involve moving the plastic hinge away

from the face of the column where welds join the

beam to the column, and are represented in Figure 2. This may be accomplished by locally strengthening the

beam near the column face, or locally weakening the

beam at a location where the plastic hinge is desired.

Local strengthening can be accomplished by addition

of cover plates, side plates, vertical fins, haunches and

knee braces. Weakening of the beam away from the

face of the column can be achieved by reducing the

area of the flanges by drilling holes in the beam

flanges or by trimming the width of the flanges.

Because behavior of modified details have not worked

satisfactorily in all cases, it is necessary to confirm

designs using tests or test-validated calculations.

Recommendations are made in the Interim Guidelines

regarding the loading and instrumentation program, as well as for situations where new tests are required.

Plastic rotation capacities recommended for modified

details for use in existing buildings are 0.025 radians.

This is lower than recommended values for new

buildings (0.03 radians) since it is not expected that

existing buildings will be designed for the same level

of reliability as new structures.

Design recommendations for modified details are based on realistic estimates of actual material

properties including strain hardening effects.

Similarly, moving the plastic hinge towards the

midspan of a beam will increase the plastic rotation

demands and, when the beam is strengthened at its end, increase the moment at the face of the column . This increase in moment will require careful

evaluation of demands on the panel zone and column,

in order to avoid unwanted plastic deformations in

these locations.

2.5 Phase 2

The second phase of this project is a three year effort to research and develop Design Criteria for Steel Moment

Resisting Frame Construction for use by the various

building codes and standards. This resource document

will address the design of new steel moment frame

buildings and the rehabilitation of existing construction.

Based on a competitive evaluation and selection

process, FEMA found the SAC Joint Venture to be the

most technically qualified firm to perform this work..

Funding for Phase 2 consists of over $6 million from the

President's Northridge Earthquake Emergency Fund as well as $2.5 million of FEMA earthquake program

funds that have been designated from FEMA's Problem

Focused Studies Project.

The SAC Joint Venture has just completed the first

two tasks, selection of a Project Oversight Committee

and development of a Project Work Plan. Upon

FEMA approval of the Project Work Plan, SAC will

begin work on the main tasks, which include assessing our current knowledge of steel building performance,

conducting topical investigations, developing a series

97
of state of the art reports, conducting a comprehensive

testing program, developing an inspection program,

assess the economic, social and political impacts, and

writing the final document, "Design Criteria for Steel

Moment Frame Buildings." Completion of that

document is tentatively scheduled for the end of 1998.

3. CONCLUSIONS

While the Interim Guidelines represent current U.S.

thinking on the proper evaluation, inspection and

repair of existing WSMF buildings as well as for the

design and construction of new ones, there are clearly

many uncertainties and unresolved questions. In Phase 2 of the FEMA/SAC Steel Project additional

research and testing will be conducted to more clearly

define the parameters controlling the performance of

connections and systems, and to develop and verify

procedures for design and rehabilitation of moment

frame structures. Efforts will be made to cooperate

with related activities worldwide.

It is a serious situation when a disaster such as the

Northridge earthquake calls into question an entire

section of the building code. Both FEMA and the SAC

Joint Venture are pleased to be part of a project that

recognized and is addressing an urgent need that arose out of the Northridge disaster, and has brought together

the best of the research and engineering fields into one

national project. This is an effort that has combined the

best of both worlds; the research and testing capabilities

ofthe academic community along with the practical

design experience ofthe engineering community.

With the Interim Guideline that can now be used for

repairing damaged buildings and improving new

construction and with the work that will be done in

Phase 2, steel moment frame buildings will be able to

provide the level of seismic safety that the citizens of

this country have a right to expect.

4. REFERENCES

Federal Emergency Management Agency, Interim

Guidelines: Evaluation, Repair, Modification and

Design of Welded Steel Moment Frame Structures,

FEMA #267, August 1995, Federal Emergency

Management Agency, Washington DC.

5. ACKNOWLEDGMENTS

The information reported herein was prepared under

contract (EMW-95-K-4672) to the Federal Emergency

Management Agency. The authors are indebted to the

efforts and work of numerous individuals and

organizations that have participated in the FEMA/SAC

Steel Project. The findings, conclusions and

recommendations summarized in this paper are not

intended for use in design or evaluation of structures. They do not represent official policy of FEMA or the

SAC Joint Venture.

Pre-Northridge

Detail

Fractures

16 tests

1

H

Repaired or

Upgraded

Detail

1

17 tests

M.

h

New

Construction

Detail

1

4 tests

M.

Figure 1 - Preliminary Test Program

98
Preliminary Evaluation Detailed Evaluation

• Estimate ground

motion at site

• Proximity to fault

rupture

• Preliminary inspection

of damage to

architectural and

exposed structural

elements

• Observe damage in

neighboring buildings

• Analyze building and/or

• Inspect sample of connections in

building

- random selection

- deterministic selection

- analytic selection

• Evaluate damage state

- assign damage intensity factor,

dg, to each connection

- estimate damage index D for

structure

- probability P that damage on at

least one floor exceeds 1/3

Recovery Actions

• Building occupancy

- continued, or

- interrupted

• Repairs

- not required,

- some damaged

connections, or

- all damaged connections

• Modification of connections

or system

Table 1 - Basic Steps in Evaluation Process

Damage State Recommended Recovery Action

All cases Repair all connections found with dj > 5

P>5%orDmax >0.1 Repair all connections found with dj > 2

P>10% orDmax>0.2 Inspect all connections. Repair all

connections found with dj >2

P>25% or Dmax>0.33 Inspect all connections. Unsafe conditions

may exist, careful assessment of building

safety is required. Repair all connections

found with dj >1.

Dmax>0- 5 Inspect all connections. Unsafe

condition exists. Careful assessment of

structure is required. Repair all damage and

modify connections or system for improved

performance.

Table 2 - Recommended Recovery Actions as a Function of P and Dmax

99
\

Cover or

side plates

Single or double

haunches

Trimmed or perforated

beam flanges

Figure 2 - Typical Approaches for Modifying Connections to Move Plastic Hinge Away From Column Face

100
Recent Developments in Strong-Motion Attenuation Relationships

by

William B. Joyner and David M. Boore*

ABSTRACT

New strong-motion relationships incorporating

data recorded since 1988 show smaller spread

than older relationships. The new relationships

for soil sites agree more closely at magnitude

6.5 than at magnitude 7.5, reflecting the larger

number of data for magnitudes near 6.5.

1. INTRODUCTION

In 1988 we published a review of the subject of

strong ground motion in earthquakes (Joyner

and Boore, 1988). The general statements we

made at that time are still applicable, but so many strong-motion data have been recorded

since then that the specific attenuation

relationships described in the review should be

regarded as outdated. In this paper we describe

and compare more recent relationships.

2. NEW RELATIONSHIPS

Here we present a number of widely used

relationships applicable to ground motion from

shallow earthquakes in western North America.

Inclusion or exclusion of a relationship in our comparison signifies nothing regarding

endorsement or nonendorsement of the

relationship.

2.1 Boore et al., (1993, 1994)

Our earlier relationships (Joyner and Boore,

1981, 1982) for horizontal peak acceleration,

velocity, and response spectra were based on

data obtained through 1980. Many more data

have been recorded since then. Three

California earthquakes in particular (1989 Loma

Prieta, 1992 Petrolia, and 1992 Landers) have

provided data for a range of magnitude and

distance, critical for engineering design, which

was poorly represented in our previous work.

Our long-term goals are to develop relationships

incorporating all of the data recorded since our

earlier work and to extend the period range

covered by the relationships, which will require

reprocessing all of the data. Pending the

completion of this work we produced an interim

report (Boore et al., 1993, 1994) using the

earlier data set augmented by the three large,

recent California earthquakes.

The new relationships give horizontal peak

acceleration and response spectra from 0.1 to

2.0 seconds for 2, 5, 10, and 20 percent

damping in terms of moment magnitude,

earthquake distance and site conditions. In an unpublished but widely distributed short note,

we modified the relationships to differentiate

three earthquake mechanism categories: (1)

strike-slip, (2) reverse-slip or oblique-reverseslip, or (3) mechanism unspecified. There are no earthquakes in the data sets with normal or normal-oblique mechanisms. The data sets are

restricted to earthquakes in western North

America with moment magnitude greater than

5.0 and to shallow earthquakes, defined as those for which the fault rupture lies mainly

above a depth of 20 km.

Because the rupture surface for earthquakes

may extend over tens or hundreds of kilometers,

* U.S. Geological Survey, Menlo Park,

California 94025

101
there is ambiguity in defining the source

distance for a strong-motion record. Various

measures of source distance have been used in

the development of attenuation relationships, as

illustrated in Figure 1. We use the closest

distance to the vertical projection of the rupture

surface on the surface of the earth (D5 in

Figure 1). Differently stated, we use the closest

horizontal distance from the station to the area on the earth's surface that lies directly above

the rupture. Figure 2 shows the distribution of

the strong-motion data in terms of distance and

moment magnitude and the role of the three

added earthquakes in filling the gap in the older

data at short distance and large magnitude.

Boore et al. (1993) account for site conditions

by assigning sites to four categories based on

the average shear-wave velocity to a depth of

30 m. In an improvement on that scheme,

Boore et al. (1994) express the effect of site

conditions directly in terms of the average

shear-wave velocity to 30 m. The coefficients

of average shear-wave velocity they obtained

compare very well with those determined

independently by Midorikawa et al. (1994) from

Japanese data as well as with the coefficients

proposed by Borcherdt (1994) for determining

short-period and mid-period amplification

factors in building codes (Figure 3). In order to

compare the relationships of Boore et al. (1994)

with others, values of shear-wave velocity to 30

m are needed for an average rock site and an

average soil site. Figure 4 shows downhole

shear-wave velocity measurements for rock and

soil sites (some at sites where strong-motion

records have been obtained, some not). An

analysis of downhole data (Boore and Joyner,

1996), most of which is shown in Figure 4,

gives values of 620 m/sec for the average rock

site and 310 m/sec for the average soil site.

The standard error of an individual estimate

from our relationships ranges between values

corresponding to factors of 1.5 and 2.0,

depending upon period and damping. These

values are comparable to those for other

relationships.

Our new relationship for peak horizontal

acceleration is compared on Figure 5 with our

earlier relationship (Joyner and Boore, 1982).

On Figure 6 our new acceleration relationship

for strike-slip earthquakes is compared with

data recorded in the 1995 Hyogoken-Nambu

(Kobe) earthquake. Unlike the typical

Japanese earthquakes, which are subduction- zone events, the Hyogoken-Nambu earthquake

was a shallow crustal event, like the typical

California earthquake. For that reason the

general agreement of the recorded data with the curves on Figure 6 should not be surprising.

On Figure 7 our new relationships for

horizontal response spectra is compared with

our earlier relationships (Joyner and Boore,

1982).

There are three important ways in which our

relationships differ from those of most other

authors. The first of the three is the choice of

distance definition, described above. The

second important difference is that our

relationships have the same magnitude scaling

at all source distances. Most other

relationships have smaller magnitude scaling at short distances than at long distances (e.g.

Campbell and Bozorgnia, 1994). We have

examined the question of magnitude scaling at short distances by Monte Carlo simulation and

also by examining the magnitude dependence of

residuals to our relationships at distances less

than 10 km. Both approaches showed no

statistically significant differences in magnitude

scaling at short distances. The third important

difference is that we do the regression analysis

of response spectra at each period

independently. Most other analysts do

regression analysis on spectral ordinates

normalized by peak acceleration and then

multiply the results by the value of peak

acceleration given by regression analysis of

acceleration data.

102
2.2 Abrahamson and Silva (1995, written

communication, 1995)

The Abrahamson and Silva relationships give

horizontal and vertical peak acceleration,

velocity, displacement, and response spectral

values for periods of 0.01 to 10.0 seconds and

damping of 0.5, 1, 2, 5, 7, 10, 15, and 20

percent of critical. The magnitude measure is moment magnitude and the definition of

distance is the closest distance to the rupture

surface (D4 in Figure 1). Site conditions are characterized as rock or deep soil and source mechanisms as strike-slip, reverse-slip, or

reverse-oblique-slip. The relationships take

account of whether a site is on the hanging wall or the footwall of a dipping fault. The

relationships are based on 543 recordings of 98

shallow earthquakes recorded world-wide

between 1935 and 1994 for the magnitude

range of 4.5 to 7.5. Further information can be

found on the web site:

http://www.holonet.net/abrahamson

2.3 Campbell and Bozorgnia (1994)

The Campbell and Bozorgnia relationship gives

peak horizontal acceleration in terms of moment

magnitude, distance, source mechanism, and

site conditions. The definition of distance is the closest distance to the seismogenic rupture,

which is the part of the rupture that lies below

a depth of about 3 km. Sites are assigned to

one of three categories— 1) soft rock, 2) hard

rock, or 3) deep soil— and source mechanisms

to one of two categories- (1) strike-slip and

normal-slip or (2) reverse-slip and reverseoblique-slip. The Campbell and Bozorgnia

relationship is based on 645 recordings of 47

shallow earthquakes recorded world-wide

between 1957 and 1993 within a distance of 60

km.

2.4 Idriss (1993, written communication, 1994)

The Idriss relationships give horizontal peak

acceleration (Idriss, written communication,

1994) and response spectral values (Idriss,

1993) from 0.03 to 5.0 seconds for 5 percent

damping at rock sites in terms of magnitude,

distance, and source mechanism. Moment

magnitude is used in the relationship for peak

acceleration (Idriss, 1994, written

communication). For response spectra local

magnitude MLis used for magnitude

(unspecified) less than or equal to 6, and

surface-wave magnitude Ms is used for

magnitude (unspecified) greater than 6. This

prescription gives a result generally similar to

moment magnitude. The definition of distance

is the closest distance to the rupture surface (D4

in Figure 1). Source mechanisms are assigned

to one of three categories— 1) strike-slip, 2)

reverse-slip, or 3) reverse-oblique-slip.

2.5 Geomatrix (1995; Sadigh et al., 1993)

The Geomatrix relationships give horizontal and

vertical peak acceleration and response spectral

values for 5 percent damping at rock sites in

terms of moment magnitude, distance, and

source mechanism. The horizontal spectral

values are provided at periods between 0.05 and

7.5 seconds and the vertical values for periods

between 0.04 and 3.0 seconds. The equations

are given by Sadigh et al. (1993) with

modifications described by Geomatrix (1995).

The definition of distance is the closest distance

to the rupture surface (D4 in Figure 1). Source

mechanisms are assigned to one of three

categories— 1) strike-slip, 2) reverse-slip, or 3)

reverse-oblique-slip.

3. COMPARISON OF RELATIONSHIPS

On Figure 8 relationships are compared for

peak horizontal acceleration for strike-slip

earthquakes at soil sites for magnitudes 6.5 and

7.5. Figure 8 shows two features that appear in

all of the following comparisons. At all distances the spread among different

relationships is much smaller for magnitude 6.5

than for magnitude 7.5, reflecting the fact that

many more data are available for magnitudes

103
near 6.5 than for magnitudes near 7.5. The

second feature to note is the very different

magnitude scaling at short distance compared to

long distance for the Campbell and Bozorgnia

(1994) relationship, the modestly different

scaling for the Abrahamson and Silva (written

communication, 1995) relationship, and the

same scaling for the Boore et al. (1994)

relationship. The different relationships for 0.3

second horizontal response at soil sites are compared on Figure 9 and for 1.0 second

response on Figure 10. The different

relationships for 1.0 second response at rock

sites are compared on Figure 11. The

maximum spread between different relationships

shown on Figures 7 through 10 is significantly

smaller than shown by Joyner and Boore

(1988), presumably reflecting the influence of

the additional strong-motion data recorded since

1988.

4. REFERENCES

1. Abrahamson, N. A., and Silva, W. (1995).

"A Consistent Set of Ground Motion

Attenuation Relations Including Data from the

1994 Northridge Earthquake." (Abstract)

Seismological Research Letters, v. 66, p.23.

2. Boore, D. M., Joyner, W. B., and Fumal, T.

E. (1993). "Estimation of Response Spectra

and Peak Acceleration from Western North

American Earthquakes: An Interim Report."

U.S. Geological Survey Open-File Report

93-509, 72 p.

3. Boore, D. M., Joyner, W. B., and Fumal, T.

E. (1994). "Estimation of Response Spectra

and Peak Acceleration from Western North

American Earthquakes: An Interim Report, Part

2." U.S. Geological Survey Open-File Report

94- 127, 40 p.

4. Boore, D.M. and Joyner, W.B. (1994).

"Prediction of Ground Motion in North

America." Proc. ATC-35 Seminar on New

Developments in Earthquake Ground Motion

Estimation and Implications for Engineering

Design Practice, Applied Technology Council,

Redwood City, California, ATC 35-1, p. 6-1 -- 6-41.

5. Boore, D. M., and Joyner, W. B. (1996).

"Site Amplification for Generic Rock Sites." In

preparation.

6. Borcherdt, R.D. (1994). "Estimates of SiteDependent Response Spectra for Design

(Methodology and Justification)." Earthquake

Spectra, v. 10, p. 617-653.

7. Campbell, K. W., and Bozorgnia, Y. (1994).

"Near-Source Attenuation of Peak Horizontal

Acceleration from Worldwide Accelerograms

Recorded from 1957 to 1993." Fifth U.S.

National Conference on Earthquake

Engineering, Chicago, Illinois, July 10-- 14,

1994.

8. Geomatrix (1995). "Adjustments to Rock

Response Spectra for Caltrans Toll Bridges in

Northern California." A Report to the Business,

Transportation, and Housing Agency (Caltrans),

Division of Structures, Sacramento, California.

9. Hamada, M., Isoyama, R., and Wakamatsu,

K. (1995). "The 1995 Hyogoken-Nambu (Kobe)

Earthquake - Liquefaction, Ground

Displacement and Soil Conditions in Hanshin

Area." Association for Development of

Earthquake Prediction, Wasada University

School of Science and Engineering, Japan

Engineering Consultants.

10. Idriss, I. M. (1993). "Procedures for

Selecting Earthquake Ground Motions at Rock

Sites." A Report to the National Institute of

Standards and Technology, Department of Civil

and Environmental Engineering, University of

California, Davis, revised 1993.

11. Joyner, W. B., and Boore, D. M. (1981).

"Peak Horizontal Acceleration and Velocity

from Strong-Motion Records Including Records

104
from the 1979 Imperial Valley, California,

Earthquake." Bulletin Seismological Society of

America, v. 71, p. 201 1--2038.

12. Joyner, W. B., and Boore, D. M. (1982).

"Prediction of Earthquake Response Spectra."

U.S. Geological Survey Open-File Report

82-977, 16 p.

13. Joyner, W. B., and Boore, D. M. (1988).

"Measurement, Characterization and Prediction

of Strong Ground Motion." Proceedings of

Conference Earthquake Engineering and Soil

Dynamics II, Geotechnical Engineering

Division, American Society of Civil Engineers,

Park City, Utah, June 27--30, 1988, p. 43-102.

14. Midorikawa, S., Matsuoka, M., amd

Sakugawa, K. (1994). "Site Effects on StrongMotion Records Observed During the 1987

Chiba-Ken-Toho-Oki, Japan Earthquake." Proc.

9th Japan Earthq. Eng. Symp., v. 3, p. E085--

E090.

15. Sadigh, K., Chang, C.-Y., Abrahamson, N.

A., Chiou, S. J., and Power, M. S. (1993).

"Specification of Long-Period Ground Motions:

Updated Attenuation Relationships for Rock

Site Conditions and Adjustment Factors for

Near-Fault Effects." Proceedings of ATC-17-1

Seminar on Seismic Isolation, Passive Energy

Dissipation, and Active Control, San Francisco,

California, March 1 1—12, 1993, Applied

Technology Council, v. 1: Seismic Isolation

Systems, p. 59-70.

105
0>

Cl

CO

Oo

co

cn • coo

<D

L_Eo

CO

CDD

COD

CD

CDOCo

CO

cOo

CDCoN

CO

cn

0)

L_

COI JZ

a>D

CLCOo

0)

Q.O

oD

V-D

CO

o

oo

CD <D o o cD

CDO

CC

D

O

CO CO "6 "o "o

CO "O

CD CD O CO

CO CO

CD

CO

CD CO CO o _o

LJO

OO I I I I I r- cn k) ^" m a o a a o

E8

.a. IEcococi

T33I

60C— OS 1

~" 2 g

.2

£

Q£

g

.1 8

106
CD

D)

CO

8

7-

c

o 6-

E

o

j 1 1 1

1

Peak Acceleration

A

A

A

A

A AAA

AAftA AA/A

• • •

• • ••

• • •*> • •

0

"D3

05

0

E

o

8

7-

6-

_i i 1 1 1 1 1 1 i i i 1 1 1 1

1

Response Spectra

(2) A

A

-i—i i i 1

1

1

10'

A A Aft aa/aa

•• • • •

• • • •

• ••

• old data

a new data

T 11 T I I I I 1 1 11 I I I I I 101 10'

Closest Distance (km)

Figure 2. The distribution of the Western North American data in magnitude and distance space (each

point represents a recording). The data points labeled "old data" were used in previous studies (Joyner

and Boore, 1981, 1982); the "new data" were added in the recent work of Boore et al. (1993). The

points in the top and bottom frames were used in developing equations for peak acceleration and

response spectra, respectively. (From Boore et al., 1993.)

107
J I I I I I I J I I I I I I J L

5 % - damped pseudo-velocity response spectra

-0.2 -

-0.4 -

A -0.6-

-0.8 -

-1 - — BJF94, random comp., unsmoothed

- - - • BJF94, larger comp., unsmoothed •—• Midorikawa etal (1994)

« i Borcherdt (1994)

•1.2 -i—i—i—i i i -i— i

i—i—i i i

0.1 0.2 1

Period (sec)

Figure 3. The coefficient that controls the shear-velocity dependence of response spectral amplification,

as determined for California data (BJF94 = Boore et al, 1994) and Japanese data (Midorikawa et al,

1994). Also shown are the coefficients proposed by Borcherdt (1994) for determining short-period and

mid-period amplification factors in building codes (these were determined from Fourier spectral

amplitude spectra of recordings from the 1989 Loma Prieta earthquake). (From Boore et al, 1994.)

108
40

Avg. Vel. (m/s)

Figure 4. Histograms of shear velocity averaged over the upper 30m, from downhole surveys in

boreholes. Many of the boreholes did not quite reach 30m, and for the sake of estimating representative

velocity distributions, small extrapolations of the travel time to 30m were used in computing the average

velocities. (From Boore and Joyner, 1996.)

109
1 10 20

O)

0.2 -

CO

I 0.1

o

o

<

CO0

Q_

0.02 -

0.01

_i_L

Fault type not specified — Boore etal, 1 994; V30 = 31 0 m/s

Joyner and Boore, 1982; soil

t—i—i—i—

r

10 20

100

-i 1—i—

r

- 0.2

0.1

-0.02

0.01

100

Distance (km)

Figure 5. Peak horizontal acceleration at soil sites for moment magnitude 6.5 and 7.5 as given by Boore

et. al (1994) and Joyner and Boore (1982).

110
10"

10'

10l 101 10'

-I I I I I I 1 J I I -1 *

I I I I I 10(

CO

X

X

0

CDO

O

<

J*

CO

CD

Q_

M = 6.8

Boore etal: strike-slip fault, random component

- - Boore et al, 1994: = 620 m/s - Boore et al, 1994: = 310 m/s

- - - Boore et al, 1 994: = 1 55 m/s

x Kobe, Site I • Kobe, Site II o Kobe, Site III 10' i i I IT

10

-1 10'

1 I I I I I I I 101T 1 1 1—I I I I 10'

- 10"1

10'

Distance (km)

Figure 6. Peak horizontal acceleration for moment magnitude 6.8 at sites where the average shear-wave

velocity to 30 m is 620 m/sec (rock), 310 m/sec (firm soil), and 155 m/sec (soft soil) as given by Boore

et al. (1994) compared with data recorded in the Hyogoken-Nambu (Kobe) earthquake for site class I

(rock), II (diluvium and stiff alluvium), and III (soft alluvium and reclaimed ground) (Hamada et al.,

1995).

111
|j 200

•f 100

o

>

CD

>

as

CD

DC

O

"D3

CD

CO

CL

20 -

10 -

j i i i_ i t i J L

D =

i—i—

r

Boore et al; V30 = 31 0 m/s

Joyner and Boore; soil

0.1 0.2 1 2

- 200

- 100

20

- 10

Period (sec)

Figure 7. Horizontal pseudo-relative-velocity response spectra at soil sites for magnitude 6.5 and 7.5

at zero distance, as given by Boore et al. (1994) and Joyner and Boore (1982).

112
Distance (km)

Figure 8. Peak horizontal acceleration at soil sites for moment magnitude 6.5 and 7.5 as given by Boore

et. al (1994), Abrahamson and Silva (written communication, 1995), and Campbell and Bozorgnia

(1994).

113
1 2 10 20 100

i i i i i i i i_j i i [

T = 0.3 sec

1 1

1 1 1—i—' ' '

I

1

' ' ' 1—''M

1 2 10 20 100

Distance (km)

Figure 9. Horizontal pseudo-absolute-acceleration response at 0.3 sec at soil sites for moment magnitude

6.5 and 7.5 as given by Boore et. al (1994) and Abrahamson and Silva (written

communication, 1995)

114
Distance (km)

Figure 10. Horizontal pseudo-absolute-acceleration response at 1.0 sec at soil sites for moment

magnitude 6.5 and 7.5 as given by Boore et. al (1994) and Abrahamson and Silva (written

communication, 1995)

115
Distance (km)

Figure 11. Horizontal pseudo-absolute-acceleration response at 1.0 sec at rock sites for moment

magnitude 6.5 and 7.5 as given by Boore et. al (1994), Abrahamson and Silva (written communication,

1995), Idriss (written communication, 1994), and Geomatrix (1995, Sadigh et al., 1993).

116
TOPOGRAPHICAL EFFECT IN THE EVALUATION

OF DESIGN GROUND MOTION

by

I. OKAWA", T. KASHIMA", S. KOYAMA0

ABSTRACT

The surface geological condition is, more or less, taken

into account in most seismic structural design

provisions or requirements. In view of the shortcoming

of the effect of topography in the current seismic design

practice, however, a research project was set up for

establishing a methodology of local site effect including

the topographical effect. We investigated the

representative patterns of subsurface topography in urban

area of Japan. The topography was roughly classified

into four categories, slope, hill, sediment-filled valley,

dipping base layer. For slope, the angle and the height

of the slope vary with its location. The general tendency

and dimensions of each topography were provided for

districts. The literature survey was also conducted to obtain the actual behaviors of ground during past

earthquakes clearly justifying the topographical effect on

earthquake-induced ground motions and the subsequent

structural damage. The analytical study was conducted to find the effectiveness of each of the topography

parameters such as angle and heights. A Ricker's

wavelet input is used to make problems as simple as possible. In this paper, research on the amplification

characteristics of surface ground including topographical

effect on seismic motion is shown with some

considerations found in the interim compilation of

analytical results for the final evaluation technique.

KEYWORDS: Ricker's Wavelet, Slope Angle,

Sediment-filled Valley, Damage

1. INTRODUCTION

It is long been pointed out that geological condition has

influence on earthquake ground motion. The

methodology to evaluate it, however, is still under

development. It is partly because the topography or

surface geology has not been studied well from the view

point of dynamic characteristics of ground, and partly

because the property has not been verified through

measurement, that is, earthquake recording, and also

because the analytical investigation needed large

computer power. In addition, there are site specific

situation in the seismic environment of each site, such

as the seismicity and the wave propagation.

The past earthquake damage might be mostly

due to the amplification of irregular soil conditions or

surface topography. With these in mind, we started the

study for incorporating the effect of topography into the

evaluation of design ground motion. As this study is now in its final stage, the evaluation methodology is now being constructed as the final fruit. In this study, we summarized the feature of

topography for each area in Japan, and determine the

representative topographies of urban area in Japan, that

is, Tokyo, Osaka, Nagoya, Yokohama, etc., and

consequently, constructed analytical models for the

topographies, and performed model computation with

several parameters, that is, input condition, the

dimension of topography, slope angle of base

inclination. We made a preliminary methodology for

the evaluation by incorporating the findings from

measurement into the analytical results. We are now

constructing the final version of the evaluation

technique, although it is not included in this paper.

A guideline for design earthquake motion for the

buildings requiring dynamic analysis was proposed by

the Building Research Institute and the Building Center

of Japan. (BRI et al., 1992.) The design response

spectrum specified for horizontal motion hS(T) is to be

computed using the following expression.

HS(T) = C. HB(T). HL(T).HG(Ti)

where

£ : Seismic activity coefficient

HB(T) : Basic response spectrum for horizontal motion

HL(T) : Coefficient of longer period component for

horizontal motion

HG(Tj) : Coefficient of soil amplification for horizontal

motion

The computed spectrum is further modified with

the coefficients of both liquefaction and topographical

effect, when the possibility of liquefaction in sandy soil

is high, and/or when the effect of surface land form can

not be ignored. The expression is thus.

HS'fO = hS(T)-HP(T)-HI(T)

where,

Hl(T) : Coefficient of topographical effect for

horizontal motion

HP(T) : Coefficient of liquefaction for

1) Building Research Institute, Ministry of Construction

Tatehara-1, Tsukuba, Ibaraki 305, Japan

117
horizontal motion

2. SURFACE GEOLOGY AND TOPOGRAPHY IN URBAN AREAS OF JAPAN

The geometrical irregularity in geology is classified into

two types. One is the irregularity of surface landform,

and the other is the irregularity of interfaces of soil

layers. The former includes the slope (cliff), hill, and

valley, etc. The latter includes the sediment-filled

valley, the dipping base layer and horizontal

discontinuity of layers due to fault, etc. The representative soil conditions of (1) Slope,

(2) Hill, (3) Sediment-filled Valley, and (4) Dipping

Base Layer were selected as shown in Fig.l and

examined the distribution of the following parameters

for each of the urban areas in Japan. From the survey,

average values of simplified topography parameters are summarized for some district of Japan.

A location has its specific geological structure. Consequently, the seismic motion thus transmitted

through such strata has site specific properties in ground

motions. We examined the current status of the deep

structure of areas in Japan, such as Kanto plain, Osaka

plain, Noubi plain, Sendai plain, Niigata plain, etc. The deep subsurface structure has not been well

investigated in Japan.

Damage and topography

For the following earthquakes, damage had been

presumably caused by the influence of topography.

During the Tokachi-oki, 1968 earthquake (M=7.8),

major damage occurred to the building located along the

hill edge, especially in the direction orthogonal to the

slope. (Yamahara, 1971) During the Miyagi-ken-oki,

1978 earthquake (M=7.4), major damage occurred to the

wooden houses built on the alluvial lowland and to the

housing site on reclaimed hilly zone. During the Friuli,

Northern Italy, 1976 earthquake (M=5.5), major damage

occurred at the hilly district on the alluvial fan, at cliff edge, and on the steep hill. (AIJ, 1979)

During the Chile earthquake (Ms=7.8), a ridge effect is observed at the hilly zone close to the epicenter. (Celebi,

1987) From the Michoacan, 1985 earthquake (Ms=8.1),

major damage occurred in the transition zone between

rock site and deep sediment-filled zone. During the

Whittier Narrows, 1987 earthquake (ML=5.9), a locally

large MMI was observed. There is an analysis that explains the damage due to the irregular soil condition.

(Kawase, 1990)

The results of earthquake recording,

microtremor measurement, seismic intensity

distribution, and damage features are summarized as follows. Cliff is herein defined as the steeper form of

slope.

Slope: The dominant frequency and the amplification

vary with the height of the hill and the dominant

frequency varies with the incident angle of the input

motion. The dominant frequency of microtremor and

amplitude have good correlation with seismic intensities estimated from questionnaires.

Cliff: It was pointed out by the earthquake recording that amplification becomes larger to the direction orthogonal

to the cliff than the horizontally parallel direction to the

cliff and the difference becomes less for potion away

from the edge. It is also shown that major damage is observed for the direction orthogonal to the cliff. Mountain (Hill') : It was pointed out from earthquake

recording that larger amplification is found at the top of

the mountain and the dominant frequency differs between

top and skirt of the mountain.

Sediment-filled Valley: It was pointed out from

earthquake recording that the amplification and the

dominant frequency vary with the thickness of alluvial

deposit, and a surface wave is observed generated from

the edge of the valley. This tendency is observed for

seismic intensity distribution and the good correlation

between thickness of the alluvium and the seismic

intensity.

Dippin g Base Layer: It was pointed out from earthquake

observation that amplification and dominant frequency

changes with the depth of overlying layer at the dipping

portion, a surface wave is observed generated from the

edge of the base layer, and furthermore, dominant

frequency at the surface wary with the incident angle at the base layer.

For deep subsurface structure, the following has

been pointed out.

(1) Surface wave will be generated reflecting the deep

subsurface structure of the Kanto plain (basin) due to the

shallow earthquake occurring around Kanto plain except

for the rather deeper ones. This surface wave holds

specific dominant period corresponding to the deep

structure of the transmitting path.

(2) Many researchers support that the propagation of the

surface wave is not straight forward from the epicenter to the station, but the propagation or reflection occurs at the edge of the Kanto basin.

(3) The state of the art on the deep subsurface structure

in Kanto plain is not sufficient for the evaluation of

design motions for long period building.

3. DESCRIPTION OF THE ANALYTICAL

MODEL REPRESENTING THE

TOPOGRAPHY

The computer analyses were conducted for two sets of

representative topographies, that is, slope and sedimentfilled valley. The analytical model for slope came from

the existing landform at Kushiro District Meteorological

Observatory. Kushiro landform was selected for the

analyses. There had been dense earthquake recording

arrays and microtremor arrays. There has been

comparative earthquake recording after Kushiro

earthquake. The geological data around the site is known by the survey on the site.

118
The two dimensional models of topographies were used for the analyses. The analytical method used

.

here was finite element method for SV wave, finite difference method and boundary element method for SH

wave. The nonlinearity in soil property was not

considered in this study. The input motion for the

analysis is so called Ricker's wavelet. The result was

compared with the result by the one-dimensional

propagation model (denoted as ID hereafter). The

comparison was also made for identifying the

geographical position showing larger topographical

effect on seismic motion. Further, the degree of

simplification in forming the analytical model, the kinds

of seismic waves (P-SV, SH waves), influence of

impedance ratio and slope angle were also examined by

the parametric computations.

Computational results for slope

The model for computer analyses for slope are shown in Fig.2. The model- 1 is the realistic model derived from

the soil data. The model-2 is the simplified model of

model- 1 to check the modeling scheme. The model-3 and

model-4 are used to check the influence of slope angle

and impedance ratio of layers. The dominant frequency

of the input motion defined as exposed bedrock motion

is 4Hz as shown in Fig.3. For slope model, the

incident SH or SV wave is applied to the two

dimensional soil models. Comparisons were made

between two amplification ratios. One is the ratio

between surface response and input motion for two

dimensional case (2D hereafter). The other is those for one dimensional model for each location.

Since the influence of slope was found only at the limited portions closer to the lower part of the slope,

the comparative study was mainly made for the uphill,

the upper part of the hill. The analytical methods used

herein are 2-dimensional FEM and BEM. We had

acknowledged rough coincidence between the

computational results for two methods. Figure 4 shows

the spatial distribution of transfer functions of model1

for SH and SV wave incidence cases together with ID

case. The difference in transfer function is larger for

portions closer to the slope edge and becomes smaller

for distances away from the edge. When using the

simplified model-2, a similar tendency with model- 1 is observed in Fig.5. Two transfer functions for model1

and model-2 is compared in Fig.6 for SV. This

difference is caused by the model simplification. The

result shows that the modeling scheme has an influence on the transfer function. When the transfer function is normalizedwith the one with the ID basis, (denoted as normalized transfer function hereafter), the difference

between two normalized transfer functions becomes

smaller as shown in Fig.7. The ground response at the

upper part of the hill consists of two types of waves,

that is, the body wave propagating from the lower layer,

and the wave propagating horizontally from the slope.

The latter can be recognized as separate wave from

portions much away from the edge. This wave is called a later phase (wave). The propagation velocity for the

later phase is the phase velocity of the Love wave for

the SH wave incidence, and the phase velocity of the

Rayleigh wave for the SV wave incidence. Namely, the

later phase generated at the edge of each topography is dominated with surface wave. For portion away from

the edge, the generated later phase will be attenuated, and

there might be small difference in amplification

compared with the amplification from ID basis. It is seen from the analysis using SH and SV, that the peak

frequency of the transfer function is 3.5 to 6 Hz. The

maximum amplitudes of the computed responses

fluctuate along the hilltop. The maximum transfer

function occurs at 18 and 100 meters from the edge for

SH case, and at 40 meters for SV case. The maximum

amplitude in time history occurs at about 10 meters

from the edge for SH wave incidence, and at 20 meters

from the edge for SV wave incidence. The maximum

amplitude approximately 1.5 times larger than the one

for ID case. The influence of the different propagating

velocity of the later phase is more apparent in SH wave

incidence than in SV wave incidence for larger amplitude

variation.

The influences of slope angle upon the

normalized transfer function and the maximum

amplitude in time domain are shown in Fig.8 and Fig.9,

respectively. The influence of slope angle ranging 25 to 60 degrees becomes larger in transfer function for the

closer to the edge and for higher frequency. However,

there is a small influence of slope angle in both of

transfer function and time domain response.

The maximum amplitudes of ground responses

in reference with the results of ID results are shown in

Fig. 10 for 4 cases of impedance ratios. It is seen from

the figure that the influence of impedance ratio is relatively small.

Computational results for sediment-filled vallev

The analytical model for sediment-filled valley is shown

in Fig.l 1 The variations of the models are summarized

in Table 2 The model is symmetrical. The shapes of

the models were determined with reference of actual

landforms and considering the frequency content, number

of elements, dimension of element, computation

efficiency. These models are basically constructed as two dimensional half space with trapezoidal alluvial

deposit. The impedance ratio between two soil deposits

is assigned as 3 for standard case. The Ricker's wavelet is also used as input

motion for this case, however, the dominant frequency is 2.17 Hz and the peak amplitude occurs at 1.5 second

from the beginning in the time domain as shown in

Fig. 12. The analytical methods used are finite difference

method and boundary element method for SH wave

incidence cases and finite element method is used for SV

incidence case. The division of elements was made so

1 1 9
that we may discuss the dynamic properties less than 10

Hz.

The analyses were made for three types of

models with different horizontal widths of the valley. It

is seen from Fig. 13 that the secondary surface wave

generated at the edge of the valley propagates in horizontal direction, and repeats reflection, gradually

attenuates. The overlapping of waves changes the pattern

of distribution of peak values. As valley becomes

narrow, the peak amplitude tends to take its largest at the middle point of the valley. The transfer functions for

ID and 2D cases are compared at several points of the

surface. The fluctuation of the transfer function with

frequency is large when the damping is smaller. When

the damping is larger, however, the transfer functions for

both cases become smoother and are comparable each

other at points away from the dipping base. For areas close to the edge of the valley, the difference becomes

eminent. For cases with narrower valley, the influence

of the valley edge extends to the closer point to the

valley edge.

4. CONCLUSIONS

The typical site conditions in urban area considering the

irregularity in topography was surveyed.

The topographically irregular soil conditions in urban area in Japan are specified and represented with

dimension of landform, angle of slope, distance from the

edge, etc. The spatial variation of ground motion

properties such as maximum amplitudes, transfer function and predominant periods are computed for each

of cases. ACKNOWLEDGMENT

This research has been implemented as part of

the research project of the Ministry of Construction,

Government of Japan. For the project, a research

committee (Chairperson: Dr. Tadao Minami, Professor

of the University of Tokyo) was organized in the Japan

Institute of Construction Engineering (JICE). Under the

auspices of the committee, a working group (Head: Dr. Kazuoh Seo, Professor of Tokyo Institute of

Technology) was organized. The literature survey, model

formation and computations were conducted by the

members of working group.

REFERENCES

AD (1987). Reconnaissance Report of 1985 Mexico

Earthquake, (in Japanese)

BRI and BCJ (1992). A Guideline for Composing

Design Earthquake Ground Motion for Dynamic

Analysis of Buildings, Appendix of Annual Report

of the Committee on Technical Guideline for Design

Earthquake Ground Motion

BRI, JICE (1993). Annual Report of the Project for

1992 fiscal year, (in Japanese)

BRI, JICE (1994). Annual Report of the Project for

1993 fiscal year, (in Japanese)

Celebi, M., (1987). Topographical and Geological

Amplifications Determined from Strong-Motion and

Aftershock Records of the 3 March 1985 Chile

earthquake, B.S.S.A., 77, 1147-1167

Kawase, H. (1990). Topography effect at the critical SV-wave incidence: Possible Explanation of Damage

Pattern by the Whittier Narrows, California

earthquake of 1 October 1987, B.S.S.A., 80, pp. 1-22

Yamahara, H., (1971). Soils and School, Planning of

School Building, AU, (in Japanese)

Table 1. Average values of simplified topography parameters for areas of Japan

Tokyo Osaka Nagoya Yokohama Kawasaki

Slope 0(deg)

H(m)

10

10-20

< 5

10-30

40-60

5-15

20-40

5-20

Hill L(m)

H(m)

D(m)

100 - 500

5-10

300-900

<500

< 10

200-500

100 - 400

5-20

200 - 1300

Sedimentfilled Vallev

L(m)

H(m)

100 - 400

3-4

vary 5- 16

Dipping

Base Layer

L(m)

6(deg)

100-300 < 1

100 - 400

0.2 - 0.6

500

0.4 - 0.8

Table 2 Analytical model of sediment-filled valley

Sediment-filled Valley Bedrock

width

(m)

depth

(m)

side inclination side boundary

(deg) (m)

lower boundary

(m)

Model-A

Model-B

Model-C

200

100

50

15

15

15

45

45

45

200

200

200

75

75

75

120
(c) Sediment-filled Valley * *

(d) Dipping Base Layer

Fig.l Representative topography

T30m

Model1

T30m J

Model-2

y V«-140m/i,Vp-230m/»,H-7m / V«-300m/s Vp-1130m/i,H-9m

V«-600m/t,Vp-2850m/t

Vt-2S0m/«.Vp- 1 1 30m/i.H-1 6m

Vt-600m/«.Vp-2850m/«

I 16m

<t> = 25, 35, 45, 60 degrees

V«-250m/j.Vp-1130m/i.H-16m

Model-3

Tl6m

Model-4

Fig.2 Analytical model for slope

35

Vf600m/».Vp-28SOm/j

Vsb = 250. 500, 750, 1000 m/s

Vi-250m/t.Vp-1130m/i.H-!6m

ViB. Vp-28S0m/

• i I—i i i i

I I

' 2 3

TIME(SEC) 0.0 10°

0.0 2.0 40 6.0 8.0 10.0 12.0 FREOUENCVIHJ)

Fig.3 Ricker's wavelet as acceleration input motion at exposed bedrock

121
«' * m - m — m

NOUONHJ-l N0113HDJ-1

- £

r* — e

N01X3Nnj'l

at

noi iONnj 'i KOUONnji

- S

NOii3Nnj-i

NOI lONfli 1

- g

Nouonnj-i

»» — o

NOIiONflJ '1

122
*

•** o

/

* 1

1 -\A

/ £

/ >\

f\

V \

m

:

oucnc

$

ac

r*

o

0t e*

• • ...

• r O

»- - s

sue N * a

M

•

Ollttl

. S

01 UK OUT*

0I1TH

gun

- S

3

w

O

01 UK

1

—I- XU

* S

C

i3 (N

"S 3 ^

IS o

° s

o i

i

i

rs " H .-2 -

H O

T3

.3 «

S e

Oh 3

r0U»» 0I1VK

^ • «iNouoNnj'i

N0I13«IU'1

7

i

i

-

sV

rw — o

N0I13W1J1

HOUONni'i

a

^

f

- sw

_ 3

HOIlONdj'l

1-

\

C

o

4—

>

o

a

.3 <N

(O

T3

a 3

d

1 Q O

3 S o zL

is V 13

-7? T3

1) Q

$ t-l Oh ,0

VO

nouonnj-1 NOUDNnj'i NOUONflJ-l

123
1 1 It 30 01 Ofl 120 ISO

distance from the edge distance from the edge

Fig.8 Normalized transfer function Fig.9 Maximum amplitude Fig. 10 Normalized maximum amplitude

p2=1.6, V2=130m/s

p1 = 1.8, V1=400m/s

170m

200m

Fig. 1 1 Analytical model of sediment-filled valley for Model-A

i. Or

<=> 0. 5 owo> 0.0

u< -0. 5

-1.0

J

1

0 2 4.6 8 10 12 1 4

Time (sec)

Fig. 12 Ricker's wavelet as acceleration input motion at exposed bedrock

Fig. 13 Response waveform at the surface and the distribution of maxima for Model-A and Model-C

124
Preliminary Estimates of Ground-Motion Amplification

for the Northridge and Hyogo-ken Nanbu Earthquakes ofJanuary 17, 1994 and 1995

by

Roger D. Borcherdt*

ABSTRACT

Strong ground shaking generated by the

Northridge and Hyogo-ken Nanbu (HanshinAwaji) earthquakes of January 17, 1994 and

1995, resulted in combined property losses

exceeding 120 billion U.S. dollars, and 5300

deaths, either directly through shaking-induced

structural failure or indirectly through shakinginduced liquefaction and landslide failures. The

strong-motion instruments, which recorded

these earthquakes provided unprecedented sets of ground motion measurements on both stiffand soft-soil deposits. These measurements

provide important new estimates of groundmotion amplification by soil deposits at high

input shaking levels to better understand the

nature resultant earthquake losses. Preliminary

implications for site dependent design spectra

are examined. Results presented here are reproduced because of interest for this Joint

Meeting from Borcherdt (1996a and 1996b).

KEYWORDS

Site amplification, strong motion, attenuation,

site-specific design.

1. INTRODUCTION

The extensive losses to property and life from

the Northridge and Hyogo-ken Nanbu

earthquakes of January 17, 1994 and 1995 were

due in large part to strong ground shaking.

Consequently, understanding its nature in

relation to source rupture characteristics, crustal

propagation effects and effects of local site

* United States Geological Survey

Menlo Park, CA 94025

conditions is of special interest for new sitespecific building code provisions. Strongmotion stations in the Los Angeles and KobeOsaka basin regions provided unprecedented

documentation of the nature of strong shaking

near the earthquake source. The Northridge

earthquake provided forty-two recordings on

sites underlain by materials ranging from stiff

soil to hard rock within 25 km of the epicenter

with ground acceleration exceeding 0.9 g at seven sites. The Hyogo-ken Nanbu earthquake

provided more than 26 recordings within 25 km

of the extended rupture surface with many of

these sites being located on "soft" Holocene soil

deposits. These data provide an especially

important opportunity to document the nature of

strong shaking on "stiff' and "soft" soil sites at high levels of shaking near the source.

2. SOURCE RUPTURE CHARACTERISTICS

The Northridge earthquake occurred in a region

of compressional tectonics associated with a major regional bend in the strike of the San

Andreas fault. The earthquake of moment

magnitude Mw

6.7 occurred on a thrust fault,

dipping 42 degrees to the south with a strike of

122 degrees. Inversion of strong-motion and

teleseismic data (Wald and Heaton, 1994),

suggest that rupture initiated at a depth of about

18-19 km, extended to within 7-8 km of the

earth's surface along a slip vector oriented 109°.

Rupture extended updip about 20 km over a

surface about 14 x 15 km and lasted about 7

seconds yielding a seismic moment of 1.2 x 10

26

dyne-cm with the largest slip of nearly 4.0

meters occurring near the deepest portion of the

125
fault (17-18 km). Strong ground motions were

significantly larger north of the epicenter due to

the directivity associated with updip propagation

of the rupture.

The Hyogo-ken Nanbu earthquake occurred in an area of complex faulting about 200 km from

the Nankai trough, which forms the boundary

between the Philippine Sea and Eurasian plates.

The earthquake of magnitude 7.2 (JMA) and

Mw

6.9 is the largest and most severe earthquake

to affect the region this century. Source rupture

characteristics inferred from teleseismic and

regional data by Kikuchi (1995) suggest rupture

initiated at a depth of about 14.3 km on the

Nojima fault It was comprised of three

subevents of moment magnitudes 6.8, 6.3, and

6.4 each with a predominant strike-slip source mechanism. Rupture occurred along a 46 km

segment of the fault at an average depth of about

8 km and reached the surface along a 9 km

segment with maximum surface displacements

of 1.7 m right-lateral, horizontal and 1.0 m

vertical (T. Masaharu, pers. commun.,1995).

Duration of total rupture lasted about 11

seconds.

Comparison of source characteristics suggests

the earthquakes are about the same magnitude

(Northridge, Mw

6.7; Hyogo-ken Nanbu, Mw

6.9), with longer rupture duration for the

Hyogo-ken Nanbu earthquake (11 seconds as opposed to 8 or 9 seconds). The rupture

mechanism is predominantly strike slip or

horizontal for the Hyogo-ken Nanbu earthquake

compared to relatively larger vertical

components of motion associated with thrust

faulting for the Northridge earthquake. The

average depth of rupture and hypocentral depth

for the Hyogo-ken Nanbu earthquake is less than

that for the Northridge earthquake. Rupture for

the Hyogo-ken Nanbu earthquake reached the

surface on Awaji Island and may have come

within 4 km of the surface beneath Kobe City.

Rupture was inferred not to extend closer than 6

to 7 km of the surface for the Northridge event.

3. GROUND MOTION ATTENUATION

Locations and peak amplitude information for

the strong motion recordings have been

compiled from a variety of original data sources

for the Northridge earthquake (Borcherdt, 1995)

and the Hyogo-ken Nanbu earthquake

(Borcherdt, 1995b). Contour maps of peak

vertical and horizontal acceleration were derived

from the strong motion recordings of both

events and superimposed upon available digital

geologic maps. (Wentworth, et al.,1994; 1995;

Borcherdt and Wentworth, 1995). The contour

maps for the Northridge earthquake show a

general north-south elongation consistent with

the inferred thrust rupture characteristics of the

source. The contour maps for the Hyogo-ken

Nanbu earthquake suggest a linear zone of

intense ground shaking underlain by Holocene

soil deposits parallel to the surface projection of

the crustal rupture zone. The maps for both

earthquakes emphasize that the peak values

decrease rapidly away from the immediate

epicentral region with local anomalies due to

topographic and local site amplification.

Peak horizontal accelerations as reported for the

Hyogo-ken Nanbu and Northridge earthquakes

are plotted versus the closest distance to

seismogenic rupture. Attenuation curves developed by Boore et al. (1994) are superimposed with projected surface rupture

distance converted to seismogenic-rupture

distance using average depths of 4 and 6 km,

respectively (Fig. 1). Curves of Campbell and

Borzognia (1994) also are superimposed. The

peak values for the Hyogo-ken Nanbu

earthquake tend to be less than those for the

Northridge earthquake for sites within 4 to 20

km of the seismogenic source, but greater at distances exceeding about 30 km. These

differences in amplitude are roughly consistent

with differences in source mechanism and types

of geologic site conditions. The source mechanism for the Northridge earthquake is

predominantly vertical movement on a thrust

fault Such source mechanisms typically

generate higher peak accelerations near the source than comparable sized strike-slip

mechanisms. The geologic conditions beneath a

majority of the recording sites for the Hyogoken Nanbu earthquake are soft-soil (Holocene)

deposits with higher amplification capabilities

for ground motions near one second period than

126
the stiff Pleistocene deposits beneath many of

the recording sites for the Northridge

earthquake. Consequently, the high frequency

motions near the source might be expected to be

larger for the Northridge earthquake, but the

longer period motions at some distance from the source might be expected to be larger for the

Hanshin- earthquake. Attenuation curves derived by Boore, et al, 1994 when extrapolated

to soft soils using an average shear-wave

velocity of 200m/s fit the data quite well.

4. INFLUENCE OF LOCAL SITE

CONDITIONS

The relatively large accelerations recorded at several of the sites for both earthquakes raise

important questions concerning the influence of

local site conditions on high levels of incoming

high-level bedrock motions near the source. To

examine the influence of local site conditions,

the peak motions and amplitude spectra

(Northridge) for each component were

normalized by the corresponding value

determined for a nearby station on firm to hard

rock. Each ratio was also normalized by the

corresponding reciprocal ratio of distances to the

zone of largest energy release. Groups of

stations for normalization were determined from

3

i

"5

o

u

<

1

•a

10.000 f

1.000

0.100

0.010 r

0.001

Hanshin-Awaji & Northridge Earthquakes

o o o <D

- - - Alluvium (Campbell & Bozorgnia; (1994) o Northridge PGA

• Hanshin-Awaji PGA

Soft Soil, 200 m/s Boore, et aL, 1994 PGA - Std Dev

PGA + Std Dev

Hard Rock, 1350 m/s Boore, et aL, 1994

1

o o • H •

I I I I I I

10 100

Distance to Seismogenie Rupture (km)

1000

Fig. 1. Preliminary compilation of peak horizontal acceleration as a function of closest distance to

seismogenic rupture for the Northridge and Hyogo-ken Nanbu (Hanshin-Awaji) earthquakes.

127
22.5 (Northridge) and 45 (Hyogo-ken Nanbu)

degree azimuthal windows measured with

respect to polar coordinate systems oriented

parallel to the strike of the faults and segmented

into 25 km distance intervals with origin at the

epicenter. If no normalization station occurred in

the window, then the closest normalization

station in the azimutha] window was chosen (see

Borcherdt, 1994 and 1995 for detailed

discussion).

4.1 Amplification vs. Input Ground Motion

Level for "Stiff' Soil Sites (Northridge

Earthquake)

To quantify the dependence of the measured

amplifications on input ground motion level for

the Northridge earthquake, the acceleration

ratios and the average spectral amplifications are plotted as a function of the "input" or base

acceleration inferred at each site (Figs. 2a, 2b,

and 2c). The "base" acceleration is defined as the value recorded at the surface normalized by

the inferred acceleration ratio. Plots are shown

for ratios of average peak horizontal (radialtransverse) acceleration (Fig. 2a) and average

spectral ratio, Fa ,for the short-period (0.1 - 0.5

s;) band (Fig. 2b) and the average spectral ratio,

Fv ,for mid-period (0.4 - 2.0 s) band (Fig. 2c).

Geologic classifications for the sites were

determined from recent digital GIS compilations

of the surficial geology for the Los Angeles

basin (Wentworth, et al., 1995). The sites are

classified according to recently defined site

classes (Borcherdt, 1994) into a combined

"stiff' soil to "soft" rock site class, designated

SC II+III or using NEHRP designations C+D

and a "firm-hard" rock site class, designated

SC-Ib or NEHRP B. Use of the combined site

class reduces uncertainty in the site

classification, until ongoing detailed

investigations are completed at each site. Shearwave velocities for SC II+III range from 200 to

700 m/s with an average velocity inferred to be

about 363 m/s. Shear-wave velocities for the SC

lb exceed 700 m/s with an average velocity of

about 867 m/s. Rigorous definitions of the site

classes are provided by Borcherdt, 1994.

The acceleration ratios show a slight trend to

decrease with increasing "base" acceleration

level (Fig. 2a) as evident in the small negative

exponent of the power-law regression curve. The zero slope of the linear regression curve

suggests no dependency on base acceleration

level.

The average spectral ratios for the short-period

(0.1 - 0.5 s;) band (Fig. 2b) suggest a slight

tendency to increase with increasing base

acceleration level as evident in the positive

exponent and slopes for the regression curves. However, scatter in the data suggests that this

trend is probably, not meaningful. The average

spectral ratio for the mid-period band (0.4 - 2.0

s) band (Fig. 2c) tends to increase with

increasing base acceleration level. These

preliminary observations are consistent with

more detailed observations for the entire data set (Borcherdt, 1995).

4.2 Preliminary Amplification Estimates vs. Input Ground Motion Level for "Soft"

Soil Sites (Hyogo-ken Nanbu

Earthquake)

As a preliminary attempt to quantify the

response characteristics of "soft" soil deposits

(Holocene deposits) at high input-ground

motion levels for the Hyogo-ken Nanbu

earthquake, ratios of average peak horizontal

acceleration were computed with respect to the

Kobe University (KC-KUN) site for which an

average peak acceleration of 270 cm/s

2 g was

recorded at a distance of about 1.3 km from the

surface projection of the crustal rupture zone. The site is underlain by hard rock (Sugito,

1995). Geologic classifications for the other

sites were obtained from the GIS superposition

of station locations on the digital Geologic Map

of Japan (1:1,000,000 scale, Geological Survey

of Japan, 1992, Borcherdt and Wentworth,

1995). The resultant classification is preliminary

as it is dependent on both the accuracy of the

station coordinates and the published map scale.

The average of the acceleration ratios,

normalized by the reciprocal ratio of the

seismogenic-rupture distance, are plotted versus "base" or input acceleration levels. The ratios

128
are plotted only for sites underlain by Holocene

soils that are located along strike at distances

less than 22 km (Fig. 3). The mean and

standard deviation for these acceleration ratios are 2.6 and 1.3, respectively.

The ratios of average peak acceleration show a

well defined trend to decrease with increasing

"base" acceleration level (Fig. 3). This trend is

evident in both the negative slope of the linear

< 4

$ -2 «3- c 3

0

-

1• | , © 2

13

8 1 i

w

Stiff soils - soft rock;SC //+/// (C+D) / granite

y « O.OOx 1 .23 R

2 - 0.00

(a) v

y- 1.15x 00 R

2 - 0.00

0 1 2 3 4 5 6

Base Acceleration = Surface Acceleration / Acceleration Ratio (nVs2

)

5 c

o

•a8

1 4

a. I ~

15 M 3

ind

2

c» 2- E

§

1 i.«

Stiff soils - Soft rock;SC //+/// (C+D) / granite

#

^

(b)

»

y-0.12x+ 1.34 \ R

2 »0.05

y- 1.38x004 R

2 « 0.00

0 1 2 3 4 5 6

Base Acceleration = Surface Acceleration / Acceleration Ratio (m/s2

)

Stiff soils - Soft rock; SC //+/// (C+D) / granite

0 1 2 3 4 5 6

Base Acceleration = Surface Acceleration / Acceleration Ratio (m/s2

)

129
Fig. 2. Acceleration ratio (a), average spectral amplification ratio for the short-period (0. 1 - 0.5 s) band (b),

and average spectral amplification ratio for the mid-period (0.4 - 2.0 s) band (c) versus base

acceleration for "stiff soil - soft rock sites" (SC n+HI; NEHRP C+D) determined from strong-motion

recordings of the Northridge earthquake. Power law and linear regression curves fit to the data are superimposed.

Fig. 3. Preliminary acceleration ratios versus base acceleration for "soft" Holocene soil deposits (SC IV;

NEHRP E) determined from peak acceleration reported for the Hyogo-ken Nanbu earthquake. Power

law and linear regression curves fit to the data are superimposed.

regression curve and the negative exponent of

the power law regression curve. These results

must be considered as preliminary. Further

evaluation, based on detailed site evaluations are needed to improve quantification of these results

5. PRELIMINARY IMPLICATIONS FOR

SEISMIC DESIGN PROVISIONS

Improvements in estimates of site-dependent,

earthquake-resistant design spectra are under

review for incorporation into United States

building code provisions (NEHRP and UBC).

These new developments better account for the

amplification effects of local geological deposits

in earthquake resistant design. They are based

on new unambiguous definitions of site classes

and amplification factors derived from empirical

and numerical modeling results. Values for the

amplification factors as a function of input

ground motion level and site class are specified

by Borcherdt (1994).

The amplification factors Fa and Fv for shortand mid- period motion are predicted as a

function of mean shear-wave velocity v for

various input ground-motion levels, /, with

respect to a reference ground condition by the

following equations:

Fa (v,I ) = (v0 /vf\

(la)

and

FyivJ )=(V0 /V)

/"V

,

(lb)

where, mt and mv are exponents which depend

on the level of base input motion. These

equations can be used to rigorously compare

results derived here.

Results derived from the Northridge earthquake

for sites on "stiff soils - soft rock" suggest that

the amplification factors Fa and Fv do not show

a strong tendency to decrease with increasing

130
"base" acceleration. This trend as specified in

Figs. 2b and 2c by the power law regression curves is replotted in Figs. 4a and 4b.

Superimposed on these trends are the trends

implied by recent NEHRP code provisions.

(The code trends are specified by equations la

and lb with v = 363 m/s for SC - II+ffl, v0 =

867 m/s, mt = 0.35 and -0.05 and mv = 0.65 and

0.45 for input acceleration levels of 0.1 and 0.4g

;Borcherdt, 1994.)

c

I

CLE

< »

2 in o d

8.

' w T

<u 2. ei

iin

u.

Fa vs Base acceleration

(Stiff soils - Soft rock, SC //+/// (C+D)

(a)

Northridge

y = 1.38x° M

NEHRP Code

y = -0.14x + 1.50

Base Acceleration = Surface Acceleration / Acceleration Ratio (m/s )

c

o1

CL

CM w ** 2

2

I

1

II

Fv vs Base acceleration

(Stiff soils - Soft rock; SC //+/// (C+D)

(b)

Northridge

y = 1.6x°

21 NEHRP Code

y = -0.10x + 1.86

Base Acceleration = Surface Acceleration / Acceleration Ratio (m/s2

)

Fig. 4 Site-specific amplification factors versus base acceleration level for sites on stiff soils to soft rock

(SC II+UI; C+D) implied by the Northridge earthquake and the NEHRP code provisions for the shortperiod band, (a) and the mid-period band, (b).

The amplification factors implied by the

Northridge earthquake are in good agreement

with those suggested for the code provisions at the 1 m/s2 base acceleration level. For higher

levels of acceleration the factors implied by

Northridge become increasingly greater than

those proposed for the NEHRP code provisions.

However, for the "stiff soil -soft rock" sites

being considered the absolute amounts of

amplification are not large in comparison with

those for soft soils and hence, may not be of

significance for many applications. These

preliminary Northridge results suggest that for

"stiff soil - soft rock" sites similar to those in

the Los Angeles basin that Fa and Fv values of

131
(1.4, 1.8) as predicted by equation 1 for base

ground acceleration levels near 1 m/s

2 base

provide simple and conservative estimates for

site-specific design.

The preliminary results from the Hyogo-ken

Nanbu earthquake for "soft" soil sites suggest

that the acceleration ratios and hence the Fa

factors do show a well defined decrease with

increasing base acceleration level. This trend as

specified in Fig. 3 by the power law regression curve is replotted in Fig. 5. Superimnposed on

this trend is the trend implied by recent NEHRP

code provisions for "soft soils". (The code trend

is specified by equations la with v = 150 m/s

for SC IV (E), v0 = 1050 m/s for SC lb (B), m,

= 0.35 and -0.05 for input acceleration levels of

0.1 and0.4g ;Borcherdt, 1994.)

The very preliminary results reported here for

the Hyogo-ken Nanbu earthquake for "soft"

soils appear to be in relatively good agreement

with those suggested for the code provisions in

that the general trend and amounts of

amplification suggested by the power law

regression curve are similar to those in the

proposed code provisions. These preliminary

results suggest that the values of amplification

observed in the Kobe City area may be larger

than those currently specified for Fa in the

current NEHRP provisions. Additional

information on seismic velocity profiles at each

site are needed for a definitive conclusion.

8

CLE

< —

m

W 3 Em

o d

2L

•

a) 2.

2

E

3 1

iim

Fa vs Base acceleration

(Soft Soils; SC IV (E))

Hanshin-Awaji

y = 3.1xTM

NEHRP Code

y = -0.37x + 2.34

1 2 3 4 5 6

Base Acceleration = Surface Acceleration / Acceleration Ratio (m/s2

)

Fig. 5 Site-specific amplification factors versus base acceleration level for sites on soft soils (SC TV;

NEHRP E) implied by the strong ground motion recordings of the Hyogo-ken Nanbu (Hanshin-Awaji)

earthquake and the NEHRP code provisions for the short-period band band.

ACKNOWLEDGMENTS

The important efforts of various public and

private agencies in the acquisition of the strong

motion data reported herein are appreciated,

including CDMG, USGS, USC, and LA Dept of

Water and Power for the Northridge data and

Public Works Research Institute, The National

Research Institute for Earth Science and Disaster

Prevention, Science and Technology Agency,

The Disaster Prevention Research Institute,

Kyoto University, the Earthquake Research

Institute, University of Tokyo, Department of

Civil Engineering, Gifu University, the Port and

Harbor Research Institute, the Architectural

Institute of Japan, The Committee on

Earthquake Observation and Research in the

Kansai Area, Osaka Gas, Japan Railway, Japan

Highways, Japan Meteorological Agency,

Ministry of Construction, Building Research

Institute, Geological Survey of Japan and others

for the Hyogo-ken Nanbu data are appreciated.

A special note of thanks is expressed to Y.

Iwasaki, K Ohtani, M. Sugito, and M. Watabe,

and numerous other colleagues for information

and personal contributions. The efforts of C.

132
Wentworth in compiling the geologic maps

required for geologic classifications of the

strong-motion stations are appreciated. The

efforts of G. Glassmoyer in assembling the

Northridge data for spectral processing were

especially helpful.

REFERENCES

Boore, D.M., Joyner, W.B., and Fumal, T.

(1994). Estimates of response spectra and

peak accelerations from western North

American earthquakes; an interim report part

2 (with insert). U. S. Geological Survey Open

File Report 94-127, 39 pp.

Borcherdt, R.D. (1994b). Estimates of sitedependent response spectra for design

(Methodology and Justification), Earthquake

Spectra, 10, p. 617-653.

Borcherdt, R.D. (1995). Strong ground motion

generated by the Hanshin-Awaji (Hyogo-ken

Nanbu) earthquake of January 17. 1995 in

'The Hyogo-ken Nanbu earthquake of

January 17. 1995" J.P Moehle, ed.,

Earthquake Engineering Research Center

Report No. , University of California,

Berkeley, CA, in press.

Borcherdt, R.D. and Glassmoyer, G. (1992). On

the characteristics of local geology and their

influence on ground motions generated by

the Loma Prieta earthquake in the San

Francisco Bay region, California: Bulletin of

the Seismological Society of America, 82, p. 603-641.

Borcherdt, R.D. and Wentworth, C.W. (1995a).

Strong-ground motion generated by the

Northridge earthquake of January 17, 1994:

Implications for seismic design coefficients

and seismic zonation, Fifth International

Conference on Seismic Zonation, Procs.,

Nice, France, v. II, pp 964 - 971.

Borcherdt, R. D. and Wentworth, C.W. (1995b).

Preliminary map of peak horizontal ground

acceleration for the Hanshin-Awaji

earthquake of January 17,1995, Japan, Part H

of 2 - Description of mapped data sets, U. S. Geological Survey Open File Report No. 259

B, pp.

Borcherdt, R.D., 1996a, Preliminary

amplification estmates inferred from strongmotion recordings of the Northridge

earthquake of January 17, 1994, International

Workshop on Site Response, Procs., vol. 2,

pp 21-46.

Borcherdt, R. D., 1996b, Strong ground motions

generated by the Northridge and HanshinAwaji earthquakes of January 17, 1994 and

1995; Implications for site specific design

factors, Eleventh World Conference on

Earthquake Engineering, Procs., in press.

Campbell, K. W. and Bosorgnia, Y. (1994).

Near-source attenuation of peak acceleration

from worldwide accelerograms recorded

from 1957 to 1993, Proc. Fifth U.S. National

Conference on Earthquake Engineering,

Chicago Dlinois, 1994.

Geological Survey of Japan, 1992, Geological

Map of Japan, CD-ROM Version, Third

Edition, CD-ROM series of Earth Science

Databases, CDGSJ92010, Map Scale

1:1,000,000.

Kikuchi, M. (1995). The Mechanism of the

Hyogo-ken Nanbu earthquake of January

17,1995, in YCU (Yokohama City

University) Seismology Report No. 38.

Kuribashi, E. (1995). The Japan's earthquake in

Kobe of January 17,1995, The Hyogo-ken

Nanbu earthquake), a reconnaissance report,

Toyohashi University of Technology,

Research Report No. 95-1.

Nakakita, U., and Watanabe, Y. (1977). Soil

stabilization by preloading in Kobe Port

Island, in Proceedings of the 9th

International Conference on Soil Mechanics

and Foundation Engineering, Tokyo, Case

History Volume, p. 611-622.

Nakamura,Y. (1995). Waveform and its analysis

of the 1995 Hyogo-ken-Nanbu earthquake,

JR Earthquake Information No. 23c, 46 pp.

Nakamura,Y., Kazutoshi, H., Saita, J., and Sato,

S. (1995). Strong Accelerations and damage

of the 1995 Hyogo-ken Nanbu earthquake,

JR Earthquake Information No. 23b, 25 pp.

133
Prompt report on strong-motion accelerograms

No. 46 (1995). The National Research

Institute for Earth Science and Disaster

Prevention, Science and Technology Agency.

42 pp.

Wentworth, C.W., Borcherdt, R. D., and Mark,

R. (1995). Preliminary map of peak

horizontal ground acceleration for the

Hanshin-Awaji earthquake of January

17,1995, Japan, Part I of 2, U. S. Geological

Survey Open File Report No. 259 A, pp.

Porcella, R.L., Etheridge, E.C., Maley, R.P.,

and Acosta, A.V. (1994). Accelerograms

recorded at USGS national strong-motion

network stations during the Ms=6.6

Northridge, California earthquake of January

17, 1994, U. S. Geological Survey Open-File

Report 94-141, 100 pp.

Shakel, A, Huang, M., Darragh, R., Cao, T.,

Sherburne, R., Malhotra, P., Cramer, C,

Sydnor, R., Grazier, V., Maldonado, G.,

Petersen, C, and Wampole, J. (1994).

CSMIP Strong-motion records from the

Northridge, California earthquake of January

17, 1994, California Strong-Motion

Instrumentation Program Report No. OSMS

94-07, 308 pp.

Sugito, M. (1995). Characteristic of strong

ground motion, in Comprehensive Study of

the Great Hanshin Earthquake,\Jmtod

Nations Centre for Regional Development

(UNCRD), UNCRD Rept. 12, 244 pp.

Tognazzni,R.A. (1994). Processed LADWP

power system strong-motion records from

the Northridge, California earthquake of

January 17,1994, Linvall, Richter, Benuska,

No. 007-027.

Trifunac, M.D., Todorovska, M.I., and

Ivanovic\ S.S. (1994). A note on the

distribution of uncorrected peak ground

accelerations during the Northridge

earthquake of 17 January, 1994, University

of Southern California, in press.

Wald, David J. and Heaton, Thomas H. (1994).

A dislocation model of the 1994 Northridge,

California determined from strong ground

motions, U.S. Geological Survey Open-File

Rept. 94-278, 53pp.

Wentworth, CM., Borcherdt, R.D., Mark, R.K.,

and Boore, D.M. (1994). Maps of peak

acceleration recorded for the Northridge,

California, earthquake of January 17, 1994

and general geology of the epicentral region,

U.S. Geological Survey Open-File Report

94-197,3$$.

134
Damage to Communication Tunnels Caused by

Liquefaction and Other Phenomenon

During the Hanshin-Awaji Earthquake and Future Countermeasures

by

Yuzo Yamaguchi* / Masayuki Tsuchiya*

ABSTRACT

The Hanshin-Awaji Earthquake which occurred on

January 17, 1995 caused enormous damage to above ground buildings including reinforced concrete buildings. In the Kobe district, where the

damage was particularly severe, there are approximately 4.8 km of open-cut tunnels and 5.2 km of

shield tunnels. However, the cables encased in

these tunnels sustained absolutely no damage

whatsoever and the tunnels are able to sufficiently

carry out their intended functions, thus verifying

their high reliability with respect to earthquakes.

However, examining the tunnels as underground

structures showed that lateral and vertical faulting

of up to ten or more centimeters had occurred at the expansion joints of open-cut tunnels attached to buildings and shafts, etc. in areas experiencing a

seismic intensity of 7 (ruinous earthquake), and

that leakage had occurred in these areas. In addition, cracks had formed in a number of walls in

general tunnel sections. Particularly noteworthy was the fact that even among areas which experienced the same level of seismic intensity, this

damage was more severe in areas where liquefaction occurred. This thesis aims to illustrate the state of earthquake damage to open-cut tunnels and

shield tunnels, and to describe future earthquakeproofing measures centering on existing tunnels.

KEYWORDS: earthquake-proofing measures,

Hanshin-Awaji Earthquake, liquefaction, open-cut

runnel, shield tunnel

illaries, pipeline and highly-reliable conduits

which correspond to arteries, and tunnel facilities

which correspond to main arteries. (Fig. 15)

Of these, tunnel facilities are constructed near

buildings containing switchboards and other places

where cables converge, and are capable of housing

many cables by laying these cables on shelves

provided to the left and right of a central passageway. In addition, not only can people enter the

insides of tunnels to lay cables and perform maintenance work without affecting road traffic or the

local environment, but tunnels also boast the highest reliability with respect to earthquakes. Tunnel

construction methods in urban areas are broadly

divided into open-cut tunnels and shield tunnels.

Sectional views of these tunnels are shown in Figs.

1 and 2.

\

/ W \

\ , , f

V

/ N—1

1 x

Fig. 1 Sectional drawing of a open-cut tunnel

1. CONFIGURATION OF NTT'S COMMUNICATION CIVIL FACILITIES

Civil facilities which house communication cables can be broadly classified into the four categories of

wiring pipeline facilities which correspond to cap-

* NIPPON TELEGRAPH AND TELEPHONE CORPORATION'

NTT Access Network System Laboratories

1-7-1 Hanabatake, Tsukuba-city. Ibaraki. 305 Japan

135
Fig. 2 Sectional drawing of a shield tunnel

2. EXISTING EARTHQUAKE-PROOF

TUNNEL STRUCTURES

The main types of earthquake-proof structures for

current open-cut tunnels and shield tunnels are described below.

(1) Open-Cut Tunnel Expansion Joints

Expansion joints are installed in the following

types of locations in order to release the stress produced by earthquakes in the direction of the

tunnel axis. (Fig. 16)

a) Places where tunnels are attached to buildings,

approximately 1 to 2 m on the tunnel side from

the outer wall of the building

b) Places where the soil changes noticeably

c) Places where the tunnel cross section changes

noticeably

d) Places where pile foundations, etc. have been

laid and support conditions vary

Expansion joints generally adopt a collar structure. However, a socket structure is used when it is particularly necessary to integrate the tunnel with a

structure, or when integration is considered best

from a construction viewpoint. Normally, the gap

for these joints is 2 cm and asphalt materials are used as joint fillers. Joint fillers are also used to break the connection between collars and the outer

walls of tunnels.

(2) Secondary Covering of Shield Tunnels with

Earthquake-proofing Reinforcing Steel Bars

Earthquake-proofing reinforcing steel bars (rebars) are inserted within the secondary covering in

the following types of locations in order to increase

the cross sectional rigidity of tunnels to withstand

the stress generated by earthquakes. (Fig. 17)

a) Places where the soil changes suddenly

b) Places where shield tunnels are attached to

shafts

Regarding types of earthquake-proofing re-bars, in

the case of a), D13 re-bars are installed at 30 cm

intervals for 25 m to the left and right of locations

where the soil changes suddenly. In the case of b),

D22 re-bars are installed at 30 cm intervals for

12 m from the shaft.

3. DAMAGE

(1) Overview of Communication Tunnels in Kobe

Kobe City contains approximately 10 km of communication tunnels, of which 4.8 km is open-cut

tunnels and 5.2 km is shield tunnels with the oldest

tunnel constructed in 1964. The overlying layer is approximately 2 to 4 m for open-cut tunnels, and

approximately 10 to 25 m for shield tunnels.

(Fig. 18)

(2) Damage

Most of the Kobe tunnel routes run through areas which experienced a seismic intensity of 6 or 7. Although above ground buildings including reinforced concrete buildings suffered extensive damage in these areas, the damage to open-cut tunnels

was limited to cracks in some walls and leakage

caused by joint displacement. For shield tunnels,

the only damage occurred in locations where shield

tunnels had protruded approximately 1 cm into the

interior of some shafts. The results of a survey

conducted of tunnels in the Kobe Sannomiya area following the earthquake are described below.

Note that the maximum acceleration measured at the point closest to these areas was between 600 to 800 gal.

a) Damage to expansion joints

The earthquake produced offsetting in some ex- pansion joints, causing groundwater to flow into

the tunnels in some of these locations. Of the 37

joint locations surveyed, 17 locations showed evidence of offsetting. (Photo 1, 2)

Although all of the surveyed joints were located in areas experiencing a seismic intensity of 7, the

damage in areas where liquefaction occurred

136
clearly differed from that in areas where liquefaction did not occur. Fig. 19 shows the expansion

joint displacement. Here, the maximum displacement was 18 cm. Fig. 3 shows the occurrence of

joint widening by attachment location for joints

experiencing displacement of 2 cm or more. This

table reveals a clear difference between areas with

and without liquefaction.

Fig. 4 shows the occurrence of leakage of 50 (/mm

or more from joints by attachment location.

100-t— —

—

—

>/:j:?/:3 >/:i

9 /1 8

2/10

0/1

1/8

n

>

NTT

offices

ft|

Shafts

3"

Portions

with

structural

changes

EJGeneral

portions

I;

Overall

>

NTT

offices

ft |

Shafts

§•

Portions

with

i

structural

changes

si

General

portions

W

Overall

5'3

Fig. 3 Locations experiencing joint widening of

2 cm or more

r—

)^

»/:j

/:!6 /\ 8 /:3' T

0/1 0/1 0/8 0/10

1-

3

NTT

offices

f

Shafts

=:

Portions

with

structural

changes

2

General

portions

Overall

|

NTT

offices

fc> ?.

Shafts

g

Portions

with

Z.

structural

changes

%

General

portions

S.

Overall

3

Fig. 4 Joint locations experiencing leakage of

50 rf/min or more

Leakage from the joints of open-cut tunnels caused

groundwater to flow into these tunnels. This

groundwater then flowed from the open-cut tunnels to shield tunnels where it accumulated at the

lowest point. Normally, water leaking into the

tunnels is pumped up above ground by pumps

installed inside the tunnels. However, in addition to producing high amounts of leakage from expansion joints, this earthquake also caused a power

stoppage inside the tunnels thereby preventing

regular pump operation. Therefore, movable

pumps were carried into the tunnels in order to pump up this water.

b) Damage to general portions of open-cut tunnels

The earthquake also created cracks in the general

portions of tunnels. The number of cracked locations and occurrence ratio for areas with and

without liquefaction are shown in Table 1. The

amount of leakage from cracked locations was

relatively small compared to that from expansion

joints, and these locations were repaired by injecting urethane, etc.

Table 1 Damage to general portions

Open-cut

tunnel length

No. of cracked

locations

No. of cracked

locations per 100 m

Areas with

liquefaction

600 m 37 6.2

Areas without

liquefaction

480 m 1 0.2

Total 1080 m 38 3.5

c) Damage to cast iron manhole covers The earthquake motion produced offsetting in the

neck portions of cast iron manhole covers for

open-cut tunnels, resulting in leakage in some of

these locations. Of the 22 manhole covers sur- veyed, displacement was confirmed in 11 locations. Cast iron manhole covers (Fig. 20) differ

from tunnel joint structures in that the displacement resistance is the same in all directions.

Therefore, manhole covers are determined to be a useful reference for understanding characteristics

of earthquake motion (scale, direction) occurring on the ground surface. Surveys were conducted

regarding the amount and direction of manhole

cover displacement, and the sum total of the

displacement was calculated. The results showed

137
that the displacement in the north-south direction

far exceeded that in the east-west direction. This

indicates that the earthquake motion in the northsouth direction also exceeded that in the east-west

direction. The sum total of the displacement for all manhole covers is shown in Fig. 5, and photographs of actual manhole cover conditions are shown in Photo 3.

Table 2 Damage to cast iron manhole covers

Surveyed

locations

Offset

locations

Offset occur- rence ratio

Total offset (mm)

Areas with

liquefaction 13 10 77% 500

Areas

without

liquefaction

9 1 11% 50

N

A

Unit: mm

235

Fig. 5 Diagram showing the sum total of

cast iron manhole cover offset

d) Damage to shield tunnels

Of the seven locations surveyed, shield tunnels were found to have protruded about 1 to 2 cm into

the shaft portals along the direction of the tunnel

axis in three locations. However, shield tunnels

sustained no damage whatsoever in general tunnel

portions. (Photo 4)

(3) Investigation of Open-Cut Tunnel

Floating Caused by Liquefaction

As previously mentioned, tunnel damage resulting

from the Hanshin-Awaji earthquake varied widely

between areas where liquefaction did or did not

occur. In addition, when the results of level measurements inside tunnels carried out both before

and after the earthquake were compared, floating

was found to have occurred for open-cut tunnels in areas where liquefaction occurred.

a) Results of level measurements inside tunnels

Fig. 21 shows the difference in levels before and

after the earthquake as measured longitudinally

inside tunnels in areas where liquefaction occurred.

These levels were measured using the survey

points of buildings containing switchboards to which the tunnels are connected as fixed points.

These results show that while tunnels in these areas experienced floating of up to 9 cm, there was next

to no floating in areas where liquefaction did not

occur.

b) Soil conditions

© Soil data

Fig. 22 shows a bar graph for soil in areas where

liquefaction occurred. These areas have sandy soil

with an N value of 15 or less up to GL -7.5 m, and

have groundwater at roughly GL -2.0 m in contrast

to a tunnel overlying layer at GL -2. 1 m.

Fig. 23 shows a bar graph for soil in areas where

liquefaction did not occur. These areas are comprised of gravel with an N value of 25 or more

below GL -5.5 m, and have groundwater at roughly GL -2.0 m in contrast to a tunnel overlying layer at GL -1.0 m. ® Liquefaction judgments

Calculating the liquefaction coefficient F( using

the liquefaction judgment formula of the Japan

common conduit design index produces the values

shown in Table 3. These calculations were performed using the two values of 600 gal and 800 gal as the input acceleration. When Ff? is 1 or less,

liquefaction occurs. F£ = R/L: Liquefaction resistance factor

R: Soil dynamic shearing strength

L = rd-Ks-cv/ov-; Soil dynamic shearing stress Ks: Horizontal seismic coefficient

ov: Total surcharge pressure

or: Effective overburden pressure

138
Table 3 Liquefaction judgment calculation results

Soil data for areas with liquefaction

No.

(Depth)

Ff (600 gal) Ff (800 gal)

1 — —

2 0.313 0.235

3 0.252 0.204

4 0.285 0.230

5 0.273 0.220

6 0.326 0.263

7 0.458 0.294

8 0.017 0.270

9 — —

10 — —

11 0.329 0.266

12 0.294 0.238

13

14

15 0.101 0.348

16

17 0.498 0.402

18 0.001 0.279

1 data for areas without liquefaction

No.

(Depth)

Ft (600 gal) Ft (800 gal)

1 1.803 1.420

2 1.255 0.980

3

4 2.087 1.690

5 2.859 2.310

6 0.909 0.730

7 1.095 0.890

8 2.380 1.930

9 0.428 0.350

10 1.305 1.060

11 0.984 0.800

12 0.740 0.590

13 0.545 0.440

14 1.117 0.900

15 1.630 1.320

16 1.751 1.410

17 1.713 1.380

The calculation results show that even in areas without liquefaction, liquefaction occurs when the

actual seismic acceleration is substituted into the

liquefaction judgment formula.

(D Floating judgment

Open-cut tunnel floating was judged based on the

floating judgment formula of the Japan common

conduit design index, and floating calculations were performed using the two input accelerations

of 600 and 800 gal. The floating judgment formula

is as follows.

Ws + WB + Qs + QB

Fs = Us + UD

Soil surcharge (including the water weight)

Tunnel weight

Soil shearing resistance

Tunnel side friction resistance

Uplift produced by hydrostatic pressure

acting on the tunnel bottom

UD: Uplift produced by excess porewater

pressure acting on the tunnel bottom

Ground surface

Ws:

WB

Qs:

QB:

Us:

//AW//W

Fig. 6 Outline drawing for open-cut tunnel floating judgment

Note that the floating safety factor Fs must be 1 .

1

or more. Also, shear resistance and tunnel friction

resistance are not considered when the liquefaction

resistance factor (FL) is 1.0 or less. The calculation results are shown on the next page.

139
I) Open-cut tunnel floating judgment in areas where liquefaction occurred

600 gal 800 gal

Soil surcharge

(including the water weight) (tf/m) 8.32 8.32

Tunnel weight (tf/m) 4.63 4.63

Soil shearing resistance (tf/m) 0 0

Tunnel side friction resistance (tf/m) 0 0

Uplift produced by hydrostatic pressure (tf/m)

5.46 5.46

Uplift produced by the excess porewater pressure (tf/m) 14.03 14.03

Floating safety factor (Fs) 0.664 0.664

The above results indicate that tunnel floating

should occur. The results of actual level measure- ments at these points show that liquefaction caused

floating of 6 cm, thereby validating the judgment

formula.

II) Open-cut tunnel floating judgment in areas where liquefaction did not occur

600 gal 800 gal

Soil surcharge

(including the water weight) (tf/m) 3.98 3.98

Tunnel weight (tf/m) 5.33 5.33

Soil shearing resistance (tf/m) 0.58 0.58

Tunnel side friction resistance (tf/m) 4.86 3.53

Uplift produced by hydrostatic pressure (tf/m) 5.67 4.62

Uplift produced by the excess porewater pressure (tf/m) 0.95 4.25

Floating safety factor (Fs) 2.45 1.51

The above results indicate that tunnel floating

should not occur. The results of actual level measurements at these points show that floating was

not confirmed, thereby validating the judgment

formula.

o

o

G.L Pavement 2.3 tf/m

Crushed stone 2.1 tf/m 3

Soil unit weight

(sandy soil) Y= 1.8 tf/m3

o

o

m

20JL 1700

2100

_2tL£'

o

o

IX) CM

G.L

o o

0|0

n cm

I I Oo

Pavement 2.3 tf/m3

Crushed stone 2.1 tf/m J

Soil unit weight

(sandy soil) Y= 1.8 tf/m3

Zp£_

i .±7_Q_0_

21QQ

Fig. 7 Open-cut tunnel floating judgment (areas Fig. 8 Open-cut tunnel floating judgment (areas

with liquefaction) without liquefaction)

140
0 Relationship between the floating safety factor

and the amount of floating

Fig. 9 shows the relationship between the floating

safety factor and the amount of floating at three

locations in areas where liquefaction occurred and

three locations in areas where liquefaction did not

occur. These figures clearly show that the amount

of floating increases as the safety factor becomes

smaller.

(mm)

60

.£so

reO

!= 40 ©

3

O

£20

<

10

Horizontal acceleration = 600 gal

1 i

i

* •—

1

i

1 !

\

i

I

i

i

i

!

i I il

[

i

i

!

= 1

1 i \

i

|

i

' \

! \

i

•

j • • 1

0.6 0.8 1.0 1.2 1.4 1.6 1.8 2.0 2.2 2.4 2.6

Safety factor (Fs)

(mm) Horizontal acceleration = 800 gal

60

so

=S «o o

,-30

I 20 ©

< 10

1

1—1—

r

i 1 •

i iAm

i \

i i •

0.2 0.4 0.6 0.8 1.0 1.2 1.4 1.6

Safety factor (Fs)

Fig. 9 Relationship between the floating safety

factor and the amount of floating

(D Examination of floating

This section examines the section changing points

where floating was particularly noticeable. A typical section changing point where tunnels branch is shown in Fig. 10, and the corresponding outline

drawing is shown in Fig. 1 1

.

G.L

Fig. 10 Tunnel branch structure //M/

\

Ht f

Fig. 1 1 Outline drawing of a tunnel branch

At this point, the liquefaction safety factor is calculated using the relationship between HO and

Ht in order to grasp the relationship between the

shape of section changing points and floating

caused by liquefaction.

These calculations assume that

Ht = HO x (1 + 0.2t) with t = 1 to 8 In other words,

HO = 2.0 m

HI = 2.4 m, H2 = 2.8 m, H8 = 5.2 m

The installation conditions were set as indicated

below and the safety factor Fs was obtained for

each value of t. The results showing the relationship between Ht/HO and Fs appear in Fig. 12. These results indicate that the effects of floating

caused by liquefaction increase in tandem with the

degree of change between the shapes of the changing point and general portions.

Conditions

Overlying layer 2.7 m

Groundwater level -2.0 m

Tunnel internal height 2.0 m

Tunnel internal width 1.7 m

Wall thickness Lower floor 30 cm

Upper floor and

side walls 20 cm

141
a 0.70

0.55

1 1.2 1.4 1.6 1.8 2.0 2.2 2.4 2.6 (2.0) (2.4) (2.8) (3.2) (3.6) (4.0) (4.4) (4.8) (5.2)

H,/Ho

(H.)

Fig. 12 Relationship between the floating safety

factor and Ht/HO

C.L /my//

r

7r,Shear

plane

1=2. 5m

Fig. 13 Shearing force acting on joints

(4) Summary of Damage

The current survey results indicate that the probability of tunnel damage is higher in areas with

liquefaction than in areas without liquefaction, and

that even in areas with liquefaction this probability

increases in locations where the tunnel cross section changes dramatically. This is thought to be

caused by lateral flowing and floating produced by

liquefaction. However, the maximum displacement

was between 10 and 20 cm, and did not result in

serious damage other than damage to expansion

joints, etc.

Next, the shearing force Tt (force per meter in the

lengthwise direction of the tunnel) acting on the

joints was obtained using the following equation.

Tt = (Ust + UDt) - (Ws + WBt)

= 4.825 + 1.12t

Assuming the length of the section affected by

buoyancy to be €, the joints break under the

following conditions.

Tt x t > xj

Tj: Joint shearing strength

Here, the required joint shearing strengths assuming the affected section length i = 2.5 m are shown

in Table 4.

Table 4

t Ht (m) Required shearing strength (tf)

2 2.8 17.7

4 3.6 23.3

6 4.4 28.9

8 5.2 34.5

4. EARTHQUAKE-PROOFING

MEASURES

NTT has approximately 520 km of existing tunnels

throughout Japan which house a vast number of

communication cables. The Tokyo metropolitan

area in particular contains a network comprising

approximately 230 km of tunnels. In order to implement earthquake-proofing measures which

reflect the lessons learned from the Hanshin-Awaji

Earthquake, these improvements must be carried

out efficiently by accurately grasping the weak

points of this network and establishing an appropriate order of priority. In particular, appropriate

work methods must be selected when implementing earthquake-proofing measures for tunnels

which contain large numbers of communication

cables (Photo 5).

Upon analyzing the damage from the HanshinAwaji Earthquake, NTT decided to concentrate on

tunnels located in areas where liquefaction is most

likely to occur. Within these areas, the highest

priority has been placed on implementing

earthquake-proofing measures for joints where

tunnels are attached to buildings and other areas

142
where the tunnel cross section changes dramatically. Conventional liquefaction countermeasures

include methods of preventing both liquefaction

and deformation caused by liquefaction. However,

when implementing countermeasures for existing

tunnels in urban areas, these methods are limited

by traffic restrictions, effects on the surrounding

environment, work periods and construction costs,

etc. Therefore, NTT has selected the safest and

most economical work methods which can be

carried out from inside existing tunnels and which

keep the relaying of cables, etc. to a minimum.

(1) Introduction of Flexible Joints

The data detailing the damage in Kobe indicated a maximum floating caused by liquefaction of about

10 cm, and a maximum expansion joint displacement of 1 8 cm. In consideration of these figures, it was determined that fluctuations caused by earthquake motion and liquefaction could be handled by

flexible joints. Therefore, the decision was made to

install flexible joints in building and shaft attachments in areas where liquefaction is considered

likely to occur. In addition, a work method was

developed allowing these joints to be easily installed without having to relay the cables even in existing tunnels which house large numbers of

cables. The installation diagram for flexible joints

is shown in Fig. 14.

i

Fig. 14 Flexible joint installation diagram

(2) Experimental Verification

Structures which afforded the greatest possible

placement interval between flexible joint attachment bolts were investigated in consideration of

work inside existing tunnels which house large

numbers of communication cables. These joint

structures were then verified to ensure that they were capable of dealing with vertical and lateral

offsets of 20 cm and remained water resistant at a water pressure of 1 kg/cm2

. Photograph 6 shows the experimental conditions.

The bolt placement interval for conventional work

methods is 20 cm. However, these experiments

confirmed that water resistance could be maintained even when the placement interval was

increased by up to four times by redesigning the cross sectional shape of the fastener plates, etc.

5. CONCLUSION

NTT has linked buildings containing major switchboards in large cities with tunnels in order to ensure communications capabilities in the event of

earthquakes or other disasters. The high reliability

of these communication tunnel facilities with re- spect to large-scale earthquakes was verified by the

Hanshin-Awaji Earthquake.

However, in light of the fact that further improvement in the reliability of communication infra- structure facilities will be required as we move

toward an advanced information society in the

future, countermeasures must be devised to counteract joint offsetting and other damage to existing

open-cut tunnels caused by liquefaction, etc. Therefore, using the current earthquake damage

data, the accuracy of predictions as to which areas

are likely to suffer from liquefaction must be

further improved, and damage simulations con- ducted for various floating and offsetting amounts

in order to investigate efficient countermeasures.

[Bibliography]

1. Japan Common Conduit Design Index,

6.3 Liquefaction Judgments and Countermeasures for Surrounding Ground

2. Civil Engineering Handbook

3. The Japan Port and Harbor Association, Technical Standards for Port and Harbor Facilities

with Commentary

143
NTT building

Ventilation opening

iruge pit|

Open-cut tunnel

Entrance 16=-

Ventilation Entrance opening

mm

"2

^

Primary lining

Secondary lining

Highlyreliable Manhole

conduits

Pipeline Hand hole

Manhole

Shield tunnel

Fig. 15 Outline drawing of communication civil facilities

Collar Socket

<>

Waterstop

600

Waterstop

3^

Waterstop

2C on n

Unit: mm

Collar joint Socket joint Expansion joint detail drawing

Fig. 16 Expansion joint structures

Re-bars

inside

the shaft

a: 3.0m b: 4.0m c: 6.0m

D22 re-bars x 30 D22 re-bars x 15

a-a cross section b-b cross section c-c cross section

Fig. 17 Earthquake-proofing re-bars for shield tunnels

144
I Tunnels

Fig. 18 Communication tunnels inside Kobe City Fukiai

Fig. 19 Expansion joint damage

Fig. 20 Cast iron manhole cover structure

145
400 300 200 100 0

Distance from the Minato building (m)

Fig. 2 1 Open-cut tunnel floating caused by liquefaction

No.2 N value

11 3* f

Fig. 22 Soil survey data (areas with liquefaction)

No.4 N value

Fig. 23 Soil survey data (areas without liquefaction)

146
Photo 1 Expansion joint structure damage (offsetting)

Photo 2 Expansion joint structure damage

(With the concrete is removed, the waterstops are revealed.)

Photo 3 Cast iron manhole cover damage (offsetting)

147
Photo 4 Shield tunnel shaft attachment damage

(Tunnel has protruded approximately 1 cm into the interior of the shaft.)

Photo 6 Flexible joint experiment for open-cut runnels

148
EFFECT OF LOAD PATH ON DAMAGE TO CONCRETE BRIDGE PIERS

by

Andrew W. Taylor 1

, Ashraf El-Bahy2, William Stone

1 and Sashi Kunnath2

ABSTRACT

In earthquake engineering studies of

reinforced concrete (RC) bridge columns, a

controlled, cyclic lateral load pattern with

gradually increasing amplitude has

traditionally been applied to laboratory test specimens. However, in actual earthquakes

bridge columns are exposed to random cyclic

lateral loading patterns, which are much

different from typical laboratory loading

patterns. Current American Association of

State Highway and Transportation Officials

(AASHTO) and California Department of

Transportation (Caltrans) design provisions are based almost entirely on tests in which

traditional, controlled laboratory loading

patterns have been applied. The differences in

the effects of these types of loading have never been explored systematically. In this study

both types of loading (controlled, cyclic lateral

loads, and random earthquake type loads)

were applied to a series of twelve nominally

identical, one-fourth scale circular, cantilever

columns, and the differences in observed

damage were studied. In this paper the

experimental results are briefly summarized,

and preliminary findings are discussed.

KEYWORDS: bridge columns; building

technology; damage modeling; earthquake

engineering; laboratory testing; random

loading; reinforced concrete

1. INTRODUCTION

The purpose of this study is to explore both

experimentally and analytically the

relationship between load path and the

inelastic damage induced in reinforced

concrete bridge columns. Here "load path" is

meant to describe the series of lateral

displacements applied at the top of the

column. If cyclic lateral loading causes

inelastic behavior of a bridge column, the type

and extent of damage induced in the column

will depend on the order in which the

displacements are applied. That is, damage

induced by inelastic deformations is pathdependent. In this study a variety of load

paths were applied to reinforced concrete

bridge columns to investigate the nature of the

dependency of damage on load path.

Over the last 25 years, in laboratories around

the world, a large number of tests have been

performed in which cyclic lateral

displacements were applied to the top of

cantilever reinforced concrete columns.

Although the purpose of these tests was to

study the seismic response of the columns, the

displacement pattern applied to the columns in

most cases did not resemble the displacements

induced by an earthquake. The displacement

patterns were usually in the form of a sawtooth

wave, often with gradually increasing

amplitude, as illustrated in Figure 1. However,

earthquakes impart displacement patterns with

random amplitudes, as shown in Figure 2. The

seismic design provisions of the American

Association of State Highway and

Transportation Officials (AASHTO) and of the

California Department of Transportation

(Caltrans) are based almost entirely on tests in

which the applied displacement pattern had a form similar to that shown in Figure 1. The

tests conducted in this study are intended to

investigate the relationship between regular,

sawtooth displacement patterns widely applied

in the laboratory (Figure 1), and the random

displacement patterns induced by earthquakes

(Figure 2).

Structures Division, Building and Fire

Research Laboratory, National Institute of

Standards and Technology, Gaithersburg,

Maryland, USA, 20899

2Department of Civil Engineering, University

of Central Florida, Orlando, Florida, USA,

32816

149
The experimental portion of this study

consisted of laboratory tests of 12 nominally

identical bridge columns which were subjected

to a variety of cyclic lateral displacement

patterns. These displacement patterns were in

the form of sawtooth waves of constant

amplitude, sawtooth waves of gradually

increasing amplitude, and random

displacement patterns which simulated

earthquakes of various intensities.

Observations were made of the progress of

damage, both visually and with

instrumentation. When a specimen was

subjected to earthquake-type loads, several

simulated earthquakes were applied

sequentially to study the effects of multiple

earthquakes on the same bridge column

(Figure 2). The earthquake displacement

patterns applied to the columns were

determined from prior nonlinear dynamic

analyses of the columns, using recorded

earthquake accelerograms as input to the

analysis. This process is explained further in

Section 3.

The analytical portion of the study includes

the formulation and verification of damage

model hypotheses. The key feature sought in a damage model is that it apply equally to damage incurred either from a sawtooth wave

displacement pattern, or a random

displacement pattern. The data gathered from

the 12 specimens in this study enable the

testing of damage model hypotheses against

both types of displacement histories.

In this paper the results of the experimental

program are briefly summarized, and

observations relevant to damage modeling are discussed. Since analysis of the laboratory

data is still ongoing, preliminary findings are presented here. Detailed descriptions of both

the experiments and analytical study will be

presented in a forthcoming report.

2. EXPERIMENTAL PROGRAM

Figure 3 shows a test specimen and the loading

apparatus. The specimens had circular cross sections and were nominally identical. The

scale of the specimens was 1:4, and the full

scale pier on which the design was based

conformed to California Department of

Transportation (Caltrans) specifications

(Caltrans, 1992). The specimens were cast in an inverted position with an integral base

block. Figure 4 shows the arrangement of

reinforcement in the specimen and base block.

The column confining reinforcement was a continuous smooth spiral wire, which extended

through the depth of the specimen base block.

Longitudinal reinforcement consisted of 21

deformed bars distributed around the

perimeter of the spiral, extending into the base

block and terminating in hooks. The

measured strength of the concrete (using 150

mm by 300 mm cylinders) was 38 MPa. The

maximum concrete aggregate size was 13 mm.

The yield strength of the longitudinal

reinforcement was 414 MPa, and the yield

strength of the spiral reinforcement was 427

MPa.

As shown in Figure 3, additional re-usable

base block sections were joined to the

specimen base block with post tensioning rods,

and the entire base block assembly was posttensioned to the laboratory strong floor,

forming a rigid base. A constant vertical load

of 10 percent of the axial capacity of the gross

column section was applied through a 200 kN

capacity vertical servo-hydraulic actuator. A

670 kN capacity actuator was used to apply

cyclic lateral displacements. This second

actuator was controlled by a computer and

servo-hydraulic system, which were

programmed to apply specific lateral

displacement patterns to the top of the

column. The measured lateral load was

corrected to account for the lateral component

of the vertical actuator. A system of

horizontal guide rails and ball bearing rollers,

located at the top of the column, maintained

the in-plane alignment of the column during

testing. These rails and rollers are not shown

in Figure 3 for clarity. Instrumentation

consisted of load cells and displacement

transducers (LVDT's) for each hydraulic

actuator, an array of horizontal LVDT's to

measure lateral displacement over the height

of the specimen, strain gages on selected

longitudinal reinforcing bars near the base of

the column, clip-on external strain gages to

measure curvature near the base of the

column, and inclinometers to measure rotation near the base of the column.

3. DISPLACEMENT HISTORIES

Two types of displacement histories were

applied to the specimens in this study: regular,

controlled 2-D sawtooth patterns, which have

historically been applied in most laboratory

tests; and random 2-D patterns, which simulate

earthquake loading. Both types of

150
displacement histories were applied slowly, in a pseudo-dynamic mode.

The load paths of the first six specimens are

illustrated in Figure 5. Specimen 1 was

subjected to monotonically increasing lateral

displacement to failure. Since all specimens

are nominally identical, this test established

basic parameters, such as initial lateral stiffness,

displacement at first yield, and ultimate lateral

strength, which were used in planning

subsequent tests. Specimen 2 was subjected to a "standard" laboratory displacement history,

detailed in Figure 1. While there is in fact no

commonly accepted "standard" displacement

history, many researchers have used a displacement pattern similar to that shown in

Figure 1. Several cycles are applied initially at a displacement Ay, which causes first yielding

of the column longitudinal reinforcement.

These cycles are followed by a single cycle of

small amplitude which is used to measure the

change in initial tangent stiffness of the

column. As shown in Figure 1, subsequent

groups of cycles are applied with increasing

amplitudes. This continues until substantial

degradation of the lateral load carrying

capacity of the column is observed, usually

defined as the point at which the maximum

lateral load in a cycle is only 75 percent to 80

percent of the peak lateral load capacity

measured during the test. Specimens 3 to 6

were subjected to constant-amplitude cycles

(of ±2Ay, ±3Ay, ±4Ay, and ±5Ay, respectively)

until severe deterioration of the column was

observed. These tests provided data for

calibration of analytical damage models which

reflect damage accumulated during constant

amplitude cycling.

A typical displacement history for Specimens

7 to 12 is shown in Figure 2. The details of

the displacement histories for these specimens

are shown in Table 1 . These last six specimens

were used to study the effects of real

earthquake displacement histories on

accumulation of damage, particularly over a

series of events of varying intensity. The

displacement patterns applied to Specimens 7

to 12 were derived from measured earthquake

acceleration records, using the following

process. Given an earthquake acceleration

record, the deformation history at the top of

the column was computed using the inelastic

dynamic analysis program IDARC (Kunnath

et al. 1992). IDARC computes the response of

the column by taking into account the

degradation in stiffness, lateral strength and

energy dissipation capacity (pinching of the

hysteresis loops) which occur as inelastic

deformations progress over the course of the

earthquake. The degree of stiffness

degradation, strength degradation and

pinching are controlled by three parameters a,

P, and Y- These parameters are determined a

priori, either empirically, based on past

experience with tests on similar members

(Kunnath et al. 1992), or analytically using

correlation rules which relate a, P and y to the

geometric and material properties of the

column cross section (Stone and Taylor, 1991,

1993).

The earthquake records used in tests 7 to 12

were selected from a pool of available strong

ground motion records recorded at bedrock

sites in the state of California, and one record

from Mexico City. Each specimen was

subjected to a series of up to five earthquakes.

Table 1 shows that earthquakes of a range of

intensities were applied, and the order in which

they were applied was varied. The pool of

available earthquake records was subdivided

into two main categories: minor events and

major events. Minor events were intended to simulate either isolated low-intensity

earthquakes, or low intensity aftershocks

following a major earthquake. Minor events

would cause little or no damage to the

specimens in this test program, as determined

from the results of the constant amplitude tests (Specimens 3 to 6). Based on observations

from these tests, minor events were defined as those which caused three to five excursions

into the inelastic range, with ductility demands

no greater than 3Ay. Major events were

divided into two types: damaging earthquakes

and maximum credible events. Again, based

on the results of the earlier constant-amplitude

tests, damaging events were defined as those

which caused multiple excursions into the

inelastic range, with ductility demands greater

than 3Ay, but generally less than 5Ay. The

earlier tests showed that this level of ductility

demand caused deterioration of the column,

but that the column retained much of its

strength and stiffness. Maximum credible

events were those which caused cycles with

ductility demands greater than 5Ay. The

earlier tests also showed that just a few

excursions greater than 5Ay resulted in rapid

deterioration of strength and stiffness. In

some cases, where acceleration records were

not available which met the desired criteria, the

records were modified by simple amplitude

151
scaling. These scale factors are reported in

Table 1.

4. OBSERVATIONS AND DAMAGE

MODELING

Preliminary observations from the

experimental program are presented here.

The experimental data is currently being

analyzed, and detailed descriptions of the

experimental results will be presented in a forthcoming report. The experiments are discussed in terms of two categories: the

constant amplitude tests and the random

amplitude (earthquake loading) tests.

4.1 Constant Amplitude Tests

It was observed that repeated cycling of

Specimen 3 at a displacement amplitude of

±2 Ay caused almost no degradation of

stiffness and strength of the column. After the

first full cycle at ±2Ay (during which initial

yielding and cracking occurred) the hysteresis

loops remained extremely stable, lying nearly on top of one another for 150 cycles. The test was stopped at 150 cycles not because the

column failed, but because it was believed that

no further useful information could be

obtained by continuing the test. At the other

extreme, Specimen 6 was subjected to cycles

of +5Ay and exhibited a rapid decrease in

strength and stiffness, nearly completely losing

lateral stiffness and load capacity after only 5

cycles. This rapid deterioration is illustrated in

Figure 6. Specimen 5, which was cycled at ±4Ay, exhibited a gradual decrease in strength

and stiffness as cycling progressed (Figure 7),

but the decrease was not nearly as rapid as for

Specimen 6. This is illustrated in Figure 7.

Thus, for the specimens tested in this program

there appears to be a threshold ductility level,

of about 4Ay, above which deterioration is rapid and severe. It could be surmised that

earthquakes which induce displacements of

less than 4Ay in the columns tested in this

program would cause much less damage than

those which induce displacements greater than

4Ay. Indeed, subsequent testing under

earthquake displacement patterns (Specimens

7 to 12) confirmed this observation: a series of

several minor events, which caused few

excursions greater than 2Ay, would result in

very little damage to a column, while a single

earthquake with a few excursions greater than

4Ay would result in rapid deterioration of

strength and stiffness. The apparent threshold

of 4Ay applies only to the columns tested in

this program. It could be expected that

similarlydesigned circular columns, with high

ratios of confining reinforcement (such as those designed under the Caltrans

specifications), and which were dominated by

flexural rather than shearing deformations,

would also exhibit such a threshold, but the

level of the threshold would not necessarily be

4Ay.

Another significant observation from the

constant amplitude tests was that cumulative

dissipated energy (the area contained within

the hysteresis loops) is not good predictor of

column failure. It was found that the

cumulative dissipated energy at failure

depended on the amplitude of the sawtooth

wave. Figure 8 shows the accumulation of

energy to failure for Specimens 2, 4, 5 and 6

("standard" displacement pattern, ±3Ay, ±4Ay,

and ±5Ay, respectively). Results for Specimen

3 are not plotted because the specimen did not

deteriorate significantly under 150 cycles at ±2Ay, so a failure state was not achieved. In

Figure 8 it can be seen that dissipated energy

at failure is strongly dependent on the

displacement history. Therefore, because of

the highly variable nature of earthquakeinduced displacement histories, it is apparent

that cumulative energy alone is not an

acceptable measure of column damage.

4.2 Earthquake Loading Tests

A typical lateral load-displacement plot for

one of the earthquake loading tests is shown in

Figure 9. The two main observations made in

these tests are described below.

First, it was observed that "minor" events

(Table 1) had little or no effect on the ultimate

failure of a column. That is, it made no

difference if minor events occurred before or

after a damaging event: the level of damage

following the damaging event was essentially

the same in both cases. This has important

implications for the management of

earthquake damage to bridge columns. It would appear that minor seismic events should

be considered to have little effect on wellconfined, circular, flexure-dominated bridge

columns. It remains for transportation

authorities to define precisely what a "minor"

event is, but it appears acceptable to not

discount the strength and stiffness of wellconfined columns which have been subjected

to a minor earthquake.

152
Second, it was observed that failure of the

columns could be classified into two general

types: failure due to low cycle fatigue of the

longitudinal reinforcement, called here a "low

cycle fatigue failure"; and failure due to

rupture of confining reinforcement, or "confinement failure." Confinement failures

were observed to occur more frequently than

low cycle fatigue failures. While these two

classes of failure have been observed and

reported by others, it is interesting to recall

that the only variable in these tests was the

displacement history. Thus displacement

history, rather than the column configuration,

determined the failure mode. This illustrates

the importance of conducting nonlinear

dynamic analyses of bridge columns whenever

feasible, and considering a range of possible

ground motions at the bridge site, rather than

permitting a single earthquake record to

determine the design.

With regard to analytical damage modeling,

preliminary analyses of the test results indicate

that cumulative fatigue models, based on

variations of the classic Miner's rule (Miner

1945) for metals, can successfully track

damage in columns which fail due to low cycle

fatigue of the longitudinal reinforcement.

However, this same approach does not apply

well to columns which exhibit confinement

failure. Other methods of modeling

confinement failure are currently being

investigated. Furthermore, the columns tested

in this study were dominated by flexural

behavior, and none of the findings stated here

necessarily apply to short, shear-dominated

columns. A similar experimental investigation

of shear-dominated columns is planned.

5. CONCLUSIONS

The main findings of this study to date are summarized below. Since interpretation of the

test data is still ongoing, these findings must be

considered preliminary.

1) There appears to be a threshold ductility

level for well-confined, circular, flexuredominated columns above which degradation

of stiffness and strength is rapid. For the

columns tested in this study this threshold

ductility level was about 4Ay. Below this

threshold degradation was more gradual, and

for ductility demands of 2Ay or less

degradation was minimal.

2) The constant amplitude tests in this study

confirm that cumulative dissipated energy at failure is strongly path dependent. Therefore

cumulative energy by itself is not a reliable

measure of column damage.

3) Minor earthquakes appear to have very little influence on the damage exhibited by wellconfined, circular, flexure-dominated bridge

columns in subsequent major events. While

the precise definition of a "minor" event is open to interpretation, this study indicates that

the strength and stiffness of a well-confined,

circular, flexure-dominated bridge column

should not be discounted because the column

has been subjected to minor events.

4) The random loading tests indicate that

failure mode is path dependent. Nominally

identical columns failed either due to lowcycle fatigue of longitudinal reinforcement, or due to rupture of confining reinforcement.

The only variable between tests was the load

path. The relationship between load path and

failure mode is currently under study.

5) Cumulative fatigue methods, based on

Miner's rule for fatigue of metals, appear to

predict well the progress of damage in

columns which failed due to low cycle fatigue

of longitudinal bars. However, Miner's rule

does not apply to those columns which failed

due to rupture of the confining reinforcement.

Other damage models are being investigated

which may apply to confinement failure.

Furthermore, the findings of this study apply

only to well-confined, circular, flexuredominated columns. A future study is

planned to investigate well-confined, circular,

shear-dominated columns.

6. ACKNOWLEDGMENTS

The authors gratefully acknowledge the

financial support of the National Center for

Earthquake Engineering Research (NCEER),

the California Department of Transportation

(Caltrans), and the Federal Highway

Administration (FHWA).

153
7. REFERENCES

Caltrans (1992), "Bridge Design

Specifications," State of California, Department

of Transportation (Caltrans), Office of

Structure Design, Sacramento, California,

August 1986 (revisions to December 1992).

Kunnath, S.K., Reinhorn, A.M., and Lobo, R.F.

(1992), "IDARC Version 3.0, A Program for

the Inelastic Damage Analysis of Reinforced

Concrete Structures," Report NCEER-92-0022,

National Center for Earthquake Engineering

Research, State University of New York at Buffalo, August 1992.

Miner, M.A. (1945), "Cumulative Damage in

Fatigue," Journal of Applied Mechanics, Vol.

12, 1945, p. A-159.

Stone, W.C., and Taylor, A.W. (1991), "System

Identification of Spirally Reinforced Circular

Bridge Columns Subjected to Cyclic Lateral

Loading," Wind and Seismic Effects;

Proceedings, 23rd Joint Meeting UJNR, NIST

SP 820, National Institute of Standards and

Technology, Gaithersburg, Maryland,

September 1991, pp 269-293.

Stone, W.C., and Taylor, A.W. (1993), "Seismic

Performance of Circular Bridge Columns

Designed in Accordance with

AASHTO/CALTRANS Standards," BSS-170,

National Institute of Standards and

Technology, Gaithersburg, Maryland,

February 1993.

Table 1: Earthquake records applied to Specimens 7 to 12

Spec.

No.

Event

No.

Description:

Event Severity

and Number

Purpose

of Applying Event

to Column

Earthquake Record

Scale Factor

Scaled

PGA

g's

1 Damaging No. 1 Significant damage Loma Prieta 1989, Presidio 12.0 1.20

7 2 Minor No. 1 Aftershock Imperial Valley 1979, Superstition Mtn. 1.8 0.34

3 Minor No. 2 Second aftershock San Fernando 1971, 2011 Zonal Avenue 1.2 0.10

4 Damaging No. 2 Failure of column San Fernando 1971, 455 S. Figueroa St. 3.6 0.54

1 Minor No. 1 Minor damage Imperial Valley 1979, Superstition Mtn. 1.8 0.34

8 2 Minor No. 2 Minor damage San Fernando 1971, 2011 Zonal Avenue 1.2 0.10

3 Damaging No. 1 Significant damage Loma Prieta 1989, Presidio 12.0 1.20

4 Damaging No. 2 Failure of column San Fernando 1971, 455 S. Figueroa St. 3.6 0.54

1 Damaging No. 3 Significant damage San Fernando 1971, Orion Boulevard 3.25 1.43

2 Minor No. 2 Aftershock San Fernando 1971, 2011 Zonal Avenue 1.2 0.10

9 3 Damaging No. 4 Moderate damage El Centra 1940 1.0 0.35

4 Minor No. 3 Aftershock San Fernando 1971, 455 S. Figueroa St. 1.0 0.15

5 Damaging No. 3 Failure of column San Fernando 1971, Orion Boulevard 3.25 1.43

1 Minor No. 2 Minor damage San Fernando 1971, 2011 Zonal Avenue 1.2 0.10

2 Damaging No. 4 Moderate damage El Centra 1940 1.0 0.35

10 3 Minor No. 3 Aftershock San Fernando 1971, 455 S. Figueroa St. 1.0 0.15

4 Damaging No. 3 Significant damage San Fernando 1971, Orion Boulevard 3.25 1.43

5 Damaging No. 3 Failure of column San Fernando 1971, Orion Boulevard 3.25 1.43

1 Damaging No. 4 Significant damage Northridge 1994, VA Hospital 1.0 0.42

1

1

2 Minor No. 4 Aftershock Northridge 1994, Griffith Observatory 1.0 0.26

3 Minor No. 5 Minor damage Taft 1952 1.0 0.36

4 Damaging No. 5 Failure of column Mexico City 1985, SCT 1.0 0.17

1 Minor No. 4 Minor damage Northridge 1994, Griffith Observatory 1.0 0.26

! 12 2 Minor No. 5 Minor damage Taft 1952 1.0 0.36

3 Damaging No. 4 Significant damage Northridge 1994, VA Hospital 1.0 0.42

4 Damaging No. 5 Failure of column Mexico City 1985, SCT 1.0 0.17

154
Figure 1: Typical laboratory displacement history, applied to Specimen 2

i i i i i i i i i i i i i i I i i i i I i i i i I i i i i 0 50 100 150 200 250 300

Half Cycles

Figure 2: Typical earthquake displacement histories, applied to Specimen 12

155
REACTION FRAME

« l

LVDTs

200 kN VERTICAL

ACTUATOR

HEAD CAP

CLIP

GAGES

+/- 150 mm STROKE

ACTUATOR

-305 mm 670 kN CAPACITY

INCLINOMETERS

.STRAIN

1 / GAGES

CM

lO

t--

END BLOCKS SPECIMEN

Figure 3: Load frame and instrumentation for cyclic lateral load tests

Spiral 4.1mm Dia. 6 19mm Clear

Spacing

21 # 3 (9.5mm Dia.)

Longitudinal bars

9 § 5 (15.9mm Dia.)

-U Shape

-10J 5 (15.9mm Dia.)

-2 | 3 (9.5mm Dia.)

Hoops

5

16 f 5 (19mm Dia.)

Hoops

-2 # 3 (9.5mm Dia.)

Hoops

Elevation Side View

Figure 4: Reinforcement details for column specimen and base block

156
No. Load

Regimen

X-Direction Load

History

Y-Dlrection Load

History

Plan view of

Load Path

1 Monotonic

"Pushover" Test

y

W x

2 "Standard"

displacement

pattern vvAAA/

y v v v vy

3

2Ay

y V V V V ^^^^^ A

4

3Ay

y

5

4Ay WAV y

6

5Ay

y WvW

Figure 5: Displacement histories of Specimens 1 to 6

Lateral Displacement, mm Lateral Displacement, mm

Figure 6: Lateral load-displacement plot, Spec. 6 Figure 7: Lateral load-displacement plot, Spec. 5

157
I

Number of Half Cycles

Figure 8: Cumulative energy to failure for Specimens 2, 4, 5 and 6

-120 -80 -40 0 40 80 120

Lateral displacement, mm

Figure 9: Lateral load-displacement plot for Specimen 12

158
Guide Specifications for Reconstruction and Repair of Highway Bridges

Which Suffered Damage due to the Hyogo-ken Nanbu Earthquake

by

Kazuhiko KAWASHIMA ° , Kazuhiro NISHIKAWA 2 )

, Masanori NAKANO 3 5

Shigeki UNJOH A)

, Yoshitomi KIMURA 5) and Jun-ichi HOSHIKUMA 65

ABSTRACT

The Hanshin/Awaji Earthquake (Hyogo-ken

Nanbu Earthquake) of January 17, 1995 caused

destructive damage to highway bridges. Obviously this was the first experience to cause destructive damage

to structures since the 1923 Great Kanto Earthquake.

Various tentative measures were taken for seismic

design of reconstruction and repair of highway bridges

which suffered damage due to the earthquake, seismic

design of new bridges, and seismic strengthening of

existing bridges.

This paper summarizes the "Guide Specifications

for Reconstruction and Repair of Highway Bridges

Which Suffered Damage due to the Hyogo-ken Nanbu

Earthquake.

KEYWORDS : Hyogo-ken Nanbu Earthquake, Guide

Specifications. Highway Bridges. Reconstructrion and

Repair

1. INTRODUCTION

Highway bridges in Japan had been considered

safe even against extreme earthquakes such as the

Great Kanto Earthquake (M7.9) in 1923, because

various past bitter experiences have been accumulated

to formulate the seismic design method. Large lateral

force coefficient ranging from 0.2g to 0.3g has been

adopted in the allowable design approach. Various

provisions for preventing damage due to instability of

soils such as soil liquefaction have been used.

Furthermore, design details including the falling-down

prevention devices have been adopted.

In fact, reflecting those provisions, numbers of

highway bridges which caused complete collapse of

the superstructures was only 15 since the 1923 Great

Kanto Earthquake. Based on this evidence, it has been

regarded by the engineers and researchers in bridge

seismic engineering that the seismic damage of

highway bridges had been decreasing in recent year

(3).

The Hanshin/Awaji Earthquake took place at Kobe and Awaji Island on January 17, 1995, exactly one year after the Northridge. California, USA,

Earthquake, and it caused destructive damage to highway bridges. Falling-down and nearly

falling-down of superstructures occurred at 9 sites, and

other destructive damage occurred at 16 sites. Destructive earthquakes had not occurred at urban

areas in recent years, and the earthquake revealed that

there are various critical issues in seismic design and

seismic strengthening of bridges in urban areas. After the earthquake, the "Committee for

Investigation on the Damage of Highway Bridges

Caused by the Hyogo-ken Nanbu Earthquake

(Hanshin/Awaji Earthquake)" was formulated in the

Ministry of Construction to survey the damage and

clarify the factors which contributed to the damage.

The Committee was chaired by T. Iwasaki. Executive

Director, Civil Engineering Research Laboratory.

Members list is presented in an Appendix.

1) Professor, Department of Civil Engineering. Tokyo

Institute of Technology

2) Head, Bridge Division. Structure and Bridge

Department, Public Works Research Institute,

Ministry of Construction

3) Head, Foundation Engineering Division. Structure

and Bridge Department, ditto

4) Senior Research Engineer, Earthquake Engineering

Division. Earthquake Disaster Prevention

Department, ditto

5) Senior Research Engineer, Foundation Engineering

Division, Structure and Bridge Department, ditto

6) Research Engineer. Earthquake Engineering

Division, Earthquake Disaster Prevention

Department, ditto

159
On February 27. 1995. the Committee approved

the Guide Specifications for Reconstruction and

Repair of Highway Bridges Which Suffered Damage

due to the Hyogo-ken Nanbu Earthquake (13). and the

Ministry of Construction noticed on the same day that

the reconstruction and repair of the highway bridges

which suffered damage by the H/A Earthquake shall be made by the Guide Specifications. On March 30,

1995 . the Committee compiled the Interim Report on

the Damage of Highway Bridges by the Hyogo-ken

Nanbu Earthquake.

Although when the Guide Specifications was

made, there was a plan to prepare a tentative seismic

design method which applies to new construction and

seismic strengthening of existing highway bridges

until the formal revision of the Design Specifications

of Highway Bridges (4) is made, it was decided by

the Ministry of Construction on May 25. 1995 that the

Guide Specifications shall be tentatively used as an

emergency measure for seismic design of new

highway bridges and seismic strengthening of existing

highway bridges until the Design Specifications of

Highway Bridges is revised. Tentative measures

provided in the Guide Specifications were considered

appropriate for new construction and seismic

strengthening.

In May. 1995, the Special Sub-Committee for

Seismic Countermeasures for Highway Bridges was

formulated in the Bridge Committee of the Japan

Road Association. The Sub-Committee is chaired by

K. Kawashima, Tokyo Institute of Technology, and its main role is to draft the revision of the Design

Specifications of Highway Bridges. It is planned to complete the revision in 1996 fiscal year. Because

there were various inquiries from field engineers on

the application of the Guide Specifications to new

construction and seismic strengthening, the

Sub-Committee released on June 30. 1995 the

Reference for Applying the Guide Specifications to New Bridges and Seismic Strengthening (5). It included several supplements on the way of

application of the Guide Specifications as well as

several design examples.

In December 1995. the Report on the

Damage of Highway Bridges by the Hyogo-ken

Nanbu Earthquake was issued by the Committee (12).

The Report was essentially the same with the Interim

Report, but statistical data on damage was updated,

and damage of foundations was included.

This paper summarizes die "Guide Specifications

for Reconstruction and Repair of Highway Bridges

Which Suffered Damage due to the Hyogo-ken Nanbu

Earthquake" and the "Reference for Applying the

Guide Specifications to New Highway Bridges and

Seismic Strengthening of Existing Highway Bridges."

2. SEISMIC DESIGN FOR RECONSTRUCTED

BRIDGES AND REPAIR

2.1 General

For seismic design of reconstruction of highway

bridges that suffered damage due to the

Hanshin/Awaji Earthquake, the "Guide Specifications

for Reconstruction and Repair of Highway Bridges

Which Suffered Damage due to the Hyogo-ken Nanbu

Earthquake" (13) was issued by the Ministry of

Construction on February 27. 1995 upon approval by

the "Committee for Investigation on the Damage of

Highway Bridges Caused by the Hyogo-ken Nanbu

Earthquake." The Guide Specifications was applied

only for reconstruction and repair of the highway

bridges that suffered damage due to the

Hanshin/Awaji Earthquake.

The table of contents of the Guide Specifications

for Reconstruction and Repair was as:

1. General

2. Basic Principle of Seismic Design

3. Dynamic Response Analysis

3.1 General

3.2 Analytical Methods and Analytical Models

3.3 Input Ground Motions

4. Menshin Design

5. Check of Dynamic Strength and Ductility of

Reinforced Concrete Piers

5.1 General

5.2 Stress-Strain Relation of Confined

Concrete

5.3 Hoops

5.4 Treatment of Statically Eccentric Bending

Moment

5.5 Treatment of P-Delta Effect

5.6 Tennination of Main Reinforcements at Mid-height

6. Check of Dynamic Strength and Ductility- of

Concrete Infilled Steel Piers

6.1 General

6.2 Evaluation of Dynamic Strength and

Ductility

6.3 Treatment of P-Delta Effect

7. Check of Dynamic Strength and Deformation

Capability of Foundations

160
7.1 General

7.2 Basic Principle for Check of Dynamic

Strength and Deformation Capability

8. Bearing Supports and Surroundings

9. Falling-down Prevention Devices

10. Treatment for Lateral Spreading of Soils due

to Soil Liquefaction

The following outlines the Guide Specifications.

2.2 Basic Principle of Seismic Design

The bridges shall be designed so that they can

withstand with enough structural safety against the

Hanshin/Awaji Earthquake. To achieve this goal,

following basic principles shall be considered.

(1) To increase the ductility of whole bridge

systems, dynamic strength and ductility shall be

assured for whole structural members in which

seismic effect is predominant. Although the check of

dynamic strength and ductility has been adopted for

reinforced concrete piers since 1990, it has not been

applied for other structural members such as steel

piers and foundations.

(2) Seismic safety against the Hanshin/Awaji

Earthquake shall be verified by dynamic response

analysis considering nonlinear behavior of structural

members.

(3) In design of elevated continuous bridges, it is appropriate to adopt the Menshin Design for

distributing lateral force of superstructure to many

substructures. The Menshin Design is close to the

seismic isolation, but the emphasis is placed to increase energy dissipating capability and to distribute

lateral force of deck to substructures (9,13).

(4) Enough tie reinforcements to assure the

ductility' shall be provided in reinforced concrete piers,

and the termination of main reinforcements at mid-height shall not be made.

(5) Concrete shall be filled in steel piers to assure dynamic strength and ductility. Steel piers designed by

the current practice developed local bucking at web

and flange plates although they were stiffened by

longitudinal stiffeners and diaphragms. This tends to cause sudden decrease of bearing capacity in lateral

direction after the peak strength and therefore less

energy dissipation is anticipated. This subsequently

deteriorates the bearing capacity of steel piers in

vertical direction. Because it is now at the stage that

technical developments are being made to avoid such

behavior, it was decided to tentatively use steel piers

with infilled concrete for reconstruction and repair.

(6) Foundations shall be designed so that they

have enough dynamic strength and deformation

capability for lateral force. The dynamic strength and

deformation capability of foundations shall be larger

than the flexural strength and ductility of piers to prevent damage at foundations.

(7) It is suggested to further use rubber bearings

because they absorb relative displacements developed

between a superstructure and substructures. In design

of bearings, correct mechanism of force transfer from

a superstructure to substructures shall be considered.

(8) The devices to prevent falling-down of a

superstructure from substructures shall be designed so

that they can assure falling-down of decks. Attention

shall be paid so as to dissipate energy and to increase

strength and deformation capability.

(9) At those sites where potential to cause lateral spreading associated with soil liquefaction is high, its

effect shall be considered in design. Because technical

information to evaluate earth pressure in laterally

spreading soils is limited, it is important to recognize

that such evidence exists and that countermeasures

shall be taken in any possible ways.

2.3 Dynamic Response Analysis

Besides the dynamic response analysis described

in the "Part V Seismic Design" of the 1990 Design

Specifications of Highway Bridges," dynamic behavior

of bridges shall be carefully clarified by nonlinear

dynamic response analysis, and result of the analysis

shall be reflected in design.

In the dynamic response analysis, the ground

motion that was recorded during the Hanshin/Awaji

Earthquake with the largest intensity of ground

acceleration shall be used as an input to assure the

seismic safety of bridges against the Hanshin/Awaji

Earthquake. The acceleration recorded at JMA Kobe

Observatory as shown in Fig. 2-1 may be regarded as

the input motion. Clarification of appropriate ground

motion needs to be made based on future research. In

Fig. 2-2, response spectra for ground motions at JR

Takatori Station (15) and Higashi Kobe Bridge (Route

5, Bay Shore Line of the Hanshin Expressway) by the

Hanshin/Awaji Earthquake. JMA Kushiro Observatory

by the 1993 Kushiro-oki Earthquake, and Sylmar

Parking Lot by the 1994 Noithridge Earthquake (I)

are presented for comparison.

2.4 Menshin Design

It is appropriate to adopt the Menshin Design to

mitigate shocks associated with deck response and to reduce lateral force by increasing energy dissipation.

161
Emphasis shall be placed to distribute lateral force to

as many substructures as possible, and to increase

energy dissipation capability. Excessive elongation of

namral period to reduce lateral force shall not be

made, because this increases response displacement of

a deck. This causes various disadvantage such as an

increase of noise and vibration associated with the

adoption of large expansion joints. The Menshin

Design shall not be used at the sites where potential

for soil liquefaction and unstability of weak clayey

materials is high. The "Manual for Menshin Design of

Highway Bridges" (8.11) issued in 1992 as an outcome of the Joint Research between the Public

Works Research Institute and private companies may

be refereed to in the Menshin Design.

For introducing the Menshin Design for existing

simply supported bridges, it is suggested to connect

simply supported girders to make the superstructures

continuous. The "Reference for Design and

Construction to Eliminate Expansion Joints from

Existing Bridges" (18) issued by the Road

Maintenance Center in 1995 may be refereed to for

this purpose.

2.5 Check of Dynamic Strength and Ductility of

Reinforced Concrete Piers

The check of dynamic strength and ductility shall be made for reinforced concrete piers after they are designed by the static lateral force method. The

current provisions for the check were expanded in the

Guide Specifications as:

(1) For evaluating the strength and the ductility

of piers, the effect of confinement of concrete by ties has not been precisely considered in the current Design Specifications. A new stress-strain relation of

confined concrete as shown in Fig. 2-3 was introduced

(7). The "ultimate strain" was defined as the strain

where concrete strength decreases 20% from its peak

value.

(2) Enough ties shall be provided to assure

ductility. Ties shall be of deformed bars with diameter

larger than 13 mm. The minimum interval of ties was

decreased from 30 cm to 15 cm to increase ductility

and shear strength. The interval of ties shall not be

suddenly changed but smoothly changed to prevent

sudden decrease of strength along piers.

Intermediate ties were newly introduced in

rectangular piers to assure the confinement. Diameter

and material of the intermediate ties shall be the same

with those of the ties, and shall be placed at each

level where ties are placed. The spacing of

intermediate ties in lateral direction shall be less than

100 cm.

(3) In the inverted "L" (or. "C") piers, eccentric

bending moment is induced in piers due to the dead

weight of decks. Such effect shall be considered in the

check of dynamic strength and ductility.

(4) For slender piers, the P-delta effect shall be

considered in the check of dynamic strength and

ductility.

(6) Termination of main reinforcement at mid-height shall not be made, in principle, to prevent

shear failure at the terminated points.

2.6 Check of Dynamic Strength and Ductility of

Concrete-Infilled Steel Piers

Although various researches have been made for

dynamic strength and ductility of steel piers, they are

not still sufficient to provide general stipulations for

the check of dynamic strength and ductility of steel

piers. Therefore, it was suggested to tentatively infill

concrete in steel piers for reconstruction and repair.

The procedure for the check of dynamic strength and

ductility of concrete infilled steel piers follows 2.5

with some appropriate modifications.

2.7 Check of Dynamic Strength and Ductility of

Foundations

Because the allowable stress design approach has

been used for foundations, appropriate design methods

for the check of dynamic strength and ductility

(deformation capability) have not yet developed for

foundations. It was therefore proposed to check the

safety of foundations tentatively assuming that the

flexural strength of piers evaluated in 4.5 shall be

applied to the footing as a seismic lateral force to the

foundation. In this evaluation, the safety factor for the

allowable stress design were increased.

2.8 Design of Bearings and Surroundings

Bearings have several functions, i.e.. they need to

support dead weight of a superstructure and live load,

allow relative movement due to elongation/shrinkage

of decks, support seismic lateral force, and prevent

excessive lateral movement and uplift during

earthquakes. Because various functions concentrate at

bearings, this makes the structure of bearings

complicated. Thus, the bearings tend to be damaged

during earthquakes. It was therefore suggested to

provide separate structural components to share the

function of bearings. This may be made by providing

additional members to prevent the uplift or excessive

162
relative movement.

Adoption of rubber bearings, in particular

Menshin bearings that has energy dissipating

capability, was recommended. Because the rubber

bearings allow relative movement to occur in various

directions, they are generally free from damage.

Bearings and surrounding parts of a

superstructure shall be designed considering the

mechanism of lateral force transfer from

superstructure to substructures. For example, when tall

bearings such as pin/roller type bearings are adopted,

a bending moment is induced associated with the

distance between the gravity center of a deck where

lateral force is applied and the bottom of bearings

where lateral force is supported. Although such

bending moment has been ignored in the past, this

shall be considered to properly design bearings and

surroundings.

Bearings shall be designed so that the ultimate

strength is larger than the force developed during the

extreme earthquakes.

2.9 Devices to Prevent Falling-down of

Superstructure from Substructure

In the current Design Specifications, it is

stipulated that at both ends of superstructure, either

the seat length S e or the installation of the devices

for preventing falling-down of a superstructure from

substructures shall be adopted. In the Guide

Specifications, it was described to provide both the

seat length and the installation of falling-down

prevention devices. The falling-down prevention

devices shall be so designed that they are safe against

the lateral force not only in longitudinal direction but

in transverse direction, and that the portion of the

decks where the falling-down prevention devices are connected to should have enough strength to prevent

rupture. It was suggested to adopt the falling-down

prevention devices with energy dissipation capability.

2.10 Countermeasures against Lateral Spreading

associated with Soil Liquefaction

At the sites where foundations are close to rivers

and channels and where the potential of liquefaction is high and the liquefiable sandy layer is inclined, effect

of lateral spreading shall be taken into account in

design. Because it is difficult at this moment to properly evaluate the earth pressure due to the lateral

spreading, general engineering judgment is required.

3. REFERENCE FOR APPLYING GUIDE

SPECIFICATIONS TO NEW HIGHWAY

BRIDGES AND SEISMIC STRENGTHENING

OF EXISTING HIGHWAY BRIDGES

3.1 General

For increasing seismic safety of the highway

bridges which suffered damage by the Hanshin/Awaji

Earthquake, various new drastic changes were

tentatively introduced in the Guide Specifications for

Reconstruction and Repair of Highway Bridges Which

Suffered Damage due to the Hyogo-ken Nanbu

Earthquake. Although intensified review of design

could be made when it was applied to the bridges

only in the Hanshin area, it may not be so easy for

field design engineers to following up the new Guide

Specifications when the Guide Specifications is used

for seismic design of all new highway bridges and

seismic strengthening of existing highway bridges.

Based on such demand, the Reference for Applying

the Guide Specifications to New Bridges and Seismic

Strengthening of Existing Bridges was issued on June

30, 1995 by the Sub-Committee for Seismic

Countermeasures for Highway Bridges, Japan Road

Association (5). Several educational seminars were

held with use of the Reference at major cities such as Tokyo, Osaka, Nagoya, Hiroshima, Fukuoka and

Sapporo, with the participants over 3,000.

The table of contents of the Reference was as:

I . Application of the Guide Specifications

U . Examples of Seismic Design for New Bridges

1. Supplements

1.1 Ground Motion for Nonlinear Dynamic

Response Analysis and Seismic Design

Force

1.2 Check of Dynamic Strength and

Deformation Capability of Foundations

1.3 Design of Falling-down Prevention

Devices

1.4 Countermeasures for Soil Liquefaction

1.5 Others

2. Examples of Seismic Design

2.1 Design Example for a New Bridge

Supported by Reinforced Concrete Piers

2.2 Design Example for a New Bridge

Supported by Concrete Infilled Steel Piers

2.3 Design Example for a Menshin Bridge

UI .Examples of Seismic Strengthening of Existing

Bridges

1. Supplements

163
1.1 Seismic Strengthening of Reinforced

Concrete Piers

1.2 Falling-down Prevention Devices for

Existing Bridges

2. Examples of Seismic Strengthening

2.1 Design Example of Seismic Strengthening

for Reinforced Concrete Piers

2.2 Design Example of Falling-down

Prevention Devices

3.2 Application of the Guide Specifications

The Reference classified the application of the

Guide Specifications as shown in Table 3-1 based on

the importance of the roads. All items of the Guide

Specifications are applied for bridges on extremely

important roads, while some items which prevent

brittle failure of structural components are applied for

bridges on important roads. For example, for bridges on the important roads, the items for menshin design,

tie reinforcements, termination of longitudinal

reinforcements, type of bearings, falling-down

prevention devices and countermeasures for soil

liquefaction are applied, while the remaining items

such as die design force, concrete infilled steel

bridges, and ductility check for foundations are not

applied.

3.3 Ground Accelerations Used for Nonlinear

Dynamic Response Analysis and Seismic

Coefficient for Check of Ductility and Dynamic

Strength

As reference ground motions used for nonlinear

dynamic response analysis, the strong ground motion

accelerations recorded by the H/A Earthquake at JMA

Kobe Observatory. JR Takatori Station (15). and

Higashi-Kobe Bridge (Route 5. Bay Shore Line of the

Hanshin Expressway) were suggested for analysis of

bridges at the Type I. II and III soil condition,

respectively.

In addition to the seismic coefficient specified in

the current Design Specifications for the check of

ductility and dynamic strength, it was also suggested

to use the seismic coefficient as

khc — Cz'khcO (3-1)

where. k h c =seismic coefficient for check of

ductility and dynamic strength, k h c o =standard

seismic coefficient and given by Table 3-2, and c z =zoning coefficient.

The standard seismic coefficient in Table 3-2 was

tentatively provided by taking envelopes of the

acceleration response spectra (damping ratio of 0.05)

of the records measured by die Hanshin/Awaji

Earthquake. The modification of the response spectra

in terms of the damping ratio vs. natural period

relation was incorporated (10). At mid-range natural

period, was assumed to be larger at Type I ground

condition (Stiff sites) dian Type II (moderate) and

Type III (soft soil sites) ground conditions. This

reflects the fact that surface ground accelerations were

smaller than the accelerations in underground at soft

soil sites.

3.4 Seismic Design of Foundations

The check of ductility and dynamic strength

(deformation capability) of foundations was more

quantitatively presented. In pile foundations, the check

is made as follows:

First, a foundation is analyzed by die

displacement mediod, in which a foundation is so

idealized that a rigid footing is supported by piles

which are elastically supported by soils. The flexural

strength of a pier shall be applied as a seismic force

to foundations at the bottom of die footing togedier

with the dead weight of superstructure, pier and soils on the footing. Safety of the foundation shall be

checked so that 1) vertical load induced in piles shall be less than half of their ultimate bearing capacity. 2)

lateral displacement of footing shall be less tiian about

30 mm, and 3) bending moment developed in piles

shall be less than their yielding moment.

When unreasonable size was required in die

foundation designed by the displacement method, die

nonlinear method shall be used. In the nonlinear

method, a foundation is idealized as shown in Fig.

3-1. Nonlinear parameters for vertical strength, uplift,

lateral movement and soil springs shall be evaluated

according to the Part IV Foundations of die

Design Specifications of Highway Bndges.

Nonlinearity of piles shall be evaluated in the same way to reinforced concrete piers.

3.5 Design of Falling-down Prevention Devices

It was tentatively suggested that die following

force shall be used as a design force of falling-down

prevention devices.

R (3-2)

where R d represents a reaction force at the bearing.

If the reaction force is different between the adjacent

164
girders, the larger value shall be used in Eq. (3-2) for

restrainers which tie together two approaching girders.

Allowable stress should not be increased in design of

falling-down prevention devices.

Because R d times twice of the seismic

coefficient has been so far used, Eq. (3-2) provides

almost twice as large lateral force than the previous

design force. Because the allowable stress has been

increased 50%, actual design force for falling-down

prevention devices was increased almost 3 times.

Rigid type restrainers should not be used to connect two adjacent girders if the size of the two

girders is significantly different. The size of the two

adjacent girders should be regarded significantly

different when the reaction forces of the two girders

are different by a factor of 2 or the fundamental

natural periods of the two structural systems are

different by a factor of 1.5.

It was suggested to provide some redundancy in

the seat length in the following conditions:

(1) skew bridge (tentatively, bridges with skew

angle less than 60 degree) and curved bridges

(tentatively, bridges with curvature less than

100m and angle formed by both ends less

than 30 degree)

(2) bridges supported by slender piers with long

natural period (tentatively, bridges with

natural period longer than 2 second)

(3) bridges which tend to have large lateral displacement in foundations due to soil

liquefaction/ lateral spreading

(4) bridges supported by slender columns with

extremely short seat length

It was also suggested to provide falling-down

prevention devices not only in longitudinal direction

but also in transverse direction in the following

bridges:

(1) skew bridge (tentatively, bridges with skew

angle less than 60 degree) and curved bridges

(tentatively, bridges with curvature less than

100m and angle formed by both ends less

than 30 degree)

(2) bridges with Gerber joints

(3) bridges supported by slender columns with

very short seat length

3.6 Termination of Longitudinal Reinforcements

It was described in the Guide Specifications that

longitudinal reinforcements shall not be terminated at mid-heights. This was acceptable because there were

few bridges which were supported by tall piers in the

damaged area. However. when the Guide

Specifications are used nationwide, there are bridges

which are supported by tall piers. Therefore, the

following minimum heights were suggested if the

termination is inevitable.

h i - H(l-M y i /2M y b )+b (3-3)

where, h i =i-th height from the bottom where

longitudinal reinforcements are terminated. H =height

of a pier, M y i =yielding moment of a pier at i-th terminated point, M y b =yielding bending moment of

a pier at bottom, and b=smaller dimension between

longitudinal and transverse widths of a pier.

Furthermore, it was suggested as follows:

(1) Longitudinal reinforcements should not be

terminated between the bottom and 2 times b

from it.

(2) Longitudinal reinforcements should not be at once reduced more that 1/3.

(3) Outside longitudinal reinforcements should not

be terminated.

3.7 Intermediate Ties for Circular Reinforced

Concrete Columns

It was suggested in the Guide Specifications that

the intermediate ties are not required for circular

reinforced concrete piers. However, they are effective

to increase shear strength even in the circular piers,

while confinement effect may be small. It was

therefore suggested to tentatively provide the

intermediate ties at every other ties in circular

reinforced concrete piers. In the circular piers, the

intermediate ties can be considered to contribute to shear strength, but it should not be considered to evaluate ductility in terms of the tie reinforcement

ratio. It was described in the Guide Specifications that

the intermediate ties can be considered to evaluate

shear strength and the confinement effect in the

rectangular piers.

3.8 Seismic Strengthening of Existing Reinforced

Concrete Columns

Because damage concentrated to single reinforced

concrete piers/columns with small concrete section,

seismic strengthening is being initiated for those

columns, which were designed by the pre- 1980

Design Specifications, at extremely important bridges.

Main purpose of the seismic strengthening of

reinforced concrete columns is to increase their shear

strength, in particular in the piers with termination of

165
longitudinal reinforcements without enough anchoring

length. This increases ductility of columns, because

premature shear failure could be avoided.

However if only ductility of piers is increased,

residual displacements developed at piers after an

earthquake may increase. Therefore the flexural

strength should also be increased. However the

increase of flexural strength of piers tends to increase

the seismic force transferred from the piers to the

foundations. It was found from an analysis to various

types of foundations that failure of the foundations by

increasing the seismic force may not be significant if

the increasing rate of the flexural strength of piers is

less than 2. It is therefore suggested to increase the

flexural strength of piers within this limit so that it does not cause serious damage to foundations.

For such requirements, seismic strengthening by

Steel Jackets with Controlled Increase of Flexural

Strength was suggested. This uses steel jacket

surrounding the existing columns as shown in Fig.

3-2. Epoxy resin or non shrinkage concrete mortar are

injected between the concrete surface ands the steel

jacket. A small gap is provided at the bottom of piers

between the steel jacket and the top of footing. This

prevents to excessively increase the flexural strength.

To increase the flexural strength of columns in a

controlled manner, anchor bolts are provided at the

bottom of the steel jacket. They are drilled into the

footing. By selecting appropriate number and size of

the anchor bolts, the degree of increase of the flexural

strength of piers may be controlled. The gap is required to trigger the flexural failure at the bottom of

columns. A series of loading tests are being conducted

at the Public Works Research Institute to check the

appropriate gap and number of anchor bolts. Table 3-3

shows a tentatively suggested thickness of steel jackets and size and number of anchor bolts. They are

for reinforced concrete columns with a/b less than 3,

in which a and b represent the width of column in

transverse and longitudinal direction, respectively. The

size and number of anchor bolts were evaluated so

that the increasing rate of flexural strength of columns

is less than about 2.

4. CONCLUDING REMARKS

The Hanshin/Awaji Earthquake was the first earthquake that hit an urban area in recent years, and

revealed that various technical developments are

required in both seismic design and seismic

strengthening for mitigating the damage of highway

bridges against extreme inland earthquakes.

This paper summarizes the Guide Specifications

for Reconstruction and Repair of Highway Bridges

Which Suffered Damage due to the Hyogo-ken Nanbu

Earthquake and the Reference for Applying Guide

Specifications to New Highway Bridges and Seismic

Strengthening of Existing Highway Bridges. Tliis was

one of a series of actions taken for seismic design and

seismic strengthening of highway bridges in about a year since the earthquake.

The Special Sub-Committee for Seismic

Countermeasures for Highway Bridges in the Bridge

Committee of the Japan Road Association is drafting

the revision of the Design Specifications of Highway

Bridges. This will include various lessons learned

from the Hanshin/Awaji earthquake. It is planned to complete the revision in 1996 fiscal year.

ACKNOWLEDGMENTS

The contents presented here were the works

conducted at the Seismic Design Working Group in

the Ministry of Construction under the general

guidance by the "Committee for Investigation on the

Damage of Highway Bridges Caused by the

Hyogo-ken Nanbu Earthuqake." The authors thanks

the kind guidance and supports from the all Committee members. Special appreciation goes to Dr.

T. Iwasaki, chairman of the committee. Messer. N.

Miayata, Y. Iwasaki. and S. Kikuchi. Headquaters of

the Ministiry of Construction.

APPENDIX

Committee Members for Investigation on The Damage

of Highway Bridges Caused by The Hyogo-ken

Nanbu Earthquake

Dr. T. Iwasaki, Exective Director, Civil Engineering

Research Laboratory (Chairman)

Dr. Y. Fujino,Professor, Univesrity of Tokyo

Dr. H. Iemura, Professor. Kyoto Univertsity

Dr. S. Ikeda, Professor, Yokohama National

University

Dr. H. Kameda, Professor, Kyoto University

Dr. T. Katayama, Professor. University of Tokyo

Dr. K. Kawashima, Professor. Tokyo Institute of

Technology

Dr. Y. Onishi, Professor, Kyoto University

Mr. S. Saeki, Honshu-Shikoku Bridge Authority

Dr. K. Toki. Professor, Kyoto University

166
REFERENCES

1) California Department of Conservation: Processed

CSMIP Strong Motion Records from the

Northridge. California Earthquake of January 17,

1994, Office of Strong Motion Studies. 1994

2) Hoshikuma, J.. Kawashima, K., Nagaya, K.: A

Stress-Strain Model for Reinforced Concrete

Bridge Piers Confined by Hoop Reinforcement.

Proc. 2nd US-Japan Workshop on Seismic

Retrofit of Bridges. Berkeley, CA. USA, 1994

3) Japan Road Association: Guide Specifications for

Earthquake Hazard Mitigation for Road

Transportation Facilities, Pre Earthquake

Countenneasures. 1987

4) Japan Road Association: Design Specifications of

Highway Bridges- Part I Common Part. Part II

Steel Bridges, Part III Concrete Bridges, Part IV

Substructures, and Part V Seismic Design, 1990

5) Japan Road Association: Reference for Applying

the "Guide Specifications for Reconstruction and

Repair of Highway Bridges Which Suffered

Damage due to the Hyogo-ken Nanbu

Earthquake" to New Highway Bridges and

Seismic Strengthening of Existing Highway

Bridges. 1995

6) Japan Society of Civil Engineers: Standard

Specifications of Concrete, 1931. 1940, 1949,

1956. 1967, 1974, 1977, 1980

7) Kawashima, K., Unjoh. S. and Iida, H. : Seismic

Inspection and Seismic Strengthening of Highway

Bridges in Japan. Proc. 4th US-Japan Workshop

on Earthquake Disaster Prevention for Lifeline

Systems, Los Angeles. CA. USA. 1991

8) Kawashima, K., Hasegawa. K., and Nagashima.

H: Manual for Menshin Design of Highway

Bridges. Proc. 24th Joint Meeting. Panel on Wind

and Seismic Effects. UJNR, Gaithersburg, MD,

USA. 1992

9) Kawashima, K. and Hasegawa, K: New Seismic

Design Specifications of Highway Bridges in

Japan, Earthquake Spectra. Vol.10. No.2, 1994

10) Kawashima, K.: Seismic Design Force of

Bridges, Proc. 2nd International Workshop on

Seismic Design and Retrofitting of Reinforced

Concrete Bridges. Queenstown New Zealand,

1994

11) Ministry of Construction: Manual for Menshin

Design of Highway Bridges, 1992

12) Ministry of Construction: Report on the Damage

of Highway Bridges by the Hyogo-ken Nanbu

Earthquake, Committee for Investigation on the

Damage of Highway Bridges Caused by the

Hyogo-ken Nanbu Earthquake. 1995

13) Ministry of Construction: Guide Specifications

for Reconstruction and Repair of Highway

Bridges Which Suffered Damage due to the

Hyogo-ken Nanbu Earthquake. 1995

14) Miyaji, A.: Road Disaster Prevention Measures.

Journal of Road. Japan Road Association. 1990

15) Nakamura. Y.: Waveform and Its Analysis of the

1995 Hyogo-ken Nanbu Earthquake. JR

Earthquake Information. No. 23c. Railway

Technical Research Institute. 1995

16) Narita, N., Murakami, M. and Asanuma. H:

Report of the Investigation on Earthquake

Damage to Shizunai Bridge. 15th Joint Meeting.

US-Japan Panel on Wind and Seismic Effects.

UJNR. Tsukuba, Japan. 1983

17) Ozaka. Y., Suzuki, M.. Miyamoto. M. and

Kobayashi. S.: Evaluation of Shear Strength of

RC Beams with Termination of Longitudinal

Reinforcements, Proc. JSCE, 378/V-l, 1987

18) Road Maintenance Center: Reference for Design

and Construction to Eliminate Expansion Joints

from Existing Bridges. 1995

19) Takeda. T.. Sozen. M. A., and Nielsen. N. N.: Reinforced Concrete Response to Simulate

Earthquakes, Journal of Structural Engineering.

ASCE, 96-12. 1970

167
Table 3-1 Application of the Guide Specifications

Table 3-2 Standard Coefficient k h c o

Table 3-3 Steel Jacket and Anchor Bolts

Fig. 2-1 Accelerations Measured at JMA Kobe Observatory and Computed

(a) N-S Component

(b) E-W Component

(c) U-D Component

Fig. 2-2 Response Spectra (h=0.05) at JMA Kobe Observatory (NS) in

comparison with the Ground Earthquake, JMA Kushiro Observatory

by the 1993 Kushiro-oki Earthquake, and Sylmar Parking Lot by the

1994 Northridge Earthquake

Fig. 2-3 Stress-Strain Relation for Concrete Confined by Ties

Fig. 3-1 Idealization of Pile Foundation in the Nonlinear Analysis

Fig. 3-2 Seismic Strengthening of Reinforced Concrete Piers by Steel Jacket with

Controlled Increase of Flexural Strength

168
Table 3-1 Application of the Guide Specifications

Type of Roads and Bridges

Double Deckers, Overcrossings on

Roads and Railways, Extremely

Important Bridges from Disaster

Prevention and Road Network

Others

Expressways, Urban Expressways,

Designated Urban Expressways,

Honshu Shikoku Bridges,

Designated National Highways

Apply all items, in principle Apply all items, in principle

Non-designated National Highways,

Ken Roads, City, Town and Village

Roads

Apply all items, in principle Apply partially, in principle

Table 3-2 Standard Coefficient khco

Group I r£e <o.3s

khc0 =4.46TEQ

2' 3

0.3^7*Ee ^0.7s

khc0 =2.0

0.7s < TEQ

khcQ =l.24TEQ ^'3

Group II TEQ <OAs

khc0 =3.22TEQ

2/3

0.4^ TEQ ^ 1.2 s

khcQ =1.75

l.2s<TEQ ^c0 =2.23r£2

-4/3

Group ED TEQ <0.5s

khc0 =2.3STEQ

m

o.5^teq as- 1.5 S

khc0 =L50

i.5s<rfQ

khcQ =2.57TEQ

-*'3

Table 3-3 Steel Jacket and Anchor Bolts

Columns/Piers Steel Jackets Anchor Bolts

aA)S2 SS 400, t= 9 mm

2<a/b^3 SD 295, D35 etc 250 mm

Columns supporting Lateral Force SS 400, t= 12 mm

through Fixed Bearings and with a/b^3

169
N-S Component

10

TIME(S)

(a) N-S Component

20

E-W Component

10

TIME(S)

15 20

20 r

z 10

OJo

-10

CL

<2 -20

Q

<

i

i

i

1 1 1 L_iu 1 1 1 1 I | | |

1 '

1 1 1

10

TIME(S)

(b) E-W Component

15 20

170
I

20

1- 10

UJ

UJ u o

-1 0 -

a.W -20

Q

0

U-D Component

-i i i t_

10

TIME(S)

15

_L

10

TIME(S)

1 5

20

20

(c) U-D Component

Fig. 2-1 Accelerations Measured at JMA Kobe Observatory and Computed

171
3500

_l

O 3000

<

oo 2500 ZO

h-<

2000

LU_l

UJ 1 500

U

< 1000

LU

CO

z:o 500

Q.OO

LU 0

— JMA Kobe Observatory

JRTakatori

; Higashi Kobe Bridge

-• JMA Kushiro Observatory

Sylmar

_l 1 ) 1 1 1 1 L

0.5 1 1.5 2 2.5

NATURAL PERIOD T (SEC)

(a) Response Acceleration Spectra

450

uj 400

oo2 350

o

> 300

OO

>- 250

O 200

f

£ 150 If

1 f

LU 100

o i u ; °- 50 oo ou

LU fa' ~ 0

! \ I

v

-

0.5 1 1.5 2 2.5 3

NATURAL PERIOD T (SEC)

(b) Response Velocity Spectra

Fig. 2-2 Response Spectra (h=0.05) at JMA Kobe Observatory (NS) in

comparison with the Ground Earthquake, JMA Kushiro Observatory

by the 1 993 Kushiro-oki Earthquake, and Sylmar Parking Lot by the

1994 Northridge Earthquake

Stress

Fig. 2-3 Stress-Strain Relation for Concrete Confined by Ties

172
cd •M

03

CO>u

03

i_CD M

03

-Q irCo

-4-> _03CD q:

4—' c

cdE

CD

CJ

_03

Q.

CO Q

CD T3O

03C< >

CO

CD

CD M03>

=3 Q_ u _. cd

CO CD > 4—' C >4- cd O E c o o z -M

'a? cd o- a: co

4—'

_03CD Cd

CD

i_3

03>3

CJ

CO>

4—1 c

CDEo

CO

ol

CD •M

CD

i_CJco

CJ

T3

CDUoc

'CD on

CO

T3 >4—O

"03CJ

'+->

cd > c

CO

o > 4—'

03

u CD C£ o

u_

4-< c

~03 CD

CJ

E '+-" CD i_

CD

CJ

03 > Q.

CO b

CO • I—I

CO>^

S3

.51

CO

.3ao8

-ao

CO

CmOao• 1-1 t3 .a

COi • i—i

173
I

Fig. 3-2 Seismic Strengthening of Reinforced Concrete Piers by Steel Jacket with

Controlled Increase of Flexural Strength

174
Comparison of U.S. and Japanese Highway Bridge Column Seismic Design

U.S.

W.Phillip Yen1

, Research Structural Engr.

James D. Cooper1

, Chief of Structures Div

Abstract

This paper presents the preliminary

results of a U.S. -Japan cooperative research

project which compares the seismic design

procedures and results for a bridge pier

under the same design force levels. These

design were based on the current bridge

design specifications of the two countries.

Design results, such as column sizes and

main and transverse reinforcement, are compared, and the design approaches are discussed. Following these comparative

designs, one-sixth scale models of the two

prototype designs will be built and tested on

a shake table.

1. INTRODUCTION

1.1 Background

Recently, three large destructive earthquakes

have occurred in the United States and Japan.

These earthquakes, the Loma Prieta

earthquake in 1989, the Northridge

Japan

Hisanori Otsuka2

, Head of Earthquake Eng. Div

Toru Terayama2

, Senior Research Engr.

.

earthquake in 1994, and the Kobe earthquake

in 1995, have severely damaged a number of

highway bridges, and have cost many lives

and billions of dollars. Although, most of

the damaged bridges were designed and

constructed prior to the implementation of

modern seismic design codes, significant

damages also occurred in some bridges built

to more recent codes. Column failure due to

insufficient ductility and inadequate shear

reinforcement was found to be the most

important factor causing bridge collapse or

failure.

In order to better evaluate the current

bridge seismic codes, and to improve bridge

performance under earthquakes, a cooperative research study was initiated in

the Third US-JAPAN workshop on

Advanced Technology in Highway

Engineering, September, 1994. The two

primary tasks proposed were to design a

single column bridge pier using both US and

Japanese seismic design codes, and to

construct scale models of each design for

shake-table testing at the Public Works

1 Structures Division, Turner-Fairbank Highway Research Center

Federal Highway Administration

McLean, Virginia 22101

2 Earthquake Engineering Division

Public Works Research Institute

Tsukuba Science City, Japan

175
Research Institute (PWRI) Lab in Japan. The

methods used for the prototype bridge

column designs are based on the current

seismic design specification of both

countries. The U.S. side used the newly

revised (Spring 1995) Division I-A: Seismic

Design Provision of the American

Association of State Highway Officials'

(AASHTO) Standard Specification for

Highway Bridges (15 edition), and the

Japanese side used the Seismic Design Codes

of the Japan Road Association's (JRA)

Specification for Highway Bridges (1990).

1.2 Objective

The objective of this cooperative research

study are 1) to better assess the current

seismic design codes in column design, and

2) to improve bridge performance under

earthquakes. The shaking-table tests of the

scale models will give an indication of the

adequacy of current design criteria for bridge

piers. In this paper, design procedures for

single column bridge piers under comparable

seismic forces are discussed, and the two set

of design results are compared. The

difference in the two design approaches are

also addressed.

2. DESIGN CONDITIONS

2.1 Bridge Descriptions

A bridge representative of the most

common type in both countries was selected

and is shown in Figure 1. This bridge

consists of two continuous spans on tangent

alignment with seat type abutments at the

ends. The superstructure is assumed to have

a reinforced concrete deck, and steel

stringers with hinged bearings. The pier is

assumed to have a hammer head cap, a single

circular column, and a spread footing. In

this study, only the column design is

addressed.

2.2 Seismic Design Conditions

Design Approaches

In order to get a meaningful

comparison between the US and Japanese

seismic design methods, design assumption

need to be identical. However, this is not

feasible in every instance because the two

design approaches are different. For

example, a major difference between the two

codes is in the way design strengths are

specified. The JRA Code uses allowable

stress, while the AASHTO code is based on

ultimate strength. The JRA code permits

increasing the allowable stress by a factor of

0.5 under seismic forces. However, it also

requires a check for the ultimate horizontal

strength, and uses a coefficient K,,,. as high as

1.0. Khc is the seismic design coefficient

used to check ultimate horizontal strength

during an earthquake (JRA Codes 1990).

Another difference is that forces and

displacements for the "design" earthquake as specified in AASHTO are based on a low

probability of their being exceeded within the

normal life expectancy of a bridge. The

AASHTO code specifies probabilisticallybased effective peak accelerations throughout

the U.S. The JRA code uses a regional

seismicity classification map to account for

the intensity of design earthquake motions.

This map was developed on the basis of

statistical analyses from past strong motion

data (Kawashima et al, 1989) which were

then consolidated into three regions. This

required elimination of very localized

discrepancies within a region. Reasons for

176
this difference may be due, in part, to the

frequency of occurrence and the type of

significant earthquakes that have occurred

historically in both countries.

For this study, these seismic

conditions were chosen to be as close as possible. Table 1 shows the seismic design

conditions assigned for both US and Japanese

design methods. Acceleration coefficient,

site effects and inertia forces conditions are discussed as follows:

A. Acceleration Coefficient

AASHTO

JRA

A=0.4

Kho =0.2/Kheo =1.0

Since the two codes are quite

different in their design approaches,

assigning the same coefficient in both

methods would be meaningless.

Consequently, the same maximum response

spectrum (design force) was applied for this

design case. JRA requires a check for the

ultimate horizontal strength by using the

coefficient Khc up to 1.0 in addition to the

standard horizontal seismic design coefficient

Kh0 =0.2. This l.Og force level for is

equivalent to a value of A = 0.4 in the

AASHTO design. Utilizing single mode

analysis gives a maximum value for the

elastic seismic response coefficient Cs of

2.5A in AASHTO.

1 2AS

( Note: C = — ; where S and T 5 ^2/3

represent site effects coefficient and the

fundamental period of the bridge)

TG can be

In both design codes, the effects of

site conditions on bridge response are determined from a site coefficient based on

soil profile (or ground) types. In AASHTO,

soil profiles are divided into four different

soil types. These classification are based on

the type of soil, (such as rock, stiff clay, soft

to medium-stiff clay and silts), the depth of

the soil, and shear wave velocity. A qualified

geotechnical engineer's help is often needed

to determine the soil profile type. The JRA

states that ground type in seismic design shall

generally be classified as one of three types.

The determination of the ground type is by a

factor TG which considers soil strata

thickness and average shear wave velocity,

and is expressed in seconds,

obtained by a simple formula

= 42 —'-

Where Ht is the thickness of the i-th soil

layer (m) and Vsi represents the average

elastic shear wave velocity of the i-th soil

layer.

A soil profile with a depth of 45.9 ft (14 m) and an average shear wave velocity of

984ft/sec (300m/s) was assumed in this

design case. With the JRA method, the

ground type is clearly Type I. Under the

AASHTO method, Classifying the soil from

this information could be difficult since it

falls somewhere between Type II and Type

III. However, Type III was selected because

of the shear wave velocity.

B. Site Effects C. Inertia Forces

AASHTO: Type III JRA : Type I

Both codes agree that inertia forces

from the superstructure are assumed to be

177
acting on the bridge seat in the longitudinal

direction, and at the bottom of the deck

(height = 2.30m above the bridge seat) in

the transverse direction. The pier is

assumed to take the inertia of the two full

spans' weight in the longitudinal direction,

and half the weight of each span in the

transverse direction.

3. SEISMIC DESIGN

3.1 AASHTO Method

In designing the bridge pier assumed,

a single mode spectral analysis method was

used since this is a right bridge of less than

7 spans with a fairly constant cross section,

and can be considered regular. The seismic

performance category (SPC) of this bridge is

D because the acceleration coefficient (A) is

0.4 and the bridge is assumed to be essential.

Elastic forces obtained from the single mode

analysis were reduced by an R factor

(Response Modification Factor). This design

strategy permits these elastic forces to be

reduced substantially to obtain design forces

for a column member that will yield first.

For a single column pier design, this R factor

is 3. Seismic loadings for both transverse and

longitudinal directions were analyzed to find

the most critical forces for design. In this

case, longitudinal loading with the

fundamental natural period TL = 1.25 was

the critical design load. SAP90 was used to

check the final design by reanalyzing the

structures with a moment-release joint at the

pier top, and the results agreed very well.

3.2 JRA Method

In the JRA method, design loads are obtained from a horizontal seismic

coefficient method based on the allowable

stress approach. A standard horizontal

seismic coefficient Kh0 = 0.2 for design was

used in this case. An increased design load

was then used to check the ultimate

horizontal strength and minimum ductility.

The purpose of the ductility requirement is to

ensure that brittle failure of reinforced

concrete substructure elements does not

occur under seismic loading. This capacity

check used a coefficient for seismic checking

=1.0. The force level was then lowered

by dividing by a ductility reduction factor

(2fx-l)'

Am

, in which /u, is the ductility of the

bridge column, and is a function of a strainstress model (Hoshikuma and Kawashima

1992).

4. DESIGN RESULTS and

COMPARISON

As part of the comparison of results

and exchange of design examples, a parametric study of different sizes was

performed by both sides to make the

comparison between the two design codes

more meaningful. The AASHTO bridge

column sizes were 7 ft (2.1 m) and 9 ft (2.7m) in diameter, while the JRA columns

were 2.7m ( 9ft), and 3.2 m (10.5 ft) in

diameter. Design detail of the bridge piers

from both design codes are shown in Figure

2 (US) and Figure 3 (Japanese). Design loads

including natural periods, axial loads, and

seismic horizontal forces are compared in

Table 2. Column sizes and steel

reinforcement in the longitudinal direction

and transverse directions are compared in

Table 3. The comparisons shown are only

for the smallest sizes for each code.

5. CONCLUSIONS AND REMARKS

This paper presented a comparison of results

178
from seismic design of a typical bridge

column by both US and Japanese methods.

The results here do not seek to show the

stronger or weaker design but rather to

compare the different design approaches, and

to learn from each. From the natural periods

compared in Table 3, the AASHTO pier is

more flexible than the JRA pier in this case. These designs may also reflect the difference

in the strong ground motion, which is

affected most by the soil profile type.

Yen, W. P., Cooper, J.D., and Friedland, I. M. 1995, " Comparison of the U.S. and

Japanese Highway Bridge Seismic Design

Criteria", Proceedings of the 27th UJNR

Panel Meeting on Winds and Seismic

Effects, Tsukuba City, Japan.

REFERENCES:

AASHTO, 1995, Interim Revisions to the

AASHTO Standard Specifications for

Highway Bridges, "Division I-A: Seismic

Design," American Association of State

Highway and Transportation Officials,

Washington, DC.

Japan Road Association, 1990,

"Specifications for Highway Bridges," Part

V: Seismic Design.

Kawashima, K., 1994, "Seismic Design

Force of Bridges," Proceedings of the

Second International Workshop on Seismic

Design of Bridges, Queenstown, New

Zealand.

Kawashima, K., and Hasegawa, K., 1994,

"New Design Specification for Highway

Bridges in Japan," Earthquake Spectra, Vol.

10, No. 2.

Sultan, M, and Kawashima, K. 1994,

"Comparison of the Seismic Design of

Highway Bridges in California and in

Japan", Recent Selected publications of

Earthquake Eng. Div., PWRI, Japan.

(Technical Memorandum of PWRI No.

3276)

179
j

I

I

I

I

Table 1. Seismic Design Conditions

i

i

Boundary Condition AASHTO JRA

Design Method Single Mode Spectral

Analysis Method

Horizontal Seismic

Coefficient Method

Acceleration Coefficient

(Max. Response Spectrum

Coefficient =1.0)

A=0.4 Kho =0.2 (Design)

Khco =1.0 (Checking Ultimate

Horizontal Strength)

oun riuiiic i ypc/ VJIUIUIUType

(H = 14m Shear Wave

Velocity = 300m/sec)

^nil Prnfi1<* Tvnp TTT

S = 1.5

rilrminH Tvnp T ( T —fl 1 Re"*

CG = 0.85

Connection Type Hinge Type Fixed Bearing Hinge Type Fixed Bearing

Inertia Forces

Acting on bridge seat

(Long, direction)

Acting at the bottom of the

deck (Trans. Direction)

Acting on bridge seat

(Long, direction)

Acting at the bottom of the

deck (Trans. Direction)

Table 2. Comparison of Seismic Design Forces

Description AASHTO Method JRA Method

Fundamental Natural Period

(longitudinal direction)

1.29 sec 0.93 sec

Axial Load 1196 kips 507 tons

Seismic Horizontal Force 0.62 X Ws 0.7 X Wj

Design Force Reduction

Factor

3 2.08

180
Table 3. Comparison of Design Results

Characterization AASHTO Method JRA Method

Column Size Diameter = 7 ft (2.1m) Diameter = 2.7 m ( 9 ft)

Main Reinforcement

(longitudinal direction)

56-#18 bars

(56-D57)

2X54-D51 bars

(2X54-#16 Double layer)

Transverse Reinforcement #7@3"(D22@75)

(Spiral type)

D22 @150 (#7@6")

(Hoop type)

Concrete Gross Area 38.48 ft

2 (3.57m2) 5.73m2 (61.63 ft

2

)

Figure 1. Bridge Column Dimensions

181
182
Si

; I

, i—

uL

ai "I

:

?! : i si

^1 tfii *i a i o o I o

:|S|J

*

H

tS 9

£ 5

513 i.io- 5TTT >

-wnn so:

.

W!

-5ST7T

50? 000 c O0S i I -04 l#*0

I

183

RESIDUAL DISPLACEMENT RESPONSE SPECTRUM AND

ITS APPLICATION TO REINFORCED CONCRETE BRIDGE PIERS

Kazuhiko Kawashima*, Gregory A. MacRae^, Jyun-ichi Hoshikuma-*

and Kazuhiro Nagaya^

ABSTRACT

A structure subjected to strong ground

shaking may develop large residual

displacements after the shaking due to

nonlinear response. The residual

displacement causes various problems for

repair and reconstruction of structures

after an earthquake. This paper proposes

residual displacement response spectra,

which describe the residual displacement

developed in a one-degree-of-freedom

oscillator with various natural periods,

damping ratios, ductility factors and

bilinear factors. Characteristics of the

residual displacement ratio response

spectra which is normalized by the

maximum residual displacement was

clarified based on 63 components of

ground motion, and a design value is

proposed for the residual displacement

response spectra. An example of

application is also presented.

Key Words: Seismic design, Nonlinear

behavior, Ductility

1. INTRODUCTION

The Hanshin-Awaji Earthquake of

January 17, 1995 caused destructive

damage to bridge structures. About 250

reinforced concrete columns and 20

stiffened steel columns totally collapsed

due to the strong shaking (MOC Report

1995). There were various types and

amounts of damage. It should be noted

that some bridge columns did not collapse

but tilted and had large residual

displacements at the top. Due to the

difficulty to set the superstructures back

to the original alignment and level, most

of them were required to be rebuilt. This

clearly indicates the importance in seismic

design of structures to consider the

residual displacement associated with

nonlinear behavior during an extreme

ground motion. A series of shaking table

tests conducted at the Public Works

Research Institute for model bridges

supported by reinforced concrete

columns and steel columns have shown

that large residual displacements can be

developed under extreme ground motions

(for example, Kawashima, MacRae and

Hasegawa, 1992, Kawashima and

Hasegawa, 1994a). The force reduction

from the elastic force level should be

carefully determined not only from the

peak response displacement but also from

the residual displacement developed in a

structure after an earthquake.

1) Professor, Department of Civil

Engineering Tokyo Institute of

Technology

2) Assistant Professor, University of

Washington

3) Research Engineer, Public Works

Research Institute, Ministry of

Construction

4) Assistant Research Engineer, Public

Works Research Institute, Ministry of

Construction

185
Various studies have been made on the

nonlinear behavior of structures subjected

to strong ground motions and it is well

known that the residual displacement

increases as the ductility of structures

increases. However, the dependence of

the residual displacement on the ductility

under strong ground shaking has not yet

been precisely investigated.

This paper proposes a residual

displacement response spectrum, which

presents the residual displacement

developed in a one-degree-of-freedom

oscillator with various natural periods,

damping ratios, ductility and bilinear

factors, and presents an application of the

residual displacement response spectra.

2. RESIDUAL DISPLACEMENT

RESPONSE SPECTRUM

A one-degree-of-freedom oscillator is

assumed to be supported by a spring

with a nonlinear hy steretic restoring force

characteristic. The oscillator is subjected

to a ground shaking at its base. Although

any hy steretic restoring force model may

be assumed in the spring a bilinear model

as shown in Fig 1 is assumed here for

simplicity ofthe analysis.

The spring changes its stiffness from

an initial (elastic) stiffness to a second

(post-yielding) stiffness hi at a yield

displacement uy

. The bilinear factor r

and the ductility factor ju are defined as

fl = ^^

(2)

Uy

where is the maximum

displacement of the oscillator.

To represent the energy dissipation

contributed from sources other than the

hy steretic energy dissipation, such as

radiation of energy from a foundation to

surrounding soils, a viscous damping ratio

h is considered in the analysis in addition

to the hy steretic damping associated with

the nonlinear restoring force.

When an oscillator is subjected to a ground motion at its base, a residual

displacement ur may be developed

after the shaking has finished. The

magnitude of residual displacement ur depends on various factors such as the

ductility factor ju, natural period T,

damping ratio h, bilinear factor r and the

intensity of ground shaking Because the

displacement sometimes does not come to

its rest point even at the end of ground

acceleration records, a time of zero

acceleration was added to the end of each

record to compute the final rest point of

the displacement.

Thus the residual displacements can be

computed from each strong motion record

for oscillators with various values of

ductility factor ju, natural period T,

damping ratio h, and bilinear factor r (MacRae, Kawashima 1993). The residual

displacement plotted against natural

period T, damping ratio h, bilinear factor r and ductility factor fx is proposed here

to be designated as a residual

displacement response spectrum Srd

from its analogy to the original definition

ofthe response spectrum (Housner 1959).

Target ductility factors fij were

selected here as 2, 4 and 6, and the yield

force F

y

was obtained so that the

ductility factor ju computed corresponds

to the target ductility factor fi T . This

requires some iterations as shown in Fig

2. First one assumes F

y

to compute

"max for a given ground motion

acceleration. If the ductility factor ju

computed by Eq. (2) is not close enough

186
ur max

to a new value of F

y

is chosen to

compute Hmax until the difference

between ju and \ij becomes sufficiently

small. Generally 5 iterations are sufficient

to have a convergence with an error of

less than 0.1%.

For any hysteretic curves, there may

be a maximum value of the residual

displacement. As shown in Fig 3, the

residual displacement takes its maximum

value when oscillator unloads from the

maximum displacement umax . If it is

assumed that the motion of oscillator

comes to rest when the restoring force

becomes zero, the maximum residual

displacement urmax may be written as (jU-l)(l-r)M

y r{n -1)<1

{(l-r)lr}u

y

r(/*-l)s=l

(3)

It is convenient to normalize the

residual displacement ur by urmax as SRDR =k /fVmaxl (4)

Srdr ls designated here as a residual

displacement ratio. The plots of

against natural period T, damping ratio h,

bilinear factor r and ductility factor fx are

designated here as residual displacement

ratio response spectra.

In the following the natural period J is

computed from the initial stiffness

and the mass of the oscillator. The

viscous damping ratio h is assumed to be

0.05 that is typical for usual structures.

3. RESIDUAL DISPLACEMENT

RESPONSE SPECTRA FOR A

TYPICAL RECORD

Fig 4 shows a residual displacement

ur computed for an oscillator with natural

period T=ls subjected to a free field

ground acceleration recorded in the

vicinity of Itajima bridge during the

Hyuga-nada earthquake in 1968. The

record was obtained by an earthquake

with magnitude 7.2 at 195 km from the

epicenter. The ductility factor ii was

assumed to be 4. It is seen in Fig 4 that

the residual displacement starts to

develop at about 16 seconds, when the

ground acceleration starts to become

larger. It is important to note that ur increases as the bilinear factor r decreases.

In particular, the residual displacement is

quite large when r=-0.05. It is interesting

to note that although there are some

responses in which the magnitude of

residual displacement gradually increases or even changes direction by accumulation

of drift as shown in Fig 4(d), most of

residual displacements are developed by

one single shift in drift as shown in Fig 4

(b) and (c).

The residual displacement was

computed in the same manner for

oscillators with various natural periods,

and the residual displacement ratio

response spectrum SRDR defined by Eq.

(4) was evaluated for the Itajima record as shown in Fig 5. The absolute acceleration

response spectrum Sa with a damping

ratio of 0.05 is presented in Fig 5 for

comparison. It is seen from Sa that the

predominant periods of ground motion

are about 0.2s and 0.7s. The residual

displacement ratio response spectrum

Srdr depends significantly on the natural

period. Reflecting the results described in

Fig 4, Srdr is almost 1.0 when r is -

0.05 while Srdr ls less than 0.6 when r

is 0. 1 . It is interesting to note that Srdr

tend to have peak values at the natural

periods of about 0.2-0.3 s and 1 s which

are close with the predominant periods of

ground motion.

187
4. ANALYSIS OF RESIDUAL

RESPONSE SPECTRA

The residual displacement ratio

response spectrum SRDR was

evaluated for 63 components of ground

acceleration. They were recorded by

earthquakes with magnitude ranging from

6.5 and 7.9 at free-field ground surface.

Any records obtained at basement and

first floor of buildings were not included.

Fig 6 shows the earthquake magnitude

and the epicentral distance of the records.

About 71 % of the records were induced

by the earthquakes with magnitude larger

than 7. The ground condition was

classified into three groups as shown in

Table 1 based on the natural period of

ground T

g (JRA 1990). The number of

the records at Group I (stiff), II (medium)

and IH (soft) sites are 13, 37 and 13,

respectively. Fig 7 shows the peak

acceleration of records used in the

analysis. About 1/3 of the total records

have peak accelerations larger than 60 gal

(cm/s2

).

(1) Effect of Earthquake Magnitude

and Epicentral Distance

Fig 8 shows how the residual

displacement ratio response spectra

Srdr depend on the earthquake

magnitude M and the epicentral distance

Aof the records. Only Srdr for T =\s

and jU =4 are presented here, because

other conditions show the similar results.

The earthquake magnitude is classified

into 6.5<;M<7, 7<M<7.5, and 7.5=sM

It is apparent from Fig 8 that the

residual displacement ratio response

spectra SRDR does not consistently

depend on the earthquake magnitude M

and the epicentral distance A

.

(2) Soil Condition, Natural Period and

Ductility Factor

The residual displacement ratio

response spectra Srdr were averaged

over all earthquake magnitudes and

epicentral distances for three bilinear

factors and the three soil conditions. Figs.

9 and 10 show the mean value m and the

standard deviation a, respectively, for

residual displacement ratio response

spectra SRDR which were computed for

specific combinations of natural period

and ductility factor.

The dependence of the mean value m

of the residual displacement ratio

response spectra Srdr on the bilinear

factor r is considerable. This effect will be

precisely discussed later.

It is interesting to note that the soil

condition is less significant for the

residual displacement ratio response

spectra Srdr. The effect of natural

period on S/y)/? is generally small, in

particular when r is -0.05 and 0.0. When r

is 0.1, Srdr seems to have slightly

smaller values at natural period of about

0.1-0.25 in Group I and 0.5-15 in Groups

II and IH. This may be related to the

predominant period of ground motions in

each soil condition. However from a

practical point of view, it may be

considered that the dependence of the

residual displacement ratio response

spectra on natural period is less

significant. It should be noted that the

$RDR °f eacn individual record depends

significantly on natural period as shown

in Fig. 5. The mean value m of residual

displacement ratio response spectra is

less sensitive to ductility factor when r is

0.0. However, m increases as ductility

factor increases at r larger than 0 , while

the opposite is true at r less than 0.

188
The standard deviation a of residual

displacement ratio response spectra

generally increases as the ductility factor

increases. When r is 0, o is almost

independent of the ductility factor, and is

about 0.3 regardless of the natural period

and ground condition.

(3) Effect of Bilinear Factor

As described previously, the bilinear

factor r is the most significant parameter

controlling the residual displacement ratio

response spectra SRr>R . Therefore it is

examined how SRpR depends on the

bilinear factor r in terms of the vs.

r relation. The effect of soil condition,

natural period and ductility factor on the

Srdr vs t relation is also re-examined

based on the Srqr vs. r relation.

Fig 11 shows the Srdr vs. r relation

taking the soil condition as a parameter.

The result for a combination of T=\s and

ju=4 is presented here, because other

cases exhibit similar results. While

is about 0.2-0.3 or less at rsO.l, it

increases sharply as r approaches zero. Srdr is almost 1.0 at r<z -0.05. Generally

Srdr at r>0 is slightly larger at the

Group I sites than the Group II and III

sites. Fig. 11 shows how important the

bilinear factor r is on the residual

displacement response spectra Srqr, in

particular SrdR at rsO.

Fig 12 shows the Srdr vs. r relation

taking the natural period as a parameter.

Because the dependence of the Srdr vs. r relation on the soil condition was less

significant as shown in Fig 11, SgjyR

averaged over all the soil conditions is

presented here. The general trend of the

Srdr vs r relation is the same with that

presented in Fig 11. Srdr at longer

natural periods is slightly larger when

r>0, while they are slightly less when

r < 0 . The dependence of the Srdr vs. r

relation on natural period is not

significant.

Fig 13 shows the Srdr vs. r relation

taking the ductility factor as a parameter.

Again the general trend of the Srdr vs. r

relation is the same as that presented in

Figs. 11 and 12. Srdr with fi-2 is

slightly larger than Srdr with fi =4 and

6 when r>0, and is slightly less when

r <0.

From Figs. 11-13, it is apparent that

the residual displacement ratio response

spectra SRDR is small when the bilinear

factor r is larger than about 0.05, and it sharply increases as r approaches zero. Srdr 1S almost unity when r is less than

1. This general trend on the Srqr vs. r

relation is not dependent on soil condition,

natural period and ductility factor.

5 Srdr proposed for seismic

DESIGN

Based on those findings, it was decided

to disregard the effect of natural period,

ground condition and ductility factor on

the residual displacement ratio response

spectra 5/^^ and to consider only the

effect of bilinear factor in evaluating the

residual displacement ratio response

spectra for seismic design. Because the

scatter of the residual displacement ratio

response spectra around the mean value is

considerable, one standard deviation from

the mean value is included. Fig 14 shows

the residual displacement ratio response

spectra Sj^ji thus evaluated, and this is

proposed to be used in design.

6. APPLICATION OF RESIDUAL

DISPLACEMENT RESPONSE

SPECTRA

189
The residual displacement response

spectra of Fig 14 can be used for any

structures where the hysteresis loop

shape may be reasonably approximated

by a bilinear model.

An example of the application of this

method is given for six reinforced concrete

bridge columns as shown in Table 2.

They were designed in accordance with

the equivalent static seismic design

method with the lateral force coefficient

of 0.2 and the allowable design approach

(JRA 1990, Kawashima and Hasegawa

1994b). The column height was 15m and

20m, and the deck length was 30m and

40m. The deck was assumed to be simply

supported plate girder with 7m (2 lanes)

wide. The columns were assumed to have

rectangular cross sections 7m wide in the

transverse direction. Shear span ratios

were either 7.1 or 9 in the longitudinal

direction. These columns were assumed

to be subjected to an earthquake

excitation in longitudinal direction. The

natural periods of columns assuming the

uncracked concrete stiffness ranged from

0.63 sto 1.11s.

If it is assumed here that the hysteresis

loops of the reinforced concrete columns

can be idealized by the Takeda model

(Takeda, Sozen and Nielsen, 1970), the

lateral force vs. displacement relationship

after the initial cycle may be bilinear as shown in Fig 15 with the initial stiffness

ki and the post-yield stiffness &2 It should be noted that although the original

Takeda model assumes trilinear hysteresis,

bilinear is assumed here by disregarding

the effect of cracking These parameters

may be evaluated based on the primary

curve of the cracked columns.

The primary curve was evaluated

here according to the Seismic Design

Specifications of Bridges of the Japan

Road Association (JRA, 1990), and the

yield force F^, ultimate force Fum ,

yield displacement , and the ultimate

displacement of the columns were

obtained as shown in Table 3. From those

values, the initial stiffness &lm and the

second stiffness ^

m on the primary

curve are obtained asp

F F , 1 um rym

k2m -

(5)

(6)

The stiffness of the bilinear loop k\

and &2 fr°m tne Takeda model

corresponding to the maximum

displacement = nu

y

are obtained

from Fig. 1 5 as

h = rki

where

r =

max

(7)

(8) — (9)

2f^m uym^l ~ ^max

^max TM {l + rmU*m -l)} (10)

The yielding displacement iiy and

ductility factor \i are

Uy

J" =

_

1 max

t*m uym

(JU>1) (11)

(12)

The coefficient a in Eq. (7)

represents the degree of stiffness

degradation, and is assumed here to be 0.5

(Priestley, MacRae and Tao 1993).

Residual displacements were

computed by arbitrarily assuming a

ductility factor \im of 10. Based on \i m

and the parameters for primary curve, the

bilinear factor r can be evaluated by Eq.

(9). Because the maximum residual

displacement wrmax and the residual

190
displacement response spectra Srdr by

Eq. (4) can be evaluated by Eq. (3) and

Fig 14, respectively, the residual

displacement developed at the top of

columns |^| is obtained as shown in

Table 4.

Table 4 shows that the bilinear factor r

for the columns is in the range of 0.21 to

0.25. The residual displacement ratio

response spectra from Fig 14 is

about 0.3 considering one standard

deviation from the mean value for safety

.

The residual displacement ur predicted

to occur at the top of the columns ranges

from 6cm to 16.9cm. Those correspond

to 0.18-0.20 of the peak displacement

!*muym > and are larger in the more flexible

columns with large .

It should be noted that reinforced

concrete columns which are subjected to a pulse type motion and which do not

deform approximately the same amount

in positive and negative directions may

have larger residual displacements because

its hysteresis loop does not match the

bilinear approximation as shown in Fig

15 (Takeda, Sozen and Nielsen 1970).

Also, structures with lower bilinear

factors r or eccentrically loaded reinforced

concrete cantilever columns are likely to

have significantly greater residual

displacements than those presented

herein.

7. CONCLUSIONS

For evaluating the residual

displacements developed at structures

subjected to large ground motions,

residual displacement response spectra

5/j£> and residual displacement ratio

response spectra Srdr were proposed,

and their characteristics were clarified

based on 63 components of free-field

ground accelerations. An application of

the residual displacement ratio response

spectra Srqr to cantilevered bridge

columns was also presented. Based on the

results presented herein, the following

conclusions may be deduced:

(1) The residual displacement ratio

response spectrum SRDR , which is

defined as residual displacement response

spectra Srd normalized by the maximum

residual displacement «rmax given by Eq.

(3), is almost independent of the

earthquake magnitude, the epicentral

distance of the acceleration data and the

soil conditions.

(2) The residual displacement ratio

response spectra 5^ depend

significantly on the bilinear factor r. The

residual displacement ratio response

spectra is small at r>0, but it sharply

increases as r approaches 0. It is almost

1.0atr<-0.05.

(3) The dependence of natural period,

soil condition and ductility factor on the

residual displacement ratio response

spectra Srdr vs. bilinear factor r relation

is not significant. A design value for the

residual displacement ratio response

spectra Srdr was proposed as a parameter of the bilinear factor r as shown in Fig 14, based on the averaged

value of the residual displacement ratio

response spectra for 63 ground

acceleration records. To consider the

significant scatter around the mean value,

one standard deviation from the mean

value was included in the proposed

relation.

(4) Residual displacements developed

at the crest of reinforced concrete bridge

columns were computed based on the

residual displacement ratio response

191
spectra proposed herein. The maximum

residual displacement at the pier crest is

16.9cm for a 20m high column.

REFERENCES

1) Housner, G. W. (1959), "Behavior of

structures during earthquakes." Proc.

ASCE, 85, Em 4, 109-129

2) Japan Road Association (1990), "Part

V Seismic design, Design specifications

of highway bridges."

3) Kawashima, K., MacRae, G. A. and

Hasegawa, K. (1992), "The strength and

ductility of steel bridge piers based on

loading tests" Journal of Research, 29,

Public Works Research Institute,

Tsukuba, Japan.

4) MacRae, G. A. and Kawashima, K.

(1993) "The seismic response of bilinear

oscillators using Japanese earthquake

records," Journal of Research, 30, Public

Works Research Institute, Tsukuba,

Japan.

5) Kawashima, K. and Hasegawa, K.

(1994a), "Experimental investigation on

nonlinear seismic response of bridge

columns and accuracy of equal energy

assumption." Proc. Japan Society of

Civil Engineering, 483/1-26, 137-146

6) Kawashima, K. and Hasegawa, K.

(1994b), "New seismic design

specifications of highway bridges in

Japan." Earthquake Spectra, 10 (2), 333-

356

7) Ministry of Construction. (1995).

"Report on damage of highway bridges

by the Hyogo-ken Nanbu Earthquake,"

Committee for Investigation on the

Damage of Highway Bridges Caused by

the Hyogo-ken Nanbu Earthquake,

Tokyo, Japan.

8) Priestley, M. J. N., MacRae, G. A. and

Tao, J. (1993) "P-A effect in seismic

regions," Structural Systems Research

Project Report, No. 93-5, University of

California, San Diego.

9) Takeda, T., Sozen M. A. and Nielsen

N. N. (1970), "Reinforced concrete

response to simulated earthquakes,"

Journal of Structural Engineering ASCE,

96-12, 2557-2573

192
Table 1 Definition of Ground Condition

Case Hpierht (hi) Shear Span

Ratio

Width (m) Span Length

(m)

Natural

Period (s)

1 15 7.1 2.1 30 0.63

2 15 7.1 2.1 40 0.75

3 15 9 1.67 30 0.91

4 15 9 1.67 40 1.00

5 20 7.1 2.8 40 0.86

6 20 9 2.22 40 1.11

Table 2 Bridge Columns Analyzed

Natural Period Ground

Condition

Geological Definition

TG<0.2s Group I Tertiary or older rock, or diluvium with H< 10m

0.2ssTq<0.6s Group II Diluvium with Ha 10m, or alluvium with H<25m

including soft layer with

H<5m

0.6s sTQ Group m Other than the above, usually

soft alluvium or reclaimed

land

Table 3 Yield Displacement, Natural Period and Bilinear Factor

Case uym (cm) Uum (cm) J>(kN) &lm (kN /cm) k2m (kN/cm)

1 2.94 53.5 1,217 1,429 414.0 4.18 0.0101

2 4.12 47.2 1,497 1,766 363.5 6.23 0.0171

3 6.03 50.0 1,099 1,271 182.2 3.92 0.0215

4 7.00 48.4 1,344 1,563 192.0 5.30 0.0276

5 4.98 70.7 1,780 2,159 357.4 5.77 0.0162

6 9.4 70.9 1,740 2,679 185.0 5.52 0.0298

Table 4 Residual Displacement Developed at Top of Columns

Case it, (kN/cm) u

y (cm) r "r max (cm) SRDR ur (cm)

1 5.80 130.9 5.07 0.21 19.26 0.31 6.0

2 5.48 115.0 7.52 0.22 26.16 0.31 8.0

3 5.30 57.6 11.38 0.23 37.54 0.30 11.3

4 5.07 60.7 13.82 0.25 42.36 0.29 12.3

5 5.52 113.0 9.02 0.22 31.76 0.31 9.7

6 4.99 58.5 18.85 0.25 56.30 0.30 16.9

193
Displacement u

Fig. 2 Iteration Required to Compute Target Ductility Factor

194
(a)r(^-l)<l

Uy Ur max

Displacement

at End of

Analysis

u

Umax

(b) Km -1)^1

Fig. 3 Maximum Residual Displacement

195
15M

C

o

13

o

o

<

10 20 30 40

Time (sec)

(a) Input Acceleration

50 60

30

Time (sec)

(b)r=-0.05

60

=t 4

Time (sec)

(d) r=0.1

Fig. 4 Residual Displacement for Itajima Record (7Ms)

196
0.1 0.2 0.5 1 2 3

Natural Period T(sec)

(a) Response Acceleration

Spectrum

0.1 0.2 0.5 1 2 3

Natural Period T(sec)

(b) r=-0.05

I

o

•nA

V

a

V

o

J2

p.ws

1'53v

1

0. 8

0.6

0.4

0.2

0

r 1

. .

1 1 1 1 . J T —p-

*-•//

/•' \ !A t

• 'V- \/' V- '

•

\\ V.' V i •

'

\\ .1' \ \ i \ -

\7 r A \

"

* — X 1 V \ r/ \ i i \ / *\ \ .

i / * * \ i \

\

• 1

\ /' *\*

v \-

i .. i .... i i i

00_o1

c

e

uo

J2s

13I

<&

i i r

0. 1 0. 2 0. 5 1 2 3

Natural Period T(sec)

(c) r=0

0.2 0.5 1 2 3

Natural Period T(sec)

(d) r=0.1

Fig. 5 Residual Displacement Ratio Response Spectra Srqr for Itajima

Record

197
6.5 6.75

Magnitude M

7 7.25 7.5 7.75

20

10

I

- 19

16

13 12

3

-

1

1

5 6. 75 7 7.25 7.5 7.75 8

Magnitude M

Fig. 6 Magnitude of Earthquake and Epicentral Distance of Ground

Accelerations Used in Analysis

30

20

20

17

_ 6

8

2 3 i i 2 3

"

(NTji(DOOOIN'^<0000 oooooooooo > I I ' 1 r—l r—I OJ

I I I I I I I I I I I oo cd oo o -^r co oo co —I '—' .—t '—I .—I CO

Peak Acceleration (gal)

Fig. 7 Peak Ground Accelerations

198
o

B

CD

I—%—I

CO

•a

'55

Oh

1

0.8

0.6

0.4

0.2

0

1 1

1

1 o

o

CD

LEGEND

6. 5£M<7.0

7.0^M<7.5

7.5£M

j i_ '

i

5 10 20 50 100 200 500

Epicentral Distance A (km)

(a) r=-0.05

o

•43

CD

CD

co

3

'53v

1

0.8

0.6

0.4

0.2

0

q 1 i—, |-r-ncnf GO | O

D

o

0 o

Q

LEGEND

-I 6.5£M<7.0

7.0£M<7.5

7.5£M

o <b

d L

*

*o

Q

5 10 20 50 100 200 500

Epicentral Distance A (km)

(b) r=0

o

(2

i

0.8

cdB 0.6

cdus

•a 0.2

CU

PS"

0

LEGEND

6.5£M<7.0

7.0^M<7.5

7.5s£M *

o ® a

o

„o D

5 10 20 50 100 200 500

Epicentral Distance A (km)

(c) r=0.1

Fig. 8 Effect of Earthquake Magnitude and Epicentral Distance on Residual

Displacement Ratio Response Spectra Srdr for T=ls and p. =4

199
Group I Group II Group m

V>

v2

0.1 0.2 0.5 1 2 3

Natural Period T(sec)

0. 1 0. 2 0. 5 1 2 3

Natural Period 7"(sec)

(a) r=-0.05

0. 1 0. 2 0. 5 1 2 3

Natural Period r(sec)

0. 1 0. 2 0. 5 1 2 3

Natural Period 7Xsec)

(b) r=0

| 0.8

Og 0.6

s

0.2

0

,

1 1

1 1

I III] | |

LECEND

li-2

i \

V'

/\ - ^

;

—i i_ i

1 i

u

a5

0. 1 0. 2 0. 5 1 2 3

Natural Period 7Xsec)

1

0. 8

0.6

0.4

0.2

0

i —

r

1 1-| m-r| 1 rLEGEND wZ

H--4

" \

W=6

\

1 \ /

-..

-u i_ '

1 L_

0.1 0.2 0.5 1 2 3

Natural Period T(sec)

| 0.8

toO

a o.6

"a>

5 0.4 uS

0.2

0

v

A » ~..._.

/ \ / \ V s

-

LECEND

U'-Z

u--4

. 1 i i

. i .... i i i

0.1 0.2 0.5 1 2 3

Natural Period 7(sec)

1

0.8

o

u 0.6 3

>

5 0.4 v2

0.2

n 1

'

i

1—r—

» \

I A

» \

V v' V

"

1\ * \

V /-./ \\ \ / * \ \ / *\ \ / * \

LEGEND u =2

V »1

r<

_//_=6_

i. . I .... 1 1 1

0.1 0.2 0.5 1 2 3

Natural Period T(sec)

| 0.8

«o

g 0.6

•a>

§

0.4 s

0.2

-n r—1——r-• i i i

LEGEND u-2

U--K

" --"\

\\ .'\

\

\

. \ - / \

'

v -

. i i

» V

i

.'' '

1 . . 1 1 1_

0. 1 0. 2 0. 5 1 2 3

Natural Period T(sec)

(c) r=0.1

Fig. 9 Mean Value of Residual Displacement Ratio Response Spectra Srdr

200
"> 0.6

S3 0.4

rt

2 0-2

Group I

1

LEGEND

0. 1 0. 2 0. 5 1 2 3

Natural Period T(sec)

0. 1 0. 2 0. 5 1 2 3

Natural Period T(sec)

"> 0.6

1 0.4 1- re "g2 0.2

LEGEND u-l lt z 4 U =6

0. 1 0. 2 0. 5 1 2 3

Natural Period T(sec)

Group II

3

"> 0.6 LEGEND

0.1 0.2 0.5 1 2 3

Natural Period T(sec)

(a) r=-0.05

0.2 0.5 1 2 3

Natural Period T(sec)

(b) r=0

Group DI

w " 0. 1 0. 2 0. 5 1 2 3

Natural Period T(sec)

I

to

C5O

•a

revQ

0.6

0.4

0.2

0

0. 1 0. 2 0. 5 1 2 3

Natural Period T(sec)

0. 1 0. 2 0. 5 1 2 3

Natural Period T(sec)

0. 1 0. 2 0. 5 1 2 3

Natural Period T(sec)

(c) r=0.1

Fig. 10 Standard Deviation of Residual Displacement Ratio Response Spectra

SRDR

201
Bilinear Factor r

Fig. 1 1 Effect of Ground Condition on the Residual Displacement Ratio

Response Spectra SRDR vs. Bilinear Factor rfor T=ls and ju=4

0

I 1 1 1 1 L_

-0.2 -0.1 0 0.1 0.2 0.3

Bilinear Factor r

Fig. 12 Effect of Natural Period on the Residual Displacement Ratio

Response Spectra Srjjr vs. Bilinear Factor r

202
0

I 1 1 1 1 1_

-0. 2 -0. 1 0 0. 1 0. 2 0.

3

Bilinear Factor r

Fig. 13 Effect of Ductility Factor on the Residual Displacement Ratio

Response Spectra Srdr vs. Bilinear Factor r

Fig. 14 Residual Displacement Ratio Response Spectra Srqr vs. Bilinear

Factor r Proposed for Design

203
o 1 max

£

Fym

4m

max l^trMym Uum

Displacement u

Fig. 15 Hysteresis Loop for Reinforced Concrete Columns

204
ANALYSIS OF DAMAGED FOUNDATIONS OF HIGHWAY BRIDGES

IN THE 1995 HYOGO-KEN NANBU EARTHQUAKE

by

Micbio Okahara 13

, Masanori Nakano 2 5

, Yoshitomi Kimura s)

ABSTRACT

A bore hole television system was used to survey damage to the foundations of road bridges

caused by the Hyogo-ken Nanbu Earthquake. The

foundation bodies of some bridges were cracked, but no foundation subsidence, steel reinforcement

rupture, nor cover concrete separation damage

occurred. But residual displacement was found in some pile foundations near the shoreline where

liquefaction caused ground flow. Loading tests were

performed on piles which cracked, the results were

analyzed, and the analytical results confirmed that

the damage to the pile bodies had only a slight effect on the horizontal strength of the foundations.

KEYWORDS : Earthquake damage, Hyogo-ken

Nanbu Earthquake, Highway bridges, Foundations,

Horizontal strength

1. INTRODUCTION

The Hyogo-ken Nanbu Earthquake of January

17, 1995 caused extensive damage to road bridges:

collapsed bridge piers, fallen bridges, etc.

1

5

. It was

impossible to survey bridge foundation damage

immediately after the earthquake because the

foundations are underground, but the state of damage

to the foundations was surveyed in conjunction with

restoration work. At the result, it was confirmed that a number of foundations were cracked or suffered

horizontal residual displacement

This paper presents an outline of the state of

damage to foundations of road bridges based on the

results of the foundation survey and describes

characteristics of the damage. It also reports on

vertical loading testing, horizontal loading testing,

and pile bending testing performed using damaged

pile foundation to clarify the effect of the pile body

damage on the horizontal strength of the foundations.

2. STATE OF THE DAMAGE TO BRIDGE PIER

FOUNDATIONS

2.1 Survey of the Damage to Bridge Pier

Foundations

A survey of the damage to bridge pier

foundations was conducted on the No.3 Kobe Line

and No.5 Wangan Line of the Hanshin Expressway,

at the Hamate Bypass of National Highway No.2,

and on the Meishin and Chugoku Expressways. But

the survey of the Meishin and Chugoku Expressways was limited to seven viaducts where the damage to the bridge piers was relatively severe (Onishi,

Mizudo, Moribe, Tsuto, and Kawaraginishi Viaducts on the Meishin Expressway and the Takarazuka and

Kouda Viaducts on the Chugoku Expressway).

Figure 1 shows location of the surveyed bridges.

During the planning of the survey on the

Hamate Bypass on National Highway No.2, the No.3

Kobe Line of the Hanshin Expressway, the Meishin

Expressway, and the Chugoku Expressway, the

bridge piers actually surveyed were selected

considering the structure of the piers, the degree of

damage to each one, local factors, and so on, so that

the survey would not be concentrated on a narrow

range of pier categories or degrees of damage. On

the other hand, on the No.S Wangan Line of the

Hanshin Expressway on the other hand, the bridge

piers selected for the survey were mostly at locations

where liquefaction caused substantial ground flow.

Table 1 shows the number of bridge piers in the

survey area and the number that were actually

surveyed. Most of the bridge pier foundations

surveyed were pile foundations (cast-in-place pile

foundations). It was assumed that spread foundations

1) Research Coordinator for Advanced Technology,

Planning and Research Administration, Public

Works Research Institute, Ministry of

Construction.

2) Head, Foundation Engineering Division, Structure

and Bridge Department, ditto.

3) Senior Research Engineer, Foundation

Engineering Division, Structure and Bridge

Department, ditto.

205
would suffer little damage because they are constructed on good quality ground and evidence was not found in the surrounding ground that these

foundations had undergone large deformation, but as a precaution, 17 spread foundations were included in

the survey. Most of the caisson foundations surveyed were subjected to the effects of ground flow caused

by liquefaction and those supporting bridge piers

with large residual horizontal displacement Five

diaphragm wall foundations were selected for

inclusion in the survey, because there are few

foundations of this kind in the survey area and even where ground flow occurred in the ground

surrounding them, almost no residual horizontal

displacement appeared in the bridge piers.

The survey of the spread foundations was

performed by excavating the ground to the top

surface of the footing to visually check for cracking

of the footing. Because it would have been difficult

to visually survey the pile foundations (cast-in-pile

foundations), caisson foundations, and diaphragm

wall foundations for cracking by digging down to each foundation, the survey was done with a bore

hole television system To do this, core boring was

performed from the top of the footing into the pile or body of the foundation, and a borehole camera was

lowered into this hole to inspect the interior of the

foundation for cracking. The width of cracks

observed by a bore hole television system have been

assessed as more than three times the width of the

cracks found when the foundation was actually

excavated and inspected visually, because during the

boring of the concrete, concrete around the cracks is removed, and the illumination used during camera

observations produces shadows around the cracking.

For the final damage assessments, the actual crack

width was considered to be about 1/3 the width

found using the bore hole television system (see

Photograph 1). Some of the bridge piers on the No.S Wangan

Line of the Hanshin Expressway were found to have

residual horizontal displacement caused by ground

flow around the bridge pier under the effects of

liquefaction. For this reason, after the earthquake, a

survey of bridge pier positions was carried out using

the artificial satellite system called the Global

Positioning System (GPS) and the findings were compared with management documents to assess the

residual horizontal displacement of the bridge piers.

The method used to assess the residual horizontal

displacement of the bridge piers is discussed in detail

as part of 3.

2.2 Damage to Bridge Pier Foundations

1) Damage to Pile Foundations

The degree of damage to pile foundations is categorizes in four ranks from rank a to rank d, as shown in Table 2. The pile foundations did not suffer severe settlement, pile body failure, or failure of their

reinforcement for which would effect their overall

stability, some of them suffered residual horizontal

displacement caused by ground flow under the

effects of liquefaction and bending cracking of pile

bodies.

Table 3 presents the damage to pile foundations on each road surveyed. It reveals that pile

foundations suffered the most serious damage on

No.5 Wangan Line of the Hanshin Expressway, with

about 11% of those surveyed on this road suffering

damage assessed as rank b. This damage is assumed

to have been caused by liquefaction or by the ground

flow generated by liquefaction. On roads other than

the No.S Wangan Line of the Hanshin Expressway,

only minor damage in rank c and d was discovered.

2) Damage to Spread Foundations

No cracking or other damage was found on the

17 spread foundations surveyed on the No.S Kobe

Line of the Hanshin Expressway and the Chugoku

Expressway.

3) Damage to Caisson Foundations

Eight caisson foundations were surveyed on the

No.5 Wangan Line of the Hanshin Expressway, two on the No.3 Kobe Line of the same expressway, and

five on the Hamate Bypass of the National Highway

No.2. On the No.S Wangan Line of the Hanshin

Expressway, one caisson foundation located near a

canal was damaged; it suffered bending cracking

equivalent to rank c damage to a pile foundation

This damage is assumed to have been caused by the

relatively large residual horizontal displacement of

the foundation and ground deformation caused by

ground flow in the surrounding ground. But aside

from this one case, the caisson foundations were undamaged, or if they were damaged, it was

extremely minor.

4) Damage to Diaphragm Wall Foundations

The results of a survey of five diaphragm wall

foundations on the No.5 Wangan Line of the

206
Hanshin Expressway revealed that the foundation

bodies suffered bending cracking equivalent to rank

d damage to a pile foundation, but this damage was

very slight.

3. CHARACTERISTICS OF THE DAMAGE TO

BRIDGE PIER FOUNDATIONS

3.1 State of Crack of Pile Bodies

The No.5 Wangan Line is constructed on

reclaimed land with a reclaimed stratum thickness

ranging from 9 m to 22 m. The Hyogo-Ken Nanbu

Earthquake caused liquefaction in the ground

surrounding many foundations and shifted the ground

near canals towards the canals. These events resulted

in large residual horizontal displacement of bridge

piers and bending cracking of pile bodies of pile

foundations.

Figure 2 shows examples of cracking of pile

bodies. Sample (a) shows the widest crack

discovered during the survey. This pile body crack was 4mm wide at its widest point, and many of them

were found within the reclamation soil stratum,

particularly in the top of the pile. Some cracks also

occurred in the alluvial clay stratum. A study within a range of 5 m from the bottom surface of the

footing reveals that the sub-total of the crack width

equals 8.6 mm(4+1.3+1.3+1.0+1.0), and when

calculated in terms of the unit length of the pile, it is

1.7 mm/m. Considering the fact that the diameter of

the piles is 1.5 m, this can be considered as small

damage. Figure 2 (b) shows the pile foundation for

the bridge pier with the largest residual horizontal

displacement. Within a range of about 6 m from the

bottom surface of the footing, there are 12 cracks

with a width of 0.5 mm or greater, but the widest of

these cracks is rather small, at only 2 mm.

Table 4 shows the results of the mode of cracks

in 17 pile foundations with the rank b degree of

damage, the severest rank observed in all the pile

foundations surveyed on the No.5 Wangan Line. In

the reclaimed stratum, all 17 foundations were

cracked. The average values for the 17 foundations

are 10 cracks, crack width of 1.5mm, and a

reclaimed stratum thickness of 14 m. The damage to

these foundation piles can be defined as extremely

slight. Cracks also appeared in the alluvial clay

stratum, but far fewer than in the reclaimed stratum.

3.2 Residual Horizontal Displacement of the Piers

Along the No.5 Wangan Line, liquefaction

accompanied by ground flow occurred in the ground

around bridge piers near the shoreline, causing

considerable residual horizontal displacement to the

bridge piers. For this reason, the analysis included a

study and clarification of the relationship of the

horizontal displacement of the ground around the

bridge piers with the residual horizontal displacement

of the bridge piers.

The horizontal displacement of the ground was

found using the results of an aerial photography

survey carried out by Hamada et al.

23 They used

three survey stations established on Rokko Island as

control points for the aerial photography survey, but

for this study, six survey stations established for the

survey of the bridge pier locations for the No.5

Wangan Line of the Hanshin Expressway were used

as the control points. As examples, Figure 3 and

Figure 4 present results showing the horizontal

displacement of the ground and the residual

horizontal displacement of the bridge piers of the

Rokko Island Bridge and its approach that connect

with Rokko Island and Uozaki Hama. While the

aerial photography survey of Hamada et al. found the

displacement vectors of manhole covers and other

fixed objects on the ground surface, in addition to

that, this survey found the displacement vectors at

intersection points of a grid crossed at intervals of 20

m The residual horizontal displacement of bridge

piers, found from another survey performed after the

earthquake, represents the displacement of bridge

pier body at approximately 1 m height above the

ground.

When the horizontal displacement vectors of the

ground and bridge piers are found in this way,

almost along No.5 Wangan Line they include about

20 cm displacement evenly in the south-west

direction. And triangulation point near the No.5

Wangan Line also shifted the same distance to the

south-west This displacement can be interpreted in

various ways: as ground displacement caused by

fault movement or as a wide-area ground movement,

but because uniform displacement occurred over a

certain range, the horizontal displacement of the

ground and the residual horizontal displacement of

the bridge piers were modified after deducting this

value. The horizontal displacement vectors found in

the way include a certain degree of error because of

differences in the precision of the photographic

surveys performed before and after the earthquake,

and of variations in the way the aerial photographs.

For this analysis, the precision of the measurement of

207
the horizontal displacement of the ground is

estimated to be ± 30cm, and the bridge pier

residual horizontal displacement measurement

precision is estimated to be ± 10cm

Figure 5 shows the relationship of the residual

horizontal displacement of the bridge piers with the

horizontal displacement of the ground around the

bridge piers in Rokko Island, Uozaki Hama, and

Fukae Hama. The results are considerably scattered,

but overall, the residual horizontal displacement of

the bridge piers tends to increase as the horizontal

displacement of the surrounding ground rises. But

the extent of the displacement of bridge piers which

occurs when certain ground displacement occurs

varies according to the rigidity of the foundation, and

the residual horizontal displacement is smaller in the

case of caisson foundations or diaphragm wall

foundations: types with greater stiffness than pile

foundations.

4. HORIZONTAL STRENGTH OF DAMAGED

PILE FOUNDATIONS

4.1 Pile Loading Tests To confirm the effect of pile damage on overall

pile foundations, loading testings were performed on

piles of the No.3 Kobe Line of the Hanshin

Expressway. The degree of damage to the pile

foundation was assessed as rank c damage:a crack

with a width of about 2 mm was found near the top

of the pile as shown in Photograph 2 but its concrete was not crushed and its reinforcement bars did not

swell out

The loading testing included three tests : pile

body bending.vertical loading,and horizontal loading

tests. Three of the 19 piles were cut for use as test

piles. Specimens were taken from these cut test piles

for use in compressive strength testing of the

concrete cores and for tensile strength testing of the

reinforcement bar. Table 5 shows the results of the

compressive strength testing of the concrete cores. The average value of the strength of the concrete of

the pile bodies was about 1.8 times the design

strength of 240kgf/cm 2 and represents considerable

strength. Table 6 presents the results of the tensile

testing of the reinforcement bar. The results, that the

yield points of three of the five reinforcement bar are

not clear, shows that these reinforcement bars

received a stress history that exceeds the yield point.

An examination of arrangement of the reinforcement

at the cut section of the piles bodies showed they

were as shown as designed

Pile body bending testing was performed on the

pile tops where cracking appeared as shown in

Figure 6. Test piles were obtained by cutting the pile

bodies at a depth of 2.8 m below the bottom surface

of the footing, then horizontal loading was done at a

depth of 2.5 m below the footing using adjacent piles

as the reaction piles. The loading, which was

one-direction multi-stage loading, was performed up

to 25 tf, equivalent to about 60% of the calculated

yield bending moment of the tops of the piles. Figure

7 shows the relationship of the horizontal load with

the horizontal displacement at the loading point. In

order to grasp the effects of the decline in stiffness

of the top of the piles,the relationship of the

horizontal loading with the horizontal displacement

of the piles when the pile tops are assumed to be

sound was found through analysis, and this was

compared with the results of the loading testing. The

analysis of a sound pile referred to here was

performed using the stress-strain relationship of

concrete indicated in the "Guide Specifications for

Reconstruction and Repair of Highway Bridges

Which Suffered Damage due to the Hyogo-ken

Nanbu Earthquake" (Ministry of Construction,

Feb. 24, 1995), and in this analysis the horizontal

displacement by pulling the main reinforcement bars

out of the footing is included. According to the

material testing described above, the yield strength of

the reinforcement bar and the compressive strength

of the concrete in the pile body were assessed as 3,750kgf/cm 2 and 425kgf/cm 2 respectively. The

horizontal displacement of the loading testing is

greater than the analytical value for the same

horizontal load when the pile tops are considered to be sound, and the stiffness of the pile tops declines.

To perform the vertical loading testing and the

horizontal loading testing, two piles were cut at a

depth of 1.2 m from the bottom surface of the

footing, and the piles below this level used as the

specimens for the two tests. Vertical loading was

applied to the top of one pile and horizontal loading

to the top of the other. Figure 8 shows the amount of

loading - amount of settlement relationship for the

vertical loading testing. It confirmed a bearing

capacity as high as 450 tf : a value exceeding the

design ultimate bearing capacity of 436 tf calculated

by the present Specifications for Highway Bridges.

The pile top settled 1.3 cm much less than 10cm,

which is 10% of the pile diameter, and the amount of

208
settlement usually forecast when calculating the

ultimate bearing capacity of piles, which confirms

that the ultimate bearing capacity of the pile exceeds

the design ultimate bearing capacity. Figure 9

presents the amount of loading - horizontal

displacement curve obtained from the horizontal

loading testing. This figure shows that the value of

the coefficient of ground reaction in the horizontal

direction found assuming that the horizontal

displacement is 1% of the pile diameters is 1.9 kgf/cm 3

, which coincides closely with 2.3kgf/cm 3

, the coefficient of ground reaction in the horizontal

direction estimated from the N value obtained from a

standard penetration test according to Specifications

for Highway Bridges.

4.2 Evaluation of the Strength of the Pile

Foundations

In order to evaluate the effect of the earthquake

damage to piles on the behavior of overall

foundations, the foundations which underwent the

loading testing were used as the object of an analysis

performed to find the resistance properties of pile

foundations employing a method which accounts for

the non-linear behavior of foundations. The

analytical model is shown in Figure 10. The resistance properties in the axial direction

of a pile are modelled as a bi-linear model with the

ultimate bearing capacity and the ultimate pull-out

force as the upper limit values by treating the axial

direction spring constant Kv of the pile found from

the vertical loading testing as the initial gradient The

ground resistance in the direction perpendicular to the axial direction of the pile is modelled as a

bi-linear model with the upper limit p hu and the

initial gradient km calculated from the following

formulae.

kHd=77n*ak«k.H (1)

p hu = r} P

• a p

• p u (2)

Where:

kH<i: Design horizontal subgrade reaction

coefficient (kgf/cm *

)

p hu : Upper limit value of the horizontal

subgrade reaction (kgf/cm

2

) k h : The horizontal subgrade reaction coefficient

(kgf/cm 5 ) during the earthquake, it is found

from the Specifications for Highway

Bridges. Here it is set based on the

horizontal loading testing of the pile.

p u : The passive earth pressure strength (kgf/cm

2 ) during the earthquake, it is found using

Coulomb's passive earth pressure coefficient

based on the ground constant obtained from

the soil testing. Here it considers the weight

of the soil up to the ground surface to be the

overburden load.

a k : Corrective coefficient of the horizontal

subgrade reaction coefficient of a single pile

a P : Corrective coefficient of the upper limit

value of the horizontal subgrade reaction of a single pile

rj k : Corrective coefficient of the horizontal

subgrade reaction coefficient accounting for

the pile group effect

rj p : Corrective coefficient of the upper limit

value of the horizontal subgrade reaction

accounting for the pile group effect

The values used for a * and a P are shown in

Table 7. These were found from an analysis of the

results of horizontal loading testing of existing single

piles.

The corrective coefficient of the horizontal

subgrade reaction coefficient accounting for the pile

group effect T) k used was the following value.

V * = 2 / 3 (3)

The corrective coefficient of the upper limit

value of the horizontal subgrade reaction accounting

for the pile group effect r) p used was the following

value.

Cohesive Soil : rj „ = 1.0 (4)

Sandy Soil :

Tj P

• a P = Center interval between the

piles in the direction

perpendicular to the loading /

pile diameter ( ^ a P ) (5) In the case of piles other than the front row of

piles in sandy ground, it is 1/2 of the value shown in

formula (5). The value of rj k and 7} P for these

piles was found based on the results of the analysis

of loading testing of pile groups.

The ground resistance and bending stiffness of

the piles were estimated based on data obtained from

the loading testing and material testing described

above. Because the bending test of the pile indicates

that as shown in Figure 9, earthquake reduces the

bending stiffness of the pile top, the bending stiffness

of the pile top is reduced as shown in Figure 1 1 to a

value about 2/3 of that when the pile top is assumed

to be sound, so that analytical result of the horizontal

load - horizontal displacement relationship of a pile

will be equivalent to the bending testing results. Figure 12 shows the results of the above

209
described analysis of the horizontal load - horizontal

displacement relationship of the overall pile

foundation. A comparison of this relationship when

the pile has not been damaged with the relationship

in a case where the pile top is assumed to have been

damaged reveals that in the damaged case, the

horizontal displacement under the same horizontal

loading is slightly larger, but that the maximum

horizontal bearing strength of the pile foundation

hardly declines at all, demonstrating that damage to

the pile body has only a slight effect on the overall

pile foundatioa

5. CONCLUSIONS

The results of survey of the damage to bridge

pier foundations by the Hyogo-ken Nanbu

Earthquake have been studied and organized and an

outline of the damage and its characteristics

prepared. Loading testing was performed on cracked

piles to clarify the effects of the damage to the pile

bodies on the horizontal strength of the foundations.

The main results obtained from these survey are summarized as follows:

(1) The foundations did not suffer severe settlement,

pile body failure, or failure of their reinforcement

for which would effect their overall stability.

(2) Near the channels where liquefaction caused

ground flow, some foundations suffered horizontal

residual displacement There was little residual

displacement in the case of caisson foundations,

diaphragm wall foundations, with a high degree of

stiffness, even those at locations where equally severe ground flow occurred.

(3) Vertical loading testing, horizontal loading

testing, and pile bending testing were performed on damaged pile foundatioa The results revealed

that despite a decline in stiffness at the tops of the

cracked piles, the earthquake did not effect the

vertical bearing strength or horizontal resistance of

the piles.

(4) In order to evaluate the effect of the damage to

pile bodies on the behavior of overall foundations,

the foundation which underwent the loading

testing were used as the object of an analysis

performed to find the resistance properties of the

pile foundations. When the damage to the pile

bodies was accounted for, the horizontal

displacement produced by the same horizontal load

increased slightly, but there was almost no decline

in the maximum horizontal strength of the pile

foundatioa

In order to perform structural design accounting

for an earthquake as severe as the Hyogo-ken Nanbu

Earthquake, it is essential to study the displacement

properties and dynamic strength of foundations

during earthquakes. A method of evaluating

displacement properties and dynamic strength during

earthquakes that accounts for the non-linear

properties of ground resistance must be developed.

This paper is quote from the Report on the

Damage of Highway Bridges by the Hyogo-ken

Nanbu Earthquake, Committee for Investigation on

the Damage of Highway Bridges by the Hyogo-ken

Nanbu Earthquake 65

.

REFERENCE

1) Djima,T., Nakaoka,T. and Koga,Y. : Infrastructure

Damage by the 1995 Hyogoken Nanbu

Earthquake, UJNR Wind and Seismic Effects

Proceedings of the 27th Joint Meeting, 1995

.

2) Hamada,M, IsoyamaJR. and WakamatsuJC : The

1995 Hyogoken-Nanbu (Kobe) Earthquake,

Liquefaction, Ground Displacement and Soil

Condition in Hanshin Area, Association for

Development of Earthquake Predictioa The

School of Science and Engineering, Waseda

University, Japan Engineering Consultants, 1995.

3) Ministry of Construction : Guide Specifications

for Reconstruction and Repair of Highway Bridges

which Suffered Damage due to the Hyogo-ken

Nanbu Earthquake, 1995.

4) Japan Road Association : Specifications for

Highway Bridges, 1994.

5)Ministry of Construction : Report on the Damage

of Highway Bridges by the Hyogo-ken Nanbu

Earthquake, Committee for Investigation on the

Damage of Highway Bridges by the Hyogo-ken

Nanbu Earthquake, 1995.

210
Table 1 Number of Bridge Piers in the Survey Area and Number of Bridge Piers Surveyed

(Number of Bridge Piers)

Road Category

Type of Foundation

Spread Totals

Foundation

Caisson

Foundation

Pile

Foundation

Diaphragm

Wall

Foundation

No.3 Kobe Line

Hanshin

Expressway

All Bridge

Piers in the

Area Surveyed

44/ 4%1 929/84%1 OC 0%"> 1106 (100%)

Bridge Piers

Surveyed

12 2 109 0 123

No.5 Wangan

Line, Hanshin

Expressway

All Bridge

Piers in the

Area Surveyed

\J( u /o) J4\ l J /o) Ofifi/JMO/A zou^o i/o) 13/ ij\ 4%1 1/0) 345 (100%)

Bridge Piers

Surveyed

0 8 153 5 166

National

Highway No.2,

Hamate Bypass

All Bridge

Piers in the

Area Surveyed

n/\J\ o%\ \J /O) 1 <J\£ so 1I/O) %\ J 1 \ 1 y /O) 72 (100%)

Bridge Piers

Surveyed

0 5 20 0 25

Meishin

Expressway

and

Chugoku

Expressway

All Bridge

Piers in the

Area Surveyed

152(22%) 0( 0%) 532(78%) 0( 0%) 684 (100%)

Bridge Piers

Surveyed

5 0 21 0 26

Table 2 Categorization of Damage to Pile Foundation

Degree of

Damage

Definition

a Settlement of the foundation accompanied by considerable residual horizontal displacement.

b

Considerable residual horizontal displacement of the foundation.

Bending cracking of the pile body.

c Minor bending cracking of the pile body.

d No damage to the pile or slight bending cracking.

Table 3 Degree of Damage to Pile Foundations (Number of Bridge Piers)

Road

Degree of Damage

a b c d m

No3.Kobe Line,

Hanshin

Expressway

0 ( 0%) 0 (0%) 17 (16%) 92 (84%) 109 (100%)

No.5 wangan

Line, Hanshin

Expressway

0 ( 0%) 17 (11%) 57 (37%) 79 (52%) 153 (100%)

National

Highway No.2,

Hamate Bypass

0 ( 0%) 0 (0%) 10 (50%) 10 (50%) 20 (100%)

Meishin Expressway

and

Chugoku Expressway

0 ( 0%) 0 ( 0%) 0 ( 0%) 21 (100%) 21 (100%)

211
Table 4 Occurrence of Cracking in Pile Bodies at 17 Pile Foundations With Damage in Rank b

(No.5 Wangan Line of the Hanshin Expressway)

Stratum

Number of Bridge

Piers with Cracking

Average Stratum

Thickness (m)

Average Number of

Cracks in the Stratum

Average Crack

Width (mm)

Reclamation

Stratum

1 7 1 4 1 0 1 . 5

Alluvial Clay

Stratum

3 6 3 1 . 2

Table 5 Results of the Compressive Strength Test of the Concrete Core of the Pile Tops (No. 126 Pier)

Core No. Compressive Strength (kgf/cm

2 )

1 3 9 6

2 4 2 9

3 4 5 0

Average 4 2 5

Value

Table 6 Result of the Tensile Test of the Reinforcement Bars (D29) of the Pile Tops (No. 126 Pier)

Number of Yield Strength Tensile Strength

Reinforcement Bar (kgf/cm

2 ) (kgf/cm

2 )

1 3,860 5,790

2 3,640 5,390

3 No yield point 5,540

4 ir 5,350

5 n 5,350

Average 3,750 5,480

Standard Values >3,000 4,900 ~ 6,300

*

* ) Predicted value in current JIS standard is 4,500 to 6,100 kgf/cm 2

Table 7 Corrective Coefficient

Type of soil Layers a k a P

Sandy Soil 1 . 5 3 . 0

Cohesive Soil 1 . 5 1 . 5

Note: a P =1.0 in Soft Layers with N 2

212
213
214
256 259 280

243

335

23/

L

0

I 100m

Numerical values at the tips of arrows

= Horizontal displacement of the ground (cm)

Numerical values in the circles = Residual horizontal displacement of the bridge piers (cm)

Figure 3 Horizontal Displacement of Ground and Residual Horizontal Displacement of Bridge Piers at Rokko Island

(Horizontal Displacement of the Ground Obtained by Modifying the Results of a Survey by Hamada et al.

2

' )

215
E

to

216
200

E

o

&>

U

00

E

w

o

JO

Q.

v>

ca

oX

150

100

50

O Pile FoundMions

| Cliuon Found*torn

0 Ditphnfin Will Found»ltoro

o o

ft °

o o °a °

CD ($0

° o o0 o Oo<?

o cr

o o

0 50 100 150 200 250

Horizontal Displacement of Ground Around Bridge Piers (cm)

Figure 5 Ground Horizontal Displacement - Bridge Pier Residual Horizontal Displacement Relationship (For Bridge Piers

on the No.5 Wangan Line of the Hanshin Expressway at Rokko Island, Uozaki Hama, and Fukae Hama)

217
-c

>

Testing Pile

CSJ

CO'

/

Loading

—I I I Reaction Pile —

u

Figure 6 Pile Bending Test (No. 126 Pier)

80

70

60

1

50

i 40

I 30

s

20

10

0

0.0

Analytical Value When the Pile Top

- is Assumed to be Sound -

0. 5. 1.0 1.

5

Horizontal Displacement (cm)

2.0

Figure 7 Horizontal Load - Horizontal Displacement at the Loading Point During Pile Bending Testing

(No. 126 Pier)

218
Amount of Loading (tf)

100 200 300 400 500

I 0.4

cv

E

jo

sV

C/3

c

3

O

E

<

Figure 8 Amount of Loading - Amount of Settlement Curve During the Vertical Loading Test of the Piles (No. 126 Pier)

50

40

00c

'•Bes J3 30

(**o

c3

I 20

10

0

Calculated Value by Curt

Specifications for Highwa

k H = 2.3kgf/cm3 —

y bridge^ k g -L9k«f/cm3

;

J

/ / / / -S

f //

r

' / •/ /

//

f/7

f \i . fl /

II /

1 • / '

r » '

1*

90 II

' 1

f/7: /

, /

' /

"

Trh T

• II*// 1 * /

' (\M 1 ' f '

\ r r I * r I- I-

/ * * / ' ' r

'

0.0 1.0 2. 3.0

0

Horizontal Displacement (cm)

Figure 9 Amount of Loading - Horizontal Displacement Curve for the Horizontal Loading test of the Piles

(No. 126 Pier)

219
220
80

70

u

Horizontal Displacement (cm)

Figure 1 1 Horizontal Load - Horizontal Displacement Curve at the Loading Point During the Pile Bending Test

(No. 126 Pier)

Ho

1

.§uOX

2000

1500

1000

500

10 20

Horizontal Displacement 6 (cm)

30

Figure 12 Effect of Pile Body Damage on the Horizontal Load Horizontal Displacement Curve of the Foundation

(No. 126 Pier)

221
(a) Bore Hole Television System (b) Excavating

Photograph 1 Survey the Damage to Foundations

Photograph 2 Damage to Pile Body (No. 126 Pier)

222
Evaluation of Fault Rupture Geometry on Design Ground Motions

by

D. R. H. O'Connell and J. P. Ake

U. S. Bureau of Reclamation

Seismotectonics and Geophysics Group

Denver Technical Center

Denver, CO

ABSTRACT

Simulations of complex fault rupture and associated near-field ground motions and existing strong

motion recordings are used to illustrate the importance of fault rupture geometry on near-field

ground motion estimates for engineering applications. This study confirms other recent investigations of near-field ground motions from the 1989

Loma Prieta, 1 994 Northridge, and 1 995 Kobe

earthquakes: in the near-field, directivity and

focal mechanism can be more important than distance in controlling engineering ground motions at a particular site. Specifically, observed updip nearfield peak velocities exceeding 100 cm/sec from

thrust faults are a direct consequence of coincident

directivity and maximal Sv radiation patterns.

1.0 INTRODUCTION

One goal of strong-motion seismology has been

the prediction of ground motions for engineering

applications. Our task is to predict the site-specific

ground motions in advance of the occurrence of a

future earthquake given some knowledge of the

regional geology and the specific fault that may

rupture. Several recent earthquakes have provided

extensive data that expands our understanding of

some of the important physical phenomena. In this

paper we will utilize a complex fault rupture simulation technique to develop estimates of ground

motions for sites in the near-field of the 1994

Northridge earthquake and compare those results

to observed strong-motion data. Using this methodology we will then illustrate the prediction of

ground motions at a site near Ventura, California

for future earthquakes on the Red Mountain fault.

2.0 Methodology

To synthesize near-field ground motions from a

finite rupturing fault we use a kinematic approach

(Spudich and Archuleta, 1987). This method

requires specification of the temporal rupture characteristics on the fault. Ground velocity, dU/dt,

depends on the convolution of the

U

time evolution of the slip time function. S,-.-, and

the Green functions, Gy.-, the impulse responses

between the site of interest and positions on the

fault as described by equation 1 (see Figure 1 ). We

assume that fault rupture initiates at some point on

the fault (the hypocenter) and proceeds outward

along the fault surface. The pertinent parameters

are the fault length and width, the direction of slip

(rake), the hypocenter position on the fault, the

position of receiver site relative to the fault, the

slip time function at each point on the fault rupture, the velocity at which the fault ruptures away

from the hypocenter, and the velocity at which the

fault rupture stops along the rupture surface. The

FATT-VSDASP-RRH model (described below) is used to generate the fault parameters.

The reflectivity method of Kind (1978) is used to calculate Gy and the tau-p frequency domain distance interpolation method of Henry and others

(1980) and Kappus and others (1990) is used to obtain Gy at required source-receiver distances.

Approximate depth interpolation between Green

functions consists of increasing the distance by a small amount to increase the S-P time of the shallower Green function. Since Green functions are spaced 0.1 km apart in depth, this is a reasonable

approximation.

The computation demands scale as f

5

, where/is

frequency, effectively limiting the maximum feasible frequency for the calculations to 1 0 Hz. Consequently, this method is most applicable to the S223
wave portion of ground motion, the method cannot

reproduce the frequency bandwidth of typical high

frequency P-wave phases.The method is unlikely

to produce realistic vertical peak ground accelerations that are controlled by S->P converted phases

that typically contain substantial spectral content

above 10 Hz.

The Kostrov slip function provides a realistic

description of dynamic slip on complex faults and

is commonly used in rupture simulations (Spudich

and Archuleta, 1987). We use the approximation

to the Kostrov slip function of O'Connell and Ake

(1995), s(r,t). This slip velocity function represents the dynamic solution for a circular constant

stress-drop rupture in a homogeneous medium.

Consequently it is designated a "quasi-dynamic"

slip function for more general rupture geometries

and material properties.

The velocity at a site relative to the fault rupture is calculated in the frequency domain as,

n,m

it(x,v)= Y s • .(r,co)G . .(r,(D)R](r)D (r,co) (2)

y y y y

ij=0

where i

;y(r,co) are the frequency-domain equivalents of s(r,t), G(y(r,co) are vectors of the

impulse response traction Green functions,

R,y(£) are vectors containing the rupture orientation and rake dependent radiation coefficients, and

Djj(r,(i)) are the rupture-time and travel-time ^

delays calculated using the FATT-VSDASP-RRH

model, which will be discussed below.

2.1 Kinematic Description of Finite Fault Rupture- FATT-VSDASP-RRH Model

The FATT-VSDASP-RRH model consists of fractal asperity and travel-time (FATT) randomizations that includes variable high effective stress

circular asperities (VSDASP) imbedded in a fault

with a background of lower effective stresses combined with fractal randomized rupture and healing

velocities (RRH). Fractal randomization of stress drops, rupture, healing and S-wave velocities is consistent with the investigations of Frankel and

Clayton (1986) and Frankel (1991). Frankel

(1991) showed that a fractal distribution of asperities and stress drops can explain many aspects of

earthquake rupture and observed ground motions.

High stress drop asperities provide the dominant

high frequency radiation of seismic energy. Asperities have regions around their perimeter that provide smooth transitions of effective stresses from

the asperity interiors to the background regions of

the fault (Figure 2). The transition regions also

allow for transitional changes in average rupture

and healing velocities. Average rupture and healing velocities and randomization amplitudes are independently specified for the asperity interiors.

Additionally, asperities are allowed to heal from

their perimeters inward. This provides quite heterogeneous distributions of rise time on the fault. Self-healing asperities can have short rise times,

while background regions of the fault that are far

from healing boundaries (fault edges) can have

substantially longer rise times, consistent with the

results of Mikumo and Miyatake (1987; 1993) and

Fukuyama and Mikumo (1993). Short rise times in

the asperities provide high amplitude and frequency radiation consistent with Heaton's (1990)

observation of relatively short rise times for rupture models of large earthquakes. Longer rise

times in the lower effective stress background

region provide sufficient additional seismic

moment to provide total moments consistent with

observed moment magnitudes. If short rise times

are assumed everywhere on a fault then the asperities are required to provide most of the moment

and estimated effective stresses are unreasonably

high. This parameterization provides a means to explain low and high frequency observations of

large earthquakes, but requires less extreme effective stresses in the asperities than the constant rise

time model. The Von Karman fractal randomization of Frankel and Clayton (1986) is used to ran- domize background effective stress variations,

rupture and healing velocities in the RHH model

described below, and S-wave velocities. The same

randomization is used for all rupture quantities, as

it would be expected that high stresses and velocities would be spatially coincident.

Any realistic rupture initiates at a given time and

stops at a some later time. To describe the RRH

224
process we define starting, stopping, and healing

times for all points of a fault rupture. The rupture

velocity, Vnis allowed to vary randomly between

0.65 K,. and 1.0^v

, where Vs is local shear-wave

velocity Variations in ^.simulate the effect of

inhomogeneous rupture propagation that strongly

influences high frequency motions. The stopping

position function consists of all or part of the final

edges of the rupture or asperities where stopping

begins on the fault. The healing velocity, V/v is allowed to vary between 0.9^ and 1.1 Vs

. At each

point on the fault, the slip velocity is linearly

reduced to zero over a healing interval of 0.05sec.

A healing causality condition is imposed such that

healing points can only heal portions of the fault

that have rupture arrival times less than the healing

points' rupture times. This allows slip to go to zero

at the edges of the fault and produces realistic slips

in the interior of the fault.

Frankel and Clayton ( 1 986) showed that the Von

Karman slowness randomization can explain

observed crustal travel-time variations and seismic

scattering behavior. A similar fractal distribution

of stress drops on finite rupture faults can explain

the w spectrum radiated by large earthquakes

(Frankel, 1991; Herrero and Benard, 1994). Consequently, it provides an adequate means to ran- domize stress drops and S-wave velocities along a

finite fault. The Von Karman fractal distribution of

Frankel and Clayton (1986) with a correlation distance of 1 0 km is used to generate two-dimensional randomizations of S-wave velocities on the

fault. This randomization is superimposed on the

vertically heterogeneous S-wave velocity structure

appropriate for the site of interest and a twodimensional finite-difference travel time routine is used to calculate all rupture and healing travel

times.

2.2 Isochron Tools For Understanding NearField Ground Motions

Point source summation seismograms are difficult

to interpret. It is difficult to identify the regions of

the fault that contribute to specific portions of a seismogram. Bernard and Madariaga (1984) and

Spudich and Frazer (1984) developed the isochron

integration method to compute near-field seismograms for finite fault rupture models. Isochrons are defined as all the positions on the fault that contribute seismic energy at a specific receiver position at the same time. By plotting isochrons

projected on a fault, times of large amplitudes in a seismogram can be associated with specific

regions of fault rupture or healing.

The simplest way to employ the isochron method

in the near field is to assume that all significant

seismic radiation from the fault consists of first Swave arrivals. This assumption is verified in subsequent sections. A further simplification is to use a

simple slip velocity pulse. Let f(t) be the slip

function, For simplicity we assume

f(t) = 8(t — t^)-8(t — //,) where t,. is rupture

time, and this healing time. Then, all seismic radiation from a fault can be described with rupture

and healing isochrons. Velocity and accelerations

from rupture or healing can be calculated (Spudich

and Frazer, 1 984) from

v(x,t))=f(t)® \

(s-G)cdl (3)

y(t.x)

where c is isochron velocity, s is slip velocity

(either rupture or healing), G is the ray theory

Green function.

The FAT-VSDASP-RRH source model sometimes

produces isochron caustics (regions of a fault with

infinite isochron velocities) that can produce unrealistic peak accelerations using isochron integration (O'Connell and others. 1995). Spudich and

Frazer ( 1984) showed that c can be eliminated

from (3) by integrating along an isochron over a

finite time window defined by the isochrons t- dt

and t + dt. By limiting the integration to frequencies lower than 10 Hz {dt = 0.05s) extreme peak

accelerations are reduced, making the isochron

method useful for qualitatively evaluating accelerations. This approach is used here with a first order

approximation where (3) is reduced to a point source summation of over the isochron strip corre- sponding to the finite time window. We exploit this

feature to efficiently calculate velocities for a large

225
number of receiver positions relative to a fault. For

the isochron integrations G is approximated as

G(,,*)=

FM). jV( X,y

(5)

4nJp(x)p(^(x)^f/2

J(x-^)

2

where x is the receiver position, i; is position on the

fault, F(x, £,) is the source radiation term,

W-(x, t) is the free surface amplification factor, p

is density, and (3 is S-wave velocity. In this approximation, only first S-wave arrivals are included in

the calculation. Although ray spreading factors are simply approximated by the inverse sourcereceiver distance, an S-wave velocity model developed for this area is used to calculate, (3(^) , takeoff angles for F(x, ^) , and incidence angles for

W(x, to incorporate first-order geometric

effects of vertically heterogeneous velocity structure. This produces realistic partitioning of the

direct SV arrival between vertical and horizontal

components of ground motion.

Since isochrons are central to understanding point- source summation seismograms, we provide

explicit expressions for rupture and healing isoch- rons to illustrate how both source and propagation

factors can affect seismograms. The arrival times

of rupture at a specific receiver are

Tr(x)=t^)+tr^) (6)

where x is the receiver position, % are all fault positions, rp are S-wave propagation times between

the receiver and all fault positions, and t is the

rupture time at all fault positions. The arrival times

of healing at a specific receiver are

Th(x)=Tr(x)+R&) (7)

where R are the rise times at all fault positions.

From (6) and (7) it is clear that both fault rupture

variations and S-wave propagation time variations

effect isochron velocities. Often, most attention

has been focused on the effects of variations of

fault rupture in (6) and (7). Equations (6) and (7)

help illustrate how S-wave propagation time variations caused by anisotropy and heterogeneity can

significantly influence peak near-field velocities

and accelerations.

It is interesting to interpret (3) and (4) in the context of point-source summation seismograms.

When isochron velocities become large on a substantial area of a fault it simply means that all point sources sum at nearly the same time at the

receiver and velocities remain finite, but potentially large. In this way isochrons are a powerful

tool to dissect waveforms produced by point- source summation as well as understanding existing seismograms when information is available

about the rupture process. Times of large waveform amplitudes can be directly associated with

the regions of the fault that have corresponding

isochrons and large isochron and slip velocities.

2.3 Anisotropy and Heterogeneity in S-Wave

Propagation

An analysis of the distribution of S-wave arrival

times acquired from small earthquakes recorded

close to our site of interest suggests that a random

fluctuation of arrival times should be incorporated

into the 1 D reflectivity Green functions to account

for 3D heterogeneous velocity structure. Since the

S-wave arrivals dominate strong motion, incorporation of S-wave travel time variations associated

with anisotropy and heterogeneity into the Green

functions should produce more realistic phasing of

simulated ground motions.

The influences of anisotropy are clearly evident in

the waveform data acquired at this site. Inspection

of the three-component waveforms from about 20

earthquakes occurring north of the surface trace of

the Red Mountain fault reveal a pattern of S-wave

splitting consistent with fracture induced anisotropy associated with east-west faults as observed

by Li (1994) in the eastern Transverse Ranges. Swaves polarized approximately orthogonal to local

faults tend to be delayed relative to those polarized

parallel to the faults. To approximate this behavior, one function is used to apply time shifts to the

radial and vertical Green function components and

a different function is used to apply the anisotropy

time shift to the transverse Green function component.

226
3.0 FATT-VSDASP-RRH MODEL VALIDATION: NORTHRIDGE NEAR-FIELD

GROUND MOTIONS

One means to evaluate the ability of the FATTVSDASP-RRH model to produce realistic accelerations and velocities is to compare the synthetic

waveforms to empirical strong motion recordings.

The January 17, 1994 Mw 6.7 Northridge, California, earthquake occurred in the same tectonic setting as our site of interest and on a fault with a geometry similar to the Red Mountain thrust

source. Rupture occurred on a 45° south-dipping

fault with an average rake of 105° (Jones and others, 1994; Wald and others, 1996; Dreger, 1994).

The earthquake was recorded by numerous strong

motion instruments. Several near-field strong

motion stations (Newhall - LA County fire stations

and Sylmar - County hospital parking lot, for

example) are located in updip positions relative to

the earthquake rupture that are quite similar to our

site's position relative to the Red Mountain fault. The velocity structure near the Northridge earthquake (Haase and Hauksson, 1995) is similar to

the velocity structure near our site of interest. Consequently, synthetic FATT-VSDASP-RRH waveforms can be compared to selected recordings

from the Northridge earthquake to evaluate the

ability of the FATT-VSDASP-RRH model to produce realistic accelerations and velocities. It should be noted that all of the results presented

here have were developed using data (S-wave

travel time variations and scattering functions)

specific to our site of interest (near Ventura). Only

the fault rupture parameters are specific to the

Northridge event. No site-specific information for

the strong-motion stations was used in the validation exercise.

The Northridge earthquake ruptured from about 20

km depth to about 5-7 km depth and the fault

length was about 1 5-20 km (Wald and others;

1995, Hudnut and others, 1994; Dreger; 1994).

The Newhall and Sylmar stations are located about

1 2 km west and 1 km east, respectively, from the

hypocentral strike position in the rupture plane and

are located on the upthrown side of the fault (Hudnut and others, 1994: Wald and others, 1995).

We have produced a FATT-VSDASP-RRH rupture

model for Northridge consistent with the source models of Wald and others ( 1 996) and Zeng and

Anderson (1996). Corresponding synthetic ground

motions for several strong motion recording sites

of the 1994 Northridge earthquake reproduce

observed horizontal and vertical velocities and

horizontal accelerations quite well. Additionally,

we have used the isochron method to compute estimates of peak horizontal velocity over a 3600 km2

near-field grid for comparison with the actual

observations of peak horizontal velocities. Synthetic and observed ground motions are also compared with estimates computed from existing

strong-motion attenuation relationships.

One useful by-product of the rapid computation of

isochron results is the ability to evaluate the spatial

and temporal evolution of ground motions. Figure

3 shows contours of simulated peak horizontal

velocity at 1 s intervals in fault coordinates for the

Northridge rupture simulation. The positions of

several relevant strong-motion sites are shown as

well. Near the northwest corner of the fault (near

stations Sylmar and Rinaldi) very large motions

are both predicted and observed in a short time

interval between 7.8 and 9.8 seconds. To the north

of the rupture surface (near the Newhall station)

the maximum motions are slightly smaller but the

duration of strong motions is considerably longer,

9.8 to 13.8 seconds.

To help interpret the individual strong motion simulations produced with the FATT-VSDASP-RRH

rupture model as well as the observed recordings

we have computed isochron plots for individual

stations. An example for station NWS (north of the

fault) is shown on Figures 4 and 5. Figure 4 illustrates rupture isochrons and Figure 5 healing isochrons. Comparison of the peak velocity

contributions between the two figures illustrates

that most of the site velocity is contributed by the

rupture process. The large distance between the

isochrons in Figure 4 illustrates that a large portion

of the fault surface contributes arrivals in a small

time window (i.e. high isochron velocities). Slip

velocities from the large asperity located closest

to station NWS contributes the large pulse

observed on the synthetic records. There is clearly

more high frequency energy in the synthetic arrivals than in the observed recordings. The Green

227
functions were computed for the velocity structure

in the western Transverse Range and exhibit less

attenuation than is anticipated for the region near

Northridge.

A common methodology for predicting ground

motions for engineering analysis has been the use of empirical attenuation relationships. Recent

attenuation relationships relate ground motion

parameters of interest to earthquake magnitude

and some source-site distance measure along with

fault-type, site conditions, and in some cases fault

geometry (c.f. Boore and others, 1993 and 1994;

Campbell and Borzorgnia, 1994; Sadigh and others, 1993; Abrahamson and Silva, 1996). Often,

based upon the results of these relationships, one

or more existing strong-motion recordings are selected for use in the dynamic analysis of bridges,

dams or other structures. The most recent of these

relationships provide a reasonable, averaged estimate of the expected ground motions for many

scenarios. However, for situations in close proximity to the fault surface the fault normal component

of motion may be significantly larger than the fault

parallel component (Somervtlle and others, 1 996).

Likewise in cases where a site is located near an

asperity on the fault surface motions can be larger

than predicted. Recent observations from the 1 992

Landers, 1994 Northridge, and 1995 Kobe earthquakes show that rupture directivity significantly

influences observed ground motions. To evaluate

the ability of empirical attenuation relationships to estimate future ground motions in near-field situations we have computed peak acceleration and

velocity estimates for the Northridge earthquake over the same grid as for the isochron methodology. The relationship of Abrahamson and Silva

(1996) was used to produce estimates of acceleration values and the relationship of Campbell

( 1 990) was used to produce the peak velocity values. The spatial distribution of observed peak horizontal velocity is compared to the isochron

predictions in Figure 6 and Campbell's (1990)

empirical prediction in Figure 7. The observational data shown in Figures 6 and 7 is from Wald

and others (1996).

The empirical relationships are smoothed averages

of a number of observations and thus do not possess the spatial variability evident in the isochron

simulations. The isochron results suggest very

high local gradients in the velocity field updip of

the fault surface, (see Figure 6). Changes of 70

cm/sec in peak velocity are observed over distances of ~5km. The simulated peak velocities

agree quite well with the observations at stations

PARD, Newhall, NWS, Rinaldi, and Sylmar. The

empirical predictions of Abrahamson and Silva

(1996) for peak accelerations give reasonable

agreement with the simulations and observations

in smoothed sense. The velocity predictions of

Campbell (1990) seriously underpredict the simulations and observations in the updip region even

at the 84-percentile level (Figure 7).

Ground motion predictions in a "back-azimuth"

(i.e. towards the south) are considerably lower than

observed values. This appears to be due to at least

three factors, (1) stations in the southern San

Fernando Valley and northwestern Los Angeles

Basin did not experience forward directivity

effects. (2) because of the dip of the fault these

sites had a larger source-site distance than up-dip

stations. (3) significant three-dimensional basin

effects appear to have enhanced late-time arrivals

at these stations. Olsen and Archuleta ( 1 996) have

investigated this observation with a three-dimensional finite-element approach and suggest significant basin effects in these areas.

4.0 PREDICTION OF FUTURE GROUND

MOTIONS

Obviously we can not know a priori the rupture

geometry, distribution ofasperities and slip history

on the fault surface at sites where we must predict

ground motions from a fault for design purposes.

We have chosen to use the FATT-VSDASP-RRH

reflectivity model to predict motions for a site near

the Red Mountain Fault north of Ventura, California (Figure 8). The approach we have chosen to use is computation of a large number of plausible

simulations developed using the FATT-VSDASPRRH model to develop a distribution of ground

motion values at the site of interest. From this distribution, motions can be selected that are consistent with a desired level of conservatism. An

example of the resulting distribution is shown in

Figure 9 for PHA and PHV. Figure 10 displays

the relationship between the hypocentral depth and

228
strike position and resulting ground motions at the

site of interest. This figure illustrates a fundamental observation, large ground motions can be

expected when the hypocenter is located relatively

deep and downdip from the site and maximal Swave radiation patterns are located between the

hypocenter and the site (thrust faulting). This zone

defines a region of hypocenter positions where a

substantial portion of the total fault area (as high as 40%) can contribute coherent motions at the site in

a very short (~lsec) time window.

5.0 CONCLUSIONS

The results of this study confirm other recent

investigations of near-field ground motions from

the 1989 Loma Prieta, 1992 Landers. 1994

Northridge, and 1995 Kobe earthquakes. Specifically, in the near-field, directivity and focal mechanisms (maximal SV radiation patterns associated

with thrust faulting) can be more important than

distance in determining engineering ground

motions at a particular site.

6.0 REFERENCES

Abrahamson, N.A., and Somerville, P.G., 1996,

Effects of the hanging wall and footwall on ground motions recorded during the Northridge

earthquake, Bulletin of the Seismological Society of America, 86, S93-S99.

Abrahamson, N. A., and Silva, W., 1996, Preliminary attenuation relations for horizontal

response spectra including data from the 1 994

Northridge earthquake, submitted to Bull. Seismol. Soc. America.

Bernard. P.. and Madariaga, R.. 1984. A new

asymptotic method for the modeling of nearfield accelerograms, Bulletin of the Seismological Society of America, 74, 539-557.

Boore, D. M., Joyner, W. B., and Fumal, T. M.,

1993, Estimation of response spectra and peak

accelerations from western North America

earthquakes: an interim report, U.S. Geological

Survey Open-File Report No. 93-509, 72 pp.

Boore, D. M., Joyner, W. B., and Fumal, T. M.,

1 994, Estimation of response spectra and peak

accelerations from western North America

earthquakes: an interim report, Part II, U.S.

Geological Survey Open-File Report No. 94-

127. 40 pp.

Campbell, K.W., and Bozorgnia, Y., 1994, Nearsource attenuation of peak horizontal acceleration from world-wide accelerograms recorded

from 1957 to 1993, Proc. Fifth U.S. National

Conference on Earthquake Engineering, Chicago, IL.

Campbell, K.W., 1990, Reanalysis of strong

ground motion recordings; empirical prediction

of near-source ground motion for Diablo Canyon Power Plant site, San Luis Obisbo, California, Report to the U.S. Nuclear Regulatory

Commission.

Dreger, D. S., 1994, Empirical Green's function

study of the January 17, 1994 Northridge mainshock (Mw 6.7), Geophys. Res. Lett. v. 21,

2633-2636.

Frankel, A., and Clayton, R.W., 1986, Finite difference simulations of seismic scattering:

Implications for propagation of short-period

seismic waves in the crust and models of

crustal heterogeneity, Journal of Geophysical

Research, 91.6465-6489.

Frankel, A., 1991, High-frequency spectral falloff

of earthquakes, fractal dimension of complex

rupture, b value, and the scaling of strengths on

faults, Journal of Geophysical Research. 96,

6291-6302.

Frankel, A., 1993, Three-dimensional simulations

of ground motions in the San Bernardino Valley, California, for hypothetical earthquakes on

the San Andreas Fault, Bulletin of the Seismological Society of America, 83, 1042-1063.

Fukuyama, E., Mikumo, T, 1993. Dynamic rupture analysis: Inversion for the source process

of the 1990 Izu-Oshima, Japan, earthquake (M

= 6.5), Journal of Geophysical Research, v. 98,

6529-6542.

Hauksson, E., Jones, L.M., and Hutton, K., 1995.

The 1 994 Northridge earthquake sequence in

229
California: Seismological and tectonic aspects,

Journal of Geophysical Research, 100, 12,335-

12.355.

Heaton, T.H., 1990. Evidence for and implications

of self-healing pulses of slip in earthquake rupture, Physics Earth Planetary Interiors, 64, 1-

20.

Heaton, T.H., and Wald, D., 1994. Strong ground

motions from the Northridge earthquake; were

they anomalous or a warning of things to come?, EOS, 75, 175.

Henry, M., Orcutt, J. A., and Parker, R.L., 1980. A

new method for slant stacking refraction data,

Geophysical Research Letters, 7, 1073-1076.

Kappus, M.E., Harding, A.J., and Orcutt, J. A.,

1990. A comparison of tau-p transform methods, Geophysics, v. 55, 1202-1215.

Kind, R., 1978. The reflectivity method for a buried source. Journal of Geophysics, 44, 603-612.

Mikumo, T., and Miyatake, T., 1987, Numerical

modeling of realistic fault rupture processes, in; Seismic strong motion synthetics, B.A. Bolt,

ed., Academic Press, New York, 328 pp.

Mikumo, T., and Miyatake, T., 1993, Dynamic

rupture processes on a dipping fault, and estimates of stress drop and strength excess from

the results of waveform inversion. Geophysical

Journal International, v. 112, 481-496.

Li, Y.G., 1995. Shear-wave splitting observations

and implications in the Los Angeles Basin and

adjacent areas, California, Seismological

Research Letters, 66. 39.

O'Connell, D.R., and Ake, J.P., 1995, Ground

motion analysis for Hoover dam, U.S. Bureau

of Reclamation Seismotectonic Report 94-1,

1 14 pp.

O'Connell, D.R., Ake, J.P, and Block. L., 1995,

Ground motion analysis for Bradbury dam,

U.S. Bureau of Reclamation Seismotectonic

Report 94-5, 126 pp.

Olsen, K.B., and Archuleta, R.J., in press, 3-D

simulation of earthquakes on the Los Angeles

fault system, Bulletin of the Seismological

Society of America.

Sadigh, K.. Abrahamson, C.Y., Chiou, S.J., and

Power. M.S., 1993, Specification of longperiod ground motions: Updated attenuation

relationships for rock site conditions and

adjustment factors for near-fault effects, Proceedings of ATC- 1 7- 1 Seminar on Seismic Isolation, Passive Energy Dissipation, and Active

Control, vol. I, San Francisco, California.

Somerville. P.G., 1993, Engineering applications

of strong motion simulation: Tectonophysics,

218,195-219.

Spudich, P., and Frazer, L.N., 1984. Use of ray theory to calculate high-frequency radiation from

earthquake sources having spatially variable

rupture velocity and stress drop, Bulletin of the

Seismological Society of America, 74, 2061-

2082.

Spudich, P., and Frazer, L.N., 1987. Errata for

"Use of ray theory to calculate high-frequency

radiation from earthquake sources having spatially variable rupture velocity and stress drop",

Bulletin of the Seismological Society of America, 77. 2245.

Spudich, P., and Archuleta, R.J., 1987. Techniques

for earthquake ground-motion calculation with

applications to source parameterization to finite

faults, in; Seismic strong motion synthetics,

B.A. Bolt, ed., Academic Press, New York, 328

pp.

Wald. D.J.. Heaton, T.H, and Hudnut, K.W., 1996.

The slip history of the 1994 Northridge, California, earthquake determined from strongmotion, teleseismic, GPS, and leveling data,

Bulletin of the Seismological Society America, v. 86 no. IB. pp. S49-S70

Zeng, Y, and Anderson, J.G., 1995. A composite

source modeling of the 1994 Northridge earthquake using genetic algorithm, Seismological

Research Letters, 66, 52.

230
nm

u

Map view

Site

Green

Function

Site

Ground

Velocity

U

Time Time Time

Figure 1. Schematic diagram of finite fault rupture calculation of ground motions. Three discrete subfault

elements in the summation are shown. Rings and arrows emanating from the hypocenter represent the

time evolution of the rupture.The Green functions actually consist of eight components of ground motion

and three components of site ground velocities are calculated. Large arrows denote fault slip orientation

which is shown as predominantly reverse slip with a small component of right-lateral strike slip.

231
I

Figure 2. Schematic representation of asperities on a fault. The darkest regions are asperity interiors with

high effective stresses, the lighter regions are asperity transitions regions over which effective stresses

smoothly decrease from the values in the asperity interiors to background effective stresses. The unshaded

portion of the fault is the background region of lower effective stresses. The hypocenter is shown as a star.

!

232
40

30 \

20 {

10 \

0

-10

-20

T= 7.8

Rinaldi

SylmarZ

\ Newhall A

NWSA

40 J

30

20

10

0

-10

-20

T= 8.8 T= 9.8

A W : Wi t sB] ; sat*

40

30

20

10

0

-10

-20

-20 -10 0 10 20 30 40

1=10,8.

-20 -10 0 10 20 30 40

T = 11.8

20 -10 0 10 20 30 40

T= 12.8

-20 -10 0 10 20 30 40 -20 -10 0 10 20 30 40

T = 13.8 _ I=MfL

-20 -10 0 10 20 30 40

T = 15.8

40 J-

-20 -10 0 10 20 30 40

T = 16.8

20 -10 0 10 20 30 40

T = 17.8

-20 -10 0 10 20 30 40

T= 18.8

4-

"40 -20 -10 0 10 20 30 40 -20 -10 0 10 20 30 -20 -10 0 10 20 30 40

Figure 3. Contours of simulated peak horizontal velocity at 1s intervals (maximum is magenta) in fault co- ordinate map view. Position (0,0) is the updip projection of the northwest comer of the Northridge fault to

the surface, the rectangle is the map view of the fault plane, the epicenter is the star, time from initiation of

rupture is shown above each upper left corner, and stations are triangles labeled in the first frame.

233
Strike Position (km) (cm/sec)

Figure 4. Northridge rupture isochron plot for station NWS is shown at the bottom with gray shading showing the site velocity contribution from each fault element. Observed (black) and synthetic (gray) 46° horizontal velocity (top) and acceleration (middle) records are shown with the solid arrows from the isochron

plot showing the arrivals produced on the eastern end of the fault prior to 12.0s and dashed arrows showing the contribution from the dashed line region between 12s and 13s that produce the largest peak amplitudes. Hypocenter position is shown as the star. The position of NWS is shown as the filled triangle.

234
CD F

g o -200 F-

-400

NWSA

NW

c

o'4—

"(0

o

Q_

Q_

oQ

iti

; se

*>

- - :

j

"20

0.15

0.10

0.05

Site Velocity

(cm/sec)

5 10

Strike Position (km)

Figure 5. Same as Fig. 4-9, except healing isochrons for station NWS are shown at the bottom and the

316° horizontal velocity (top) and acceleration (middle) are shown. The same velocity scale is used to emphasize that healing contributions from the fault to site velocity are much smaller than from rupture. The

solid arrow shows the healing between 11.5s and 12s from the asperity near the hypocenter and the

dashed arrows shown that healing phases between 1 2s and 1 3.5s come from six asperities located in quite

distinct portions of the fault.

235
Distance From Fault Surface Projection (km)

Distance From Fault Surface Projection (km)

Figure 6. Contours of observed Northridge peak velocities in cm/sec (top) and isochron predicted peak velocities (bottom) for a 60x60km2 area. The surface projection of the Northridge fault plane (Wald and others, 1996) is shown as the dashed rectangle and the epicenter is the star. Station names indicate positions

of observed values.

236
Distance From Fault Surface Projection (km)

Figure 7. Contours of observed Northridge peak velocities in cm/sec (top) and emperical prediction peak

velocities (bottom, from Campbell (1990)) for a 60x60km2 area. The surface projection of the Northridge

fault plane (Wald and others, 1996) is shown as the dashed rectangle and the epicenter is the star. Station

names indicate positions of observed values.

237
Casitas Dam

Southwest A

Cross Sectional View

Figure 8. Schematic illustration of relationship between site of interest for ground motion computations (indicated by solid, dark triangle) and Red Mountain fault. Estimated magnitude of Red Mountain fault, Mw

6.9, based on average values suggested by rupture length and rupture area relations. Seismogenic depth

and dip of fault based on geologic information and microseismicity.

238
Red Mountain PHA From 1920 Rupture Simulations

0.12

o

I—

0.10

0.08

o 0.06

o

0.04

0.02

0.00

JL

Median = 699 cm/sec2

-Vi 84% = 953 cm/sec2

n 1

1

500 1000 1500 2000 ^b00

Peak Acceleration (cm/sec2

)

Red Mountain PHV From 1920 Rupture Simulations

0.10

« 0.08h

O

^0.06h

c

o

o

05

LL

0.04 :

0.02 1

0.00

n Median = '90 cm/sec

84% = 1 28 cm/sec -

0 50 100 1 50 200 250

Peak Velocity (cm/sec)

Figure 9. Histograms of peak horizontal acceleration (top) and peak horizontal velocity from 1920 simulations of the Red Mountain fault using the FATT-VSDASP-RRH rupture model. Ten strike positions and

eight hypocentral depths were used. A total of 24 rupture simulations were calculated for each position

(eight asperity models times three travel-time randomizations) to produce the 1920 simulations.

239
Strike Position (km)

Figure 10. Contours of median peak horizontal velocities (top) and horizontal accelerations (bottom) as a

function of hypocenteral depth and strike position for 1920 rupture simulaitons of the Red Mountain fault. Ten strike positions and eight hypocentral depths were used. A total of 24 rupture simulations were calculated for each position (eight asperity models times three travel-time randomizations).

240
Damage to Port Facilities of KOBE Port by The 1995 Hyogo-Ken Nanbu Earthquake and Their

Reconstruction

by

Tatsuo Uwabe*JCen Oikawa**, Tatsuo Wako*** and Takahiro Sugano****

ABSTRACT

Port facilities of Kobe port were seriously damaged by the

1995 Hyogo-Ken Nanbu Earthquake, particularly for

gravity type quaywalls except earthquake-proof ones at Maya wharf. Immediately after earthquake, reconstruction

plan to recover the function of Kobe port within 2 years

was decided based on 1) to achieve quick recover the

function of port function and 2)reconstruction with

consideration on future function.

KEYWORDS: Earthquake damage; Port facilities; Reconstruction; Earthquake ground motion

1 .INTRODUCTION

The 1995 Hyogo-Ken Nanbu earthquake which took place

early in the morning on January 17,1995 brought

tremendous damage to port facilities of Kobe port, a

representative international trading port in Japan, and gave a severe effect to economical activity both in Japan and

Asia. Immediately after the earthquake, Ministry of

Transport and Kobe city (port management body) carried

out a survey to evaluate the damages and prepared are

construction plan. Total damage of port facilities estimated 550 billion yen.

This paper describes earthquake ground motion, damage to

port and their reconstruction.

2.EARTHQUAKE GROUND MOTION

The 1995 Hyogo-Ken Nanbu Earthquake of M7.2, which

hypocenter is located at the northern part of Awajishima

Island, is the right-under type one caused by movement of a active fault which runs below Awajishima Island and

Kobe as shown Fig. 1 with seismic intensities recorded in Osaka Bay area. Earthquake motion measured by SMACEC type accelerograph placed on the ground surface at Kobe Port Construction Office is shown in Fig.2. The

maximum acceleration for the horizontal direction is 502

cm/s2(N45W component) and 283 cm/s2 (UD

component) for the vertical one. Motion of this

earthquake is characterized by 1) its relative short

duration of main shock of 5 to 10 seconds, 2) horizontal

motion being predominant to north-south direction.

3.0UTLINE OF KOBE PORT

Kobe port, one of the major ports in Japan together with

Yokohama port has been

developed since beginning of 20th century. Shin-ko piers

No. 1 to No. 6, Hyogo pier and central pier has already

been developed before the world war n. These facilities which were seriously damaged by the war were immediately restored, Shin-ko piers No.7 and No.8 and

Maya wharf were newly developed after the war.

Construction of Port-Island, first large scaled man-made

island in Japan, was started from 1966 and completed in

1981 followed by reclamation of Rokkou-Island. which was completed in 1990. Twenty one container berths

which are mostly located at two man-made islands were

under operations in Kobe port. Annual amounts of

containers at Kobe port reaches 2,700,000 container,

which is about 30 % of the total amount of Japan, and

keeps No. 1 position in Japan. Fig.3 shows plain view of

Kobe port.

4.DAMAGE TO PORT FACILITIES

Damage to port facilities which are quaywall, breakwater

and reclaimed

land are outlined in the followings.

*Chief of Earthquake Disaster Prevention Laboratory,

Port and Harbour Research Institute, Ministry of

Transport, Nagase 3-1-1, Yokosuka, Japan 239

**The 3rd District Port Construction Bureau, MOT,

Onohamacho 7-30,Chuo-ku, Kobe, Japan 651 *** Head of Geotechnical Engineering Division, PHRI,

MOT( Former The 3rd District Port Construction Bureau)

****Chief of Structural dynamics laboratory, PHRI,

MOT

241
(1)Quaywall

Structures of quaywalls at Kobe port are mostly gravity

type such as concrete caisson and block with a few steel

structure jetties. Design of these structures in Japan is

carried out under the 'Technical Standards for Port and

Harbour Facilities^". Seismic design is generally

conducted by seismic coefficient method and seismic

coefficient is determined by regional earthquake activity,

ground condition and importance of the structures. Clear

differences were observed in damage according to seismic

coefficient adopted in design and direction of face line. Quaywalls designed for small seismic coefficient of 0.15

to 0.18 were mostly damaged seriously while earthquakeproof berths with 0.25 at Maya wharf shown in Fig.4 were not subjected to any significant damage. It was also

observed that the damage of the quaywalls directed to east- west is larger than that directed to north-south.

Gravity type quaywalls slid to offshore side reaching lm

to 5m and subsided lm to 2m subsidence of 3m ~ 4m at

the back site due to lateral deformation of the quaywall as

indicated in Fig. 5. Buckling and de-railing caused by

crotch-tear generated by sliding of the quaywall 1 were also

observed for gantry cranes.

(2)Breakwater

Damage observed for breakwater were mostly settlement

reaching 2m in maximum.

Differential displacement and settlement were rather small.

(3)Reclaimed land

Liquefaction took place at most and limited parts of PortIsland and Rokkou-Island, respectively, which were

reclaimed by well graded Masa soil, de-composed granite,

of which gradation curve is shown in Fig.6 together with

poorly graded sands with high liquefaction potential used

for reclamation in other areas. Though catastrophic failure observed for structures in Niigata earthquake did not take

place, settlement reaching 0.5m in maximum was

entirely observed and trafficability was seriously damaged

by liquefied sand covering the ground surface.

5.RECONSTRUCTION OF KOBE PORT

(1) Fundamental approach

In order to recover the function of Kobe port as the hubport in Asia together with

taking into its future function for 21th century, it was

determined that the reconstruction should be done with

considerations on l)quick recovery of port function, 2)

increase of eanhquake-proof,3) cooperation with

reconstruction of city area and 4) reconstruction as the

international hub-port.

1)Qirick recovery of port function

In order to reduce the effect of damage port function,

provisional use as aimed fa slightly damaged quaywall

which are restored with a slight repairing. For facilities

seriously damaged, it was planned to enable their use step

by step. It is also important to repair the breakwater

function quickly in order to maintain the safety of

facilities from storm surge and tide.

2)Increase of earthquake-proof

Earthquake-proof is increased for damaged structures according to their priority. In restoration of quaywalL

increase of seismic coefficient and combination of

structures with different seismic response are taken into

account also according to the priority of the facilities. Earthquake-proof quaywall is constructed at various

locations among container, ferry and existing wharf. A

value of 0.25 is used as the design seismic coefficient for

the earthquake - proof quaywalls, considering that Maya

quaywall actually resisted the earthquake force without

destruction. For ordinary quaywall, the design seismic

coefficient is increased from 0.1 8to 0.2.

3)Cooperation with reconstruction of city area.

In order to cooperate with the reconstruction work for

city, waste materials in the city are used for reclamation.

4)Reconstruction as the international hub-port

Construction of large scaled container berth together with

redevelopment of existing wharf was planned to be

progressed in order to satisfy the function as the

international hub-port.

(2) Increase of earthquake-proof quaywall

In restoration oft the structures seriously damaged, design

for recovery was done based on following three plans,

1) Construction of new quaywall in front of the damaged

one. This design was adopted when the damaged quaywalls can no longer be used as they were originally and removal of

displaced structures is impossible. A typical restored

cross-section is shown in Fig.7 The gravity caissoncomposite quaywall was changed to piled jetty type

quaywall.

2) Reduction of earth pressure of the back site. This design was adopted when damage is rather small. The

existing quaywalls can resist only by reducing the active

earth pressure behind them. Cement mixing method or

replacement method using blast furnace slag is used to

reduce the earth pressure by adding cohesion to the fill

242
material behind the revetment. An example of this method

is shown in Fig. 8.

3) Reconstruction of the quaywall with temporary release

of caisson with considerations on degree of damage and

locations o structures. This design was adopted when no space exists in front of

them. In case of reconstruction of the gravity caisson, the

sliding resistance must be increased, for example, by

placing an asphalt mat beneath the caisson. An example

of this method is shown in Fig.9.

(3) Schedule for reconstruction works

It was firstly decided to restore the function of Kobe port

within 2 years. Among 21

container berths provisional use of 8 berths was planed to be achieved until end of June, 1995 with temporary

repairing as shown in Fig. 10. Full-scaled reconstruction was commenced from April, 1995 and complete within 2

years. Seven ferry berths and 50 % of other quaywalls were restored until end of December, 1995. Function of

important breakwaters was recovered by the end of

September, 1995 against the attack of typhoon.

6. SUMMARY

Port facilities of Kobe port were seriously damaged by the

1995 Hyogo-Ken Nanbu Earthquake, particularly for

gravity type quaywalls except earthquake-proof ones at Maya wharf. Many bridges connecting the city and man- made island were also damaged. Immediately after

earthquake, reconstruction plan to recover the function of

Kobe port within 2 years was decided based on 1) to achieve quick recover the function of port function and

2)reconstruction with consideration on future function.

Full-scaled reconstruction are now under way at the entire

area of Kobe port and some of the container berths were

already under provisional use.

REFERENCES

1)Ministry of Transport(ecL):Technical Standards for Port

and Harbour Facilities in Japan, new edition, Overseas

Coastal Area Development Institute of Japan, 1991

2)Wako,T. : Damage to port facilities of Kobe Port by

the 1995 Hyogoken-Nanbu earthquake and their

reconstruction plan, Proc. Int. Symp. on Coastal Ocean

Space Utilization, 1995

3)Kamon,T., T. Wako, K. Isemura, K. Sawa, M.

Mimura, K. Tateyama and S. Kobayashi : Geotechnical

Disasters on the waterfront,Soils and Foundations,

Special Issue on Geotechnical Aspects of the January 17

1995 Hyogoken-Nanbu Earthquake, 1996

4)Wako, T. , K. Oikawa, Y. Matunaga and E. Naruse : Damage to Kobe Port Facilities by the Hyogoken-nanbu

Earthquake and their Reconstruction, Tsuchi-to-Kiso, The

Japanese Geotechnical Society, Vol.44,No.3,1996

243
UNIT: ca/s'

Honshu Island

Aaagasaki

472T

Kobe Port 311 ( 502

283

472(EN)

(UD)

©

Hypocenter

Seto-lsland

Sea

Osaka Bay

Awaj ishina

Osaka

178(SB)

103(UD]

>® Wakayaaa

159 (NE)

67(UD)

Shikoku

I s land

Fig. 1 Location of hypocenter and seismic intensities recorded in Osaka Bay area

500

to

o

id

I—D

-500

500

H -500

500

-500

LID

Max. 283

10 20

Time (s)

30

Fig. 2 Strong-motion acceleration measured on ground surface

at Kobe Port Construction Office

244
Ferry Berth

Container Berth

Fig. 3 Plain view ofKobe Port

Face line

+4.0

20.00 10.00

H.W.L. +1.7

L.W.L. +0.0

+2.0

-10.3

\ Back filling

^Shed

Filling sand

ri 14.4m

Gravel

Backfilling

-13.0

Original Cellular Block

Quayw&ll

-13.0

Asphalt mat Replaced Sand ' -\ -16.0

Rubble mound

Fig. 4 Earthquake-proof quaywall at Maya wharf

245
Horizontal

displacement Subsidence

g Sealf.VB.1

'• j Caisson

Rubble Mound

Reclaimed Fill

Replaced Sand

Cohesive

Soil

Fig. 5 Typical damage observed for gravity type quaywall

—

0

•

0.01 0.1 Grain Size (m)

i

1.0

i 10

Clay Silt Sand Gravel

Fig. 6 Gradation curve ofMasa soil used for reclamation in Kobe

no

L.W.L. i0.0-

Steel pipe 4 - 1.300

I - 32.5 I- 19

-J3LSL

Armour

Stone.

_j Bearing Layer

Fig. 7 Restoration of quaywall in front of the damaged one

246
New Face Line

(Previous Face Line

, J]—

!

Fig. 8 Restoration of quaywall using cement-treated soil

[.Face Line

j 42.00l(Compacticxi)

Fig.9 Restoration of quaywall using existing caisson

247
Fig.10 Provisional use of container berth and location for

the use of waste material in Kobe port

248
Major Factors of Damages of Reinforced Concrete Buildings

due to the 1995 Hyogo-ken Nanbu Earthquake

by

H. Hiraishi 1}

, T. Kaminosono2)

, M. Teshigawara 3) and S. Nakata 4)

ABSTRACT

Reinforced concrete buildings were severely damaged

due to the 1995 Hyogo-ken Nanbu Earthquake, and

some of them completely collapsed. Most of these

damaged buildings were those designed by the

previous Japanese design code. Particularly the

damage of medium high-rise buildings designed by

the previous code was distinguished The number of

damaged buildings which were designed by the current Japanese design code revised in 1981 and were

classified as "severely damaged" or "collapsed" is very

few except for the buildings with particular structural

layout. This paper describes the major factors on two

destructive typical damages of reinforced concrete

buildings. One is the complete collapse of mid-height

story of medium high-rise buildings from seven to twelve stories which were found only for the

buildings designed by the previous code. The other is

the collapse of piloti of buildings with the soft first

story which were found for the buildings designed by

both the previous and current codes. Through this

studies, it was concluded that, the location of the story

which suffered damage like pancake collapse depended

on vibration mode as well as strength distribution

along the height of the buildings and insufficient

lateral capacity and ductility of columns at that story

caused such collapse, and that there were many factors

which caused severe damage of piloti buildings, and

reinforced concrete walls considered as non-structural

walls were one of most important factors. The paper

also proposes some design recommendation to avoid

such catastrophic collapse.

KEYWORDS : 1995 Hyogo-ken Nanbu earthquake;

earthquake damage; reinforced concrete buildings;

collapse of mid-height story; piloti.

COMPLETE COLLAPSE OF A MID-HEIGHT

STORY OF MEDIUM HI-RISE BUILDINGS

Damage example

Photograph 1 shows a typical damage example of the

eight-story office building which was designed by the

previous code. This building had reinforced concrete framing in the upper four stories supported by steelencased reinforced concrete(SRC) framing in the

lower four stories, and steel in the column was

developed up to the mid-height of the fifth story

columns. The sixth story completely collapsed.

Except sixth story, the damage level of the story was

slight or small, as a whole, although some damages

were observed such as the fracture of joint steel at the

splice of steel due to tension, damage of column bases

due to pulling out of anchor bolts at the base plate of

steel, and damages of columns at the fifth story due to

falling down of sixth story. The size of columns at the

sixth and seventh stories was 65 cmX65 cm and

longitudinal steel bars in these columns were eight

round bars with a diameter of 22 mm although the size

and longitudinal steel bars in fifth story columns were 75cm X 75cm and twenty round bars with a diameter

of 25 mm, respectively.

1) Associate Director for Composite Structures

Research, Building Research Institute (BRI),

Ministry of Construction, Japanese Government

Tatehara 1, Tsukuba City, Ibaraki Pref. 305,

JAPAN

2) Head, Construction Techniques Division,

Production Dept. BRI.

3) Senior Research Associate, International Institute

of Seismology and Earthquake Engineering. BRI.

4) Director, Production Dept. BRI.

249
The ratio of the amount of cross sectional area of the

longitudinal steel bars to the cross sectional area of

the sixth story column was about 0.8% which was the

minimum requirement specified in the previous

codeAnd the amount of the longitudinal steel bars in

the sixth story column was about one fourth as small

as that in the fifth story column.

Photo. 1. Complete collapse of mid-height story of a

eight story building.

Design shear forces specified in the current and

previous codes.

This damage pattern was found only in the buildings

designed by the previous code. Therefore, in this

section, the design shear force specified in the

previous code is discussed by compared with that in

the current code.

In the previous code, allowable stress design was

specified for the horizontal earthquake force Fi at i-th

floor was given by Eq. (1).

F; = Z • G K • W; (i)

where,

Z = seismic zone coefficient (= 0.8 ~ 1.0 and

1.0 for Kobe),

G = soil-structure coefficient (= 0.8 - 1.0 and

1.0 in the most area in Kobe defined as the

largest value of VII of Japanese seismic

intensity scale),

K = seismic coefficient (=0.2 to height of 16m

and below, and increased by 0.01 for every

4m above), and

Wi = sum of dead and live loads of i-th story.

In the current code, the lateral seismic force in current code is given by Eq. (2).

Q = Z • Rt

• Aj • C0 (3) where,

Qi = seismic shear force at i-th story, n = number of stories,

Ci = story shear coefficient,

Z = seismic zoning factor ( = 0.7 ~ 1.0 and 1.0

for Kobe),

Rt = vibration characteristics factor and 1.0 if

natural period of a building is less than

0.6 and the type of soil is so soft as that of most area in Kobe defined as VII of the

Japanese seismic intensity scale,

Ai = vertical distribution factor, and

Co = standard shear coefficient, and not less than 0.2 for the strong earthquakes which

can occur several times during the

lifetime of the building, and not less than

1.0 for the severe earthquakes which

could occur once in the life time.

The stresses in the structural members shall be less than allowable stress for strong earthquakes given by

Eq (2) and the lateral load carrying capacity of each

story shall be greater than Eq. (4) for severe

earthquakes.

Q,=A-C0|w,

<4)

Where,

Qun = required ultimate capacity for lateral

load of a story,

Ds = structural characteristics factor and 0.3

for most ductile reinforced concrete

frames, and

Fes = a shape factor which considers the

rigidity and eccentricity factors and 1 .0

for the building whose structural

layout is regularly and well-balanced.

Equations (2) and (4) are simply rewritten by Eq. (5)

for strong earthquakes and Eq (6) for severe

earthquakes respectively, for the most of low-rise and

250
medium high-rise buildings in Kobe, if the minimum

values of Co, Ds and Fes are considered.

Qi = 0.2 A; • \ W;

Qun = 0.3 Aj • \ Wi

(5)

(6)

In the following discussions, it is assumed that the mass of each story is the same, and the story height is 3.0m. Figure 1 shows story shear coefficients

calculated by Eqs. (1), (5) and (6), and values of Eq. (1)

times 1.5. By comparing the design story shear

coefficients by Eqs. (1) and (5) because same

allowable stress design is specified for these shear

forces, in the both previous and current codes, it is

easily found that design shear force specified in the

previous codes is considerably smaller than that in the

current code, at the upper stories of a building. And if

the ultimate capacity for lateral load of each story of a building may be 1.5 times that defined by Eq. (1) and

Eq. (6) is minimum requirement for severe

earthquakes, the shortage of the ultimate capacity of a

story designed by the previous code is much

remarkable as the location of the story of a building is

higher. However, the ultimate capacity of the stories

near the uppermost story of a building designed by the

previous code may be more than minimum ultimate

capacity required by current code, because of

minimum requirement such as size of members,

minimum reinforcement, etc.

Estimation of Ultimate Capacity for lateral Load of

Buildings Designed by the Previous Code

The ultimate capacity of columns of a story is assumed to be sum of ultimate lateral capacity of

columns of a story considered, because the failure

mechanism of this damage pattern is the complete

collapse of a certain story. It is also assumed that the

shear capacity of these columns is higher than their

flexural capacity .

The flexural capacity of a column is given by Eq. (7).

Mu = 0.8 a, • q • D + 0.5 N- d(i N

y I b-D- Oh

(7)

where,

Mu = flexural strength of a column (N mm),

at = amount of cross sectional area of

longitudinal tensile steel bars (mm2),

sy = tensile yield strength of longitudinal

steel bars (MPa),

N = axial force in a column (N),

sB = compressive strength of concrete (MPa),

b = width of a column (mm), and

D = depth of a column (mm).

Equation (7) is rewritten by Eq. (8)

,

if N / (b • D • qb) is assumed t0 be 0-L

Mu = 0.8 a, • av

* D + 0.45 N D

(8)

The shear Qm of a column at the flexural strength at both ends is given by Eq. (9).

Qm = 2 Mu / ho (9)

where,

ho = clear height of a column (mm).

The ultimate capacity for lateral load of i-th story of a

building is given by dividing sum of Qm of each

column at i-th story by their sum of axial force, and

the shear coefficient mCi of i-th story at the ultimate

capacity is given by Eq. (10). Here, the following

assumptions are used.

i) The dimensions of depth and width of every

column are the same.

ii) at in each column in a certain story considered is

the same.

iii) D/ho is 1/3 for every column.

iv) The ratio of the sum of cross sectional area of the

columns to the floor area at a certain story

considered is 1/100.

v) The mean unit mass of a building is 1.2 t / m2.

vi) sy is 360 MPa.

251
I :

i:

i:

j:£

(

j

i

\

•I

~- Design star/ shear coeff. (Current code) — Required minimum value of ultimate capacity '— Design story shear coeff. (Previous code) — Design story shear coeff. X 1.5

0.2 0.4 0.6 0.3 1.0

Design story shear coeff.

Fig.l. Design story shear coefficients specified in the previous and current codes.

Story

10

9

8

7

6

5

Minimum strength

required b y the

• current code

Shortage of strength

Strength estimated from flexural strength

of columns with pt of 0.2%

Strength estimated from flexural strength of

columns with pt of 0.8% for lower seven stories

Story Shear Coefficient

0.2 0. 0.6 0.8 I.O

Fig.2. Estimated ultimate capacity of the buildings designed by the previous code

and minimum ultimate capacity required by the current code.

252
»Q = —"»2_ iP, + 0.3 n + 1 - 1 (10)

where,

iPt = at / (bd) of the column at i-th story,

n = number of stories of a building, and

i = 1+ number of stories of a building

above i-the story.

The first term and the second term in the right-hand

side of Eq. (10) represent terms related to the

longitudinal steel bars and axial force, respectively.

The ultimate lateral capacity of 0.3 of story shear

coefficient provided by the axial force may be large

enough to satisfy allowable stress for about 0.2 of

story shear coefficient specified in the previous code.

This results means that longitudinal steel bars of

most of columns designed by the previous code were

arranged by referring to their minimum requirement.

However, when the axial force is reduced to be half due

to up-down input motion, the second term become

0.15. Thus the ultimate capacity of buildings is remarkably affected by up-down motion.

estimated value of ultimate lateral capacity of lower

seven-story in which pt of columns is assumed to be

0.8% considering steel plate in columns is also

shown. The shortage of ultimate lateral capacity of

mid-height stories is furthermore distinguished The

building shown in Photograph 1 is the typical damage

example in which reinforced concrete sixth story

above lower SRC stories completely collapsed.

Figure 3 shows the case studies for 10, 8, 6 and 4-

story buildings. The solid line shows the story shear

coefficient of ultimate lateral strength given by Eq.

(10) in which pt is 0.3%, and the dashed lines show

a Ai which has Ai distribution profile and comes in

contact with the solid line. The multiplier a

corresponds to Ds. The more the number of story of a

building is, the higher the relative contact point is and

the smaller the a is. The relative contact point and

the value of a is largely affected by the shear

distribution profile during earthquakes. It is expected

that this profile is closer to Ai distribution rather than

design shear distribution specified in the previous

code.

Discussion

Equation (10) is the result based on flexural strength

of columns considering damage patterns caused by the

earthquake. If columns will fail before they reach

flexural yielding or beams reach their ultimate

strength, the ultimate capacity for lateral load of the

story become smaller value than that given by Eq.

(10) , so values that given by Eq. (10) is one of the

possible maximum value of the ultimate capacity.

Figure 2 shows the result of pt of 0.3% which

corresponds to the ratio of the minimum requirement

of 0.8 % of longitudinal steel bars by comparing the

minimum requirement of ultimate capacity specified

in the current code given by Eq(6). The shortage of the

ultimate capacity comparing to 0.3 of Ds is remarkable at mid-height stories of the building.

In existing buildings, the dimensions of columns and

amount of the longitudinal steel bars in columns are

usually lager in the lower stories, and in Japan the

lower stories of a building are commonly constructed

by SRC if a building is higher than 21m. Such

construction practice enhances the ultimate lateral

capacity of lower stories of the building. In Fig. 2, the

COLLAPSE OF PILOTI BUILDINGS

Damage Example

Photograph 2 shows a five-story building whose first

story completely collapsed This building was

designed by the current design code. There are

reinforced concrete walls in the upper four stories.

This type of structures is called as piloti buildings and

they suffered severe damage including those designed

by the current code. Any severe damage was not

observed in the upper four stories. The structural slit where thickness of concrete wall was half of that of

the in-panel wall was placed between columns and

in-panel walls. In the structural design, these in-panel

walls were considered as non-structural walls and their

structural effect was not considered . However, these

walls might have worked as shear walls during the

earthquake. Four of five piloti buildings which were

designed by the current code and were identified as

completely collapse by Building Research Institute

had similar non-structural walls in the upper stories.

253
J I I

' ' I

1 1 1 i 1 1 0 0.2 0.4 0.6 0.8 i.O 0 0.2 0.4 0.6 0.8 1.0 0 0.2 0.4 0.6 0.8 1.0

Story shear coeff. Story shear coeff. Story shear coeff.

Fig. 3. Estimated ultimate capacity of 10, 8, 6 and 4-story buildings with 0.2% of p,.

254
Earthquake

force

Building without

first soft story

Earthquake

force

Buildingwith

pilot!

Fig. 4. Examples of earthquake force distribution of

building with and without first soft story.

2) Axial force and shear force.

During earthquake response, forces and moments in

the piloti columns changes significantly even after Photo. 2. Complete collapse of a piloti building. yield mechanism is formed and it is not easy to find

appropriate design forces and moments.

Factors for Damage

There are many factors which might have cased such

damages. These factors are discussed below.

1) Shear distribution profile.

Response shear distribution profile along the

height of the building due to input earthquake motion

may be top heavy distribution like Ai distribution if

the structural layout of the buildings are wellbalanced and regularly shaped However, this

distribution profile is not appropriate for design shear

for piloti buildings because of soft stiffness of the

piloti story (see Fig. 4).

3) Rigidity Factor and eccentricity Factor.

In the current design code, the effects of non-uniform

rigidity in the vertical direction and eccentricity

between centers of mass and rigidity on structural

performance are specified. However, the rigidity of

members, particularly that of shear walls remarkably

changes depending on damage degree. Therefore, these

effects continuously change during dynamic response

because the members in the piloti story may suffer

damages such as cracks of concrete and yielding of

steel bars, while those in the upper stories may suffer

little damage. In the structural design, such change of

the rigidity should be considered.

4) Seismic performance of buildings with story

collapse mechanism

Input energy is dissipated mostly in the soft story of

piloti buildings, so piloti columns should be ductile

enough . This required ductility of columns needs to be further researched. Another subjects to be

considered is restoring force characteristics of

columns subjected to high axial force. In these

columns such as piloti columns, the load carrying

capacity remarkable decreases after the maximum

capacity even if they fail in flexure (see Fig. 5). If the

P-I effect is considered it is easily found that this

descendng slope after the maximum capacity should

be emphasized to be considered because it affects

255
significantly the seismic performance of buildings.

On the other hand, buildings with beam yielding

mechanism have not such problem because the load

carrying capacity continuously increases even after beam yielding due to effects of strain hardening of

steel bars, yielding of slab reinforcement, etc.

Essentially, piloti columns should be designed to remain before maximum capacity.

Fig. 5. Idealized restoring force versus deformation

relations of the buildings.

CONCLUSIONS

This paper described two types of destructive damage

patterns of reinforced concrete buildings which might

have caused mass loss of human lives: complete

collapse of mid-height story and collapse of piloti of

the soft first story. The followings were concluded.

1. The location of the story which suffered severe damage like pancake collapse depends on vibration

mode as well as ultimate capacity distribution along

the height of the buildings. Insufficient lateral

capacity and ductility of columns at certain story

caused such collapse.

2. There are many factors which caused collapse of

piloti columns. Essentially, piloti columns should be

designed to remain before their maximum strength.

256
PUSH OVER LABORATORY TESTING OF UNREINFORCED MASONRY INFILLS

by

Ghassan Al-Chaar1

, Steve Sw eney

1

, and Pamalee Brady1

ABSTRACT:

The U. S. Army Construction Engineering

Research Laboratories, USACERL, Performed

push over laboratory tests on three models of

reinforced concrete frames with masonry infill. These tests where part of a study of the

ultimate strength boundary and the residual

strength of infilled concrete frames. This

paper describes the failure mechanisms of the

specimens, and includes a brief discussion on

major parameters that describe the behavior.

1. INTRODUCTION:

Test assemblies consist of masonry infilled

R/C frame. In analysis, the redundant

constraints between the infill and the frame

yield more unknown forces than equilibrium

equations can resolve, and therefor is a

statically indeterminate problem. Thus, the

composite behavior of such structural entities

is best revealed experimentally. The

USACERL push over tests is part of a larger

research effort to further the understanding of

the composite behavior of R/C frame with

masonry infills. Dynamic testing of the two

small scale specimens has been reported

previously (Al-Chaar, 1993) and called halfscale strong and weak frames infilled with

half-scale brick masonry units. The third

model tested is similar in shape to the weak

and strong frames, it is twice the scale, and it

is referred to herein as the large frame. Scaled

concrete block units were used to infill the

large frame opening. All three frames were

used for testing previously and were partially

damaged as manifested in small cracks within

the frame. The infill in the weak and strong

frame also was damaged in previous tests,

while the infill in the large frame was

undamaged. The previous tests, however,

reached the lower boundary of the inelastic

limit of the frames. Table 1 below describes

the dimensions and the material properties of

the three models. In-plane monotonic loading,

stroke-control push over tests were carried out

to further understand the inelastic behavior of

these physical models as they are subjected to

drift ratio where the models are severely

damaged beyond repair. These tests provided

information beyond the elastic limit of the

models which was not achieved in the dynamic

tests.

2. EXPERIMENTAL SETUP:

As shown in Figure 1, two stroke-controlled

hydraulic actuators were installed on steel

reaction frames. The actuators were capable of

applying in-plane lateral monotonic loading on

the center of the top beam in each specimen.

A stroke of nearly 6 inches was imposed

slowly, over 15 minutes, while other

dependent parameters were recorded in two

second intervals. Some of the major

Research Structural Engineers, U. S. Army Construction Engineering Research

Laboratories, P. O. Box 9005 Champaign, Illinois 61826-9005

257
parameters acquired during the test were

loads, drift, diagonal deformation, and joint

rotation. Greased Teflon plates were installed

between the actuators and the beam to

eliminate any shear effect resulting from

friction between the two components. Modes

of failure were observed during the tests,

indicated by crack propagation in the infill panels and the formation of cracks and hinges

in the R/C frames.

3. FAILURE MECHANISMS OF

INFILLED R/C FRAMES:

The three models failure mechanism are compared to the family of failure mechanism

types presented by Mehrabi and Shing, showed

in Figure 2, (Mehrabi, 1996).

3.1 The strong frame model:

The failure mechanism observed in the infill was a diagonal tension crack propagating

along the diagonal strut. The stair crack

initiated at the upper corner of the windward

side ofthe specimen. As seen in Figure 3, the

crack propagated 45 degrees toward the

center and branched into two main horizontal

cracks, then they merged into one crack

toward the opposite corner. This crack

pattern deviated slightly from types identified

by Mehrabi. This is probably due in part to the

pre-cracks presented from previous out-ofplane tests. However, hinges in the R/C frame

formed similar to type E4 or E6 described by

Mehrabi. Figure 6 shows the location of the

hinges.

3.2 The weak frame model:

Stair step cracks initiated again at the infill upper corner of the windward side of the

specimen. As seen in Figures 5 and 6, these

cracks propagated wider and branched into

three main horizontal cracks, then they merged

in two cracks toward the opposite corner of

the infill. This mode of failure is identified as type C5, with diagonal cracks and two hinges

in the windward column, (Mehrabi, 1996).

Again, because this model was previously

tested out-of-plane, cracks from previous

tests likely caused the formation of secondary

cracks under in-plane testing. As the stroke

increased, hinges initiated at the bottom of the

windward column followed by a wide crack in

the leeward column, and in the beam

approximately to one beam depth away from

the windward column. A dead zone remained

intact below the bottom cracks. Figure 7

shows the R/C frame deformations of the

weak frame model.

3.3. The large frame model:

Failure type Dl as described by Mehrabi was

clearly observed. The failure surface clearly

indicated poor bond between the mortar and

the concrete blocks. Also, as illustrated in

Figure 8, zones in triangles 1 and 2 shows

vertical tension cracks unlike zone 3 that

exhibited pure joint slippage between all courses.

4. LOAD VS DRIFT OF THE MODELS:

4.1 The strong frame model:

As shown in Figure 9, the initial strength of the

strong frame model was 28.2 kips at driftheight ratio of 4.7%. The structure, beyond

the ultimate strength, exhibited ductile

behavior important in dynamic capacity

evaluation. A drift-height ratio increased

nearly three times at approximately 30%

reduction in strength of the strong frame

258
model. The residual strength at 15% drift is

approximately 40% the ultimate strength.

These values indicated a high energy

dissipation capability of this type of model,

characteristics that are important in resisting

seismic loads.

4.2 The weak frame model:

The weak frame load-drift curve exhibited a

different behavior from the strong frame

model. An initial strength of 12.8 kips at 1.6

% drift ratio was achieved. The load resistance

between 1.6% and 3.6 % drift ratio decreases

by 32% and than increased to 130% of the

initial strength, at 10% drift. The first peak

explained the infill initial strength, where the

second peak indicates the composite resistance

of the infill and the frame after the formation

of severe cracks on the infill. The failure

mechanism ofthe weak frame appears to show

poor interactions in the early deformation of

the model and overall poor composite

resistance. Although this model residual

strength at the second peak strength is only

60% of the strong frame model ultimate

strength, its still exhibits ductile characteristic,

but at a lower level of load resistance and

residual strength.

4.3 The large frame model:

The large frame load versus % drift ratio curve

in Figure 9, shows a very low initial strength

of 1 1.9 kips at 1.3 % drift ratio. This is due in

part for the low prism strength of the concrete

masonry infill blocks. The similarity in shapes

ofthe load-drift ratio curves of the strong and

large frames is very apparent despite their

different modes of failure. This observation

may lead to develop a function that can be

applied to a family of curves for similar infill and frame relative rigidities. Due to 6 inches

stroke limitation, the drift ratio beyond 9%

was not carried out.

All three frames were able to achieve 9% drift

height ratio value, way beyond the initial

strength of the three models. At this drift

state, the ratios of the residual strengths to

initial strengths are approximately 90%,

120%, and 60% for the strong frame model,

weak frame model and large frame model,

respectively. Accountability for variations in

models parameters such as prism strength,

and scaling between the models, will be

encompassed in future work.

5. DIAGONAL DEFORMATION OF

THE INFILLS:

The diagonal deformations of the infills for the

three models were measured and plotted

against the corresponding drifts and loads.

The load versus diagonal deformation curves,

shown in Figure 10, were limited to a stroke

of 2.3" where the limits of the LVDT were

reached. However, from these curves, it can

be seen that the compression and tension

deformations were uniform. From Figure 11,

the diagonal deformations within the first 2

inches of loading were linearly deformed.

Thus, diagonal deformation of the infill can be

modeled as a linear inelastic change in shape

that it is a characteristic of the composite

behavior ofthe R/C frame with infill.

6. JOINTS ROTATIONS OF R/C

FRAMES:

To study the rotations in selected critical

sections at which plastic hinges may form,

experimental measurements in the tested

models were compared with rotations

computed from analysis on identical R/C

frames without infills, using RISA2D

259
structural program. Critical sections 1 and 2

are located above the windward and leeward

column base joints, respectively. Sections 3,

4, are located in the beam 4" from the face of

the windward and leeward columns,

respectively.

6.1. Strong Frame Model:

An analytical R/C frame identical to that of the

strong frame model, was subjected to an inplane lateral load. The rotation in sections 1,

2, 3, and 4, were computed to be 0.00257,

0.00250, 0.00345, and 0.00329, respectively.

These results are compared to the test results

shown in Figure 12. The largest rotations

from the test were in section 3, the windward

end ofthe beam, and section 2, the base of the

leeward column, the smallest rotation was in

section 1. This is not unexpected in that with

the direction ofloading, the windward column

was braced by the infill, but on the leeward

side the infill and column separated. The infill therefore created a non-symmetrical moment

redistribution in the frame. One potential

benefit however of the composite action of

the infill and the R/C frame is a redistribution

of the moment caused by the infill from critical

sections to less critical section of the frame

and transformation of some loading moments

to loading in shear. This however can also

have an adverse effect as well. Simple

calculations yielded elastic rotation capacity of

columns and the beam in the strong frame

model on the order of 1 . 04 x 1 0"3 radians. All

measured rotations on the model significantly

exceeded the calculated elastic rotation value.

6.2 Weak Frame Model:

Weak frame rotation varied significantly from

the strong frame and large frames models.

The rotations measured in the weak frame

were large relative to rotations measured in the

in strong frame. The magnitudes of weak

frame rotations did not vary as seen in both the

strong and large frames. The rotation beyond

the elastic limit remained linear. The formed

diagonal strut resulted in shear failure at the

bottom of the leeward column and the

windward end of the beam. It appears that in

the stiffer strong frame, the rotation was

dominated by shear in the diagonal strut, while

weak frame rotation was dominated by

moment.

6.3 Large Frame Model:

The rotations of sections under consideration

behaved similarly to the strong frame in terms

of reduction observed in the rotations at the

beam windward end and the bottom of the

leeward column. It should be reminded that

these same two frames had very similar load

versus drift behavior as well.

7. Summary and Conclusion:

The three models were tested under stroke

controlled monotonic loading. The failure

mechanism of three models compared to

Mehrabi's failure mechanism types and all agreed to one of Mehrabi's type with minor

differences seen in the weak and strong frames

due to the presence of minor cracks from outof-plane loading.

The residual strength and the drift capability of

a model can be used as an index to compare

performance of a model. Exact comparison

among the three models must account for

scaling effects and configuration, which are not accounted for in this paper.

The behavior of these models needs to be

studied beyond the initial strength. Ultimate

260
strength capacity, and residual strength need to

be identified in terms of the models

parameters.

The outstanding composite behavior of R/C

frames with masonry infills, is obviously due

the ability of the models to redistribute loads

that supposedly resisted by the R/C frame or

the infill along. The rotations of sections

where hinges may form, were controlled by the

relative stiffness between the R/C frames and

their infills. Finally, although the all model

constructed from non-ductile components, the

composite behavior exhibited remarkable

ductile behavior, characteristic that is highly

desirable in design of structures in high

seismic zones.

8. REFERENCES:

1. Al-Chair, G., Angel, R, and Abrams, D. P.,

" Dynamic Testing of Unreinforced Brick

Masonry Infills," Proceeding of Symposium on

Advances in Engineering for Concrete and

Masonry Structures, April 24-28, 1994.

2. Mehrabi, B. A., Shing, P. B, and Schuller,

P. M., "Experimental Evaluation ofMasonryInfilled RC Frames," ASCE, Journal of

Structural Engineering MAR. 1996 VOL. 122

No. 3.

Table 1: Summary ofDimensions and Material Pro Derties of Tested Models:

Parameters Half-Scale Strong

Frame Model

Half-Scale Weak

Frame

Large Frame

The height of column, h 35" 35" 56.125"

Length ofthe beam, L 54" 54" 72"

Infill thick x height x width 1.75" x 32" x 48" 1.75" x 32" x 49" 4" x 52.25" x 64"

Masonry Units' dimensions l"x4"xl.75 l"x4"xl.75 4"x8"x4

Masonry unit/ compressive strength clay clay concrete/ 1041 psi

Mortar type/ mortar cube strength N N N/ 792 psi

Column width/depth, Wc , dc 6"x6" 5"x5" 8"x 5"

Beam width/depth, dc 5"x6" 4"X6" 7.75" x 5"

Infill fm (E'm) 5000 psi 5000 psi 190 psi

Concrete fc, E'c 4000 psi 4000 psi 4220 psi 3700 ksi

Beam stirrups 6. ga 6ga. 3/16" dia.

Column ties 6ga. 6ga. 3/8" dia.

Beam Longitudinal Reinforcement 3-#3 (top),3-#3

(bot.)

3-#3 (top),2-#3

(bot.)

3-3/8 " dia. (top), 2-3/8 " dia.

(bot.) !

Column Longitudinal Reinf. 6-#3 4-#3 4-3/8" dia.

261
Reaction Frame

y: Yo-YoGage

R: LVDT

L: Load Cell

S: Stoke

ST : Strain

Figure 1: Push Over Test Setup and Instrumentations

Possible failure L*Ur*l lo.d direction mechanisms

:

A Flexure! B Mldheight crack C Diagonal crack D Horizontal ilip E Comer crushing

Figure 2: Failure Mechanism of Infilled Frame (Mehrabi, 1996)

262
Figure 3 : Strong Frame Deformation After the Figure 4: Strong Frame Hinges on the R/C

Push Over Test. Frame

Figure 5: Weak Frame Deformation After the Figure 6: Weak Frame Deformation After the

Push Over Test (Front Side) Push Over Test (Back Side)

263
264
STRONG FRAME

Load vs Diagonal Deformations

a

•a

(8O

30

24

18

12

> /

/ I Ul

w« if. Tensi

I

on

(0

(So

18

15

12

9

6

3

-2 -1.6 -1.2 -.8 -.4 0 0.4 0.8 Diagonal Deformation, (in)

WEAK FRAME

Load Vs. Diagonal Deformations

1.2 1.6

J D2

, » -

» - -

-1.8 -1.5 -1.2 -.9 -.6 -.3 0 0.3 0.6

Diagonal Deformation, (in.)

LARGE FRAME

Load vs Diagonal Deformations

0.9 1.2 1.5

-.8 -.4 0 0.4 0.8 Diagonal Deformation, (in)

Figure 10: Load Vs. Diagonal Deformation for the Strong, Weak, Large Frame Models

265
STRONG FRAME

Diagonal Desfromation Vs. Drift

Tension

Q 0 0.6 1.2 1.8 2.4 3 3.6 4.2 4.8 5.4 6

Drift, (in.)

c

o

Q

15

c

o

to

1.8

1.2

0.6

0

-.6

-1.2

1.8

WEAK FRAME

Diagonal Deformation Vs. Drift

Compression

ion

[_ i

Tens

\

i

[

j

0.5 1.5 2.5 3

Drift, (in.)

3.5 4.5

c 3

c 2

o

"is 1 a o

i -2 " -3

LARGE FRAME

Diagonal Deformation Vs. Drift

Corner

Tensio

ession n =di—

Q 0 0.6 1.2 1.8 2.4 3 3.6 4.2 48 5.4

Drift, (in.)

Figure 11: Diagonal Deformation Vs Drift for the Strong, Weak, Large Frame Models

266
(sdpO 'peon

o

6

(sdpO 'peon

CO

LLI 5 So

< (0

E

03

o

Io

3

o

>

(QO

o

c

LU

i

c

I

\

I —

§

1 c

h

—

I

1 o

\

8 "o I 2

8 c

d .2

(0 **o

IT

(sdnj) 'pco-j

LU .2

(0

3

O

o

w

c —' o

o -°- «

a:

in

o

(A >

(0o

o

m

cfi

Limit

(C

Elastic

itation

o

8 c

d .2

8 o

o cc 8

d

» S £ 2

(sdi>j) 'peo-j

CN

267
E

©

o o

til *-> 5 (0

o 5 on * «0 < >

UJ _

o

o

o

c

LU

(0

00o

COo

00o

d

CN

COo

COo

d

o

(0o

a:

tfi lO lO

r*« ^ cn m m m

1^ <N*

«8

e

o

o

tomiooiomioo *~ cn cn

(sdn\) 'peon (sdi>|) 'peon

c

o

m 'S

< o

OS (2

u_<

ill ...o

(A

E

(0

©

CO

©

c

LU

"S

to

c

§

COo

d

coqd

ao<ro

d

CN

COo

d

COo

d

AS

o

o

a:

*t> IT) tO

r«-' *~ cm o w m »oCN

(sdpi) 'peon

c

o

Ml *2

< o

a: q;

u_^ co < > UJ _

reo

c

E

O

o

"E

to

C

©

o

E

o

m

T

\

\

00o

fl- eaa

00T

a

d

CN

coa

d

coT—o

d

e

o

CO *»o

Of

lO CN OJ CO CO

(sdi>j) 'peon

C*1

2

3

ao

268
(sdpt) 'peoi

siu.

UJ< cE

zsooofa. CO5c59oEom

(sdpi) 'peon

UJ52

LL

LUO

a:< o C O • 2 "S

CO CO ° I

0) > « a £

CO «_ o o J Eorom

bS -6 ° 1 ins I 8

<0O

od

CN

(sd^) 'peon

UJ

UJO

a:< co

"J

COo

ccw>

COo

-I

E3OUa

COoE©

tSom

o

(0coo

(Sdlif) *peo-j

3b

269

Designs and Techniques for the Restoration

of the Earthquake-damaged

Hanshin Expressway Kobe Route

Hisamori Ohtsuka'\ Hiroyuki Nakajima2

', Hidenao HayashP and Kenji Kosa'

ABSTRACT

The Great Hanshin Earthquake that occurred at dawn

on January 17, 1995, caused tremendous damage to

the Hanshin Expressways. The most severely

affected was the Kobe Route (Route 3) in the Hyogo

prefectural area where piers over a length of 635

meters toppled and four other elevated sections

collapsed. The restoration work is being carried out

in accordance with the "Specifications for the

Restoration of Highway Bridges stricken by the

Great Hanshin Earthquake", following the basic

policies described below.

a) Improve the deformation performance and

strength of individual structures, so that the entire

bridge can withstand seismic force.

b) Reuse as many components as possible by

repairing and reinforcing existing structures. Restoration of a total of 27.7 km of the damaged

section is now in progress. Complete restoration of

the entire Kobe Route is scheduled for October 1996.

KEYWORDS: deformation, ductility, earthquake,

reinforced concrete

has been carried out in accordance with the

"Specifications for the Restoration of Highway

Bridges Stricken by the Great Hanshin Earthquake",

following the basic policies described below.

a) Improving the deformation performance and

strength of individual structures, so that the entire

bridge can withstand seismic force.

The ductility and lateralforce bearing capacity of

substructures have been increased. To reduce the

seismic force acting on substructures,

superstructures have been improved by connecting

girders or making them continuous, installing

isolation or horizontal reaction-dispersing

bearings, and employing light-weight steel decks

for bridge floors.

b) As many components as possible have been reused

by repairing and reinforcing existing structures

depending on the results of damage investigation.

Foundations were reused since they received

relatively minor damage. Piers and girders with

light damage were repaired and reinforced to improve their seismic performance.

1. INTRODUCTION

The Great Hanshin Earthquake that occurred at dawn

on January 17, 1995, caused tremendous damage to the Hanshin area. The Hanshin Expressways also

sustained serious damage. The most severely affected

was the Kobe Route (Route 3) in the Hyogo

prefectural area where piers over a length of 635

meters fell sideways and four other elevated sections

collapsed. Before the disaster, an average of 115,000

vehicles used the Kobe Route every day, accounting

for 40% of the east-west traffic between Kobe and

Osaka. This main traffic artery is now being repaired

and reconstructed, and will be restored to complete

service as rapidly as possible. The restoration work

1) Head, Earthquake Engineering Division, RW.R.I.

2) Section Chief, Design Section, Engineering

Department, Hanshin Expressway Public

Corporation

3) Section Chief, Research and Design Section, Kobe

Route Reconstruction Division, Hanshin

Expressway Public Corporation

4) Senior Engineer, Construction Section, Kobe

Route Reconstruction Division, Hanshin

Expressway Public Corporation

271
Restoration of a total of 27.7 km of the damaged

sections is now in progress. The Kyobashi-Maya

section in the center of Kobe city was put back into

service in February 1996, and 13.3 km of the FukaeWakamiya section will be available this coming

August. Complete restoration of the entire Kobe

Route is scheduled for October 1996. Following the

emergency safety measures taken immediately after

the earthquake, collapsed girders and piers were removed and many damaged super- and substructures

have been repaired, reinforced or rebuilt. The

Corporation is currently working on the last stages of

construction such as pavements, noise barriers and

other facilities related to roadways.

Techniques used in restoring the expressways are described below according to the type of structure.

2. OUTLINE OF THE DAMAGE

The Kobe Route suffered extensive damage during

the earthquake. One 635-meter elevated section was

toppled at the border between Kobe and Ashiya

cities, four other sections collapsed and, as seen in

Table 1, about 60% of the total of 1,106 piers were damaged. Figure 1 shows the areas where major

damage occurred. Emergency investigation was

carried out immediately after the earthquake. Using

the results from this, damaged structures were ranked on a scale between As, A, B, C and D, in accordance

with the criteria specified in the Handbook of

Earthquake Countermeasures for Roads (Disaster

Restoration) which judge the severity of damage

caused during a disaster. Of the piers on the Kobe

Route, 943 are made of reinforced concrete, 15% of

which were ranked As or A, i.e. functional damage,

and 35% ranked B or C, i.e. relatively light damage.

Flexural and shear failure was typically found in

columns at the cut-off point or at ground level.

Although the number of severely affected steel piers was relatively low at 1 1 , buckling or cracking around

manholes and peeling of coatings on welds were

observed in about 80% of steel piers. About 7% of

superstructures fell and buckling of the girder ends was frequently found.

3. DESIGNS OF RECONSTRUCTED

REINFORCED CONCRETE PIERS

Designs of reinforced concrete piers were checked as follows at three different levels.

a) Level 1 : Design of the allowable stress, using a seismic coefficient method in accordance with

"Part V Seismic Design, the Specifications for the

Design of Highway Bridges".

The design horizontal seismic coefficient was

increased from the previous value of 0.2 to 0.25.

b) Level 2: Check on the lateral force bearing

capacity, in accordance with the "Part V, the

Specifications for the Design of Highway

Bridges".

The lateral force bearing capacity, which was not

taken into account in the old design, was checked.

For Group II ground (alluvium and diluvium), for

example, the seismic coefficient used in the check

was 0.85.

c) Level 3-1: Check on the lateral force bearing

capacity in accordance with the Restoration

Specifications.

The stress-strain curve of concrete is different

from Level 2 which takes into account the lateral

force confinement effect of reinforcements. For

Group II ground (alluvium and diluvium), for

example, the seismic coefficient to be used in the

check was ,1.75.

d) Level 3-2: Nonlinear dynamic analysis using the

seismic records of those areas where the maximum

acceleration was observed during the Great

Hanshin Earthquake.

Dynamic response was checked with typical

waveforms for various types of ground, using

records from measuring sites such as the Kobe

Marine Meteorological Observatory and the

Takatori Station of West Japan Railways.

Results of these investigations are shown in Fig. 2,

comparing the old and new designs of a typical pier

cross-section. The diameter of a circular column was

increased to 3.4 m from the previous 3.0 m. The old

hoop ties were 13 mm in diameter and installed at

intervals of 30 cm, but as a result of the above

studies, they were upgraded to 22 mm in diameter

272
and placed at intervals of 12.5 cm.

Before determining the designs and construction

techniques for reinforced concrete piers, studies were

carried out in three categories as described in the tree diagram (Fig. 3): a) experimental studies on the

effectiveness of various reinforcing methods; b)

confirmation of the practicability of techniques such

as adding bars and concrete to piers, through

experiments using full-scale test models prepared in

accordance with the Restoration Specifications; and

c) establishment of the "Construction manual for the

substructures of the Hanshin Expressway Kobe

Route" to specify standard techniques and assure

consistent quality control. Items a) and b) are described below.

4. EXPERIMENTAL STUDIES ON THE

STRENGTHENING TECHNIQUES OF

REINFORCED CONCRETE PIERS

(1) Test Models

Piers were restored by the following two procedures

depending on the degree of damage (Fig. 4): while

reusing existing footings, severely-damaged piers were removed and new ones were constructed;

relatively less damaged piers were repaired and

reinforced. As shown in the experimental conditions

and parameters of Table 2, four types of reinforcing

methods were tested by experiment using a total of

15 specimens.

a) Steel plate jacketing: this is an attempt to increase

the bearing capacity and confinement effect of

existing piers by simply encasing them with steel

plates.

b) Reinforced concrete jacketing: both the bearing

capacity and confinement effect are expected to improve.

c) A combination of steel plate jacketing and

reinforced concrete jacketing: the reinforced

concrete jacket increases the bearing capacity,

while the steel plate jacket installed around it but

not anchored to the footing increases the

confinement effect.

d) Rebuilding of reinforced concrete piers: a new

reinforced concrete pier is unified with an existing

footing by means of post-installed anchors.

To investigate the effect of reinforcement damage on

the pier's bearing capacity, test models for a) to d)

with different degrees of damage were strengthened

with the reinforced concrete jacketing and the

combined jacketing of reinforced concrete and steel

plate. All test pieces were one-sixth scale models of a

real bridge with square cross-sections, and loads were applied to them as indicated in Fig. 5.

(2) Experimental Results and Discussion

Figure 6 shows the typical relationship between the

load applied to a test model and the amount of

horizontal displacement, and Fig. 7 is a typical loaddisplacement envelope linking every first applied

load under the reversed cyclic lateral loading. Results

of both calculations and experiments are shown in

Table 3. In the experiments the maximum bearing

capacities of the standard model were about 24.5 tf and those of the reinforced models, except the model

strengthened by the steel plate jackets, were around

36 tf, suggesting an increase of about 1.5 times in

bearing capacity. One of the steel plate jacket models

was anchored to the footing; it exhibited a maximum

bearing capacity of 35.1 tf, 1.4 times the standard

model. The experimental maximum bearing

capacities closely agreed with the corresponding

analysis values. Figure 8 shows the ductility factors

of the test models, with the displacement when the

load had dropped to 80% of its maximum being taken

as the ultimate value. The experimental values, even

after modification for the loss due to the pull-out of

reinforcements, satisfied the analysis values derived

from the Restoration Specifications. The ductility

factor exceeded 10 for some of the reinforced

concrete jacket models, well above the analysis

values. Figure 9 shows the relationship between the

ductility factor and the hoop tie ratio in the reinforced

concrete jacket models. With the higher hoop tie

ratio, the ductility factor increased from 6 to around

9, suggesting that the increased number of hoop ties

contributes to the improvement of ductility. Taking

into account that the hoop tie ratio is usually limited

to about 0.7% in actual structures, the ductility factor

of 10 might be a reasonable upper limit for normal

reinforcement with reinforced concrete. Figure 10

shows the energy absorption area enclosed by the

load-horizontal displacement envelope of each

model. There is a similar tendency in energy

absorption area with the ductility factor, with the

reinforced concrete jacket model having outstanding

values. Figure 1 1 shows the relationship between the

273
hoop tie ratio and energy absorption ratio in the

reinforced concrete jacket model, where the

absorption ratio increased from 15 to 22 with the

increase in the hoop tie ratio. Increasing the number

of hoop ties contributes to the improvement of

ductility.

5. EXPERIMENTS FOR CONFIRMING THE

PRACTICABILITY OF RECONSTRUCTION

TECHNIQUES FOR REINFORCED

CONCRETE PIERS BASED ON THE

RESTORATION SPECIFICATIONS

Reconstruction techniques were tested for their

practicability using full-scale test models. Two had

rectangular cross-sections of 3 x 3 m and one had a

circular cross-section with a diameter of 3 m. They

were all 6 m tall. Comparison was carried out

between the three methods in Fig. 12 for installing

hoop ties during reconstruction and also between the

other three methods in Fig. 13 for the installation of

intermediate hoop ties. Normal concrete (slump: 8

cm) was compared with flowing concrete (slump: 1

8

cm) containing super AE plasticizer to increase

workability. After these comparative studies, the

following techniques and materials were chosen as

the standard for restoration work:

a) For installing hoop ties during reconstruction, the

4-section (rectangular) and 3-section (circular)

hooking methods which provide the best

practicability, or the 2-section flare welding

method that can be implemented even if main

reinforcements are supported, were used.

b) For adding intermediate hoop ties, two acuteangled hooks bent at one end were used in place of one acute-angled hook at both ends since the latter hook is difficult to insert because of space

constraints.

c) Normal concrete was used since this type proved

successful in tests.

6. RECONSTRUCTION OF REINFORCED

CONCRETE PIERS

A composite structure was selected as a standard

structure for rebuilding reinforced concrete piers.

Reinforced concrete columns were designed to

satisfy the Restoration Specifications, and steel

beams were employed to reduce the construction

time. Figure 14 shows the structure of a rebuilt pier.

Before reconstruction, old damaged piers were removed approximately to ground level (1.5 m above

the top of the footing). After removing the old

concrete from a column, the axial main

reinforcements were erected, reusing the old intact

bars, while installing additional reinforcements with

post-installed anchorages.

Hoop ties with a diameter of 22 mm were then

installed at intervals of 12.5 cm.

The column was erected to a height 2.5 m below the

base of the beam, beam supports were assembled,

and the steel beam was installed. Column axial main

reinforcement bars were inserted through the holes in

the lower flanges at the corners of the steel beam, and

the column was filled with concrete.

7. REPAIR AND REINFORCEMENT OF

REINFORCED CONCRETE PIERS

The combined use of reinforced concrete and steel

plate jackets was selected as the standard method for

strengthening existing reinforced concrete piers. The

steel plate jacket in this technique is intended to increase the ductility and shear bearing capacity of

the pier, so it was not anchored to the footing. Details

of the reinforcement are shown in Fig. 15. The steel

plate jacket was regarded as hoop ties in the design.

The reinforced concrete jacket was designed to have

a thickness of at least 30 cm from the surface of the

existing concrete cover. After impregnating resin into

cracks, the surface of the existing concrete was

chipped. Any loose parts were removed, and bars

which had buckled, failed or yielded during the

earthquake were replaced. The reinforcement bars were then installed by post-installed anchorage.

These additional reinforcement bars were of the same

type and diameter as the existing main

reinforcements, and laid at intervals of between 20

and 30 cm. Hoop ties and then steel plates were

installed. Studs were used to unite the steel plate

jacket to the concrete jacket. Finally, concrete was

injected into the space between the steel plate jacket

and the old concrete. Flowing concrete was used in

this technique.

274
8. REPAIR AND REINFORCEMENT OF STEEL

STRUCTURES

(1) Steel Piers

Steel piers were removed and reconstructed where

columns had been substantially buckled or cracks had

developed around the circumference. Where repair

by partial replacement of elements was judged

possible, the load of the superstructure was

temporarily removed, damaged sections were cut out

and removed, and new elements were installed in

their place. There are several measures to increase

the lateral force bearing capacity of steel piers:

adding ribs to the interior of a pier, placing concrete

inside the pier, and so on. In the restoration of the

Kobe Route, the latter technique was used in

accordance with the Restoration Specifications.

(2) Steel Girders

Reconstructed girders were designed as normal new

girders, but steel decks were employed as the floor

material to reduce the weight of the superstructure.

Main girder buckling was often found where end

cross beams had knee braces, so a full-web structure

is now employed for the end cross beams to reinforce

the girders. Steel girders were made continuous or connected to each other to make the structure more

seismic-resistant.

Steel bearings were replaced with isolation rubber

bearings so that horizontal forces are dispersed and

damped. To help prevent a bridge from falling off its supports, the conventional plate-type device was

replaced with a PC cable type that can also move

perpendicular to the bridge axis.

9. REPAIR AND REINFORCEMENT OF

FOUNDATIONS

Foundations were checked for damage over the full length of their piles by core boring. Although small

cracks were found in some pile caps, the overall

damage was light. Loading tests were performed on

the most severely damaged piles. The results

suggested that their load bearing capacity had not

been reduced. To assure safety, however, cracks were

impregnated with resin so reinforcement bars inside

the piles are protected from corrosion. All Benoto

pile foundations were checked for their bearing

capacity under seismic loads by a method that takes

nonlinear behavior into account, in accordance with

the "Reference for Application of the Specifications

for Restoration of the Highway Bridges Damaged by

the Great Hanshin Earthquake". Five foundations

whose bearing capacity was judged insufficient were

reinforced with additional piles.

10. DESIGNS AND TECHNIQUES FOR THE

REBUILDING OF CONTINUOUS SECTIONS

(1) Fukae Section [a Mushroom-shaped Pier (Pilz)

Section]

There were two sections on the Kobe Route where

structures toppled or collapsed over 635 m and 686

m, respectively. The old structures in these sections were removed completely except for the foundations,

and all were reconstructed. One of the two was the

Pilz section that was widely reported on TV and in

the newspapers, where a 635-meter stretch of 18

spans of the unified super- and substructures toppled.

In the reconstruction design, the existing main

reinforcements in the base of the columns was

ignored since it was considered to have been

subjected to extraordinary stress, so new main

reinforcements 35-mm in diameter are now unified

with the footing by post-installed anchorage. The

shape of the piers was changed from a circular

column of 3 m diameter to a wall-type column 3x6

m (Figs. 19 and 20) to accept post-installed anchors

inserted at 20-cm intervals.

The columns were slightly rounded to match the

landscape. The diameter of hoop ties was increased

from 13 mm to 16 mm, and their interval was

reduced from 30 cm to 12.5 cm. The intermediate

hoop ties were laid at intervals of 60 cm. Steel floor

slabs were employed so that the old, reused

foundations would not be overloaded. Girders were made continuous where possible, and isolation

rubber bearings were used. The new superstructure is

to be 9 spans of twin continuous box girders with

steel decks.

(2) Benten Section (Continuous Rigid Frame

Bridge)

This section is a 686-meter viaduct of 24 spans above

275
1

National Road No. 2. The old structure consisted of

30-meter spans of simple composite or continuous

non-composite I girders with cross beams supported

by independent single reinforced concrete columns

(Fig. 22). Since the columns were sited in the median

strip and sidewalks, they were only 1 m wide and

steel bearings were used to connect them to the

superstructure. The concrete columns failed due to shear during the earthquake, causing three girders to collapse and five girders to buckle. Rigid connection

to the base would not have been sufficient to assure adequate bearing capacity of the existing

foundations, so the restoration design adopted a multi-span continuous rigid frame structure with

isolation bearings installed in the bases of the piers.

The number of continuous spans was set at 19, taking

into account the movement due to thermal stress and

earthquakes and the torsional deformation (the end

piers are single columns) at the ends of a continuous

girder.

11. CLOSING WORDS

As stated so far, the working conditions are tough and

there are many difficult technical problems

surrounding the restoration of the Kobe Route.

However, due to around-the-clock work and

tremendous efforts by the staff as well as contractors,

it has been possible to drastically reduce the

construction period from the initial estimate of three

years. Work is progressing toward complete

restoration of the entire network by October 1996.

The authors express their deep gratitude to all the

people concerned and ask for continuous support and

cooperation.

276
Table 1 Ranking of damaged piers

As A B C D Total

Steel piers 3 8 12 112 28 163

Concrete piers 64 78 102 225 474 943

Total 67 86 114 337 502 1,106

Figure 2 Results of checks on reinforced concrete piers

Experiments on the effects of the reinforcing method

Studies on restoration methodology

Practicability tests on the arrangement

of bars

Establishment of the "Construction

manual for the substructures of the

Hanshin Expressway Kobe Route"

Figure 3 Investigation procedures

277

Table 2 Experimental conditions and parameters

Test model

No.

Reinforcement

technique

Existing pier cross-section

Cross-section

after reinforcement

Reinforcement

ratio Hoon lip

Design horizontal vicfflit* rrv

efficient (Khco)

Reinforcing steel anchorage nncnoragc icngui Remarks

I Existing Damage rank C

2 Existing P - 0.51% Pw - 0.11% 0.78 Damage rank C

3 Existing Damage rank B

4 Existing Damage rank B

5 Steel plate jacket P - 0.85% 1.50

Post-installed

resin anchor

To the footing bottom bar*

The bottom ends of the steel plates are anchored.

6 Steel plate jacket

Square, 60 on x 60 cm

P - 0.61%

Pw - 0.64%

1.41

Vg - 0 cm

(Vg; clearance between the

steel plate and the footing) 7 Steel plate jacket Vg - 5 cm

8 Reinforced concrete

jacket Square, 60 cm x 60 cm Pw - 0.70% 2.17

Post-installed

resin anchor

To the footing bottom bar*

Test model No. 1 reused

after the experiment

9

Reinforced concrete

jacket P-0.61%

Post-installed

resin anchor

To the footing bottom bar*

Test model No. 3 reused

after the experiment

10

Steel plate +

reinforced concrete Pw - 0.65% 1.97

Post-installed

resin anchor

To the footing bottom bar*

Test model No. 2 reused after the experiment; Vg - 5 cm

II Steel plate +

Square, 70 cm x 70 cm

Post-installed To the footing Test model No. 4 reused after reinforced concrete resin anchor bottom bar* the experiment; Vg - 5 cm

12 Reinforced concrete

rebuilding

Before concrete

casting

To the footing bottom bar*

Standard test model

13 Reinforced concrete

ICUUMUIIIg rphnt lr) in a P - 0.61%

Pw - 0.27% 1.42 Post- installed

resin anchor

To the footing bottom bar*

14 Reinforced concrete

rebuilding

Post-installed

resin anchor

Length required from the top bar**

12' Reinforced concrete

rebuilding Pw - 0.58% 2.21 Before concrete

casting

To the footing bottom bar*

2G-resistant test model

* One half the thickness of the footing plus the anchorage length needed for calculation ** The thickness of the covering on top of the footing plus the anchorage length needed for calculation

Table 3 Results of calculation and experiments

Test model

No.

Pier crosssection (cm)Reinforcement

ratio Hoop tie

Experimental results Calculation results

(tf)

8, (mm)

Pu

(tf)

Su (mm)

Condition at the completion of experiment

Py

(tf) (mm)

Pu

(tf) (mm)

Existing piers

1

60 x 60

DIO

(13+15)

p - 0.61%

D6

4 - etc 200 pw -0.11%

20.6 16.6 24.3 51.7 Cracks only

2 19.8 16.6 24.3 51.4 Cracks only

3 20.8 16.8 24.7 87.6 Concrete spalled

4 20.1 16.7 24.7 87.6 Concrete spalled

Steel plate jacket

5 60 x 60

(+ PI t -

1.6 mm)

No. 1 + PII.6 p - 0.85% D6

4 - ctc200 +P1 - 1.6 pw - 0.64%

27.0 16.4 35.1 122.2

Steel plate broken,

bars buckled or broken

26.6 14.1 31.7 51.0 3.62

6 D10

(13 + 15) p - 0.61%

22.9 17.2 26.0 92.5

Steel plate bulging,

bars buckled or broken

9.5 13.2 23.8 63.1 47.8

7 23.1 17.2 25.0 94.9

Steel plate bulging,

bars buckled or broken

9.7 119 24.0 61.1 4.74

Reinforced concrete jacket

8

70 x 70 (Added thickness: 5 cm) DIO

(13 + 15 + 9)

D6 x 4 -

ctc200 + DIO x 2 -

ctc40 pw - 0.70%

28.1 17.0 37.9 147.3 Core concrete fractured,

bars buckled or broken

23.4 10.2 36.3 81.8 8.02

9 27.4 17.0 38.8 96J

Core concrete fractured,

bars buckled or broken 23.5 10.2 36.6 81.7 8.01

Combined use of reinforced concrete and

steel plate jackets

10 70 x 70 (RC t - 50 + PI t -

1.6 mm)

p - 0.61% 4 - ctc200 + 2 - ctc80 + PI - 1.6 pw - 0.66%

27.8 15.0 38.8 113.2

Steel plate bulging,

bars buckled or broken

23.7 10.1 36.4 75.2 7.80

II 26.6 15.0 38.7 94.8

Steel plate bulging,

bars buckled or broken

23.2 10.2 36.0 77.9 7.50

Reconstruction

12

70 x 70

DIO

(13 + 15 + 9) p - 0.61%

D6

9 - ctcl50 pw - 0.27%

28.7 18.4 37.0 112.2

Core concrete fractured, bars buckled or broken

28.8 11.7 35.8 36.5 3.12

13 29.5 18.4 .37.4 128.1

Core concrete fractured, bars buckJed or broken

28.8 11.7 36.1 36.5 3.12

14 31.1 18.4 37.5 118.1

Core concrete fractured, bars buckled or broken

28.8 IIJ 36.1 36.6 3.24

12' 9 - ctc70 pw - 0.58%

28.8 16.7 36.6 120.3

Core concrete fractured, bars buckJed or broken

28.4 11.9 35.8 58.1 4.88

279
• Damage rank As

or A.

• Significantly tilted, causing functional

problems.

Remove existing

piers and reconstruct.

Reconstruction of reinforced concrete

Piers

• Damage rank B,

CorD.

Reuse existing

piers.

Combined use of reinforced concrete and steel plate jackets, or steel plate jacket only

Figure 4 Pier restoration flow diagram

Axial force load actuator (200tf)

Lateral force load actuator C±125tf ±500im)

o

r

o

o

6000

2800

(600:

1335

1400

e

o1

06

Figure 5 The test model showing details of load application

50

Analysis value

Experimental value

-200 -150 -100 -50 0 50 100 150 200

Horizontal displacement (mm)

Figure 6 Load and horizontal displacement

280

Modification for pull-out of reinforcement

1 1 1

i 1 1 ' ' » 1 1 1 1 1 109.8 147.3

0 50 100 150 200

Amount of displacement at loaded point

Test model No. 8

Figure 7 Load-displacement envelope
y- Analysis value A- Experimental value o Modification for pull-out

12

10

1

«

Q

2

0

o

o

I I

OS

U =i.33i'Pw + 5.8

Data from past experiments

5 6 7 8 9 10 11 12 13 14 12'

Test model Nos.

Figure 8 Ductility factor plot

0.0 0.2 0.4

Hoop tic ratio

0.6 0.8 1.0

7

6 ^ 5 E

i 4

CO „ £ 3

CO

2

1

0

1

1

! i

i

i i

—- Analysis value -6 Experimental value --©- Modification for pull-out

Figure 9 Hoop tie ratio and ductility factor

28

24

i i

x V ±

":c i,,.4l-

: 20

* 16

5 6 7 8 9 10 11 12 13 14 12'

Test model Nos.

P. LO

X) ^ &5 8

15 4

0

i i

1 A>

. i A.

j

! A

i >" i

E =12. 41-Pw-j- 12.S9

|A Data from past experiments

0.0 0.2 0.4 0.6 0.8 1.0

Hoop tie ratio ( % )

Figure 10 Energy absorption area plot Flgure n Hoop tle ratio and energy absorption ratio

1) 2-section hooking 2) 4-section hooking 3) 2-section flare welding

1) 2-section hooking 2) 3-section hooking 3) 2-section flare welding

Figl2 Methods of installing hoops ties

281
r -)

1) Acute-angled,

semicirular hooking

2) U-shaped section 3) Acute-angled, section

hooking hooking

Figl3 Methods of installing intermediate hoop ties

Reinforcement bars

Reinforcement bars

Steel plate jacket —

Anchors installed after construction ; e

Hodptles

Hoop ties

main reinforcement

Construction joints finished Anchors installed after construction

60e

1 'I U|J ic)J Iqfj

Figure 14 Reinforced concrete pier reconstruction

Reinforcement bars

Ttr

Existing main reinforcement

i P

1

a"

r Hoop ties

R

R

no

300

Thickness of new concrete

Steel plate jacket

Reinforcement bars

Figure 15 Reinforced concrete pier repair and reinforcement

282
Temporary pedestal

strengthening Joint RC slab Asphalt povement

d4

Motion of girders

Elastic bearing

Figure 16 Connecting girders Figurel7 Bridge restrainers

Figurel8 Injection of cement slurry into piles

3 300 14000

Figure 19 General view of a rebuilt pier

283
6000

110

ro

eg

8•—

i

ro

oo

CM

ID

ro

rvj

O

o

oCM300 5 400

-r

250 8 X 250 = 2000 1 280

300

n8 X 250 = 2 000 250 110

•

i« HI ^ •V"• X• »V'

1.

.

If

•\

•\

•\

1 i-J

1

•

•

•

•I

•I•f

i

\ »— ] • / •/ •/

4

i \

• /

» -*

Figure 20 Arrangement of bars in a column

Figure 21 Artist's impression of the Fukae section

400 8000 800 8000 400 400 8000 800 8000 400

fi A r

Qf ool kanrmn

p A. r L. - - .

sidew ewtUK si LatlT e roa

side'vaR _r lane road

£j g

3 lane roa

Isolation bearjnjilsolation bearing

lane road

Fig. 22 (Existing structure of the Fig. 23 (New structure of the Benten

Benten section) section)

284
Action and Concept on Performance-Based Design and Engineering System in Japan

by

Hiroyuki YamanouchP , Hisashi Okada2)

, Yuji Ohashi

ABSTRACT

This paper describes a framework on performance-based structural design/engineering

system that should be realized in the future, and

introduces a current activity that is underway

toward the new structural design/engineering

system at the Building Research Institute (BRI)

in cooperation with many professional

communities related to building structures.

Also, the paper presents the relationship

between the similar action at the SEAOC and

the BRI in terms of structural performance

concepts. Further, recent workshops on future

design codes with coordination of a study

group of the U.S. is introduced.

INTRODUCTION

The current structural design methods are based

on either strength or ductility, or otherwise the

both of a structure. Thus, overall structural

performance including others than strength and

ductility has not yet become conscious in the

mind of not only a structural engineer but also a

researcher. Namely, the global structural

performance before and after structural design

is not always clearly described by a structural

engineer. For instance, looking at the

structural performance against a specified level

of seismic forces, a structural engineer cannot

say the expected performance such as remaining inter-story drift, damage degree of non

structural elements and lifelines, reparability

and so on. This situation is strongly owing to the current prescriptive design codes that bind a

structural engineer without design flexibility.

The current design codes do not clearly require

target structural performance, whereas design

procedures and rules are definitely and

prescriptively specified as code provisions.

Therefore, under this inflexible system, the

performance of a designed structure cannot be

clearly predicted and described even by a

structural engineer who designed the structure.

Furthermore, structural design itself has

become "calculation" just to keep restrictedly

the prescriptive specifications under code

provisions.

In the future design system, on the contrary,

the target or objective structural performance

should firsdy be defined as a clear picture, and

then in order to attain the performance, a

structural engineer may choose an appropriate

design method and procedure. Thus, the most

important thing toward the future design system

is to facilitate a new design system including

codes, where a structural engineer can

determine target performance and realize it by

choosing an appropriate design procedure. It is

not until this is realized that a structural

engineer can clearly explain the objective and

expected performance of a structure designed

by him or her to a client.

STRUCTURAL PERFORMANCE

The performance of a structure is not limited to structural safety; the word of structural

performance has even wider concepts. That is,

the concepts includes every structure-related

performance; safety, reliability, durability,

human comfort, maintainability, constructibility, aesthetics and so on. Furthermore,

performance against various loads and external

forces should be taken into account The basic

definition and description materials of structural

performance have already been proposed by the

author elsewhere (1).

1) Director, Structural Department Building

Research Institute, Ministry of Construction

(BRI), Tukuba City, 305 Japan

2) Head, Aero Dynamics Division, Structural

Dept., BRI

3) Head, Structural Dynamics Division,

Structural Dept., BRI

285
PERFORMANCE CODE AND SPECIFICATION CODE

In recent years, structural technology of

buildings makes progress rapidly. Various

technology such as new materials, base

isolation system and response control system is

developing.

However, because that current structural codes

of buildings in Japan are regulated for

conventional structural materials and structural

methods and most of the codes are specification

codes that describe in detail exactly what

materials are to be used, the size and spacing,

current codes cannot match the new materials

and new structural methods.

Therefore, it becomes necessary to convert

structural codes from specification codes into

performance codes that prescribe the objective

to be accomplished and to allow broad leeway

to the structural designers in selecting structural

and structural methods that achieve required

performance.

BRI ACTION TOWARD NEW

SYSTEM

Considering significant needs for the new

design system on the basis of performance, the

Building Research Institute, Ministry of

Construction, has initiated a three-year National

Comprehensive R&D Program entitled " New

Structural Engineering System" from the fiscal

year of 1995, in coordination with structurerelated communities such as those of structural

engineers, contractors, industries, academy and

so on.

The most critical mission of the Program is to show possible ways to a new structural

design/engineering world where a structural

engineer can clearly explain the objective and

expected performance of a structure designed

by him or her. For this, a lot of issues must be

studied, discussed and solved; as a result of

preliminary studies for formulating the

Program, the major subjects to be dealt with in

the Program have been identified, and the

committees for the action have been organized,

as shown in Table I.

Therefore, in this project, we examine the next

two technical subjects.

1. Develop performance based structural design

system (Here, we examine not only seismic

performance but also other performance.)

2. Convert specification codes into performance

codes in the Building Standard Law

Further, we examine social subject

3. Propose social system such as qualification

of structural engineers, building confirmation procedure in the Building Standard Law

and so on for performance based design

system.

As shown in Table 1. we organized technical

committee (convener is Prof. Okada of Tokyo

University). The technical committee steers

and manages this project Under the technical

committee, we settled three sub committees.

The tasks of the sub committees are as follows.

Sub committee 1 : Convener is Prof. Akiyama

of Tokyo University.

Subject of the sub committee is "Evaluation of

performance ". This sub committee examines

concept of performance of structure, definition

of performance, measure of performance,

design seismic action, design wind load and so

on. Under this subcommittee, some working

groups are settled.

Sub committee 2 : Convener is Prof. Aoki of

Tokyo Institute of Technology.

Subject of this sub committee is "Object

performance level". This sub committee

examines opinion of clients and users of

buildings, importance category of buildings,

acceptable risk and so on.

Sub committee 3 : Convener is Dr. Yano of

Nikken Sekkei Corporation.

Subject of this sub committee is "Social

system". This sub committee examines

construction procedures, design review system,

qualification of structural engineers, role of

building officials and so on. COMPARISON OF TARGET PERFORMANCE MATRICES PROPOSED

BY BRI/KOZAI-CLUB AND BY

SEAOC

286
Before the initiation of the BRI National

Program, BRI and Kozai-Club (KC) have

already terminated a two-year joint research

work (1992-1994) on a new structural

engineering system based on performance

modeling steel building structures (2). In the

work, definition of performance, objective

performance, criteria for determining performance and so on were broadly studied and

discussed.

As a result of these studies, several matrices

determining target performance for design

(Performance Matrix) have been proposed. As

an example, Fig. 1 shows the proposed

performance matrix for major structural

skeletons (MS) against earthquakes. This

matrix consists of four intensity levels of

earthquake in terms of return periods or

probabilities of exceedance in 50 years, and

structural grades. The structural grade A is of

critical building structures such as fire stations,

hospitals, public buildings and so on. The

grade D would be for temporary buildings and

grade C for ordinary buildings under a minimum performance level. Further, the

grade B indicates a fairly high level of

performance. Each element of the matrix

designates required performance corresponding

to given conditions. The BRI National R&D

Program has well refereed to these results.

Also, in the United States, similar work has

already been done; the Vision 2000 Committee

of the SEAOC has recommended the performance based structural design procedures, as the

results of the phase I work (3). In the work,

the Committee has proposed performance

objectives as a form of matrix. Fig. 2

introduced this performance matrix as it is.

Looking at the above two matrices (Figs. 1 and

2), it can be seen that both concepts are very

similar in spite of the difference in terminology

or wording used in those matrices. Fig. 3

shows this similarity very well; the MSPerformance Matrix shown in Fig. 1 was

transferred into the Vision 2000 form in Fig. 2.

Then, again in Fig. 3, the structural grades A to D are depicted as diagonal lines similar to those

in Fig. 2. By this comparison, it can be said

that in the course of the BRI/KC and SEAOC

activities, they have been considering basically

an almost same thing in terms of performance,

and have reached similar results with somewhat

different expressions. This is a noteworthy

fact.

NEED FOR INTERNATIONAL COORDINATION/COOPERATION

In accomplishing a new structural design world

in each country, it is hoped that international

harmony and consistency in the conceptual

framework on the new system will be attained

considering the progressing borderless world.

For this aim, an "international committee" is

under consideration to be formed in the BRI

National Program.

Partly for an initial step of the above action, a Japan-U.S. Workshop on Seismic Building

Codes (Performance Based Seismic Engineering of Buildings) was held at the Earthquake

Engineering Research Center, University of

California, Richmond, CA. The main

objectives of the workshop were:

1) to discuss the adequacy and efficiency of

present Japanese and U.S. seismic codes in

light of the experience derived from recent

(1985-1995) significant earthquakes

(particularly the 1994 Northridge and the

1995 Kobe earthquakes), and from studies

conducted in the last ten years, and to identify problems whose solutions need improvement,

2) to discuss what can be done to improve

present seismic code approaches, with

particular attention to the interim recommendations for engineering procedures that the

Vision 2000 Committee of the SEAOC has

recommended for obtaining buildings with

predictable and defined performance; and

3) to discuss the development of a work plan

and the organization that will be needed to formulate the basic concepts and framework

that can be used in Japan and the U.S. to develop practical seismic code regulations

whose applications will result in facilities

whose seismic performance will be predictable with a reasonable degree of confidence

under the different types of earthquake

ground motions that can occur during their

service life.

287
The resolutions and recommendations of the

Workshop are fully introduced in Appendix.

The recommendations of the Workshop should

be thought very significant to break through

current blocked situations in the structural

design/engineering world and to make progress

in international harmony related to a future

structural engineering system based on

performance.

Just one year after the 1995 Hyogo-ken-nanbu

Earthquake, and two years after the 1994

Northridge Earthquake, seventh U.S.-Japan

workshop on the improvement of structural

design and construction practice was held. At

that time we held a special meeting with some

members of SEAOC and ATC to discuss the

international coordination and cooperation as second step. In this meeting, the recommendation was re-affirmed and following items were

recommended.

1) Each side should establish a balanced high

level policy/coordination committee with no

more than seven members plus ex-officio

members from participating governmental

agencies.

2) Formal cooperation should be continued

between Japan and U.S. Next meeting

should be held as soon as possible after

formation of the high level committee : preferably within six months from this

meeting.

REFERENCES

(1) H. Yamanouchi, "An Approach to

Performance-Based Design System," the 5

the U.S.-Japan Workshop on the Improvement of Building Structural Design and

Construction Practices, ATC-JSCA Joint

Workshop, San Diego, Sept. 8-9, 1992.

(2) BRI/Kozai-Club Joint Research Committee,

"Report on New Structural Design System,"

Feb. 1995 (in Japanese).

(3) The Committee of Vision 2000,

"Performance Based Seismic Engineering of

Buildings," SEAOC, April 3, 1995.

288
Table 1 R/D Project for Performance Based Design

Development of Performance Based Structural

Design System in Japan

Three-year Project from 1995 Fiscal Year

Main Subjects

• Develop Performance Based Structural Design System

(Not only seismic performance but also other performance)

• Convert specific code to performance code inBuilding Standard Law

• Propose Social System for Performance Based Code

Pre-study of Performance Based Design from 1992 In cooperation with the

Kozai Club (Non-profit corporation consisting of steel makers)

Organization of R/D Project on Performance

Based Design System in Japan

Technical Committee

Convener: Professor Okada: Tokyo Univ.

Sub Committee 1

Evaluation of Performance

Convener : Professor Akiyama : Tokyo Univ.

Concept of performance of structure

Definition of performance

Measure of performance

Design seismic action, design loads and so on,

Under this sub committee, there are some Working Groups.

Sub Committee 2

Object Performance Level

Convener : Professor Aoki: Tokyo Institute of Tech,

Opinion of clients and users of buildings

Importance category of buildings

Acceptable risk and so on. Sub Committee 3

Social System

Convener : Dr. Yano: Nikken Sekkei

Construction procedures

Design review

Qualification of engineers

Role of building officials and so on. An International Committee (under consideration)

289
EQ Level

Return

Period 20 years 100 years 500 years

L 3

1,000 years

Probability of

Exceedance

in 50 years

92% 39% 9.5% 4.9%

e. g. for

Tokyo 0.8m /s* 2m/s* 5m/s 1 7m/s*

"O- S

C3£

3

O

3w *

55

No damage No damage Continued use

Usable after

repair

B

No damage Continued useUsable after

repair

Non

collapse

Continued use

Usable after

repair

Non

collapse

Usable after

repair

Non

collapse

Fig. 1 MS-Performance Matrix for EQ

Fully Operational Operational Life Safe Near Collapse

0)>

c

1(03

O"

Si

"E

CO LU

Fraquant

(43 year)

Occasional

(72 year)

Rare

(475 year)

Very Rare

(970 year)

Fig. 2 Recommended Performance Objectives for Buildings

Earthquake Performance Level (Vision 2000, SEAOC)

No Damage Continued

Use

Usable after

Repair

Non

Collapse

L, O

D

L,

L 2

L, 0 O

Fig. 3 MS-Performance Matrix with Vision 2000 Form

290
U.S. -Japan Workshop on Performance Based Seismic Engineering of Buildings

13-15 November 1995

Earthquake Engineering Research Center

University of California, Richmond, California

RESOLUTION

After two and one-half days of presentations and detailed discussions about the experiences

derived from significant earthquakes which occurred from 1985 to 1995, the participants of the

Japan-U.S. Workshop on Seismic Building Code agreed upon the following findings and

recommendations.

Findings:

a. There should be international cooperation in formulating the conceptual framework for the

development of more reliable seismic codes; Japan and the U.S. should take the lead in

organizing such cooperation.

b. Buildings designed and constructed following modem codes showed improved performance for

life-safety. However

1. Northridge and Kobe were the first major earthquakes in large urban areas;

2. destructive power of ground motions exceeded prior experience;

3. severe damage and/or collapse of residential (non-engineered) buildings was the number one

cause of excessive deaths and injuries;

4. property damage and disruption of activities in urban areas was excessive;

5. public expectations were not met;

6. design and construction quality was not adequate;

7. societal impact was widespread due to failure of infrastructure, making many buildings non

functional;

8. irregular and low-redundancy buildings had problems;

9. steel and RC joints, as well as RC column reinforcement (confinement), need detailing study

and material/fabrication research;

10. non-structural building systems and contents damage were amajor cause of function

disruption;

1 1 . building separation and pile foundation need improvements;

12. inventory of older existing buildings poses the biggest economic/public safety risk;

13. retrofit design and construction procedures need further study and improvement;

14. the concept of weak/soft story needs detailed study and research.

Recommendations

a. Future seismic code of buildings should be performance based.

b. Seismic design and construction should consider all aspects of building engineering.

c. A conceptual framework should be developed. The effort can most efficiendy and effectively be

accomplished by:

1. joint cooperation and coordination initially between the U.S. and Japan, then with other

affected countries;

2. strong balanced partnership between researchers and practitioners;

3. include all relevant disciplines;

4. include multi-hazards.

d. Implementation of the conceptual framework into the development of reliable simple practical

code will require first to identify the research results that are available, and second to clearly

define and prioritize the focused research needed to achieve reliable code regulations. Short

291
term, intermediate term, and long term research and procedures for practitioner involvement must

be included.

e. Adequate education and/or training of engineers, professionals, and technical personnel who take

part in design and construction projects must be considered.

f. Education of and communication with the public, government, and owners regarding the

limitations of current code procedures for earthquake resistant design and construction, and

proposed performance based design, must be started as soon as possible.

g. Formal cooperation should be continued between Japan and the U.S.

U.S. Participants Japan Participants

Craig Barnes Nagahide Kani

Vitelmo V. Bertero Yuji Ohashi

Eric Elsesser Toshio Okoshi

John Hooper Masakazu Ozaki

James Jirsa Yuki Sakai

Jack Moehle Shunsuke Sugano

Richard Phillips Takayuki Teramoto

Roland Sharpe Hiroyuki Yamanouchi

Andrew Whittker Katsumi Yano

292
Composites for Construction

by

John B. Scalzi 1

A new synthetic material is making inroads

into the construction industry. This material

is commonly referred to as Advanced

Composites or Polymer Composites. It is

being used, at present, primarily as a support

material in the repair and retrofit of existing

structures. As a primary/ structural

component it is being used on offshore

structures, pedestrian bridges, and many

non-structural applications. The trend is to

use the material in various ways for highway

bridges, such as repair, retrofit and even new

construction.

The material consists of many fibers of

different types such as: glass, carbon, and

aramids, in a matrix of various binders such

as; epoxies, polyesters, and resins. The

material is produced in many forms and

shapes; rebars, structural shapes, prestressing

tendons, cables, cloth or sheets/ laminates,

tubes, and grids of two and three

dimensional configurations. These products

are produced by the pultrusion process of

pulling the fibers through a bath of resin

material and then they are hardened by a heat

process.

The attractive features of the new materials

for construction are: lightweight, high

strength, non-conducting, non-magnetic, and

corrosion resistant. Some of the materials

have high stiffness and may be used to

increase the stiffness when used in

combination with materials of less stiffness.

The feature of being non-magnetic makes it an ideal material for hospital operating

rooms where non-magnetic material is

desired for the structural material. The

concept is technology transfer from the

aerospace and pleasure boat industries where

these produces have been used as a

structural material for a long time.

Although the National Science Foundation

has been supporting individual researchers

for many years, it was not until 1984 that the

researchers, together, with industry

representatives developed a detailed fiveyear plan of research projects to determine

the mechanical and physical characteristics of

the materials. The purpose of the program is

to guide future researchers to the type of

research projects that fill the need to bring

these products to the construction market.

The plan also would assist to avoid

duplication of investigations. As a result, the

coordinated data would be very beneficial to

develop standards and design specifications

for the materials. These are needed to bring

these products into practical applications.

Each product needs to be researched for

material characterization, ductility,

environmental effects, buckling, bending,

shear, temperature effects, connections,

anchorages, methods of testing the products,

Program Director

Structures and Building Systems

National Science Foundation

293
nondestructive techniques and behavior in

total systems.

An important task for all products, because

of their brittleness, is the philosophy of

designing structures with all ductile

materials, such as, in combination with

concrete. The question of sufficient ductility

is very important for the safety of bridges

and in other types of structures which

endanger the safety of the public. Analytical

methods are needed to predict the behavior

of the total system when using any of the

composites. Many questions need to be

answered before the construction industry

will accept these new composite products.

The planned program is an attempt to guide

the research activity in the right direction.

This research is required to develop

sufficient data to develop the standards and

design specifications for practical

applications. Although many projects are

currently underway at many universities it is extremely important that coordination takes

place.

An international meeting was held in June,

1996 at Bologna, Italy where many

researchers and industry representatives

discussed the research that is needed and

discussed cooperative projects. At this

meeting were engineers from Japan, Europe,

Canada, the United States, and Latin

American countries. This is the first

international workshop of composite

engineers and set the tone for future

meetings.

Advanced composites are the new material

for construction for the 21st century. They

will be used, at first, for repairs and retrofit

of existing structures before being used as new replacements. Primarily, these materials

will be used wherever corrosion resistance is a major concern, thus reducing maintenance

costs and extending the life of the structure.

In the light of these new materials for

construction and the great amount of

research data that is needed, this author

recommends the UJNR Panel on Wind and

Seismic Effects consider the establishment of

a new task committee on composites, or

include the composite material discussions in

one of the existing Task Committee

structure.

294
Arrangement of Kyoshin Net

By

Keiichi OHTANI D, Shigeo KINOSHITA 2\ Hiroyuki FUJIWARA 3)

ABSTRACT

Having said that, however, it is not unreasonable to then say any area in Japan could suffer a

destructive earthquake. The gathering of records

about strong earthquake motions is still too slow

and the official reporting system is still inadequate.

It is therefore necessary to formulate policy for the

placement of strong motion seismographs in

locations that will uniformly cover the country.

Telephone lines need to be used to facilitate the

collection of records, such records being

comprehensively processed in the one location,

published and used. To this end, the establishment

of 1,000 strong motion measurement sites is being

accelerated across the country, and a "Kyoshin

Net" is being constructed, linked by telephone line

with a processing system via a computer located in

a control center. This paper is described the

performance and data collection and release system

of Kyoshin Net.

Key words: Kyoshin Net, Strong-motion

seismograph, Digital recording, Internet

INTRODUCTION

The need to know what type of vibrations occurred

in the various locations struck by an earthquake is regarded ad important information for planning

countermeasures and investigating the cause of

damage following an earthquake. Moreover, to more rationally implement earthquake disaster

prevention measures and earthquake-proof design,

it is essential to be able to estimate the precise

nature of the vibrations in any given area. Having

said that, however, it is not unreasonable to then

say any area in Japan could suffer a destructive

earthquake. The gathering of records about strong

earthquake motions is still too slow and the official reporting system is still inadequate. It is therefore

necessary to formulate policy for the placement of

strong motion seismographs in locations that will

uniformly cover the country. Telephone lines need

to be use to facilitate the collection of records, such

records being comprehensively processed in the one location, published and used.

For this object, we constructed the "Kyoshin Net"

(Strong-motion earthquake measurement network)

which is consisted 1,000 measurement locations

with 3-components accelerometer covered all over

the country, linked by telephone line with a processing system via a computer located in a control center, the National Research Institute for

Earth Science and Disaster Prevention. When a big earthquake will occur anywhere, the control

signal dispatched from control center to the

measurement locations, the recorded signals are

collected to the computer in a center, after that,

the computer is edited and analyzed that data.

After the completion of this network, we can obtain

the good strong motion record with no over-scale.

If we assume that the Hyogoken-Nanbu earthquake

will again occur, about 150 records will be obtained

by this network. It is hope that we can possible to use estimation the real cause of damage and

making the restoration planning through the rapid

opening of the strong-motion data with this

network. And, by the accumulation of data, it will

be possible to upgrade the aseismic design of

structures and the countermeasures of earthquake

disaster prevention.

1) Director, Disaster Prevention Research Division,

National Research Institute for Earth Science

and Disaster Prevention (NIED), Science and

Technology Agency (STA)

2) Head, Earthquake and Volcanic Disaster

Prevention Laboratory, NIED, STA

3) Researcher, Earthquake and Volcanic Disaster

Prevention Laboratory, NIED, STA

295
MEASUREMENT STATION

The Kyoshin Net is a distributed sesimograph

network and it consists of 1,000 stations. The

average station to station distance is about 25 km,

All stations were constructed on free field. The

measurement station is installed on a site 3 meters

square. It commonly consists of a house made by

fiber reinforced plastic, a concrete base on which a seismograph is installed, facilities for electric

power and a telephone line with lighting arresters,

and a fence (Photo. 1). The house is designed to withstand snow of 4 meters depth. Fig. 1 shows

the drawing of the station. The figure of right hand

side shows the detailed drawing of a base for cold area where the temperature becomes less than - 20° C, the base on which a seismograph is

installed is constructed about 80 cm below the

ground surface.

SEISMOGRAPH

The Kyoshin Net used a new developed digital

strong-motion seismograph (Photo. 2). It has a frequency range from 0 to 30 Hz at a sampling

frequency of 100 Hz. This seismograph uses a 24-

bit A/D converter and thus the value of 108 dB

means the effective dynamic range. Fig. 2 shows a block diagram of recording system. A seismometer

is a tri-axial force-balance accelerometer with a natural frequency of 450 Hz and a damping factor of

0.7. Seismic signals obtained from this

seismometer are fed into amplifier and the to analog filter. This analog filter is an antialius filter

for A to D converter and consists of two-stage RC

filter. The output of A to D converter is a seismic

signal with a sampling frequency of 800 Hz

(Converter clock is 1.6384 MHz, thus

oversampling ratio is 2,048.) Next digital filter is a decimation filter and decreases the sampling rate from 800 Hz. to 100 Hz. This digital filter is a three-pole Batterworth filter with a cutoff

frequency of 30 Hz. In the present seismograph,

the design of decimation filter is free, so that w can make a causal recording system. Causality is an

important factor for seismic wave recordings.

Finally, seismic signals are stored on IC memory of

8 MB capacity.

The other specifications are as follows;

Resolution of accelerometer 15 mGal

Resolution of A/D converter 18 bits

Recording capacity Approx. 150 minutes

Communication port 2 ports of RS-232C

Trigger level From 0.1 to 10 Gals

Trigger condition OR/AND combination of

individual channel triggers

Slow code accuracy Within 1 ms

Calibration signal 100 Gals (1 Hz)

Operating temperature -5° C to 50° C

Waterproof Waterproof to 3 meters depth

Weight About 7 kg

Power consumption About 16 W at 100 V AC

Dimensions 180 (H) X 380 (L) X 280 (W)

Fig. 3 shows the overall frequency characteristics

of seismograph. The corner frequency is 30 Hz

and the phase delay at 10 Hz is about 40 degrees.

DATA COLLECTION

The procedures of data collection from each station to the control center are as follows;

1) Data collection at the occurrence of a big

earthquake

We will collect the data as soon as possible. Our

system is consisted by the dial-up means. The data

collect the following 2 steps.

(1) The maximum acceleration in which stored at the header part collect from the station where is be

able to record the seismic signals.

(2) After that, the time history data collect from

the same station as mentioned above.

2) Periodically data collection

We will call the station periodically (one time per

month), and collect the data.

DATA OPENING

The procedures of data opening are as follows;

296
1) Opening the maximum acceleration

When a big earthquake will occur, we will report

the distribution of maximum acceleration (map and

table) within one day after. This information will

sent by Fax. to pre-registered organizations (mas

media, federal and local government and several

specific institutions).

2) Opening the whole data

when a big earthquake will occur, we will open the

whole data which include the time history through

the STA Net (Internet) within about one week.

We also open the whole data collected by the

periodically procedure by same way at the finishing

the data editing. The minimum value of the

maximum acceleration of each record is supposed

to 10 Gals.

DATA OPENING ON THE INTERNET

One main feature of the Kyoshin Net is the

release on the Internet. Fig. 4 shows the home

page of the Kyoshin Net. This page has 4

message boxes such as introduction, how to use, news and others, and 4 data boxes such as site

information, strong motion data, distribution map of

maximum acceleration and characteristics of

seismograph. Namely, we will send 4 kinds of data on the Internet.

We make three kinds of data files, UNIX, DOS and

ASCII (Fig. 5). You can select one of these files. UNIX and DOS files are compressed by the

standard methods. Each file name is constructed

by site code, year, month, day, hour, minutes, point

component and point compression code.

The Internet address of our home page is as

follows;

http://www.k-net.bosai.go.jp/

CONCLUSION

The Kyoshin Net is constructed by the

supplemental budget of 1995. The construction

started in June, 1996 and completed in March, 1996.

Now, we are storing the site information to the

computer system. We will officially open the

Kyoshin Net home page from June, 1996.

There is no restriction on using the Kyoshin Net

information and data. If you use all or part of the

information including strong-motion data on the

Internet, you are requested to acknowledge the

data source and to send a copy of your publication

to our office.

297
Photo 1. Measurement station for the Kyoshin Net

298
Photo 2. Seismograph for the Kyoshin Net

299
Pole

FEP80^

150 100 800 100 150

Cover(FPR )

Concrete base

Base for cold area

At sites where the temperature

becomes less than -20 C, the base on which a seismograph is installed

is constructed about 80cm below

the ground surface.

Fig. 1. Drawing of the station

Block Diagram

Fig. 2. Block Diagram of seismograph

300
Fig. 3. Response characteristics of seismograph

Strong-Motion Data

Please clicfe one of three data types.

UNIX

(file-name ; CCCCCCYYMDDHHm HS, gi)

(file-name: CCCCCCYYMHDDHHBB. NS. lzh)

(file-name : CCCCCCYYMHDDHHKM. HS)

copy of your publication to the following address:

Earthquake and Volcanic Disaster Prevention Laboratory

lafxonal Research Institute for Earth Science and Disaster Prevention

3-1, Tennodai, Tsukuba-shi, Ibaraki-kenn 306, Japan

Fig. 5. Three kinds of data files

301
Fig. 4. Home page of the Kyoshin Net

302
Seismic Stability for Sliding of Rockfill Dams

by

Tadahiko SAKAMOTO1 *, Tadahiko FUJISAWA2)

, Akira NAKAMURA3) and Tomoya IWASHITA4)

ABSTRACT

The Hyogoken-Nambu Earthquake (M. 7.2)

caused no major damage to dams that affected their

safety. Other infrastructures, however, were badly

damaged. We therefore evaluated the seismic

capacity of rockfill dams designed by the current

design method.

We estimated that the maximum horizontal

acceleration at dam site rocks near the earthquake

source fault was about 220 gal from the attenuation of

maximum accelerations observed at dam foundations

during the quake. We conducted a dynamic analysis of

the Minoogawa Dam located near the aftershock

distribution zone (earthquake source fault) using the

observed acceleration records. We then assessed the

accuracy of the dynamic analysis method. Finally,

we made a dynamic analysis and stability analysis of

the rockfill dam models using the current design

method. This showed that the dam body of rockfill

dams constructed by the current design method would

be safe enough for sliding even if the incident

earthquake motion was a little stronger than that

which occurred near the source fault of this

earthquake.

Key Words : Rockfill Dam, Seismic Design,

Earthquake Motion, Dynamic Analysis, HyogokenNambu Earthquake

1. INTRODUCTION

Many infrastructures and buildings were damaged by the Hyogoken-Nambu Earthquake of

magnitude 7.2, which occurred on January 17, 1995.

In the case of dams, special safety inspections were

carried out by site offices within the river reaches

administered under the River Act immediately after

the earthquake at a total of 251 dams. Later,

engineers of the Public Works Research Institute,

Ministry of Construction conducted detailed on-site

investigations, mainly at the dams which suffered

minor damage.

The results confirmed that all the dams

generally behaved well and did not require

protective countermeasures. However, some

slight damage, such as minor cracks in the crest

paving, was found at a few dams. The amount of

drainage water increased slightly after the

earthquake at eight dams, but the total amount was

small and either fell or stabilized later (Matsumoto

etal., 1996).

In this paper, we reevaluated the seismic

capacity of rockfill dams constructed by the current

design criteria, in the light of the serious damage of

other infrastructures and buildings. We estimated

the characteristics and strength of the earthquake

motion near the earthquake source fault from the

records of accelerations observed at the foundation of

dams. We conducted the dynamic analysis of the

Minoogawa Dam, with a height of 47 m and located near the earthquake source fault, using the

acceleration records observed at the dam to confirm

the accuracy of the analysis. We then evaluated the

stability for sliding of rockfill dams against an

incident earthquake motion of slightly greater

magnitude near this earthquake source fault by

dynamic analyses of model dams.

2. EARTHQUAKE MOTIONS AT DAM

FOUNDATIONS DURING HYOGOKENNAMBU EARTHQUAKE

(1) Peak Acceleration

Figures 1 and 2 show the distributions of the

horizontal peak (either of the peak accelerations in the

1) Director General of Public Works Research

Institute, Ministry of Construction

Tsukuba Science City, 305 Japan

2) Director of Dam Department, ditto

3) Former Head of Fill Dam Division, ditto

4) Research Engineer, Fill Dam Division, ditto

303
direction of the stream or of the dam axis, whichever

is greater) and vertical peak accelerations, respectively,

observed at the foundation of dams during the

Hyogoken-Nambu Earthquake. These values include

the values obtained at the gallery under embankment

dams or at the lowest gallery in concrete dams.

The attenuation of the horizontal peak

accelerations observed at the dam foundations during

the earthquake with epicentral distance is shown in

Figure 3, and that of the vertical peak accelerations in

Figure 4. These figures also show peak accelerations

measured on soil sites (Committee for Investigation on

the Damage of Highway Bridges Caused by the

Hyogoken-Nambu Earthquake, 1995). The figures

indicate that peak accelerations at the dam sites

were substantially smaller than those at the soil

sites for the same epicentral distance. It can be

seen that the earthquake motion at the foundation of a

structure constructed on a hard rock site, such as a dam, is considerably smaller than that on a soil site consisting of alluvial or diluvial deposits.

The attenuation of horizontal peak accelerations

at dam sites with distance from the earthquake source

faults is shown in Figure 5, and that of vertical peak

accelerations in Figure 6. Here, the earthquake

source faults, which indicate the Nojima Fault on

Awaji Island and the distribution line of aftershocks

occurring on the day of the main shock in Kobe, are shown in Figure 7. The dam's distance from the

earthquake source fault indicates the shortest distance

from the fault to each dam site. The Hitokura Dam,

at which a horizontal peak acceleration of 1 83 gal was

recorded during the earthquake, is located about 10

km from the fault. An acceleration record at a rock

site besides dam sites was observed at a point 15 m

underground which is located 10 km from the fault. The horizontal peak acceleration (component in the

EW direction) was 213 gal (Izawa et al., 1995). The

elastic wave (P wave) velocity at the point is 2.67

km/s, which is almost equal to those at dam sites. The earthquake motion on a rock site was assessed

from the acceleration records shown in Figure 5 and

those observed in rock sites close to the earthquake

source fault. As a result, it was estimated that the

maximum horizontal acceleration in a hard rock site, on which a dam could be constructed, caused by this

earthquake was about 220 gal (Committee on

Evaluation of Earthquake Resistance of Dams, 1995).

Figure 8 shows the ratio of the vertical peak

acceleration to horizontal peak acceleration at dam

foundations, which ranges from 1/3 to 1/1. As the

peak acceleration rises, the ratio tends to decrease.

(2) Acceleration Response Spectrum

Figure 9 shows the response spectra of the

acceleration in the stream direction observed at the

foundations of 25 dams during the Hyogoken-Nambu

Earthquake at a damping ratio of 10%. The spectra

are normalized so that the peak accelerations are the

unit. The average value of the response spectra for a

natural period ranging from 0.1 to 0.6 seconds is about 2. The response spectra decrease rapidly at natural periods above 0.6 seconds.

(3) Natural Period of Rockfill Dams

Using the acceleration waves observed at eight

rockfill dams, we calculated the ratios of Fourier

spectrum at the crest to that at the foundation, which is a frequency response function. We then obtained the

fundamental natural period of dam bodies from the

predominant frequency of the frequency response

function. Figure 10 shows the relationship between

fundamental natural period T and height of dams H.

The relationship is nearly linear and within the

empirical equation (1) (Okamoto, 1984) , shown in

Figure 10, from the earthquake observations and

vibration tests.

7/[sec]= (0.35-0.65) x^gl (1)

3. EVALUATION OF SEISMIC CAPACITY

OF ROCKFILL DAMS

(1) Seismic Design of Dams in Japan

The structural safety design of dams has been

prescribed by the provisions of the "Cabinet Order

Concerning Structural Standards for River

Administration Facilities, etc." and the "Regulation for

Enforcement of Cabinet Order Concerning Structural

Standards for River Administration Facilities, etc." These stipulate that the seismic design of dams should

be based on the seismic coefficient method.

The concept of seismic design of dams based on

the seismic coefficient method was proposed by Dr.

Nagaho Mononobe in 1925 in the aftermath of the

Great Kanto Earthquake of 1923. Komaki Dam

(concrete gravity dam), which was constructed in

1930, was the first dam designed by the seismic

coefficient method. The seismic design method was

304
put in a statutory form by the publication of the

"Design Criteria for Dams" established by the

Japanese National Committee on Large Dams in 1953.

Later in 1971, the criteria were partly revised by the

publication of the "Design Criteria for Dams

(Revised)". In 1976, the Ministry of Construction

enacted the "Regulation for Enforcement of Cabinet

Order Concerning Structural Standards for River

Administration Facilities, etc." as the legally binding

design standards for dams. In 1981, the design

seismic coefficient was reviewed by the publication of

the "Regulation for Enforcement of Cabinet Order

Concerning Structural Standards for River

Administration Facilities, etc. (Revised)", which has

up to the present been in force. In the "seismic coefficient method," the effect of

the inertia force that is produced by the dam body

vibration is evaluated as a static horizontal force in

proportion to the weight of each section of the dam

body. The proportional coefficient is called the design

seismic coefficient. The lowest value of the coefficient

for rockfill dams is 0.10 to 0.15 in accordance with

seismic activity specified in the "Regulation for

Enforcement of Cabinet Order Concerning Structural

Standards for River Administration Facilities, etc." This stipulates that the safety of a rockfill dam should

be secured by ensuring that the safety factor for

sliding of the dam body is at least 1 .2 in the structural

design based on the seismic coefficient method.

Rockfill dams designed by the seismic

coefficient method have not suffered any damage

affecting their structural safely from any earthquakes

in the past, including the Hyogoken-Nambu

Earthquake. Rockfill dams designed by the seismic

coefficient method are thus judged to have a high

seismic capacity.

(2) Dynamic Analysis of Minoogawa Dam

Minoogawa Dam is the dam showed the greatest

earthquake acceleration record of all embankment

dams during the Hyogoken-Nambu Earthquake. We

analyzed for the dam using the acceleration records

observed at the dam to confirm the accuracy of using

dynamic analysis to estimate the seismic capacity of

rockfill dams.

(a) Profile ofMinoogawa Dam

Minoogawa Dam was completed in 1982. It is a 47.0 m-high zoned rockfill dam with a central earth

core. The foundation rock of the dam is composed

of sandstone and slate of the Tamba group in the

Paleozoic or Mesozoic era and is hard, with its P wave

velocity, Vp, exceeding 4 km/s approximately. This

dam was designed by the seismic coefficient method,

and its design seismic coefficient was 0.15. Figure

1 1 shows its standard cross section.

This dam is located 48 km from the epicenter

and 1 1 km northeast of the east end of the aftershock

zone. Seismometers are installed at the foundation

(gallery) and the crest, as shown in Figure 1 1

.

During the Hyogoken-Nambu Earthquake they

recorded acceleration waves with peaks of about 130

gal in the stream and dam axis directions at the

foundation, and a peak of about 240 gal in the stream

direction and a peak of about 400 gal in the dam axis

direction at the crest. The quake caused no major

damage. Minor transverse cracks occurred on the

crest pavement and the checkpoints on the crest sank

only 11 mm at the maximum in comparison with

measurements three weeks before the quake.

b) Analysis method and conditions (for details, refer to the paper by Nakamura and Iwashita in 1 996)

Dynamic analysis was adopted for the complex

response analysis by the equivalent linear method.

We set the density of each zone of the dam body

based on the results of the control tests conducted

during the banking of the dam, as shown in Table 1

.

The initial shear modulus Go, with average effective

principal stress were set as shown in Figure 12. The

shear strain dependency of the shear modulus ratio

G/Go, and hysteresis damping ratio hh were set as shown in Figure 5. Those of the rock materials were

obtained from large-scale cyclic triaxial tests. The

Go, G/Go and hh of filter materials were set the same

as those of inner rockfill materials (Rock H). We

used the test results for the core materials of another

rockfill dam as the Go, G/Go and hh of the core

material. We set the Poisson's ratio using the

empirical expression (Sawada et al., 1975) based on

measurements of rockfill dams. As the incident

earthquake waves, we used the acceleration wave,

shown in Figure 14, observed at the gallery of

Minoogawa Dam during the Hyogoken-Nambu

Earthquake.

To take into consideration energy losses because

of internal viscous damping, uneven incident waves at the dam foundation, dynamic interaction between the

dam body, foundation and reservoir and so on, an

equivalent damping for these effects must be added to the hysteresis damping ratio hh of the dam body

material. We conducted dynamic analyses for two

cases where equivalent damping ratio of 15% and

305
10% was added to hysteresis damping ratio hh.

c) Results ofdynamic analysis

We compare the analyses and the observation of

the time histories (shown in Figure 1 5) and the Fourier

amplitude spectra (shown in Figure 16) of the

response acceleration at the crest. In Figure 15, the

analysis response wave of the stream direction

component agrees well with the observation in the

first half of the principal motion (t = 4 to 5.5 sec).

The amplitude of the analysis response wave becomes

smaller than that of the observation in the second half

of the principal motion. The amplitude of the

analysis response wave of the vertical direction

component is smaller overall than that of the

observation. In Figure 16, the Fourier spectra of the

analysis in the stream direction component are almost

equal to that of the observation at frequencies up to

1.2 Hz. The spectrum of the analysis is larger than

that of the observation at frequencies of 1.2 to 2 Hz

and is smaller above 2 Hz. The Fourier spectra of

the analysis in the vertical direction component are smaller than that of the observation above 3 Hz,

whereas both of the spectra are almost equal up to about 3 Hz. The disagreements between the analysis

and observation are largely caused by threedimensional factors such as incident waves of other

directions and the topography of the dam site. The

observed accelerations in the stream direction are considerably larger in the second half of the principal

earthquake motion, which agrees with the time zone when the acceleration in the dam axis direction at the

crest exceeded 300 gal. This indicates the great

effect of three-dimensional vibration.

The comparison of both the cases of adding a damping ratio of 10 and 15% to the hysteresis

damping ratio hh in Figures 1 5 and 1 6 showed that the

analysis result in the case of adding 15% agrees better

with the observation, accurately reproducing the

behavior of the dam during the earthquake.

(3) Seismic Capacity of Rockfill Dams Close to Source Fault of Hyogoken-Nambu

Earthquake

a) Incident earthquake wave

As noted in section 2, the maximum horizontal

peak acceleration at the rock foundation of dams

during the Hyogoken-Nambu Earthquake is estimated

to have been about 220 gal. Incident waves with a

horizontal maximum acceleration of 250 gal were,

therefore, used in the dynamic analysis to provide an

adequate margin of safety. These waves can be

considered to be greater than the amplitude of

earthquake motions close to the source fault of the

Hyogoken-Nambu Earthquake.

We selected as the incident earthquake waves

four acceleration records (ACC-1 to ACC-4) observed

at the dam foundations near the epicenter, the peak

acceleration of which was raised to 250 gal. Our

selection took account of the fact that the proportion

of rise in the amplitude should be as small as possible

and that the selected acceleration waves should have

various frequency properties. The vertical

acceleration waves were to be raised in the same

proportion as their corresponding horizontal ones. Here, we used the incident waves without

modification of their frequency characteristics and the

duration of the observations. Figure 17 shows the

acceleration response spectra of the four selected

incident earthquake waves with a damping ratio of

10%.

b) Dynamic analysis conditions

We prepared dam models to evaluate the seismic

capacity of rockfill dams with reference to Shichigashuku Dam (completion in 1991, height of 90

m, design seismic coefficient of 0.15) designed in

accordance with the current design method. The

finite element mesh of the dam model is shown in

Figure 18. The gradient of the upstream slope is 1 : 2.6 and that of the downstream slope is 1 : 2.0. The

heights of the dam models to be adopted are 63, 110

and 150 m. The water should be stored up to the

normal water level. Direct time integration using the equivalent

linear method was used for the dynamic analysis. In the dynamic analysis, the physical properties of the

model are assumed to be only rock materials, which

account for the greater part of dam bodies. The

physical properties of the material for the analysis were set on the basis of the results of the control

tests during construction and large-scale triaxial tests

for the Shichigashuku Dam. We set the wet density

at 1880 kg/m3

, saturated density at 2080 kg/m3 and

Poisson's ratio V at 0.35. The initial shear modulus

Go was set according to the depth D from the surface

of the dam as shown in Figure 19. Figure 20 shows

the curves representing the shear strain dependency of

the shear modulus ratio G/Go and the hysteresis

damping ratio hh of the rock material. The used

damping ratio was obtained by adding 15% to the

hysteresis damping ratio of the rock material, based on

the comparison study described above for the

306
observed and analytical response of the Minoogawa

Dam during the Hyogoken-Nambu Earthquake.

Rayleigh damping was used considering the

dependence on frequency (Iwashita et al., 1995).

c) Resisting acceleration for sliding

We conducted convergent static stability

analyses for the potential circular slide masses as shown in Figure 18. We calculated the accelerations

acting on the slide masses (resisting acceleration ares)

that would cause the safety factor Fs against sliding to become 1.0 as expressed by equation (2). The

resisting acceleration ares is the horizontal acceleration

(force) that would cause the potential slide mass to begin to slide.

N

Ores m x rvi m J = Fs = 1.0 (2)

Where, i / is the shear strength of the dam body

materials, r is the radius of the potential slide circle, L

is the length of circular-arc of the slide mass, i is the

element number, N is the number of elements in the

potential slide mass, and r vi is the vertical height from

the center of the slide circle to the center of gravity of

element i. We analyzed the stability for sliding of only the

upstream slope, because the safety factor against

sliding in the downstream slope is usually greater than

that in the upstream slope. The simple slice method

was used for the stability analysis.

Table 2 shows shear strength parameters of the

materials for the stability analysis. The shear

strength of the rock material was decided from the

results of large-scale monotonic loading triaxial tests using the rock materials carried from the quarry sites. We adopted the shear strength X /of the coarse grain

materials (rock and filter) approximated by the

equation (3) as an exponential function. Equation

(3) expresses the dependence of internal friction angle

0 on normal stress o n acting on the sliding surface.

Tf = Po- Ax (on I Po)1 (3)

Where, Po is 1 MPa.

(d) Evaluation ofstabilityfor sliding

The seismic inertia force acting on a potential

slide mass is expressed by the product of the mass and

the acceleration of the slide mass. We evaluated the

stability for sliding using "peak average acceleration

of slide mass", a(t)max. The peak average

acceleration of slide mass is the maximum value

during the duration in the sliding direction of aave(t)

which is the value of the peak seismic inertia force

moment divided by the mass moment of each slide

mass by equation (4):

N / N

Ctave{t)= X fl //;(/) X /V;) / ^(mxrvi) (4)

; = 1 i=l

Where, aHi(t) is the horizontal response acceleration

time series acting on an element i by dynamic analysis.

Figure 21 shows the distribution of peak average

accelerations of slide masses with different y/H for

three heights of dam model. This figure also shows

the distribution of resisting accelerations for each

potential slide mass ares. Figure 21 shows that the

peak average accelerations are large values in the

upper part of the dam, but the values are within the

resisting acceleration in all cases. This indicates that

rockfill dams are safe against sliding. Moreover, the

peak average acceleration of slide mass, which

represents the maximum value during the duration of

an earthquake, does not continue but acts on the dam

body only in the peaking instant. Therefore, rockfill

dams are actually safer than the results shown in

Figure 21 suggest.

4. CONCLUSIONS

The Hyogoken-Nambu Earthquake of January 1 7,

1995, which was an intra-plate earthquake, occurred

just below urban areas causing severe damage in the

Hanshin region and on Awaji Island. However, no dam was seriously damaged.

The peak accelerations of earthquake motion

observed at rock sites of dams were sufficiently

smaller than those on soil sites consisting of alluvial or diluvial deposits. From the attenuation of the peak

accelerations observed at dam foundations, we

estimated that the maximum horizontal acceleration

that would have been caused close to the earthquake

source fault was about 220 gal. The construction of

dams on hard bed rock sites is considered to be one of

the reasons why this earthquake did not inflict major

damage on any dam.

We conducted a dynamic FEM analysis of the

Minoogawa Dam located near the earthquake source

fault using the observed acceleration wave. The

response acceleration wave of the analysis reproduced

307
the observed wave with good accuracy.

We evaluated the seismic capacity of rockfill

dams designed by the current design method based on

the seismic coefficient method. We prepared FEM

model dams and four incident earthquake waves based on the observations and conducted the analysis using

them. The analysis results confirmed that the rockfill

dams designed by the current design method have

sufficient safety for sliding even against the magnitude

of shaking that occurred close to the source fault of

the Hyogoken-Nambu Earthquake.

Dams, as large-scale structures, must have

sufficient stability even during earthquakes. The

seismic design technology for dams and the system for

observing earthquakes must be improved.

ACKNOWLEDGMENTS

We would like to extend our deep gratitude to the many persons of Osaka and Hyogo Prefectures,

Water Resources Development Public Corporation

and Kansai Electric Power Corporation for their

cooperation in collecting earthquake acceleration

records.

REFERENCES

Committee for Investigation on the Damage of

Highway Bridges Caused by the Hyogoken-Nambu

Earthquake, with T. Iwasaki as the chair (1995):

Report on the Damage ofHighway Bridges Caused by

the Hyogoken-Nambu Earthquake (in Japanese)

Committee on Evaluation of Earthquake Resistance of

Dams, with C. Tamura as the chair (1995): Report of

Committee on Evaluation ofEarthquake Resistance of

Dams (in Japanese)

Development Division, River Bureau, Ministry of

Construction (1991): Seismic Design Standards for

Embankment Dams (Draft)

Izawa, S. et al. (1995): Measured Earthquake Records

in a Rock Site During Hyogoken-Nambu Earthquake,

Architectural Institute of Japan Kinki Chapter

Architectural Research Meeting (Structure) (in

Japanese)

Iwashita, T., Yasuda, N., Nakamura, A., Matsumoto,

N., Takeda, O. (1995): Dynamic Analysis of Rockfill

Dams Considering Frequency Independent

Characteristics of Radiation Damping Ratio, First

International Conference on Geotechnical

Engineering, Vol.1, Balkema, pp. 511-516

Matsumoto, N., Nakamura, A., Sasaki, T., Iwashita, T. (1996): Effects on Dams, A Special Issue ofSoils and

Foundations on Geotechnical Aspects of the January

17 1995 Hyogoken-Nambu Earthquake, JGS, pp.273-

281

Nakamura, A. and Iwashita, T. (1996): Behavior and

Stability of a Rockfill Dam by Dynamic Analysis

During the 1995 Hyogoken-Nambu Earthquake,

Proceedings of Conference on the Great Disaster in

Hanshin and Awaji, JSCE, pp.44 1-448 (with

English summary)

Okamoto, S. (1984): Introduction to Earthquake

Engineering, pp. 474-479, University of Tokyo Press

Sawada, Y. and Takahashi, T. (1975): Study on the

Material Properties and the Earthquake, Proceedings

of the fourth Japan Earthquake Engineering

Symposium, pp. 695-702

308
Table 1 Properties of materials of the Minoogawa

Dam model

Wet Saturated

Zoning Density Density

P /(kg/m

3 ) p ^/(kg/m

3 ) Rock I 2120 2320

Rockll 2030 2270

Filter 2230 2350

Core 2180 2180

Table 2 Shear strength parameters for stability analysis of model dams

Zoning

Wet Density

(kg/m

3 )

Saturated Density

(kg/m 3 )

Strength Parameter*

A (or c) b (or CP)

Rock 1880 2080 A=1.128 b=0.804

Filter 2130 2240 A=0.808 b=0.908

Core 2220 2230 c=98 kPa CP=35°

* rf = Po-A-(g n/Po)

b or rf = c + cr n tan <£>, where rf is the shear strength,

Po is 1 MPa, crn is the normal stress on the sliding plane, c is cohesion,

and 0 is internal friction angle.

309
Figure 1 Horizontal peak acceleration at dam foundations

Figure 2 Vertical peak acceleration at dam foundations

310
COc

g

'to i— _g

O)oo

<

CO

CD 0_

"cc

-*—

'

c

oNd

oX

1000

500

300

200

100

50

20

10

5

1

o

0 o o

0 0 o %„

o

DAM SITE

stream direction

dam axis direction

SOIL SITE

o

°o,

'o

3fr n

5 10 20 30 50 100 200 300

Horizontal Distance to the Epicenter(km)

COc

g

to_g0u

o<

CL

75g

'C>

1000

500

300

200

100

50

20

10

5

2 h

1

DAM SITE

SOIL SITE

o

o

o o

CP

o

8?o

10 20 30 50 100 200 300

Horizontal Distance to the Epicenter (km)

Figure 3 Attenuation of horizontal peak acceleration

with epicentral distance

Figure 4 Attenuation of vertical peak acceleration with

epicentral distance

I 100 c

Q

16

In

CO8

<

COCD 0.

c

o

,NOI

0 50 100 150 200 250 300

Horizontal Distance to the Earthquake Source Faults (km)

0 50 100 150 200 250 300

Horizontal Distance to the Earthquake Source Faults (km)

Figure 5 Attenuation of horizontal peak acceleration

with distance to the earthquake source fault

Figure 6 Attenuation of vertical peak acceleration with

distance to the earthquake source fault

311
0 20 KM

1 1 I

200

cd

o 150

CDO

O

<

CO

CD Q_

100

16 50

o

CD >

0

0 50 100 150 200

Horizontal Peak Acceleration (gal)

Figure 7 Earthquake source fault Figure 8 Ratio of vertical peak acceleration to horizontal peak acceleration

E3k.O

0)D. to

d)

<oc

o

Cl

</> 0) occ

o

«)

aio

o<

I

(0Ei—oZ

10

0.01

0.001

h=10%

Record of Each Dam Mean Value

0.01 0.1 1 Period (sec)

10

0.8

o

CDin

0.6

a"

o

0) a.

15 0.4

ra Z

15c 0.2

tt> E

ro 73c

3

0

ROCKFILL DAM after Okamolo

0 20 40 60 80 100 120 140

Height of Dams, H (m)

Figure 9 Acceleration response spectra at dam

foundations

Figure 10 Relationship between natural period and

dam height

312
EL. 337.0m

Figure 1 1 Cross section of the Minoogawa Dam

1,000

CO Oo

CD

CO a

o

to

<x> s:COH

'c

10 20 50 100 200 500 1,000 2,000

Average Efective Principal Stress, crm (kPa)

Figure 12 Initial shear modulus with average effective

principal stress for the Minoogawa Dam model

o

CDO

o"

aj a.

in _2

"DO

CO

cu _cCO

I0 3 10- io

Shear Strain, y

0 I

10*

Figure 13 Shear strain dependency of shear modulus

ratio and hysteresis damping ratio for the Minoogawa

Dam model

CO

O

w

CDO

U

<

Stream Direction

llj^Vv/^1 ^'1 ^1 ^^

CO so U)

2 o 2

8 * 3

10

(sec)

Vertical Direction

| ||^|yi/^^

10

(sec)

Figure 14 Incident acceleration waves observed at

gallery of the Minoogawa Dam

313
300

200

(0

o

1

o o

-100 -

-200

-300

Observation

Analysis

h=hys+15%

h=hys+10%

Stream Direction

(sec)

(sec)

10

Figure 15 Response acceleration waves at the crest of the Minoogawa Dam

250

E

i

o.

La

0) •co

LL

200 -

150 -

100

50

Stream Direction

Observation

Analysis

h=hys+15%

h=hys+10%

sad

4 6 Frequency (Hz)

10

Vertical Direction

Observation

4 6 Frequency (Hz)

10

Figure 16 Response Fourier spectra at the crest of the Minoogawa Dam

314
Figure 1 7 Acceleration response spectra of incident acceleration waves for dynamic analysis of model dams

Figure 1 8 Finite element mesh of model dam and potential slide masses

315
Figure 19 Distribution of initial shear modulus with Figure 20 Shear strain dependency of shear modulus

depth from the slope surface ratio and hysteresis damping ratio for model dams

Peak Average Acceleration of Potential Slide Mass (gal)

Figure 21 Peak average accelerations and resisting

acceleration of potential slide masses for three heights

of dam models

316
SUMMARY OF JOINT COOPERATIVE

RESEARCH PROGRAMS
I

I

i

!

I

i

i
U.S.-Japan Cooperative Earthquake Research Program

on Composite and Hybrid Structures - Research Progress and Current Status

by

Subhash C. Goell and Isao Nishiyama^

ABSTRACT

A five-year research program on Composite

and Hybrid Structures as Phase 5 of the U.S.-

Japan Cooperative Earthquake Research

Program was recommended to be initiated in

1993 in both countries. The research work in

Japan started in fiscal year 1993. However,

fuller participation of researchers on both

sides started in early summer 1995. The

sponsorship of the program is by the National

Science Foundation in the U.S. and by the

Ministry of Construction along with a number

of industry groups in Japan. This paper

presents a brief status report of the program

and research progress made todate on both

sides. Because of diverse and broad scope of

the subject area, the research program is organized into the following four groups:

Concrete Filled Tube Column Systems (CFT);

Reinforced Concrete (RC) and Steel

Reinforced Concrete (SRC) Column Systems

(RCS); and RC/SRC Wall Systems (HWS); New

Materials, Elements and Systems (RFI). A

theme structure with well selected layout,

geometry and design loads for the research is also presented, which provides a common

focus for various systems to be studied, and

also a common prototype structure from

which the components and sub-assemblages

are drawn.

Key Words:

Concrete Filled Steel Tube;

RC Column with Steel Girder;

RCS Core Wall with Exterior Steel Frame;

New Materials Elements and Systems.

1. INTRODUCTION

The U.S.-Japan Cooperative Earthquake

Research Program began in 1979 following

the recommendations as outlined in the final

report of the U.S.-Japan Planning Group for

the program (Ref. 1). The overall objective of

the total program is to improve seismic safety

practices in both countries through

cooperative studies to determine the

relationship among full-scale tests, small-scale

tests, component tests, and related analytical

and design implication studies. First four

phases of the program have been Reinforced

Concrete (RC) Structures, Steel Structures,

Masonry Structures, and Precast Concrete

Structures. Research on Mixed Steel/RC

Structures was identified as an important phase

of the program as recommended in the

planning group report.

It is widely recognized that innovative uses of

two or more different materials in a structure

leads to more efficient system for resisting

seismic forces. During the past ten years the use of composite and hybrid structures has

increased in the U.S. and Japan. In spite of

some research and development work by

Japanese construction companies, not enough

is known at present regarding their seismic

behavior or performance. Design procedures

and codes for use in typical design offices are not well developed at present.

Considering the importance of developing

design guidelines (a unified code

development) for typical composite and

hybrid structures that are used in current

practice, and of developing new and innovative

composite structural elements and hybrid

systems using advanced new materials and/or

devices, a five-year research program on

Composite and Hybrid Structures was recommended as the fifth phase of the

ongoing U.S.-Japan Cooperative Earthquake

Research Program. The recommendations

were based on a number of technical meetings

of the U.S. and Japan Planning Groups and

finally a Joint Planning Group Workshop held

in Berkeley on September 10-12, 1992 (Ref.

2).

1 Professor, Dept. of Civil and Environmental

Engrg., University of Michigan, Ann Arbor,

MI 48109-2125, USA

2 Coordinator for International Research

Cooperation, Building Research Institute,

Ministry opf Construction, Tsukuba, Japan.

319
Because of diverse and broad scope of the

subject area, the research program is organized

into the following four groups: Concrete Filled

Tube Column Systems (CFT); RC/Steel

Reinforced Concrete Column Systems (RCS);

and RC/SRC Wall Systems (HWS); New

Materials, Elements and Systems (RFI). A

theme structure provides a common focus of

the program to facilitate cross comparison

between various system types identified and to

derive structural elements and sub-assemblages

for detailed studies.

2. PROGRAM MANAGEMENT,

COORDINATION AND CURRENT

STATUS

The research program is planned for a period

of five years in both countries, and funding

has been provided accordingly. Research

work on the Japanese side started in the spring

of 1993. However, fuller participation and

coordination of research work on both sides

started in early summer 1995 when the first

group of U.S. research projects were awarded.

Cooperation and coordination of work by all participants are most essential to successful

completion of this program. Various

committees have been formed for this

purpose. A Joint Technical Sub-Committee

(JTSC) in each of the four components of the

research program provides technical advice

and coordination. Each JTSC has a co- chairman from each side with membership

including all researchers in that area. These

groups meet as often as needed. All

participants and institutions are also part of the

bigger Joint Technical Coordinating

Committee (JTCC) to review progress and

discuss scientific and technical issues on a common basis. This committee meets once a year with the last (second) JTCC meeting held

in June 1995. The third JTCC meeting is scheduled to be held in November 1996 soon

after the award of second year U.S. research

projects. The individual JTSC groups will

hold their meetings in the summer 1996. A

smaller group called Joint Steering Committee

(JSC), which consists of key representatives

from the JTCC, oversees the entire program

and provides guidance on issues that are common to the four components of the

program. The overall program has a technical

coordinator and a co-chairman from each

side. The management chart with principal

supporting groups for the program is shown in

Figure 1. The progress of the research

program as reported in this paper is based on

proceedings of the latest US-TCC meeting

held on February 7, 1996.

In addition to sharing of technical information

and ideas among researchers, exchange of

personnel involved is strongly encouraged.

This will occur more now as the research

projects on both sides are well under way.

Active participation of practitioners and

various industry representatives in planning,

coordination and execution of the research

work is also a strong feature of this program.

3. PROGRAM OBJECTIVES

Since the subject area of the Composite and

Hybrid Structures is diverse and broad, the

research program was organized into the

following four groups: CFT Column Systems;

RC/SRC Column Systems; and RC/SRC Wall

Systems; New Materials, Elements and

Systems. The first three groups aim at developing practical design guidelines (a

unified code development) for typical

composite and hybrid structures currently

used. The last group aims at developing new

and innovative composite structural elements

and hybrid systems using advanced new

materials and/or devices.

4. THEME STRUCTURE

The theme structure provides a common

focus and a strong link between various

components of the research program, facilitate cross comparisons between various systems to

be studied, and also provides a common

prototype structure from which the structural

elements and sub-assemblages for detailed

studies are drawn. The following criteria were

used for selection of the theme structure:

- The building plan was selected so as to

reflect the benefits that the hybrid structures

provide over non-hybrid structures.

- The basic plan is common to various types

of hybrid structures. The number of stories

and the make-up and combination of

structural components and details can be

varied.

- Rules to determine dead load, live load and

seismic forces to be used in design

simulation are common in the U.S. and

Japan Groups.

The basic plan selected for the program is representative of typical office buildings. As

shown in Figure 2, it provides office space of

320
12.8 meters in span length outside the central

core. The central core is used as utility space

for elevators, stairs, etc. Floor area of the

central core is determined on the basis of a twenty story office building. The core area may be changed with the number of stories if

desired. Six types of theme structures are

selected as shown in Table 1. The reasons for

selecting numbers of stories as 3, 6, 9, 10, 12,

24 or 40 for different structure types are as follows:

- Six, twelve and twenty four stories have been

used frequently in Japan for RC, SRC and

steel construction, respectively.

- Six is selected as the standard (key) number

of stories for RC (OPEN)-06 structure.

Three and nine are selected for comparison.

- Twelve is selected as the key number for (SSRQ+W-12 (S-RQ+W-12 structures.

- Twenty four is selected as the key number

for CFT(OPEN)-24 structure. Ten and forty

may be necessary for comparison.

- Buildings more than forty stories may be

investigated under New Materials, Elements

and Systems.

Plan dimensions and layout for the above six

structure types are shown in Figures 3-8.

5. RESEARCH ON CFT SYSTEMS

5.1 Research Progress in Japan

A number of experimental and analytical

studies on CFT systems have been carried out

and reported in the past. A literature survey of

those studies and available data has been

conducted and the summary report and data

preparation is under way. Further

investigations have been planned and are in

progress to verify and expand on the existing

knowledge. Stub-column tests on circular and

square sections under concentric and eccentric

loading have been completed. Half of the

planned 28 beam-column specimens have

been tested. Concrete and steel strengths and

tube wall slenderness are the main variables. A

number of trial designs of the theme structure

of 10, 24 and 40 stories have been completed

and design comparisions made. Results

obtained on member tests will be used in the

design implication studies to derive more

rational models for analysis and design. Main

items of study are: confinement effects and

load shared between the concrete and steel.

5.2 Research Progress in U.S.

A number of research projects were funded

during the first year. These projects are focusing on the behavior of CFT column-toWF beam moment connections. The

specimens are nearly full size and include

circular and square tube sections for columns.

The connection details include the use of

diaphragms, shear studs, welds, and bolts. The

main variables are the connection details to

transfer beam forces into the column. The

objectives are:

1) To determine the force transfer

mechanism, examine the effects of various

connection details on this mechanism, and on

the connection strength, stiffness and ductility

under cyclic loading.

2) Develop and evaluate practical and

economical connection details for seismic

resistance.

3) Formulate practical design guidelines.

Some testing work also includes study of

beam-columns under combined axial force

and bending. Analytical studies accompany

the experimental work. The analytical models

include complex finite element formulations

in an effort to develop simpler empirical or mechanics based models for practical design

work. Analytical work is also aiming at studying the bond and shrinkage effects, and

confinement effects in CFT members

subjected to different load conditions.

6. RESEARCH ON RCS SYSTEMS

6.1 Research Progress in Japan

In recent years the Japanese practice in RCS

structures has moved from using heavy steel

shapes encased in RC for columns to no

structural steel in RC columns with steel beams

passing through the RC columns. A data base

of over 300 tests on RCS sub-assemblages with

30 different joint details has been prepared.

However, most of those tests were carried out

on strong column/connection-weak beam

combinations which give little information on

the force transfer mechanisms in the joint

panel regions at large ductility levels.

Therefore, tests in this program focus on weak

joint specimens in order to obtain information

on strength and ductility of the beam-tocolumns. Bond link tests have been completed

and modeled into 3-D FEM. Through column

and through beam tests on beam-to-column

321
subassemblages have been half completed, and

3-D FEM analyses are in progress. Structural

design of 6 and 12 story moment frames of

the theme structure has been completed

including the inelastic dynamic analyses.

6.2 Research Progress in U.S.

RCS structures in the U.S. generally consist of

columns made of light steel shapes encased in

RC. The light steel shape provides ease of

erection and connection to the steel beam.

Some earlier tests on RCS interior frame

connections have been carried out and

reported in the literature. In one research

project under this program exterior

connections are being studied. The joint detail

variables are: lateral ties, beam bearing plates, use of FRC in place of stirrups. Effect of floor

slab is also included in the study.

7. RESEARCH ON HWS SYSTEMS

7.1 Research progress in Japan

The structural system of core wall with exterior

steel frame is not too common in Japan and

the design methodology is not well establishes

at present. Therefore, trial designs of 6, 12

and 24 story theme structures with RC/SRC

coupled wall have been carried out by using

existing practice. The response of the

resulting structures is being studied through

inelastic dynamic analyses. In the meantime,

three one-third scale T-shaped shear wall

specimens representing the lowest story of the

12 story theme structure have been tested

under different ratios of shear force to

moment and varying axial force. Also, testing

work on a one-third scale 2-D model of the 1

2

story coupled shear wall is almost complete.

Large rotations of approximately 6% in the

coupling beams were measured at wall drifts of

1.5%.

7.2 Research progress in U.S.

In one research project seismic behavior of

steel/composite coupling beams and their

connections to the RC shear wall has been

studied experimentally and analytically. Four

one-third scale specimens were tested which

represented a portion of the wall and one half

length of the coupling beam from a 20 story

theme structure. The effect of concrete

encasement of the steel beam and details of the

connection in the embedded portion have

been studied.

8. RESEARCH ON NEW MATERIALS,

ELEMENTS AND SYSTEMS (RFI)

The research work in this group aims at developing and application of new and

advanced materials in innovative composite

structural elements and systems (Research For

Innovation - RFI). Thus, the studies in this

group are more of feasibility type than those

in the previous three groups where the

objectives are to develop detailed guidelines

for practical design work.

8.1 Research Progress in Japan

Two research groups on "Fiber Reinforced

Plastic (FRP) elements" and "High

Performance Concrete (HPC)" have been

organized. The team includes research groups

from a large number of industry partners.

Feasibility studies are in progress on FRP-RC

panels; application of FRP for repair and

strengthening of RC members; FRP in

electrical facilities; light weight concrete with

high strength, ductility and stiffness; and

intelligent materials.

8.2 Research Progress in U.S.

Significant amount of research work has been

devoted to advanced cementitious and plastic

composite materials. Albeit, the available

information needs to be expanded to

application in seismic load conditions. In one

project under this program work is in progress

on developing an innovative structural system

consisting of steel open web steel joists

encased in fiber reinforced concrete

eliminating need for conventional labor

intensive reinforcing bars or shear connectors.

In another project application of fiber

reinforced concrete in the joint regions of RCS

systems is being studied.

9. CONCLUDING REMARKS

A brief overview of the progress of U.S.-Japan

Cooperative Earthquake Research Program on

Composite and Hybrid Structures has been

presented in this paper. This five-year

program started in 1993 in Japan and in 1995

in the U.S. The two year time lag is being

taken as an opportunity to plan the future

research work in a more careful and useful

manner in order to derive the most benefit

from the work that has already been

322
completed on both sides. The overall

coordination efforts, cooperation and

exchange of information have been excellent.

More active exchange of research personnel

during the remaining period of the program

will be most beneficial and pursued

accordingly.

REFERENCES

1. "Recommendations for a U.S. -Japan

Cooperative Research Program Utilizing

Large-Scale Testing Facilities," Report No.

UCB/EERC 79-26, September 1979.

2. "Recommendation for U.S. -Japan

Cooperative Research Program - Phase 5

Composite and Hybrid Structures," Report

No. UMCEE92-29, November 1992.

Japan Side U. S. Side

MOC 4 NSF

BCS Univ.

BCJ

KC

Cooperative

Work

JSCA

BRI

JTCC Hybrid

Domestic

Promotion

Committee

Japan JTCC

Hybrid

Secretary

Meeting

Japan Sub-JTCC

1. CFT

2. RCS

3. System

4. RFI

US JTCC

Hybrid

Joint Meeting

US Sub-JTCC

Figure 1. Organization of Cooperative Research

323
Name iNumocr

of

ClArt**

Exterior

Column

Interior Column

/ocsrn

and

MISC.

RQOPEN)-

06

3-6-9 RC RC/S

u or op/or Wall TT cm (S-SRQ+

W-12

12 s SRC/

SRC

Wall

(S-RQ+W1212 s RC/RC Wall

CFT(OPEN)

-24

10-24-

40

CFT CFT/S

CFT+B/HD

-24/

24 CFT CFT/S brace/

hingedevice

Table 1 . Six Types of Theme Structures
RC-Column and S-Beam

1 1

^-S-Beam 1

I *

| 6.400 1 6.400| 6.400| 6.400

1 6.40o| 6.40p|

Figure 3. RC(OPEN)-06

<5\

I 6.400| 6.400| 6.400| 6.400

1 6.40p| 6.400|

Figure 4. RC+W-06

6.400| 6.400| 6.400| 6.40o| 6.40o| 6.400

Figure 5. (S-SRQ+W-12

S-Column pnd S-Qeam

RC-Cblumn

| 6.400| 6.40cj 6.400[ 6.4001 6.40c| 6.40o|

Figure 6. (S-RC)+W-12

m#r~n CFT-Column and S-Beam

IT*

CFT-Column N

S-Beam

Seam

6.400| 6.400| 6.400 1 6.400

1 6.40p| 6.400

1

Figure 7. CFT(OPEN)-24

Ti#n CFT-Column and S-Beam

CFT-Column

Brace or Hinge Device

S-Beam

1 u I

t-"--t

6.400| 6.40o| 6.400 1 6.400 1 6.400

1 6.40o|

Figure 8. CFT+B/HD-24

325

Design of Large-scale Liquefaction Experiment System

by

K. Ishihara *

T. Kagawa **

N. Ogawa ***

C. Minowa ***

K. Sakai  

A. Abe ** ***

ABSTRACT

A large-scale liquefaction experiment system is developed to investigate the liquefaction

phenomenon and the behaviour of structures during

liquefaction. The system consists of a shear box

installed on the multi-purpose shaking table, grit

chamber, elevated water tank, temporary storage,

regulating yard and storage for the soil. The overall

dimension of the shear box is 12m in length (vibrating

direction), 3.5m in width, and 6m in height. The ratio

of the total weight of the shear box to that of the soil

(about 400t) is 9-12%. The allowable shear strain is ± 50%; however, the limit of the deformation is 1.5m

Tests can be performed repeatedly by boiling of the

soil. At the end of a test series, the soil will be

evacuated rapidly (with an evacuation rate of about

1 t/min) as slurry.

KEYWORDS : liquefaction; shear box

1. INTRODUCTION

The present paper describes the design of large-scale

liquefaction experiment system. As shown in Figure

1, the significant feature of the system is the largeness

II as

L=11.6m (W=3.1m)

large-scale shear box

(vibrating direction)

maximum deformation

maximum de:

—existing shear box

region

Figure 1 Comparison of shear box size

of the size and the maximum deformation compared

with existing shear boxes.

2. SYSTEM DESCRIPTION

The system, as shown in Figure 2, consists of shear

box, shaking table, grit chamber, temporary soil

storage, elevated water tank, regulating yard and

storage for the soil. In this order, the soil circulates

in the system during the vibration test. The shear box is connected to the grit chamber

through a hosepipe while the grit chamber is

connected to the temporary soil storage through a

soil conveyor. Among the shear box, the grit chamber

temporary

soil storage

grit chamt er f

water

tank

shear box

sh iking table 12Xl2n

regulatipg

yard

soil

storage

Figure 2 Layout of the system

* Science University of Tokyo, 2641, Yamazaki,

Noda, Japan

** Wayne University, Detroit, MI 48202, USA

*** National Research'Institute for Earth Science

Disaster Prevention Science and Technology

Agency, 3-1, Tennodai, Tukuba, Japan

  Kiso-Jiban Consultants Co., Ltd., 1-11-5,

Kudan-kita, Chiyoda-ku, Tokyo, Japan

** ***Tokyo Soil Research Co., Ltd., 2-1-12,

Umezono, Tukuba, Japan

327
and the water tank, water supply and drainage

system is incorporated. Between the soil storage and

the shear box, a system is being fabricated which

transports and sprays dry sand into the shear box

by means of compressed air.

In order to operate the system efficiently, the

following items are studied :

- soil preparation in the shear box

- regulation of the shear box and soil for the

subsequent test without soil evacuation

- soil evacuation after the test

In addition, as the system utilizes a multi-purpose

shaking table, the shear box is designed so that it can be assembled and disassembled for every test

series.

3. STRUCTURAL DESIGN OF FRAME

support

system

Figure 3 Shear box and lateral support system

The external dimension of the shear box is chosen to be 12m in length (vibration direction), 3.5m in width,

and 6m in height considering the constraints of the

size of shaking table of 12m x 12m, the rated load

(about 500t), and the height of the overhead crane of

8.5m. The most important design conditions of the

shear frames are as follows:

Concerning the height of the frame, although a thinner plate is desirable for smooth shear

deformation, steel H-beam (H-200x200) of 20cm in

height and sliding layer of 1cm in height are adopted

considering various constraints such as weight, etc. The allowable shear displacement between frames

is ± 50% of the height. Considering the stability of

shear box, the displacement is limited at 10.5cm by

stoppers (slots and shear bolts) installed at the four

corners of frames while the overall maximum shear

displacement is limited at 1.5m by the bumpers of

lateral support system . The rigidity of the frame is an important factor for a large-scale system in order

to keep the integrity of the frame and the smooth

horizontal movement against the variation of

pressures in the shear box due to liquefaction. As a

result of the examination, a lateral support system

is adopted to limit the lateral displacements, as shown in Figure 3. The weight ratio of frame to soil is in the range of 9

to 12%1)

/ which is the ratio of an existing shear box

sliding layer

(resin)

inside of the

shear box

steel H-beam

/ f 200 N- / roller \+

steel H-beam

/

side view front view

Figure 4 Roller between frames

of 1.2m x 0.8m x 1.0m. The shear resistance (friction)

of each frame is ideally the same as the soil (it would

be quite complex to take into account the soil

irregularity and the flowability due to liquefaction).

In order to approximate the resistance to that of the

soil, rollers are adopted as shown in Figure 4. The

coefficient of friction between frames may be several

percent.

Furthermore, the design is performed to prevent

swaying and rocking of frame and self-oscillation of

structural materials.

328
4. STRUCTURAL DESIGN OF

SHEAR BOX

For static study, the following items are examined

considering loads such as weight of structure and

soil, static earth pressure, dynamic earth pressure,

water pressure, earth pressure under liquefaction

(18m) and inertial force of 0.5G :

- load distribution in the shear box

- deformation and stress of lateral support system

(considering spring constants and impact force of

stoppers)

- deformation and stress of shear frame (at initial state and at 1.5m deformation state, with and without

spring of lateral support system)

- stability of sliding and tilting of the shear box

The maximum deformations of frame in the state of

soil setup, vibration and 1.5m deformation of shear

box after liquefaction are within 3mm (the allowable

deformation is 5mm).

For dynamic study, the natural frequencies and eigen

modes are obtained for structural materials, frames,

and lateral support system with and without soil. The natural frequencies are above 20Hz, confirming

the safety of the system.

5. WATER SUPPLY AND DRAINAGE

Water supply rate of 2 t/min is enough for the

boilling of soil (injection from the bottom, the top

and the sides of the box). At the end of a test series,

it is necessary to evacuate the soil and to disassemble

the shear box immediately. For this purpose, a slurry

evacuation method is adopted. The method, which

consists of injecting water into the shear box to make

the slurry and again in the downstream of the

evacuation valve ( <j> 200) to produce negative

pressure, would enable the evacuation of the soil

with an evacuation rate of about It/min.

6. CONCLUSION

The design of large-scale shear box (total weight

about 500t) and preliminary tests (vibration test with

a height of 4m of shear box, and test of friction

between frames with jack) are performed.

The main results are as follows :

- the ratio of the weight of the shear box to that of

soil is in the range of 9 to 12%

- the frame deformation at the maximum shear box

deformation of 1.5m is within 5mm

- the feasibility of soil reuse by boiling as a results of

preliminary tests is verified

- the efficiency of soil evacuation method is

confirmed

Basic performance tests of the system and then

liquefaction tests will be carried out to extract and

resolve the problems that may arise.

REFERENCE

1) Kokusho, T., Dynamic Soil Properties and

Nonlinear Seismic Response of Ground, EPRI-Report

No. 301 (1982)

329

WIND ENGINEERING
I I!

', IjjIIIi a

']I ; IHIi|

i'' !
Influence of Vent Configurations on Bridge Response to Wind

by

Harold R. Bosch*

ABSTRACT

Deck vents and open-grate sidewalks have often

been used to improve or enhance the

aerodynamic performance of long-span bridge

structures. The new Tacoma Narrows Bridge is a good example of such an application. This

paper presents the results of a wind tunnel study

to investigate the significance of existing

sidewalk grates in the overall stability of an Hshaped deck section for a suspension bridge.

The results of tests to evaluate additional deck

venting will also be presented. The tests indicate that the sidewalk grates do, in fact, play an important role in the behavior of this bridge.

The status of the windward vent appears to be

most significant. Deck venting, on the other

hand, does not appear to uniformly enhance

stability.

KEYWORDS: aerodynamics; bridges; longspan; suspension; vents; wind engineering.

1. INTRODUCTION

Flexible or slender structures can be sensitive to

wind forces and exhibit undesirable windinduced oscillations. The engineer can often

adjust the shape or modify edge details to

enhance the structure's aerodynamic

performance. For long-span bridges, one

method of improving stability has been to install

vents in the deck, shoulder, or sidewalk areas. The new Tacoma Narrows Bridge, for example,

has 5 sets of vents in the deck which run the

length of the bridge. The Deer Isle-Sedgwick

Bridge, which will be the focus of this paper,

has sidewalk vents that extend along each edge

of the roadway. Since its construction in 1939,

the Deer Isle bridge has been quite sensitive to

the wind. There have been observations that

bridge oscillations are more pronounced when

the sidewalk grates are covered with snow. To

confirm these observations and to establish the

significance of the vents for this structure's

stability, a series of tests have been conducted in

the wind tunnel. Since the bridge deck was

scheduled for replacement, tests have also been

conducted to explore the utility of installing

additional vents in the traffic lanes. Both

dynamic response tests and static force tests have

been performed in smooth oncoming flow. This

paper will describe the experiments and test

facilities, present results of the response and

force measurements, and discuss these results.

2. BRIDGE DESCRIPTION

The Deer Isle-Sedgwick Bridge is a girderstiffened suspension bridge similar in cross

section to the original Tacoma Narrows Bridge

(Figure 1). The structure consists of a 1080 ft (329m) main span, with side spans of 484 ft (148m) and approach spans of 100 ft (40m), for a total bridge length of 2308 ft (704m). The

two-lane deck is 20 ft (6m) wide and 4.5 in

(11cm) thick. Stiffening girders are 6.5 ft (2m)

deep and spaced 23.5 ft (7m) apart.

3. AERODYNAMICS LABORATORY

The research was conducted in the

Aerodynamics Laboratory at TFHRC.

Established in the late 1950's the Laboratory has

been used primarily to ensure the aerodynamic

stability of long span bridges. The facility

includes a large lab area containing a wind

tunnel, several unique test fixtures,

instrumentation systems, and lab test equipment

as well as an adjacent office area housing staff,

computer systems, reference materials, and

* Research Structural Engineer, FHWA TurnerFairbank Highway Research Center, 6300

Georgetown Pike, Mclean, VA 22101

333
records. The G.S. Vincent wind tunnel is a low

velocity, open-circuit, laminar flow design. Air

enters a double inlet section, passes through a

diffusing section containing a series of 8 screens,

and exits a 6 by 6-ft (1.8 by 1.8-m) nozzle into

the test area. The air flow is quite laminar and

well controlled at speeds up to 50 fps (15m/s).

For experiments where turbulent flow is desired, an active turbulence generator is inserted into

the circuit between the wind tunnel nozzle and

the test area. With this system, large scale

turbulence can be simulated in the vertical and

along-wind directions, independently. The

vertical angle of the wind can be changed by

rotating the wind tunnel nozzle up and down or by rotating the model around its longitudinal

axis as needed.

4. SECTION MODEL

A section model is a scaled, rigid, and

geometrically faithful copy of a typical length of

the prototype structure. One basic section

model, built to a scale of 1/25, was used for the

tests reported here (Figure 2). The alternate

configurations studied were achieved through

simple modification of the base model. All

flanges, stiffeners, floor beams, and bracing

were carefully designed and fabricated to

replicate prototype details. Special attention was

given to the sidewalk grates to ensure a proper

solidity ratio. The length of the model was set

at 5 ft (1.5m). The model width (B) was 11.25

in (28.6cm), web to web, and the depth was

3.125 in (7.9cm).

5. RESPONSE MEASUREMENTS

5.1 Procedure

The rigid section model was suspended in the

wind tunnel test section by means of four coil

springs. The size and stiffness of the springs

was adjusted to provide the desired vertical and

torsional vibration frequencies. For the basic

test configuration, these frequencies were set to

2.05 and 5.73 Hz, respectively. Large elliptical

end plates were attached to the the model to ensure two dimensional flow near the ends.

Horizontal restraining wires were attached to the

ends of the model to prohibit longitudinal and

lateral movement of the model. No attempt was

made to block either the vertical or torsional

degree-of-freedom during testing. Mechanical

damping for the test setup ranged from about 0.

1

to 1.5 percent of critical. The results from tests conducted at low damping levels only will be

reported here. The section model was subjected

to smooth oncoming flow at a fixed angle of

attack of 0 degrees. The measurements were

taken at a series of steady wind velocities

ranging from 0 to 37 fps (0-11.3m/s). For some

tests, however, the velocity range was limited

due to violent, and potentially damaging,

response of the model in the wind flow. At

each new velocity, the wind flow and model

activity were allowed to stabilize prior to data

collection. Model response in its two degreesof-freedom was sampled at a rate of 50 samples

per second for a period of 20 to 30 seconds.

Dynamic data was filtered to remove any

undesirable electronic noise prior to recording.

For each test event, a 10-second data window

was selected for analysis. The two channels of

motion data were demeaned and detrended to

leave only the dynamic portion of each signal.

Next, the root-mean-square (RMS) values were

calculated from the "conditioned" time histories.

Finally, the vertical and torsional calibration

factors were applied.

5.2 Results

Dynamic response measurements were conducted

on the unmodified deck section, three alternate

shapes, and a proposed deck modification under

a variety of conditions. Only results for the

unmodified section, with sidewalk grates open or

closed and with deck vents present or not

present, will be discussed here. For the

unmodified section with the sidewalk vents open

(the normal condition), the model exhibits a

large RMS rotation response of 1.42 degrees

which initiates at a reduced velocity (V/NB) of

about 2.10 with the peak centered around 2.75

(Figure 3). In the vertical degree-of-freedom,

the model displays two distinct regions of

activity centered at a V/NB of 1.75 and 5.75,

with the beginnings of a third evident at 8.75

(Figure 4). Note that this test had to be

334
terminated at 18 fps (5.5m/s) due to rapidly

growing vertical motion.

Since the sidewalk grates are sometimes covered

with snow during the winter months, tests were

conducted with the grates sealed off. For this

condition, the model exhibits a small region of

RMS rotational response of 0.6 degrees centered

at a V/NB of 1.20 and a larger one initiating at 2.00 and climbing to an amplitude of about 2.50

degrees (Figure 3). In the vertical vibration

mode, the model displays three zones of activity

centered at 1.65, 3.10, and 6.25 with

nondimensional RMS levels of .032, .034, and

.022, respectively (Figure 4). At the conclusion

of this test vertical activity was decreasing while

torsional activity was on the increase.

Since the existing bridge deck was scheduled for

replacement, venting of the new deck was

considered as a means for enhancing the

aerodynamic behavior of this bridge. Tests were

conducted with the deck vented in the center of

each traffic lane. Although venting of the deck

does reduce the level of vertical motion, it does

not eliminate the motion entirely (Figure 5). On

the other hand, this venting tends to augment the

torsional motion resulting in vibration levels

more than double those observed previously

(Figure 6). Divergent vertical response was not

noticeably affected by the deck venting and

reappeared at about 17 fps (5.2m/s).

6. FORCE MEASUREMENTS

6.1 Procedure

To measure steady-state wind forces, the section

model was carefully mounted in the forcebalance. This force-balance supports the model

in a horizontal orientation and measures the 3

force components of lift (L), drag (D), and

pitching moment (M). The balance is capable of

rotating the section model about its longitudinal

axis to vary the wind angle of attack. For the

tests reported here, the attack angle ranges from

-15 to + 15 degrees. Large circular end plates are installed on the balance to ensure two

dimensional flow near the model ends. Each

model configuration was tested in smooth

oncoming flow at a minimum of 9 distinct wind

speeds and all tests were performed twice to ensure repeatability of the results. Wind speeds

were carefully chosen to provide the maximum

range of Reynolds number possible and to avoid

regions where undesirable vibrations would

develop in the model. In general, wind speeds

ranged from 5 to 16 fps (1.5-4.9m/s) and the

same speed settings were not necessarily selected

for each configuration. During each test,

readings from the balance and other laboratory

instrumentation were continuously sampled at a

rate of 100 Hz for a period of at least 83

seconds. All of the data from the force-balance

was low pass filtered at 30 Hz prior to

recording. For each test, an 83-second data

window was selected for analysis. Calibration

factors were applied to each channel of force

data, mean forces per unit span were calculated,

and these forces converted to nondimensional

force coefficients.

6.2 Results

The drag and lift forces, which are in the wind

coordinate system, are often translated to chord

(X) and normal (Z) forces, in the bridge

coordinate system, using the following:

X = D cos a - L sin a

Z = L cos a + D sin a

where a is the wind angle or angle of attack.

The orientation of these forces is illustrated in

Figure 7. All forces are then reduced to

dimensionless coefficients defined as:

Cd = 2 D / (p U

2 B)

C, = 2 L / (p U

2 B)

Cm = 2 M / (p U

2 B2

)

Cx = 2 X / (p U

2 B)

C, = 2 Z / (p U

2 B)

where p is air density; U is oncoming wind

velocity; and B is deck width.

Steady-state force measurements were conducted

on the unmodified deck section and four

alternate shapes under a variety of conditions.

Only the results for the shape designated

335
Alternate A will be presented here (Figure 8).

This example was chosen since the bridge was

recently changed to this shape. To evaluate the

significance of the sidewalk grates for this

shape, the following test cases were studied:

Case 1-both grates open, Case 2-both grates

closed, Case 3-windward grate closed, and Case

4-leeward grate closed. As noted earlier, the

grates can become blocked when snow is present

on the bridge.

For Alternate A with both sidewalk grates open,

the normal coefficient exhibits a gentle positive

slope with a value of about .34 for a horizontal

wind (Figure 9). When both grates are closed,

the slope becomes steep over most of the test range, indicating greater sensitivity to wind

angle change, and C2 for a horizontal wind is reduced to .06. When only the leeward grate is closed, the change is not quite so dramatic; the

slope increases a small amount and Cz at zero degrees drops to about .23. For the case where

the windward grate is closed, however, the

normal coefficient changes in a different

manner. Here, the slope of the Cz curve turns

significantly negative for small wind angles.

This is not a desirable characteristic and can lead

to vertical stability problems.

For Alternate A with both grates open, the

moment coefficient is positive and exhibits a

positive slope except for wind angles above +6

degrees (Figure 10). This is also true for the

case where the leeward grate is closed. When

the windward grate is closed or both grates are

closed, however, Cm takes on negative values for

wind angles below about -11 degrees. When the

windward grate only is closed, the moment

coefficient exhibits a slight negative slope for

small positive wind angles. This characteristic

could indicate torsional stability problems.

In general, the chord coefficient takes on similar

values, at modest wind angles, for the four cases investigated (Figure 11). For a wind angle

equal to zero degrees, Cx tends to decrease

slightly when both sidewalk grates are closed.

At large wind angles, however, the decrease in

chord coefficient is more apparent with closure

of both grates resulting in the most change.

7. CONCLUSIONS

For the unmodified deck section, the results

indicate that the sidewalk grates perform a

significant role in the dynamic behavior of this

structure — they weaken the potential vortexinduced response of the bridge. Blockage of

these vents results in more frequent response,

greater bridge motion amplitudes, and increased

activity at lower wind speeds. Although the

addition of vents in the roadway does reduce

vortex-induced motion, it does not eliminate the

motion entirely. With the deck venting,

torsional motion amplitudes were more than

doubled and divergent vertical response was not

delayed. For the fairing-modified deck section,

Alternate A, some changes were observed in the

wind forces depending on the status of the

sidewalk grates. In general, it appears that the

windward vent is more important than the

leeward one. When this vent is blocked, the

normal and moment coefficients exhibit negative

slopes for small angles of attack. These results

tend to confirm observations made at the bridge

site.

336
Little Ditr ItU

8«4gwiefc

130 ft (39.6m)

THT

(Elevation)

484 ft (147.5m)

1080 ft (329.2m)

484 ft (14?. 5m)

130 ft (39.6m)

(Plan)

SUSPENDERS

•"(15cm)

BRACKET FOR LATERAL STAY AT SELECTED

PANEL POINTS

10 ft (3n) 10 ft (3m)

OPEN GRATE SCEVKALK 1 ^4 1/2* CONC SLAB

24VtFF.B.

23'-«* CtoCGIROERS

Figure 1. Deer Isle - Sedgwick Bridge.

24 ft (7.3m)

11.812*

i

60.00"

a 126 ki 7.8 cm

.5626 In 1.4 cm

Figure 2. Bridge Section Model.

337

3.125" -41
3. 00 ,

2. 50 .

.

10

111

III £ 2. 00

111

I „ 1.50

0)zKi_i

111 J 1.00

Z

<

0. 50 .

.

0. 00

O CURB OPEN

X CURB CLOSED

X

X X X

X

X

X

X

X

x o

o

o 0 00

o

X 0

X

x o

ro i oflililitrvTT¥rnTT riAr""rtTinninnnraf1tI1—> <

Ifl B

WIND TUNNEL VELOCITY - FPS

in S

Figure 3. Torsional Response - Sidewalk Vents.

0. 035 ,

0. 030 .

.

0 CURB OPEN

X CURB CLOSED

0. 025 .

.

ID^ 0. 020 .

.

n

COX

K

i 0. 015 :

:

0. 010 ..

x * x x

x

* o

_0

oo

cX° O

o rfbo

X X

0. 005 .

.

0. 000 OP i o

o x °o

x O o ° o

0

fjo

0^»q %°,o*? * J ,*

00

O Oqo

in s in WIND TUNNEL VELOCITY - FPS

(VI

Figure 4. Vertical Response - Sidewalk Vents.

338
0. 035 _

0. 030 .

.

0 DECK SOLID

X. DECK. VENTED

0. 025 .

.

CDx 0. 020 .

.

n

01zK

z 0. 015

0. 010

o

X

0o

0

tPoo

0. 005 .

.

0. 000

x°o 0 0

CO <08

0 Oqq

in s io WIND TUNNEL VELOCITY - FPS

s

Figure 5. Vertical Response - Roadway Vents.

3. 00

2. 50

COU

ill K 2. 00 U

111 Q

O DECK SOLID

X DECK VENTED

n 1- 50

10zK

XO

o

O 00

1. 00 ..

0. 50 .

.

0

0. 00

10 WIND TUNNEL VELOCITY - FPS

Figure 6. Torsional Response - Roadway Vents.

339
Figure 7. Orientation of Wind Forces.

1

0.75

0.5

0.25

O 0

-0.25

-0.5

-0.75

Figure 8. Configuration Alternate A.

AERODYNAMIC FORCE COEFFICIENTS

1

-15

s •

• CASE 1

0

m 0

0

0 r _

0 «*•'• St..

\ CASE 2

" •'

- CASE 3

-J—

" x CASE 4

s

s I.I, I.I.

-10 -5 0 5 10

Alpha ( Deg

)

Figure 9. Normal Force Coefficients.

340

15
AERODYNAMIC FORCE COEFFICIENTS

15 -5 0 5

Alpha ( Deg

)

Figure 10. Pitching Moment Coefficients.

Figure 11. Chord Force Coefficients.

341

Aerodynamic Characteristics of Slotted Box Girders

by

Hiroshi SATO1

* and Katsuya OGIHARA2)

ABSTRACT

Aerodynamic stability is one of the most

important themes in the design of a super longspan bridge. To improve the aerodynamic

stability of a super long-span bridge, a series of

wind tunnel studies and analytical studies were

conducted for slotted box girders. According to

the test results and analytical results, it was found

that slot at the center of girder was effective to improve the aerodynamic stability. It was also

found from section model test that devices like

center barrier and guide vanes increased flutter speed considerably.

KEYWORD: flutter, super long-span bridge,

slotted box girder, unsteady aerodynamic force

1. INTRODUCTION

The Akashi Kaikyo Bridge, which is now under

construction in Japan, will have the main span

length of 1990m. In Japan and in the world, there

are several plans or ideas to construct longer

bridges such as the Messina Straits Bridge. In the

design of such super long-span bridges,

aerodynamic stability is one of the most important

themes. The approaches to improve aerodynamic

stability can be classified into structural one and

aerodynamic one. In this paper described is the

latter approach, namely to improve aerodynamic

characteristics of stiffening girders using slot.

2. EXPERIMENT ON THE EFFECT OF

LOCATION AND SIZE OF SLOT

The effect of location and size of slot on

aerodynamic characteristics was examined

through section model wind tunnel test. Considering a super long-span bridge which has

center span length of 3,000m with two side spans

of 1,500m, the structural conditions were assumed

[1,2,3,4] as shown in Table-1. As for the cross

section of the box girder, two types were

considered. Model-A was 40m wide, 4m deep,

and Model-B was 45m wide, 3m deep. The

cross section of these two types are shown in Fig.-

1. In the models, gird rails and any other small

attachments which are usually installed on bridges were neglected to evaluate the basic

characteristics of slotted box girder. Slots were

represented by removing some deck plates of the

model. The tests were conducted in smooth

flow. The test cases are listed in the Table-2 with

corresponding cross sections and test results.

(1)Model-A

From the test results of the Model-A, it was found

that the slot at the center increased the flutter

onset wind speed by 12 m/s(case-l, case-5). On

the other hand, it was found that the slots at the

edge of the girder lowered the flutter onset wind

speed(case-l, case-8). In case- 10, where slots were installed at the center and edge of the girder,

flutter onset wind speed was very low. This

shows that the location of slot should be carefully

chosen. As for the cases without fairing, the slot

at the center of the girder decreased flutter onset

wind speed from 27m/s(case-14) to 18m/s(case11). This implies that suitable fairing as well as center slot is required to increase flutter speed.

(2)Model-B

From the test results of the Model-B, it was found

that the flutter onset wind speed was increased

with the width of slot at the center of the

girder(case-l 6, 1 8,20).

3. ANALYTICAL STUDY ON THE EFFECT

OF SLOT

In order to explain the effect of slot at the center

of girder, preliminary analysis was conducted.

For the analysis, aerodynamic forces acting on the

each box of the girder were calculated using the

Theodorsen's function. The aerodynamic

interference between the 2 boxes was neglected.

Using the above aerodynamic forces, two degreeof-freedom flutter analysis was conducted by U-g

1) Head, Structure Division, Structure and Bridge

Dept., Public Works Research Institute,

Ministry of Construction

2) Senior Research Engineer, Structure Division,

ditto

343
method [5,6]. The result of the flutter analysis

shown in Fig.-2 indicates that the flutter onset

wind speed increases with size of slot. The

differences between the analysis and the

experiment seems to be caused by aerodynamic

interference between the 2 boxes.

4. IMPROVEMENT OF AERODYNAMIC

CHARACTERISTICS OF GIRDER

Although wide slot at the center of the girder

improves flutter characteristics, narrower slot

would be preferable from the viewpoint of

construction cost of towers and foundations. To

improve aerodynamic characteristics, the effect of

some devices was studied by section model tests. The model used in this study was the Model-B

with gird rails.

The tested devices are illustrated in Fig.-3. The

result is shown in Fig. -4, where the flutter speed

measured from section model tests was

normalized by the flutter speed predicted by the

Selberg's formula. It can be seen that a center

barrier and guide vanes improved flutter

characteristics very well.

However, the flutter speed was not so high when

angle of attack was -3 deg. as is shown in Fig.-4.

It was found that the gird rails at the bottom deck

increased the flutter speed considerably at this

angle of attack (Fig.-5).

These devices were applied to 3 slotted box

girders with different girder depth(D) as is shown

in Fig.-6.

The flutter speed of these girders are shown in

Fig.-7. The aerodynamic stability was excellent

for the girders with D=3m, and D=7m.

However, the aerodynamic stability of the girder

with D=5m was lower than the other girders.

When the slot ratio b/B* of the girder with D=3m

was decreased from 25% to 20%, the flutter onset

wind speed dropped considerably. It seems that

the slot wider than 25% of the girder width is required to maintain excellent aerodynamic

stability in this case.

5. COEFFICIENTS OF THE UNSTEADY

AERODYNAMIC FORCES

In order to investigate the aerodynamic

characteristics of the slotted box girder more

precisely, unsteady aerodynamic forces were

measured for the girder with D=3m. The

measurement was made by forced oscillation

method [7]. Coefficients of unsteady

aerodynamic forces were defined as follows.

L = 7t p {b2[LZRco 2Z+LZI coZ']

+ b

3 [L

0

R

co 2 0+L0JO) 0']

+ b

2[LYRa> 2Y+LYi<oY']}

M= 7T p {b3 [MZRco 2Z+ MZI coZ']

+ b4[M0 R

a) 2 0+M0TCO 0']

+ b

3 [MYRC*> 2Y+MYiWY']}

D =- 7T p {bd[DZR co 2Z+DZj co Z]

+ db2[D 0 R

co 2 0 +D 0jto 0

']

+ db[DYR<*> 2Y+DYi<*>Y']}

where

p : Air density

b,d : Girder width and depth

CO : Angular frequency

Z, 0 ,Y: Vertical, torsional and horizontal

(along wind) displacement

Z', 0 ',Y': Vertical, torsional and horizontal

(along wind) velocity

L,M,D: Lift, Moment and Drag force

L**, M**, D**: Coefficients of unsteady

aerodynamic forces (For example, MZj is

the coefficient of imaginary part of

unsteady aerodynamic moment due to

vertical motion.)

Some of the calculated coefficients are shown in

Fig.-8. They are compared with those of single

flat plate, and with those of flat plate with slot at

the center. The unsteady aerodynamic forces for

these plates were derived from the Theodorsen's

function. In the calculation for the slotted flat

plate, aerodynamic interference between the 2

plates was neglected. As a whole, the unsteady

aerodynamic forces of the slotted girder seems to be similar to those of the slotted plate rather than

those of the single plate.

6. DISCUSSION

Among these 18 coefficients, the most important

coefficients are MZj

, M q j, L q r and M q r.

344
For 2-degrees of freedom system, the relationship

between unsteady aerodynamic forces and some

vibration properties can be expressed by the

following formulas [8].

(5a^-7t 2MZIX/u-7C 2M 0I/w

X=2o/ 6 o/B= n L e R

/(-l+(fz/f g f a 2)/n

a

2 =(fg/{)2 =l+KMg R/V

Here, 6* a is aerodynamic damping in

logarithmic decrement, and X is amplitude ratio,

and o is frequency ratio. They were derived by

assuming that absolute value of aerodynamic

damping and phase angle are small, and that

absolute value of the amplitude ratio X is small.

As is shown here, M q r

affects the frequency

ratio a . L q r and o affect the amplitude ratio

X. M^i , M q J and X affect the aerodynamic

damping.

In Fig.-8, the measured M q r of the slotted box

girders is compared with that of single flat plate,

and that of slotted flat plate. From the figure, it can be said that M g r of the slotted plate is

smaller than the single plate, and that M q r of

the slotted box girder is smaller than the slotted

plate.

As for L q r, the three coefficients are almost

identical.

M^i will cause negative damping. From the

figure, it can be said that the absolute value of

Mzi for the slotted plate is smaller than the single

plate. The absolute value of M^j for the slotted

box girder is smaller than the slotted plate.

M q j will cause positive damping. M q j for

the slotted plate is almost identical with that for

single plate. The absolute value of M q j for the

slotted box girder is smaller than the others.

The aerodynamic damping for the single plate, the

slotted plate and for the slotted box girder was

calculated from the above formulas, and is shown

in Fig.-9. The aerodynamic damping turns

negative at small reduced frequency. The critical

reduced frequency for the slotted plate and slotted

box girder is smaller that that for the single plate.

It means that the flutter speed for these slotted

plates and slotted box girders are higher than that

for single plate.

The characteristics of unsteady aerodynamic

forces of the slotted box girder are summarized

here. Since M q r is small, the frequency ratio

o becomes small. Since (fz/f g ) o is less than 1

and small, and since L q r is positive, the

amplitude ratio X becomes negative and its absolute value becomes small. Since Mzj is

negative and its absolute value is small, the

critical wind speed for flutter becomes high.

7. FLUTTER ANALYSIS FOR A SUPER

LONG-SPAN BRIDGE

Using all the measured coefficients of unsteady

aerodynamic forces, flutter analysis of a super

long-span bridge was conducted (Table-3). The

main span length of the super long-span bridge

was 2,500 m, and the side span was 1,250 m. For

comparison, flutter analysis was also conducted

using unsteady aerodynamic forces of the single

plate derived from Theodorsen's function. The

structural conditions for both cases were same.

Flutter speed for the slotted box girder was as high as about 80 m/s as is shown in Fig.-lO. On

the other hand, the critical wind speed for single

flat plate was as low as about 40 m/s.

8. CONCLUSION

It was found that box girder with slot at the center

had good flutter characteristics, and it could be

improved by some devices like center barrier and

guide vane. It seems that the slotted box girder

would be one of the possible stiffening girders for

super long-span bridges.

REFERENCES

[1] Nakamura, S. et. al.. "A proposal on super

long-span bridge with stiffing girder", Bridge

and Foundation, 92-12, 1992, pp37-42 (in

Japanese)

[2] Kawada, T., "Recent Suspension Bridge",

Rikou Tosho, Tokyo, 1987 (in Japanese).

[3] Kimura M., "On cable system for long span

suspension bridge with center span of

3,000m", Proc. of the 47th Annual

Conference of the Japan Society of Civil

345
Engineering, 1992, ppl018-1019 (in

Japanese).

[4] Suzuki et al, "Study on Structure of Super

long-span bridge with aerodynamic stability",

Proc. of the 48th Annual Conference of the

Japan Society of Civil Engineering, 1993,

pp838-839 (in Japanese).

[5] Bisplinghoff R. L., Ashley H., and Halfman R.

L. "Aeroelasticity" Addison Wesley.

[6] Washizu K, "Aeroelasticity", Kyouritsu

Shuppan, Tokyo, 1957 (in Japanese).

[7] Okubo, T., Narita N. and Yokoyama K.,

"Some approaches for improving wind

stability of cable-stayed girder bridges", Proc.

of the 4th International Conference on Wind

Effects on Buildings and Structures, 1975,

pp241-249.

[8] Nakamura, Y. and Watanabe, K., "Flutter

analysis of 2-degrees of freedom for

suspension bridges", Proc. of the 32nd

Annual Conference of the Japan Society of

Civil Engineering, 1977, pp231-232 (in

Japanese).

346
Tab I e-1 Assumed Structural Condition

Mode 1 -A Mode 1 -B

Weight tf/dl 46 46

Pol or Moment tf •m

, s2/ni 1,280 1,620

of Inertia

Natural Vertical Hz 0.050 0.050

Frequency Tortional Hz 0.105 0.105

Girder Width

^

m 40 45

Girder Depth m 4 3

347
Table-2 Test Case and Result

cass Mode 1 Type

Slot Pattern

Ang 1 e of Attack

Slot

Ratio'

Fairirv

F 1 u 1 1 6 r On 8 6 t

Wind Speed**

1 A <i : : : ; ; i [> 0° 0% Yes 3 9m/s

2 A <i i i ; : ; > 0° 0% Yes 3 8m/s

3 A <J i i ! ! : : > 0° 0% Yes 3 7m/s

4 A «_J ! i. ! : j_J> 0° 0% Yes 3 7m/s

5 A <l i : : ! : ! I> 0° 20% Yes 5 1 m/s

6 A <l i : i ! : : l> 0° 20% Yes 3 7m/s

7 A '

' M> 0* 20% Yes 4 7 m/s

8 A <l i i : : ! i i> o- 20% Yes 2 5 m/s

9 A <U L_J LJ U> 0* 20% Yes 2 1 m/s

1 0 A < [_j o g> 0' 40% Yes 1 9 m/s

1 1 A I: : : : : :

| 0' 20% No 1 8 m/s

1 2 A <i i i :

i

; : i> -3° 20% Yes 5 3 m/s

1 3 A <l ! : i ; !

i l> + 3* 20% Yes 5 3 m/s

1 4 A 1

• i :

• ; ;

| 0° 0% No 2 7 m/s

1 5 A Ui i i

: U 0° 20% No 1 6 m/s

1 6 B <l 1 1 1 1 III It 0° 0% Yes 3 9 m/s

1 7 B <l 1 1 1 1 1 1 1 1 0 0° 13% Yes 3 8m/s

1 8 B <l 1 1 1 1 1 II lb 0* 20% Yes 5 4m/s

1 9 B < I> 0° 33% Yes 5 1 m/s

2 0 B <l 1 1 1 1 1 1 l> 0° 47% Yes 8 0 m/s

2 1 B <1 1 II 1 1 1 1 1 l> 0° 26% Yes 6 2 m/s

2 2 B <1 1 1 1 1 1 II lb 0° 40% Yes 5 7 m/s

2 3 B < II i.o 0° 13% Yes 4 6 m/s

2 4 B <] 1 1 1 1 1 II lt> cr 47% Yes 6 4 m/s

2 5 B < t> 0' 73% Yes 7 4m/s

2 6 B <i i i i i r hi i> 0° 33% Yes 6 8 m/s

2 7 B <i > 0° 40% Yes 6 9 m/s

* : Slot Width / Girder Width ** : Shown as wind speed for the real bridge

348
Tab I e-3

Resul t of Flutter Analysis

Span Length 1, 250 + 1, 250 + 1, 250 m

Natural

Frequency

Bending Ist-symm. 0. 055Hz

Torsion Ist-symm. 0. 097Hz

Flutter Onset

Wind Speed

Single Plate 34. 54m/s

Slotted Box 84. 18m/s

349
4 0m

< II 1 1 II

Model-A

4 m

4 4 4 8 8 4 4 4

<

4 5m Model- B

> 3m

2. 6 3 6 3 6 6 3 6 3 2. 6

Lines in box show segment of test model

Figure-1 Cross Section of Girder

35

XJ

CD£30

CO

.E 25

£ 20

+-•

°- 15

•o

CDO

Analysis

Experiment

t

0. 2 0. 4

Slot Ratio (b/B)

0.6

Figure-2 Result of Flutter Analysi

350
Guide Vane

Gird Rail

/ \

Center Barrier

<

Guide Vane Gird Rail

>

b(=0.25B*)

B

5

Figure-3 Tested Devices

351
CB+GV

Figure-4 Effect of Tested Devices

DO

s—CD -Q

CD

to \

o

Z5

1.5

0.5

0

-3

•Without gi rd Rai

I

iWith Gird Rai

I

0 3

Angle of Attack

Figure-5 Effect of Gird Rails

at the Bottom Deck

352
<

B*=45.3m

0.25B*

1

^> [ D=3m

<

0.25B*

1 J ~>T =5m

B*=45.3m

0.25B*

1

D=7m

Figure-6 Cross Section Considered

1.5

60

k_

03 -Q

<D 5 1 \

o

Z>

0.5

o

-3

D=3m, b/B*=20%

•

I D=3m

D=5m

D=7m j

0 3

Angle of Attack

Figure-7 Effect of Gireder Depth

and Slot Width

353
MZR

_ ncomsENSiKLE run)

.. nctootSEiuunio futi)

0.05 0.1 0.15 0.2 0.25 Reduced Frequency

0.00

M0R

- nCMOKENSIKlE PUIE)

... ncooctta(norTa mti>

CASJtED

0.05 0.1 0.15 0.2 Reduced Frequency 0.25

MZI

- ntOOOKEJKSIM

... ncoWtSEK SLOTTED fUTE) QICilSKD

0.05 0.1 0.15 0.2 Reduced Frequency0.25

Figure-8 Cofficients of Unsteady

Aerodynamic Forces

354
LOGARITHMIC DECREMENT <3"

1.0

0.5

0.0

-0.5

-1.0

r

1 1

1

<?t1l ID 1 fNF • THFfinriRSFMSINGLE PLATE) dotted line : theodorsemslotted plate) • : measured

' tSt " -

•

O 0.05 0.10 0.15 0.20 0.25

Reduced Frequency

Figure-9 Calculated Aerodynamic

Dampi ng

V-5

B=45.3m,D=3m

00

T3

03c

>> -ao

! a

0.10

0.05

-0.00

-0.05

-0.10

Single Plate

i # Slotted Box

i •

T

t

T

T

1

0 20 40 60 80 100 120

VelocityGn/s)

Figure-10 Result of Flutter Analysis

355

Wind Engineering Research in the United States

by

Ahsan KAREEM1 and Eleanore SABADELL2

1. INTRODUCTION

This paper summarizes wind engineering

research at various institutions in the U.S.

Most of the information has been derived

from the papers presented at the Sixth and

Seventh U.S. National Conferences held in

1989 and 1993 at Houston and Los Angeles,

respectively, the Ninth International

Conference on Wind Engineering held in

1995 at New Delhi, and Hurricanes of 92 in

Miami. The research activity is discussed

here under the following topics:

1 . Bluff Body Aerodynamics and

Unsteady Load Effects

2. Performance of Building Envelope

3. Damage Control and Mitigation

4. Random Data Analysis, Modeling and

System Identification

5. Tall Buildings and Engineering

Structures

6. Mitigation and Control of Structural

Response

7. Bridge Aerodynamics and

Aeroelasticity

8. Knowledge-Based Expert Systems and

Computational Wind Engineering

9. Hurricane Wind Speeds,

Regionalization of Extreme Winds,

Vortices and Cyclonic Winds

10. Wind Engineering Research

Applications

2. BLUFF BODY AERODYNAMICS

AND UNSTEADY LOAD

EFFECTS

Over the past several decades, much

attention has been focussed on the

aerodynamics of streamline bodies; however,

our knowledge of the aerodynamics of

sharp-edged bluff bodies in turbulent

boundary layers is far from complete. Efforts

in this direction are in progress at several

institutions, e.g., University of Notre Dame,

Colorado State University, Clemson

University, University of Minnesota, Virginia

Polytechnic Institute, Johns Hopkins

University and Florida Atlantic.

The various research topics undertaken in the

area of bluff body aerodynamics involve

extensive measurements of the pressure field

around prisms, cylinders and bridge deck

sections. These studies have examined the

influence of scale and intensity of turbulence

on pressure fluctuations. Measurements and

analyses of the space-time structure of

random pressure fields and area-averaged

loads acting on the surface of prismatic and

cylindrical bodies of different aspect ratios,

exposed to simulated turbulent or turbulent

boundary layer flows, have helped to improve

our understanding of the complex

fluid-structure interactions.

1 Professor, Department of Civil Engineering, Notre

Dame University

2 Director, Natural and Technological Hazards

Mitigation Program, National Science Foundation

357
Several wind tunnel facilities in the U.S. have

been involved in the development of

measurement schemes to monitor unsteady

load effects on such bluff bodies. These

include both measuring fluctuating integral or local loads on buildings and bridge models

and similar measurements on the full-scale

structure. The measurement schemes include

area-averaged pneumatic manifolding

schemes, ultra-sensitive high frequency force

balances, and second generation force

balance to account for multiple level force

measurements. The use of high speed

scanning systems to monitor a large number

of pressure taps are being developed that

have the potential of replacing high

sensitivity force balances. The senior writer

has, in addition to the preceding schemes,

utilized piezoelectric films to monitor area averaged loads, but the complexity of

installation and the film's sensitivity to

temperature, in addition to pressure, has

limited its applications.

3. PERFORMANCE OF THE

BUILDING ENVELOPE

Several institutions in the U.S. are actively

involved in researching the structural

resistance to wind hazards. These institutions

include: Texas Tech University, Texas A&M

University, Clemson University, the

University of Florida, Universities of

Missouri at Columbia and Rolla, University

of Notre Dame, University of Washington,

Colorado State University, Johns Hopkins

University, State University of New York at

Buffalo, University of Hawaii, Wind

Engineering Research Council, Roofing

Industry Research Institute, Lawrence

Livermore Laboratory, Glass Industry,

Underwriters Laboratory, and Model

Building Codes. The research effort spans

areas concerning documentation of damage

from past disaster studies to performance of

building envelopes, i.e. cladding, roofing and

glazing, and quantification. A further

development of the research topics in

structural glazing, cladding, and curtain walls

and roofing follows.

Efforts to better design and to improve the

performance of structural glazing, cladding

and curtain walls are taking place at several

academic institutions, within the glass

industry, and among curtain wall

consultants. Problems with performance of

such systems are associated with the special

nature of hurricane winds, being turbulent,

sustained, varying in direction, and are often

containing wind-borne debris. Measures to

improve the performance of glazing in these

respects are being investigated to ensure

integrity of the building envelope following

impact from wind-borne debris and the

subsequent repetitive action of wind gusts.

For the application of repetitive wind

pressure with desired intensity and temporal

characteristics on a portion of cladding,

pneumatic chambers are being utilized. The

University of Missouri at Rolla and Texas

Tech are involved in glass studies, whereas

Clemson University is pursuing both

envelope and roofing studies.

Some attention in research has also been

devoted to sealant performance. While

previous studies had indicated that sealants

are capable of resisting cyclic loadings,

current studies have been considering their

performance under the joint effects of cyclic

loading and environmental factors. A joint

research effort between Texas Tech

University and Colorado State University,

called Cooperative Research Program in

Wind Engineering (CPWE), has investigated

the failure strength of sealants under random

wind loading. Still, while improvements in

358
glazing and cladding quality are emphasized,

researchers insist that integration of the

designs of glass, mullions, frames, and frame

anchorage would improve overall cladding

performance.

Most of the studies concerning structural

roofing have been initiated following major

U.S. hurricanes in which it was observed that

the primary failure of roofs during such

extreme events occurred near roof corners and ridges. Wind tunnel studies, coupled

with post-disaster investigation and full-scale

measurements have proven to be vital tools

in developing better roof designs. In one

study, an analysis of hurricane damage was

used to determine the best roof design under

hurricane conditions and the impact of slope

on roof performance. Another study,

following the failure of classrooms in

Western Samoa in 1992, utilized wind

tunnel simulations to investigate the use of

"parasol" roofs, which leave an opening of

10% of the wall height at the top of the wall

to reduce uplift wind forces and improve

indoor comfort by venting hot air. Some

work has also been done in the area of

fatigue characteristics of roof pressures, using

the rainflow count method as opposed to the

upcrossing method to observe the

characteristics of these fatigue characteristics.

Post-disaster studies have also been useful in

determining the negative aspects of roofing

structures, pointing out the need to fasten

metal edge flashing, copings, gutters,

aggregate, and rooftop air conditioners and

equipment which may be ripped from the

roof damaging it or becoming wind-borne

missiles themselves. In one instance, a local

filtering technique using the Haar function

was applied to investigate the relationship

between the approach flow and the

dislodging of loose-laid roof pavers. Future

roofing studies at Clemson will focus on

measurements of flow over the roofs of scale

models in a boundary layer wind tunnel using

a three component fiberoptic laser Doppler

anemometer system.

Most of the interest in low-rise building

testing and codes has been spawned by the

destructive path of the recent U.S.

hurricanes, e.g. Andrew, Hugo. Based on the

observations of engineers during the

aftermath, suggestions have been made on

the improvements to low-rise buildings,

which were neglected in most engineering

study until now. Many wind tunnel

simulations are underway to develop

improvements in building design to be

incorporated into building codes. One study

at Clemson utilized a device, BRERWULF,

to apply fluctuating pressures to cladding

panels to study its effects under loading.

Current research at the University of Notre

Dame will also turn toward the development

of a dynamic wind load simulator to

investigate the performance of building

components and systems under dynamic

space-time variation of loads. Such tools were

used to develop retrofit roofs to withstand

severe winds and improve cladding so it may

better function as a structural resistance and

load path. Recent studies stress the

improvement of the building envelope,

consideration of increased internal pressures

due to envelope rupture, and the need for

retrofit measures to improve the connections

on existing structures.

While some research projects have devoted

themselves entirely to roofing, low-rise

building tests, or cladding tests, one joint

project has undertaken investigations of all

three topics and others, thus warranting

appropriate attention. A joint research effort

Texas Tech University (TTU) and Colorado

359
State University (CSU), CPWE, as mentioned before, has developed a five year

program funded by the National Science

Foundation to investigate various topics in

wind engineering, of which include areas of

cladding, roofing, and low-rise structure

testing. TTU possesses a field laboratory,

WERFUL (Wind Engineering Research Field

Laboratory), established in 1987, ideally

suited for the study of low-slope roofing

systems and the measurement of wind speeds

and the associated wind-induced structural

response of a one story test building. Not

only has WERFUL proven to be a source of

data for the study of single-ply membrane

roofs, but it has also proven to be a vital tool

in determining the average wind pressure

acting on a roof purlin and the wind

pressures above and below loose-laid concrete

pavers over an adhered single-ply membrane

roof covering. WERFUL has also been used

to measure wind-induced pressure

coefficients across the building centerline and

the internal pressure coefficients for no

opening, static openings of 1%, 2%, and 5%

and sudden openings of 2%, providing insight

into the increased internal pressures

following the rupture of the building

envelope. All this data has gone on to provide

a basis for the development of analytical

models for the stochastic modeling of wind

speeds and the determination of

wind-induced pressures and admittance

functions of the modified quasi-steady

theory. Other work has also been devoted to

structural sealants, the details of which were

mentioned earlier, as well as information on

other topics which will be alluded to in the

appropriate sections.

Colorado State University's contributions to

this joint effort are mainly in the

investigation of fundamental and applied

topics of wind engineering using physical

modeling, performed in their long-test

section boundary layer wind tunnel,

analytical solutions, and more recently,

numerical simulations. The data collected

from wind tunnel simulations at CSU may

then be compared to the full-scale

measurements from WERFUL at Texas Tech.

In addition, CSU is also involved in

cooperative research with Japan.

4. DAMAGE CONTROL AND

MITIGATION

Several studies are focussing on efforts to

control or mitigation of damage by

eliminating poor design or construction

practice, which emphasizes neither the

provision of redundancies nor the elimination

of weak links in structural systems that can

lead to failure at wind speeds well below

design values. Some of key items of interest

include improved wind code specification,

better code enforcement, learning from post

disaster investigations, need for redundancy

and ductility in manufactured homes,

improved standards for building envelope

systems, utilization of expert systems,

computer graphics technology and self-help

retrofit activities.

Aside from the concentrated efforts of

individuals at several institutions devoted to

damage control and mitigation, there are focussed group efforts to improve

performance of structures under extreme

wind events. The two leading examples are CSU/TTU Cooperative Program, mentioned

earlier, and a program at Clemson

University. A brief description of the program

at Clemson is given here. Funded through the

State of South Carolina and the Federal

Emergency Management Agency (FEMA) the

focus of this program is to mitigate wind

hazards to low rise buildings. The major tasks

360
include: evaluation of loads resulting from

strong winds; evaluation of the resistance of

enclosures; stability of low rise wood and

masonry buildings; and improvement of

codes; standards, and inspection practices.

The project is slated to span five years, with

work beginning at Clemson University in the

summer of 1991.

On a larger scale, the American Association

of Wind Engineering (AAWE), formerly

known as the Wind Engineering Research

Council, Inc. (WERC), established in 1970,

has also taken an active role in promoting

wind hazard mitigation. AAWE hopes to

implement a plan for legislative action that

would foster the initiation of a National

Wind Hazards Reduction Program. Such a

program would provide support for wind

engineering in the U.S. through improvement

of understanding and codes, providing

advanced methods for predicting wind

loading, and encourage the development of

new technology. AAWE also hopes to

accomplish numerous tasks, among which

include initiation of an active sharing of

technology within the field primarily through

their publication, "Wind Engineer,"

promotion and prioritization of wind

engineering research, and leadership in

learning from wind storms through

post-storm data collection and analysis.

The dynamic load effects, structural

reliability and risk assessment are receiving

an increased attention from researchers while

the need for practical application of these

concepts to design is increasing. Dynamic

effects are treated in codes primarily through

gust factor approach, which has been the

focus of concentrated effort particularly to

improve our understanding of hurricane wind

fields. The dynamic loads introduced by the

acrosswind and torsion effects are being

studied for developing simplified procedures

for code applications, primarily with a

reliability based format. Studies focussing on

this aspect for both the low and high-rise

construction are helping to provide a basis for

preliminary discussions on reliability-based

design.

The area of risk assessment has some basic

focus on gathering data after a disaster and

using such findings to develop a model to

assess the risk of failure of constructed

faculties and provide a quantitative measure

to estimate the extent of damage under a

particular intensity of wind storm. This

includes damage to structure as well as building contents due to damaging storm

winds and wave action. Various

subcommittees in the U.S., mcluding the

U.S. Subcommittee on Natural Disasters

Reduction (SNDR), have made it their goal

to reduce human fatalities and suffering,

environmental damage, and economic loss

caused by natural disasters.

Computational schemes, through this effort, are being developed to forecast the

probability and severity of hazardous events,

namely hurricanes, over varying temporal

and spatial scales. These schemes are based

on a synergistic combination of essential

ingredients of risk assessment, with the

damage potential is being assessed through

damage probability matrix, neural networks

and fuzzy logic, as well as the use of

probability trees. Artificial neural networks

have been also developed to predict the

damage of individual buildings. Work has

been devoted to developing ways of

expressing frequency, probability or level of

confidence, and probability of frequency for

quantifying the likelihood of an event. Fur

hurricane models, features would include

relations of the geophysical environment,

361
including global warming, and features of the

hurricane itself. Methods considered in such

an analysis include the Monte Carlo

simulation and the extended bootstrap

simulation, as well as numerical models

which predict the sea surface elevations and

currents due to mesoscale processes like tides

and storm surge. The overall risk assessment

is also being considered in the light of

projected consequences of global climate

change. Studies in this area are being done at

the University of Notre Dame, Clemson,

Texas Tech, Texas A&M, University of

Washington, Stanford, National Institute of

Standards and Technology and Johns

Hopkins University.

In addition to university involvement, risk

assessment has been approached by other

organizations and private firms. The INEL

Advanced Combined Environments Test

Station (ACETS) Program hopes to develop a sound, scientific, engineering and economic

basis for evaluating the ability of a building

to resist wind loads and wind effects which

will lead to reduced losses in wind storms. Its

first goal is to establish realistic test methods

for evaluating the performance of the

building components, connections, and

systems in severe wind storms. Then, INEL

will focus on the development and

improvement of assessment tools which will

be used to identify key areas of building

deficiency as well as identify appropriate

levels of economic expenditure to upgrade

construction practices or for expenditures for

retrofitting. Thus a cost/benefit assessment

scheme will be developed. Other groups

involved in risk assessment include State

Farm Fire and Casualty Co. and the Texas

Department of Insurance, funding a program

for Texas A&M University.

Previous discussion of risk assessment has

focused on the property damage and its

relationship to the hazard. However, a

society must view this topic from a holistic

point of view and extend it to ascertain the

impact of disasters on the socio-economics of

the region. The use of "multiple-objective

planning1

' to place non-monetary value on a

damaged good or lost service as a result of

disaster, stressing that losses must also be

evaluated at both a regional and local level, is

appropriate. Thus, through a

"multiple-objective benefit-cost framework,"

both economic and environmental losses due

to a disaster may be estimated. Such an

analysis of damage would also consider

several levels of impact, including geological,

ecological, atmospheric, and human, through

the "with-without principle." This principle

considers all changes in the system that

occurred with the disaster and all changes

that would have occurred had the disaster

not taken place. Meanwhile, the calculation

of economic damages considers both the

interruptions of production processes and the

damages to assets. Such damages are direct

damages. Secondary damages must also be

assessed, which include the effects of the

disaster which trigger a sudden reduction in

demands for outputs within the effected

region. Other considerations include those of

cultural asset loss, i.e. the loss of historic sites or other sides which would be desirably

passed on to future generations. While

market values derived through appraisals are often used to estimate such losses, typically

these losses are evaluated at a variety of

different levels depending on the

"significance" of the site.

5. RANDOM DATA ANALYSIS

MODELLING AND SYSTEM

IDENTIFICATION

The analysis, modeling and system

362
identification of random data are very

important aspects in wind engineering

studies. Recent advances in stochastic

analysis and modeling provide useful tools

towards improved representation and

interpretation of data. The studies in this area focus on time series modeling of random

data, e.g., auto regressive and moving

averages modeling (ARJMA), estimation of

frequency response of structures,

identification of load effects on structures,

simulation of stationary, nonstationary and

non-Gaussian signals, conditional simulation

and wavelet transforms. Activities under this

topic are being pursued at different levels of

concentration at various institutions involved

in wind engineering research, e.g. Cornell

University, University of Notre Dame, Johns

Hopkins, Colorado State University,

Princeton and Rice Universities.

Computer simulations have been developed

for multiple wind pressure time series which

are strongly non-Gaussian due to skewed

spikiness in fluctuations and have strong

cross-correlations of fluctuating features.

Such simulations produce the exact sample

cross-spectra as well as the auto-spectra of

the target signals and have great flexibility in

controlling non-Gaussian fluctuating features

and some statistical properties of simulated

signals, while preserving the target auto- and

cross-spectral characteristics. However, this

approach leads to generation of deterministic

time histories that represent non-Gaussian

features. Simulations of non-Gaussian

features of wind pressure fluctuations under

separated flow regions using higher-order

spectral transformations and transient events

by wavelet transforms have been performed.

This approach leads to stochastic processes

which represent non-Gaussian and

non-stationary features, respectively. This is

one of the current focuses of wind-related

research at the University of Notre Dame. In

the area of stationary processes, ARMA

techniques have also been used in the

simulation of the fluctuating component of

wind and wind-induced pressures.

6. TALL BUILDINGS AND

ENGINEERING STRUCTURES

In the area of tall buildings most of the

innovations in structural system development

come from industry, whereas, system

response analysis, based on analytical

considerations and wind tunnel studies, is

generally conducted in universities (Colorado

State University, CPP, Clemson, University

of Notre Dame). Some of the work is

focussed on research while other work is

development related. In the research area,

some work is still being done to refine the

gust factor approach and to package it in a

closed form expressions rather than the

current version that utilizes plots. One such

approach has been adopted in ASCE 7-95,

ASCE standard minimum design loads for

buildings and other structures. The

quantification of alongwind, acrosswind, and

torsional loads for a wide range of buildings

of different cross-sections and aspect ratios is

essential for developing coupled

lateral-torsional analysis of buildings. The

analysis is being simplified for code

application at the University of Notre Dame.

The torsional response of tall buildings has

also been considered, since it has been

observed that the torsional response can have

a great effect on the total response of the

eccentric building. There may also be some

contribution to the response of an eccentric

building, in the way of interference effects,

by adjacent buildings. Such interactions have

been studied through wind tunnel

simulations, observing that, while an isolated

363
building experiences inherent turbulence, the

same building with interference from

surrounding structures was found to manifest

wake excitation. Special concern has also

been given to situations in which the

frequency of vortex shedding from the

interfering bunding coincided with the

natural frequency of the eccentric principle

building, causing torsional resonant buffeting

to the principle building.

The research in the area of engineering

structures includes towers, transmission lines,

chimneys, which has been actively pursued in

both academia and industry. For example, in

the study of powerlines, researchers have

attempted to estimate the electric drag

coefficients used in determining wind forces

on transmission line structures via numerical

methods based on 2D Navier Stokes

equations for viscous, incompressible, steady

flow. Many studies have also devoted their

efforts to improvement of instrumentation on

power lines and towers for data collection

and often utilize free-air wind tunnels to

compare the drag coefficients estimated in

wind tunnel studies with actual open air

results. Furthermore, efforts are also being

devoted to improved design of sign, signal,

and luminare structures, which, in recent

years, have failed under high winds injuring

pedestrians and commuters below. Studies

have utilized the data from wind tunnel

studies to establish the static drag coefficients

and root mean square estimates of the

lift-coefficients of tapered cylinders in order

to develop alternative, more aerodynamic

cross-sectional shapes (e.g. octagonal) and

designs capable of resisting cyclic loads in

order to avoid future failures.

More recently, offshore platforms are receiving increased attention due to

structural damage to deck structures and

appurtenances in the Gulf of Mexico in the

aftermath of Hurricane Andrew. As a result

of the new breed of deepwater platforms,

which are very sensitive to the gusting action

of wind, focussed activity on the behavior of

such platforms under the combined action of

wind, waves and currents has developed. For

such behavior to be studied, the wind fields

surrounding such structures must be

accurately modeled. The wind field, however,

is not so much influenced by the form of the

sea surface, but rather is sensitive to the

energy loss and rate of momentum transfer

due to surface friction. However, if a

relationship between on-land wind

characteristics and over-water wind

characteristics is developed, existing on-land

models may be somehow extended over the

water to predict the wind fields. There are some difficulties with these attempts for

spatial separations which may exceed the

turbulence length scales, since the correlation

function typically used by land-based

structures inadequately describes the spatial

structure of the wind field over the ocean.

Further development and clarification of this

approach is one topic for current study. The

senior author is also involved in modeling the

stochastic response of tension leg platforms

to random wind and wave fields. In these

studies, both the time and frequency domain

procedures have been developed utilizing the

Cray Y-MP to predict the nonlinear platform

motion.

Under an Office of Naval Research (ONR)

initiative entitled "Reliability of Nonlinear

Ocean Structures Under Stochastic

Excitation," several investigators are studying

the response of ocean systems to extreme

wind and other loadings. The senior author

has a project concerning response statistics of

ocean structures under wind, wave, and

current loads. The major focus of this study

364
rests upon the advancement of techniques for

the determination of extreme response of

moored floating ocean structures to wind,

wave, and current loads. These improved

techniques are then to be implemented in

predicting the performance of such structures

in different ocean environments and, hence,

in predicting loading patterns. Particularly,

the present effort will serve to quantify

structural survivability and serviceability as well as prescribed motion control strategies

for improved structural reliability. Other

universities and institutions that are addressing wind related topics include

Rutgers, Florida Atlantic, Stanford, NIST,

and several Naval laboratories.

Proposed future off-shore structure

investigations include improving

quantification of windfield models, better

modeling of shielding effects, investigation of

overturning caused by uplift forces,

understanding of overall wave profiles, better

design for gust loading factors, and

investigation of the vortex shedding-induced

vibration on cranes and flare booms. Most

research of this nature is being conducted by

the R&D groups of oil companies. At the

university level, work is being done at the

University of Notre Dame, Texas A&M

University, and University of Texas.

7. MITIGATION AND CONTROL

OF STRUCTURAL RESPONSE

The current trend toward butidings of ever

increasing heights and lighter facades has led

to the construction of relatively flexible

structures possessing quite low damping. As a

result, the serviceability of these buildings is

affected by the excessive accelerations

experienced at the higher levels in wind

storms which may cause considerable

discomfort to the building occupants. Ever

since the evolution of high-rise construction,

efforts, both in industry and academia, to

improve the performance of buildings have

developed. Such efforts are devoted to a wide

range of topics from considering alternative

structural systems to aerodynamic

modifications with particular emphasis on

damping systems. Research in active and

passive systems, which includes active mass

dampers, tuned mass dampers, tuned liquid

dampers and hybrid tuned liquid dampers in

single and/or multiple damper configuration,

is currently being undertaken. While passive

damper study has investigated multiple

damper configurations, the area of active

dampers has considered frequency domain

based on H2 control strategies for wind

excited structures. Further research in the

areas of tuned sloshing dampers is exploring

the use of screens, nets, floating styrofoam

beads, and baffles to increase the damping

ability. Semi-active systems are also being

considered in which the building motion is

monitored and fed into a filter designed to

represent the features of the optimally tuned

damper. For example, semi-active tuned

sloshing dampers may be incorporated with a

feedback system, which adjusts a moveable

sliding screen over a fixed screen or regulates

opening vanes to achieve optimal damping.

Research in both active and passive systems

for wind applications is being carried out at

the University of Notre Dame. Other

institutions involved in active/passive systems

studies include 3M Corporation, Colorado

State University, CPP, University of

Michigan and SUNY at Buffalo. The

American Society of Civil Engineering has

two special task committees, one has just

completed a study concerning motion

perception criteria and motion mitigation,

another studying related to damping systems.

The senior writer is chairing the task

committee and it includes membership from

Japan.

365
The primary focus of this task committee is

to address three important aspects of

structural damping systems: a survey of

inherent damping in buildings and bridges,

the various methods of assessing damping

levels from ambient data, and investigation

of the different active and passive damping

systems and their performance (Japanese

building industry's experience). This would

also require technology transfer through

documentation of design procedures for

various damping devices, including passive,

active, or semi-active systems, utilizing fluid

and viscoelastic dampers, as well as inertial

systems like mass dampers.

A joint research program in structural control

between several institutions from the U.S.

and China concerns the full-scale

implementation of passive and active control

devices in the Nanjing Tower, a free-standing

tower approximately 1000 ft tall with two

levels of observation decks, to minimize the

effects of wind and earthquake loads. This

project pools the knowledge of experts in

structural, wind, earthquake, and control

engineering to design, install, and evaluate

the performance of the passive/active devices.

The field testing and performance analysis

will provide useful insight into the validity of

modeling procedures involving wind-induced

response of both the tower alone and with

the addition of control devices, as well as, assess the adequacy of the passive tuned

liquid dampers and the need for future

developments.

In 1992, the National Science Foundation

initiated a five-year research program on

structural control for safety, performance,

and hazard mitigation. The objective of this

initiative was to address a number of

important technological, as well as, practical

issues of cross-disciplinary nature in order to

forge close ties between industry, research

and practice and to accelerate the process of

implementation. The scope of this initiative

includes the areas of earthquakes, natural

and man-made hazard mitigation, controls,

and structural and building systems. Typical

areas of research funded by the program

range from sensor technology, actuator

dynamics, control concepts, damage control,

uncertain and nonlinear systems, and the

actions of wind and earthquakes. Several

wind related projects are funded under this

initiative which include sloshing dampers, the

next generation of liquid dampers, and

general topics in the control of wind excited

structures.

Base isolation systems are also being explored

in which passive energy absorption

mechanisms are installed that significantly

reduce the base shears felt by the building

during a major event. Through such

measures, the natural frequency of the

structural system is shifted below those of the

dominant excitation source, such as an

earthquake or wind loads. Such isolators

feature elastomeric rubber bearings, with or

without lead cores, which support the

building superstructure, allowing deflection

and sustaining the shear that would result if

the bunding were fixed during an earthquake

and may be used in conjunction with viscous

dampers at the top of the building. The

development of such systems has relied upon

investigations of building response in both

wind tunnels and on shaking tables.

8. BRIDGE AERODYNAMICS AND

AEROELASTICITY

Both theoretical and experimental studies are being conducted to improve our

understanding of the interaction of turbulent

winds with the cable stayed and suspension

366
bridges. While theoretical analysis is focussing on the effect of turbulence on

flutter speed of long span bridges,

experimental work has turned toward

improved modeling and quantification of

aerodynamic forces, considering their

sensitivity to turbulence characteristics and

the geometric form of the bridge section. A

thorough understanding of these effects is essential for improvements in both design

and theory. Some recent studies have raised

questions about the validity of some

assumptions made in modeling, i.e. the

validity of a taut-strip model vs. a section

model. In addition, the assumption that the

correlation time of excitation is short

compared with the relaxation time of the

dynamic system, which allows the use of the

Markov property, may not be appropriate

when noise is modeled properly. Proper

modeling of noise insists that when

fluctuations are introduced, they cannot be

modeled as white noise, but instead as colored. This is an essential consideration

since colored noise does indeed influence the

system stability quite differently than colored

noise. Section model studies are also

underway which investigate the effect of

modifications of the deck cross-section,

including the addition of plates, streamlined

deflectors, and railings, to enhance

aerodynamic performance as well as insure

aeroelastic stability.

A variety of theoretical studies have been

initiated in hopes of developing

mathematical simulations of the pressure

forces acting on and excitation of bridge

decks. Recent studies have focused on

aerodynamic stability of bridges in buffeting

and self-excited loads using flutter

derivatives, vortex-induced vibration

modeling of bridge decks, and the

modification of linear analysis to nonlinear

cases. One study has attempted to develop

theoretical relations for bridge deck

excitation by utilizing modifications of airfoil

theory and exploiting the analogy of a

streamlined bridge deck to a airfoil. In

addition, the use of the Ibrahim Time

Domain Method, the force oscillation

method, and the system-identification

technique in a single study confirm that good

consistency in detennining flutter derivatives

may be achieved by using any of these three

approaches. Researchers have now also found

ways of utilizing computational wind

engineering to determine the wind flow

around and the forces acting on a bridge

deck. Such techniques, one of which uses the

2D unsteady, incompressible, laminar,

Navier-Stokes equations in a body-fitted,

cuivilinear coordinate system, allows

designers to weed out bad bride deck designs

before wind tunnel tests are applied.

Furthermore, some time has been devoted to

exploring the validity of the Discrete Vortex

Method, finding it to be effective in

simulating flow around a suspension bridge

for modest angles and a Reynolds Number of

200. Full-scale studies are also underway to

validate the theoretical and wind tunnel

findings. Johns Hopkins University, the

Department of Transportation, and the

University of Notre Dame are engaged in

topics under this area. In the past, some of

the work in this field has been conducted at

Princeton and Colorado State University.

Future studies should be devoted to

investigating length scale effects and the

dynamics of the vortex formation near

separation. Furthermore, a multi-disciplinary

research project to investigate the effects of

turbulence length scale and intensity on the

aerodynamics of bridge deck sections

utilizing actively generated turbulence flow

fields is currently underway. The study

367
utilizes active turbulence generation devices,

consisting of controlled air jets and oscillating screens in a large cross-section wind tunnel to

identify the effects of properly scaled

turbulence on the aerodynamics of a variety

of generic deck sections. The use of

oscillating grids in this research allows for the

simulation of turbulent length scales that

cannot be produced with passive

turbulence-generation techniques, while the

jets add energy throughout the spectrum to

supply energy that the grids are incapable of

supplying and to ensure that the flows

contain the appropriate fine-scale turbulence.

Thus, this technique, by allowing for large

models, permits the inclusion of additional

deck geometric details that are often omitted

due to scaling requirements. In order to

determine unsteady buffeting and self-excited

aerodynamic loads on the deck sections,

synchronous pressure, pneumatically

averaged pressure, and force measurements

are collected. The study also relies upon

system identification techniques to evaluate

flutter derivatives by expressing equations of

motion in a state-of-space form and utilizing

the Gauss-Newton algorithm for

minimization.

9. KNOWLEDGE BASED EXPERT

SYSTEMS (KBES) AND

COMPUTATIONAL WIND

ENGINEERING

Knowledge based expert-systems can offer

intelligent assistance to designers and

planners in performing a vast array of tasks.

Though KBES are not without their

limitations, the potential applications of

KBESs in wind engineering are many. They

may include analysis and prediction of

extreme winds, risk and damage assessment,

design and synthesis, monitoring and control,

urban and coastal community planning, and

decision making for hurricane response.

Currently most of the effort in expert system

development is focussed on damage and risk

assessment utilizing expert system shells. The

University of Washington, Stanford, Texas

Tech, National Institute of Standards and

Technology and Clemson University are

active in this area.

The NIST staff is developing an interactive

computer program addressing the wind

loading provisions of the ASCE standard

7-95. The software and the instructions for

the user were developed by an expert system

consultant under contract with NIST, which

provided guidelines on the structure of the

interactive program and assisted the software

contractor in the interpretation of the

standard wind loading provisions. The

software was subjected to extensive

verification.

Recent advances in computational fluid

dynamics are noteworthy, but major

challenges remain in simulating flows

common to wind engineering where both the

structural geometry and the flow field are complex. Initial work in CWE has been

focussed on a model which has succeeded in

reproducing some of the integral features of

flow field but some discrepancies between

numerical results and experimental data still

remain. The flow field is modeled by the

Navier-Stokes equations with appropriate

boundary conditions. Turbulence models

have been developed to simulate high

Reynolds number flows that are not

achievable with direct numerical simulation

(DNS) due to computer limitations or to

simulate a flow with the same Reynold's

number similar to DNS but with less

computational effort. Currently the most

widely used turbulence model is the k-epsilon

method, which has the advantage of lower

368
computational effort, but with a lower level

of accuracy. The large eddy simulation (LES)

approach has emerged as a more attractive

scheme which has the promise of providing

improved results with reasonable

computational effort. In LES, fluid motion

scales larger than the filter size are resolved

and solved directly, while the scales smaller

than the filter size are modeled by a subgrid

scale model. The most commonly used LES

simulation relies on the Smagorinsky model

for subgrid scale viscosity. Coherent flow

structures are responsible for the large scale

pressure field rather than the eddies

associated with the transport of momentum

and energy, therefore, a scheme that models

essentially coherent structures is appropriate

for wind engineering applications. It is

referred to as coherent structure capturing

(CSC) to distinguish it from LES. One should

not expect CSC to simulate flow field

characteristics that are associated with small

scales. The LES for capturing of coherent

structures has shown improved accuracy,

upon comparison of predictions and

experimental data, but not without high

penalty in computational effort.

The development of different approaches are underway at various other institutions to

improve our modeling capability of wind

engineering related problems (e.g., Colorado

State University, University of Notre Dame,

Texas Tech, University of Washington,

University of Arkansas and Johns Hopkins

University).

10. HURRICANE WIND SPEEDS,

REGIONALIZATION OF

EXTREME WINDS, VORTICES

AND CYCLONIC WINDS

Hurricane wind speeds may be quantified by

measurement of the footprints of its wind

field near shore and after landfall. Based on

these measurements, recent studies have

shown that the gust factors in hurricane wind

field are higher than those presently used for

extratropical storms. This is in part attributed

to the movement of convective cell that

introduces both nonstationarity as well as non-Gaussianity, thus precluding use of

conventional analysis. On the simulation

front, Monte Carlo based simulations of

hurricanes, based on phenomenological

models, along with historical data on the

wind field and filling of hurricane, provide

very useful information for design.

Alternative simulations based on empirical

simulation schemes are also being studied

which do not strictly account for the

assumption of independence among the

modeling parameters. In essence, the

empirical simulation technique avoids, in a nonparametric way, the whole issue of joint

probabilities among parameters by utilizing

information based on actual historical data.

Studies have also been conducted to compare

Shapiro's model, new gust factor curve and

radial profile exponent equal to 0.5, with the

ASCE 7-88 model for hurricanes into each

model by feeding statistical information from

previous hurricanes and seeing how well it

simulates the winds experienced during the

actual event, concluding that the Shapiro

model, in most instances, appears to be

superior to the more conservative ASCE 7-88

model. Continuous wind fields are also being

computed from pressure data from previous

hurricanes through the use of a finite-element

representation called SAFER. The applied

Research Associates, National Institute of

Standards and Technology, Hurricane

Research Laboratory, Wind Engineering

Research Council, University of Notre Dame,

University of Wyoming and University of

Hawaii are involved in research concerning

this topic.

369
A new method, using computational fluid

dynamics, for numerical simulation of the

turbulent flow field around a structure with

oblique wind angles based on a composite

grid, has been proposed in a recent study.

The method consists of a composite grid

technique which connects two or more grids

supported by a fortified solution algorithm to

simulate 3D flow fields for various wind

angles. Another numerical model of air flows

created by the TTU/CSU joint research effort

has produced a model capable of predicting

the detailed flow field surrounding a building or other structure. The technique adopts a version of the CSU Regional Atmospheric

Modeling System (CSU-RAMS) which is a nonhydrostatic compressible model designed

for large eddy simulation. Other studies have

applied techniques like discrete vortex

methods which seem to show promise over

grid based techniques.

In order to provide a better estimate of

extreme winds at an ungauged site, regional

zoning for extreme wind data collected at

various weather stations in a large region is

being developed. The applications are focussing on regional and the entire U.S.

Colorado State University, Texas Tech

University, University of Hawaii and

National Institute of Standards and

Technology are working toward development

of such regional zoning throughout the U.S.

The CSU/TTU joint effort has also been

active in extreme wind research, updating the

ASCE 7 wind speed map and investigating

the sampling errors in extreme wind records.

While hurricane damage has captured the

attention of the wind engineering

community, there is always concern for

tornado-type winds. Studies of tornado wind

forces have been conducted in low-speed,

open-circuit wind tunnels coupled with an

attachment to provide circulation of the flow,

as angular momentum is introduced by the

rotating of a wire screen. Such studies have

proven the necessity of introducing surface

roughness, via sandpaper, to wind tunnel

simulations to accurately reproduce the

vortex flow, delaying the transition to

multiple vortices at low swirl ratios.

TTU/CSU program has also been active in

this area, attempting to identify major

parameters and examine scaling relations and

techniques for simulation of high-wind

phenomena. These efforts have produced

parameters for meaningful wind tunnel

simulations and developed a tornado or

microbust generator. Even through such

efforts, it is still difficult to study tornado

effects in wind tunnels because of the

frequent changes in wind speed and

direction. Consequently, efforts are now

being devoted to exploring the use of

computational fluid dynamics to study

tornado effects. Such CFD models apply the

3D, incompressible, unsteady Navier Stokes

equations, integrated by the control volume

procedure, with convective terms

approximated using hybrid upwinding, solved

on a rectangular grid system. There is,

however, the need for future study of

turbulence models like the k-epsilon model or the Reynolds stress model, focusing on their

lack of proper information on the turbulence

statistics of the tornado close to the

boundary layer.

Besides research related to the effects of

severe weather on infrastructure pursued at

universities, efforts by the federal

government to better detect and define

events during these storms are conducted at

NOAA's Hurricane Research Division in

Miami and Severe Storm Laboratory in

Norman, Oklahoma. The thrust of work at

these laboratories focuses more on the

370
meteorological aspects which are useful for

wind engineering applications.

1 1 . WIND ENGINEERING

RESEARCH APPLICATIONS

One of the major objectives of research in

wind engineering is to improve our

understanding of the wind load effects on

structures for subsequent implementation in

codes and standards. So in this context the

major application of research is for the

development of codes. Building codes in each

locality are responsible for the design and

construction of buildings. Most communities

across the United States adopt, in large part,

one of the three model building codes: the

National Building Code of the Building

Officials and Code Administrators

International (BOCA), the Standard Building

Code, or the Uniform Building Code.

The provisions for wind loads in these model

building codes are patterned after the ASCE

standard on Minimum Design Loads for

Buildings and Other Structures, ANSI/ASCE

7-93. Recently, a major effort has been

completed toward the next revision of ASCE

7-93. A number of university researchers,

practicing engineers and code officials have

been working together to draw information

from recent research findings for inclusion in

the revised version of ASCE standard.

12. CONCLUDING REMARKS

In this article, an attempt has been made to

summarize the wind engineering related

activities in the U.S. Additional details can

be obtained by contacting individuals in the

several institutions mentioned in this paper.

The author sincerely regrets an inadvertent

omission of any institution from the different

wind related activities listed in this paper.

Most of the information has been drawn

from the bibliography and through personal

contacts.

13. ACKNOWLEDGMENTS

The authors are deeply indebted to Ms.

Tracy Kijewski for her assistance in the

preparation and typing of this manuscript.

The senior author was supported in part by a NSF Grant # CMS95-22 145.

14. BIBLIOGRAPHY

1. Hurricanes of 1992, "Andrew and

Iniki One Year Later," Proceedings of

a Symposium Organized by The

American Society of Civil Engineers

(Preprints).

2. Hurricane Alicia: One Year Later,

Proceedings of the Specialty

Conference, ASCE, 1985 (A. Kareem,

editor).

3. Proceedings of the Sixth U.S.

National Conference on Wind

Engineering, Vol. I &II, 1989 (A.

Kareem, editor).

4. Proceedings of the Seventh U.S.

National Conference on Wind

Engineering, Vol. I &II, 1993 (G.C.

Hart, editor).

5. "Wind and the Built Environment,"

U.S. Needs in Wind Engineering and

Hazard Mitigation, Panel on the

Assessment of Wind Engineering

Issues in the U.S., National Academy

Press, Washington, D.C., 1993.

6. Kareem, A., "Recent Advances in

Wind Engineering: Part I," Journal of

Wind Engineering and Industrial

Aerodynamics, Vol. 36, 1990.

7. Cermak, J.E., "Recent Advances in

Wind Engineering: Part II," Journal of

371
Wind Engineering and Industrial

Aerodynamics, Vol. 36, 1990.

8. Howe and Cochrne, "Natural Hazard

Damage Handbook," Colorado State

University, 1991.

9. "Cooperative Program in Wind

Engineering," Inaugural Meeting of

the International Wind Engineering

Forum, Tokyo, 1994.

10. "Building an International

Community if Structural Engineers,"

Proceedings of Structures Congress

XIV, vol.2, ASCE (S.K. Ghosh &J. Mohammodi, editors).

1 1 . "Engineering for Wind Hazard

Mitigation," Project Summary by the

Wind Engineering Research Council,

Inc., National Science Foundation

Form 1358, 1994.

12. "Wind Loading and Capacities of

Components, Connections and

Systems" (G229).

1 3 . Summary Papers for the Ninth

International Conference on Wind

Engineering, 1995 (Krishna,

Chairman).

14. Schechter, M.M., Schechter, E. and

Simiu, E., "Developmental

Computer-Based Version of ASCE

7-95 Standard Provisions for Wind

Loads," NIST Technical Note 1415,

1996.

372
Estimation of relevant internal pressure and wind loads on cladding

by

H. Okada and L. Kong

Aerodynamics Division, Structural Engineering Department

Building Research Institute ,Ministry of Construction

ABSTRACT

Wind loads on claddings are estimated by the

difference between external pressures and internal

pressures on them. Therefore, to determine both

the building external and internal pressures

accurately is very important. In various building

regulations and standards, the design wind loads on cladding are defined as the difference between

maximum external and maximum internal

pressures ( in a sense of the absolute values). This comes from the assumptions that external and

internal pressures have different signs from each

other and the correlation between them is perfect.

However, external and internal pressures do not

always have different signs and the correlation is not always likely to be perfect either. So far, the

assumptions resulted in either over-estimation or under-estimation of wind loads on cladding. In

spite of this situation, there has been relatively

very little attention to the internal pressure. The

effect of correlation between the internal and

external pressures on cladding has not been

investigated at all. In this study, the effects of the

internal pressures and correlation between

external and internal pressures on wind loads on

claddings, are investigated. Wind tunnel tests are conducted in a simulated natural wind conditions

by using building models which have various

combinations of openings. From the tests, the

difference in the wind load estimation between the

ordinary method and the accurate methods is found to be quite large. A new concept of gust

factor for the building internal pressures is introduced to provide the accurate design load.

KEYWORDS: Wind load on cladding, Internal

pressure, Correlation between internal and

external pressure

1. INTRODUCTION

Each year, strong windstorms cause serious

damages in the world. Damages caused by

Typhoon 9119, 9313 and etc. in Japan, Hurricane

Andrew, Iniki and etc. in USA are still fresh in our memory. Looking back on the building damages,

almost all are on the external elements such as

roofings, window glasses, external walls and roof

structures, etc. Damages on the claddings

produced large increasing of internal pressure and

lead the serious collapse of the building. It also

produced very dangerous flying debris. Such

damages warn us to pay more attention on wind

resistant design of claddings.

The purpose of this work is to examine the

wind loads on cladding which are formulated by a combination of external and internal pressures.

Internal pressures are made from the external

pressures through various openings on external

elements of the building. The openings on a

building can be classified as two kinds, the

dominant type and leakage. A theoretical model is

set up to obtain the characteristics of internal

pressures through small openings. The model is verified by a speaker test. To obtain experimental

data of wind loads on cladding, a wind tunnel test

is conducted in a simulated natural wind. In the

test, internal, external pressures on claddings and

their correlation are measured. Wind loads on

claddings are calculated by the ordinary method

and the accurate method from the experimental

data. The results are compared with each other.

2. THEORY OF WIND LOAD ON CLADDING

Mean wind internal pressure

It is well-known that mean wind pressure can be

estimated by the mass flow balance in and out of

the building. For a building with a number of N

openings, the equation can be written as:

±Qj =0 CD

From which a relation between the external

and internal pressures is established:

fi sign(Pej.-Pi )(A

J \Pej -Pl f) = 0 (2)

The sign of each term depends whether it is inflow or outflow. In the equation, power index of

1/2 becomes 1 if sizes of the openings are very

small and the effect of viscosity can not be

373
neglected.

Fluctuating component ofinternal pressure

The relation of the fluctuating components of

external and internal pressures was developed

from the Bernoulli equation based on the internal

flow condition inside the opening by Liu and

Saathoff[l] without considering the viscous effect. The relationship can be written by the following

Helmholtz model:

pLe V0 d

2 P, nKAP„ dt

2 + 0.5/7

K dPt dP,

n

2 K2 A2 Pa2 dt dt + p = p„

(3)

Where,

Le = effective opening length

, which is defined as: Le = L0 +

L0 = length of opening

Pi = internal pressure

Pe = external pressure

p = air density

K orifice discharge coefficient n =poritropic exponent

Pa atmospheric pressure

A = opening area V0 = volume of building

A very important parameter, the Helmholtz

frequency can be written as: s _ _L l ^K^A > 2*\pLt V0 which determines largely the characteristics of the

wind induced internal pressures. However, for

long and narrow openings, the viscous effect

should be a dominant factor. In order to take into

consideration of the viscous effect, a similar

relation as the above Helmholtz model is developed from the Navier-Stokes equation for the

flow inside the opening:

pLe V0 d

2 Pt nKAP„ dt

2 + 0.94/7

V

2 dPt n

2 K

2 A2 P2 dt dt

(4)

It can be easily seen that the only difference

between the two equations is the damping

coefficient of the second term, which is now

0.94 [2] instead of 0.5 in the original one. Therefore, the effect of adding the viscous effect

is the same as to increase the damping, i.e., a loss

of energy from the flow.

Verification of the theory is done by a speaker system, which is used to generate

external pressure functions in a sinusoidal form at different amplitudes and frequencies. A box with a dimension of 90mm x 90mm x 55mm was used to simulate the room in which one wall can be

replaced by plates with different sizes of openings.

Pressure transducers are connected to both the

speaker and the box. During the tests, the box was

fixed to the speaker, so pressures from inside and

outside of the box, which represent respectively

the room internal and external pressures, can be

measured and recorded simultaneously by a data

acquisition system. Several test pieces with

different thicknesses and different sizes of

openings were used. Here results for 3 of them

with rectangular shape, which are listed in table 1, are presented.

Figure 1 shows Pt / Pe between the theory

and experiment at Pe = 7$N / m

2 for R10L. The

two theoretical results are for the two damping

coefficients 0.5 and 0.94. The Helmholtz

frequency is about 52Hz. Experimentally, the

figure shows a roughly 15% amplification of the

internal pressure comparing to the external

pressure around the Helmholtz frequency.

Apparently the viscous effect should be included

as the result without the viscous term shows a much larger value than the experimental data.

However the present theory obtained a very close

solution.

Figure 2 is another example of PJ Pe between

the theory and the speaker test for R05L. It is interesting to see that the peak of Pt /Pe disappeared completely in the test. This means

that once the opening is very small, there is no

need to worry the occurrence of the Helmholtz

resonance. Obviously, even with the viscous effect

the results of the theory and experiment are totally

different. It is believed that flow physics in the

opening has changed and the effect of viscosity is not only confined to a thin layer near the wall but

the whole flow region.

374
Another important parameter regarding to the

dynamic characteristics of the opening is the

phase difference between the internal and external

pressures. Figure 3 shows the results for R20L.

The theoretical solutions on the figure are obtained with the damping coefficient 0.94.

Though there is some difference when / > 200

between the experiment and theory, they generally

show a good agreement between the present

theory and experiment. It should be pointed out

that if the old damping coefficient 0.5 is used, a much larger phase value, thus a larger deviation

from the experiment would be obtained. The

sudden phase decrease at f=300 shown in the

graph is considered to be due to the speaker resonance at the vicinity of this particular f value,

which resulted in the irregularity of the signal.

Table 1 : Test models

models openings thickness

R05L 0.5mm X 10mm 15mm

R10L 1.0mm X 10mm 15mm

R20L 2.0mm X 10mm 15mm

Pl/pe

Experiment

Theory, damping coeff. 0.5

Theory, damping coeff. 0.94

200 300 Frequency f

500

Figure 1 : Comparison of Pt IPt ratio between

theory and experiment for R10L ,Pe=75N/m2

Pl/Pe

-Pe = 16 N/m2, theory

Pe = 8 N/m2, theory

Pe = 16 N/m2, experiment

Pe = 8 N/m2, experiment

100 200 300 Frequency f

400 500

Figure 2: Comparison of PJPe ratio between

theory with damping coefficient =0.94 and

experiment for R05L

theory, Pe=78N/m2

theory,Pe=16N/m2

theory, Pe=8N/m2

expt, Pe=78N/m2

O expt, Pe=16N/m2

O expt, Pe=8N/m2

100 200 300 Frequency f

400 500

Figure 3: Phase difference between the internal

and external pressures for R20L with C0 =0.94,

experiment and theory with 0.94 damping

coefficient

Wind load on cladding

As pointed out in the previous sections, the total

loads on cladding depend on not only the absolute

internal and external pressure values but also the

correlation between them. However almost all standards

375
and regulations in the world give the design wind loads on cladding as the following equation or eqations based on a similar concept.

Fd =Ge Ie -GM (5)

Where,

Fd = design wind loads on cladding

Ge = gust response factor of the external wind pressure

G = gust response factor of the internal wind pressure

P = mean external wind pressure

p. = mean internal wind pressure

This formula comes from the idea that the signs of

external and internal pressures are different from each

other and maximum loads on cladding are given by the

difference between the maximum external pressure and

the maximum internal pressure. However, Okada and

Chino[3] indicated in their theoretical analysis that this is not accurate because the signs of external and internal

pressures are not always different from each other.

Furthermore it does not take the correlation of the

internal and external pressures into consideration. If the

external and internal pressures are not in perfect

correlation, this formula would result in either over- estimation or under-estimation of wind loads. In their

analysis, they proposed a new concept of internal

pressure gust factor in the sense that if G,Pj is

calculated in the following way, equation (5) would give

the real maximum loading.

G,P, = 7> + Pe (Ge - 1)

sign(Pe - P: )gf af (6)

Where,

g

= peak factor for difference between the external and

internal pressures

cr = rms of difference between the internal and

external pressures

and,

af = ^]+a]-2ae

, (7)

Where,

ac = rms of the external pressure

a = rms of the internal pressure

a = covariance between the external and internal

pressures

Note that the definition of the internal pressure gust

factor here is not the same as the traditional concept

which gives the maximum internal pressure.

By using the above theory and various

experimental data of external pressures, Chino and

Okada obtained the internal pressure values and their

gust factors for different building configurations. Based

on the results, they suggested that internal pressures

between 0 to -0.4 and a gust factor of 1.3 should be

adopted in building design[3],[4]. These values are from a theoretical analysis based on test results of only

external pressures. The assumption behind it is that

leakages are evenly distributed on all building walls and

there are no dominant openings on any of the walls. It is believed that most buildings are close to this condition.

These proposed values correspond to a 10 minutes

averaging time in full scale test and a 0.5 second for the

instantaneous values. In order to obtain a close

simulation of the full scale test, the following wind

tunnel test selects the same time reduction ratio.

3. EXPERIMENTAL ARRANGEMENT

To check the building internal pressure in a real wind condition, three sizes of building models are made

for wind tunnel tests. They have two kinds of

openings on their claddings. One of the opening types is made to simulate the air leakage characteristics of

external walls at different equivalent opening ratios. The other is to simulate the dominant openings. Table 2

gives the model dimensions, where (W) in the table

indicates the windward side of the model. The opening

for the leakage type changes with the number of walls

having leakages. The leakages are all 1mm in diameter.

There are total 13 leakage openings on the smaller size

wall and 19 on the larger one. The equivalent leakage

opening ratios for all models, which is defined as the area of opening over the gross area of the wall, are about the same at 0.05%. This value is very common

for buildings in Japan. For the dominant opening, even

for the same model, it varies depending on the size of

the opening area. The three sizes of dominant openings

are 12mm (large), 8mm(medium) and 4mm(small) in

diameter. They are only opened on the windward side

and centered at a height of 125mm on the model. The

arrangements of openings and leakages are given in

Table 3. W, S, S, L means leakages respectively on the

windward, side, side and leeward faces of the model. In

this paper, only results from tests listed on the table are presented.

The wind tunnel is a boundary layer type with a

sectional dimension of 3m X 2.5m. Spires are used at

the test section entrance and roughness are distributed

along the floor to generate a simulated atmospheric

boundary layer flow in the power law profile with an

index of 1/4 in the following form:

* ~ v

0

This simulated atmospheric boundary layer flow

corresponds to an urban area condition. The boundary

layer thickness is around 1000mm. Large number of

376
pressure taps are mounted on both inside and outside of

the models. In the tests, the external and internal

pressures on the walls are measured simultaneously by a multi-pressure measuring system.

Table 2 Wind tunnel test models, dimension unit: mm

Model No. height width depth

No. 1 200 100 100

No. 2 200 100 (W) 150

No. 3 200 150 (W) 100

maximum loading is observed. From the test, it can be

seen that when the maximum load is on the side wall, it appears on the front corner. If the maximum load is on

the front wall, it normally locates at the up center part.

C^is the mean internal pressure coefficient. The last

line shows on which wall the maximum load exists at the corresponding test condition and model

configuration.

As expected, these values are very close to the

proposed ones. Maximum loads for all three cases are on the front wall. If using the internal pressure value of

-0.4 and a gust factor of 1.3 as proposed in the building

code to estimate the design loads, the design will be

safe.

Table 5 Leakages on all faces and without dominant

opening

test number gust factor Cpi maximum load

No. 1 " 1.196 ~ -0.37 front wall

No. 11 1.266 -0.31 front wall

No. 12 ~ 1.292 " -0.35 front wall

Error in the wind load estimations by traditional M>ay

Here two examples, test 1 and test 27, are given to see why the formula FJ = G P - GP has overestimation of wind loads for most cases and why it has

under-estimation of wind loads when the internal

pressure has the same sign as the external pressure if Gfi uses the traditional gust factor concept. The

maximum load results for test 1 are shown in Figure 3

and for test 27 are given in Figure 4. The maximum

loads estimated by using the new gust factor concept

give the real maximum loads, which are compared with

those estimated by using the traditional concept. For

both cases, the maximum loads act on the side walls.

All values of maximum loads on the graphs are in the

form of coefficient as CFD , which is normalized by the

mean wind velocity pressure at the height of model. The

traditional way is indeed to result either a higher loading

for test 1 or a smaller loading for test 27 on the side

walls. This proved that in estimation of the maximum

wind loads, it is not just the separated values of the

external and internal pressures that matter, it is also

how they correlate each other.

The effect of external-internal pressure correlation on

the wind loads

The differences between the wind loads estimated

from the perfect correlation assumption and from the

real loads are shown in figures 3 and 4. Here result of

wind loads estimated from the no correlation

assumption are also given in Figures 5 and 6. They are

Table 3 Building model test conditions

test number

dimension dominant permeability

testl 100 X 100X200 no W,S,S,L

test 11 100(W)X 150X200 no W,S,S,L

test 12 100 x 150(W)X200 no W,S,S,L

test23 100 X 150(W)X200 W(large) W,S,S,L

test27 100 X 150(W)X200 W(medium) W,S,S,L

test28 100 X 150(W)X200 W(small) W,S,S,L

4. RESULTS ANALYSIS

Mean internal pressure

Based on the relation between the internal and

external pressures by using the condition of mass flow

balance in and out of the building, internal pressure is calculated from the external pressures. The comparison

with the measured internal pressures are given in Table

4 for the three models. As expected, they are very close

to each other. For all tests with or without dominant

openings, with evenly distributed or with unevenly

distributed leakages, the internal pressures from the

experiments are all very close to their corresponding

theoretical values.

Table 4 Comparison of the internal pressure coefficients

test No. models C experiment Cpi , theory

No. 1 No. 1 -0.37 -0.40

No. 11 No. 2 -0.31 -0.32

No. 12 No. 3 -0.35 -0.40

No.23 No. 3 0.54 0.55

No.27 No.3 0.40 0.43

No.28 No.3 -0.04 -0.06

Gustfactor of internal pressure

Table 5 shows a group of experimental results at the test conditions of evenly distributed leakages and no

dominant openings for all three models. These

conditions are the same as in the proposals. The tabled

values are chosen from the place on the model where a

377
compared with the real wind loads as well. Underestimation of wind loads is observed in the plus external

pressure zone and over- estimation in the minus

external pressure zone in test 1 . In test 27 the over and

under estimations are found in the contrary zones to test

1. The overestimation in test 27 is quite large.

5 67 8

with new gust factor

with old gust factor

Figure 3: Comparison of the maximum loads estimated from

different ways for test 1

1.5

1 0.5

0

-0.5 S -1 0-1.5

-2

-2.5

-3

-3.5

-4

7 8 9 10

-with new gust factor

with old gust factor

10 15 20

pressure tap positions

Figure 4 : Comparison of the maximum loads estimated from

different ways for test 27

Figure 5: Comparison of the maximum load estimations by

assuming no correlation of external and internal pressure with

the real wind loads for test 1

with covariance

without covariance

10 15

pressure tap positions

20

Figure 6: Comparison of the maximum load estimations by

assuming no correlation of external and internal pressure with

the real wind loads for test 27

5. CONCLUSION

The present investigation proves two points, the

importance of the internal pressure in building design and

the importance of the correlation between the internal and

external pressures in estimating the maximum wind loads.

Both of them have large effects on the implication of

building safety. Though the over-estimation of maximum

wind loads may only result in a higher cost in building

design which is really undesired, the under-estimation of

wind loads obviously would be unacceptable. On the

other hand, the internal pressure value and gust factor

proposed by Okada and Chino is proved to be correct. Usually, buildings have about the same permeabilities on

all of its walls, so the internal pressure values and the gust

factor in the proposal are very practical and close to real

situation.

REFERENCES

[l]Liu H. and Saathoff P. J. /'Building Internal Pressure: Sudden Change", Journal of Engineering Mechanics

378
Division, ASCE, Apr. ,1981

[2] H. Okada and L. Kong, "Investigation of the dynamic

characteristics of wind induced building internal

pressures", to be presented at the Third International

Colloquium on BBAA, July 1996, Virginia, USA.

[3] H. Okada and N. Chino, " Wind - induced internal

pressures in buildings, part 2 gust response factor of

internal pressure", Journal of Wind Engineering, Japan

Association for Wind Engineering, No. 58, 1994.

[4] N. Chino and H. Okada, " Wind - induced internal

pressures in buildings, part 1 mean internal pressures",

Journal of Wind Engineering, Japan Association for Wind

Engineering, No. 56, 1993.

379

Descriptive Epidemiology of Wind-related Mortality

by

Josephine Malilay1

, Annemarie Wasley1, Leslie A. Fiedler1

ABSTRACT

To examine the effects of wind in deaths

attributed to recent tropical cyclones and

tornadoes in the United States, we reviewed

174 deaths reported by medical examiners or coroners (ME/Cs) in Morbidity and

Mortality Weekly Reports during 1989-

1995. Of these 174, we found 99 windrelated deaths, including all 62 of the deaths

attributed to tornadoes and 37 of the 112

deaths attributed to tropical cyclones. Ofthe

wind-related deaths attributed to cyclones,

the major causes were multiple blunt trauma,

crushing injuries, head trauma, and

compressional asphyxia. ME/C's cited

structural failure about five times as often in

wind-related deaths as in non-wind-related

deaths (OR = 5; 95% CI = 1.8-14). On the

basis of our findings, we recommend that 1)

structural failure during high winds be

explored further; 2) uniform methods be

established for collecting information about

disaster-related deaths in which wind is a

contributing factor; and 3) epidemiologic

studies of cyclones and tornadoes include

input from wind and construction engineers.

KEYWORDS: wind, tropical cyclones,

tornadoes, health effects of disasters

1. INTRODUCTION

Through postdisaster investigations of

natural disasters, such as hurricanes,

tornadoes, and severe storms,

epidemiologists and public health officials

have identified certain risk factors for death

or injury, such as occupancy of a mobile

home or motor vehicle during the disaster.

The risk factors identified have in turn

formed the basis for interventions designed

to reduce the number of deaths and injuries

in subsequent disasters.

Wind is usually regarded as an underlying

hazard of a disaster event, rather than as a

disaster event per se (1). The mechanism of

injury in wind-related deaths varies by the

type of disaster. In tropical cyclones and

nontornadic thunderstorms, most injuries are caused by extremely high winds. Injuries in

particular from such high winds most often

occurred indoors and were largely attributed

to 1) multiple lacerations, abrasions, and

fractures from flying debris; and 2) traumatic

asphyxiation from massive trauma caused by

collapsing walls and roofs (2).

In contrast, in a study of severe westerly

gales (with wind speeds up to 96 miles per

hour) in the United Kingdom, researchers

found that most deaths and injuries occurred

outdoors. Several fatalities occurred when

people were blown off roofs or off ladders

(3). Results of another study, again in the

United Kingdom, showed that injuries were

most likely to occur to people who were

1 National Center for Environmental Health, Centers for Disease Control and Prevention, Atlanta, Georgia

30341-3724

381
outdoors during winds that exceeded 60

knots; most of 70 injuries to 58 persons were

attributed to flying debris, falling masonry, or being blown over by the wind. Only three

deaths were associated with structural

collapse (4).

In this paper, we examined the effects of

extreme wind in deaths attributed to recent

tropical cyclones and tornadoes and

characterize this wind-related mortality. We

also call attention to other disciplines, such

as engineering and meteorology, which can

contribute additional information that would

be useful in designing appropriate strategies

to reduce fatalities during future windrelated disasters.

2. METHODS

We selected four tropical cyclones and three

tornadoes for which medical

examiner/coroner (ME/C) data were

available from the Morbidity and Mortality

Weekly Report of the Centers for Disease

Control and Prevention (5-13): Hurricane

Hugo (1989), Hurricane Andrew (1992),

Hurricane Marilyn (1995), and Hurricane

Opal (1995), and tornadoes in Illinois

(1990), Kansas (1992), and Alabama (1994).

We examined descriptive information

provided for each person who died as a

result of these storms: demographic

information (age and sex), cause of death,

circumstances of death, and when the death

occurred in relation to the storm's impact,

namely before the storm arrived (preimpact),

during the storm (impact), or after the storm

passed (postimpact). We defined "windrelated" deaths as those associated with the

effects of wind, either directly or indirectly,

according to circumstances described by the

ME/C for those deaths.

3. RESULTS

Medical examiner/coroners reported a total

of 174 deaths related to tropical cyclones or

tornadoes, 1 12 (64.4 percent) related to

tropical cyclones and 62 related to

tornadoes. Ofthe 112 cyclone-related

deaths, 37 were attributed to high wind and

the remaining 75 were attributed to storm

surge or other elements. All 62 tornadorelated deaths were wind-related.

We focused further analyses on cyclonerelated deaths in which wind was a

contributing factor (Table 1). Demographic

data were available for all but 1 of the 37

wind-related fatalities attributed to tropical

cyclones Hugo in South Carolina, Andrew in

Florida and Louisiana, Marilyn in the U.S.

Virgin Islands, and Opal in the southeastern

U.S. The average age of these decedents

was 45 years (range: 1 to 80 years); 28 (76

percent) were male.

Thirty-one of 37 wind-related fatalities

occurred during the impact phase, when

wind speeds were at or near their highest

(Table 2). Two deaths occurred before the

impact phase: one person died of crush

injuries from a tornado that preceded a

hurricane, and one person sustained multiple

blunt injuries to the head when her mobile

home was hit by a tornado that preceded the

cyclone. Four deaths occurred after the

impact: one person was asphyxiated under a

falling tree, two sustained head injuries when

they were hit by falling trees, and one was

killed by blunt force trauma when his

motorcycle ran into a fallen tree.

Of the 37 wind-related deaths, 14 were

382
caused by multiple blunt trauma, 7 by

crushing injuries, 9 by head trauma, and 7 by

compressional asphyxia (Table 3). Structural

failure was cited more often by ME/Cs as being involved in wind-related deaths than in

non-wind-related deaths (odds ratio = 5;

95% confidence limits =1.8-14). Windrelated structural failures cited by ME/Cs

included "collapsing house, mobile home, or townhouse," "roof of home caving in," "tree

falling onto house or mobile home," and

"tree falling through roof of residence."

Further, structural failure was also

mentioned as a factor in 10 of 75 non-windrelated deaths. Descriptions of such nonwind-related structural failures included

"storm surge in mobile home leading to

drowning," "house fire started by burning

candle," and "explosion caused by propane

cooking device due to power outage."

4. DISCUSSION

Prior to Hurricane Hugo in 1989, high

proportions of cyclone-related deaths were

attributed to drowning from storm surge

(14). In recent years, however, improved

systems for issuing and disseminating

warnings^ more evacuations from areas in the

projected track of hurricanes, and increased

public attention to hurricane warnings have

led to a decrease in the proportion of

cyclone-related deaths that are surge-related.

Deaths are now more often associated

directly or indirectly with high winds.

ME/Cs cited structural failure as a

contributing factor in about one of two

cyclone-related deaths attributed to high

winds, but in only about one of eight deaths

attributed to storm surge. Because structural

failure was cited in 26 of 1 12 deaths,

particularly in wind-related deaths, additional

information about the circumstances of these

deaths ~ including estimated wind speeds at

the time of death and building characteristics

-- should be collected in future disasters.

Wind-related deaths also occurred as a result

of tornadoes, including tornadoes that

preceded cyclones. Presumably, most deaths

happened during the impact phase, when the

winds were at their highest. Any health

outcomes during the pre- and postimpact

phases have yet to be determined, because

"preimpact," "impact," and "postimpact"

phases have not been defined for tornadoes.

We used information provided by ME/Cs to

CDC. This death investigation system is

capable of obtaining information about

victims within several days of a natural

disaster; however, the system would not

include disaster-related deaths that occurred

during the months after the event. One

advantage of this system is that for most

cases, particularly for those related to

tropical cyclones, we were able to obtain

detailed information about the circumstances

of death information that would not be

available on a death certificate and which

usually would be issued several weeks after a

death.

Results from this study indicated the need for

uniformity in collecting information about

disaster-related deaths. Although we found

consistency in the manner in which

hurricane-related deaths were reported by

ME/Cs, we found a broad range in the

manner in which tornado-related deaths were

reported, ranging from simple counts of

deaths in a particular location, such as a school or church, to total numbers of deaths

related to a tornado. A line listing of

tornado-related deaths, much like those

collected for cyclone-related deaths, would

383
be useful for comparing the effects of

different disasters, particularly those

involving extreme wind.

Finally, we acknowledge the difficulty in

determining the risk of mortality, windrelated or otherwise, in the population

exposed to the event. Although the

population census may be available for a

disaster-affected area, estimating the

exposed population from this source may not

be feasible because residents may evacuate, or conversely, return to an area at any given

time. Field methods and statistical analyses

have been developed to address this issue

(15).

5. CONCLUSIONS

As the number of deaths related to storm

surge has decreased in recent years (14), the

proportion of cyclone-related deaths

attributable to wind has increased. These

wind-related deaths, as well as wind-related

injuries, need to be further characterized in

order to explain the mechanisms by which

wind acts as an environmental threat. In

future investigations of the effects of high

wind, structural failure and its components

should be further explored in order to

determine appropriate predictors for windrelated mortality, and more precise wind data

should be collected, including data on mean

wind speeds, wind speed thresholds above

which the risk for injury or death increases.

Therefore, we recommend that

epidemiologic studies be designed so as to

involve the participation of other disciplines,

including wind engineering and construction

engineering. Only with the help of people

from these disciplines can disaster

epidemiologists fully address the factors that

predict wind-related outcomes and design

interventions that will prevent future windrelated fatalities.

6. ACKNOWLEDGMENT

We thank Debra Combs, Surveillance and

Programs Branch, NCEH, CDC, for

reviewing this manuscript.

7. REFERENCES

1 . McDonald JR. Individual

preparedness planning for safety and

windstorm damage mitigation. In: Majumdar SK, Forbes GS, Miller

EW, Schmalz RF, eds. Natural and

Technological Disasters: Causes,

Effects and Preventive Measures.

PA: The Pennsylvania Academy of

Science, 1991:384-99.

2. Eckert WG. Medicolegal aspects of

tornadic storms in Kansas, U.S.A.

Am J Forensic Med Pathol

1991;12:281-5.

3 . Illingworth RN, Illingworth KA.

Multiple casualties caused by a gale.

Arch Emerg Med 1984;1:57-60.

4. Cugnoni HL, Whitworth I. Injuries

related to wind speed. Ann R Coll of

Surg Engl 1992;74:294-6.

5. CDC. Medical examiner/coroner

reports of deaths associated with

Hurricane Hugo-South Carolina.

MMWR 1989;38:754-62.

6. CDC. Deaths associated with

Hurricane Hugo—Puerto Rico.

MMWR 1989;38:680-2.

384
7. CDC. Update: work-related

electrocutions associated with

Hurricane Hugo—Puerto Rico.

MMWR 1989;38:718-20,725.

8. CDC. Preliminary report: medical

examiner reports of deaths associated

with Hurricane Andrew—Florida,

August 1992. MMWR 1992;41:641-

4.

9. CDC. Injuries and illnesses related to

Hurricane Andrew—Louisiana,

1992;42:242-3,249-51.

10. CDC. Deaths associated with

Hurricanes Marilyn and Opal—United

States, September-October 1995.

MMWR 1996;45:32-8.

11. CDC. Tornado disaster—Illinois,

1990. MMWR 1991;40:33-6.

12. CDC. Tornado disaster—Kansas,

1991. MMWR 1992;41:181-3.

13. CDC. Tornado disaster—Alabama,

March 27, 1994. MMWR

1994;43:356-9.

14. French JG. Hurricanes. In: Gregg

MB, ed. Public health consequences

of disasters. Atlanta: Centers for

Disease Control, 1989.

15. Malilay J, Flanders WD, Brogan D.

A modified cluster sampling method

for post-disaster rapid assessment of

needs. Bull World Health Organ. In

press.

385
Table 1 . Wind-related fatalities for selected tropical cyclones

Event

Wind speed, miles

per

hour

Wind-related

deaths

n (%)

Total deaths

n (%)

Hugo, 1989:

Puerto Rico

South Carolina

Andrew, 1992:

Florida

Louisiana

100-104

135, tidal surge 12-

17 ft

sustained to 145;

gusts to 164

0 (0.0)

9 (24.4)

13 (35.1)

1 (2.7)

9 (8.0)

35 (31.3)

14 (12.5)

17* (15.2)

Marilyn, 1995:

US Virgin Islands

Puerto Rico

Opal, 1995:

Florida, Georgia,

Alabama, North

Carolina

105

115, storm surge

10-15 feet

1 (2.7)

0 (0.0)

13 (35.1)

8 (7.1)

2 (1.8)

27 (24.1)

Total 37 (100.0) 112(100.0)

* Complete information was available for 9 deaths.

386
Table 2. Wind-related fatalities occurring before, during, and after selected tropical cyclones

Event

Disaster phase*

Preimpact

Impact Postimpact

Total

deaths

n <%)

Hugo, 1989:

South Carolina

Andrew, 1992:

Florida

Louisiana

Marilyn, 1995:

US Virgin Islands

Opal, 1995:

Florida, Georgia,

Alabama, North

Carolina

Total

0

0

l

f

0

1

13

0

1

10

0

0

0

2

2 (5.4) 31(83.8) 4 (10.8)

9 (24.4)

13(35.1)

1 (2.7)

1 (2.7)

13(35.1)

37(100.0)

* Deaths are categorized as occurring before the tropical cyclone made landfall (preimpact),

during the storm (impact), or after the storm passed (postimpact).

t Of 18 reported deaths with incomplete information, 1 death was known to have occurred

during the pre-impact phase.

387
Table 3. Causes of 37 wind-related fatalities from selected tropical cyclones

Cause of death Number Circumstances

Preimpact phase:

-Multiple blunt

injuries to head

-Crush injury

Impact phase:

-All penetrating trauma

and/or crush injuries

-All multiple blunt trauma

and/or head trauma

-Asphyxiation

Postimpact phase:

-Asphyxiation

-Head injury

-Blunt force trauma

1

1

12

2

1

13

1

1

12

4

1

1

12

1

-Mobile home hit by pre-hurricane

tornado

-Tornado preceding hurricane

-Collapsing house, townhouse, or mobile

home

-House or mobile home struck by falling

tree

-Inside residence

-Killed by flying debris outside home, in

boat

-Roof collapse

-Blown over while working on train

-Pinned by tree

-Car hit by falling tree

-Struck by beam in home

-In boat

-Hit by tree

-Collapsed truck trailer used as shelter

-Collapsing trailer

-Struck by falling tree

-Mobile home struck by falling tree

-Trapped under tree

-Struck by tree

-Hit tree while riding motorcycle

388
JAPAN-SIDE PAPERS PRESENTED

AT MINI-SYMPOSIA
I

I

I

!
NEW EVALUATION TECHNIQUE OF WIND LOADS

ON

BUILDING EXTERNAL ELEMENTS

H. Okada and L. Kong

Aerodynamics division, Structural Department

Building Research Institute

Tsukuba, Japan 305

ABSTRACT

Building external elements such as roofing, roofs, external walls, finishing of external walls and windows etc. are often received serious damages due to strong winds. The damages on the external elements are the most conspicuous

among the damages of buildings due to strong winds. So we should design them with extra attentions. However we still do not have enough information to obtain the accurate wind loads. Wind loads on building external elements are related to the pressures on both sides and the correlation between the two pressures. There are fairly amount of data regarding to the pressures on their external sides. But the pressure data on the opposite side and the correlation between the two

pressures are quite few. The two pressures are referred as external and internal pressures in this paper. The internal pressure should be created from the external pressures through various openings between the external elements. Both

aspects of the internal pressure, its average and fluctuation, affect the load on the elements significantly. The present

investigation studied both, particularly on its dynamic characteristics. An equation relates the internal pressure to the

external pressure is developed from the Navier-Stokes equation. On solving the equation, a simple but very accurate method is introduced to obtain an analytical solution. The speaker test designed in the investigation verified the

theoretical analysis. The procedure described here is applied to a simple building external model. From the application, it

is very interesting to see that the space between the external elements and the building cladding has great role on wind

loads on building external elements. And the area of openings between the external elements is found to have great

effects on the wind loads.

Keywords: building external element, internal pressure, internal space, damping coefficient

INTRODUCTION

Each year, strong windstorms cause serious building damages everywhere in the world. Among the

various kinds of damages, the one on the external elements is the most conspicuous. The damage on a building should be remarkably decreased after the damage on the external elements is reduced. To do that,

we have to investigate the wind loads on them. The wind loads can be estimated with the pressures on

their both sides, which are called the internal pressure and external pressure in this paper. Only when the

internal pressure and its dynamic characteristics are accurately determined, the wind loads on external

elements can be accurately estimated. Since the internal pressure is created from the external pressure,

the creating process should be examined clearly. The purpose of this paper is to describe how to estimate

the wind loads on building external elements. A theoretical model is set up to obtain the characteristics of

the internal pressure by integrating the Navier-Stoke equation through the opening. Speaker test are conducted to verify the theory at different conditions. The procedures on the evaluation of the wind loads

is applied to a simple building external model.

WIND LOAD ON BUILDING EXTERNAL ELEMENT

Wind loads on building external elements are given with the external and internal pressures and the

correlation between the two pressures. The external elements on roof are classified into three types based on the setting way on the buildings. The types are explained by illustrations shown in figure 1. The first type actually is part of the building cladding itself. V-beam roof belong to this type. Internal pressure on

this type is the internal pressure in the building. That is the one generally called as internal pressure. In the

391
second type, internal pressures is the pressure in the space between the external element and the sheathing

boards. Clay roof tile and lose laid roof block, etc. belong to this type. The last type has almost no space

between the element and the sheathing board. Internal pressure in this type is likely to be the one enclosed

at the time of its setting work. But it is true in this type as well as the other types that the internal pressure

is created from the external pressure. Flat seam roofing is one example of the 3rd type. Regardless of the

types, the wind loads is given as the following equation . F = Pe -Pi +sign(Pe -Pi )g/a/ (1)

Where, F : wind load on building external element

Pe : mean external pressure

Pt: mean internal pressure

gf : peak factor of wind load

(jf \ rms of difference between internal and external pressures, which is given by the following

equation.

af = Je; +CT,

2 -2a„ (2)

Where, ae: rms of external pressure

<7i : rms of internal pressure

aei : covariance of external and internal pressures

Internal pressure is created from the external pressure through the opening or gap between adjacent

external elements . If the transfer function is given as H(co), relation between fourie spectra of internal and

external pressures are given by the following equation.

X.(a>) = H(<o)XJ<o) (3)

Where, Xi((o) : fourie spectrum of internal pressure

Xe ((o) : fourie spectrum of external pressure

Rms's of external and internal pressures and covariance of external and internal pressures are given by the

following equations.

ae

2 =\Xe (o))Xe '(co)d6) = \Se (co)d(o (4)

<t,

2 = \H{co)H-{(o)Xe {co)Xe \co)dco = \\H(co)\

2 Se (co)da> (5)

aJ =

f

X/<o)H(<a)X;((o)d(a (6)

Where, Se ((o) :power spectrum of external pressure, and

super script * indicates the conjugate.

space

opening

no space

no opening

Type 1 Type 2 Type 3

Figure 1 Types of building external element

392
MEAN INTERNAL PRESSURE

It is well-known that mean internal pressure can be estimated by the mass flow balance in and out of

the openings. For a internal space connecting to the external space with a number of N openings, the

equation can be written as:

From which a relation between the external and internal pressures is established:

fi sign(PeJ -Pl)(AJ \P

eJ -Pl f) = 0 ( g)

The sign of each term depends whether it is inflow or outflow. In the equation, power index of 1/2

becomes 1 if sizes of the openings are very small and the effect of viscosity can not be neglected.

FLUCTUATION COMPONENT OF INTERNAL PRESSURE

The following equation, obtained by Liu and Saathoffll] through the unsteady Bernouli equation,

governs the relation between the internal and external pressures of a building in a dominant opening

condition. This equation may applied to the problem described here.

nKAPa dr

Where A is the opening area, V0 is the volume of the building, Pa is the atmosphere pressure, p is

the air density, n is the polytropic exponent and K is the discharge coefficient of the opening. This

equation is similar to the well known Helmholtz resonator in acoustics. Theoretical analysis and wind

tunnel tests conducted by previous investigators [2, 3] showed that in a natural wind condition, the so1 InKP A

called Helmholtz frequency / = — 2— determines the characteristics of the wind induced internal

2n \ pLe V0

+ 0.5p

V

2 dp, dp,

n

2 K2 A

2 P2 dt dt + P,=P. (9)

pressures. Le in the equation is the effective opening length defined as Le = L0 + J— and LQ is the

actual opening length.

Previously, it has been largely believed that viscous effect is not an influential factor in the analysis

for building openings. This is right if the opening is large and shallow so the boundary layer inside is very

small, such as a large opening on a glass window. Most openings on buildings do occur like this. However,

if the opening happens to be a very narrow one between adjacent external elements, e.g. a thin rectangular

type, the previous assumption becomes invalid and the viscous effect is expected to be important. Based

on this consideration, a similar relation as equation (9) is developed.

Schematically, the flow inside a rectangular opening can be simplified as in figure 2. For

simplification, flow variation in the z direction is ignored and the flow is approximated as a 2-D case. The

Navier-Stokes equation for a two-dimensional flow in ^-direction can be written as:

du du du 1 dp + 14 + V ~ —+V\

dt dx dy p dx \dx dy

' d u o u + (10)

Because of the symmetry of the flow inside a rectangular opening, only half of the flow from the

centerline is considered. The boundary layer flow in the opening is considered to be laminar because it only represents the inlet part of a straight channel flow. The velocity distributions of u and v have been

393
given by Schlichting[4]. Here the first order of his solution is taken for simplicity. By curve fitting, the

data is fit into a power series:

Velocity in x direction: u = H0 (0.0026n 3 -0.0581rf +0.4251n) (11)

Iny direction: v = j^-(o.OOU?]4 -0.025/;

3 + O.I2677

2 -0.0064 77) (12)

Where n = yJ—W . n = 5 is approximately the border of the boundary layer. Integrate equation (10)

V x

in x direction from entrance of the opening to inside of the room and substitute u, v in it, the following

equation is obtained,

(13)

dt p

Then from the flow mass continuity equation, a relation between velocity u0 and pressure p can be

written as u = K dP

0 KAnpa dt

9Le V0

Put it in equation (13), the following form is established:

nKAP. dt

2

V; dp, dp,

n

2 K2 A2 P; dt dt

+P, =p. (14)

The new equation has a damping coefficient of 0.94. So the effect of adding the viscous term and the

other convective term v — is the same as to increase the damping. Physically this is correct since the

existence of these terms consume more energy from the flow. In the derivation, some of the integration are evaluated by their averages. For

example, [ v —dx - vLe. This simplification not only represents a general picture of the flow, but

Je dy dy

also simplifies the problem greatly. Actually, since the average characteristics in the opening is examined,

averaging techniques can be used in both x and y directions.

T> U=Uo,Tl=5

Figure 2: Flow conditions inside a rectangular opening on a building wall

SIMPLIFIED METHOD

A simplified method is used to solve equation (14) when the external pressure is in a sinusoidal form

nKPAy

, where C0 p, = P, sin(d)t) . Then p = P sin((ot + <|>) . Let A. = is either

pL' V° and A7 r< e = Cp \ )

1 nKPa A2 0 \

0.5 or 0.94. Substitute both pe and pi into the equation, approximate |co5-(cor + (J))| as 1 and combine the

394
terms on the left side: Rsin((£>t + $ + <&)= Pe sin(
t) , where R = 1j(A2 (o

2 f

P* + (-A] <o

2 P2 and

O = tan

l-A(o 2

1 71 A resonance would occur at the Helmholtz frequency co = ,

where <b = — . At this critical JX 2

condition, a significant amplification of the internal pressure will be induced. Therefore if the wind has a

strong component at this frequency, the building internal pressure will increase dramatically and the load on cladding would increase accordingly. More seriously, if part of the building is flexible and has such a

structural frequency, resonance will be induced and there would be a danger of collapse for that part of the

building.

Finally, Pi can be solved and presented in an analytic form,

P

2 =

-(I- A^)

2 + ^l-A^)4 + 4(A2 v>)

2

PJ_ 2{A^)2 (15)

precise, Pe=78

simplified, Pe=78 A precise, Pe=16

simplified, Pe=16

200 300

frequency f

500

Figure 3: Comparison of pt / pt ratio between the

precise and simple methods for C80L with C0 =

0.5

simplified method, Pe=78

precise method, Pe=78

simplified method, Pe=16

precise method, Pe=16

200 300 400

frequency f

Figure 4: Comparison of the phase difference

between the precise and simple methods for C80L

with C„ = 0.5

Qualitatively, neglecting the term cas((M+<()) is similar to increase the damping, which would

decrease the internal pressure amplitude. The exact solution of the equation is also obtained by the RungeKutta method. The results between the simplified method and the exact method are compared in figures 3

and 4 for two external pressure amplitudes Pe = 16 N / m

2 and 78N / m2

. The opening area of the model

C80L is 50.27mm2 and the thickness is 15mm, which gives a Helmholtz frequency about 1 10Hz. Here both

calculations adopted the damping coefficient of 0.5. Figures 3 and 4 show that at most frequencies, there are hardly any difference between the solutions

of the simplified method and the precise method for both the pressure ratio and phase value. The only

noticeable difference is the peak area close to the Helmholtz frequency for the smaller external pressure

Pt =\6N / m

2

. As expected, Pt / Pe ratios around the Helmholtz frequency for the simplified method is

smaller than that of the precise method due to increase of the damping. From figure 3, it is also amazing to see that depending on the external pressure amplitude, Pi could be as high as 6 times of Pe

.

395
SPEAKER TEST AND COMPARISON WITH THE THEORY

Table 1 : Test models

models openings thickness

R05L 0.5mm X 10mm 15mm

R10L 1.0mm X 10mm 15mm

R20L 2.0mm X 10mm 15mm

C80L ^8mm 15mm

Pi/pe

Experiment

Theory, damping coeff. 0.5

" Theory, damping coeff. 0.94

200 300 Frequency f

400 500

Figure 6: Comparison of Pt / Pe ratio between

theory and experiment for R10L

2

1.8

1.6

1.4

1.2

Pi/Pe

,

0.8

0.6

0.4

0.2

0

-Pe = 16 N/m2, theory

Pe = 8 N/m2, theory

Pe = 16 N/m2, experiment

Pe = 8 N/m2, experiment

100 200 300 Frequency f

400 500

Figure 5: Comparison of PJPe ratio between

theory and experiment for R05L with C0 =0.94

3.5

-theory,Pe=78N/m2

theory,Pe=16N/m2

•theory,Pe=8N/m2

expt, Pe=78N/m2

expt, Pe=16N/m2

expt, Pe=8N/m2

400 500

Figure 7: Phase difference between the internal

and external pressures for R20L with C0 =0.94,

experiment and theory

The speaker system was used to generate an external pressure function in the sinusoidal form at

different amplitudes and frequencies. A box with a dimension of 90mm X 90mm X 55mm was used to simulate the room in which one wall can be replaced by plates with different sizes of openings. In the tests, pressures from both inside and outside of the box are measured. Quite a few test pieces were used. Here

results for 4 of them (Table 1) with two different kinds of shapes, rectangular and round, were presented.

Figure 4 shows Pt / Pe between the theory and experiment at Pe =7&N / m

2 for R10L. The two

theoretical results are for the two damping coefficients 0.5 and 0.94. The Helmholtz frequency is about

52Hz. Experimentally, the figure shows a roughly 15% amplification of the internal pressure comparing to the external pressure around the Helmholtz frequency. Apparently the viscous effect should be included as the result without the viscous term shows a much larger value than the experimental data. However the

present theory obtained a very close solution.

Figure 5 is another example of Pt / Pe between the theory and the speaker test for R05L. It is

interesting to see that the peak of Pi / Pe disappeared completely in the test. This means that once the

396
opening is very small, there is no need to worry the occurrence of the Helmholtz resonance. Obviously, even with the viscous effect the results of the theory and experiment are totally different. It is believed that

flow physics in the opening has changed and the effect of viscosity is not only confined to a thin layer near

the wall but the whole flow region.

Another important parameter regarding to the dynamic characteristics of the opening is the phase

difference between the internal and external pressures. Figure 7 shows the results for R20L. The

theoretical solutions on the figure are obtained with the damping coefficient 0.94. Though there is some

difference when / > 200 between the experiment and theory, they generally show a good agreement

between the present theory and experiment. It should be pointed out that if the old damping coefficient 0.5

is used, a much larger phase value, thus a larger deviation from the experiment would be obtained. The

sudden phase decrease at /=300 shown in the graph is considered to be due to the speaker resonance at the vicinity of this particular f value, which resulted in the irregularity of the signal.

In this session, wind loads on a simple model of building external elements are obtained by using the

procedure and concept shown in the previous sessions. The model is shown in figure 8. Its volume of

internal space and total area of openings between adjacent elements are listed in table 3. External pressure

is assumed to be a sinusoidal function having a frequency of co and amplitude of 100kgf/m in the

following equation.

The results are shown in figures 9 to 11. The ratio of internal pressure to external pressure is shown

in figures 9(a) and (b). The ratio is 1 in the lower frequency range, and it becomes smaller than 1 when

frequency becomes larger than about 5 to 10 Hz. The frequency becomes low with the increasing of

internal volume or the decreasing of area between the adjacent elements. Figure 10 shows phase angle

between external and internal pressures. The phase angle is 0 in the lower frequency range. The higher the

frequency is, the lager the phase angle would be. Amplitude of wind load is shown as a ratio to the

external pressure amplitude in figure 11. Wind load in lower frequency is found to be 0 because the phase

angle between internal and external pressures is 0. These results show that wind loads on the external

elements can be reduced by adjusting the volume of internal space and the areas of openings between

adjacent elements. Note that in the present calculation in order to require wind loads on the building

external elements accurately, phase angle between external and internal pressure, etc. has been taken into

consideration.

APPLICATION

pe (t) = 100 sin g* (16)

external element Table 3 Combinations of total area of openings

and volume of internal space for application

models (O designates models used here)

V(mms

) 10 | 20 | 40

V=intemal space

Figure 8: A model of building external elements

100x10s O

397
1.2

1

0.8

b

06

6" 0.4

0.2

0

0.01

-V0=0.0025m3

Vo=O.005m3

•V0=0.010m3

0.1 1 10 Frequency (Hz)

.2

100

1.2

1

0.8 u$0.6

b

0.4

0.2

0

•S Aj=80cm: £ A,=40cm2 2 A1=20cm2

0.01 0.1 1 10 Frequency (Hz)

(b) V0=0.005m3 (a) EAi=40 cm'

Figure 9 Ratio of internal to external pressure amplitudes

100

3.14

"S 1.57

00

0.00

1

£ -157

-3.14

0. )1

•Vo=0.0025m3

Vo=0.0050m

3

•Vo=0.0010m

3

0.1 10 100 Frequency (Hz)

1.2

1

0.8

& 0.6

& 0.4

0.2

0

• V0=0.0025m3 V0=O.005m3

•V0=0.010m3

Figure 10: Phase angle between external

and internal pressures

0.01 0.1 1 10 Frequency (Hz)

Figure 1 1 : Ratio of load to external pressure

amplitude

100

CONCLUSIONS

Some of the important aspects from the current study are summarized:(l) A fundamental equation is described for estimation on wind loads on building external elements; (2) A new equation with a damping

coefficient 0.94, similar to Liu's equation, is developed to obtain a better solution of the internal pressure.

The simple method, employed to solve the equation, is very accurate to obtain both the amplitude and

phase value of the internal pressure; (3) The speaker test can be regarded as a simplified version of the

wind tunnel test. Not only it verified the theory, it is also used to predict the dynamic pressure

transmission characteristics of various openings;(4)Wind loads on the building external elements are changed by adjusting area of openings between adjacent elements and volume of the internal space

between the elements and sheathing board.

REFERENCES

[1] Liu, H. and Saathoff, P.J., "Building Internal Pressure: Sudden Change", Journal of the Engineering

Mechanics Division, ASCE, Vol. 107, No. 2, 1981.

[2] Holmes, J.D., "Mean and Fluctuating Internal Pressures Induced by Wind", Proc. of the 5th

International Conference on Wind Engineering, Fort Collins, Colorado, July 1979.

[3] Liu, H. and Rhee, K.H., "Helmholtz Oscillation in Building Models", Journal of Wind Engineering and

Industrial Aerodynamics, Vol. 24, 1986, pp. 95-1 15.

[4] Schlichting, H, "Boundary Layer Theory", McGraw Hill Series in Mechanical Engineering, 1968.

398
WIND TUNNEL STUDIES ON LONG-SPAN BRIDGES

CONDUCTED AT THE LARGE BOUNDARY LAYER WIND TUNNEL IN

TSUKUBA

by

Hiroshi SATO

Head, Structure Division, Structure and Bridge Department,

Public Works Research Institute, Tsukuba, JAPAN

ABSTRACT

The Large Boundary Layer Wind Tunnel, whose

test section is 41m wide, 4m high and 30m long,

was constructed at Public Works Research

Institute (PWRI) in 1991, as one of cooperative

studies between PWRI and Honshu-Shikoku

Bridge Authority. By using this facility, wind

tunnel studies for the Akashi Kaikyo (strait)

Bridge, which will be the longest suspension

bridge in the world, and for the Tatara bridge,

which will be the longest cable-stayed bridge in

the world, have been conducted.

From the wind tunnel studies, the safety of these

bridges were confirmed. In addition, some new

properties on flutter and gust response of super

long-span bridges were found.

KEYWORDS

Super long-span suspension bridge, Super longspan cable-stayed bridge, Flutter, Gust

response, Full aeroelastic model, Wind tunnel

study, Turbulent flow, Surrounding Islands

1. INTRODUCTION

The study on wind effects is very important for

the design of super long-span bridges. The Akashi

Kaikyo Bridge (Fig.l, Photo 1), which is now

under construction, will have the main span length

of 1990m, which is 580m longer than the Humber

Bridge, the longest bridge in the world at present.

In Japan, the safety of proposed design of longspan bridges to wind-induced instabilities has

been confirmed mainly by spring-mounted rigid

sectional model test in smooth flow. Since the

span length of the Akashi Kaikyo Bridge is much

longer than the existent longest bridge in the

world, it was thought that the wind-induced

instabilities should be confirmed by more accurate

wind tunnel study, namely full aeroelastic model

study.

From these reasons, cooperative study on wind

resistant design of super long-span bridges was

started in 1989 by PWRI, Ministry of

Construction and Honshu-Shikoku Bridge

Authority (HSBA). The Large Boundary Layer

Wind Tunnel was constructed at PWRI in 1991 as one of the cooperative efforts between PWRI and

HSBA, and wind effects on the Akashi Kaikyo

Bridge were investigated in smooth flow and in

turbulent flow using the full aeroelastic model,

whose scale ratio was 1/100.

The Tatara Bridge (Fig. 2, Photo 2), which is also

under construction in Japan, will have the main

span length of 890m. The bridge will be the

world's longest cable-stayed bridge in future.

Since the Tatara bridge has such a long span, it was thought that the aerodynamic behavior should

be confirmed through full aeroelastic model test at the Large Boundary Layer Wind Tunnel in PWRI.

The Tatara bridge will be surrounded by islands.

Some of the islands have high mountains up to 400m above the sea level. Therefore, it was

considered that topographical effect should be

studied using topographical model.

In this paper described are the Large Boundary

Layer Wind Tunnel and the wind tunnel studies

for the Akashi Kaikyo Bridge and the Tatara

Bridge conducted at the Large Boundary Layer

Wind Tunnel.

2. LARGE BOUNDARY LAYER WIND

TUNNEL [1]

2.1 REQUIRED DIMENSIONS AND

PERFORMANCE OF THE WIND TUNNEL

At first, the model scale of 1/100 was decided so

that it might simulate the characteristics of the

bridge and that its scale might become as small as

399
possible. Then the width 41m and height 4m of

the test section were decided to put whole the

model in it. The length of the test section 30m

were selected in order to generate boundary layer

turbulent flow.

The maximum wind speed of 12 m/s was required

considering the Froude number similarity and

energy loss due to turbulence generation.

The required properties of smooth flow were as follows:

a) Turbulence Intensity ; less than 1 %

b) Mean Wind Speed Distribution ; less

than 3 %

2.2 PILOT WIND TUNNEL TEST

In order to determine the dimensions of the wind

tunnel and the building, a pilot wind tunnel test was carried out. Its scale was 1/16. The pilot wind

tunnel is shown in Photo 3. The main findings are as follows:

1) Specifications of the contraction nozzle ; a) Contraction ratio 2 could satisfy the

required flow properties.

b) Two steel screens were required.

c) One of steel screens should be installed at the entrance of the bell-mouth.

2) Building

;

a) The minimum length of 75 m was required.

b) The minimum height of 19.5 m was

required.

3) Others

;

a) Corner-vanes at the entrance were very

effective. However, those at the exit were

not. b) The flow was little affected by the inside

trusses supporting the ceiling of building.

c) The flow was little muddled by the

streamline-section posts of corner-vanes.

2.3 FINAL PLAN OF THE WIND TUNNEL

The general layout of the wind tunnel facility

selected as a final plan is shown in Fig.3. The

following items were determined considering the

usage:

a) Three test sections were prepared: upstream

section for smooth flow test, downstream

section for turbulent flow test and center

section for test of the model with yaw

angle.

b) Sensor rooms were arranged at both sides

of the test section.

c) Size and number of fans were selected as 1.8m in diameter and 36, respectively.

d) Wind speeds should be controlled by

rotation cycles of fans.

2.4 PERFORMANCE OF THE WIND

TUNNEL

Properties of a smooth flow in this wind tunnel were measured as follows:

a) Turbulence intensity; less than 0.5 %

b) Variation in mean speed; less than 3 %

Consequently, the required properties such as low

turbulence and uniformity of the flow were

sufficiently satisfied.

3. WIND TUNNEL STUDY FOR THE

AKASHI KAIKYO BRIDGE [1]

3.1 AEROELASTIC FULL MODEL OF THE

AKASHI KAIKYO BRIDGE

1) Modeling of Stiffening Truss Girder

In modeling of the stiffening truss girder,

attention was paid to realizing the similarity of the

stiffness without providing stiffening spine at the

center of cross section. Two types of truss models were studied. One was the rigid block model

connected by springs, and the other was the

wholly elastic model in which four main chords were used as stiffening bars. As for the latter, it was found difficult to support floor decks at the

accurate positions. Since it was known that the

position of floor decks affects flutter

characteristics of the bridge very much, the rigid

block model type was selected.

In this type of model, each rigid block was

composed of every four truss panels to simulate

the geometrical shape of the actual truss girder,

and springs installed between blocks to represent

the mechanical characteristics. Based on the

preliminary study, V-shape springs were selected.

The model is shown in Fig.4.

The effects of springs and clearances between

the blocks on the flutter characteristics were

studied through sectional model test, and their

effects were found to be small . The blocks were

designed as rigid as required by using light new

material of carbon fiber plate for the truss members. The detailed modeling of the truss girder was one of reasons why geometrical scale

400
of 1/100 was adopted.

2) Modeling of Tower

The model of tower was composed of a frame of

stiffening bars and exterior wooden forms by

which the geometrical shape of actual structure was simulated. The bending stiffness in the plane

of bridge and torsional stiffness about tower axis were simulated by choosing an appropriate

section of stiffening bars. The bending stiffness in

the plane of tower was simulated by replacing the

effect of actual diagonal members to that of lateral members, namely, by cutting at their connections.

The axial stiffness of tower columns, however,

could not be simulated correctly in this method,

and it resulted small error in torsional

displacement of the girder.

Fig.2 shows a frame of stiffening bars of the

tower model. The level of similarity was checked

by analyzing natural frequencies and static

deflection of the total bridge structure. The results were almost satisfactory.

3) Modeling of Cables

As for the modeling of the cables, the mass, drag

force and axial stiffness should be simulated,

considering the effect of Reynolds number on

drag force. Therefore, model cable was composed

of a uniform wire and weight pieces. The

uniform wire was used for similarity of axial

stiffness, and the weight pieces for similarity of mass and drag force. The dimension and interval

of weight pieces were determined from drag force

measured in wind tunnel test.

Aerodynamic interaction between stiffening truss and cables near the center of main span was also

considered. From the section model test results,

it was determined to attach geometrically

simulated weight pieces near the center of the

main span. Fig.2 shows the modeling of cables.

4) Connections

Connections of each element such as end links,

tower links, cable anchors, tower bases etc. were

carefully designed so as to realize the same

functions as those of the actual bridge and to keep

structural damping as small as possible.

5) Structural Properties of the Model

The completed full model of the Akashi Kaikyo

Bridge is shown in Photo 4. Natural frequencies

measured at dynamic tests of the model agreed

well with the required values, which meant that

the model was fabricated with good accuracy.

Structural damping of the model was small

enough.

3.2 EXPERIMENTS IN SMOOTH FLOW

1) Static Torsional Displacement

Relation between wind velocity and static

torsional displacement at the mid-point of center span is shown in Fig.5. At wind velocity of 6 m/s,

that is equivalent to design wind speed of 60 m/s

for the prototype bridge, torsional displacement of

-2 degrees was observed. Although the observed

torsional displacement was a little larger than the

analytical value due to such reasons as error in simulation of longitudinal stiffness of towers, this remarkable displacement seems to be one of

distinctive features of super long-span bridges.

In addition, it was found that major cause of

torsional displacement was not the aerodynamic

moment but torque due to tension of inclined

suspending rope caused by the aerodynamic drag

force acting to the stiffening truss.

2) Flutter Characteristics

a) Damping

Relation between wind speed and logarithmic

decrement is shown in Fig.6. The damping of

torsional mode decreased drastically when wind

speed exceeded 7 m/s. Wind speed of 8.5 m/s (85

m/s for prototype) was flutter onset speed. Here,

the vibrational mode was not pure torsion, but

coupled with vertical bending mode. In case of the

section model test of the attack angle of 0 degree,

damping decreased from wind speed of 2 m/s, and

flutter occurred at 6 m/s.

b) Frequency

As is shown in Fig.7, torsional frequency

decreased as wind speed became higher, and at onset wind speed the frequency became 90% of

natural frequency. On the other hand, frequencies

of bending mode scarcely decreased. This

tendency of frequencies was almost same as that of section model tests.

c) Rotation Center

Fig.8 illustrates the position of rotation center measured along bridge axis. The rotation center

lay on the windward side at the midspan, on the

401
leeward side at quarter point of center span, and on the windward side again at the middle of side

spans. Flutter observed in the section model test was also coupled one in which the rotation center

lay on the windward side, and the behavior was

almost similar to that of the midspan of the full model.

d) Vibrational Modes

Torsional and vertical bending response at every

1/8 cycle are illustrated in Fig.9. As for the

torsion, flutter mode was almost same as the first symmetric one in windless condition. As for the

vertical bending, however, flutter mode did not

agree with any single mode in windless condition,

but very complicated.

3) Distinctive Feature of Flutter of Super LongSpan Bridges

In case of ordinary suspension bridges, which

have center span of about 1,000 m, static torsional

displacement is negligibly small. It was found in

the full aeroelastic model test of the Akashi

Kaikyo Bridge, however, that relative angle of

attack cased by the static displacement was not

negligibly small, and that it varied along the

bridge axis. Since angle of attack affects flutter

characteristics, the effects of torsional

displacement on the aerodynamic stability must be

considered in case of super long-span bridges,

which have center span of 2,000 m or over.

In addition, the flutter of the Akashi Kaikyo

Bridge was not stall flutter but coupled flutter,

and the vertical bending mode during flutter was

not single natural mode but complicated one. Therefore it does not seem that aerodynamic

stability of super long-span suspension bridges can be predicted directly from spring-mounted

rigid model test.

4) Comparative Flutter Analysis

Since spring-mounted rigid model could not be

compared directly with the present full model test,

flutter analysis was conducted considering the

effect of the static torsional displacement,

including higher natural modes as well as the first natural mode, and using measured aerodynamic

derivatives.

At first, the following aerodynamic derivatives were considered:

Drag due to Along-wind motion;

quasi-steady theory

Lift due to Heaving motion;

measured

Lift due to Torsional motion;

measured

Pitching moment due to Heaving motion;

measured

Pitching moment due to Torsional motion;

measured

The results are shown in Fig.6 and 7. The

agreement with full model test was not good.

Since the change of drag with angle of attack was

not small at the relative angle of attack caused by

static torsional displacement near the flutter speed, it was thought that the other aerodynamic

derivatives should be included in the flutter

analysis.

In the second analysis, the following aerodynamic

derivatives were considered as well as the above

ones:Drag due to Heaving motion;

measured

Drag due to Torsional motion;

measured

Lift due to Along-wind motion;

quasi-steady theory

Pitching moment due to Along-wind motion;

quasi-steady theory

The results are shown in the same figures. The

agreement was fairly well. From additional

analysis, it was found that drag due to heaving

motion and torsional motion was more effective

than lift and pitching moment due to along-wind

motion.

3.3 EXPERIMENTS IN TURBULENT FLOW

1) Test Condition

The same aeroelastic full model as used in smooth

flow test was used.

The turbulent flow was generated by spires and

roughness. In the simulation of turbulent flow, it was thought that turbulent intensity and integral

scale of turbulence were the most important

parameters to be similar to natural wind. The

intensity of turbulence Iu for main wind direction

at the deck height was assumed to be 10%.

Several spires and some arrangements of

roughness blocks were tested, and the most

suitable spires and roughness blocks were

selected.

402
The main properties of the turbulent flow were as follows:

power extent of wind speed profile; 1/5.1

intensity of turbulence; Iu=9.6% Iw=6.8%

integral scale of turbulence;

Lxu=l.lm, Lxw=0.4m

Although the power extent of vertical distribution

of mean wind speed was a little larger and the

integral scale of turbulence was a little smaller,

the properties of the simulated turbulent flow were thought to be fairly satisfactory.

2) Test Results

It seems obvious that the random vibrations

observed at relatively low wind speeds were

caused by approaching turbulence. As wind speed

increases, however, characteristics of the vibration

changes. The power near the torsional 1st mode

becomes dominant at higher wind speeds. The

increase rate of the vibrational amplitude with

wind speed becomes larger at higher wind speed.

It can be seen clearly in Fig. 10, where

displacement and wind speeds were plotted in

logarithmic scales. These changes in vibrational

characteristics seem to be caused by negative

aerodynamic damping, which result in flutter. The

apparent damping of the model in wind, which is sum of structural damping and aerodynamic

damping, was estimated by random decrement

method [2], and is shown in Fig. 11. It is clear that

the apparent damping becomes small at high wind

speeds.

3) Effect of Turbulence on Flutter

The vibrations of the bridge model at high wind

speed in turbulent flow were caused by both gust

and negative aerodynamic damping. It can be said

that the observed vibrations are mixture of gust

response and flutter. It is difficult to separate

vibrational amplitude caused by flutter from that

caused by gust. Therefore, the effect of turbulence on flutter was investigated in viewpoint of

damping rather than vibrational amplitude.

The relationship between wind speed and

apparent damping of torsional vibration in

turbulent flow is shown in Fig. 11, compared with

that in smooth flow. Since the difference in

apparent damping between in turbulent flow and

in smooth flow is small, it can be said that the

effect of turbulence on flutter is small in case of

the Akashi Kaikyo Bridge.

4) Comparison between Calculated Gust

Responses and Observed Ones

a) Calculation Method

For comparison, gust responses of horizontal

bending, vertical bending and torsion were

calculated according to the ordinary methods in

frequency domain [3]. In the calculation of

torsional gust responses, the effect of horizontal

bending modes was considered.

As for wind and structural properties, the

observed data were used. As for coefficients of

drag, lift and aerodynamic moment, the data

measured in sectional model test in smooth flow were used. Aerodynamic forces caused by gust were calculated according to quasi-steady theory,

corrected by aerodynamic admittance and spatial

correlation as follows:

i) aerodynamic admittance for drag [3]

Xu2(fd)=2[kfd-l+exp(-kfd)]/(kfd)2

ii) aerodynamic admittance for lift and moment

[4]

Xw2(fb)=(a+7Tfb)

/[a+( n a+ 1) n fb+2 7t ( 7t fb)2]

iii) spatial correlation [5]

R(fb)=exp(-kfb I xl-x2 I /B)

where,

a : constant (=0.1811)

B : width of stiffening truss D : depth of stiffening truss

f : frequency

fb=fB/U

fd=fD/U

k : decay factor (assumed to be 8)

U : mean wind speed

xl, x2 : coordinate along bridge axis

Aerodynamic damping for horizontal and vertical

bending vibration was calculated according to quasi-steady theory. As for torsion, aerodynamic

damping was assumed to be zero.

b) Comparison

The gust responses calculated at the design wind

speed are shown in Table 1, compared with the

responses observed in experiment. The agreement

in vertical bending and in torsion was fairly good,

however, the observed horizontal bending

responses were much smaller than calculated

ones.

One of the causes of this discrepancy is thought to be the spatial correlation which was assumed as

403
the exponential function of fb I xl-x2 I /B. Since

the measured spatial correlation of wind speed

(root co-coherence function, Fig. 12) did not tend

to unity as frequency became 0 when separation

of measurement points were large, the calculation

might lead to an overestimation as was pointed

out in ref.[6] and [7]. Another possible cause of

the discrepancy is thought to be aerodynamic

admittance.

c) Improved Calculation

The aerodynamic admittance for this stiffening

truss was measured. Using the measured

aerodynamic admittance and the spatial

correlation based on the turbulent flow of the

wind tunnel, gust responses were calculated again.

The result is shown in Table-1, which agrees with

experimantal result better than the previous

calculation.

4. STUDY FOR THE TATARA BRIDGE

4.1 EXPERIMENT IN SMOOTH FLOW [8]

1) Aeroelastic Full Model of the Tatara Bridge

The model was fabricated with a geometric scale

of 1/70. Except for the cables, the model was

designed by well known methods, i.e. steel frame

was designed to achieve proper stiffness,

additional mass was installed to simulate mass

and polar moment of inertia, and the frame and

additional mass were covered with exterior

wooden forms to simulate geometrical shape.

Froude number similitude was adopted to the

design of the model. This means wind speed

ratio was 8.4

.

Attention was paid to the design of the cables.

As well as the number of the cable, mass and

tensile force of each cable was simulated. The

drag force acting on each cable was also similar to the prototype bridge. Piano wires were used to simulate axial stiffness. In order to give proper mass to each cable, cylindrical mass pieces were

placed at a certain interval. The mass pieces were so designed that the cables might have

proper mass and drag force. For the similarity of

drag force, the difference in Reynolds number

were taken into account. Similar method was

used for the design of cables in the model of

Akashi Kaikyo Bridge. Tensile force in each

cable was measured from its natural frequency,

and adjusted by turning turnbuckle installed at the

end of cable.

During the preliminary wind tunnel test, it was

found that the stay cables of the model vibrated in

wind. At certain wind speed, the vibrational

amplitude became so large that stay cables might

touch each other. In the real bridge, the windinduced vibration of stay cables will be eliminated

by installing dampers or by some other methods.

Therefore the vibration of each cable in the model

was eliminated by installing thin extra cables.

Through preliminary wind tunnel study, it was

found that the thin extra cables would not affect

aerodynamic characteristics of the model so

significantly.

Photo-5 shows the model of the Tatara bridge in

the Large Boundary Layer Wind Tunnel. The

structural characteristis of the model were as follows.

a) Static characteristics

Static loading tests were carried out in order to check the static characteristic of the model. Either

in holizontal, vertical and torsional loading, the

results showed good agreement with calculated

results.

b) Dynamic characteristics

Free vibration tests in windless condition were

carried out to check the dynamic characteristics of

the model. It was confirmed that the model

represented the prototype bridge well. The

damping observed was a little larger than the

target value of 0.02.

2) Wind-Induced Vibration

The followings are results of the wind tunnel tests

in smooth flow with the angle of attack 0 deg.

a)Vibrational amplitude

No divergent vibration such as flutter was

observed up to the reference wind speed of flutter. When the reduced wind speed was about 1.0,

small vertical bending vortex-induced vibration

was observed. The vibrational mode was 1st

symmetric one. Usually, when 1st symmetric

vortex-induced vibration is observed, vortexinduced vibration of higher modes are expected in

higher wind speeds. However, in this case, no

higher modes vibration was observed.

b) Change of damping

The decrease of damping at the reduced wind

404
speed of 1.0 is coincident with the appearance of

vortex-induced vibration. At higher wind

speed, the damping increased with the wind

speed.

c)Change of frequency

The frequency of torsion, vertical and horizontal

bending scarcely changed.

4.2 EXPERIMENT WITH ISLAND MODEL

[9]

1) Wind Characteristics at the Construction Site

In order to clarify wind characteristics such as mean wind speed, wind direction, attack angle,

and turbulence intensity, an anemometer was

installed at the construction site of the Tatara

Bridge. Wind tunnel test was also carried out

using 1/2000 scale topographical model of the

terrain around construction site (See Fig. 13). The

measured data at the site was used in order to

calibrate the topographical wind tunnel test (See

Fig. 14).

Fig. 15 shows the wind flow pattern measured in

the wind tunnel test. From the figure, it can be

seen that wind characteristics at the site are very

complicated. For example, in the case of wind

direction of NNW and SSE, a topographical effect

is hardly recognizable. In other words, wind speed

and turbulence intensity were scarcely changed by

the topography. On the contrary, in the case of

wind direction such as NE and W, wind

characteristics were changed significantly by the

topography. Particularly, in the case of NE

direction, the turbulence intensity was

considerably increased by the topography, but the

wind speed was not remarkably decreased. In the

design of the long-span bridge, these wind

characteristics were thought to be critical. Therefore, it was thought that the topographical

effects on the bridge must be studied in the wind

tunnel test with bridge model and topographical

model.

2) Aeroelastic Full Model of the Tatara Bridge

The scale ratio of the model was determined as to 1/200 considering the size of the wind-tunnel,-

integral scale of turbulence and topographical

model to be represented inside the wind tunnel.

Froude number similitude was not adopted to the

design of the model, because the wind speed ratio

14 according to Froude number similitude seemed

too large to measure wind-induced vibration

accurately. The model was fabricated a little

stiffer than the Froude number similitude

required. The wind speed ratio was about 8.

The model was designed by well known methods,

i.e. steel frame was designed to achieve proper

stiffness, additional mass was installed to simulate mass and polar moment of inertia, and the frame

and additional mass were covered with exterior

wooden forms to simulate geometrical shape.

Each three cables of the real bridge was

represented by one piano wire in the model. Mass

of the cables were equally distributed to the tower

and the girder.

Photo 6 show the model of the Tatara bridge in

the Large Boundary Layer Wind Tunnel. Free

vibration tests in windless condition were carried

out to check the dynamic characteristics of the

model. The damping of vertical bending mode

was a little larger than the target value of 0.02,

and the damping of lateral bending mode was a

little smaller.

3) Wind Tunnel Experiments of the Bridge and

Island

Wind tunnel experiments were carried out for

three cases of wind direction, namely 18, 36 and

180 degrees. Wind direction was defined

clockwise from the north direction normal to bridge axis. Three directions were chosen because

they were thought to be critical from the design

point of view. That is, the mean wind speed for 18

and 36 degrees was not noticeably decreased,

while turbulence intensity was considerably

increased. The wind direction of 180 degrees,

normal to the bridge axis, was the prevailing wind

at the site. The increase in turbulence intensity

due to Hakata Island located in the south of the

bridge site was expected to be small.

Wind tunnel experiments were conducted both

with and without topographical models. In both

cases, approaching flow had slight turbulence

generated by spires and roughness blocks.

Fig. 16 shows the rms value of the lateral gust

response of the girder at the design wind speed.

The experimental results with the topographical

models are compared with those without

topographical models in these figures. In the case of 18 and 180 degrees, the topographical effect is hardly recognizable because there is little difference in the rms values. On the other hand, in

the case of 36 degrees, the topography increased

405
the gust response significantly.

4) Discussion

According to the quasi-steady theory,

aerodynamic force due to approaching turbulence

increases with square of mean wind speed and

turbulence intensity. Fig. 17 shows the wind

characteristics along the bridge axis in the three

wind directions. In the case of 36 degrees, a

significant increase in turbulence intensity along

the bridge axis is recognized. However, the

increase in turbulence intensity is associated with

decrease in mean wind speed.

Another factor that influences the gust response of

long-span bridges is correlation of fluctuating

wind speed at different places. Table 2 shows

turbulence intensity and integral scale of

turbulence at the mid-point of the center span in case of 36 degree. It can be seen that the integral

scale as well as intensity was increased very much

by the topographical model. Fig. 18 shows co- coherence of fluctuating wind speed at two places on the bridge axis. It can be seen that the

correlation is much higher in case of 36 degree

than in case of 180 degree.

From these measurements, it seems that the large

gust response in case of wind direction 36 degree

was caused by high turbulence intensity and high

space-wise correlation of the turbulent flow

generated by the topographical model.

5. CONCLUDING REMARKS

1) The Large Boundary Layer Wind Tunnel,

whose test section is 41m wide, 4m high and 30m

long, was designed and constructed based on

test results of 1/16 pilot wind tunnel. The low

turbulence and uniformity of wind speed of the

smooth flow were satisfactory.

2) Wind Tunnel Study for the Akashi Kaikyo

Bridge

a) The 1/100 aeroelastic full model of the Akashi

Kaikyo Bridge was designed so that the similarity

of shape, mass distribution and stiffness

distribution might be satisfied. The stiffening truss was modeled without stiffening spine at the center

of cross section. The bending stiffness in the

plane of tower was simulated as well as the

bending stiffness in the plane of bridge and

torsional stiffness about the axis of tower. In modeling of the cables, mass, drag force and axial

stiffness were simulated, and aerodynamic

interaction with stiffening truss was considered.

b) In the smooth flow test, remarkable static

torsional displacement was observed. It was

caused mainly by aerodynamic drag force acting

to the stiffening truss. Coupled flutter was

observed at the wind speed of 8.5m/s (85m/s for

real bridge), and its vertical bending vibrational

mode was not similar to any of natural mode,

while its torsional vibrational mode was similar to the first symmetric natural mode. Therefore it does not seem that aerodynamic stability of super

long-span suspension bridges can be predicted

directly from spring-mounted rigid model test.

c) In order to predict flutter of super long-span

bridges by analytical way, the effect of the static

torsional displacement should be considered, and

higher natural modes as well as the first natural

mode should be included. Besides, aerodynamic

derivatives such as Drag due to Heaving motion,

Drag due to Torsional motion, Lift due to Alongwind motion, and Pitching moment due to Alongwind motion should be included in addition to the

conventional aerodynamic derivatives.

d) In the turbulent flow test, gust responses were

observed in horizontal bending, vertical bending

and torsional mode. As wind speed increased, the

vibration of torsional 1st mode became dominant,

and apparent damping became smaller. This seems to be the effect of flutter. Apparent

damping in the turbulent flow was compared with

that in the smooth flow. It was found that the

effect of turbulence on flutter was small for this

bridge model.

e) The observed gust responses were compared

with the calculated ones. As for vertical bending

and torsional responses, the agreement was fairly

good, however, the observed horizontal bending

responses were much smaller than calculated ones. The more accurate aerodynamic admittance

or spatial correlation should be measured and

applied to the calculation of gust responses.

3) Wind Tunnel Study for the Tatara Bridge

a) The aerodynamic stability of the Tatara bridge

was investigated in smooth flow. No divergent

vibration such as flutter was observed. This

means the Tatara bridge is aerodynamically

stable. The only wind induced vibration

observed was vortex-induced vibration with small

amplitude.

b) The gust response of the Tatara bridge was

406
investigated through wind tunnel test using 1/200

full aeroelastic model in turbulent flow generated

by the model of surrounding islands. When wind

direction was 18 and 180 degrees clockwise from

the north direction normal to bridge axis, the

topographical effect was hardly recognizable. On

the other hand, in the case of 36 degrees, the

topography increased the gust response

significantly.

c) The main reasons for the large gust response

measured in case of wind direction 36 degree

seem to be high turbulence intensity and high

space-wise correlation caused by the

topographical model.

[8] Sato H., Toriumi R., Kitagawa M. and

Katsuchi H., Wind tunnel experiment for a super

long-span cable-stayed bridge in smooth flow,

Proc. of 26th Joint Panel Meeting on Wind and

Seismic Effects, UJNR, 1994.

[9] Sato H., Kitagawa M. Toriumi R., and

Katsuchi H., Wind-induced vibrations of a super

long-span cable-stayed bridge in turbulent flow

generated by surrounding islands, Proc. of 27th

Joint Panel Meeting on Wind and Seismic Effects,

UJNR, 1995.

ACKNOWLEDGEMENT

The authors would like to express their thanks to

all of members of the study committee for wind

resistant design of the Honshu-Shikoku bridges

and its subcommittee for wind tunnel works.

Discussions at the committees were essential part

of this paper.

REFERENCES

[1] Sumiyoshi Y., Endo T., Miyata T., Sato H.

and Kitagawa M., Experiments for the Akashi

Kaikyo Bridge in a Large Boundary Layer Wind

Tunnel, Proc. of International Seminar on

Utilization of Large Boundary Layer Wind

Tunnel, Tsukuba, Japan, 1993.

[2] Technical Committee on the Wind Resistant

Design of the Honshu-Shikoku Bridge, Final

reports on the observation of large scale bridge

model (Part 2), JSCE, 1978 (in Japanese)

[3] Davenport A.G., Buffeting of a suspension

bridge by storm winds, Proc. ASCE, Vol. 88,

ST3, 1962

[4] Fung, Y.C., An Introduction to the Theory of

Aeroelasticity, John Wiley & Sons, 1955

[5] Davenport A.G., The spectrum of horizontal

gustiness near the ground in high winds, Q.J. Roy.

Met. Soc, Vol. 87, 1961

[6] ESDU, Characteristics of atmospheric

turbulence near the ground, Part III : variations in

space and time for strong winds (neutral

atmosphere), 1975

[7] Irwin H.P.A.H., Wind tunnel and analytical

investigations of the response of Lions' Gate

Bridge to a turbulent wind, Laboratory Technical

Report LTR-LA-210, National Research Council

Canada, 1977

407
Table 1 Gust Responses

Horizontal

(mm)

Vertical

(mm)

Torsional

(deg)

Measured 20 12 1.4

Calculated

(Conventional

calculation)

83 13 2.9

Calculated

(Improved

calculation)

38 11 1.5

Table 2 Turbulence Intensity and Integral Scale of Turbulence

Wind Characterinstics

(Center of center span)

with

topographical model

0 = 3

6°

without

topographical model

j3 = 3 6°

Turbulence Intensity

Iu (%) 15.6 9.3

Iw (%) 8.7 5.8

Integral Scale

of Turbulence

Lx u (m) 1.21(242) 0.58(116)

Lx w (m) 0.25(50) 0.31(62)

( ):converted real scale

408
KOBE

960

3 910

1 990 960 AWAII

ISLAND

(A) GENERAL VIEW

23

35.50

10.75 . 10.75 2.50

ROAD LEVEL

05

MAIN TOWER

STIFFENING GIRDER

(B) MAIN TOWER AND STIFFENING GIRDER

(unit:m)

Fig. 1 Akashi-Kaikyo Bridge

409
fc— oo

0 -QZZ

E—

C_> toCO

CO

coo

CJ

es

aso

•i-H

H

(3JD E

410
OOS'61

411
116

Cutting

Point

Cable Model (General Part)

282

1

r—

1

0J

Jro

j

284

Cable Model (Near the Center of Main Span)

Tower Model

( Unit : mm)

Fig. 4 Full model of the Akashi-Kaikyo Bridge

412
Wind Speed (m/sj

Fig. 5 Torsional displacement by wind (midspan)

413
V) w > >

( zh ) /Couanbajj

i }uaui9Joap otuiij^t jb3ot

414
415
toVD

0>o

00

o

(0o

10 1

10 (

10

-1

L/2 point at Mid-span

O

o

A

Average Displacement

RMS

Maximum Displacement

Maximum Vibrational Amplitude

-7

uA

3*

Jt

10

2 5

Wind Speed (m/s)

8 10

Fig. 10 Displacement in turbulent flow (logarithmic scales)

0. 20

0. 15

• 0.10

o

<u

o

0. 05

I 0.(

-0. 05

-0. 10

i i

:Turbulent Flow

A .'Smooth Flow

(Torsional Component)

i

A

^ —

* A A A < A

A

8 10

Fig. 11

lind Speed (m/s)

Damping of model (turbulent flow, smooth flow)

416
1.0

0.5

0

Separat ion of

Measurement Points

O : 2 0 mm

A 40 mm

80 mm

o 1 5 0 mm

V : 300 mm

X : 500 mm

X x

vx

XXxxxxXxxxxx><XXxXXx

xxxxXxxxxx><xxxXXn^xxxx

(Horizontal Component, Midspan)

0 0.02 0.04 0.06 0.08 0.1

Reduced Frequency

Fig. 12 Spectral correlation of wind speed

417
Region of topographical

modeling 0 1U lit 3L

Fig. 13 Location of the Tatara Bridge and

region of topographical modeling

c —

» 3

1 30H

: Wind tunnel test x : Field Measurement

—i

N

Wind direction

Fig. 14 Turbulent intensity at the site of the Tatara Bridge

418
419
Yawed angle: 18 deg.

>

CO

1.6m r

0.8m

1A 2P

Lateral response

• • •

O : with topography

# : without topography

3P 4P

Yawed angle: 36 deg.

*i l-6m

>

S3 0.8m

o

1A 2P

Lateral response

o

o o

• • •

O : with topography

# : without topography

3P 4P

Yawed angle: 180 deg.

1.6m

co 0.8m

1A 2P

Lateral response

O : with topography

# : without topography

3P 4P

Fig. 16 RMS value of girder's gust response

420
Yawed angle: 18 deg.

oD

\

D

1A

40 r

2P

- 20H

Ratio of wind speed

between aproach flow(Uo)

and girder's height(U)

3P 4P

Turbulence intensity. Iu

• • •

1A 2P 3P 4P

Yawed angle: 36 deg.

1

-

o\

Ratio of wind speed

between aproach flow(Uo)

and girder's height(U)

• • • • •

o

1—

1A 2P 3P 4P

40 r

- 20

Turbulence intensity: Iu

• • • • •

1A 2P 3P 4P

Yawed angle: 180 deg.

o\

a

Ratio of wind speed

between aproach flow(Uo)

and girder's height(U) • • ••• • • •

1A

40 r

2P

p

20

3P 4P

Turbulence intensity: Iu

1A 2P 3P 4P

Fig. 17 Wind characteristics along bridge axis

421
© v> © v>

<—i o o 9DU9J91100-00

o

o v> o v» - d <? 90U9J9qO0-O0

O «1 ° "»?

90U9J9qO0-O0

422
Photo 1 Akashi-Kaikyo Bridge

423
Photo 3 Pilot wind tunnel test

Photo 4 Full model test of the Akashi-Kaikyo Bridge

424
Photo 5 Model of the Tatara Bridge for the wind tunnel study

in smooth flow

Photo 6 Model of the Tatara Bridge for the wind tunnel study

in turbulent flow

425

A SEISMIC RETROFITTING METHOD FOR REINFORCED CONCRETE

BRIDGE PIERS AND THEIR ANALYSIS

by

Hisanori Otuska*, Jun-ichi Hoshikuma**, and Kazuhiro Nagaya***

ABSTRACT

A new steel-jacketing strengthening method for

enhanced flexural performance of existing reinforce

concrete bridge piers was proposed in this paper. A

series of large-scale specimens were cyclic-loaded,

to examine the strengthening effect. Test resulsts

showed that the strengthening method was effective

in enhancing both the flexural strength and the

ductility. Also, effects of steel-jacket details were

studied based on the loadig tests. Furthermore, an

analytical model for the steel-jacketing reinforced

concrete piers was proposed. The flexural strength

and the ductility analyzed by the model showed

good agreements with the test results.

KEYWORDS: seismic retrofitting, steel jacketing,

reinforced concrete bridge piers, cyclic loading test,

flexural strength, ductility

1. INTRODUCTION

When the Hyogo-ken Nambu Earthquake hit the

Hanshin and Awaji districts in the early morning of

January 17, 1995, civil engineering structures,

including highway bridges, were severely damaged

that had never been seen in previous earthquakes.

With regard to damage to highway bridges, it was

noted that the piers of some bridges that had been

designed in accordance with the 1964 or 1971

seismic specifications suffered severe damage; the

provisions in those standards for the cut-off of

longitudinal reinforcement required for piers were

revised in 1980(1). The reason for the damage was

that the quake's ground motion had properties that

produced greater effects on highway bridges than

had previously been observed. It has been estimated

that severely damaged bridges received seismic

forces larger than had been assumed in their design

(1).

To make a bridge sufficiently resistant to earthquakes such as this one, its structure has to be

made tougher by increasing the strength and

ductility of the bridge as a total system (2). This

concept of seismic design for a bridge also applies

to the seismic retrofitting of existing bridges: it is necessary to adopt a retrofitting method that can be

expected to enhance both the flexural strength and

the ductility of the piers. However, when the

flexural strength of piers is enhanced, the seismic

forces propagated into the foundation are increased

all the more. Therefore, if the calculated strength of

the foundation of a bridge is smaller than the

expected of its retrofitted piers, it is necessary to

carry out more wide-ranging retrofitting of the

bridge by retrofitting its foundation as well. On the

other hand, if the ductility of the piers alone is enhanced, while their flexural strength is left as it is,

this may lead to substantial plastic deformation

during a major earthquake, which would cause a

residual deformation that could not be repaired (3).

It is therefore important to work out a retrofitting

design that is capable of enhancing the flexural

strength of the piers as required, within the range

that can be supported by the foundation, and to make this enhancement match the improvement in

the ductility.

Head, Earthquake Engineering Division,

Earthquake Disaster Prevention Department,

Public Works Research Institute, Ministry of

Construction

Research Engineer, ditto

Assistant Research Engineer, ditto

427
This paper proposes the steel jacketing method, with

a controlled increase in flexural strength allowing

well-balanced enhancement of both the flexural

strength and the ductility of each pier in a reinforced

concrete bridge. It also describes the results of a

cyclic loading test using large-scale models

constructed in order to examine the effects of

enhancing the strength and ductility of bridge piers

using this retrofitting method. Furthermore, the

analytical method considering the confinement of

concrete is checked the validness by comparing the

test results.

2. SEISMIC RETROFITTING OF REINFORCED

CONCRETE PIERS USING A STEEL JACKET

WITH ANCHOR BARS

The steel jacket method, providing a controlled

increase in flexural strength, involves steel plates

around the reinforced concrete bridge piers, and

bonding the steel jacket to the pier with epoxy resin or non-shrinkage mortar, while anchoring the jacket

to the footing with anchor bars, as shown in Figure

1. This retrofitting method mainly aims to compensate for the inadequate cut-off of the

longitudinal reinforcement and, at the same time, to enhance the flexural strength and ductility of the

pier. The structure has several important

characteristics, including the anchoring of the steel

jacket to the footing with anchor bars, the vertical

gap between the lower end of steel jacket and the

top of the footing, and the use of H-beams fastened

at the lower end of the jacket. These characteristics

are described below.

First, the anchor bars play a part in enhancing the

flexural strength of the pier, as well as in controlling

the seismic forces propagated into the footing. This

is because the increase in the flexural strength of the

pier resulting from the retrofitting method can be

controlled by adjusting the number and diameter of

the anchor bars installed. The vertical gap between

the lower end of the steel jacket and the top of the

footing is provided because it is expected that, when

it receives a large seismic force, the steel jacket may

come into contact with the footing, which may then

buckle under the unexpectedly large compressive

force and receive damage. However, when the

length of the vertical gap is too large, the confining

effect of the jacket at the bottom of pier is reduced.

The length should therefore be as small as possible.

Piers with a rectangular section also have H-beams

installed around them at the lower end of the jacket.

The reason for this is that, when the base of the pier

suffers from a deformation with a large curvature,

the jacket at the plastic hinge zone swells out, and

the confining effect of the jacket is liable to be lost. When the lower end of the jacket is attached with

the H-beams, however, then even under large

deformations it still functions as a lateral reinforcement.

3. CYCLIC LOADING TEST FOR MODELS OF

RETROFITTED REINFORCED CONCRETE

PIERS

3-1. Specimens

The specimens used are shown in Figure 2. Their

section was a square of 60 cm x 60 cm. The height

from the base to the loading point was 3.01 m. The

shear span ratio was 5.0. As longitudinal

reinforcements, SD295/D10 bars were arranged, and

the ratio of longitudinal reinforcement was 1.58%

(the ratio of tensile reinforcement was 0.55%). As

hoop reinforcements, SD295/D6 hoops were

arranged at intervals of 20 cm. In the specimens

used in the test, the longitudinal reinforcement was

not cut off at mid-height. This was because previous

studies (4) had found that the steel jacket method

can prevent damage to the cut-off sections of

longitudinal reinforcement, thus it was decided that

effect of this method in retrofitting the cut-off

sections was not to be examined in this test.

When the steel jacket was applied to the specimen

to be retrofitted, two steel plates with a thickness of

1.6 mm (SS400) were each bent into a channel

-shape, and then welded into a jacket. Finally, the

jacket was slipped from above over the specimen.

As described above, however, a vertical gap of 5-15

cm was set between the lower end of steel jacket

and the top of the footing (see Photo 1). In addition,

H-beams (100 x 100 x 6 x 8, SS400) were used to

strengthen the lower end of the jacket. Epoxy resin was poured in between the reinforced concrete pier

and the steel jacket, with a thickness of 4 mm, to bond them.

The numbers and diameters of the anchor bars were

set in such a way that the flexural strength of the

428
specimen could increase by approximately 30%. As

described above, with this retrofitting method, any

increase in the strength of the pier has to be set so

that its resulting strength will not exceed that of the

foundation. In this test, it was arbitrarily decided that

the pier strength should be increased by about 30%.

It was found that, using the calculation method

described later, when anchor bars equivalent to about 60% of the total strength of the steel jacket were used, the pier strength increased by

approximately 30%. Consequently, it was decided

that twenty anchor bars, made by processing SD345

/D13 reinforcements into an Ml2 screw bolt

(effective section area: 0.843 cm

2 ), should be

arranged at intervals of 12 cm. The bars were

anchored with resin anchors 55.6 mm outside the

external edge of the concrete. The upper ends of die

anchor bars were fastened to the webs of the

H-beams by double nuts, as shown in Photo 2. In

this structure, the axial force produced in the anchor

bars must be propagated smoothly via the H-beams

to the steel jacket. In the specimens for the test,

therefore, the H-beams and the steel jacket were

joined with six rib plates (thickness: 6 mm) per side,

as shown in Figure 2b and Photo 1 . The anchor bars were anchored to the footing for a length of 65 cm.

In order to examine the retrofitting effect produced

by the steel jacket, and the effect of vertical gaps of

varying length between the lower end of jacket and

the top of the footing, one as-built specimen (P-l)

and three retrofitted specimens (P-2 ~ P-4) were

used in this test. The vertical gap length in specimen

P-2 was 10 cm, while those of specimens P-3 and

P-4 were 5 cm and 15 cm, respectively.

3-2. Loading Procedure

Photo 3 shows the way in which each specimen was

loaded. While an axial force equivalent to a

dead-load reaction was applied to each specimen

laid down, a level load was applied to it by an

actuator installed on a reaction wall. An axial force

of 55 tf was used in this test. This is equivalent to a compressive stress of 15.3 kgf/cm 2

.

The yield displacement at the loading point was

determined here when the longitudinal reinforcement

of the base of the as-built specimen yielded, or the

anchor bars of each retrofitted specimen yielded.

Alternative displacements were given to the

specimens by the displacement control. The loading

displacement was made by a sine wave; die loading

speed 3 cm/sec; and the number of loading cycles

per loading step was to be three. The ultimate

displacement during the test was defined here as the

displacement when the strength was deteriorated to

the yield strength.

4. RETROFITTING EFFECT OF STEEL JACKET

WITH ANCHOR BARS

4-1. Hysteresis loop and aggravation of damage

Figure 3 shows the hysteresis loops of the

relationships between lateral load and the

displacement at the loading points of all the

specimens. Figure 4 illustrates the damage to specimens P-l and P-2 was aggravated.

In the case of P-L the as-built specimen, based on

the value for the longitudinal reinforcement strain at

the base, the yield displacement was measured to be

16.0 mm. Its peak strength was 23.1 tf. This was

maintained until 4 5 y was loaded, when its cover

concrete started to spall off. In the stage of 5 6 y

loading, the core concrete started to suffer damage,

and the hysteresis loop began to become unstable.

When even greater loads continued to be applied,

damage to the core concrete was aggravated, with

the level strength gradually reduced.

In the case of P-2, a retrofitted specimen, based on

the yield strain of its anchor bars, the yield

displacement was measured at 12.7 mm. In comparison with P-l, its yield displacement was

smaller, because the flexural stiffness was increased

by the steel jacket. Its peak strength was 31.8 tf. This was maintained until 6 d y was loaded. Its anchor bars started to buckle under the load of 4 6

y. In the stage of the 6 6 y loading, seven anchor

bars were broken. At the same time, the external

edge of the concrete pier, in the 10cm vertical gap

provided at the base, began to fail due to the cyclic

pressure, widi the lateral strength reduced. Under the

7 6 y loading, a total of ten anchor bars broke, and

two longitudinal reinforcements of the pier were also

broken, with the lateral strength reduced below the

yield strength. Even under the final 8 6 y loading,

the steel jacket did not suffer from deformations

such as buckling or swelling out.

429
4-2. Strength and ductility

Figure 5 compares the envelope curves of the

hysteresis loops in the four specimens. Table 1 compares their strength and ductility. The

comparison between P-l and P-2 shows that the

peak strength and the ductility factor of the pier were enhanced by the steel jacket, with a controlled

increase in flexural strength. Also, while the effect

of the vertical gap length on the peak strength was

only slight, the smaller the length, the larger was the

ductility ratio. Based on the result of this test, it is recommended to make the vertical gap length small,

at least about 5 cm, in the case of piers with

dimensions similar to these specimens.

Figure 6 shows the relationship between equivalent

stiffness and lateral displacement. Equivalent

stiffness was defined here as being the gradient of

the straight line connecting the points of the

maximum and minimum displacements in the first loading loop in each loading step. The figure shows

that the equivalent stiffnesses of P-2 to P-4 under

the 1 6 y loading were about 1.5 times greater

than that of P-l. The reason for this will be that the

flexural stiffnesses of the retrofitted specimens were

greater than that of the as-built one, because of their

increased stiffness due to the anchor bars, and

because the steel jackets covering the whole height

of the piers also functioned as a longitudinal

reinforcement. However, the effect of the difference

in the vertical gap length was not noticeable over

the range of 5 to 15 cm.

4-3. Energy absorption

Figure 7 compares the specimens with regard to accumulated energy absorption. Energy absorption in

each loading step was computed here by integrating

the hysteresis loop concerned. The figure shows that

in the case of P-l, the as-built specimen, from

around the time of 5 6 y, when the core concrete

started to suffer damage, the energy absorption in

each loading step began to decrease. In the case of

P-2 to P-4, on the other hand, energy absorption was

stable even after 5 d y. It was also found that the

smaller the vertical gap length, the greater was the

energy absorption under loads above 7 6 y loading

step.

Figure 8 shows effect of the retrofitting on the

equivalent damping ratio. The ratio computed was

based here on the hysteresis loop in the first loading

in each loading step. While the ratio of P-l was

greater in the range of lateral displacement below 5

<5 y, it began to decrease when the displacement

increased beyond that. On the other hand, the

equivalent damping ratios of P-2 to P-4 rose constantly as the lateral displacement increased,

indicating that stable damping performance was

obtained with the retrofitting method. It can also be

seen that the damping performance of P-4, with a

vertical gap length of 15 cm, was slightly poorer

than those of P-2 and P-3 after 7 6 y loading step.

4-4. Strain Performance

Figure 9 compares the reinforcement and the steel jacket of P-2 with regard to the peak axial strain on

the tensile side of the section at a height of 20 cm

from the base at each loading step. Under the load

for 1 6 y, the axial strain of the jacket was

0.0838%, and it had not yet yielded. Since the

design had foreseen that five anchor bars on the

tensile side would yield under this load, it is estimated from the yield strength (observed value:

3,916 kgf/cm 2

) and section area of the anchor bars

that an axial strength, Na, of the value obtained by

the following formula was at work:

Na = 3,916 x (0.843 x 5) = 1.65 x 10

4 kgf

(1)

Assuming this axial strength to be the tensile

strength propagated to the tensile side of the steel

jacket, the strain produced in the jacket is computed

as follows:

£ j = 1.65 x 10

4 / (0.16 x 60) x 2.1 x 10

6

= 818x10" 6 (2)

As can be seen, the strain on the jacket thus

obtained agrees well with the observed value

mentioned above. Consequently, judging from the

strains on the anchor bars and the steel jacket under

the elastic loading, it can be said that the jacket

functioned effectively as a longitudinal

reinforcement of a strength equivalent to that of the

anchor bars. Also, because the jacket did not suffer

from any visible plastic deformation, such as

swelling out under the loads for 1 6 y or more

displacements, it can be expected that the jacket will

produce the proper effect of confining the concrete. The axial strain on the jacket peaked under the load

430
for 6 6 y. The displacement was equivalent to the one immediately before the anchor bars were

broken, and the strength began to decrease after

remaining stable near its maximum. On the other

hand, Figure 10 compares P-l and P-2 with regard

to the strain of the hoop reinforcement positioned at

the section at a height of 40 cm from the base. It can be seen from the figure that the strain produced

in the hoop reinforcement of P-2 was smaller at every loading step, indicating that the jacket

functioned effectively as a hoop reinforcement

confining the lateral deformation of the concrete

pier.

5. ANALYSIS OF STRENGTH AND DUCTILITY

OF REINFORCED CONCRETE PIERS

RETROFITTED BY STEEL JACKET WITH

ANCHOR BARS

5-1. Analysis Procedure

To make a seismic design procudure, it is necessary

to estimate the lateral load and displacement relation

of the retrofitted pier adequately. The relation

between lateral load and displacement is calculated

by a method based on the moment-curvature relation

described in the seismic design for reinforced

concrete piers (5). In this paper, a model of analysis

of the retrofitting effect based on this method is

therefore used.

The test results mentioned above show that the

strength and ductility of the pier are enhanced by the

steel jacket with anchor bars. Specifically, the

anchor bars and the jacket function as longitudinal

reinforcements. Only forces equivalent to the

strength of the anchor bars act on the jacket, and the

jacket also functions as a hoop reinforcement to confine the concrete. On the basis of these test

results, the following assumptions will be made in

analyzing the strength and ductility of retrofitted

reinforced concrete piers.

(1) Only the strength of the anchor bars can be

counted as a effective longitudinal reinforcement by

the steel jacket. Reinforcement bars are assumed to be positioned at the external edge of the concrete, as shown in Figure 11.

(2) The whole sectionl area of the jacket works as a hoop reinforcement. This effect, combined with that

of the existing hoop reinforcement, can be taken into

account the stress-strain curve.

(3) The vertical gap length is understood to be better

as small as possible, and the steel jacket and the

footing will not expected to come into contact if the

pier becomes deformed, the confining strength of the

jacket can be considered to be propagated into the

concrete in this section. The confining effect on the

gap section should be the same as that on the

section covered by the jacket.

With regard to the modeling of the position of the

anchor bars, there are two potential approaches. One

is to assume that the anchor bars are positioned

where they are actually anchored in the base (55.6

mm outside the external edge of the concrete pier, in

the case of P-2); and the other is to assume that they

are positioned on the external edge of the concrete

pier, where the steel jacket is placed. In the former

case, while the conditions in the actual structure and

the analysis model correspond to one other in

relation to the vertical gap section with a length

from the pier base of 5 to 15 cm, in the section

covered by the jacket the flexural stiffness tends to be overestimated, because the effective height of the

anchor bars is greater than it actually is. It was

therefore decided that, in the design model, the

anchor bars should be arranged on the external edge

of the concrete pier.

In this analysis, the model of the stress-strain curve

in concrete proposed by Hoshikuma et al. (6) was

used, since it allows the confining effect of the hoop

reinforcement to be taken into consideration.

However, the strain produced when compressive

stress is reduced to 80% of its peak is considered as

the ultimate strain (2). This is equivalent to the

strain produced when the concrete starts to spall off

due to the compression.

5-2. Analyzing retrofitting effect upon specimens

To verify the method of analysis, the strength and

ductility of P-2 were analyzed, and a comparison

between these results and the test results is carried

out. Values obtained from a materials test were used

in the analysis.

The volumetric ratio of lateral reinforcement of the

reinforced concrete pier, p s, was computed by the

following equation, converting the steel jacket into

hoop reinforcement:

431
where:

Ah: section area of a hoop reinforcement (cm 2 )

s: spacing between hoop reinforcements (cm)

dh: effective length of hoop reinforcement (2) (cm)

dj: effective length of steel jacket (cm)

t: thickness of steel jacket (cm)

fyh: yield strength of hoop reinforcement (kgf/cm

2 )

fyj: yield strength of steel jacket (kgf/cm

2 )

In Eq. (3), the first term is the volumetric ratio of

the lateral reinforcement of the pier, while the

second term gives the volumetric ratio of the steel jacket as converted into hoop reinforcement. The

effective length of the jacket, dj, can be assumed

here as the side length of the section of the concrete

pier, in the case of structures like this, in which steel

plates are placed like a jacket. Consequently, in the case of the specimens for this test, the effective

length of the jacket, dj, was 60 cm. Hence, the

volumetric ratio of lateral reinforcement of P-l, the

as-built specimen, p s, was computed to be

0.00253. The ratio of P-2, a retrofitted specimen,

p s, was 0.0115. Figure 12 shows the stress-strain curve for the concrete pier obtained by using these

values.

recorded when the strain of the most external edge

of the concrete pier reached its ultimate strain The

ultimate displacement of P-2 obtained by the

analysis method was 82.5 mm. To compare this with

the damage caused in the test when a lateral displacement of this extent was effected, the anchor

bars began to buckle, reaching the stage of damage

immediately before breaking, and the concrete

started to spall off, with the peak strength starting to

fall after having remained stable. This damage

degree corresponds to the state defined as "ultimate"

in the analysis. On the other hand, when P-2 was so deformed that its strength had been reduced to yield

strength, the anchor bars had alrea 6 y broken, and

the longitudinal reinforcement was also about to break. Based on these considerations, the ultimate

displacement obtained by the method had a value

suitable for the design, from the points of view of

residual strength, degree of difficulty of restoration,

etc. It should be noted that the plastic hinge length

would be different between P-l and P-2, as shown

in Figure 4. In above analysis, the ductility is calculated without taking account of the effect of the

plastic hinge length, which may cause the under- or

overestimate of the ultimate displacement.

6. CONCLUSIONS

Using this stress-strain curve for the concrete, an

analysis was made of the moment-curvature relation

of the section, and the lateral load and displacement

relation at the loading point was computed. Figure

13 compares the analysis results and the envelope

curve of the loading hysteresis loop for P-l and P-2. In the analysis, the peak strengths of P-l and P-2 were computed to be 24.1 tf and 31.1 tf,

respectively. These figures both agreed well with the

test results. It was concluded, therefore, that the

method of analysis was able to estimate the strength

of the retrofitted piers with sufficient precision.

On the other hand, the ductility factors of P-l and

P-2 calculated in this analysis were 3.0 and 5.3,

respectively. The corresponding test results were

both higher than these analysis values. This is due to a difference in definition of ultimate stage. In the

loading test, on the one hand, ultimate displacement was defined as the lateral displacement recorded

when the strength was reduced to the yield strength. In the analysis, on the other hand, ultimate

displacement was defined as the lateral displacement

We have proposed here a method of designing a

steel jacket with anchor bars as a method of

enhancing the ductility of reinforced concrete piers

while controlling the increase in its strength We

carried out a lateral loading test using large models,

including models retrofitted with this method. We

also examined a way of analyzing the retrofitting

effect, i.e., the effect of enhancing the strength and

ductility with this method The following is the

summary of results of the study:

(1) The test clearly showed that the ductility of a

pier can be enhanced while controlling the increase

in its strength using a steel jacket with anchor bars.

The pier's energy absorption and damping

performance were also enhanced.

(2) It was confirmed that the jacket and hoop

reinforcement were strained in such a way that the

steel jacket functions effectively as a longitudinal

and hoop reinforcement, with a strength equivalent

to that of the anchor bars.

(3) A comparison of the three cases with a vertical

gap between the lower end of jacket and the top of

the base—5cm, 10cm and 15cm~showed that, while

432
the peak strengths were all equal, the smaller the

vertical gap length, the larger the ductility factor.

(4) A method of analyzing the retrofitting effect of

this type of steel jacket with a controlled increase in

flexural strength was described. It was confirmed

that the strength calculated according to this method

of analysis corresponded well with the observed test

results. On the basis of the results of the test (number of cycles of loading: 3), the ultimate

displacement obtained by this analysis is equivalent

to the stage in which the anchor bars buckle, having

reached the stage of damage that immediately

precedes breaking. At this stage, concrete starts to

spall off, and the peak strength begins to deteriorate

after having remained stable for a time.

REFERENCES

1) Ministry of Construction: Report on the Damage

of Highway Bridges by the Hyogo-ken Nanbu

Earthquake, Committee for Investigation on the

Damage of Highway Bridges Caused by the

Hyogo-ken Nanbu Earthquake, 1995

2) Ministry of Construction: Guide Specifications

for Reconstruction and Repair of Highway

Bridges Which Suffered Damage due to the

Hyogo-ken Nambu Earthquake, Feb. 1995

3) Kazuhiko Kawashima, G.A. MacRae, Jun-ichi

Hoshikuma and Kazuhiro Nagaya: Residual

Displacement Response Spectrum and its Application to Evaluate Residual Displacement

Developed in Reinforced Concrete Bridge

Columns, 28th Joint Meeting of U.S.-Japan Panel on Wind and Seismic Effects,UJNR, May 1996

4) Kazuhiko Kawashima, Shigeki Unjo and

Hiroyuki Iida: Seismic Inspection and Seismic

Strengthening of Reinforced Concrete Piers at Mid-height Where Main Reinforcement is Terminated, First U.S.-Japan Workshop on

Seismic Retrofit of Bridges, UJNR, Tsukuba,

Dec. 1990

5) Design Specifications for Highway Bridges, V

Seismic Design, Japan Road Association, Feb.

1990

6) Jun-ichi Hoshikuma, Kazuhiko Kawashima and

Kazuhiro Nagaya: A Stress-Strain Model for

Reinforced Concrete Bridge Piers Confined by

Hoop Reinforcement, Proceedings of the Second

International Workshop, Seismic Design and

Retrofitting of Reinforced Concrete Bridges, New

Zealand, Aug. 1994

433
Fig. 1 Illustraion of Retrofit by Steel-Jacketing with Anchor Bars

to

JiOJL

tf

Lateral Force

Steel Jacket (l=l.6,SS400)

12mm Dia. Bars

50-150

Elevation

®

3

it

3(3)40 3@4p

^0 ^80=320^

^fj

^0 600 21 -24.

40

Existing Column Cross-Section

"

ON

-n

u.§>

IOo

II &

00o

:=r—=:i

a

•

0 :- =;i

- 0^ 4®! 20=480 ^6 D

55.6 • 55.6

Retroftted Column Cross-Section

Epoxy In tit I

Detail of Stiffened Collar

Fig. 2 Details of Test Specimens (unit: mm)

434
0 %

435
Lateral Displacement (mm)

-150 -100 -50 0 50 100 150

-6 -4 -2 0 2 4 6

Lateral Displacement (in)

(c) P-l Specimen

Lateral Displacement (mm)

-150 -100 -50 0 50 100 150

-6 -4 -2 0 2 4 6

Lateral Displacement (in)

(d) P-2 Specimen

Fig. 3 Hysteresis Roop of Lateral Load and Displacement Relation

436
5 5v 6 5, 7 5, 8 5.

(a) P-l Specimen

A

1

1 5,

, L

1 |

l

i '

i /1

1

il

i i

!

S" **

!

3 5.

•h h5 5,

1

75,

25,

4 5,

6 5,

T I

'

t 1 I

1

1

,

i

>

:

i

!

1

: il 1; 1: 1 i

i i 1 1

t ' 1 i

I i j

i 1 1

! i i

li 1 i r 1 1

' r

8 5,

f

—»--.-•[ —j

*

i il

| 1! 1 P

! : i

:

ill i!

| r \ ' 1

!

': i

i nm1

(b) P-2 Specimen

Fig.4 Damage Process Observed in Specimens

437
u

o

a

u

H-3

40

30 -

20 -

10 -

Fig.

0

-10

-20

-30

-40

-150 -100 -50 0 50 100 150

Lateral Displacement (mm)

5 Comparison of Envelope of Hysteresis Roop

LEGEND

P-l

P-2

2 4 6 8 10 12

Lateral Displacement ( x dy)

Fig. 6 Effect of Retrofit on Equivalent Stiffness

Table 1 Observed Strength and Ductility

Specimen P-l P-2 P-3 P-4

Yield Strength (tf) 18.8 21.3 20.2 21.5

Strength Peak Strength (tf) 23.1 31.8 31.7 31.9

PWPy 1.23 1.49 1.57 1.48

Yield Displacement (mm) 16.0 12.7 12.7 12.7

Ductility Ultimate Displacement (mm) 87.0 96.5 110.9 96.0

Ductility Factor 5.4 7.6 8.7 7.6

438
60

S 0

50 -

30

20

10

LEGEND — — P-l

P-2

- - • - P-3 _f 0 — P-4 Jr * ¥ 0 /•' '

1 /' s - A* — /' y

t /

y

-—^ 1 1 1 1

10

0. 6

12

Lateral Displacement ( x dy)

OS

&D

a

'EL

E

a

Q

c

>

'3

cr w

Lateral Displacement ( x dy)

Fig. 7 Effect of Retrofit on Accumulated

Energy Absorption

Fig. 8 Effect of Retrofit on Equivalent

Damping Ratio

Fig. 9 Axial Strain of Longitudinal Reinforcement Fig. 10 Comparison of Strain of Hoop

and Steel Plate in Specime P-2

439
Gap Section Retrofitted Section

; : Concrete

HlllllllUl R

Concrete

Anchor Bar

/Slccl Jacket A Modeling

Whole Section

Anchor Bar

Concrete

*—,

,

1

1 ,

Edes=40157-

189373

£cu=0. 00p42

ifcu^Q. 00271 \ I ecc=0. 00339J

0 0.001 0.002 0.003 0.004 0.005 0.006

Strain

Fig. 1 1 Analyzing Model of Steel Jacketing with

Anchor Bars

Fig. 12 Stress-strain Curve Used in Analysis

80

a 60 u©

fa

OSz

03 -J

40 -

20 -.'

Lateral Displacement (mm)

50 100 150

i

1 r—

Pu=53. lkips

5u=l. 71 in

/ Py =43. Skips

• 5 y=0. 57in

Pc=15. 4kips

6^=0. 07in

LEGEND — Experiment " Analysis

12 3 4 5 6 Lateral Displacement (in)

300

200

100

u

o

(>.

os

03

80

</5

4>u

o

CB -1

40 -

20

Lateral Displacement (mm)

50 100 150

12 3 4 5 6 Lateral Displacement (in)

1 1 1

Pu=68. 6kips LEGEND

6u=3. 25 in Experiment

^^^^^^ ^^^^ • Analysis

/ Py=54. 2kips \

/ 5 y=0. 59in

- ••

•

•

•

•

» Pc=16. 3kips

J 5^=0. 07in

'ill 1 1 I

- 300

200

100

ou

"5

OS -J

(a) P-l Specimen (b) P-2 Specimen

Fig. 13 Comparison of Lateral Load and Displacemetn Relation

440
Experimental Study on the Bearing Capacity of

Cast-in-Place Concrete Piles Damaged by the

Hanshin Earthquake

Hiroyuki Nakajima", Hidenao Hayashi21 and Kenji Kosa3)

ABSTRACT

To evaluate the actual behavior of pile foundations, a loading test was conducted on full-scale foundations in a box arrangement of nine cast-in-place piles. The

results show that as loading was increased until the

maximum jack stroke of 43 cm was reached, the

resisting load increased constantly. After the Great

Hanshin Earthquake, another loading test was

conducted on actual piles whose tops had cracked

during the earthquake to evaluate the degree of damage.

The bending load test results allowed us to assume that

the piles had been subjected to a maximum

displacement of 28y. An analysis that takes the nonlinear behavior into consideration lead to the conclusion

that these damaged pile tops still had the same level of

rigidity and deformation capacity as sound pile

foundations when considered as a complete pile

foundation system. The above has clarified that the

tested pile foundations possess a rigidity and

deformation capacity far greater than the levels required

by the specifications.

KEYWORDS: bearing capacity, cast-in-place concrete

piles, ultimate behavior, load deformation

1. INTRODUCTION

To establish design procedures for pile foundations of

bridges that can resist big earthquakes, it is necessary to understand the behavior of full-scale bridge structures during an earthquake or when they are subjected to large deformation. To evaluate the actual behavior of

pile foundations in September 1993, we conducted a loading test on full-scale pile foundations in a box

arrangement of nine cast-in-place piles, each having a diameter of 1,200 mm. In March 1995 after the Great

Hanshin Earthquake, another loading test was

conducted on actual piles whose tops had cracked

during the earthquake to evaluate the degree of damage.

The measurement results from these two tests are reported here.

2. HORIZONTAL LOAD TEST ON A FULL-SCALE

MODEL OF NINE PILES"

2. 1 Outline of the Test

The loading test on a full-scale model was conducted in

1993 on recently-reclaimed land (around 1990) at Kishiwada Port, Osaka Prefecture. The land is a 13-

meter layer mainly of gravel having a maximum

diameter of 300 mm. N values of the experimental site ranged from 3 to 20 and the deformation coefficient

obtained by the in-boring horizontal load test was

around Eb = 10 kgf/cm2

, indicating that the ground is

fairly loose. Under the reclaimed layer lies a diluvial

deposit consisting of alternating cohesive and sandy

soils.

The horizontal load test was conducted on both single

piles and a nine-pile group (see Fig. 1 for pile

arrangement). The footing of the tested pile group was

9.0 x 9.0 x 2.3 m. To eliminate resistance from the

surrounding soil during the test of the pile group, soil was removed from the area 10 m in front and 5 m on

both sides of the footing to a depth of 2.3 m, down to the bottom of the footing, so the test ground and the

footing bottom were level. Horizontal load was applied

in a single direction, multi-cycle method, 90 cm above

the footing bottom using six hydraulic jacks, with the

adjacent footing as the reaction force.

1) Section Chief, Design Section, Engineering Department, Hanshin Expressway Public Corporation

2) Section Chief, Research and Design Section, Kobe

Route Reconstruction Division, Hanshin Expressway

Public Corporation

3) Senior Engineer, Construction Section, Hanshin

Expressway Public Corporation

441
2.2 Results of Displacement Measurement

Figure 2 shows the relationship between the load and

displacement at a point 20 cm above the bottom end of

the footing. As loading was increased until the

maximum jack stroke of 43 cm was reached, the

resisting load also increased, attaining a maximum load

of 2,050 tf. Elastic behavior was observed in the loaddisplacement relationship up to around 600 tf, but after

the displacement reached about 10% of the pile

diameter, the increase in displacement became steep.

The logarithmic scale figure shows a bend around 800

tf, where the pile is believed to have started cracking

and its rigidity began to drop.

The load-displacement curve shown in Fig. 2 was

obtained by a method1 * that takes the non-linear

behavior of foundations based on the references for

application of the Restoration Specifications21 into

consideration. In the analysis, the following values were

used: strength of concrete oc = 350 kgf/cm2

, strength of

steel as = 3,800 kgf/cm2

, filling layer c = 0, <j) = 38°;

diluvial deposit c = 9 tf/m2

, 0 = 0; all of which were

obtained from on-site tests, and correction factors for

the single pile were Ok = 1.5, otpi = 2.5, Op2 = 1.25,

aP3=1.25, with a correction factor for pile group effect

of T| = 1.0. The ultimate load and displacement obtained

by the analysis were considerably smaller than those

from the tests, indicating that the analysis provides

values erring on the safe side.

Figure 3 shows the measurement results of footing

displacement. It can be seen from the figure that the

footing rotated while lifting. This is because the rotation

of the footing increased the compression force on the

piles in the front row, causing them to stick in the

ground while the piles in the rear row were subjected to an increasing tensile force and tended to lift. However,

the front-row piles that started to settle were then braked

by the resistance of the soil at the footing bottom, and

eventually all the piles lifted.

2.3 Displacement Distribution on the Ground Surface

and Observation Results of the Pile Tops

Figures 4 and 5 show vertical and horizontal

displacements respectively on the ground around the

test piles which were measured right after the load test. These values should be smaller than the actual

displacement during loading. However they were

determined to represent the true displacement

distribution on the ground surface under the maximum

load because contact displacement gauges showed

nearly equal values during and after loading. Surface

displacement extended over about 6D, or 7 meters, from

the piles in the front row. No distinct sliding surface was found but a 10-cm elevation was formed just in

front of the front row.

The ground around the footing was excavated about a

further 1 m after the load test to check for cracks in the

pile tops. Figure 6 shows the crack conditions in the pile

tops observed from the side. Flexural cracks were found

in the pile tops, a little below the bottom of the footing.

They measured about 5 mm extending up to 80% of the cross section of the pile even after the removal of the

load. This indicates that the ultimate stage of failure had

been passed.

3. LOADING TEST ON PILES DAMAGED BY THE

EARTHQUAKE

3.1 Profile of Loaded Piles

Loaded piles were the most severely damaged

foundation piles of the Hanshin Expressway No. 3 Kobe

Route in the Great Hanshin Earthquake. Visual

observation and investigation using borehole cameras

before the load test had revealed that the degree of

damage to the 19 piles was relatively consistent with

cracks of 2 mm in width being observed in the pile tops;

but no concrete was crushed and no reinforcements were buckled.

These cracks were narrower and less damaging than

those in the piles at Kishiwada test site (see Section 2).

A material test was conducted by sampling three core

concrete specimens and five reinforcement specimens

from the pile tops. The test results were as follows: the

mean concrete strength was 425 kgf/cm2

, about 1.8 times the design strength (ack = 240 kgf/cm2 ). The

reinforcement specimens taken from the compressed

442
section of the pile tops had a yield strength of 3,750

kgf/cm2 and a fracture strength of 5,590 kgf/cm2

, exceeding the respective reference values of 3,000

kgf/cm2 and 4,500 - 6,100 kgf/cm2

. Meanwhile,

specimens from the cracked areas had a fracture

strength of 5,400 kgf/cm2 which is high enough, but did

not have any clear yield point. This indicates that the

reinforcements had been subjected to a stress history

exceeding the yield point.

3.2 Pile Bending Test

A pile bending test was conducted on a test specimen

cut from a pile 2.8 m below the bottom end of the

footing (Fig. 7). Two adjacent piles were used as the

reaction body. A 50-tf hydraulic jack was installed

between an H-shaped steel girder laid between the two

adjacent piles and the test specimen to apply a

horizontal load to the point 2.5 m below the footing.

Multi-cycle loading was applied from one direction.

The specimen was loaded up to a maximum of 25 tf,

equivalent to 60% of the calculated bending moment to cause the pile top section to yield. As shown in Fig. 8,

the pile's horizontal displacement, curvature and load

intensity were measured in the test. Figure 9 shows the

load-displacement curves at the point where the load was applied. They are linear while the load is being

applied and little residual displacement is observed even

after load removal. This leads to the assumption that the

damaged piles have not been subjected to the ultimate

load. The rigidity obtained by the M-<j) relationship

shown in Fig. 10 is smaller than the calculated rigidity

at the yield point, indicating that the reinforcement in

the pile top has reached the yielding range. Calculation

of the displacement based on the assumption that

bending rigidity obtained from the test pile is equal to

the inclination of line (A) that can be deduced by the

load-deformation history rules, suggests that a maximum displacement of about 25y was generated.

As the next step, load-deformation curves were plotted

by using a method2

' that takes the non-linear behavior of

the foundation into consideration in order to evaluate

the soundness of the entire pile foundation system that

has cracked.

Here the rigidity of the damaged pile at the pile top over

1.5 m was reduced to the level obtained from Fig. 9. Figure 10 shows the load-displacement curves. The

degree of deformation of the damaged pile foundation is

slightly bigger than that of the undamaged one, but

almost no decrease in ultimate strength was observed.

Those piles whose rigidity had dropped due to a deformation of about 28y led to the decrease in initial

rigidity inclination and little change in strength and

deformation in the ultimate state.

3.3 Horizontal Load Test

For the horizontal load test, the pile located in the center of the footing was cut 1 .2 m below the footing bottom

and the section below was used as a test specimen. A

horizontal load was applied to the pile top. Two piles

adjacent to the test pile were used as reaction piles.

Because the pile was to be used for a new bridge, the

maximum applied load was set to 45 tf since it was the

level where reinforcement stress in the area of

maximum moment became 1,800 kgf/cm2 when the test

pile was regarded as a pile with a free top. Figure 13

shows the load-displacement curves derived from the

horizontal load test. Calculation based on these curves

gives a horizontal subgrade reaction coefficient of 1 .9 kgf/cm2 when the horizontal displacement is 1% of the

pile diameter. The horizontal subgrade reaction

coefficient can be assumed to be 2.3 kgf/cm2 using the

N values obtained in the standard penetration test designated in the Highway Bridge Specifications. The

method used in the Highway Bridge Specifications

gives mean values; when the data dispersion is taken

into account, our calculated value is believed to lie approximately in the same range, and is not an abnormal value caused by the Earthquake.

3.4 Vertical Load Test

Using the same test pile from the horizontal load test, a

vertical load test was conducted employing the footing's

own weight as the reaction force. The applied load was

450 tf, the ultimate bearing capacity calculated based on

the Highway Bridge Specifications. The loaddisplacement curves obtained are shown in Fig. 14,

which indicates almost elastic behavior with bend. It is

clear, therefore, that the load applied was below the

ultimate bearing capacity. The amount of settlement was 1.3 cm, well below the permitted level (10% of the

pile diameter = 10 cm). As a result, it is obvious that

the design bearing capacity specified by the Highway

Bridge Specifications was obtained.

443
4. CONCLUSIONS REFERENCES:

To evaluate the pile behavior under an extreme

deforming force, a series of load tests were conducted

and the following results were obtained:

1 . In the load test using a pile group (a box arrangement

of nine piles), as loading was increased until the

maximum jack stroke of 43 cm was reached, the

resisting load increased constantly also. After the

test, a crack 5 mm wide was found in the pile top

section, indicating that the ultimate state had been

reached.

2. Horizontal and vertical load tests were conducted

with the piles that had cracked in their top section

because of the earthquake. The obtained test results

roughly satisfied the ultimate bearing capacity and

horizontal subgrade reaction coefficient that were

obtained from the Highway Bridge Specifications.

3. The bending load test results using the earthquakedamaged pile tops allowed us to assume that they had

been subjected to a maximum displacement of 28y. Meanwhile an analysis taking non-linear behavior

into consideration lead to the conclusion that these

damaged pile tops still had the same level of rigidity

and deformation capacity as sound pile foundations

as an entire pile foundation system.

4. All the above has clarified that the tested pile

foundations possess a rigidity and deformation

capacity far greater than the levels required by the

specifications.

1) K. Kosa, M. Kimura, Y. Morita and K. Itoh: Full-Scale Lateral Displacement Tests on Grouped

Cast-in-Place Concrete Piles, Soil and Foundations,

pp. 43-46, May 1995.

2) Japan Road Association: References for Application

of the Specifications for Restoration of the Highway

Bridges Damaged by the Great Hanshin Earthquake

(draft), June 1995.

ACKNOWLEDGMENTS

We express special thanks to the members of the

"Subcommittee on Substructures, Technical Research

Committee of the Hanshin Expressway Public

Corporation" and the "Subcommittee on Substructures,

Disaster Prevention Committee of the Hanshin

Expressway Public Corporation" for their kind advice in completing the load tests.

444
Nine-pile group oo

Single pile o

o

o?0 <^ <=>o

01 oo

CO

oo 13 000 3 000

1 500 1 500

9 000-

® o o t o

® © 0

so

® ® o o]o

1885

-8 500-

o

CO

co o o o

o o o -9

000-

o o o

870

-9 000-

Reclaimed layer 10-

Diluvial deposit

Units: mm

Fig. 1 Arrangement of test piles

.'{.000 1

1 O Measured value immediately after application ot the preset load

• Measured value 1 5 minutes alter application of the preset load

- Analysis curve

2.000

T3

COo

.000

'JO 30

Horizontal displacement S (cm)

Fig. 2 Load-horizontal displacement relationship

V-6 P=2 OOOtf

Displacement gauge

V-2

T '1l 000

1 40 20 0

Sh (cm)

t4„

0

4

E

o

Fig. 3 Footing displacement behavior

445
2.30 -20.00

2.30

o

CO

CM

oo

oCM

O

CO

cm

Measurement mesh

0 0 5 \

-9.00

oo

X

oLO

0

1

T

4

5

10 3

10

12

> 4© © ©

Specimen pile group -©---©-©-

© © ©

1—1.20X7 = 8.40 —

,

1.50X4 = 6.00

1

Reaction body

o oo

Units: Distances: m. displacement: cm (positive numbers indicate "lift")

Fig. 4 Vertical displacement distribution of the ground surface measured immediately after load removal

. 2.30 n a n„ 2.30 20.00-

o

CO

CM

oo

o

CM

o

CO

CM

Measurement mesh

0

9.00-

oo

10 0

r—

I

11 0

•o — x0

o

in

J_0

-1\ 2

® © ®

Specimen pile group

® ® ®

c=

1.20X7 = 8.40

1.50X4 = 6.00

"T

Reaction body

o o O O

—I

Units: Distances: m, displacement: cm (positive numbers Indicate forward")

Fig. 5 Horizontal displacement distribution of the ground surface measured immediately after load removal

446
6o

r

Osaka

side

"HI"}--

No.3 No.2 No.l

Fig. 6 Cracks in the pile tops

1

Excavation level

I

Fraction

pile

it

» , i I V. ( \ I \

I 1 I \ '•_ / Test pile

V.-/

Footing

# £_

Kobe

side-T^r

J

5.

Jo

— o

-f'MST

_jl"^ajliqg

o 1,000

o

od

-Legend

-Borehole TV

Visual observation point

Fig. 7 Bending load test

0

-1

-2

-3

-4

-5

-6

-7

-8

-9

-10

-11

cn

o

o

H—V

s:

cV

E

o

o

XIu

x:

£

o

a

Q

Measurement of horizontal displacement; Measurement of crack curvature Measurement of angle of rotation and mean curvature

Fig. 8 Test pile measuring points

447
Fig. 9 Load-displacement curves of the test pile Fig \q Moment-mean curvature of the pile top section

3

o

60 />, = 41.0tf J> = 5.3nm

60

20

Pu =60.9tf

Su — 23.3 mm

20

10 20

Displacement (nun)

Line A: Targeting the maximum value

40 Line B: Equal to the rigidity at the yield point

60 From the Highway Bridge Specifications

Fig. 1 1 Load-deformation history

2000

- Sound pile top

"

' Damaged pile top

I i i i i i i 10 20

Horizontal displacement S (cm)

Fig. 12 Load-horizontal displacement curves

448
Fig. 13 Load-displacement curves (horizontal load)

Fig. 14 Load-displacement curves (vertical load)

449

MANUSCRIPTS AUTHORED FOR

PANEL MEETING BUT NOT

PRESENTED ORALLY
3i!Iij
ANALYSES OF LIQUEFACTION IN THE 1995 HYOGOKEN-NANBU EARTHQUAKE

by

Osamu MATSUO *

Takao SHIMAZU *

*

Tatsuya TSUTSUMI *

*

ABSTRACT

This paper presents analysis results of field

performance data on liquefaction during the 1995

Hyogoken-Nanbu earthquake. The paper contains the

following: (l)Grain size distribution of soil deposits at

liquefied spots in the earthquake is compiled. (2)Two

case records of liquefaction of gravelly soil deposits are

presented. Finally, (3)Based on the observation of

liquefaction occurrence, the relationship between the

standard penetration test (SPT) N-value, soil gradation

and dynamic shear stress ratio induced during

earthquakes is analysed.

1. INTRODUCTION

The 1995 Hyogoken-Nanbu earthquake caused

extensive liquefaction in a wide area of offshore

reclaimed lands and natural deposits. Ground flow

associated with soil liquefaction took place at waterfront

in reclaimed lands as well. One of the important aspects

in geotechnical engineering was liquefaction of gravelly

soils in reclaimed lands. This phenomenon has attracted

geotechnical engineers' interest because liquefaction of

gravelly soils had been considered to be exceptional.

Another aspect was whether evaluation methods of

liquefaction potential currently used in practices could

well explain the performance of the ground during such

a strong earthquake motion as the Hyogoken-Nanbu

earthquake.

Having these geotechnical concerns, the authors

reviewed liquefaction phenomena during the earthquake

from a couple of viewpoints.

2. GRAIN SIZE DISTRIBUTIONS OF LIQUEFIED

SOILS

During past decades, a number of papers have

reported grain size distribution of ejected soils due to

liquefaction in past earthquakes. It has been recognized,

however, that soil gradation of sand boils is never

identical to that of liquefied soils at depths because the

liquefied soils are segregated in the course of ejection.

This section summarizes grain size distributions of

penetration core samples retrieved from the supposedly

liquefied soil deposits.

Figure 1 shows location of borings that were

performed along the Hanshin Expressways after the

Hyogoken-Nanbu earthquake. The two routes of

expressways shown in the figure were viaducts and a number of piers were collapsed during the earthquake.

Figure 2 summarizes the grain size distributions of the

subsoils at the boring locations. In the figure, mean grain

size of the liquefied soils ranges from 0.1 to 1 mm for

the Kobe Line, and from 0.3 to 10 mm for the Harbor

Line. Fines contents are below 40-35 % for both the

routes. In the Kobe Line, soil gradation of the

non-liquefied soils is coarser than that of the liquefied

soils. As shown in Fig.l, the Route No.5 runs through

offshore reclaimed lands. The reclaimed lands were

constructed with weathered granite soil, called "Masado",

that had been taken from the Rokko Mountains. The soil

contains a wide range of grain size particles, from fine

silt to gravel. Figure 3(a) plots the relationship of fines

content and mean grain size. The relationship of

Holocene alluvial sands are comparatively shown in

Fig.3(b)

1

. The figure demonstrates that the Masado

soil contains more fines than alluvial sands with the same mean grain size.

3. LIQUEFACTION OF GRAVELLY SOILS

In the Hyogoken-Nanbu earthquake, liquefaction

of gravelly soils in the reclaimed lands attracted

geotechnical engineers' attention, because the soils called

Masado with very large uniformity coefficient had been

considered insusceptible to liquefaction. A couple of case

records of liquefaction of gravelly soils are presented in

the following. One case is for a reclaimed land and the

other is for a natural deposit.

* Head, Soil Dynamics Division

Earthquake Disaster Prevention Research Center

Public Works Research Institute, Tsukuba, 305 Japan

** Researcher, ditto.

453
3.1 CASE 1: The Koshien-hama Reclaimed Land

In the Koshien-hama reclaimed land locating

offshore Nishinomiya, occurrence of sand boils was

widespread and caisson-type revetments displaced

seaward. Figure 4 demonstrates a boring log obtained

after the quake at a site where significant amount of

ejected sand was observed. As shown in the figure, a

thickness from surface down to 14 m is composed

mostly of gravelly soil, or the Masado soil.The layer

below 14 m depth is a marine clay, the original seabed

material. The thin clay layer intersecting the Masado

layer was formed with dredged material during a certain

period of reclamation interval. Because the presumably liquefied Masado layer

contained gravel, the large penetration test (LPT) was

performed in addition to the standard penetration test

(SPT). Main differences of both tests are presented in

Table 1. The LPT is considered more applicable to

gravelly soils than the SPT because of their difference in

the shoe diameter

z)

. In Fig.4, the blowcounts in the

Masado layer are as small as between 2 to 10 for both

the tests. The correlation between the LPT and the SPT

blowcounts are given in Figure 5. The LPT N-values,

Nd, are less than the SPT N-values, N. Previous studies

have shown that the ratio of N to Nd is about 1.5 for

sand and it is getting larger with increasing gravel

content

2)

. In Fig.5, however, the ratio is about 1.5 or

less. The grain size distribution of the Masado is

presented in Figure 6 that was obtained by seiving

large-scale undisturbed samples (30 cm diameter). In the

figure, volcanic gravelly soils that liquefied during the

1993 Hokkaido Nansei-oki earthquake

3

, and clean

sands that liquefied during the 1964 Niigata earthquake

4)

are also plotted for reference. It is seen in Fig.6 that mean grain size, maximum grain size and gravel content

of the Masado soil are as large as 3 mm, 150 mm and

60 %, respectively. It appears in the figure that in spite

of the large gravel content the SPT blowcounts are not

significantly affected by the containment of gravel.

3.2 CASE 2: Mukogawa River

A naturally deposited sand gravel liquefied in the

Mukogawa River that originates from the Rokko

Mountains and flows down through Nishinomiya. Sand

boils were observed in the flood plain in the extent of

120 m by 80 m area. The site is located at the eastern edge of the seismic intensity of 7 zone, the largest

intensity (the Japan Meteorological Agency; JMA). An

abutment and piers of the Mukogawa-shinbashi bridge

locating exactly at the site had suffered displacement of a

few centimeters.

Figure 7 shows the boring logs that were obtained

in the liquefied area after the quake. The whole depth of

13 m bored are composed of sand gravel with cobbles.

According to the data recorded during foundation

construction of the Mukogawa-shinbashi bridge several

years ago, the cobbles were as large as 150 to 300 mm

in diameter. Grain size distribution of the sand gravel

measured from the SPT core samples is demonstrated in

Figure 8. It is observed that the maximum grain size is

about 30 mm, which is much smaller than the observed one during the foundation construction. Needless to say,

this difference is due to the shoe diameter of 30 mm.

In Figure 7, the SPT blow counts are 8 to 15 in

the upper 4 m and are increasing downwards. The

ground water level is 2 m below the ground surface. It is

conceived that the soil at a depth of 2 to 4 m must have

liquefied during the earthquake. It should be noted that

there is not any impermeable soil cap over the depth that

prevents the upward dissipation of excess pore-water

pressure.

4. ANALYSIS OF LIQUEFACTION AND

NON-LIQUEFACTION

4.1 Analysis Procedures

(1) Data used in the analysis

In order to eliminate local soil effects, past eight

earthquakes in Japan including the Hyogoken-Nanbu

earthquake were surveyed. The number of boring data

compiled are summarized in Table 2. Among a total of

216 data, 145 are for liquefied sites and 71 for

non-liquefied sites.

(2) Identification of liquefaction or non-liquefaction at a

siteJudgement of liquefaction occurrence at a boring

site was made primarily based on visible observation of

surface features such as sand boils, cracking, or

settlement. Liquefaction map compiled by Hamada , et

al.

5) was used for this purpose. A site with sand boils

in a nearby area was classified as "liquefied". Even at a

site where sand boils were not observed, if there were

ground movements or settlements, it was also considered

as "liquefied".

(3) Identification of liquefied depth

For a site where liquefaction occurrence was

presumed, liquefied depth must also be presumed. The

depth was selected according to the following criteria:

a) below the ground water level and within 8 m deep

454
from the ground surface,

3

)

b-1) mean grain size D so is larger than 0.02 mm ,

or

b-2) D 6 o is between 0.02 mm and 10 mm, and

fines content Fc is less than 35 %, and

c) the measured SPT N-value is the second smallest.

In the above criterion a), the depth of 8 m was

adopted considering that sand boils must have ejected

from a relatively small depth. In the criterion b), the mean grain size D so of larger than 0.02 mm comes

from an existing design manual

4>

. The fines content Fc

of less than 35 % is based also on another design

manual

6)

. The criterion c) comes from the

consideration that the smallest SPT N-value might

represent locally soft layer with very limited thickness.

(4) Estimation of maximum shear stress at depth

The peak ground surface acceleration a s m a x at a

boring site was at first estimated from the neighbouring

strong motion records. The maximum horizontal shear

stress r dmax was calculated as follows.

X dmax =r d

• O V •( a smax /g) (1)

The maximum shear stress ratio L m a x at depth was obtained by

Lmax = X dmax / O V (2)

where

r d : reduction factor with depth

(r d = 1 - 0.015- z; z: depth in meter), a v , a v ' : total and effective overburden

stresses, and

g: gravitational acceleration.

4.2 Results

Figure 9(a) plots all the 216 data according to the

criteria a), b-1) and c) as described in the above 4.1(3).

Figure 9(b) is a result obtained by employing the

criterion b-2) instead of b-1). Figure 9(c) is a result

obtained by adding to the result of Figure 9(b) the

following engineering judgements.

(a) In cases that a loose sand layer of small thickness was underlain by a dense sand deposit and accordingly

the second smallest N i -value represented the lower

dense deposit, the smallest N i -value representing the

overlaying loose layer was adopted.

(b) In cases that even the second smallest N 1 -value was considered to be representing a locally loose layer at

a certain depth, an average N 1 -value in the nearby

depth was adopted.

(c) In cases of "non-liquefied" sites that the ground

water level was very deep or the sand deposit was

capped by a thick impermeable layer, those site data were discarded, because the existence of the thick layer

might have hindered the development of surface

evidence.

In the figures, the horizontal axis, N 1 , is given as

follows

7>

:

Nl=1.7-N/( aw' + 0.7) (3)

where

N i : SPT blowcount corrected for overburden

stress to 1 kgf/cm

2

, N : raw SPT blowcount, and a v ' : effective overburden stress, in kgf/cm

2

.

In Figs.9, the black plots represent "liquefied" data

and the open plots represent "non-liquefied" data. It

appears that the two data groups are considerably well

separated. The results also suggest us that soils with

corrected SPT N-value of 20-25 or greater will not

liquefy against such a strong ground motion as that

during the Hyogoken-Nanbu earthquake.

In order to see the effect of fines content, the data were separately plotted as in Figure 10. Lines bordering

the "liquefied" plots are superimposed in the figures. The

bordering lines imply the relationship between the N i -value and the liquefaction resistance. The terminology

"liquefaction" used here may be defined as that

liquefaction of subsoils may cause any surface evidence

such as sand boils, cracking or settlement, as explained

earlier. It is observed in the figures that at the same level

of earthquake intensity as expressed by L m a x , the

bordering N i -value for sand with fines is lower than

that for clean sand. This supports the view that soils with more fines are more resistant to liquefaction for the same

penetration resistance

8>

.

5. CONCLUSIONS

The study of liquefaction during the 1995

Hyogoken Nanbu earthquake presented in this paper may

be concluded as follows.

(1) Soil gradations of subsoil materials were compiled

along two expressways. It was found that mean grain

size of the liquefied soils ranged from 0.1 to 10 mm and

fines content was within 40 %. Non-liquefied soils were

coarser than liquefied soils.

(2) Gravelly soils liquefied in a wide area of reclaimed

455
lands. Liquefaction also occurred in a natural alluvial

deposit. The maximum grain size of the liquefied

naturally depositted sand gravel was as large as 150-300

mm.

(3) A total of 216 boring data, composed of 145

liquefied data and 71 non-liquefied data, was analysed in

terms of the maximum shear stress ratio and the

corrected SPT N-value. It was found that the corrected

SPT N-values of liquefied soils were less than 20-25

and that the critical SPT N-value decreased with

increasing fines content.

ACKNOWLEDGEMENT

The authors are grateful to the Hanshin

Expressway Public Corporation for providing the soil

gradation data. Their gratitude is also extended to the

personnels concerned of Takarazuka City Government

for their permission to present the boring data regarding

the Mukogawa-shinbashi bridge.

REFERENCES

1) Yasuda, S.: Effects of soil gradation on liquefaction

resistance, Proc. of the 38th Annual Meeting of JSCE,

pp.377-378, 1983 (in Japanese).

2) The Japanese Geotechnical Society: The Geotechnical

Survey Manual, 1996 (in Japanese).

3) Tanaka, Y., et al.: Liquefaction of gravelly soils

during the 1993 Hokkaido Nansei-oki earthquake, 1994

(in Japanese).

4) Japan Road Association: Specifications for Highway

Bridges, 1990.

5) Hamada, M., et al.: The 1995 Hyogoken-Nanbu

(Kobe) Earthquake - Liquefaction, Ground Displacement

and Soil Condition in Hanshin Area, Co-published by

Association for Development of Earthquake Prediction,

Waseda University and Japan Engineering Consultants,

1995.

6) The Japanese Architecture Society: The Design

Manual for Building Foundations, pp.163-169, 1994 (in

Japanese).

7) Meyerhoff, G.G.: Discussion of Session 1, Proc. 4th

ICSMFE, London, Vol.3, 1957.

8) Tokimatsu, K., et al.: Empirical correlation of soil

liquefaction based on SPT N-value and fines content,

Soils & Foundations, Vol.23, No.4, pp.56-74, 1983.

456
Table 1 Main Specifications of SPT and LPT 2>

Standard Penetration Test (SPT)

Large Penetration Test (LPT)

Tube length (mm) 810 700

Inner diameter of shoe (mm) 35 50

Outer diameter of shoe (mm) 51 73

Rod section area (cm

2 ) 5.6 10.2

Mass of hammer (kg) 63.5 100

Falling height (cm) 75 150

Penetration length

for measuring blowcount (cm) 30 30

Symbol for blowcount N N d

Table 2 Summary of Case Records Used in the Analysis

Earthquake

Number of boring data

Liquefied Non-liquefied Total

Niigata (1964) 19 9 28

Tokachi-oki (1968) 3 0 3

Miyagiken-oki (1978) 16 22 38

Nihonkai-chubu (1983) 36 20 56

Chibaken Toho-oki (1987) 1 1 2

Kushiro-oki (1993) 3 2 5

Hokkaido Nansei-oki (1993) 4 2 6

Hyogoken Nanbu (1995) 63 15 78

Total 145 71 216

457
Figure 1 Location of the Borings Along the Hanshin Expressways

Kobe Line ( Route No.3 ) Harbor Line ( Route No.5 )

Particle size ( mm ) Particle size ( mm )

Particle size (mm) Particle size (mm)

( N: Number of data )

Figure 2 Grain Size Distribution of the Hanshin Expressway Subsoils

458
100

80

o

- 60

4—

•

c

e 40

o

o

c/> S> 20

i r 1 r

0

0.01

o

N=72

o o

• Reclaimed soils

O Natural deposit

o

68 rfT •

ia

0.03 0.05 0.1 0.3 0.5 1.0 3.0 5.0 10

Mean grain size : Dso ( mm )

(a) Deposits in the Reclaimed Lands Along the Hanshin Expressways

0.01 0.030.05 0.1 0.3 0.5 1.0

Mean grain size : Dso ( mm )

(b) Holocene Alluvial Deposits 1 '

3.0 5.0

Figure 3 Correlation between Fines Content and Mean Grain Size

459
Figure 4 The Boring Log at the Koshien-hama

Reclaimed Land

0.001 0.01 0.1 1 10 100 1000

Particle size ( mm )

Figure 6 Grain Size Distribution of Masado Soil in the Koshien-hama Reclaimed Land

460
Boring No.B1 Boring No.B2

Boring LogiSPT N-value Boring Log STP N-value

0 20 40 60

Pa (/) <D

bbl V 4 0

0

O a Q ith <3

O Q) 0

>

(0

D)

T3C

(0

CO

20 40 60

J il l

- 5

.Depth for grain size

'analysis

( SPT core samples were used

)

10

Gravelly

sand

o o

•&

?Q

&

o o.

• a:

30

0'/ ,0

oO

.0 0 o

b

(/> 0)

-Q

O

O

x:

>

(0cn o

c

ro CO

(0 O)

(Or

03 Ol

1—

r

I

B2-1

I B2-2

Figure 7 Boring Logs at the Mukogawa River Flood Plain

j j j

i i i i i I j ji i i i i i I i j i i i i i i

0. 01 0.1 1 10

Particle size ( mm )

100

Figure 8 Grain Size Distribution of Sand Gravel at the Mukogawa River

Flood Plain

461
(b)

1.2

2 0.8

(A

Vt

Z 0.6

TO

0) JC

(/) _ . E

E

'§ 0.2

5 0.O2£D5O£1Onw

FC£35%

20 40 60

Corrected SPT N-value : Ni

(c)

o

lo

0)« 0.6 h t_

(0 |o

E

SO

5

. *

I • i i °

I

[

I j

- • • •

o : • o. . ^---oO i

^ , ...:<>—

t y* 8 ; __

j

i i i

o o

o

o

0.02£D50£10wi

FCSJ35%

20 40 60

Corrected SPT N-value : Ni

Figure 9 Plot of the Liquefied and Non-liquefied Case Records

462
(b)

1.2

I

1

O

1 0.8

(A

OT I 0.6

qo>w 0.4

E

I 0.2

(3 S

i

1

1 10£FC<20%

r ••

1

I

1

!

i

1

•• • • 1

i 1

i

i

•

• •J** • • * ° 0

o

>

w— 1 - n

!

! < ; : i

i

i i

i i

1 1 1 1 1 1 20 40 60

Corrected SPT N-value : Ni

1.2

o

1 0.8

inI

% 0.6

i_

5P0

«5 0.4

E

E „ „

1 9

i

1 o

-

I ! !

- •! ' t i

1 o

•

/ /_ _ ....

- .'•/ ; ^--o

-

; 203FC£35%

20 40 60

Corrected SPT N-value : Ni

Figure 10 Plot of the Liquefied and Non-liquefied Case Records

463

Photogrammetric Measurement of Three Dimensional Ground Surface

Displacement Caused by the Kobe Earthquake

by

Hiroshi MURAKAMI and Minoru HOSHINO

Geographical Survey Institute, Japan

ABSTRACT

The Hyogoken-Nanbu (Kobe) Earthquake in

1995 made the Nojima Fault appear on the

ground in Awaji Island, Japan. The fault was a

right-lateral fault with a reverse component. The

maximum displacement measured on the ground

along the fault was 1 .7 meter in horizontal

(right-lateral) and 1.3 meter in vertical direction

in Nojima-Hirabayashi.

The detailed measurement of ground surface

displacement around a fault has been only

available as a relative movement of one side of

the fault against the other. The authors

succeeded in measuring three dimensional

displacement of 880 points around the Nojima

fault by employing air photos of before and

after the earthquake.

The characteristics of the displacement are: 1)

on the northwestern side ofthe fault, the

horizontal displacement is directed toward east,

but the magnitude and direction of the vertical

displacement change in a complex manner; 2)

most horizontal displacement on the

southeastern side is more than one meter and

directed toward south or southeast. In addition,

large displacement can be found even in the areas beyond one kilometer from the fault line.

KEYWORDS: Awaji Island, Hyogoken-Nanbu

(Kobe) Earthquake, Nojima Seismic Fault,

Photogrammetric Displacement Measurement.

1. INTRODUCTION

The Hyogoken-Nanbu (Kobe) Earthquake

(M=7.2, Epicenter: 34.36 degree N, 135.02

degree E, depth = 17.9km; Japan

Meteorological Agency) occurred on January 17,

1995 and caused the Nojima Fault to appear on

the ground in Awaji Island (Ohta, et al., 1995;

Nakata, et al., 1995; Awata, et al., 1995;

Suzuki, et al., 1995). The Geographical Survey

Institute surveyed the fault on the ground just

after the earthquake and found that it was a

right-lateral fault with a reverse component. The

maximum displacement measured on the ground

along the fault was 1 .7 meter in horizontal

(right-lateral) and 1.3 meter in vertical

directions in Nojima-Hirabayashi of Hokudancho of Hyogo Prefecture (Ohta et al., 1995).

The detailed measurement of ground surface

displacement around a fault has been only

available as a relative movement of one side of

the fault against the other. Ground surface

around a fault measured in detail for a spatial

large extent of an area might be useful for

modeling the crustal movement of the

earthquake.

The authors applied a photogrammetric

approach measuring three dimensional

displacement around the Nojima fault by

employing air photos of before and after the

earthquake. This paper describes the result of

the measurement and discusses the feasibility of

applying photogrammetric approach to ground

surface displacement measurement caused by

earthquakes (Hoshino, et al., 1996).

2. STUDY AREA AND MEASUREMENT

The study area covers the north eastern part of

Awaji Island, Japan and encompasses an area of

approximately 20 km2, 13 km long along the

Nojima fault and 5 km wide (Figure 1). Air

photos covering the study area of before and

after the earthquake (before- and after-quake

photos) were available at the scale of 1 : 8,000

and 1:5,000, respectively. The before-quake

black and white photos were taken by the local

government, Hokudan-cho in October, 1983 and

the after-quake color photos by PASCO

Corporation, Japan on the next day of the

earthquake, January 18, 1995. Ground features

clearly defined on the before-quake photos (e.g.,

road intersections, house roof tops, foot bases of

electric power towers and bridges) were

identified on the after-quake photos and their

three dimensional coordinates measured on both

465
photos to obtain the displacement vectors for

each measured point. A total of 880 points were

identified and measured on these photos.

The orientation parameters of before-quake

photos were available as the result of aerial

triangulation conducted by Hokudan-cho. The

accuracies of aerial triangulation was 26 and 30

cm for horizontal and vertical directions,

respectively. For the horizontal control of afterquake photos, thirteen horizontal GPS control

points were established for this study as shown

in Figure 1 by employing two existing

permanent GPS control points (Takao and

Toshima) which had been established by the

Geographical Survey Institute (GSI) of Japan as part of their permanent GPS control point

network to monitor the crustal movement. The

vertical control was obtained through a level

survey starting from an existing benchmark,

Iwaya which had also been established by GSI

(Figure 1). GSI's level survey after the

earthquake revealed a 1 .96 cm uplift of this

benchmark. The aerial triangulation errors of

the after-quake photos were 5 and 7 cm on the

ground for horizontal and vertical directions,

respectively. Out of the 880 points measured in

this study, 545 points were selected from house

roof tops, 115 from paddy fields, 102 from road

intersections, and 108 from others. Not many

ground features were found appropriate for

measurement on the southeastern side of the

fault which is covered with forests and does not

have many man-made features. Some houses

were severely damaged and distorted and

considered inappropriate for the measurement

because photogrammetric approach only allows

to measure house roofs instead of ground

surface itself. Houses with too much damage

were plotted on "maps of houses to be

dismantled" by the local governments based on

the owners' requests. A total of 1 18 measured

houses were identified in the maps, and

considered to have too much distortion, and

their data were removed from the measurement

result.

3. RESULT

A part of the photogrammetric measurement

result of Nojima-Hirabayashi area, where the

largest relative ground surface displacement

was observed, is shown in detail in Figure 2. This figure clearly shows the difference of

displacement patterns between northwestern and

southeastern side ofthe fault. The northwestern

side moved eastward with subsidence and the

southeastern side southward with uplift.

Figures 3 and 4 show the result of the

horizontal and vertical ground surface

measurement of the whole study area. The

direction of horizontal displacement is not

parallel to the fault line, and it is east- or northeastward on the northwestern side ofthe

fault and south- or southeastward on the

southeastern side. The vertical displacement of

the southeastern side of the fault is mostly

upward while the northwestern side has both

downward and upward displacement areas

distributed in a complex manner, especially,

downward and upward displacement areas appearing alternately along the fault. These

figures illustrate that the displacement is not

localized along the fault line, but large

displacement of more than one meter can be

found even in the areas over one km apart from

the fault line. The significance of the photogrammetric

measurement as opposed to ground survey is seen in Nashimoto where relative uplift of the

northwestern side by 40 cm was observed on the

ground (Figure 5). Figure 4 shows this area was

uplifted on both side of the fault instead of only one side.

Both horizontal and vertical displacement

patterns are summarized in Figure 6.

4. DISCUSSIONS

4.1. Accuracy of Measured Data

The aerial triangulation accuracy for beforeand after-quake photos were 26 cm for

horizontal and 30 cm for vertical directions and

5 cm for horizontal and 7 cm for vertical

directions, respectively. However, these

accuracies do not include errors associated with

the operator's misidentification and/or

observation errors of the objects. Since there

was no other independent source of direct

measurement for the displacement caused by the

earthquake, it was difficult to validate the total

466
accuracy of the ground surface measurement

conducted in this study. However, the ground

survey measurement along the fault can be

considered as an independent data to estimate

the accuracy of the measured displacement.

Four ground survey points, Nojima-Hirabayashi,

Nojima-Ohkawa, Nashimoto and Ogura were

selected for the accuracy estimation. As no

object measured on the air photos was located

exactly on the fault, the area within 15 m from

each ground survey point encompassing both

sides of the fault was considered uniformly

moved and the photogrammetric measurement

data within the area were used to calculate the

relative displacement along the fault for

horizontal and vertical directions. The difference

of the relative horizontal displacement between

direct ground survey and photogrammetric

measurement was largest in Ogura with 58 cm

and smallest in Nashimoto with 12 cm. The

difference of relative vertical displacement was

largest in Nashimoto with 49 cm and smallest in

Ogura with 6 cm. The root mean square error (RMSE) of the photogrammetric measurement

based on the relative ground surface

displacement was 33 cm for both horizontal and

vertical directions. Since these errors are comparable to those of aerial triangulation, the

operator error is considered relatively small

compared to the error introduced during the

aerial triangulation.

4.2. Feasibility of Measuring Houses

Successful measurement of ground surface

displacement caused by an earthquake heavily

depends on the density of man-made features

firmly fixed to the ground and clearly

identifiable on both air photos before and after

the earthquake. Natural features such as trees,

rivers and shorelines are subject to changes over

time. Unfortunately, there are not many features

that satisfy these conditions. Building and house

roof tops are the most clearly identifiable

features on air photos, and the measurement in

this study relied on their accuracy in terms of

representing the ground surface displacement.

As discussed above, however, it is very difficult

to visually tell whether a particular house or

building is heavily damaged or not from vertical

air photos.

In this study, "maps of houses to be dismantled"

were employed to remove damaged houses from

the measurement result assuming that houses

with distortion of about 30 cm, the accuracy of

the photogrammetrc measurement, should be

dismantled by the owners' request. Although this

assumption may not be perfect, the

measurement result of houses is comparable to

that ofthe other features, and considered usable

in this study.

5. CONCLUSIONS

This study focused on photogrammetric

measurement ofthe ground surface

displacement caused by the Kobe earthquake

and the following conclusions were derived:

i) The direction of horizontal displacement is not parallel to the fault line, and it is east- or northeastward on the northwestern side of

the fault and south- or southeastward on the

southeastern side of the fault;

ii) The ground surface displacement is not

localized along the fault line, but large

displacement of more than one meter can be

found even in the areas over 1 km apart from

the fault line; and

iii) The overall accuracy of the

photogrammetric measurement was

approximately 33 cm for both horizontal and

vertical directions and small enough

compared to the actual displacement of more

than 1 m around the fault line. The

photogrammetric measurement approach

adopted in this study should be considered

successful in detecting the ground surface

displacement around the fault. However, this success depends on the magnitude of ground

surface displacement and the scale and

quality of air photos available for the

measurement. Unfortunately, air photos of

large scale are not necessarily available for

the area with large ground surface

displacement caused by an earthquake.

ACKNOWLEDGMENTS

The authors wish to thank Dr. Takashi Tada of

GSI and Dr. Kohei Tanaka of the National

Research Institute for Earth Science and

467
Disaster Prevention for their suggestions and

comments for this study. Hokudan-cho of

Hyogo-Prefecture, Japan kindly provided

valuable maps and materials used in this study.

Mr. Masayoshi Maeshima and Mr. Ryozo

Onozuka of GSI assisted in the data analyses.

This study was funded by the Ministry of

Construction, Japan as part of their

Comprehensive R&D Project program.

K. Sango, 1996. Ground Surface

Displacement around Nojima Fault

Emerged at 1995 Hyogo-ken Nanbu

Earthquake(Kobe). GEOGRAPHICAL

REVIEW OF JAPAN, Vol.69(Ser.A),

No.7, Julyl996(in print).

REFERENCES

Awta, Y. , T. Yoshida, K. Mizuno,

Y. Miyati, K. Sgiyama, A .Sangawa,

T. Kamai, I. Isobe, K. Okamura,

E. Tukuda, F. Yoshida, H. Kamata and

N. Matsuyama, 1995. Surface faults and

geologic hazards with the Hyogokennanbu earthquake. Proceedings of

Preliminary Research Result on

Hyogoken-Nanbu Earthquake of January

17, 1995.

Ohta, Y. , M. Horino and Survey Group on

Disasters, Geographical Survey Institute,

1995. Notes on the Earthquake Fault on Awaji Island, Associated with the 1995

Hyogoken-nanbu Earthquake and Related

Disasters. Journal of Geography, 104(1)

143-155

Suzuki, K. ,Y. Umino,

M. Horino, J. Kisanuki, M. Hoshino,

J. Iwahashi, H. Mizukishi, T. Nemoto,

O. Nakano and G. Iida, 1995.

Geomorphological Phenomena and

Damage in the Northern Part of Awaji

Island Caused by the 1995 Hyogokennanbu Earthquake. GSI Journal, 83, 34-

51

Nakata, T. , K. Yomogida, J. Okada,

T. Sakamoto, K. Asahi and N. Chida,

1995. Surface Fault Ruptures Associated

with 1995 Hyogoken-Nanbu Earthquake , Journal of Geography 104(1) 127-142

Lin, A. , H. Imiya, S. Uda, K. Iimura, T.

Misawa, T. Yoshida, Y. Abemitsu, T.

Wada and K. Kawai, 1995. Investigation

of the Nojima Earthquake Fault Occurred

on Awaji Island in the Southern Hyogo

Prefecture Earthquake.

Hoshino, M. , H. Murakami, H. Mizukoshi and

468
Figure 1. Study Area Control Point Distribution and

Level Servey Networks.

469
0 250 500 m

I i I i 1

Figure 2. Ground surface displacement

around No j i ma-Hi rabayashi.

470
Figure 3. Distribution of horizontal

displacement.

471
NOJ I MAEZAK

Legend

Vertical displacement

UP

O 1.5 m

o 1 m

• 0.3 m

DOWN

-1m

0 - 0.3m

NOJ I MAH I RABAYASH

N0JIMAT0D0R0KI,

o o o

9 °-*P "O"

o_

NASH I MOTC

TOSH I MA

5

1

TONOUCHIi

D H3, u o On OO n

9

°

o

o.O

Q3 ID

°.° o

a n"

o a»o Q oton

° C o° n O

o

o

3Q DD

O a

a

'a t

an. q

°o

CP

a-°D

m O

CP

= a Pti

a

CP

•ti

'

iP

P55

<9 °o.

N

CP

3 km i.i.i

\

Figure 4. Distribution of Vertical displacement.

472
km

O-i

1-

2-

3-

4-

5-

6-

7-

8-

North-side South-side Right- lateral

(Up) I (Up) I

0.5 0.0 0.5 1 .0 1 .5 m 0.0 0.5 1 .0 1 .5 2.0 m

Beach

Noj imaezaki

Noj ima

hi rabayashi

Noj imaokawa

Noj ima

hikinoura

Nash i moto

Tosh i mahokubu

Figure 5. Displacement by ground survey.

473
Area where the horizontal displacement is more than or equal to 1 m.

<**- Area where the horizontal displacemen is less than 1m.

Figure 6. Generalized distribution of

horizontal and vertical displacement.

474
A Real Time System for Forecasting Storm Surges Caused by the Typhoons

by

Tatsuo KONISHI

ABSTRACT

A dynamical-numerical model and a new

procedure for forecasting storm surges on a real time basis are proposed. The model is a

traditional two-dimensional storm surge equation, on the other hand, the parameters of storm model

are determined by using the geostationary

meteorological satellite data for estimating radius

of maximum winds of the typhoon, in addition to the forecaster's information about current and

predicted positions and intensity of the typhoon.

The results of hindcast and experimental forecast

due to this method with application to past

typhoons are shown in this paper.

KEYWORDS: storm surge; typhoon; real time

forecast; model; wave set-up

1. INTRODUCTION

As shown in Konishi(1995a), the frequency of

high storm surges in Japan has increased recently,

being associated with the same tendency of the

frequency of typhoons hitting in Japan.

Quantitative forecast of storm surges before event

is very important for disaster prevention. At

present, the Japan Meteorological Agency (JMA)

applies an empirical formula to the real time

forecast of storm surges (JMA.1994). The

method has rather good accuracy, but also has a

limitation such that we can not forecast storm

surges at a point where there is no storm surge

record in the past. Therefore, a dynamicalnumerical model must be developed for the

forecast of future storm surges.

The purpose of this paper is to propose a procedure of storm surge forecast on a real time

basis and to show the results of hindcast and

experimental forecast and the associated problems.

2. STORM SURGE MODEL

The method of numerical simulation utilized in

the present study is the same as that in Konishi

(1989). The two-dimensional equations of

motion and continuity are expressed as

dt ax p p

8N J^-^ T —+ fM- -gD

dt dy

Q

)

t W by

P

.(1)

.(2)

w

*..TM_»JL >(3)

dt dx dy

where the origin, O, is within the equilibrium

level of the sea surface, Ox points east, Oy north,

Oz is directed vertically upwards. We define u and v as the x and y components of velocity and

then M and N are the x and y components of

volume flux given by

o o M - judz , N - jvdz , -D -D

respectively. { is the sea surface displacement

from the mean sea surface level, D is the depth, g

is the gravity acceleration, and f is the Coriolis

parameter. Parameterizations of r s x and r b x are given as follows:

t s x = P aCd|W|w x , r b x = PwCdw|V|v x , where p a is the density of the atmosphere, Cd is the drag coefficient at the sea surface, W is the

wind velocity vector, Wx is the x component of

W, p w is the density of sea water, Cdw is the

drag coefficient at the sea bottom, V is the mean

water velocity vector, and Vx

is the x component

of V. T

v

and are parameterized in a

similar way. Water head C 0 is originated from

the effect of atmospheric pressure and the relation

is P a g C 0 =-(Ps-Pn). Ps and Pn are the

surface pressure of the atmosphere and the

peripheral pressure of the typhoon, respectively.

* Typhoon Research Department

Meteorological Research Institute, JMA

1-4, Nagamine, Tsukuba, Ibaraki, 305, Japan

Present Address: Oceanographical Division,

Marine Department, JMA 1-3-4, Otemachi,

Chiyoda-ku, Tokyo, 100, Japan

475
Drag coefficients, Cd and CdWj_are assumed to be

3.5 X 10 and 2.6 X 10

*~

, respectively.

The values are selected from discussion of past

studies, especially, Konishi and Tsuji (1995).

Boundary conditions are based on Konishi (1989).

On the coastal boundary, the condition of zero normal velocity is applied. As open boundary

conditions, modified radiation conditions are used.

3. TYPHOON MODEL

In the state of the art, the accuracy of prediction of

a typhoon track is not sufficient for storm surge

prediction, especially for disaster prevention.

Therefore, several calculations must be done to predict storm surges for the possible courses of the

track. For this, we are obliged to use a parametric typhoon model in place of the results of numerical weather prediction. Fujita's formula

(Fujita,1952) was supposed as the distribution of

atmospheric pressure, which is given by

Ps-Pn —; ——=, (4)

where AP is the centra] pressure depression of a typhoon, r is the distance from the center, r0

is the

parameter which is related to the radial

distribution of the atmospheric pressure. Wind in the typhoon has been calculated for hindcast of

storm surges by Miyazaki's formula (Miyazaki et al.,1961) written in the form,

W - q • G' +C

2

• C • exp(-Jt /rr

e

), (5)

where G

1 is a wind vector, C is the velocity

vector of the center of the typhoon and re is an

attenuation factor. The magnitude of rt is normally assigned to 500km in the formula and

the same value was set for Tt in the present study.

G' has the same magnitude as the gradient wind

vector G and the direction of G' is turned

inward of the typhoon at a certain angle from G

to take the effect of friction between the

atmosphere and the earth's surface into account. The magnitude of G is calculated by the equation,

1 1/2 ^ 4

.y

fr dps 2

-1 + l+( ,(6)

where the notations are the same as those in Section 2 and Ps is given by equation (4). There

are insufficient real-time data to allow a direct

analysis of the central region of most typhoons.

Therefore, equation (5) is simplified and supposed

to be written as W - C'(G'+C • exp(-ji r/r

e

)) (7)

The inflow angle between G' and G is supposed

to be 30 degrees as in many past studies (e.g.

Unoki et a!., 1964). Based on these suppositions,

the parameters we must determine in real time are Pn, A P, r0 , and C of equations (4) and (7) and

positions of the typhoon. The latter gives the

typhoon velocity C in equation (7).

At fixed times, the forecaster in charge determines

the position of the typhoon center and the

intensity of the typhoon, and predicts the course and intensity change of the typhoon 12 or 24

hours later. Information on the present intensity

includes central pressure, Pc, maximum surface

wind, Wm, radius of over 50-kt winds, R50, and

mean radius of lOOOhPa, R1000. We assign the

interpolated positions between the present position

and that 12 or 24 hours later to those in the

numerical model. Storm surges are sensitive on

the wind structure near the typhoon center. In order to include information on the structure in estimating r0 , we intend to evaluate the radius of

the maximum wind, Rm, which has not been

determined in our real-time forecast system.

Chubachi(1988) performed a regression analysis

between the positions of maximum surface winds

from aircraft reconnaissance due to the Joint

Typhoon Warning Center (JTWC) at Guam and

the observed cloud patterns of typhoons from the

geostationary meteorological satellite "Himawari".

He made three types of the regression equations

for three categories of cloud patterns, EYE, CDO,

and OTHERS. EYE pattern means that the eye

of a typhoon is clearly found in CDO(Central

Dense Overcast). CDO pattern means that no eye

but CDO is found in the satellite image.

OTHERS indicates that there is no eye and no

organized cloud image. Good correlations, about

0.6 to 0.7, were obtained for EYE and CDO

patterns, and the regression equations of the

position of maximum winds are

Rm=3.857*E

-0.271*Mw7+0.531*Cus+0.227*Pc

479.613 for EYE pattern (8)

Rm=1.452*Cuw

-0.470*Mw7+0.061'Tyn+27.552

for CDO pattern (9),

where E is the length of the major axis of the eye

in unit of 0.1 degrees, Mw7 is the maximum wind

476
speed of 700 hPa in unit of kt, Cus and Cow are

the extensions of the convection cloud towards

south and west respectively in unit of 0.1 degrees,

Pc is the centra] pressure in unit of hPa, Tyn is the

latitude of the typhoon center in unit of 0.1

degrees. The Meteorological Satellite Center of

the JMA determines these cloud parameters

except Mw7 routinely at 6-h intervals. So we can

use them for forecast of storm surges. However, no relevant data for Mw7 are found. The

maximum wind on 700 hPa is a little stronger

than that at the surface; on the other hand the

observed wind by an aircraft is weaker than the

maximum wind. These have opposite tendencies

to each other, and it is expected that Mw7 may be

nearly equal to Wm. The maximum surface wind

determined by the forecaster, Wm, is applied to Mw7. Now we can evaluate all terms in the

right-hand sides of equations (8) and (9); and for

the patterns of EYE and CDO, Rm can be

estimated reasonably. Using the estimation of Rm

and the information of the forecaster with

equations (4), (6), and (7), the magnitudes of r0

and C were determined in the following way.

Firstly, the value of r0 was selected, which

minimizes the square error between Rm and the

position of maximum wind calculated by equation

(7) and between R1000 and the calculated one. Next, using the r

0

, C was determined by

minimizing the square error between wind speeds

of Wm and 50kt and the corresponding calculated

values.

For the category of OTHERS, we can not obtain a proper value of Rm, and without information on Rm, the parameters are determined only from the

forecasters' information, such as Wm, R50, and

R1000. This can be validated from the following

viewpoint. The state of no eye and no organized

cloud, as the OTHERS pattern, means that the

spatial scales of variation of wind and pressure

fields become large and the position of maximum

wind also becomes obscure. Therefore, it is not

effective for calculations of storm surges in

OTHERS patterns. In this case, we applied a

recursive procedure to estimating the appropriate r0 , which is similar to Ueno(1994). The

procedure described above is summarized in Fig.l

4. RESULTS OF HINDCAST

APPLIED TO TOKYO BAY

Above-mentioned method was applied to Tokyo

Bay and hindcast studies were performed. Fig.2

shows the comparison of peak surges between

observations and calculations of six typhoons,

which caused big storm surges in Tokyo Bay in the past. This includes the results observed at ten

tidal stations in and around Tokyo Bay. Rather

good agreement can be found in Fig.2. The

calculation simulates the observed peak surges

within the accuracy of about 40 cm.

The temporal variations of the calculated storm

surges are shown in Fig.3 for the case of Typhoon

8506(1RMA). The calculated values agree well

with the observed in this figure, as well as in Fig.2,

but a large seiche after the peak surge at Tokyo

and some underestimation of storm surges at Minami-izu. The latter appears in all cases to some extent and improvement must be made on

this point.

5. PROBLEM 1 — EFFECT OF

WAVE SETUP

The calculation at the ports facing southerly on

open ocean has the common feature of

underestimation as at Minami-izu. One of the reasons is considered as the effect of wave setup

proposed by Longuett-Higgins and Stewart(1962).

The existence of this effect has been confirmed in the hydraulic experiments, but in situ has not,

because the wave setup becomes prominent only

in case of high waves (significant wave height

may be greater than 4m), moreover it was difficult that the detection of the effect needs the separation

of contributions by air pressure and wind stress properly from observed storm surges.

As above-mentioned method determines storm

parameters objectively, the separation can be done

reasonably. Konishi(1996) studied the wave setup

effect with this method. He computed the storm

surges at Minami-izu for six cases and subtracted

the computed values from the observed storm

surges, being called as SSD. SSD was compared

with the wave data at Irohzaki which is very close

to Minami-izu and found a good correlation

between them. Though the analysis is continuing

now, we will need a method in future, in which

the effect of wave setup is properly incorporated

for forecasting storm surges.

477
6. PROBLEM 2 — EFFECT OF

MODIFICATION OF TYPHOON

STRUCTURE

We also faces on an another problem for real time

forecast of storm surges, litis is related to the

effect of modification of typhoon structure after landing of the typhoon. Fig.4 shows the

distribution of peak surges observed around the

western part of Inland Sea of Japan in case of

Typhoon 9119(MireilIe). The simulated peak

surges with the present method are also shown in

the figure. The position of station number in

abscissa is found in Fig.5. The experimental

forecast results are shown in Fig.6, whose

calculation is only based on the information at 9

o'clock on Sep.27th 1991. After about 12 hours

later of the time, peak surges occurred. Computed

conditions are tabulated in Table 1. From fig.4

the present method simulated observed peak

surges with a good manner, on the other hand, the

forecast cases show some overestimation

especially in the eastern part of computed area. The overestimation remained, even when we use

the best track and the observed central pressures

for computation(Case 5 in Fig.6). The difference

between Case 1 of Fig.4 and Case 5 of Fig.6 is due

to the actual variation of radius of maximum

winds. In Case 1, the closer the typhoon

approaches to the analyzed area, the smaller the

radius becomes, but in the forecast case(Case 5)

the variation is not considered because the

meteorological forecast information does not

include such tendency. Detailed analysis of the

event shows that it is related to the topographic

convergence (see Kooishi(1995b)). Thus, in order

to raise the accuracy of storm surge forecast, it is important to forecast the detailed wind structure

in the stormy period at forecast time.

7. CONCLUSION

A real time system for forecasting storm surges was developed. With the application to hindcast

of past events for Tokyo Bay, the accuracy of the

method is about 40 cm of peak surges. Hereafter,

we must incorporate the effect of wave setup and

consider the procedure to include the effect of

typhoon modification in order to raise the

accuracy of storm surge prediction.

REFERENCES

Chubachi, K., 1988: The position of maximum

wind within the distance of 200 km from

typhoon center, Document of task force in

JMA for analysis of typhoon (unpublished

document).

Fujita, T., 1952: "Pressure Distribution Within

Typhoon", Geophysical Magazine, 23, 437-

451.

Japan Meteorological Agency, 1994: Tide Tables

for the Year 1995,265.

Konishi, T, 1989: "Numerical forecast of storm

surges on a real time basis", Oceanographic

Magazine, 39, 21-42.

Konishi, T., 1995a: "Storm surges in the western

part of the Inland Sea, caused by Typhoon

9119 -analyses of tidal records, documents on

flooding and field inspection", Umi to Sora

(SEA AND SKY), 70,171-188 (in Japanese).

Konishi, T, 1995b: "An experimental storm surge

prediction for the western part of the Inland

Sea with application to Typhoon 9119", Papers

in Met. and Geophy., vol.46, 947.

Konishi, T. and Tsuji, Y., 1995: "Analyses of the

storm surges in the western part of the Inland

Sea of Japan caused by Typhoon 9119"

Continental Shelf Res., 15, 17954823.

Konishi, T, 1996: "A cause of storm surges

observed at the ports facing on the open ocean

(II)", Preprints of the Spring meeting of Japan

Oceanographical Society, No.229 (in

Japanese).

Longuet-Higgins, M.S. and R.W. Stewart, 1962:

"Radiation stress and mass transport in gravity

waves, with application to 'surf beat'", Jour, of

Fluid Mech., 13, 481-504.

Miyazaki, M., and T. Ueno, and S. Unoki, 1961:

"The theoretical investigations of typhoon

surges along the Japanese coast (I)", Oceanographical Magazine, 13 (1), 51-75.

Ueno, M. 1994: A monologue consisting of four

papers on tropical cyclone motion, JMA/NPD

Technical Report, 32, 28-29.

Unoki, S., I. Isozaki and S. Otsuka 1964:

Simulations of storm surges in Tokyo Bay,

Report of the Second District Port Construction Bureau, 125pp. (in Japanese).

478
6 hourly

Satellite Center

Determination of

cloud parameters

from sat. data

Forecast Center

Identification and

forecast of typhoon

positions and intensity

EYE and CDO pat. V

Radius of maximum wind

OTHERS pat.

\

i i

>

Determination of storm model

parameters

Computation of storm surges

for possible typhoon courses

V

Fig.l: Forecasting procedures of storm surges (ongoing plan).

479
Peak Surges (Tokyo bay)

0 100 200

Observation (cm)

Fig.2: Comparison of peak surges between observations and

calculations of six typhoons in Tokyo Bay.

480
Fig.3: Temporal variations of the calculated and observed storm

surges at Tokyo, Yokosuka, Ito and Minami-izu, in the case of

Typhoon 8506(IRMA).

481
Peak surge (simulation)

0

1

JL p r e S € n b m < i t ho d

2 l m a > 6

».... w "W. n

j... Q

\

n

f

l\\\

/ .

"1/

' 7

I \ -Lt

1 / << \\

\.v

\

c b s e r i

' e d u

2 4 6 8 10 12 14 16 18 20 22

Station number

Fig.4:Comparison of simulated peak surges with those observed at

the stations around the western part of the Inland Sea in the

case of Typhoon 91 19(Mireille). Computational conditions are

given in Table 1. For the station number on the abscissa, see

Fig.5.

482
Fig.5: Domain of computation. The numeral with solid circle

corresponds to the station number in Figs.4 and 6. Contours of

the depth are drawn.

Table 1 Computational conditions of Figs. 4 and 6

Case Track Pc To C a

(hPa) (km) (Cl,C2)degree

1 Best Obs. Present method

2 Best 0 b s . (Estimated wind)*

3 Forecast 935 49 0.85 30

4 Best 935 49 0.85 30

5 Best Obs. 49 0.85 30

(^Estimated from tidal data, see Konishi and

T s u j i ( 1 9 9 5 ) .

483
Peak surge (forecast)

s

0>

d 2

o

rvJ F i > r e < : a St 4r, r a c k

4 t t el

5 B < (P

i.S

c;

t... r.a C 1

I.)...

t —A s

/ / -/r/ '\ \

/ i

(//

z

rrV

\\\ Ml

...Vv

/''

"N

\ ///

( M

''

....

"V; 7

..... ...... p

//

M7..... \... v

.../../ /

\ 1 / / 4

1 J //

\

...« /

i

ft

\ c b 3 e r i

' e d 4

ti it

\

V-

.......

2 4 6 8 10 12 14 16 18 20 22

Station number

Fig.6: Same as in Fig.4, but for forecast cases.

484
Windborne Debris in Hurricanes and Typhoons

by

Joseph E. Minor*

ABSTRACT

A major change is taking place in the way

buildings are constructed for hurricanes and

typhoons. The sustained, turbulent winds in

these severe tropical cyclones carry large

amounts of debris onto building facades,

breaking windows and subjecting the

building interior to internal pressures and

rain. Concerns with occupant safety and

with damage to building contents have

prompted changes to building codes in the

U.S. Several new code provisions require

special designs for glazed openings in order

to protect windows from being penetrated by

windborne debris during hurricanes.

Research conducted for tornado protection in

schools, certain public buildings and critical

facilities, including buildings in the nuclear

industry, has been applied to the mitigation

of damage caused by windborne debris in

hurricanes. Building products, including

shutters and impact resistant glazing, have

been developed to protect buildings from

windborne debris by rejecting debris impacts

while remaining in place throughout the

storm.

KEY WORDS: buildings; building codes;

debris; glass; hurricane; typhoon; wind.

1. INTRODUCTION

It is common practice for an architect or engineer to design a building in a given

community for the wind speed specified in

the local building code. Using conventional

methods outlined therein, this wind speed is

used to develop pressures for cladding

design. Implicit in this process is the

assumption that if the cladding is strong

enough for the pressures accompanying the

largest gust in the design windstorm, it will

be strong enough for pressures

accompanying all of the other gusts, since

these gusts will be smaller than the gust

associated with the design wind speed.

Experiences with buildings in a wide range

of different environments suggest that this

conventional approach to defining loads and

designing cladding is inadequate for tropical

cyclone prone areas. Hurricanes and

typhoons contain sustained, turbulent winds

that generate significant amounts of

windborne debris that can present a major

threat to cladding, especially architectural

glazing. Further, the gusty winds cause

repetitive loadings that can affect glazing

systems before and after impacts from

debris. In building environments where

these conditions can occur, breaching of the

building envelope produces large internal

pressures and water damage. Further, the

post-breakage behavior of glass becomes an

important consideration because the glass

must stay in the opening following breakage

to protect the envelope and not fall to the

street below. Effective designs have evolved

in response to new criteria for design that

involve debris impact and pressure cycles

representing fluctuating wind in the

hurricane or typhoon.

* University of Missouri-Rolla, Rolla,

Missouri, 65409-1170 USA

485
2. EXPERIENCES 4. WINDBORNE DEBRIS

Minor and Behr (1993) summarize events

over a 25 year period during which

architectural glazing in buildings was broken

during hurricanes. For example,

architectural glazing in several tall buildings

in Houston, Texas was broken by windborne

debris generated by Hurricane Alicia (see

Figure 1). Hurricane Andrew damaged

facades in several major buildings in south

Florida (see Figure 2). Failures of cladding

systems during hurricanes result in damage

to buildings, loss of building contents,

interruption of business, hazards to people

and marred images for buildings that have

sustained damage.

In an example from Hurricane Andrew, the

insured loss to a commercial building was

US$33 million, only US$3 million of which

was related to damage to the exterior of the

building. The balance involved damage to

the building interior, damage to building

contents, contents replacement costs,

business interruption costs and business

relocation expenses.

3. HURRICANES AND TYPHOONS

The effects of winds in hurricanes and

typhoons are especially harsh (Minor,

1984). Turbulent winds can affect a building for hours. These winds change

slowly in direction as the storm approaches

and passes over a building (see Figure 3).

Debris can be progressively dislodged from

adjacent structures, accelerated by sustained

winds and impact several elevations of a

building. Following impact, the building

can be buffeted by cyclic wind pressures for

hours before the storm moves away. This

severe environment has not been recognized

by building designers, building officials and

product manufacturers.

The first suggestion that impacts from

windborne debris were a likely cause for

failures of architectural glazing systems

during windstorms was recorded by Reed

(1970). He observed that windborne debris

was a significant factor in the relatively

large window glass breakage rates at low

elevations on high-rise buildings during the

Lubbock Storm of May 11, 1970. In

assessing failures of structures during

windstorms, Minor, Mehta and McDonald

(1972) observed that windows are

traditionally designed for wind pressures, but

that breakage from impacts by windborne

debris is the most common failure

mechanism.

In extensive wind damage surveys conducted

throughout the 1970s, a pattern became

apparent that proved to be common to all

types of extreme wind events (Minor and

Mehta, 1979). Small debris, principally roof

gravel, can be carried into all elevations of

building facades at velocities sufficiently

large to break glass. Large debris, including

framing timbers and roofing materials, can

impact the building envelope near ground

level with sufficient force to penetrate wall

coverings and break windows. Research has

defined the nature of both small and large

debris in windstorms.

3.1 Small Missiles

Minor (1974) analyzed the window

damaging mechanism during windstorms in

urban areas. He identified roof gravel as the

principal form of small debris that causes damage to whitlows in the upper floors of

high-rise buildings. Field surveys of typical

built-up roofs (conventional "tar and gravel")

established the average roof gravel size as 0.6 gm, and an average large size of roof

gravel as 5 gm. Harris (1978) defined the

486
threshold of roof gravel impact velocities

that cause breakage in the three basic types

of window glass (annealed, heat strengthened

and fully tempered). Threshold average

breakage velocities for a 5 gm steel ball

representing roof gravel impacting 6 mm

(1/4 in.) annealed, heat strengthened and

fully tempered glass were reported by Harris

as 10 m/sec (31 ft/sec), 12 m/sec (36 ft/sec),

and 20 m/sec (65 ft/sec), respectively. Thus,

all types of glass in this common thickness

can be broken by windborne roof gravel, if

impact speeds are larger than about 1/2 of

gravel carrying wind speeds of 38 m/sec (80

mph) or more.

Travel distances exceeding 30 m (> 100 ft) in

Hurricanes Alicia and Andrew permitted

roof gravel to be accelerated to large

fractions (1/2 or more) of the 38 m/sec (80

mph) wind speeds estimated to have

occurred at the building sites illustrated in

Figures 1 and 2. Heat strengthened and

fully tempered glass in windows on these

buildings were broken.

3.2 Large Missiles

In their investigations of tornadoes, Minor,

McDonald and Mehta (1978) concluded that

the most prevalent type of windborne debris

in residential areas is timber from wood

frame houses. Individual timbers were

observed to have broken windows,

penetrated walls and roofs, and impaled in

the ground. Additional windborne timbers

were attached together as parts of failed

roofing systems, timber trusses and timber

walls. These observations led to the

selection of a 2x4 timber weighing 7 kg (15

lbs) as a representative object for use in

defining impact criteria for tornado shelters

in schools and residences (DCPA, 1975).

The ease with which this 2x4 timber, which

is about 3.7 m (12 ft) long, can be propelled

by a compressed air "cannon" also led to its

becoming a standard for impact testing of

walls, shutters and window systems.

In 1974, Tropical Cyclone Tracy devastated

the City of Darwin in Australia's Northern

Territory. Walker (1991) observes: "It was

clearly demonstrated in the investigation of

Cyclone Tracy that the two major factors

contributing to the widescale damage to

housing were internal pressurization of

buildings following failure of windward

windows, generally due to windborne debris,

and fatigue failure of cladding and metal

connections under the fluctuating pressures."

A 4 kg (9 lb) 2x4 timber (about 2.7 m; 9 ft long) first appeared as a design missile in a building code in the Darwin Area Building

Manual (1976). This design missile appears

in the Australian national wind loading code

(SAA, 1989), is used both editions of the

South Florida Building Code (1993, 1994)

and in Southern Building Code Congress

International (SBCCI) Standard SSTD 12-94

(1994).

A 2x4 timber, 3.7 m (12 ft) in length, and

travelling at 1/2 of the design wind speed

was advanced as one of three "large" objects

recommended for use in the design of lower

elevation cladding on urban buildings by

Minor, et al. (1978). This recommendation

was based upon field documentation

conducted by the Institute for Disaster

Research at Texas Tech University and

research being conducted by the nuclear

industry on missile impact speeds in

tornadoes and other extreme windstorms. A

list of large missiles for use in design is

contained in Department of Energy Standard

DOE-STD-1020-94 (1994).

4. BREACHING OF THE ENVELOPE

The building envelope must remain integral

during windstorms. Breaching of the

building envelope produces several harmful

487
effects: internal pressure, exposure of

occupants to wind and rain, damage to

building contents, additional debris in the

wind stream, hazardous debris falling to the

street, disrupted business, and a blemished

image of the building. Windborne debris

has been established as a principal cause for

the breaching of the building envelope

during hurricanes. Failure of windward wall

cladding (windows and wall coverings) or doors leads to an increase in pressure inside

the building and can produce failure of the

principal structural frame. As shown in

Figure 4, internal pressures can effectively

double forces acting to lift the roof and push

side and leeward walls outward.

5. POST-BREAKAGE BEHAVIOR

The post-breakage behavior of architectural

glazing in urban high-rise buildings is

important for two reasons. First, hazards to

people on urban streets below establishes

the importance of keeping broken particles

of glass within the window opening should

breakage occur. Second, the importance of

preserving the integrity of the building

envelope (see discussion in 4. above and

Figure 4) reinforces the benefit of keeping

broken glass in the opening should debris

impact induced breakage occur. Laminated

glass (two "plies" of glass joined by a plastic

interlayer) offers opportunities to achieve

both objectives (keeping particles within the

opening and preserving the integrity of the

building envelope) following breakage.

Understandings of the hurricane

environment, and the concerns with public

safety and property protection discussed

above, have produced test protocols for

cladding designs. These protocols use

missile impact, followed by application of a

pressure spectrum, as a basis for qualifying

architectural glazing system designs for

hurricanes and typhoons. Pantelides, et al.

(1994) established a test in which roof

gravel weighing 2 gm is propelled onto the

window at a speed equal to the design wind

speed for a specific building. Following

multiple impacts, the window system is

subjected to a sequence of pressure cycles to

test its ability to remain in the opening

following breakage. Similar test protocols

are employed in SBCCI Standard SSTD 12-

94 (1994) and the two editions of the South

Florida Building Code (1993. 1994).

The post-breakage pressure spectra employed

in the above test protocols were developed

from a succession of three reports that

contain authoritative assessments of

fluctuating pressures on cladding produced

by the passage of a hurricane. Jancauskas,

et al. (1990) devised the basic analytical

model of a tropical cyclone crossing a building site and used this model to derive

cyclic pressure spectra for roof cladding.

Mahendran (1993) extended this work and

offered a simplified pressure spectrum for

use in testing. Finally, Letchford and

Norville (1994) applied these concepts to

wall cladding and outlined a 18,500 cycle

pressure spectrum that they recommended be

used to evaluate the post-breakage behavior

of glazing and external protection devices.

6. EFFECTIVE DESIGNS

Architects and engineers have used the

design concepts and test procedures outlined

above to design buildings which may be

exposed to impacts from windborne debris.

Further, the glazing, door and shutter

industries have developed new glazing

products, cladding system designs, and

external protective devices (shutters) that

have been tested according to criteria

outlined in the new building codes and test standards.

Pantelides, et al. (1993) report that a

488
laminated glass product that is anchored to

the window frame with a silicone anchor

bead can survive the small missile impacts

and the cyclic pressure spectrum (see Figure

5). An "anchored film" design in which a polyethylene terethalate (PET) film is

applied to the inner surface of monolithic

glass and anchored to the window frame

with a batten bar has also passed the small

missile test and the cyclic pressure spectrum.

The Building Code Compliance Office in

Dade County, Florida USA has approved

more than 350 products that meet the South

Florida Building Code debris impact

requirements. Most of the approved

products are shutter systems, but windows,

doors, skylights, and wall coverings are included in the approved products list. An

innovative new window system for

residential applications is comprised of a

thick layer of polyvinyl butyral (PVB)

covered by a thin layer of polyethelyne

terephthalate (PET) and applied to the inner

surface of monolithic heat-strengthened glass

(see Figure 6). Commercial size insulating

glass units with both lites made of laminated

glass have also passed the large missile

impact test and pressure cycles prescribed by

the South Florida Building Code.

6. CONCLUSION

Twenty years of meteorological research,

building performance evaluations, wind

engineering, and test method development

have produced fundamental changes in the

way the building envelope is designed to

survive hurricanes and typhoons.. Two

overlooked aspects of the windstorm

environment, windborne debris and

fluctuating pressures, have found their way

into building codes, test standards and design

practice. New impact resistant glazing

products, shutter systems and architectural

glazing have emerged in response to these

fundamental changes in the concepts of

design for hurricanes and typhoons.

7. ACKNOWLEDGMENTS

Support of wind engineering activities dating

from 1970 by the Hazards Mitigation

Section of the U.S. National Science

Foundation has been instrumental in the

performance of building evaluations, the

conduct of basic research and the

development of test methods that relate

windborne debris to performance of the

building envelope. The Institute for Disaster

Research and the Glass Research and Testing

Laboratory at Texas Tech University, and

the State of Missouri through the Missouri

Research Assistance Act and the University

of Missouri-Rolla, have also supported these

activities. The Saflex® Division of

Monsanto supported debris impact related

research projects at various times during this

period. Participation by the above

organizations in addressing the various facets

of windborne debris and the building

envelope are gratefully acknowledged.

8. REFERENCES

1. Darwin Area Building Manual. Darwin Reconstruction Commission,

Darwin, NT (Australia), 1976.

2. DCPA, "Interim Guidelines for

Building Occupant Protection from

Tornadoes and Extreme Winds,"

Defense Civil Preparedness Agency,

Washington, DC, 1975.

3. DOE, "Natural Phenomena Hazards

Design and Evaluation Criteria for

Department of Energy Facilities,"

DOE-STD-1020-94, U.S. Department

of Energy, Washington, DC, April,

1994.

489
Harris, P. L., "Effects of Thickness

and Temper on the Resistance of

Glass to Small Missile Impacts,"

thesis presented to Texas Tech

University, Lubbock, TX in partial

fulfillment of the requirements for

the degree of Master of Science,

1978.

1984), ASCE, New York, 1985, pp

151-167.

10. Minor, J. E., Beason, W.L., and

Harris, P.L., "Designing for

Windborne Missiles in Urban Areas,"

J. Structural Div.. ASCE, 104 (1978)

1749-1759.

8.

Jancauskas, E.D., Walker, G.R., 11.

Mahendran, M., Capitanio, C. and

Prien, G.D., "Computer Simulation of

the Fatigue Behaviour of Roof

Cladding During the Passage of a Tropical Cyclone," Proceedings of

the Twelfth Australasian Conference

on the Mechanics of Structures and

Materials. Queensland University of

Technology, 1990, pp 327-334.

12.

Letchford, C.W. and Norville, H.S.,

"Wind Pressure Loading Cycles for

Wall Cladding During Hurricanes,"

J. Wind Engin. and Indus. Aerodv. . Vol. 53, (1994), pp 189-206.

Mahendran, M., "Towards an

Appropriate Fatigue Loading

Sequence for Roof Claddings in 13.

Cyclone Prone Regions," Queensland

University of Technology, Brisbane

(Australia), 1993.

Minor, J. E., "Analysis of the 14.

Window Damaging Mechanism in

Windstorms," dissertation submitted

to the graduate faculty of Texas Tech

University in partial fulfillment of

requirements for the Ph.D. Degree,

Texas Tech University, May 1974. 15.

Minor, J. E., "Window Glass

Performance and Hurricane Effects,"

Proceedings. Specialty Conference:

Hurricane Alicia: One Year Later

(Galveston, TX, August 16-17,

Minor, J. E. and Behr, R. A.,

"Architectural Glazing Systems in

Hurricanes: Performance, Design

Criteria and Designs," Proceedings,

the 7th National Conference on Wind

Engineering. (University of

California, Los Angeles, June 27-30,

1993), UCLA, Los Angeles, 1993, pp

453-461.

Minor, J. E., McDonald, J. R., and

Mehta, K. C, "The Tornado: An

Engineering Oriented Perspective,"

NOAA Technical Memorandum ERL

NSSL-82, National Weather Service,

NOAA, Washington, DC, 1978

(reprinted as NOAA Technical

Memorandum NWS SR-147, 1993).

Minor, J. E. and Mehta, K. C,

"Wind Damage Observations and

Implications," J. Structural Piv. . ASCE, 105 (1979) 2279-2291.

Minor, J. E., Mehta, K. C. and

McDonald, J. R., "Failures of

Structures due to Extreme Winds," L

Structural Div. . ASCE, 98 (1972)

2455-2471.

Pantelides, C. P., Horst, A. D., and

Minor, J. E., "Postbreakage Behavior

of Heat Strengthened Laminated

Glass under Wind Effects," L

Structural Engin. . ASCE, 119 (1993)

454-467.

490
16. Pantelides, C. P., Horst, A. D., and

Minor, J. E., "Post-breakage

Behavior of Architectural Glazing in

Windstorms," J. Wind Engin, and

Indus. Aerody.. 41-44 (1992) 2425-

2435.

17. Reed, Jack. W., "Window Damage

Study of the Lubbock Tornado,"

Report No. SC-TM-70-535, Sandia

Laboratories, Albuquerque, August

1970.

18. SAA Loading Code Part 2: Wind

Loads, AS1 170.2-1989, Standards

Australia, North Sydney, NSW,

Australia, 1989.

19. SBCCI, "SBCCI Test Standard for

Determining Impact Resistance from

Windborne Debris," SSTD 12-94,

Southern Building Code Congress,

International, Birmingham, AL

(USA), December 1994, 16 pp.

20. SFBC, "Metropolitan Dade County

Edition of the South Florida Building

Code." Metropolitan Dade County,

Miami, Florida, adopted December

14, 1993.

21. SFBC, "Broward County Edition of

the South Florida Building Code." Broward County, Ft. Lauderdale,

Florida, adopted February 10, 1994.

22. Walker, G. R., "The Application of

Wind Engineering Technology to the

Mitigation of Damage to Housing

from Tropical Cyclones - An

Australian Achievement," preprint of

a paper presented at International

Conference on the Impact of

Disasters, University of California,

Los Angeles, CA (USA), July, 1991.

Figure 1. Window Glass Breakage in Downtown Houston, Texas Following Hurricane

Alicia in August 1983

491
Figure 2. Window Glass Breakage in South Florida (USA) Following Hurricane Andrew in

August 1992

Figure 3. Hurricane Wind Field Showing Winds Spiraling Inward Toward Eye (A building

at A will experience slowly changing wind directions from North to Southeast as the hurricane crosses the building from right to left.)

492
Figure 4. Breaching of the Building Envelope Results in Internal Pressures that can Double

Forces Acting to Remove Roof and Push Leeward and Side Walls Outward

, 3/16 in.

Figure 5. A Laminated Glass Unit that is Anchored to the Window Frame Will Remain in

the Opening During Pressures Cycles that Follow Breakage by Windborne Debris

1. Abrasion Resistant Coating

2. PET Film

3. PVB Film

Figure 6. An Innovative Laminated Film that is Factory Applied to Monolithic Glass can

Stop Impacts from Windborne Debris and Survive Pressure Cycles that Follow

493
]

!
LIQUEFACTION POTENTIAL OF COARSE FOUNDATION MATERIALS:

LOST CREEK DAM

by

Ronald A. Oaks1

ABSTRACT

The difficulty in using Standard Penetration Testing

(SPT) in coarse grained deposits to determine the

existence ofpotentially liquefiable soils is discussed

in this document. As an alternative method of

testing these coarse grained soils, a Becker Hammer

Penetration Testing (BPT) program was conducted.

The various details of conducting this program are discussed. In addition, the use of the Becker

Hammer as a penetration testing device is compared

with the SPT.

Three layers of potentially liquefiable soils have

been identified in the foundation at the downstream

toe ofLost Creek Dam. These layers were detected

during the Field Exploration Program [1] conducted

in 1991. Based on the results of this program, cross canyon continuity of potentially liquefiable layers

was assumed. Due to the manner in which the

overbank deposits that comprise the alluvial

foundation occurred, geologists have reasoned that

these liquefiable layers could extend upstream of the

dam [2]. If a seismic event occurred in the area of

Lost Creek Dam, these foundation soils could

liquefy resulting in an upstream and/or downstream

failure of the dam and catastrophic release of the

reservoir.

This paper presents the geologic explorations

performed to identify the layers of potentially

liquefiable materials in the foundation of Lost Creek

Dam Only the foundation materials downstream of

the dam were explored since exploration in the

reservoir was cost prohibitive. The results of the

exploration are assumed to apply to the foundation

materials upstream of the embankment. The

continuity of liquefiable layers (both

upstream/downstream and cross canyon) is discussed for the downstream foundation.

The rationale for adjustments and results of two

SPT and one Becker Hammer Penetration Test BPT

exploration program(s) are presented in this paper.

The results of geophysical testing conducted at this

site are also presented.

KEYWORDS: site investigation; foundations;

liquefaction; Standard Penetration Test; Becker

Hammer.

1. INTRODUCTION

Lost Creek Dam is located on the western slope of

the Wasatch Mountains about 12 miles northeast of

Devils Slide, Utah, and 30 miles east of Ogden,

shown on figure 1. The dam is situated in a

relatively narrow section ofLost Creek Valley about

2,000 feet downstream from the confluence of Lost

Creek and Francis Creek, where the valleys of both

streams form the reservoir basin. Total reservoir

capacity is 22,5 10 acre-feet at normal pool elevation

of6,005.0 feet. A plan view of Lost Creek Dam is shown on figure 2.

The embankment of Lost Creek Dam is a zoned

earth and rockfill structure with a crest length of

1,110 feet, a crest elevation of 6,022 feet, a crest

width of30 feet, a structural height of 248 feet, and

a hydraulic height of approximately 190 feet, as shown on figure 3. The embankment was completed

by the Bureau of Reclamation (Reclamation) in

1966 and contains approximately 1.8 million yd3 of

earthen materials.

Appurtenant structures for Lost Creek Dam consist

ofan uncontrolled concrete-lined spillway located in

the right abutment and an outlet works structure

tunneled through bedrock adjacent to the spillway.

The spillway has a design capacity of 2,455 ft

3/s;

the outlet works has a design capacity of 800 ftVs at elevation 6015.9.

'Civil Engineer, U.S. Bureau of Reclamation, Denver, CO 80225

495
KEY MAP

Figure 1. Location Map

496
LOST CREEK DAM

Figure 2. Plan View

497
Top of joint use a 6005.0-^ L-i Crest of dam

UOL UTS £160119 ~*\ S?F~ Crest £1 6022

6008

f

-£15950 " £15945

A Surface settlement

I point

-Stripping ' r>^^rJyK^,

Assumed rock surfoce^'^\afft ~3^i£^

^€15655

yXI 5850

-Thickness of Zone blanket decreases from o feet at a5850 to 5 feet at a 6000

jo--'i"*

Cnut noles * iCers. (opproi)

MAXIMUM SECTION

0 100

SCALE OF FEET

too

I

EMBANKMENT EXPLANATION

(7) Selected day, silt, sand, and grant compacted by

tamping rollers to 4-Inch layers

(?) Selected sand, gravel, and cabbies, or roc* fragments

compacted In it-hen layers by crawler type tractor

(?) Miscellaneous day, silt, sand, gravel, and cabbies, or

rock fragments compacted m it- Inch layers by

pneumatic tired roller

LOST CREEK DAM

Figure 3. Cross Section of Dam

498
2. GEOLOGY.

Alluvial deposits, ranging up to 55 feet thick, are present under the shells of the dam, both upstream

and downstream. These materials were removed in

the dam cutoff trench excavation which extends to

bedrock.

Deposits of slopewash and talus are also found

beneath the abutment slopes of the dam. These

deposits are generally very thin but reach a thickness

of up to 20 feet.

The locations of exploration holes used to

characterize the foundation are shown on figure 4.

2.1 Foundation Materials

A cross section of the geology at the toe of the dam

is shown in figure 5. The top horizon of foundation

material at the location of the 1991 exploration drill holes at the toe of the dam consists of an

approximately 5-foot thick layer of fill material

immediately downstream of the embankment at surface elevation 5830.0. This material consists of

angular limestone blocks ranging in size from

3 inches to 2 feet in diameter with minor amounts

of sand- and gravel-sized fragments.

Beneath the 5-foot fill material layer, the first

potentially liquefiable layer of alluvial material

(Qal,) is encountered. At the downstream toe it extends from approximate elevation 5825 down to

approximate elevation 5821. Material in this layer

is classified as silty sand to sandy silt (SM, ML).

The next layer of material encountered is the

alluvium (Qal) laid down by Lost Creek. It is located between approximate elevations 5821 and

5813.5. Material in this layer is classified as predominantly poorly graded gravel with silt and

sand (GP, GP-GM, SP, SP-SM, SM). Based on

limited SPT data and confirmed with the BPT, this

layer is not considered to be liquefiable.

From approximate elevation 58 13.5 to approximate

elevation 5806.5 is another liquefiable layer Qal2

. Material in this layer has been classified as sandy

silty clay and silty sand (CL-ML, SM). As

described above, this is the layer in which the lowest

SPT blow counts occurred.

Beneath Qal2 is another layer of the alluvium, Qal.

It occurs from approximate elevation 5806.5 down

to approximate elevation 5800. This material has

been designated as predominantly poorly graded

gravel with sand and cobbles (GP, GP-GM, SP).

Based on limited SPT data and confirmed with the

BPT, this material is not considered to be

liquefiable.

From elevation 5800 to 5780 is a layer designated

Qal3

. Material in this layer is classified as poorly

graded gravel with sand and cobbles (GP, GP-GM,

SP). This material was identified as being

potentially liquefiable by investigators in 1994 [2].

3. STANDARD PENETRATION TESTS

PERFORMED IN 1991.

To determine if potentially liquefiable materials

were present in the foundation below the

downstream toe of Lost Creek Dam, a field

exploration program was undertaken as part of

Reclamation's Dam Safety process in 1991. SPT's

were performed in four test holes, SPT-1 - 4, as shown on figure 4. They were completed along the

toe of the embankment using a Sprague and

Henwood skid mounted, rotary drill. This work is documented in reference [1].

3. 1 Adjustments of Raw Blow Counts, SPT 1991.

The SPT blow counts obtained were adjusted in two

ways. First, raw blow counts which appeared to be

increasing as the SPT sampler advanced down the

hole were plotted on graph paper. An increase in

blow counts as the SPT sampler is advanced

indicates that a large gravel or cobble sized particle

may be lodged in the sampling shoe and that this

may be creating more resistance than the SPT shoe.

Due to this constriction at the sampling shoe, the

blow counts would be artificially high. The

cumulative blow counts are therefore plotted versus the cumulative penetration in feet. The data used

for this analysis were collected during the drilling

program that recorded how many blows were

required to advance the SPT sampler each 0. 1 foot

within the test interval of 1.0 foot. Once the data were plotted, a linear fit was made to that portion of

the graph that shows that the blows per tenth are not

increasing or are the least erratic. From this linear

fit, the assumed number of blows that it would have

taken to drive the SPT sampler through material

499
500 i 1

unimpeded by a large gravel or cobble sized particle can be read from the graph by following the linear

fit line out to the 1.0 foot mark and reading the

corresponding number of blows. The raw data

and graphs used in making these adjustments can be

found in reference [2].

3.2 Adjustment to N,(60) Blow Counts, SPT 1991.

Once the raw blow counts requiring adjustment as described above have been evaluated, all the blow

counts (those requiring adjustment for increased

resistance, and those that did not) are entered into a spreadsheet and adjusted to N,(60) blow counts.

The process for adjusting these blow counts is outlined in reference [3], and is made based on fines

content and effective stress of the material being

penetrated by the SPT. The spreadsheet used for

performing this adjustment is documented in

reference [2] and recreated, for drill hole SPT91-2,

as figure 6.

Figure 6 also shows whether a ground motion

acceleration (A,^) of 0.32 g — the predicted

maximum acceleration associated with the

Maximum Credible Earthquake (MCE) — will

produce liquefaction of the materials tested [3].

This is done by multiplying N,(60) by O^-k^lc, to obtain a corrected blow count designated as NA

. The factors ka and kj correct for the initial static

shear stress and the overburden stress, respectively.

Values for ka and k0 were determined as described

in reference 3. This new blow count NA

is then

compared to the blow count (Nr) below which

liquefaction is expected to occur. The required blow

counts (NJare developed based on the magnitude of

the earthquake, the cyclic stress ratio (t^o,,1 ), and

the fines content of the layer in which the test is being performed. Once NR

is determined, it is compared with NA

. If the is less than then

"Yes," that particular layer is subject to liquefaction

during the MCE (see fig. 6). If the NA

is greater

than NR then "No," that particular layer is not

subject to liquefaction during the MCE.

3.3 Conclusions Related to the 1991 SPT Program.

Based on the exploration program performed in

1991, three stratigraphic layers were identified as having potentially liquefiable materials in the

foundation of the downstream toe at Lost Creek

Dam. These layers have been termed Qal,, Qal2 and

Qal3

, as discussed previously in the Geology

Section. These three layers range between

approximate depths of 5 to 9 feet, 16.5 to 23.5 feet

and 30 feet to bedrock (a maximum of 50 feet deep),

respectively, in the alluvial foundation materials

downstream of the dam Due to the large size gravel

and cobble particles encountered during the drilling

program and the inherent problems of testing these

materials with the SPT method, additional drilling

and other methods were required to define the

potentially liquefiable layers. Of particular interest

was the bottom layer, Qal3 in which refusal of the

SPT or some indication of increasing blow counts

was observed with increasing penetration for the

majority of the SPT tests in this layer. The

following sections of this paper present methods

used to evaluate this situation and identify if these

layers are indeed liquefiable.

A triggering analysis was performed in 1994 [4] for

Lost Creek Dam based on information from the

1991 exploration program. This analysis was

undertaken to provide information for deciding on

an interim restriction to the reservoir prior to making remediations to the dam to address the

seismic deficiency. Blow counts of 2 and 4 were

observed in the Qal2 layer of the foundation. The

Amtx that would produce liquefaction of the Qjil

based on the blow counts above, are 0.08 g and

0.13 g, respectively. The recurrence intervals for

seismic events capable of producing ground motions

of these magnitudes at Lost Creek Dam were

determined to be approximately 79 and 188 years,

respectively, showing how sensitive these

potentially liquefiable materials would be to a low

probability seismic event.

4. STANDARD PENETRATION TESTS

PERFORMED IN 1995.

As a followup to the SPT's performed in 1991, a second exploration program was undertaken to further define and confirm the upstream/downstream

extent of the potentially liquefiable materials at the

downstream toe of the embankment. Ofparticular

interest was the continuity of the Qal, and Qalj

layers with the hope of obtaining additional data

regarding the potentially lower layer of liquefiable

material, Qal3. The 1995 program used a Gus Pech

II truck mounted drill rig to drill 5 holes, DH95-1 -

5 as shown on figure 4. SPT's were performed only

in drill holes DH95-1 - 3. Drill holes DH95-4 and

502
tocu >-

10

to

0>Io

tlO

CMO

co

T3

s

E

CM

CO

oto2

II

CO 22

CM

CMCOtoc

.o5

_a> Ul0o3

COw

GJ

"co 2 8 8

COs 8

2.oXo

Q.O 8 8 cO 8

CU ii

21

0)

etc

05C

0) Q

CM

10

CUc

.2c

cu a.Dc

IS

co <Q

be: Ul

tu

o:o

iCOO

fc

LU2 otO

(0to

(Uh_ 55

a.

co

o

a.

oinCl

(Aa.

V)

co

cuk. 55

co

oX

c

cu2

cu a.

CM

CO

O

to>

LL) _J£"5. 5T o

_J2

to

CM

coto

To

s8

co

s

CO 88

CN8 98

to8

to CO 88

CM

co

CO 8888

CM8 8

CM8

CM

CM

CO

to 8 5 a. a co 88 8

CMi H

GO

El £§oUI5iCO3<

503
5 were used for the geophysical testing discussed in

the subsequent section.

4.1 Adjustments of Raw Blow Counts, SPT 1995.

The raw data (penetration per blow) obtained from

drill holes DH95-1 - 3 were reviewed and

adjustments were made, if required, to reflect the

number of blows required to advance the SPT

sampler through 1 foot of soil. An example of this

data, for one test interval within drill hole DH95-1

is shown on figure 7. Next, the data were adjusted

in the same manner as described previously for the

exploration program performed in 1991. However,

instead of collecting the number of blows required

to advance the SPT sampler 0. 1 foot, the amount of

penetration per blow was recorded. An example of

the graphs used for making these adjustments, for an interval of DH95-1, is shown in figure 8.

4.2 Adjustment to N, (60) Blow Counts, SPT 1995.

Since unit weights for the various layers of soil

sampled during the 1991 SPT program were

determined, a typical soil column was created by

averaging the elevations of the stratigraphy (see fig.

4). This soil column was used to calculate the

effective stress at the interval where the 1995 SPT

had been performed and make an adjustment based on this stress.

Gradation lab testing was performed on materials

sampled during the 1995 SPT program. The fines

contents of the materials tested were used to adjust

the blow counts and perform the analyses as described previously.

Liquefaction, based on an A^ of 0.32 g for the

MCE, for the various materials tested was analyzed

using a spreadsheet as described in the section

discussing the adjustments to the 1991 exploration

program.

4.3 Conclusions Related to the 1995 SPT Program.

Due to the nature of the material (some large size

gravel and cobble particles) encountered by the

SPT, there were quite a few tests where the sampler

either could not penetrate or that yielded results less

than satisfactory in terms of evaluating the overall

strength of the material. These blow counts

required a significant amount of interpretation in

order to develop a value of blows per foot for

certain straugraphic intervals tested. Based on these

reasons, the information obtained from this

exploration program was inadequate, for

determining if the lower (approximately) 30 feet of

the alluvial material in the foundation was

potentially liquefiable. The adjusted blow counts

from this program, for drill hole DH95-1, are shown in figure 9. As indicated in the conclusions

for the 1991 exploration program, a different

method(s) was required in order to determine if

potentially liquefiable layers within the foundation

ofLost Creek Dam, and particularly the lower layer

(Qal3), were potentially liquefiable.

5. GEOPHYSICS PERFORMED IN 1995.

After the SPT testing was completed, all 5 holes

were logged geophysically and cross hole

geophysics was performed in the triplet of DH95-2,

DH95-4, and DH95-5. The Bureau of Reclamation

considers shear wave velocities in soils less than

800 ft/s as potentially liquefiable; 800 to 1200 ft/s

likely not liquefiable, but requires some supporting

evidence to rule out liquefaction; and greater than

1200 ft/s, not liquefiable [3]. The results of this

work are documented in reference [5] and the

resulting velocity versus depth relations are reproduced in figure 10.

6. BECKER HAMMER TESTING PERFORMED

IN 1995.

Because the results of the SPT were inconclusive, a Becker Hammer Penetration Testing (BPT) and

sampling program was undertaken in October of

1995. The Becker Hammer, when used to perform

BPTs, typically uses a 6.6-inch O.D. drill casing and

a tapered closed end, crowd-out bit which is driven

by a diesel hammer. Blows from the Becker

Hammer are recorded in a manner similar to the

SPT blows. However, the Becker Hammer allows

for continuous penetration tests to be conducted

whereas the SPT requires a seating and clean-out

interval to be included with each successive

penetration test. In addition to reading the number

of blows per foot for BPTs, the bounce chamber

pressure resulting from the hammer striking the

anvil is also recorded to determine the amount of

energy imparted to the bit. Adjustments are then

made to obtain the normalized BPT blow counts

based on bounce chamber pressure and to

504
PENETRATION RESISTANCE TESTING

[FORM FOR RECORDING PENETRATION-PER-BLOW DATA]

PROJECT: Weber Basin

FEATURE: Lost Creek Dam

DATE: 4/8/95

DRILL HOLE: DH95-1

DRILLER: Brad Winters

SEATING INTERVAL: 19.0 to 19.5

TESTING INTERVAL: 19.5 to 20.5

SEATING TEST TEST

Blow

No.

Accum.

Penetr.

Penetr.

per

Blow

Blow

No.

Accum.

Penetr.

Penetr.

per

Blow

Blow

No.

Accum.

Penetr.

Penetr.

per

Blow

1 0 08 8 0.53 0.03 24 1.08 0.04

2 0.14 0.06 9 0.55 0.02 25 1.11 - 0.03

3 0.22 0.08 10 0.58 0.03 26 1.15 0.04

4 0.3 0.08 11 0.61 0.03 27 1.19 0.04

5 0.35 0.05 12 0.64 0.03 28 1.21 0.02

6 0.48 0.13 13 0.66 0.02 29 1.23 0.02

7 0.5 0.02 14 0.69 0.03 30 1.26 0.03

15 0.71 0.02 31 1.29 0.03

16 0.74 0.03 32 1.3.1 0.02

17 0.77 0.03 33 1.34 0.03

18 0.81 0.04 34 1.36 0.02

19 0.83 0.02 35 1.38 0.02

20 0.9 0.07 36 1.4 0.02

21 0.95 0.05 37 1.41 0.01

22 0.99 0.04 38 1.43 0.02

23 1.04 0.05 39 1.44 0.01

40 1.46 0.02

41 1.47 0.01

42 1.49 0.02

43 1.5 0.01

Figure 7. Raw Data from DH95-1

505
IliiliilllHIiffllllllllilillilllllllllii

E

Q

« i£ - £ 5 s o 5

*

p <

1 i 11 I I I E 1

pea 1 1 1 1 1 1 i 1I 1 iiii

Ifg R & 5 1 1 |

A3 ill

Ira & ft M a* 1 s i

#> i i I i & H. £, E" £

i I $ | $ & I i |

v- i

>*

*>" i 1 i I I ft >i Mi #

1 f 1 ii I 1 £ f. % i I

rr.

g $ f i f 1 i 1 a g as| f $ i N

a w 1

Iff! vi «

if 111 ?=: w$ 3 i w 1 K

Si M

/ * rl 1 I i 1 1 i * ISlllilil % i in § « I

1

fSS' St. V sV-- fill

s' i.

mmMmMmMEmmmmmMmmMmumBMBmm

EmBmEEEEmmmMmmmmmmBEmmBwmM

llpllllllllllllllllll^

liiii^IIlililiiiiiliiiln

ilMllIlillliiiilEIil^illlllilliiillii

ii^IiillllllilllliillSIIKyilllillllllii

ilSSiiliiiililiilliiilliililllilllliiill

i|il|jB|||||||jffl

m

n

O

CM

. o

COd d dCMd

c

3

O

um

(jaaj) uonejjausd aAjjcinainQ

506
01

m

in

o>

u5oo

CMO

co

CM

in >_

cm

</> 4> >_TCM

CMO

3

(0

X

roE

<

in CO

COo

CM

CO

s

CM

8 8

COo

CM8

p

cT

CO

CO

to

9 CM

00

in" in

o*

cm" CO

CM

CM

CO

CM

r-m

In" CA

CO

oCM

CO

IT)

C

o

«

UJeo

re t

3

CO

t_

re

CO

V)

v

oI

"5

a.o

re Z

to 5

o

UJ

cO

o

2

II CM

o o o o

in

8]

oto

5? 3

CO 8$ in 2?

8

m¥)C

4) Q

8

co

8

CM

Is

8

18

8

8

CO

8 8

8

8

CM

8 8 8

3

8 8

CM O

in

CO 8

in

c

o

4>c

4) 0_

"D

i

ro T>C

ro

<

o

UJ

UJ

ceo

iCOO

tf

ini

Q

*>Q.

V)<n

01

t

o

'

(0Q.

3

0)

COCM

O)NCO

CD CM s

IS

ih8

CO

CO* CM

COO

to" CO

coin

oCOm

CM8

CM

CO

8

3

8

10

8

<na.m

w

v>

CO

s

8

ro8

_4)OX

"EQ

c

o

4) 0_

co

CDCM

CO

OCMs

COoCM

CMd

8

co

8. 2r

>

UJ

D-

! 0) Q

in

co

in

co

en' coID

to

<n^ o?O

"in

r~" 5m

jto

CO

CO

to

in

O

olO

9

olO

oT

cn

COoCOin

CD

CO

s

CO

to

oCO

in

CMrCM

o>o"

CM

CO

CO

8 8

a:

CD

CM

C?

CM

CO

cn

CDO

COi

o

o

or

5>

cin

o7

co

O)

CM

§

E

£

i

I

CQ

ICO3

•o" <

507
LOST CREEK DAM, UT

CROSSHOLE S-WAVE VELOCITY PROFILE

DOWNSTREAM TOE

DH-95-5 x -2 x -4

Plol Da1«: 7-10-1995

Expanded Scale

5766.6

S761 .6 '

' '

'

' 1 ' ' ' ' 1

LEGEND Uthology

x Shear Way* (S-) — DH-95-5 x -2 x -4

12

3

ALLUVIUM ALLUVIUM

ALLUVIUM

- Qall - Qal - Qal2

4 ALLUVIUM

5 LIMESTONE - Qal - Me

Figure 10. Geophysical Survey

508
compensate for the difference in pressure due to the

increase in elevation above sea level [6]. Once the

BPT data have been adjusted, adjustments to

correlate the data to SPT N,(60) blow counts can be

made [6].

The blow count and bounce pressure chamber data

is recorded using an automated data logger. The

data logger is necessary because the blows are occurring at a rate of 1 blow about every 0.5

seconds, and it allows for the gathering of blows and

bounce chamber pressures in an orderly and

organized fashion. The data acquisition process is not completely automated since an operator is required to enter a "flag" into the data when the

operator observes that the drill string of the BPT has

been advanced one foot. The process for doing this

is to mark graduated increments of 1 foot on the

drill casing with chalk each time a new joint is added. Then these 1 foot increments are easily

discemable by the data logger operator.

Sampling, using the Becker Hammer, is conducted

using the same drill casing or pipe as is used for

BPT. However, a 6.6-inch O.D., opened ended,

crowd-in bit is used for sampling. Samples are obtained by advancing the bit into the material and

introducing compressed air to the bit via the double

walled drill pipe. The compressed air is directed

between the outer and inner walls of the drill pipe to

the bit and the sample is lifted and carried up

through the inner annulus of the drill pipe. At this

point the inner drill pipe, and thus the sample, exits

to a cyclone so that the compressed air does not

scatter the sample and any associated groundwater.

The energy used to transport the sample is dissipated in the cyclone and the sample drops out

of the bottom and is collected in a rubber tub where

it is visually classified and then sampled and bagged

for lab classification.

An AP-1000 Becker Hammer Drill Rig built by

Drill Systems Inc. and owned and operated by Great

West Drilling was used to perform BPT's in 16

holes and drill four Becker Hammer sample holes in

the foundation downstream of Lost Creek Dam.

The locations of the 16 penetration tests, BH95-1 -

16, and four sample holes, BS95-1 - 4, are shown

on figure 4.

6.1 Adjustments of Raw Blow Counts, BPT 1995.

As a check to the data logger, blow counts were also

manually counted and recorded on data sheets for

later comparisons to the record of the data logger.

An example ofthis data for Becker Penetration Test

Hole BH-95-8 is shown in figure 11. In addition, a pressure gauge was connected to the bounce

chamber and maximum bounce chamber pressures

were observed and recorded as a check against the

data logger. From this data it was found that errors

in the number of blow counts were introduced into

the data before commencing drilling at a new hole

and when a new drill casing joint was added. These

errors appeared to have been introduced into the

data logger record when pressure reversals in the

bounce chamber occurred during the addition of a newjoint of drill casing or the lowering of the anvil

back over the new joint and entire drill string.

A discussion with the designer of the data logger

was held. These discussion uncovered that

apparently additional errors were introduced when

the diesel hammer "fires" but does not create a pressure reversal of at least 1 psi in the bounce

chamber. Errors in the blow count recorded by the

data logger can also be introduced when the pressure

in the bounce chamber never exceeds the 5 psi

threshold pressure for counting a blow. These

errors were fairly apparent due to the lower

pressures observed in the data record (see fig. 1 1).

Manually recorded data which had a larger number

of blow counts than those recorded by the data

logger could be checked by looking at a detailed

time related record of each individual blow. During

times in the procedure when blows were recorded by

the data logger in time intervals greater than .75

seconds, it was assumed that the data logger missed

the number of blow counts it took to make up the

time interval recorded. After making this

adjustment to the data logger data, the total number

of blow counts was typically in agreement with the

number of blow counts that had been recorded

manually.

If errors of the types described above were

observed, the data was corrected to one of the

manually observed blow counts or an average

thereof, whichever was considered more appropriate

based on the preceding and succeeding interval blow

counts.

509
BH95-8

Depth FL^valinn TVhdf lUol 1 ivsanuai uataiog Manual

5829.1 Blow Cts. Blow Cts Blow Cts Otff. > 5 rlOSSUIV Adjust. N./601

1 5828.1 122 120 114 2 7 7 28

2 5827.1 33 32 32 7.21 6 15 3 5826.1 5 11 13 1 &2 5.202 5 5

4 5825.1 11 14 12 5.86 6 8

S 5824.1 21 21 21 6.098 6 11 6 5823.1 18 16 16 5596 6 10

7 5822.1 10 11 12 5.166 5 7 8 5821.1 14 15 14 5.609 6 8 9 5820.1 20 22 22 6.371 7 10 10 5819.1 10 11 12 5.717 6 7

11 5818.1 1 11 9 142 5.056 4 10* 6 12 5817.1 0 9 9 142 0 7 9 ip 0

13 5816.1 4 10 11 142 5.726 7 10* 7 14 5815.1 44 44 41 8.8 9 14 15 5814.1 52 52 51 9.34 9 17 16 5813.1 31 32 34 9.44 9 13 17 5812.1 39 38 36 9.43 9 14 18 5811.1 27 27 28 9.61 8 11 19 5810.1 22 23 24 9.52 10 11 20 5809.1 41 18 19 142 6.933 9 YES 18* 8 21 5808.1 20 20 19 9.3 9 10

22 5807.1 24 24 27 9.25 10 10 23 5806.1 29 j 29 30 9.59 10 10 24 5805.1 63 53 40 142 10.08 11 40* 15 25 5804.1 60 53 46 142 10.44 11 46* 15 26 5803.1 26 43 142 10.81 11 43* 17 27 5802.1 8 45 39 142 10.76 11 39* 16 28 5801.1 46 40 38 14 2 10.66 \\ 15 29 5800.1 32 39 29 1 10.3 11 12 30 5799.1 30 26 26 9.11 10 10

31 5798.1 33 34 34 10.61 10 14 32 5797.1 43 44 44 11.08 12 15 33 5796.1 63 62 63 11.41 12 22 34 5795.1 65 57 57 142 12.25 13 25 35 5794.1 59 66 62 1 12.44 13 24 36 5793.1 71 67 65 2 12.88 13 27 37 5792.1 68 71 69 13.2 14 27 38 5791.1 60 61 64 12.52 13 23 39 5790.1 62 62 61 13.49 14 26 40 5789.1 75 69 69 142 12.1 14 22

41 5788.1 74 75 71 13.11 14 28 42 5787.1 80 79 73 2 13.66 14 29 43 5786.1 78 79 84 2 14.25 15 30 44 5785.1 98 98 99 15.18 16 35 45 5784.1 235 119 116 142 16.07 17 118* 45 46 5783.1 7 100 103 142 1559 16 101* 35 47 5782.1 47 70 67 142 15.44 16 69* 29 48 5781.1 89 88 91 16.06 17 35 49 5780.1 139 140 141 16.77 18 4S 50 5779.1 1 187 174 172 1 4 2 14.3 16 173 „ 44

51 5778 1 108 105 106 14.77 16 36 52 5777.1 84 83 80 15.48 17 31 53 5776.1 104 107 110 2 1655 17 40 54 5775.1 146 147 145 16.88 17 51 55 5774.1 319 312 274 142 18.27 20 116 56 5773.1 203 203 281 2 19.43 20 88

Ip • low pressure reversal below the threshold pressure detected by the data logger ob - operator blunder related to the operation of the data logger |

I

nj - new joint of drill casing creating additional blows related to the addition of the casing

Figure 1 1 . Corrections to Raw Becker Penetration Data - Lost Creek Dam

510
During the tests there were several occasions where

operator blunders were made in setting the "foot

mark flag" in the data logger. On some occasions

this occurred because it was difficult distinguishing

the foot marks on the drill casing. On other

occasions the foot marker was pushed and since

there is a 1 second lag before the data logger display

registers that foot, it was pushed a second time

resulting in an error. These errors only occurred a few times, and when they did a note was made in the

manual record of what had transpired. This

situation was rectified by checking the manual count

and adjusting the logger data based on that count

and the blow counts of the preceding and succeeding

soil interval blow counts. All of the adjustments

discussed above have been denoted in a sample of

those tables in figure 1 1 based on the reason for

adjustment.

6.2 Adjustment to Normalized Blow Counts, BPT

1995.

As discussed above, BPT blow counts are adjusted

to normalized BPT blow counts to compensate for

the effects of changes in bounce chamber pressures

and to account for the pressure difference introduced

by performing a test at an elevation other than sea

level. BPT blow counts must be converted to an

equivalent blow count under atmospheric pressures

at sea level since all of the data related to correlating

BPTs to SPT have been developed based on this

condition. A spreadsheet was set up to perform this

correction. An example of this spreadsheet, for test hole BH95-8, is shown in figure 12. The correction

is performed based on empirical data developed by

Seed and Harder [6].

6.3 Adjustment to N,(60) Blow Counts, BPT 1995.

Once the correction for the difference in

atmospheric pressure from sea level has been

performed, the BPTs are correlated to SPTs in the

spreadsheet based on procedures developed by Seed

and Harder [6]. After the BPTs have been

converted to SPTs the same process of adjusting

blow counts as previously referenced in the

discussion on SPT is used to arrive at an N,(60)

blow count. Since the BPTs are performed with a

closed end bit and yield no information regarding

the material encountered, the typical soil column

discussed in the 1995 SPT exploration program was

used to make adjustments based on effective stress.

6.4 Conclusions Related to the 1995 BPT Program.

Results of the Becker Hammer BPT exploration

program confirmed the existence of potentially

liquefiable layers identified as Qal, and Qal2 within

the foundation of Lost Creek Dam. The program

also obtained credible blow counts for the layer

identified as Qal3 existing in the bottom 30 feet of

the alluvial foundation of Lost Creek Dam. The

adjusted blow counts in the Qal3 reveal the existence

of potentially liquefiable material in this layer.

Sampling with the Becker Hammer yielded

continuous samples of material which were as expected based on previous sampling done during

the SPT explorations. However, there was one

unusual occurrence of a very clean gravel being

recovered in BS95-1 at depth of 38 to 40 feet, which

was not consistent with what had been sampled in

the other SPT or Becker Sample holes. One

drawback to sampling with the Becker Drill is that

material is transported up the inner annulus of the

drill string using high pressure air, and it is therefore

not exactly an in-place sample and is definitely not an undisturbed sample. However, when contrasting

the Becker Drill samples with SPT samples, the

Becker Drill is capable of returning particles over 4

inches in maximum dimension. The SPT is restricted to sampling coarse grained material

smaller than the 1.5 inch diameter opening of the

sampling shoe. After discussions with the Becker

Drill operator and after observing the drill sampling,

it was concluded that the Becker Drill has sufficient

pressure to return almost any material in a timely

manner, and segregation of samples is not a

concern.

7. CONCLUSIONS.

The results ofcorrected N,(60) blow counts from all three exploration programs (SPT 1991, SPT 1995,

and BPT 1995) have been grouped together based

on the close proximity to each other as shown on

figure 4. These blow counts were then tabulated by

elevation, as shown on figure 1 3 . The results of this

study shows that in some cases there is correlation

among blow counts and in other cases there is not. Each of these holes may encounter a large gravel

particle, cobble, or boulder, thus resulting in a high

blow count as discussed previously. The SPTs are probably going to be more sensitive to such an

occurrence, but even the BPT would probably be

511
Yes/ No Nn INo No No Yes Nn!no

YeTl No Nn Yes Nn NnYe^Ye^,Ye^ INn INn 'Nn Nn Nn

CO

CJ o

>- 2

Yes Yes Yes No Yes Yes No No Yes Yes No Yes o2 No Yes Nn No No No No No No No No No No No No No

I

No |

2

CD CO fN

rsi fNCN

en

fN fN

CO <oN

in

rsi infM

inCN

STCN mfM

infMCO

fN

iri CN fN

fN

STfM

to CO CO CO CO CD CO CO CO CD

*Tin

iFmes

CD

cn toCN

fN CO CN

CN

ST CO ro cn ST O ro f~ CN CD CN

fN

1 1

DH9 n

G

CJ Q

5827 9 fN

a

iri fNa

STCN

COm

5822.6 5821.6 5818

i

5814.7 5812.3

CN

a

5810.2 <TD

5802

5801.1 5800

5797.5 5797.4 5793.7 5792.7 5790.7 5788.5 5785

5782.2

Cyclic

stress

ratio

CNOCNOfNofNOCNo

fNOCNO

0.22 0.23

0.241 0.25 0.25 0.26 0.26 0.27 0.28 0.28 0.29 0.29 0.29 0.29 0.290.30! 0.30 0.30 0.30 0.30 0.30 0.30 0.30 0.30 0.31 0.30 0.30 0.31 0.30 0.30 0.30 0.30 0.29 0.29 0.29 0.29 0.28 0.28

0.28

0.28 0.28 0.27| 0.271 0.27

0.270.270.26

0.26|

Tau ?

to

2698- 53

95'

77.80

100.86 123.92 146.99

172.2l[

199

601

224

69

251

80

278

91

306

02 ST

CO

CO

to

356

5? oST(Os

408

83 s

ri

toT

458.46 478

29

502

85

528

85 ro

IN

cna

575

11

601

00 b »ri N

o to

664

37

689

71

707.36

to

STfM

to

fs749

26

774,06 798

86

!

814

60i

639

12

863.65 878.30 892.41 905.98 91900 931.47 943.40 954

79

965

63

975

92

985

67 CO

CO

•CT 8

1003

54

1024

79

1032.631

1053

61

1060

64

1081

35]

1087

55

1093

22

1098

33

i

i

TJCC

0,99

C7> cnd

cncnocn

cno

cn

cno

Oi

cnd

cncno

cno>d 0

98 COcno 098 098 0

98 cno

r~

cno

t^

OloOlo

cno 8

o

8

o

8

o

in

cno

in

cno mOi e d c

T

7> Ol D Ci

to

cno

to

cno

fM

cno

fM

cnO

cnO

OloOlo

0.90i 8

d

8

d I

0.89! i

0

88! !

0.87! i

0

86 \

0.85 0.84 !0.83 0.82 0.81 0.80

CTl dCOdCDd d o

CDOCDOino

sTo

rio

X

roE<

no

0.32 cn

COO

0.32

csi COo

0.32 0.32 0.32 fNCOo0.32 fNCOo

0.32 0.32 fN

roo

oi

COo

0.32 0.32 fM

tod

CM

tod

0.32 0.32 CN

too

fM

COo

fN c CO c d c

N CN

d d

CM

roo

CM

too

fM

roO

fM

too

CM

too

0.32 0.32 I

0.32I

0.321 !

0.32 0.32 0.32 !0.32 0

32 ]

0.32

i

0.32 I0.32 i0.32 !0.32 fM

roO

I

0.32

,

0.32 i0.32 0.32 CNnO

fM

too

fMn

o

Csl toO

CM

roO

Cmno

ro 2

1

5

r—fN

ST m CO o o> CD t-s- o CD m o CO fN ST s— Ol Ol CD CO CO CO CN < n en fN fM O CO fN ofN

CO ofN

s—CN

CO ofNP-- fN

CN

fN

fN

fN

CDIN

CO

to

CD

IN fN mfN mCO

fN

to to

Cn

fM

CN

CTl fM

CO

to CO

Csl CD

!

to

ro

ofN

CO CD ->T CN o O 8 g

COo

COo o 8 8 901

90I

STO COt

too

fMofMq o o 8!38 8

o

cn

cno

0.99

CO

Old

COOlo

0.98 r^

Olo

IS. Olo

0.97 096 0.96 8

d

0.95 8

d

0.94 0.94 ro

OiO

CO

Olo

0.92 CN

CTl dSio

0.91 0.91 0.90 8

o

CT) COO

cn

CDo

CO

COo

CO

COoCOo o

CN

CD CD fN o o cnO

CTi o

COoCOo o 8 § o

mOToE0I

COl

CMoCMo 5 o 8 c 38 Ol

o>o

CT.

CT. O

CT1 Cn

CO

CT) O

CO

Olo

CO

CTi O

CnO

1-^ cno

cnO

COcnoCO

cno

CO

cno

0.951 m

Olo

STcnosT

CTl O

m

CTi O

n

cnO

CM

CT) O

CM

CT> OCT) OCTl O

CTl O

0.90 oCT) o

CT. COOCT) COo

a:COoCOCOO 0

87I

ID o o o o O o o O o O o o o o o o o o o O o o o O C3 O o o o o o o o o o o o o o o o O o o O o o o O o o o o o o o

<N

CM8 Meas.

1

(N1)60!

Harderl

CO

CN

in m CO o r~ CO o r— CD o rss •c fs. ro ST CO o o o tri u1 1— CD m CN o ST m CN

CN mfM

sTCN CNr—fM

r>

CN

COCN

CN

fN

CO

fN OlIN 8

mCOin

•sr 8

Ol

fM mfO

CO

ST CTsf 8 CO oST in

CD CO

CO

Cn s s s. s in

•ST ST cn

CO 8

fs- CN

rofN o

CN

r~ m CO 8

fs- o 8 CO'

fNo SI 0.97 0.96 0.94 0

93

0.92 0.90 0.89 0.88 0.87 0.66 0.85 0.84 0.83 0.82 0.81 0.81 0.80 0.79 0.78 0.77 0.77 0.76 0.75 j0.75 0.74 0.73 0.73 0.72 0.72 0.71 o

evation:

i

!

5829.1! |

SIGV ecoa.

CO sfN I

377.8| 489.8| 601.8| 713.81 836.3|

925.621 996.221

1066.821 1137.421 1208.021 1278.621 1349.221 1419.821

fN

•ST sST

1544.021 1604.621 1665.221 1725.82] 1793.621 1862.221 1930.821

1999.42 2136.62

2205.22!

2273.82I

2342.421 2411.021 2479.62

CNCM

COsCN

2616.821 2685.421 2754.021 2822.62| 2891.221

^2959.82! L3028.42I 3097.02![3165.621 !

3234.22! !3302.82! I

3371.421 s

s

CO

3508.62! [

3577.22! j3645.82! I

3714.421 i

3783.02] ]

3851.621 !3920.22! |3988.82

i

4057.42! !4126.021 i4194.621

i

Water

Surface

El

Hole: |

PORE

jPRESS.

I

o

o o o o o o o

43.68|

106.081

CO

-ST s

230.881 293.28| 355.68| 418

08|

480.481 542.

88

605.281 667.68] 730.081 792

48|

854.88| 917.28| 979.68|

1042

08

1166

88

!

1229

28|

I

1291.681

1354

081

i

1416

48| S

CO I

1541.281 |

1603.681

190

9991 1j

1728.481 |

1790.881 1

1853.281 ]1915.681 ]1978.081 ]

2040.48| ]

2102.881 !2165.281 I2227.68! I

2290.081 I

2352.48! 8

•sr STCN |

2477.28! |

2539.68

|

2602.081

COT

•st' 8CN |

2726

88|

|

2789

28]

|

2851.68| |

2914

08]

|

2976.48] I

3038

88

i

Top

ol

|

TOTAL

[STRESS otoQ.

eo

fN

OO

fN j

377.8|

!

489.8I

601.

8

713.81 836

3|

969.3|

1102.31 1235.3| 1368.3| 1501.3] 1634.3| 1767.3|

ro

2026.3

2149.3] 2272.3] 2395.3] 25183| 2648.5] 2779.5| 29105] 3041.5 33035

3434.5I

3565.51 3696.5| |

3827.5| 3958

5

4089.5] 4220.5| 4351.5| 4482.5|

!

4613.51 m

t"

•ST rST

4875.5] i5006.51 ]

5137.5]

i

5268.5] |

5399.5 (

!

5530.5! I5661.5]

:

5792.5] |

5923.5I

I

6054.5, I

6185.5 I6316.5 16447.5 |6578.5 I6709.5 |6840.5 |

6971.5 I7102.5 I

7233

5

ation

Test

|N60 I

(HARDER

I&SEEO)

i

1

17.301 9.18 3.26|

5.29 6.77

6.28| a

•<t 1

5.78

i

7.26 j5.29 4.79 0.14 5.29 11.99| s

sT

11.06| 11.99| 10.13| 10.13|

]

7.26 9.18 10.13| 10.13| cn f- CD if V u

17.30 16.44] 16.44| 12.901

!

11.06| 15.57|

!

17.30] 24.61| o

oi

IN 8

f-sJ CN

31.25|

1

32.64| 27.63|

1

31.251 Ol

CD

CD

(N

34.69]

I

36.031 38.00! 44.88

CO

^-a

CO

ST

iri T

I

38.64 8

CD

sT |

64.42

!

59.61

I

49.65

CO

STCN

]

55.50

f

70.72

]

162.68

!

125

33

Peneti NBC

Final PCJ

CD Ol CO m m

CD

CO m

T m

iri m o o fN m CN o o r-. Ol o o m ct> CO to CD CO r-- CN

CO

CO en 8 S CO s 8 ST CO

ST CDsT m 8

COm

Olm 8

fN

CO mCD

COm

mrs. 8

mCD 8

ammer

i

SEEK

D/

1

Oi

or

Factor I

QJ

' 6

OJ CD

CJ

I

= ' ' I

NBC

cnm,o

LICD O) CO in in

CO

CO in

ST min

f* in iri o in CN

!

in fN o o rs

1

Ol

I

o o in ci) CO r- ro ID CO ro

ro m

CD

to COro ro CD

ro 8 ST COCT CO

•ST r^m oQD

COmr-~- Olm

o CN

CD nCD

to

in SO

fs.

I

100 mOD OnO

Location:

Hole: BP

|

El.

Corr.

'S

Q.

10.85

.

11-10 9.51 9.79 9.20 8.69 9.22

10.11

sTCO

CO

8.56

8241.84

9.35 OlfN

13.61 13.73 13.71 13.93 13.82

r-N. f-- d

13.56 13.50 13.90 § s

sT s;

15.34 15.28 15.16

STf~

ST

13.34 O

iri

15.66 16.04 17.03 17.26 17.77| ]

18.15] 17.35| Ol

STCO |

16.86! 18.05] [18.69! 19.39] CO

•CT d

toin 8

d

I

20.79 I

21.52; I

22.35i in

sToi 8o

sTCOo

8

fM \

22.48

CM

sTCN I

25

49

Becker 03z

Closed

CN

CN

eo

CO m T—CN

CO os—sTs— ofN os— o Go o STTfNO roOl

CO fN

(NCN

CO ofN

STCN

OlCN o a

sT K

) to

ST Ol

CO fN

CO 8

ro

CO CO

sT (2 S

Olm s 8 s 2

STr- 8

CO CO

Ol

CO oOl

CO

Ol

CD Olto

CO

rs. s—

COo

•ST COsrO

CO

ST OlCO

roOfM

m

i

in

rs.

7.21 fNofN

tri

5.86 860 9

5.596 5.166 6099 6.371 5.717 5.056 6669 5

726 8.8 sTCO

oi

9.44

;

9-<3 9.61 9.52

6.933

tooi 9.25 9.59

10.08 10.81 10

76i

|

10.66] 10.3

;

?-.11|

,

10.61 11.08 ST

12.25 !

12.44; !

12.88 I

13-2! 1

12.52

Ol

•CT ri CN 13.111 8

ri

14.25 15.18 16.07 15.59 sT

iri |

16.06 16.77 14.3

14.77 15.48

in

incb

16.88 IN

CO

ro

sTOl

i

1

i i

i

i

si

.—

LU

CD

fN

Lo

5827 1

to

rN

com

mCN

CO

in

T

rsi a

ro

fN

COin

fN

fNafN

COin

ofNa

5819 1

5818

1

5817.1 5816

1

5815

1

ST

COn

5813

1

fNa

5811.1 o

a ri

5808.

8in

5806.

5805.1 5804

1

5803.1 CNgin5801.1 5800

1

8m

5798 j1

I

5797.1! j

5796,

!

5795.

sT

Olm :

5793 1

i

5792.1 5791 1

!

5790.1 ]

5789.1!

I

5788.1! !5787.1!

!

5786.1 i5785.1 j

5784

1

j

5783.1; 5782.1; |

5781.1 5780.1 5779.1 5778.1 rs.

Iss m

5776

1

5775

1

T

r~. m I

5773

1

Depth

IN T m <o CO cn o CM to T n CO CO cn O

fN rN

CNfNtofM

sr cr fN O

CDfN fMCOCMcn

fM 8 ro IN

to roCO

sTCO mCO 8

f^- cn a

Ol

CO oCT *r fN

sT COsT

•ST ST mct CO

sT •sr CD

•sr

CT) sr 8 uo

fMm

rom

STin m COm

Figure 12. Adjusted Blow Counts from BPT95-8

512
513
affected and not produce as low a blow count as may have been encountered at the same elevation in an SPT hole.

Based on the ease of operation and performance of

the Becker Hammer at Lost Creek Dam, it appears

that very valuable data was obtained related to the

strengths of foundation materials. The difficulty of

using the Becker Hammer is in correlating that

information with the information obtained from SPT

programs. The data base for relating the shear

strengths of coarse grained materials to Becker

Hammer blows is ever expanding and thus there

appears to be more and more credence given to its use.

7.1 BPT versus SPT

The Becker Hammer Penetration Test has several

advantages to the Standard Penetration Test.

First the BPT yields blow counts for every foot

tested, whereas in order to properly perform an SPT

test only one out of every 2- 1/2 feet can actually be

tested. This is because, for continuous testing, the

procedure requires that the first six inches of soil be

the seating interval, the next foot be the actual test interval, and the final foot be the clean out interval

[7]. In addition, the SPT may meet refusal and then

the clean out and seating intervals must be repeated

thus increasing the length of interval for which one

set of blow counts can be obtained.

Second, the BPT can be conducted much more

rapidly than the SPT given that approximately 1 blow every 0.5 seconds can be achieved using the

Becker Hammer. Because of this a testing program

using the Becker can be completed in a more timely

manner and economical fashion because the rig and

personnel to operate the equipment are not deployed

for as great a length of time as is required to perform a SPT program.

Finally, as related to the operations at Lost Creek

Dam, the observation was made that the equipment

used in performing BPT was less susceptible to

mechanical breakdowns than the equipment used in

performing the SPT programs. This is probably due

to the larger size drill string of the Becker (6.6-inchoutside-diameter) as opposed to the SPT with its nominal 2.0-inch-outside-diameter drill string.

One advantage the SPT does have over the Becker,

however, is that the Standard Penetration Testing is performed ahead of a cased holed. By this method

the effects of wall friction as the drill string

advances are quite negligible. In contrast Becker

Penetration uses no casing and wall friction is a

great concern. Harder and Sy have been working

for several years to develop a method for

compensating for this factor [6,8].

8. RECOMMENDATIONS.

A triggering analysis as shown on figure 12 and

similar analyses performed on the BPT data, yields

that the foundation downstream of the dam down to bedrock is susceptible to liquefaction and should be

treated. This analysis was performed using a maximum acceleration, associated with the MCE, of

0.32 g. Methods for treating the foundation at Lost

Creek Dam are discussed in references [4] and [9].

9. ACKNOWLEDGMENTS

The team work and spirit of cooperation

demonstrated by the many people involved in this

job —government, contractor, and consultants —

made it a challenging and rewarding project.

The author would like to thank all those who have

helped make this paper possible. From

Reclamation's Technical Service Center (TSC) in

Denver: Tom Brown, Geotechnical Engineer;

Cletus King, Geologist; Jack Rosenfield, Civil

Engineer/Instrumentation Specialist; Mike Stevens,

Geotechnical Engineer; and Frank McLean,

Reclamation Value Program Manager. From

Reclamation's Upper Colorado Regional Office

(Salt Lake City): Dan Grundvig, Regional

Geologist; and Mark Vandeberg, Geologist

(formerly with Reclamation's Upper Colorado

Region). From Reclamation's Provo Area Office

(Provo, Utah): Gary Dow, Geologist. Jim Benson

of Great West Drilling: owner/operator of the

Becker Hammer.

10. REFERENCES

1. "Results of Geologic Investigations and Data

Evaluations for Lost Creek Dam - Weber Basin

Project - Utah," Report Number G-480, United

States Department of Interior, Bureau of

514
Reclamation, Upper Colorado Region, Salt Lake

City, Utah, October, 1991.

2. "Geologic\Geotechnical Technical Memorandum

- Lost Creek Dam," Patty Ortiz, Technical

Memorandum No. LO-3620-2, Weber Basin

Project, Upper Colorado Region, Prepared by

Woodward-Clyde Consultants, Denver, CO,

WCC Project No. 23205, for the U. S. Bureau

of Reclamation, Denver, CO, Contract

No. 1425-2-CA-81-17610, Delivery Order

No. 3-PD-81-17610-018, August 1993.

3. "Seismic Design Analysis," Design Standard No.

13 - Embankment Design, Chapter 13, U.S.

Department of the Interior, Bureau of

Reclamation, 1989.

4. "Assessment of Corrective Action Alternatives

for Seismic Performance Deficiencies - Lost

Creek Dam," Ronald A. Oaks, Warren J. Danos,

Technical Memorandum No. LO-8314-4,

Bureau of Reclamation, Geotechnical

Engineering and Geology Division, Geotechnical

Engineering and Embankment Dams Branch,

Denver, Colorado, (DRAFT).

5. "Crosshole Seismic Wave Velocity

Measurements for Soil Liquefaction Analysis - Lost Creek Dam - Weber Basin Project - Utah,"

Ulrich Schimschal, Technical Memorandum

No. D8330-95-013, U.S. Department of the

Interior, Bureau of Reclamation, Technical

Service Center, Denver, Colorado, July 21,

1995.

6. "Determination of Penetration Resistance for

Coarse Grained Soils Using the Becker Hammer

Drill," Leslie F. Harder, Jr., H. Bolton Seed,

Report No. UCB/EERC-86/06, Earthquake

Engineering Research Center, College of

Engineering, University of California, Berkeley,

California, May 1986.

7. Earth Manual Part 2. Third Edition, U. S. Department of the Interior, Bureau of

Reclamation, Materials Engineering Branch,

Research Laboratory Services Division, Denver,

Colorado, 1990.

8. Sy, Alexander, "Energy Measurements and

Correlations of the Standard Penetration Test

(SPT) and the Becker Penetration Test (BPT),"

Ph.D. Dissertation, Supervisor: Dr. R_G.

Campanella, University of British Columbia,

Vancouver, British Columbia, Canada, October,

1993.

9. "Technically Viable Corrective Action

Alternatives for Seismic Dam Safety Deficiency

- Upper Colorado Region - Weber Basin Project

- Lost Creek Dam," Decision Memorandum

No. DEC-LO-8314-4, Bureau of Reclamation,

Geotechnical Engineering and Geology Division,

Geotechnical Engineering and Embankment

Dams Branch, Denver, Colorado, April 29, 1995

(DRAFT).

515

U. S. Guidelines for the Seismic Rehabilitation of Buildings

by

Christopher Rojahn1 and Ugo Morelli

2

ABSTRACT

An extensive 6-year project to prepare Guidelines for Seismic Rehabilitation of Existing

Buildings is currently underway in the United

States. Funded by the Federal Emergency Management Agency (FEMA) and prepared by the

Applied Technology Council for the Building

Seismic Safety Council, the Guidelines are intended to serve as a tool for design professionals, a reference document for building regulatory officials, and a foundation for the future

development and implementation of building

code provisions and standards. The Guidelines

will enable users, on a voluntary basis, to re- habilitate buildings to meet specified perform- ance levels such as Collapse Prevention, Life

Safety, and Immediate Occupancy, and performance ranges such as Limited Safety and

Damage Control, for specifically defined ground

motion levels. The Guidelines contain significant new features, including a "Simplified Rehabilitation" method, which can be applied to

certain simple, regular buildings that do not ex- ceed specified height restrictions; a "Systematic

Rehabilitation" method (the main body of the

Guidelines), which can be applied to any type of

building at any location; new Linear Static and

Nonlinear Static analysis procedures; procedures for determining acceptability of existing

components/elements; and procedures for incorporating emerging technologies. Seismic re- habilitation guidance is also provided for nonstructural components and systems. As of October 1995, the Guidelines were 75% complete.

The expected completion date for the Guidelines

project is September 1997

KEYWORDS: seismic rehabilitation; seismic

retrofit; seismic strengthening; building standards; seismic performance levels.

1. INTRODUCTION

Model codes and standards for the seismic design of new buildings have been adopted and

enforced throughout the United States for decades. To date, however, only a few communities

in the United States, such as the City of Los

Angeles, have adopted regulations for the seismic strengthening of existing buildings. In those

instances where regulations have been adopted,

they have focused on one or two of the most

vulnerable local building types (e.g., unreinforced masonry) and have been limited primarily to "risk reduction" measures, as opposed to

higher performance levels, such as life safety.

One of the primary barriers to widespread implementation of seismic strengthening/

upgrading in the United States has been the ab- sence of consensus-backed, professionally ac- cepted, nationally applicable standards for the

seismic rehabilitation (retrofit/strengthening) of

existing buildings.

In 1989, as a capstone project in a major, multiyear Federal program for mitigating earthquake

hazards of existing (non-Federal) buildings, the

Federal Emergency Management Agency

(FEMA) initiated a two-phase effort to prepare a

set of technically sound, nationally applicable

guidelines for the seismic rehabilitation of

buildings. Phase I, conducted by the Applied

Technology Council (ATC) and completed in

1992, focused on the identification and resolution of a wide range of issues that should be

considered during the actual preparation of

guidelines in Phase II. The results of the Phase I project, published in FEMA Report 237, Seismic

Rehabilitation of Buildings - Phase I, Issues

Identification and Resolution (ATC, 1992), es- tablished the initial conceptual decisions for the

Phase II guidelines writers and provided a comprehensive set of recommendations intended to

guide, but not bind, the writers. Among the nu- merous Phase I recommendations was the rec- ommendation that the guidelines provide per-

'Applied Technology Council, 555 Twin Dolphin Drive, Suite 550, Redwood City, California, 94065

2Federal Emergency Management Agency, 500

C Street, S.W., Washington, D.C. 20472

517
formance based design procedures that would

enable designers to achieve specific levels of

building performance, such as life safety and

damage control, for given levels of ground

shaking.

The Phase II effort commenced in 1992 with the

establishment of a five-year, $8-million Cooperative Agreement (No. EMW-91-K-3602) between FEMA and the National Institute of

Building Sciences (NIBS) for preparation of

Guidelines for the Seismic Rehabilitation of

Buildings (Guidelines Project). NIBS, through

its Building Seismic Safety Council (BSSC),

has overall management responsibility for the

Guidelines Project as well as direct technical

responsibility for performance of a consensus review of the Guidelines and several other specific tasks. ATC, as a subcontractor to BSSC,

has responsibility for preparing the Guidelines

and related companion documents and for revising the documents as appropriate following the consensus review process. The American Society of Civil Engineers (ASCE), as another subcontractor to BSSC, has responsibility for con- ducting Users Workshops, capturing relevant

research results, and resolving issues raised (but

not resolved) in Phase I.

A team of approximately 60 consultants

(practitioners and researchers) from various re- gions of the United States have been engaged by

ATC to develop the Guidelines (ATC-33 project). End users are participating in ASCEconducted "Users Workshops" at the 25%, 50%,

and 75% complete stages to evaluate and comment on the utility and user-friendliness of the

Guidelines. Approximately one year prior to completion, which is scheduled for September

1997, the Guidelines will undergo a consensus review by the BSSC.

The Guidelines are based on current research

knowledge and practice methods. Although no

new research is being conducted to support the

development process, an effort has been undertaken to identify and summarize relevant re- search results for use by the Guidelines writers.

In addition, the new analysis methods developed

in the project are being evaluated in case studies

of 25 buildings damaged by the January 17,

1994 Northridge, California, earthquake.

2. PURPOSE

The Guidelines for Seismic Rehabilitation of

Buildings (ATC, 1995) are intended to serve as a tool for design professionals, a reference

document for building regulatory officials, and a

foundation for the future development and implementation of building code provisions and

standards. The Guidelines do not presume that

rehabilitation will necessarily be mandated by a

state or local code; rather, they are intended to be of use to owners and their design professionals for voluntary rehabilitation efforts.

The document is intended to be used primarily

by the technical community responsible for developing and using building codes and standards

and for carrying out the design and analysis of

buildings. This primary user group includes ar- chitects, engineers, and building officials. Parts

of the Guidelines and companion Commentary

will also be useful and informative beyond the

primary technical audience to building owners,

government agencies, and policy makers.

The engineering expertise of a design professional (architect, engineer, code official) is a

prerequisite to the appropriate use of the Guidelines, and most of the provisions of the document presume the expertise of a structural engi- neer experienced in building design.

The document is neither a code nor a standard.

It is intended to be suitable both for voluntary use by owners and design professionals as well

as for adaptation and adoption into model

building codes and standards.

3. SIGNIFICANT NEW FEATURES

The 75% complete document contains several

new features that depart significantly from previous seismic codes.

1. Criteria and methods for achieving

various performance levels at various

seismicity (ground shaking) levels, defined regionally in terms of recurrence

intervals. Performance levels and ranges

considered include: Limited Safety;

Collapse Prevention; Life Safety; Damage Control; and Immediate Occupancy

(see Figure 1). The combination of these

parameters (performance level and ex- 518
pected level of ground shaking) enable

the user to seek one of the following re- habilitation objectives:

• Basic Safety Objective : Life Safety

for ground shaking with a 10%

probability of exceedance in 50

years and Collapse Prevention for

ground shaking with a 2% probability of exceedance in 50 years.

• Limited Objectives : rehabilitation

objectives less stringent than the

Basic Safety Objective, such as performance levels less than life safety

(Limited Safety) for ground shaking

with a 10% probability of exceed- ance in 50 years, or Life Safety for a ground shaking with a higher than

10% (e.g., 50%) probability of ex- ceedance in 50 years.

• Enhanced Objectives : rehabilitation

objectives that call for a higher

level of performance than the Basic

Safety Objective, such as Immediate Occupancy, or Damage Control,

for ground shaking with a 10%

probability of exceedance in 50

years, or Life Safety for ground

shaking with less than 10% (e.g.,

5%) probability of exceedance in 50

years and "Collapse Prevention" for

ground shaking with less than 2%

probability of exceedance in 50

years.

Methods for "Simplified Rehabilitation", applicable to small regular buildings primarily, although not exclusively,

in areas of low and moderate seismicity

and specified as a function of model

building type, ground shaking amplitude

(seismicity level), and desired perform- ance level (applicable to the perform- ance levels/ranges of Limited Safety,

and, in some cases, Life Safety).

Methods for "Systematic Rehabilitation", i.e., complete procedures for

considering all elements necessary to

reach a specified performance level for

any building. These procedures address

rehabilitation strategies, analysis procedures and their selection, and material

acceptance criteria.

4. New methods of analysis, including (a) a new Linear Static Procedure (LSP)

based on estimating maximum structural displacements, and (b) a new Nonlinear Static Procedure (NSP) (i.e.,

"Pushover" Analysis) that will better

account for structural behavior caused

by progressive yielding of structural

elements.

5. Detailed procedures for determining acceptability of existing components and

elements as well as designing new ones

in the four major material groups:

wood, concrete, steel, and masonry.

These depart from standard design procedures for new buildings, which generally consider all parts of the structure

to have similar characteristics.

6. Procedures for incorporating emerging

technologies in the rehabilitation design, including seismic isolation and

supplemental damping.

7. Criteria for seismic rehabilitation of

nonstructural elements, coupled with

various performance objectives and emphasizing those components important

for the Basic Safety Objective. Components are divided between those that affect response and those that do not. Both prescriptive and analysis procedures are included.

4. SCOPE

The Guidelines can be applied to all buildings

regardless of importance, occupancy, historic

features, size, or other characteristics, that by

some criteria are deficient in their ability to resist the effects of earthquake shaking. In addition to the direct effects of ground shaking, the

document also includes consideration of the effects on buildings of local ground failure such

as liquefaction. Excepted are non-building

structures such as elevated tanks or billboards

unless they are part of a building. Also excepted

are buildings or structures covered by special

codes or standards such as bridges and nuclear

power plants.

519
The document also applies both to the overall

structural system of a building, its elements, for

example, shear walls or frames, and the constituent components of elements such as a column in a frame or a boundary member in a wall.

It also applies to nonstructural components of

existing buildings—ceilings partitions, and mechanical/electrical systems. Rehabilitation

techniques for reducing seismic demand, e.g., the introduction of isolation or damping devices,

in addition to increasing strength and ductility

of systems, are also included. And, although the

document is not intended to address the design

of new buildings, it does cover new components

or elements added to existing buildings.

While the Guidelines provide detailed engineering guidance on how to conduct seismic rehabilitation analysis once the decision to rehabilitate a building has been made and while the

Guidelines could be used to conduct seismic

evaluations, guidance on initial decisionmaking

as to whether or not to undertake a rehabilitation

project is beyond the scope of the Guidelines.

Secondly, the question of when the Guidelines

should be applicable in a mandatory way to a remodeling or structural rehabilitation project is also beyond the scope of the document. Finally,

methods of reducing seismic risk that do not

physically change the building, such as reducing

the number of occupants, are not covered.

Also beyond the scope of this document are rec- ommendations regarding what Performance

Level and what severity and probability of

shaking should be selected for Rehabilitation of

any particular building, although a risk traditionally considered acceptable in the United

States is central to this document and is termed

the Basic Safety Objective (see significant new

features above). Higher and lower objectives can also be defined by the user. The Guidelines

were written under the premise that greater

flexibility is required in seismic rehabilitation

than in the design of new buildings. Once the

rehabilitation objective is established, the

Guidelines provide internally consistent procedures that include analysis and design specifications necessary to meet the rehabilitation objective.

Special or new mapping of expected seismic

ground shaking for the United States has not

been developed for the Guidelines. Although

ground shaking can be defined for any probability of occurrence, it is intended that the

seismic hazard used with this document be consistent with national mapping contained in the

National Earthquake Hazards Reduction Program (NEHRP) Provisions for new buildings.

In the 1994 NEHRP edition (FEMA, 1995) national maps are available for ground shaking

with a 10% chance of exceedance in 50 years, a 5% chance of exceedance in 50 years, and a 2%

chance of exceedance in 50 years. These probabilities correspond to shaking that is expected

to occur, on average, about once every 500 (or

more exactly, 474 years), 1000 and 2500 years.

Presently under consideration for the 1997 edition of the NEHRP Provisions for new buildings

are maps that provide response spectra ordinates

(at specified periods) for "Maximum Considered Earthquake" ground motions. These maps

specify ground motions with a 2% probability of

exceedance in 50 years that are constrained in

certain regions with caps and floors developed on the basis of engineering judgment.

Also featured in the Guidelines are descriptions

of damage states with relation to specific Performance Levels. These descriptions are intended to aid design professionals and owners

when selecting appropriate Performance Levels

for rehabilitation design and in their judgments

concerning design of rehabilitation measures.

5. CONTENTS OVERVIEW

While the Guidelines contain information that

would not be included in seismic codes for new

buildings, the Guidelines are organized much

like codes for new buildings (see Figure 2).

Following are descriptive summaries of the

contents of the 75% complete draft.

Chapter 1 provides an overview of the Guidelines, its relationship to other documents, the use of the Guidelines in the seismic rehabilitation process, and the use of the Guidelines for

local or direct risk mitigation programs.

General requirements for rehabilitating existing

buildings are presented in Chapter 2. The

framework in which these requirements are specified is purposely broad in order to accommodate rehabilitating buildings: (1) of many

different types, (2) to satisfy a broad range of

performance goals, and (3) to include consid520
eration of the variations of seismic hazards

throughout the United States. Criteria for eleven

general issues regarding the seismic rehabilitation of buildings are included:

• Rehabilitation (Risk Reduction) Objectives . Selection of desired performance

level for given earthquake severity levels.

• Performance Levels defining the ex- pected behavior of the building in the

design earthquake(s) in terms of limiting levels of damage.

• Seismic Hazard. Determination of the

design ground shaking and other site

hazards, such as landsliding, liquefaction, settlement.

• As Built Characteristics . Determination

of the basic construction and earthquake

resistive capacity of the existing building.

• Rehabilitation Methods. Selection between simplified approaches and systematic approaches.

• Rehabilitation Strategies. Selection of a

basic approach for rehabilitation, e.g., provide additional lateral load carrying

elements, base isolation.

• Analysis and Design Procedures . For

systematic rehabilitation approaches,

selection among linear static, linear dynamic, or nonlinear methods of analysis.

• General Analysis and Design Requirements. Specification of the force actions

for which given components of a building must be evaluated, and minimum design criteria for interconnection

of structural components.

• Building Interaction . Guidelines for

buildings that share elements with

neighboring structures, and buildings

with performance affected by the pres- ence of adjacent structures.

• Quality Assurance. Guidelines for assur- ing that the design intent is appropriately represented by the rehabilitation

construction.

• Alternative Materials and Methods. Guidelines for evaluating and designing

structural components not specifically

covered by other sections of the Guidelines.

Chapter 3 provides the analysis and design re- quirements for seismic rehabilitation of existing

buildings. Four alternative analysis procedures

are presented: (1) Linear Static Procedure; (2)

Linear Dynamic Procedure; (3) Nonlinear Static

Procedure; and (4) Nonlinear Dynamic Procedure. The limitation with regard to the use of

these procedures are provided in Chapter 2. The

Linear Static Procedure (LSP), which provides an approximate solution for the first mode dynamic response of a building to earthquake

ground motions and uses a newly formulated

equation to estimate base shear, is the preferred

procedure for use with buildings of wood frame

construction. Both the LSP and the Linear Dynamic Procedure (LDP), which is performed

using specified ground motion criteria and ac- cepted principles of structural dynamics, may be

used for any of the recommended rehabilitation

strategies except strategies incorporating the use

of supplemental energy dissipation systems and

some types of base isolation systems. Component actions resulting from application of the

linear analysis procedures are categorized as being either force controlled or displacement

controlled and compared against allowable capacities to determine acceptability. Linear

analysis procedures (LSP and LDP) should not

be used for buildings with highly irregular

structural systems, unless the building is capable

of responding to the design earthquake(s) in a

nearly elastic manner and the earthquake ductility demands on the building are suitably low

(guidelines for computing and evaluating the

distribution of inelastic demands are provided in

Chapter 2).

The Nonlinear Static Procedure (NSP) of

Chapter 3 is a "push-over" analysis procedure

whereby lateral loads in patterns that represent

approximately the relative inertia forces generated at locations of substantial mass are applied

to a two or three dimensional analytical model

521
of the building. This model is deflected to displacements that are larger than the maximum

displacements expected in the design earthquake, and the resulting demands are then compared with available capacities. The Nonlinear

Dynamic Procedure (NDP) is performed using

specified ground motion criteria, accepted principles of structural dynamics, and computer

software applications (as they become available). In the NDP, the force and inelastic deformation demands in the design earthquake(s) are

also computed explicitly for each component of

the structure, and acceptability is based on a comparison of computed demands and available

capacities at the performance level(s) of interest. The NDP should be considered for a struc- ture in which higher mode effects are important.

Nonlinear analysis procedures (NSP and NDP)

may be used for any structure and any rehabilitation strategy with the following exceptions

and limitations:

• When sufficient information is available

on the probable strengths of materials

and configurations of components (as

specified in Chapter 2).

• When the NDP is used, the project

should be subject to review by an independent, third-party professional engi- neer with substantial experience in

seismic design.

Chapter 4 presents the Simplified Rehabilitation

Method, which is intended for use only on a se- lected group of simple buildings being rehabilitated to a limited safety design objective (i.e., to a performance level less than the Basic Safety

Objective). The Simplified Rehabilitation

Method recognizes that the performance of certain common building types that meet specific

limitations on height and regularity can be substantially improved by simply eliminating each

of the deficiencies found using the NEHRP

Handbook for the Seismic Evaluation of Existing Buildings (FEMA, 1992). In addition to the

NEHRP Handbook deficiencies, the Guidelines

define eight newly identified additional deficiencies that must be considered in the Simplified Rehabilitation Method. The design ground

motion criteria in the Simplified Rehabilitation

Method is the same as that specified in the

NEHRP Handbook: 67% to 85% (depending on

building period) of the ground motions having a

less than 10% chance of exceedance in 50 years.

In high seismic zones, the Simplified Rehabilitation Method is limited to buildings no taller

than 2 or 3 stories, depending on building type;

in moderate zones, it is limited to buildings no

taller than 2 to 4 stories, depending on building

type; and in low seismic zones, it is limited to

buildings no taller than 2 to 6 stories, depending

on building type. Moreover, in certain locations,

the Simplified Rehabilitation Method can not be

used on certain types of steel and concrete

frame buildings with masonry infill walls and

on certain types of precast concrete frame

buildings.

Chapters 5 through 8 describe rehabilitation measures for components and elements, respectively, for the four major material groups: steel,

concrete, masonry and wood. Included in these

chapters are sections that provide (1) perspective on historical variations in material properties and connection details; (2) material properties for new and existing components; (3)

evaluation methods for in-place materials; (4)

attributes of typical system elements; and (5)

stiffness and strength acceptance criteria for

structural components, including foundation

components. The acceptance criteria are ex- pressed in terms of component demand modifiers that have been developed on the basis of

available research data on the cyclic, forcedeformation relationships of materials. The demand modifiers are used in conjunction with

acceptance criteria equations and specifications

provided in Chapter 3.

Guidelines for the application of seismic protective systems to building rehabilitation are provided in Chapter 9. Specific guidance is

provided for seismic (base) isolation systems

and for passive energy dissipation systems.

Additional, limited guidance is provided for

other special seismic systems, including active

control systems, hybrid active and passive systems, and tuned mass and liquid dampers. The

guidelines in Chapter 9 recognize that seismic

isolation and energy dissipation systems include

a wide variety of concepts and devices. In most

cases, these systems will be implemented with

some additional conventional strengthening of

the structure and, in all cases, require evaluation

of existing building elements. Chapter 9 supplements the guidelines of other chapters with

additional criteria and methods of analysis that

522
are appropriate for buildings rehabilitated with

seismic isolation systems and/or energy dissipation devices.

Chapter 10 provides geotechnical engineering

guidance regarding building foundation and

seismic-geologic sites hazards. Included are guidelines for establishing site soil characteristics and for identifying the following site

geotechnical hazards: fault rupture, liquefaction,

differential compaction, landslide and rock fall,

and flooding. Also provided are techniques for

mitigating geotechnical site hazards; criteria for

establishing soil strength capacity, stiffness and

soil-structure interaction parameters for making

foundation design evaluations; guidance on how

to estimate soil pressure on retaining walls; and

criteria for enhancing foundation strength and

stiffness.

Chapter 11 provides rehabilitation criteria for

architectural, mechanical, and electrical components and systems that are permanently installed

in buildings, including their supports and at- tachments. Contents introduced into buildings

by owners or occupants are not within the scope

of the Guidelines. This chapter contains a list of

nonstructural components that are subject to the

Guidelines, rehabilitation requirements related

to seismic zone and the collapse prevention,

life-safety and immediate occupancy perform- ance levels; general requirements for acceptance

criteria for acceleration-sensitive and deformation-sensitive components; and analysis methods. Methods for nonstructural rehabilitation,

which can be accomplished through replacement, strengthening, bracing, or attachment, are provided in the Commentary.

6. SUMMARY

An extensive 6-year project to prepare Guidelines for Seismic Rehabilitation of Existing

Buildings is currently underway in the United

States. Funded by the Federal Emergency Management Agency (FEMA) and prepared by the

Applied Technology Council for the Building

Seismic Safety Council, the Guidelines (ATC,

1995) are intended to serve as a tool for design

professionals, a reference document for building

regulatory officials, and a foundation for the

future development and implementation of

building code provisions and standards. The

Guidelines will enable users, on a voluntary basis, to rehabilitate buildings to meet specified

performance levels such as Collapse Prevention,

Life Safety, and Immediate Occupancy, and performance ranges such as Limited Safety and

Damage Control, for specifically defined ground

motion levels. The Guidelines contain significant new features, including a "Simplified Rehabilitation" method, which can be applied to certain simple, regular buildings that do not ex- ceed specified height restrictions; a "Systematic

Rehabilitation" method (the main body of the

Guidelines), which can be applied to any type of

building at any location; new Linear Static and

Nonlinear Static analysis procedures; procedures for determining acceptability of existing

components/elements; and procedures for incorporating emerging technologies. Seismic re- habilitation guidance is also provided for nonstructural components and systems.

7. REFERENCES

1. ATC, 1992, Seismic Rehabilitation of

Buildings - Phase I, Issues Identification

and Resolution, Applied Technology

Council Report ATC-28, Redwood City,

Calif, (published as FEMA Report 237,

1992).

2. ATC, 1995, Guidelines for Seismic Rehabilitation of Buildings (75% Complete

Draft), Applied Technology Council Report

ATC-33.03, Redwood City, California.

3. FEMA, 1995, NEHRP Recommended Provisions for the Development of Seismic

Regulations for New Buildings, prepared by

the Building Seismic Safety Council, published by FEMA, Washington, DC.

4. FEMA, 1992, NEHRP Handbook for the

Seismic Evaluation of Existing Buildings,

prepared by the Building Seismic Safety

Council and published as FEMA Report

178, Washington, DC.

523
Performance Levels and Ranges

Performance Level: the intended postearthquake

condition of a building; a discrete and well-defined

point on a scale measuring how much loss is caused

by earthquake damage; in addition to casualties, loss

may be in terms of property and operational loss.

Performance Range: a range or band of

performance, rather than a discrete level.

Rehabilitation Objective: The combination of a Performance Level or Range with Seismic Demand

Criteria.

higher performance

less loss

4

Operational Level

Back-up utility services

maintain functions; very

little damage

Immediate Occupancy

Level

The building receives a

"green tag" (safe to occupy) inspection rating;

any repairs are minor.

Life Safety Level

Structure remains stable

and has significant reserve capacity;

hazardous nonstructural

damage is controlled.

Collapse

Prevention Level

The building remains

standing, but only barely;

any other damage or loss

is acceptable.

Damage

Control

Range

Limited

Safety

Range

lower performance more loss

Figure 1. Performance Levels and Ranges (75% complete

524

draft)
iaDie oi moments /o complete araiii

Page No.

Preface V

i. Introduction 1-1

2. General Requirements 2-1

J. Analysis and Design 3-1

4. oimpiineo KenaDintation Metnou A 1 4-1

c

J. Steel Materials and Elements JI

6. Concrete Materials and Elements 6-1

/. Masonry Materials and Elements H 1 /-I

8. Wood, Light Steel Framing, and Associated Sheathings 8-1

9. Seismic Isolation and Energy Dissipation 9-1

10. Foundations and Geotechnical Hazards 10-1

11. Architectural, Mechanical, and Electrical Components and Systems 11-1

Glossary A-l

Figure 2. Guidelines Table of Contents (75% complete draft).

525

Northridge Earthquake Responses of Base-isolated

(USC) and Conventionally Designed (Olive View) Hospital Buildings

by

Mehmet K. Celebi*

ABSTRACT

Recorded data during the 17 January 1994

Northridge, California earthquake (M„=6.8)

from the base- isolated University of Southern

California (USC) and conventionally designed

Olive View Hospital (OVH) buildings indicate

that they were subjected to ground motions with

peak accelerations equivalent to or larger than

those postulated for their design. The data

reveals successes of the two different approaches

of design. Both buildings performed well during

the earthquake. The USC building is on rock

and OVH is on alluvial deposit. The isolators of

the USC building experienced non-linear

displacements with peak at 3.5 cm that put them

within 10 % shear strain and effectively

dissipated the energy of incoming motions

attaining equivalent viscous damping of

approximately 10 % of the critical. The

superstructure experienced reduced peakaccelerations as compared to the free-field or the

foundation and did not suffer structural or

contents damage. On the other hand, the

conservatively designed OVH (with high lateral

load resisting capability) experienced peak

horizontal acceleration of 2.31 g at the roof

compared to 0.82 g and 0.91 g, respectively, at the ground floor and free-field. The OVH

suffered limited non-structural damage. This stiff

structure was not affected by the long duration

pulses of this earthquake; however, it was

affected by the resonating frequencies of the site.

In addition to Northridge data, the response of

OVH, recorded during the 1987 Whittier

earthquake indicates that the response

characteristics of the building are significantly

different for the two events - attributable to (a)

possible non-linear behaviour due to minor

structural damage and (b) soil-structure effect.

The study exhibits these differences for the

Northridge earthquake during which the shaking

was significantly stronger than that during the

Whittier earthquake.

KEYWORDS: base-isolation, viscous damping,

hysteretic behavior, response spectra, shear wall,

peak acceleration

1. INTRODUCTION

Studying the responses of structures during

earthquakes constitutes an important element of

earthquake hazard reduction programs in order

to confirm or improve methodologies for design

and analyses of structures. Response records

retrieved by California Division of Mines and

Geology (CDMG) from (1) the base isolated

USC Hospital building in Los Angeles, at 36 km

from the epicenter, and (2) conventionally

designed OVH Building in Sylmar (California),

at 16 km from the epicenter of the (Mf =6.8)

Northridge (California) earthquake of January

17, 1994 are particularly significant (Shakal, et

al. 1994) because both buildings were tested

with input motions having (a) peak acceleration

levels equivalent to or larger than those

postulated in their respective design criteria and

(b) long period pulses typically generated by

near-fault ground motions or surface wave

motions as in the Northridge earthquake

(Scientists of the USGS and SCEC, 1994).

Long-period pulses result in large velocities and

displacements. The effective fundamental period

of an isolated structure is typically 3-5 times

longer than an otherwise fixed base structure.

On the other hand, the critical damping

*USGS (MS977), 345 Middlefield Rd.,

Menlo Park, Ca. 94015

527
percentages of the isolators are typically greater

than 10 %. Even then, how base-isolated

systems respond to such long-period motions has

been a cause for concern. The USC records are

the first set of strong-motion records from a base-isolated structure where the displacements

are large enough to show that the isolators

experienced non-linear displacement excursions

and effectively reduced the peak accelerations

and relative displacements of the superstructure.

The OVH response exhibits possibly the largest

peak accelerations at the input (ground) level

[0.9 lg] and the roof level [2.3 lg]. The OVH

building was designed in 1976 to increased level

of seismic forces as a reaction to the disastrous

fate of its predecessor, the original OVH

building that was severely damaged during the

M,=6.4 San Fernando (California) earthquake of

February 9, 1971 and was later razed. Figure 1 shows the location of the USC hospital and the

OVH building relative to the epicenter of the

Northridge earthquake and significant peak

accelerations recorded at select sites during that

earthquake. The objective of this paper is to assess the performance of both hospital buildings

during the Northridge earthquake. Response

records of OVH building recorded at 45 km

epicentral distance from the (M,=5.9) Whittier

earthquake of October 1, 1987 are also included

in this study as they reveal significant response

characteristics of this unique building (Huang, et

al., 1989). The epicenter of that earthquake is also shown in Figure 1. Further details and

references are provided in Celebi (1996a and

1996b).

2. THE BASE-ISOLATED HOSPITAL

The instrumentation and three dimensional view

of the eight story building is seen in Figure 2.

Its steel superstructure is supported by 149

isolators (34.6 cm high) on continuous concrete

spread footings. The diagonally braced perimeter

frames supported by 67 lead-rubber isolators

(55.9 cm square) are designed to carry the

lateral loads. The internal vertical load carrying

columns are supported by 82 elastomeric

isolators (66.0 cm square). The building is on an

outcrop within 15 km of the Newport-Inglewood

fault. It is designed for a maximum relative

isolator displacement of 26 cm and to ATC-6

[SI] (same as UBC 1988 and 1994) spectrum

with ZPA equal to 0.4 g ~ increased by 20 % to account for near field effects. The maximum

peak accelerations and displacements extracted

from the data, summarized in Table 1, indicate

that the free-field peak in either direction is larger than that of (a) the design level, (b) above

the isolators or the roof - clear evidence that the

isolators were effective in dissipating the

incoming vibrational energy. Response spectra

(5 % damped) of horizontal free-field motions

are compared in Figure 3a with the UBC (SI)

design response spectra scaled to 0.48 g. The

normalized free-field response spectra (5 %

damped) are compared in Figure 3b with that of

the shape of the design response for UBC (or

ATC) SI soil type. In general the shapes of the

spectra of recorded components of motions

appears to be well enveloped by the code

spectrum except for some high frequency

(> 1Hz) bands for which it is exceeded. Figures

3c and d show comparison of 5 % damped

response spectra for the locations at the roof,

above and below isolators and free-field. It is seen that below 0.5 seconds, the free-field

spectra exceeds those at the roof.

Figure 4 shows accelerations at various levels of

the building in both directions, their amplitude

spectra and relative accelerations and

displacements of the isolators. It should be noted

that the roof response increases significantly as compared to the response of the 6th floor due to change in plan above that level. In the NS

direction, the maximum relative displacement is about 3.5 cm while in the EW direction, it is about 3.0 cm (at the south end). The

displacements and relative displacements

correspond to (a) the largest drift ratio of

approximately 10 % of the allowable, (b)

maximum 10 % shear strain (maximum relative

displacement divided by the height of the

isolators), (c) stiffness K2 of the actual hysteretic curve of the isolators (Figure 5) and (d) to hysteretic damping of approximately 10 %

calculated by the relationship d = A(w)/47r(W)

where A(w) is the actual area of the hysteresis

loop at a displacement level and W is the

hypothetical elastic energy defined by the area

528
formed by the line defined by the elastic slope at the same displacement (Hudson, 1965, Jacobsen,

1960). The level of damping, approximately 10

% and 15 %, for the 1st and 2nd modes,

respectively are also extracted by system

identification technique (Qelebi, 1986a). The

first mode dominates the response of the

building. Figure 6 shows cross-spectra,

coherence function and phase angle plots of roof

and 4th. floor, and relative torsional

accelerations between the roof and the lower

level. The family of (linear and non-linear

frequencies attributed to the first and second

modes seen in the cross-spectra are coherent and

are in phase. The first mode frequency starts at

1 Hz and due to nonlinear behavior of the

isolators shifts to 0.75 Hz in the NS direction

and similarly starts with 1 Hz and shifts to 0.62

Hz in the EW direction. These clear peaks are

indicative of the fact that distinctive

instantaneous stiffnesses can be defined for the

isolators as a function of the relative isolator

displacements. Similarly, for the second mode,

the motions are 180 deg out of phase at around

1.5 Hz and shift in the frequences are also

clearly identifiable for that mode.

It is important to note that the full displacement

and dissipation capability of the isolators were

only partially exhausted during the Northridge

earthquake, suggesting that the performance of

the building and its isolators will be satisfactory

during future events at possibly higher

acceleration and displacement levels than those

recorded at the site and experienced by the

structure during the Northridge earthquake and

compatible with the seismicity of the area. A

detailed probabilistic evaluation of the seismic

hazard in Southern California and probable

earthquakes for a thirty year period (1994-2024)

has been recently reassessed (Working Group,

1995). Deliberations on this subject is beyond

the scope of this paper.

However, it is possible to estimate the extreme

situations for the building and its isolators with

the known first mode frequency which starts at

1 Hz and then with nonlinear behavior of the

isolators shifts to 0.7 Hz in the NS direction and

similarly starts at 1 Hz and shifts to .62 Hz in

the EW direction. For this, the stiffnesses for a

bilinear hysteretic curve as defined by Asher et

al. (1990, 1995) are adopted. These stiffnesses as corresponding to 1 inch (2.5 cm)

displacement are: kL1 = 8.55 t/cm (for 59) and

9.08 t/cm (for 9) of the 68 lead-rubber isolators;

and similarly, kL2 =1.98 t/cm for 59 and 2.68

t/cm for 9 lead-rubber bearings. The stiffness of

the high damping rubber (only) bearings are defined to be linear at 1 inch (2.5 cm) and

therefore, kR1 = kR2 =2.41 t/cm for 8 and 3.07

t/cm for 73 of the 81 such bearings. To facilitate

further evaluation at larger displacements, the

author, based on actual hysteresis plots of

prototype isolators, extended these to kL3 =1

t/cm for all lead-rubber bearings and kR3 =1.75

t/cm for all rubber bearings. If the

superstructure can be assumed to be rigid and

since the mass is constant, the fundamental

frequency of the system at any displacement

stage can be defined by ft = (0.5/7r)(Klj /Mlj)°-5

where Kij = 68kLi + 81kRi is the total stiffness

of the (68 lead and 81 rubber) 149 isolators.

Thus, for the first mode only:

Ku = 59(8.55) + 9(9.08) + 8(2.41) + 73(3. 07) t/cm.

Ku = 830 t/cm.,

K12 =59(1.98)+9(2.68) + 8(2.41)+73(3.07)

K12 =384 t/cm. and,

K13 =68(l) + 81(1.75)=210 t/cm.

From these, equivalent mass units corresponding

to the identified frequencies (1 and .7 Hz of the

first mode are determined): Mn = 21 units (for

fu =l Hz and Ku = 830 t/cm.) and M12 = 19.9

units (for f12 =0.7 Hz and K12 =384 t/cm.). From

this, an average equivalent mass is taken as Mij=20 units. Using this, the first mode

frequency at large displacements can be

estimated with K13 . Thus, at large displacements,

f13 ~ (0.5/tt)[210/20]°-5 ~ 0.46 Hz [~ 2.2

seconds]. At 2.2 seconds, the spectral

accelerations, conservatively estimated from the

free-field response spectrum (of Northridge

529
earthquake with known low-frequency content)

to be less than 100 cm/s/s. Assuming even a higher spectral acceleration at 200 cm/s/s (at 2.2

seconds), the displacement of an isolator can

then be estimated to be d - Sa/w2 ~ Sa/(27rf)

2 ~ 200/(2tt[.46])2 ~ 24 cm which is within the

displacement capability of the isolators and less

than the moat clearance.

3. THE OVH BUILDING

Figure 7 shows the instrumentation and three

dimensional view of the six story building and

its associated free-field. The lateral force

resisting system of the building is a mixed

design of concrete and steel shear walls. The

ground floor and second floor are typically 10

inches thick concrete shear walls that extend

along several column lines. At the 3rd level, the

plan of the building changes to a cross shape

making a four story cruciform tower with steel

shear walls surrounding the perimeter. The

foundation of the building consists of spread

footings and grade mat on the ground floor. The

seismic design criteria of the building was based

on two levels of performances defined by the

design and survivability level earthquake spectra

with ZPA of 0.52g and 0.69g respectively. The

survivability earthquake is postulated for a Richter magnitude 8.5 earthquake. The two

design response spectra (for 5% damping) are compared in Figure 8 with response spectra of

the free-field at the OVH grounds. Also shown

in the figure are the response spectra of the freefield (at epicentral distance of 45 km) motions

recorded during the 1987 Whittier earthquake.

Peak accelerations and absolute displacements

for motions recorded at the roof, ground floor,

and free-field, during both events are summarized in Table 2. The maximum drift

reached in the Northridge event was less than

0.2%.

The building is located on an alluvial fan.

Several borehole logs up to 100 m taken in the

immediate vicinity of the building indicate that

the site has approximately 3-5 m of dense to very dense silty sand and gravel, with

underlying alluvial deposits of dense to very

dense sand and silty sand with gravel and

cobbles and that the depth to bedrock is between

52-92 m (see references in Celebi [1986a]). The

thickness of the alluvium decreases towards the

hills (north of the hospital). Using three possible

profiles, site transfer functions corresponding for

two different depths and varying shear velocities

of layered substrata are calculated based on

shear wave propagation method resulting in the

fundamental site frequency (period) of

approximately 2-3 Hz (0.33-0.5 sec) [Figure 9].

Figure 10 shows acceleration time histories for

both the Northridge and Whittier events at the

roof, ground floor and the free-field and

corresponding amplitude spectra and spectral

ratios. In the NS direction, for lower amplitude

shaking during the Whittier earthquake, the

structural frequencies are > 3 Hz and for the

stronger shaking during the Nothridge

earthquake, the frequency cluster is between 2-3

Hz. In the EW direction, the lowest frequency is

at 3 Hz (Northridge) and 3.3-3.6 Hz (Whittier).

The 2 Hz (0.5 sec) is the fundamental site

frequency (period) because the spectral ratio of

roof/ground floor or roof/free-field motions is approximately 1 (the effect of the site frequency

cancels out) or small. Therefore, two effects are occurring: (1) the frequencies (periods) at the

roof are lower (higher) during the strong shaking

and (2) the structural and free-field fundamental

frequencies (periods) are very close to one

another -- evidence that the structure was in resonance between 0.3-0.4 sec (2.5-3.3 Hz)

which is also well within the site period

(frequency) of 0.33-0.5 sec (2-3 Hz) calculated

from the soil log. The building escapes most of

the low frequency (<2 Hz) energy in both

events.

Figure lla,b,d,e show the coherence, phase

angle and cross-spectra (Sxy) plots of the two

(NS) parallel motions at the roof and two

parallel (EW) motions at the roof and 4th floor

of the building for both events. The frequencies

(e.g., for Northridge, 3.0 and 2.5 Hz in the NS

and 3.3 and 2.7 Hz in the EW direction) are

530
close and are in phase. Therefore, the parallel

motions at two wings of the cruciform tower are not torsional but can possibly be attributed to the

fundamental mode of the complete structure

(higher frequency) being different than that of

the wing of the cruciform (lower frequency). To

clarify this effect, an additional sensor should be

deployed in the core of the cruciform. Figures

11c, f are prepared to investigate rocking as a form of SSI. While for the Northridge event,

rocking may have taken place in the NS

direction because of unity coherence and 0°

degree phase angle at the fundamental mode

frequency (2.5 Hz) that clearly peaks out in the

cross-spectra, the same argument cannot be

made for the EW direction motions or for the

Whittier earthquake motions because they have

low coherency and are not in phase at the

appropriate frequencies. Therefore, more

vertical sensors should be deployed on the

ground floor to clarify this effect with certainty.

Damping ratios extracted from system

identification procedures are 10-15 % (NS) and

5-10 % (EW) for Northridge and 1-4 % (NS)

and 5-8 % (EW) for Whittier. These results are

consistent with the direction and level of shaking

during the Northridge or the Whittier

earthquake. The high level of damping ratio

assessed from stronger shaking during the

Northridge earthquake may be attributed to

radiation damping of the incoming motions. This

is further explained elsewhere (Celebi, 1986b).

4. CONCLUSIONS

Both the base-isolated USC hospital and

conventionally and conservatively designed OVH

building performed well during the Northridge

earthquake of 17 January 1994. The USC data is the first set of data from any base-isolated

building that exhibits excursions into the

nonlinear range of the isolators. The drift ratios

experienced by the superstructure is less than 10

% of the allowable which should explain that (a)

there was no damage to the structure or its contents and (b) isolators performed well and

effectively dissipated (reaching 10 % hysteretic

damping) the energy of motions with

acceleration levels equivalent to the design level

accelerations. The effective performance of the

isolators also reduced the drift ratios of the

building.

The OVH Building data shows that with

fundamental mode period (frequencies) between

0.3-0.4 sec (2.5-3.3 Hz) escapes the wrath of

the long period pulses (> 1 sec) characteristic

of the Northridge ground motions. The effective

structural frequencies derived from the

Northridge and Whittier data are different and

exhibit variations attributable to non-linear

effects: (a) soil-structure interaction which is more pronounced during the strong shaking of

the Northridge earthquake. The building possibly

experienced rocking at 2.5 Hz in the NS

direction during the Northridge event and there

is possibility that radiation damping at the

foundation contributed to the response, (b)

non-linear behavior due to minor structural

damage during the Northridge earthquake. It is also likely that the cruciform wings responses

with a different frequency than that of the

overall building. Additional sensors are needed

to distinguish the response of the wings and the core and to clearly verify if rocking occurs.

The designer of the building conceived a very

strong and stiff structure (particularly in

response to the disastrous performance of the

original Olive View Hospital building during the

1971 San Fernando earthquake). However, site

frequencies were not evaluated and therefore

consequencies were not given proper attention in

the development of the design response spectrum

for the new OVMC building. The data shows

that fundamental frequencies of the building are well within the influence of the site frequency

(2-3 Hz) to cause resonance. This case study

indicates that representation of site resonating

frequencies need to be improved in development

of design response spectra. Finally, although the

performance of the OVMC may be considered to be satisfactory during the (MJ =6.8) Northridge

event when the shaking was at the postulated

peak design level accelerations, it is reasonable

531
to ponder; however, what the realistic ground

motions would be and how the building would

perform during motions generated by the

postulated M,= 8.5 design level earthquake as defined by the design criteria of this building.

5. REFERENCES

1. Asher, J., Van Volkinburg, D., Mayes,

R., Kelly, T., Sveinsson, B., and

Hussain, S., 1990, Seismic Isolation of

the USC University Hospital, PROC.

4th. U. S. National Conference on

Earthquake Engineering, Palm Springs,

Ca., May 20-24, 1990, v. 3, pp.529-

538.

2. Asher, J., Hosker, S., Ewing, R.,

Volkinburg, D., Mayes, R., andBatton,

M., 1995, Seismic performance of the

base-isolated USC Hospital in the 1994

Northridge earthquake, ASME/JSME

Joint PVP Confrence, Hawaii, (July

23-27, 1995), v. 319, pp. 147-154.

3. Celebi, M. (1996a). Succesful

performance of base-isolated building

during the 17 January 1994 Northridge

earthquake, [in print]

4. Celebi, M. (1996b). Seismic

performance of the New Olive View

Hospital during the Northridge and

Whittier earthquakes, submitted. ASCE

Journal of Structural Division.

5. Huang, M., T.,Cao, D. Parke, and A.

Shakal. (1989). Processed strong-motion

data from the Whittier, California

earthquake of October 1, 1987 (Part I

- Ground Response Records),

OSMS-89-03, California Department of

Conservation, Sacramento, Ca., July,

1989.

6. Hudson, D. (1965). Equivalent Viscous

Friction for Hysteretic Systems with

Earthquake-like excitations, PROC. III. WCEE, New Zealand.

7. Jacobsen, L. S. (1960). Damping in

Composite Structures, PROC. II. WCEE, Tokyo, Japan, pp. 1029-1040.

8. Shakal, A., M. Huang, R. Darragh, T.

Cao, R. Sherburne, P. Malhotra, C.

Cramer, R. Sydnor, V. Graizer, G.

Maldonado, C. Petersen, and J. Wampole. (1994). CSMIP strong-motion

records from the Northridge, California

earthquake of 17 January 1994, Report

No. OSMS 94-07, California

Department of Conservation,

Sacramento, Ca., February 1994.

9. Scientists of the USGS and SCEC.

(1994). The magnitude 6.7 Northridge,

California, earthquake of 17 January

1994, Science, 266, pp. 389-397,

October 21, 1994.

10. Working Group on the Probabilities of

Future Large Earthquakes in Southern

California, 1995, Seismic hazards in

Southern California: Probable

earthquakes, 1994-2024, Bull. Seism.

Soc. of Am., vol. 85, no. 2, pp. 379-

439, April 1995.

532
Table 1. Peak Accelerations and Displacements of USC Hospital

Location Max. Peak Accel, (g) Max. Peak Displ. (cm)

NS EW UP NS EW UP

Roof .21 .19 3.9 5.1

6th Fl. .11 .15 3.3 4.0

4th Fl. .10 .16 3.1 3.3

Above Isolators .13 .14 .10 2.8 3.0 1.3

Below Isolators .37 .17 .09 1.7 2.3 1.4

Free-Field .49 .22 .12 2.3 2.5 1.3

Table 2. Peak Acceleration and Displacements at OVH

Northridge Eq. Whittier Eq.

Location

A(g)

NS

D(cm) A(g)

EW

D(cm) A(g)

NS

D(cm) A(g)

EW

D(cm)

Roof 2.31 34.1 0.79 19.2 0.20 0.65 0.16 0.79

Ground Floor 0.82 28.3 0.42 18.4 0.06 0.53 0.06 0.56

Free-Field 0.91 32.6 0.61 15.2 0.06 0.55 0.05 0.51

Fig. 1. Location of USC and Olive View Hospital (OVH) relative to epicenters of the Northridge and

Whittier earthquakes and significant peak accelerations recorded during the Northridge earthquake.

533
Fig. 2. General three-dimensional schematic of the USC Hospital Building showing overall dimensions,

the vertical and horizontal unsymmetry and the instrumentation scheme.

Ca) FREE-FIELD VS. UBqSl)

10"1 10° PERIOD (SEC)

Cc3 FREE-FIELD VS. UBqSl)

O

J

a

u

02o

z

10 1 10° PERIOD (SEC)

RESP. SPECTRA (NS) - 5 %

a:rodf

b: above

isolators

Tl 11 1 1 d/V

't'beloiv' ! HIM

isolators

d:£ree-field

490 ^LH,

" 370 UP

m *LJ_LU

a,/

'

who

io-2 io-1 10° PERIOD (SEC)

Cd) RESP. SPECTRA (EW) - 5 %

101

t—r-r-nrm 1—i—ri'ini Ad ....

ild

" 220 /c /

170 s£j

90 /D

I 1 L. l-l LLU

A ;

'

y wv"k*

Mjk.. b\ ,: ...

10 1 10° PERIOD (SEC)

101

Fig. 3. Comparison of (a) free-field spectra with design response spectra, (b) normalized free-field and

design response spectra and (c, d) spectra of motions at different levels.

534
_ 500

l

0

d

-500

<-1000

USC HOSPITAL: NS

ROOF

6TH

I 1S(

ii i

i i Wf* 1

ISOLATORS

FREE-FIELD

500 "ROOF

=r o

-500

10 20 30

TIME (SEC)

USC HOSPITAL: EW

40

6TH

ABOV

BELO

FREE-FIELD

10 20 30 TIME (SEC)40

,x\0* USC HOSPITAL: NS

ROOF

&TH

»-*liS .nrVy-riniil i ir£imj _

FREQ (HZ)

»10« USC HOSPITAL: E

1

1.5

1 At , ^ABOVE ISOLATOR

05 r^'*^

FREQ (HZ)

500

d

as -500

(ABOVE-BELOW) ISOLATORS

fcS:CH9«CH5

+ £W:CH12-CH8

P £W:CHlil-CH7

' £W:CH10-CH6

10 20 30 40

TIME(SEC)

10 20 30

TIME (SEC)

Fig. 4. (a) Accelerations and (b) displacements at various levels and (c, d) their amplitude spectra,

(e, f) relative accelerations and displacement of the isolators.

Fig. 5. Hysteresis loops from the laboratory testing of a prototype lead-rubber isolator.

535
NS:ROOF(CH21)&4TH.FL(CH17) EW:ROOF(CH23)&4TH.FL(CH15) (CH24-CH22)A(CH12-CH10)

FREQ (HZ) FREQ (HZ) FREQ (HZ)

Fig. 6. Cross-spectra, coherence function and phase angle plots motions at roof and 4th floor and

torsional motions at the roof and lower level.

Fig. 7. General three-dimensional schematic of the Olive View Hospital (OVH) Building showing overall

dimensions, the instrumentation scheme and the free-field site location.

536
SYLMAR OLIVE VIEW HOSPITAL SITE: TRANSFER FUNCTION

U

E

2

4

-

3

-

0

0

1 1 1

1 1

1 1 11

!

\ A

A

i

1 1 1 n-1 : i : » : > • :

! ! / i i / \\

: : i /.:.... ,\ : : ,.'.".*„ ?. fij?

c/ i V7 V N \

; i / /H\ \ i V \ ://*. \ :/'

/

: • V / 4 » V //.' V.

Iii - ^-i-e^

J i ':

\ / \ / \ /

1 .' /:

/ ' ' • '• '

'• /// A[h(m),Vs(Km/s)]=[(7,.2).(13,.3)i(13,.5),ai,l)] !

* I ; : ; ; ; : B[h(m),Vs(KnVs)J=[(7,.2).(13,.3);(13,.5),(61,l)] 1

b=[(7,.2).(13,.3)i(14,.5),(30,.6),(27,.8)] j

i i i i i i i 2 3 4 5 6

FREQ (HZ)

8 10

Fig. 8. Site transfer functions of OVH building site.

i

i

CO

1

-

RESPONSE SPECTRA:(DESIGN & SURVTVABIUTY EQ. VS. FREE-FIELD)

I r——r~~T—| I i }

I ! r——

j—i—i i i

j

; -— 1 1

1—i—rff N=NORTHRIDGE

w=wHrrnER

iDA^=5%]

t

i:

i

:

i :

i,-JN;FREE-BELD(NS)

sbfcvrvABnirrY IeoJ

i\ DESIGN EQ

WiFFCNS&EW)

i i A I

: \ —i—*f* erit«hJ.i

10-2 10 1 10°

PERIOD (SEC)

101

Fig. 9. Comparison of design response spectra with response spectra of free-field motions recorded

the OVH site during the Northridge and Whittier earthquakes.

537
3000

ROOF(CH3>

V t* I

WEE-PL(NS)

3 10 IS TIME (SEC) NORTHRipOE (EW)

i.J/V h

HtEQ(HZ)

»10« NORTHRIDOEfEW)

i . A ROOWC

0 J FREQ (HZ)

IORTHRIDOE (EW)

S 10 IS TIME (SEC)

WHITTIER (NS> 0000- ftOOF(CH$

FREQ (HZ)

WHITTIER (N5)

10 20 TIME (SEC)

30

FREQ (HZ)

WHITTIER (N5) ROOP[CH3)/GR(CH9)

FREQ (HZ) WHmTER (EW)FREQ (HZ)

C 20

<

!

ROOF(CHiqyFF il !: | i w

FREQ (HZ) FREQ (HZ)

Fig. 10. Accelerations recorded during the Northridge and Whittier earthquakes at the roof, ground level

and free-field and corresponding amplitude spectra and spectral ratios.

[ * WrTHRIDGE: CH3 A CH2

0.8

0,6

0.4 -

0.2

0. i

1

at

a6

a4

02

o.

o

(4)

0.8

0.6

0.4

0.2

0

NORTHRIDGE: CHIP A CHI 1

FREQ (HZ)

WHTr^nER^CTOACH2

i f

f.L...ii.... . {A

!—»T"

2.7 v

v ; t i. 1 i • L*i

Jv

§\ ' \

i «

L\. A. I

V

-

!

5 JsAy

'

180

FREQ (HZ)

/HITTIER:CH10&CH11

180

180

-180

FREQ (HZ) FREQ (HZ) FREQ (HZ)

Fig. 11. Coherency, phase angle and cross-spectra plots of (a) parallel NS motions at the roof, (b) EW

motions at the roof and 4th floor, and (c) horizontal motion at the roof and vertical motion at ground

floor for the Northridge earthquake, d, e, and f are for the Whittier earthquake.

538
Recent U. S. Experience in Design and Testing of Beam-to-Column

Connections for Moment-Resisting Steel Frames

Ferdinand LoRay Schwartz *

ABSTRACT

This paper presents a compilation of recent

reports describing emergency changes to the

Uniform Building Code and resulting design

information and test data for an improved design

of beam-to-column connections for momentresisting steel building frames. The design of this

special connection or joint was developed for

construction of the U.S. Department of Energy

funded Human Genome Laboratory at the

Lawrence Berkeley National Laboratory.

An earthquake of 6.7 magnitude hit Northridge,

California, on January 17, 1994, and a similar

earthquake that devastated Kobe, Japan, exactly

one year later caused serious, unexpected cracking

in and around the rigidly welded joints of

moment-resisting steel building frames of the

strong column-weak beam design. This

traditional design concept under seismic loading

should result in elastic behavior ofthe column and

plastic behavior of the connected beam. The

discovery of cracks in and around beam-tocolumn joints raised the specter of these failures

leading to partial or complete building collapse.

As a result of the discovery of the cracked joints,

the International Conference of Building Officials

adopted an emergency change to the Uniform

Building Code requiring that new joint designs be

verified by analysis or test to confirm meeting the

required strength criteria. At the time the

emergency code changes were issued, the Human

Genome Laboratory was in the design phase. A

new design of a Special Moment-Resisting Frame

(SMRF) joint was developed and successfully

tested for use as the structural lateral system for

the facility. Since a number of other new designs

failed similar testing, the SMRF joint represents

a breakthrough in the development of beam-tocolumn joints which meet the emergency Uniform

Building Code requirements. The SMRF joint is generally less complex to fabricate and erect than

other new joint designs and is only slightly more

expensive than the "pre-Northridge" weldedjoints

which failed.

KEYWORDS: cracked welds; Northridge; joint

design; moment-resisting; seismic; steel frames.

1. INTRODUCTION

On January 17, 1994, an earthquake of 6.7

magnitude hit Northridge, California, about 20

miles from Los Angeles. Mr. Donald G. Eagling

describes the earthquake-induced structural

failures resulting from this event in the U.S.

Department of Energy's draft Seismic Safety

Manual (Ref. 1) and provides considerable

historical and technical data on structural design,

structural failures, and the new building code

requirements for ensuring resistance to

earthquakes. During the Northridge event,

horizontal ground accelerations of approximately

1 g caused major, widespread damage affecting

more than 12,000 residential and commercial

structures as well as highways and bridges.

Classic damage, including total collapse, occurred

to unreinforced masonry, non-ductile reinforced

concrete, and multistory wood frame structures.

Even some structures which were designed and

constructed in compliance with the then current

Uniform Building Code (UBC) suffered severe damage well beyond what might

* Office of Energy Research, U.S. Department

of Energy, Washington, DC 20585

by

539
be expected as a result of an earthquake of the

magnitude of the Northridge event.

Inspections of beam-to-column joints after the

event disclosed very serious failures in momentresisting steel frame beam-to-column joints of

some structures designed to the UBC. These

unexpected failures manifested themselves as

significant cracking in and around the double

sided, full penetration welded tee connection

between the beam end and the flange of the

supporting column. This "pre-Northridge" joint

design is typical for steel building frames using

the strong column-weak beam approach to

accommodating seismic loads. The column is designed to carry the cyclic lateral loads imposed

by the ground motion without exceeding the

elastic limit for the column size and material. The

beam, however, is supposed to deform plastically

above a predetermined load to prevent over

stressing or fracturing the joint or the column. In

this manner the integrity of the frame is maintained, preventing collapse of a floor or the

entire structure.

An earthquake of 6.8 magnitude devastated Kobe,

Japan, exactly one year after the Northridge event.

This earthquake destroyed about 20,000 buildings

and severely damaged another 35,000. Even

though the two earthquakes were of nearly equal

magnitude, the much greater extent of damage in

Kobe compared to Northridge was due in large

part to the much higher population and building

density in Kobe combined with a large number of

typical wood frame residential structures. Like

Northridge, however, the Kobe event caused the

same unexpected cracking failures in several

buildings with moment-resisting steel frames.

These modern buildings were designed to the

applicable building codes which are very similar

to UBC requirements (the UBC is developed with

international cooperation).

2. EVALUATION OF FAILURES

Investigations in the aftermath of the Northridge

earthquake indicated a number of potential

problems in design and construction of strong

column-weak beam joints including the following:

• The UBC includes both specific minimum

code requirements and general (intent)

provisions incorporating good engineering

practices. Many newer structures were

constructed under "design-build" contracts and

were designed only to meet the minimum

requirements of the code. Due to the

competitive nature of the construction

business, compliance with the general

provisions of the code, which might have

resulted in more conservative designs,

generally was not mandated by the contracts.

There is evidence that had the building designs

also met the intent of the general provisions, at

least some of the failures would not have

occurred.

• Fractures in the welded joints of a 4-story

commercial building were initially attributed

to poor workmanship in welding the

connections. However, review of historic test

data, combined with a series of tests performed

after the earthquake, indicated that even

connections made with standard quality

workmanship were vulnerable to this kind of

damage. However, the review did highlight

the importance of weld quality on the seismic

performance of this kind ofjoint.

• Steel overstrength has become a serious

problem for designing the strong column-weak

beam joint, particularly when different strength

steels are specified for the column and beam.

In general the standard grades of steel

produced exceed, but by a relatively small

margin, the minimum required strength for the

specified grade. The problem arises from the

steel industry's practice of down grading a heat

of steel if its strength properties do not meet

the intended minimum. The result is that the

material significantly exceeds the minimum

strength of the next lower strength grade.

When the design of a joint depends on its predictable inelastic behavior, the use of

540
overstrength beam material may not allow plastic

deformation of the beam at a low enough load to

prevent excessive load being placed on the beamto-column joint.

• The observed cracking failures in and around

the beam-to-column weld were brittle in

nature. This could be attributed to one or a combination of several factors.

- The tendency for the heat-effected zone of

a weld to be more brittle than the

surrounding base metal unless sufficient

preheat and careful post-heat cool-down are employed.

- Work hardening of the joint during the

earthquake induced strain leading to low

cycle fatigue failure.

- Typical beam-to-column joint designs with

full penetration welds between the beam

end and the column flange are so rigid that

the cross sectional area of the weld

interface cannot deform sufficiently under

large bending loads to allow the joint to

remain ductile.

Regardless of the cause, the standard "preNorthridge" welded beam-to-column joint used

for moment-resisting steel frames exhibited a

here-to-fore unnoticed and very dangerous flaw

that exists in innumerable buildings throughout

the world.

3. CODE REVISIONS

Shortly after the discovery ofthe unexpected steel

frame joint failures in the Northridge area, the

International Conference of Building Officials

(ICBO), which publishes the UBC, adopted an emergency change to the code requiring that new

joint designs be verified by analysis or test to

confirm meeting the required strength criteria.

The emergency change requires that new designs

must take into consideration the effects of both

overstrength and strain hardening of the beam

steel which were among the suspected causes of

the earthquake-damaged frame joints.

In addition, the problem is being intensively

studied by the SAC, a joint venture of the

Structural Engineers Association of California

(SEAOC), the Applied Technology Council

(ATC), and California Universities for Research

in Earthquake Engineering (CUREe). Under

emergency authorizations from the California

Office of Emergency Services and the Federal

Emergency Management Agency, SAC is performing research aimed at steel frame seismic

damage. The results, in the form of Design

Advisories and Interim Guidelines, provide

assistance to both engineers and building officials

pending development of new UBC provisions.

4. IMPROVED JOINT DESIGN

At the time that the emergency UBC changes

were issued, the new Human Genome Laboratory

at the Lawrence Berkeley National Laboratory

was being designed and was immediately

impacted by the UBC changes. Forell/Elsesser

Engineers, Inc., the structural designer for the

facility, undertook a study to develop and test a beam-to-column joint for the Special MomentResisting Frames (SMRF) selected as the

structural lateral system for this building. The

building consists of three seismically isolated

modules. Details of design and testing results are provided in Lawrence Berkeley National

Laboratory, Human Genome Laboratory Steel

Joint Test - Technical Brief (Ref. 2). The design

that evolved from this effort depends on limiting

stresses in the column and panel zone to less than

the elastic limit and on inelastic buckling of the

beam flange away from the beam-to-columnjoint.

It requires the use of materials of known

mechanical properties, high quality welding

processes, skilled crafts personnel, and diligent

quality control.

4. 1 Physical Description

The SMRF joint design is shown in Figure 1. Neither the beam flanges nor the beam web is welded directly to the column. The beam cover

plates are welded to the column flange with full

541
penetration groove welds.

The bottom cover plate is shop welded to the

column while the top cover plate is shop welded

to the beam using longitudinal fillet welds along

the beam flange edges. The width of the cover

plates and the column flange width are equal

where the plates abut the column flange. The

cover plates then taper down to a total width two

inches greater than the beam flange width. The

full width of the column flange at the cover plate

interface distributes stress in the cover

plate/column joint region as uniformly as

possible. The design allows easy removal of weld

tabs, facilitates inspection of the weld ends after

grinding, and avoids using overhead welds.

The beam web is attached to the column using a

trapezoidal web plate. The web plate is welded to the column in the shop with a full penetration,

double-sided groove weld. The web plate is welded to the beam web in the field with a threesided fillet weld. The web plate is tapered to

facilitate welding of the top and bottom sections

of the fillet weld in the field. The beam web and

web plate are predrilled in the shop to

accommodate temporary bolts used to simplify

field erection and to hold the beam in position for

welding.

Within the column, top and bottom continuity

plates are positioned on each side of the column

web to align with the beam cover plates. Each of

the four continuity plates is chamfered at the two

inside corners to prevent contact with or welding

to the column flange-web fillets. The continuity

plates are welded with full penetration groove

welds.

A temporary stiffener bar is welded diagonally

between the under side of the bottom cover plate

and the column flange during shop fabrication to

prevent inadvertent stressing of the bottom cover

plate groove weld during shipping and erection.

This stiffener bar is removed after field erection

and welding are complete.

The overall design of the beam-to-column joint

offers several additional positive features over

other "post-Northridge" designs as follows:

• The total cross-sectional weld area is reduced

resulting in both cost reduction and the

technical benefits of less heating and reduced

weld shrinkage which increase reliability.

• Beam shop fabrication time is reduced by the

absence of flange and web weld preparations

and the need for only one weld access relief at the top of the web and top flange of the beam.

The weld access relief accommodates

installing and removing the backup bar for the

top cover plate field weld.

• Field erection time is lessened with only one

down-hand full penetration groove weld (with

no interference from the beam web) for

welding the top flange cover plate to the

column, two fillet welds for welding the beam

flange to the bottom flange cover plate, and the

three-sided fillet weld for welding the web

plate to the beam web.

4.2 Design Parameters

The SMPvF joint is designed to force plastic

deformation in the beam to occur away from the

welded region. The length of the cover plates is approximately one nominal beam depth measured

from the column centerline. However, longer cover plate lengths may be required for beams of

less depth to provide sufficient length and,

therefore, strength of the fillet welds between

cover plates and the beam flanges. The beam is proportioned to satisfy stress and drift

requirements with the additional requirement that

excessive beam strength must be minimized. The

column is proportioned to satisfy UBC panel zone

strength requirements without using column web

doubler plates.

Strong column-weak beam response is confirmed

by plastic frame analysis of all possible

mechanisms, accounting for the additional

542
flexural demands imposed on the column by the

eccentric plastic hinge location.

The design parameters and specifications used for

the SMRF joint test specimens are summarized in

Table 1.

4.3 Welding Requirements

Weld specifications stipulate special provisions

for all welds used in the SMRF joint. These

include:

• Detailed welding procedure specifications;

welding inspection checklists; qualification of

welders using high-grade materials; 100

percent ultrasonic inspection in the shop and

field inspections using magnetic particle or dye

penetrant techniques.

• Flux cored arc weld (FCAW) and shielded

metal arc weld (SMAW) processes were

subject to the following limitations:

- Charpy V-Notch toughness of weld wire of

20 foot pounds at minus 20 °F minimum.

- Preheat requirements about 50 °F higher

than AWS DLL

- Post-weld control of cool-down rate if

ambient temperature goes below 50 °F.

- Requirements for peening all but root and

cap passes to provide stress relief.

- Grinding of all thermally cut surfaces.

- Removal of weld tabs and backup bars.

5. PROTOTYPE JOINT TESTS

5 . 1 Test Development

Nonlinear dynamic analysis of the building frame

was used to establish the test acceptance criteria.

Building frames were designed in accordance

with Lawrence Berkeley National Laboratory

seismic criteria which specified an allowable

stress design (ASD) static base shear of 20

percent for two buildings and 28 percent for the

third building. The analysis indicated that plastic

beam rotations of about 0.02 radian could be

expected for a Hayward seismic event, the

appropriate design basis for the Berkeley area. For the buildings being designed, the total beam

rotations are in the range of 120 percent to 140

percent of the maximum interstory drift.

Allowing for both the limited accuracy of

structural calculations and a higher than predicted

seismic event, test acceptance criteria were

established to achieve at least three cycles that

result in plastic beam rotations of 0.03 radian,

computed at the plastic hinge location.

5.2 Specimen Material Properties

Material properties were verified by mill

certificates and by independent laboratory testing

performed by Signet Testing Laboratories. The

results are provided in Table 2.

5.3 Specimen Fabrication and Testing

Two test specimens were fabricated by The

Herrick Corporation at their Stockton, California,

facility. The specimens were tested at the Davis

Hall Laboratory at the University of California at Berkeley. The testing arrangement and protocol

were identical to the SAC test procedures.

All shop welds were made using standard shop

practices. Field erection was simulated in the

shop with the column vertical and the beam

temporarily shored at its free end. Groove welds

were made with Lincoln NR-3 1 1 FCAW wire.

Fillet welds were made with Lincoln NR-232

FCAW wire. Specimens were fitted with

potentiometers, strain gauges, and strain rosettes.

During preparation of Test Specimen 2, the

technicians at Davis Hall Laboratory discovered

a four inch long by 3/16 inch deep crack in the

column flange. It was located on the flange

centerline starting at the bottom cover plate weld

and extending downward away from the weld.

The crack was discovered while grinding the area

for installation of instrumentation. Signet Testing

Laboratories determined that the crack was an

543
acceptable defect by AWS Dl.l. The testing

team elected to proceed with the test without

repairing the crack.

Each test specimen was installed in the test machine with the column and beam in the

horizontal position to accommodate the physical

design and constraints of the machine. The test arrangement is shown in Figure 2. The column

flange ends were restrained from motion in the

direction perpendicular to the plane of the column

flanges by knife edge type supports which were

placed 1 1 feet 6 inches apart and equidistant from

the centerline of the beam. The test loads were

applied to the beam flange in the direction

perpendicular to the beam flanges at a distance of

12 feet from the column centerline. The test was

performed by applying cyclic loads to the beam

to cause deflections of from ± 1 inch to ±5 inches

of the beam tip at the point of load application.

This simulated bending moments applied to the

joint by the lateral movements of a steel frame

caused by seismic events of increasing magnitude.

The test steps spanned both the point of expected

plastic deformation and the point of expected

fracture.

5.4 Test Results

Both specimens met the acceptance criteria of

three cycles of 0.03 radian plastic beam rotation.

The general results of the tests on the two

specimens are summarized in Table 3

Specimen 1 was tested up to three cycles at ±5

inches of tip displacement. The permanent

deformation of the beam web and flanges of

Specimen 1 at the end of the five test steps is shown in Figure 3.

Specimen 2 was tested up to 10 cycles at ±5

inches tip displacement whereupon a beam flange

fractured in the buckled region due to high strain

cyclic fatigue. Also, for Specimen 2, there was

evidence of minor yielding near the column crack

found prior to the tests, but the crack did not

appear to open or propagate during the tests.

5.5 Cost and Other Considerations

The cost to fabricate and erect the SMRF joint for . the Human Genome Laboratory is estimated at approximately $1000 per joint compared to approximately $500 per joint for a "preNorthridge" joint used in the same application.

This cost differential is relatively small compared

to a number of other joints designed and tested to

meet the requirements of the emergency changes

to the UBC.

The designer has cautioned that every potential

application of the SMRFjoint design, or any other

advanced design, is unique and must be qualified

for the specific loads and geometries for which its use is intended.

6. CONCLUSIONS

Brittle failures in and around the beam-to-column

joints of moment-resisting steel building frames were discovered after the Northridge, California,

earthquake early in 1994. As a result,

requirements for the design and qualification of

this type ofjoint have been greatly modified. The

SMRF joint design has been shown by both

calculation and testing to more than adequately

meet the demanding requirements for resistance

to the limiting earthquake-induced loading which

might be expected in the Lawrence Berkeley

National Laboratory area. The design offers

several unique features which improve seismic

performance over conventional, fully welded

designs of the "pre-Northridge" concept. These

include more uniform distribution of stresses in

the welds connecting the beam and column,

reduced fabrication and erection complexity,

elimination of difficult overhead welds, and

improved reliability due to reduced welding.

Finally, the SMRF joint design is cost effective

compared to a number of other new joint designs

which are being considered for use.

6. ACKNOWLEDGMENTS

The author acknowledges the expertise of Mr.

544
Donald G. Eagling, retired plant manager of the

Lawrence Berkeley National Laboratory, whose

treatise on earthquake-induced structural failures

in Chapter 3 of the U.S. Department of Energy's

draft Seismic Safety Manual (Ref. 1) was the source of the historical and technical data on the

Northridge and Kobe earthquakes, the evaluation

of the beam-to-column joint failures discovered,

and implementation of the emergency building

code changes described in this paper.

The author is indebted to Mr. Mark Jokerst, SE, of

Forell/Elsesser Engineers, Inc., for allowing the use of the design information and test data

presented in Lawrence Berkeley National

Laboratory, Human Genome Laboratory Steel

Joint Test - Technical Brief (Ref. 2) and for

providing additional information and insight on

the design, fabrication, and testing of the SMRF

Groove Weld (field)

Groove Weld

typ. 6 places

both sides of web

(shop)

11/8" Top Cover Plate

1" Continuity

Plates both sides of web

joint.

Finally, appreciation is extended to Mr. Ronald

Levis, of Roy F. Weston, Inc., for applying his

experience and knowledge of structural design to

the development, editing, and production of this

paper.

7. REFERENCES

1. Eagling, D.G. (1995). "Seismic Safety

Manual, A Practical Guide for Facility

Managers and Earthquake Engineers

(DRAFT)." U.S. Department of Energy.

2. Jokerst, M. (1995). "Lawrence Berkeley

National Laboratory, Human Genome

Laboratory Steel Joint Test - Technical Brief."

Forell/Elsesser Engineers, Inc., San Francisco,

CA.

Column

W14x283

1/2" Fillet Weld typ. 2

places (shop)

Beam

W30x90

9/16" Fil et Weld

three sid s (field)

1/2" Fillet Weld typ. 2 places (field)

1 1/8" Bottom Cover Plate

Temporary Stiffner Bar

Groove Weld (shop) (removed after erection)

Figure 1 . SMRF Joint Design

545
Figure 2. SMRF Test Assembly

Figure 3. Permanent deformation of Test Specimen 1 at end of test

546
Table 1 . Test Specimen Parameters and Specifications

Piece Parameters and Specifications

Beam ASTM A572 Grade 50; no limit placed on maximum yield strength. For design of the

connection, the mean plus one deviation of the SAC 95-01 Table A (Appendix D) data

(mean yield strength of 57.6 ksi and standard deviation of 5.1 ksi), plus a 10% strain

hardening factor, less a 5% web to flange strength conversion was used, yielding a total

effective Fy = 65.5 ksi.

Column ASTM A572 Grade 50; fine grained fully killed with grain size 6 or finer; minimum

Charpy V-notch toughness, determined at the centerline intersection of web and flange,

was 20 foot pounds at 40 °F. In all calculations, possible material overstrength was

neglected.

Cover

Plates

ASTM A572 Grade 50 with rolling grain parallel to the beam axis. Flexural stress at cover plate-column interfaces = 40ksi.

Table 2. Test Specimen Material Properties

Piece Fy, Fu, Elongation, CVN, (1) Average Tested

ksi ksi % ft-lb Grain Size by (2»

W30x90 Beam 60 71 26 Mill

Web 61.1 70.4 36.5 Lab

Flange 50.3 66.1 41.0 Lab

W14x283 Column 51 70 29 Mill

Web 48.5 67.5 32.5 Lab

Flange 45.5 67.5 32.5 21.4@40 °F #6- #7 Lab

Cover Plates 61.4 88.2 22 Mill

61.5 87.5 28.0 Lab

Web Plate 60.8 82.2 23 Mill

NOTES

(1) Charpy V-Notch test

(2) "Mill" indicates data from mill certificate; "Lab" indicates Signet Testing Lab data

547
Table 3. General Test Results

Tip

Displacement,

± inches

Op, Plastic

Beam

Rotation,

radians

Description of Beam Reactions to Test Loading

l 0.000 Very minor flaking or the paint (yield occurs at 1 . 1 inches).

2 0.005 Significant yield and plastic rotation; no buckling. Ram load peaks

at 148 kips.

3 0.014 Flange buckles before reaching ±3 inch displacement. Distorted

flanges straighten upon reversed loading. Ram load repeats up to

148 kips in first cycle; subsequent ram loads show strength

degradation.

4 0.024 Flanges and web buckle; distortion remains in constant form with

each cycle; minor movement of flanges. Ram load diminishes to

yield level (105 kips).

5 0.034 Same behavior as 4 inch displacement test. Rotation peaks at 6p =

0.039r. Ram load decreases uniformly with each cycle.

548
APPENDIX

TASK COMMITTEE A-K REPORTS

Report of Task Committee A

STRONG MOTION DATA AND APPLICATIONS

Date: 15 May 1996

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. side - Roger Borcherdt (Chairman) USGS

Jon Ake U.S. Bureau of Reclamation

GhassanAl-Chaar USACERL

David Boore USGS

William Joyner USGS

Japan side - Takahiro Sugano (Acting Chairman) PHRI

Keiichi Ohtani NIED

Noriko Kamaya J MA

1. Objective and Scope of Work

To coordinate and promote sharing of strong motion earthquake data among researchers and

practicing engineers, and to develop techniques and exchange information for evaluating the

destructive effects of earthquake motion.

The scope of work includes:

o instrumentation

o recording, processing, and analyzing strong motion data,

o engineering characterization of ground motion,

o design applications, and

o seismic zonation.

The activities of the Task Committee include:

o regular exchange of data and publications,

o creating procedures for disseminating significant strong motion digital data with regard

for the rights and expectations of (a) owner, (b) the users of data and © the earthquake

engineering community,

o planning and conducting T/C workshops and meetings, and

o coordinating relevant research activities.

2. Accomplishments

Conducted an international workshop on Site Response to Strong Earthquake Motions, 16-17

January 1996, Yokosuka, Japan; conducted technical site visits relevant to the Workshop in the

Hyogo ken-Nanbu area during 18-19 January 1996. The workshop was organized by Dr. Iai,

PHRI with support of the Japan Science and Technology Exchange Center acting on behalf of

the PHRI. More than 50 world experts discussed the response of soils under large earthquake

excitation. Workshop discussions centered on the response of soils as implied by strong-motion

recordings of the Northridge and Hyogo ken-Nanbu Earthquakes and recent results implied by

laboratory experiments and numerical modeling results. Critical issues related to the non-linear

551
behavior of soils were treated in detail. Proceedings of the workshop were prepared for

publication by March 1996.

3. Future Plans

a. T/C "A" recognizes the importance of understanding near source earthquake ground

motions, geological effects, and structural response in support of the "Natural Disaster

Reduction Initiative" of the "United States-Japan Common Agenda." Strong motion

recordings obtained from past and future earthquakes provide the basic measurements

essential for understanding these effects for improved earthquake resistant design and

retrofit of structures needed to reduce future earthquake disasters. T/C "A" in

recognizing its responsibilities for "Strong-Motion Data and Applications" recommends

the immediate establishment of coordinated programs for:

1) Strong-motion Instrumentation Array Design, Operation and Data

Interpretation based on recent design and installation of extensive surface and

borehole arrays installed by the Science and Technology Agency throughout

Japan and the U.S. Geological Survey in the San Francisco Bay region and other

areas.

2) Understanding Near Source Effects, especially a) near source attenuation

relations for fault-normal and fault-parallel motions and b) procedures to

account for source rupture types and radiation effects, including directivity, into

engineering design.

3) Understanding Geological Effects; especially a) acquisition of site- condition data

and analyses of corresponding strong motion data from Northridge and Hyogo

ken-Nanbu earthquakes for characterizing the response of various types of

geological site conditions at high levels for input rock motions for purposes of

improving ground-motion amplification factors currently used in Building Code

Provisions and earthquake resistant design.

4) Understanding structural response, especially coordinated efforts to facilitate

interpretation of strong motion measurements in buildings for resolving

important structural response problems such as those for modern steel moment

frame structures.

b. T/C "A" recognizes that extensive sets of strong-motion instrumentation installed in

both countries since the Hyogo ken-Nanbu and Northridge earthquakes provide an

important new resource to develop improved understanding of these important effects

and responses. In preparation to fully use this important new international resource and existing resources, T/C "A" recommends the development of a www (internet) page

as part of the UJNR panel page to:

• provide a list of participating agencies &. researchers

• provide internet addresses and/or hypertext links to those researchers

• list of www sites where strong motion databases are available

• provide a listing of recent strong motion recordings of interest and

• provide a list of proceedings/publications produced under the auspices of Task

Committee A of the UJNR.

c. T/C "A" endorses the recommendation of T/C "B" concerning the importance of soilstructure interaction (SSI) as a field observation. In the U.S., USGS has initiated a

project on SSI. Since the Hyogo ken-Nanbu earthquake, BRI has installed instruments

552
for SSI observations in the Tokyo metropolitan area. This task committee recommends

(1) exchange of information on SSI data and (2) the establishment of a coordinated

program.

T/C "A" recognizes the need for continued interaction between U.S. and Japanese

researchers in the forum provided by regular workshops.

T/C "A" will investigate alternative funding venues to focus on important issues such as site-response and near-source phenomena.

T/C "A" in recognizing the considerable interest in proceedings of the recent workshop

on "Site Response to Strong Earthquake Ground Motions" committee members will

expand efforts to communicate availability of proceeding volumes through professional

societies in both countries.

553
Report of Task Committee B

TESTING AND EVALUATION PROCEDURES FOR BUILDING SYSTEMS

Date: 14 May 1996

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. Side -- H. S. Lew (Chairman) NIST

Shih-Chi Liu NSF

Mehmet Celebi USGS

Erdal Safak USGS

Harish Chander DOE

Japan Side -- Keichi Ohtani (Chairman) NIED

1Objective and Scope of Work

To develop rational test procedures and to collect performance data of dynamic response of

structures through laboratory testing and prototype structures and field testing of structures insitu.

The scope of this Task Committee includes the following:

o Plans and conducts workshops and joint meetings to identify research topics and

develops joint research programs,

o Coordinates research projects carried out by various laboratories in the U.S. and Japan

and facilitates publication of research results and implementation of findings in codes

and standards.

o Facilitates exchange of research personnel, technical information and available testing

facilities.

o Develops uniform testing procedures including test loading history for case of

comparison of results of tests carried out by various researchers, and for establishment

of data base.

o Develops guidelines for interpretation of test results in consideration for design of

structures.

o Develops methodology for evaluation and interpretation of physical test results.

2. Accomplishments

a. The fourth phase of the U.S. Precast Seismic Structural Systems Program (PRESSS) has

been initiated in 1995 with several universities participating in the program. The fourth

phase includes a full scale testing and supporting subassemblage tests and is near

completion.

b. The research program on Composite and Hybrid Structures (CHS) in Japan was

initiated in 1993. Experimental and analytical work is being carried out by the Building

Research Institute and by several universities and construction companies. The Japan

Structural Consultant Association is undertaking an application study for implementing

research results into practice.

c. The research program on CHS in the U.S. was initiated in 1995. Dr. S. C. Goel,

554
University of Michigan and Dr. Steve Mahin, University of Califomia-/Berkeley are coordinating analytical and experimental studies being carried out by several

universities.

d. The second US-Japan Joint Technical Coordinating Committee (JTCC) meeting on

Composite and Hybrid Structures (CHS) was held in June 1995 at Honolulu, Hawaii.

e. The U.S. Technical Coordinating Committee meeting on CHS was held in Los Angeles,

California in February 1996. The second-year U.S. research program will be awarded

by NSF in June 1996.

f. The JTSC-RFI (Research for Innovation) was held in Ann Arbor, Michigan in February

1996, and JTSC-CFT (concrete filled tube) was held in Chicago, Illinois in April 1996.

3. Future Plans

a. The third U.S.-Japan Joint Technical Coordinating Committee meeting on CHS will be

held in November 1996 in Hong Kong.

b. T/C "B" recognizes the importance of soil-structure interaction (SSI) as a field

observation. In the U.S., USGS has initiated a project on SSI. Since the Hyogo Ken

Nanbu earthquake, BRI has installed instruments for SSI observations in the Tokyo

metropolitan area. This task committee recommends that (1) exchange of information

on SSI data and (2) the establishment of a coordinated program. The scope of the

program will be developed jointly by USGS and BRI before the 29th joint meeting in

cooperation with T/C "A" and T/C "H".

c. In support of the International Decade for Natural Disaster Reduction (IDNDR)

Program, techniques will be explored to disseminate the findings of completed Joint

Research Projects, such as masonry and precast concrete structures, to countries with

high seismic risks.

d. To maximize the usefulness of research data, it is desirable to establish standards for

test procedures, documentation and retrieval of test data. A joint workshop on "Load

Path and Test Observation Requirements for Quasi-Dynamic and Shake Table Testing"

is proposed to be held before the 29th Joint Panel Meeting to discuss this topic.

555
Report of Task Committee C

DESIGN, EVALUATION, AND IMPROVEMENT OF STRUCTURES

Date: 14 May 1996

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. side - Ken Chong (Chairman)

S.C. Liu

Larry Hultengren

Ugo Morelli

NSF

NSF

DOS

FEMA

Japan side -- Hisahiro Hiraishi (Chairman) BRI

1. Objective and Scope of Work

To develop disaster mitigation guidelines and programs that will improve the performance of

new structures and the capacity of existing structures in resisting wind and seismic forces. To

establish adequate evaluation of performance, each country will coordinate development of

condition assessment, screening, and structural analysis and design methodologies. Enhance

performance can be obtained by the intelligent renewal, repair and retrofit of existing buildings

and other structures. Structures will be analyzed and instrumented, and then evaluated after

disasters. Also, effective use of new advanced materials and structural systems, associated with

appropriate design methods, will provide excellent structural performance in the life cycle.

The scope of work includes:

o Exchanging information, planning, and hosting workshops on new design technologies,

evaluation technologies, and repair and retrofit techniques,

o Studying new materials and methods for repairing and retrofitting existing structures

and for new construction,

o Developing a uniform system for screening and analyzing wind and seismic resistance

capacity of structures in each country,

o Developing reliable condition assessment systems for new and existing structures,

o Evaluating masonry, precast concrete and composite/hybrid building structures: and

providing documentation for post-disaster evaluation of performance,

o Coordinating research projects in the U.S. and Japan to minimize duplication and

maximize benefits.

2. Accomplishments

a. U.S.-Japan Workshop on smart materials and structures supported by NSF was held at

the University of Washington during 4-5 December 1995.

b. An NSF funded International Symposium on Civil Infrastructure Systems Research was

held during 25-28 September 1995, Seoul, Korea. Researchers from U.S., Japan,

Korea, and other countries participated.

c. The Third U.S.-Japan Technical Coordinating Committee meeting on Composite and

Hybrid Structural Systems was held in June 1995 at Hawaii.

556
3. Future Plans

The following are planned for T/C "C":

a. The second workshop on "Structural Control and Intelligent Material Systems" is being

planned in the U.S. during 1996 (co-sponsored by NSF and T/C "C" and "G"). The

workshop will be held at the University of Maryland, following the ASCE Materials

Conference, November 1996.

b. Develop a reliable methodology for screening and analyzing wind and seismic resistant

capacity, and advanced instrumentation technology and expert systems to provide

condition assessment of existing structures.

c. Compile a database of advanced materials/systems that have potential for improving

structural performance of new construction, and for use in rehabilitating and

strengthening existing buildings.

d. Investigate new design methodologies based on required structural performance using

advanced materials and systems.

e. Develop design guidelines for applying advanced structural and materials system to be

used for composite and hybrid structures under the joint U.S. -Japan Cooperative

Research Program. The fourth U.S.-Japan Technical Coordinating Committee meeting

on Composite and Hybrid Structural System will be held November 1996 at Hong

Kong.

f. Continue to encourage participation by private industry, consulting engineers,

universities, national and local government agencies involved in instrumentation,

evaluation, condition assessment, and retrofit and strengthening of existing buildings,

for seismic and wind resistance.

g. In collaboration with T/C "G", to develop the next U.S.-Japan Cooperative Research

Project on Smart Structural System, a working group will be formed to plan for the

technical report and to develop a joint initiative within two years. The first meeting will

be held in conjunction with the U.S./Japan Workshop on Smart Structures and

Materials in November 1996 in Washington, D.C.

4. Other Activities

a. U.S. research sponsored by NSF, CORPS, FHWA, and NIST has developed reports on

seismic rehabilitation and retrofit strategies, including innovative techniques for

strengthening existing buildings and bridges.

b. An updated report on Civil Infrastructure Systems - Strategic Issues has been published

by NSF. Work continues on development of NSF's Civil Infrastructure Systems

Research Program, including strong emphasis on evaluation and improvement of

structures.

c. NSF is coorporating with Japan Science and Technology Agency (STA/NISTP) on

Smart Materials Research proposed by STA.

d. PWRI, BRI, NIEDC, NIST, USGS, NSF, and FHWA held a meeting during March 28-

29, 1996, at Stanford University to discuss opportunities for cooperation under the

U.S.-Japan Common Agenda, "A New Economic Partnership for the 21st Century".

e. Continue to encourage research in the retrofit and repair areas.

557
A PLAN FOR NEXT U.S.-JAPAN COOPERATIVE RESEARCH PROJECT

UJNRT/C "C" &"G"

TITLE: Development of Smart Structural Systems

OBJECTIVE: Buildings of low-cost and high-performance can be developed by optimal use of

materials in terms of their properties, cost, and availability, and by intelligent

design of their components, elements and structural systems which are durable,

functionally adaptive, safe under hazardous loading conditions have minimum

life- cycle cost. The objective is to develop buildings and other infrastructure

systems which are expected to meet the demands for high-performance, low

maintenance, long service life, and possess on-line monitoring, detection,

diagnosis, and control capabilities. New materials which exhibit exceptional

sensoring and actuating capabilities to enable the smart structural elements and

systems, shall be studied and developed. Advanced signal processing techniques

and control technologies for building and infrastructure system applications are

also essential to meet this objective. In this project, new structural systems are developed to have functions as follows:

1 . Sensoring which sense and detect the unusual conditions.

2. Processing which diagnose the sensored data and makes judgements on

the response to the unusual conditions.

3. Actuating which initiates responding actions to the unusual conditions.

OUTLINE: Development of structural systems which use high-performance and smart

materials, and intelligent engineering for design and construction. These

systems might employ resistance and self-control systems under unusually large

external load conditions.

PRESENT

SITUATION: It is expected that new innovative structural systems use new materials such as shape memory alloys, electrorheology (ER) fluid, magnetorheological (MR) fluid,

and optical fibers, etc., will be developed, in addition to the low yield point steel,

viscous damper and self-repairable concrete.

EXPECTED

RESULTS: Building systems that keep the following performance will be developed:

1 . Maintain the functions under unusually large external load effects.

2. Excellent habitability and durability.

3. Architecturally attractive new structural system.

RECOMMENDED

ACTIONS: To immediately commission a JTCC planning study for a 12-month period.

This study shall result in a technical report which lays out a detailed plan for a 5-year joint U.S.-Japan research program in smart structural systems for

implementation beginning in 1998.

558
Report of Task Committee D

EARTHQUAKE ENGINEERING FOR DAMS

Date: 14 May 1996

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. side -- William Roper (Chairman) USACE, HQ

Lucian Guthrie USACE, HQ

Francis McLean USBUREC

Richard S. Olsen USACE, WES

A.G. Franklin USACE, WES

Japan side - Michio Okahara (Chairman) PWRI

Shigetoshi Kobayashi PWRC

1. Objective and Scope of Work

To develop technical insights into better understanding of the response of dams to seismic

effects. The T/C will plan, promote, and develop research initiatives to assist in assuring

seismic safety and economical protective countermeasure against earthquake loading for these

structures. Its scope of work includes:

o Methods of analysis for seismic design of dams.

Comparison of design methods and criteria between U.S. and Japan

Development of "Design Earthquake Ground Motions" for analysis and evaluation of

dams.

Assessment of investigation and dynamic analysis methods as tools (modeling,

calculation codes, etc.).

o Dynamic characteristics of dam construction materials.

Strength and deformation characteristics during earthquakes (concrete, soil and rock),

o Analysis of observed behavior of dams during an earthquake.

Investigation of the mechanisms of damage from earthquake loading.

Application of the analysis of the observed behavior during earthquakes to the earthquake —

resistant design.

2. Accomplishments

a. Preliminary plans have been completed for the November 1996 Workshop.

b. Results/investigation and technical information about dam earthquake engineering were

exchanged as reports and publications.

3. Future Plans

a. Exchange results of research/investigation and technical information about dam

earthquake engineering, and feasibility of joint research and investigation is in

continuous consideration.

b. Exchange visits to the institutes concerned, of scientists and engineers between U.S. and

559
Japan, is to be extended for the effective communications.

According to progress of the activities (1) and (2), T/C(D) is planning to conduct a

Workshop on State-of-the-Art Review of U.S.-Japan Research on Earthquake

Engineering for Dams, at the Waterways Experiment Station in Vicksburg, Mississippi.

The proposed date for the workshop is 19-22 November 1996.

560
Report of Task Committee E

DESIGN FOR WIND AND WIND HAZARD MITIGATION

Date: 15 May 1996

Place:

Attendees:

National Institute of Standards and Technology, Gaithersburg, MD, USA

U.S. side - Co-Chair

Co-Chair

Visitor

Japan side — Co-Chair

Co-Chair

Dr. J. E. Sabadell

Dr. A. W. Taylor

Dr. J. Golden

Mr. H. Bosch

Dr. R. Marshall

Dr. Joseph Minor

Dr. H. Okada

Dr. H. Sato

NSF

NIST

NOAA

FHWA

NIST

University of Missouri, Rolla

BRI

PWRI

1. Objective and Scope of Work

To exchange technical information and to jointly plan, promote and foster research and its

dissemination to improve understanding of wind and its effects on structures, establish more

rational wind resistant design methods for structures, and to contribute to wind hazard

mitigation.

The scope of work includes:

o Characteristics of strong wind, especially boundary layer extreme winds

Wind effects (wind loading on and wind-induced response of structures)

Experimental and analytical methods to predict wind and its effects

Damage and risk assessment

Wind hazards and wind hazard mitigation

o

2. Accomplishments

Developed T/C Goals, Objectives, and Mission Statement and identified and received approval

from public and private wind related experts to serve as T/C members. The list of U.S.-side and

Japan-side members appears below:

U.S. side — Co-Chair

Co-Chair

Dr. J. E. Sabadell

Dr. A. Taylor

Dr. B. Bienkiewicz

Mr. H. Bosch

Dr. G. Chiu

Dr. W. Freeborn

Dr. J. Golden

Dr. A. Kareem

Dr. R. Marshall

Dr. K. Mehta

Dr. D. Reed

NSF

NIST

Colorado State University

FHWA

IIPLR

HUD

NOAA

University of Notre Dame

NIST

Texas Tech University

University of Washington

561
Dr. T. Reinhold

Dr. C. Smith

Dr. P. Tertell

Clemson University

MMS

FEMA

Japan side -- Co-Chair

Co-Chair

Dr. H. Okada

Dr. H. Sato

Dr. H. Akiyama

Dr. K. Kimura

Dr. N. Kinoshita

Mr. Y. Matsuno

Dr. G. Naito

Dr. T. Tamura

Dr. Y. Tamura

Dr. Y. Uematsu

Dr. H. Yamada

Dr. S. Yoshizumi

BRI

PWRI

Honshu-Shikoku Bridge

Authority

University of Tokyo

MRI

PWRI

National Defense Academy

Tokyo Institute of

Technology

Tokyo Institute of

Polytechnics

Tohoku University

Yokohama National

University

Meteorological Research

Institute

3. Future Plans

a. Exchange technical information on the following subjects:

1) Topographical effects on wind, and the use of geographic information systems (GIS)

to study these effects

2) Control of wind-induced response of structures

3) Computational fluid dynamics vs. wind tunnel tests

4) Prediction of wind-induced response of full-scale structures

5) Investigation of wind hazards, including damage assessment methodologies

b. Plan and hold a joint workshop. The proposed time of the workshop is in October or November of 1997. The proposed theme of the workshop is "The Prediction and

Control of Wind-induced Response." The proposed location of the workshop is in the

United States. Further discussions on the time, theme and location of the workshop

will take place over the next several months.

562
Report of Task Committee F

DISASTER PREVENTION METHODS FOR LIFELINE SYSTEMS

Date: 14 May 1996

Place: National Institute of Standards and Technology, Gaithersburg, Maryland

Attendees: U.S. side - Riley M. Chung (Chairman) NIST

RonAndrus NIST

James Cooper FHWA

W. Phillip Yen FHWA

Japan side - Hisanori Ohtsuka (Act. Chair) PWRI

Noriko Kamaya Japan Meteorological Agency

Kazuhiko Kawashima Tokyo Institute of Technology

Kenji Kosa Hanshin Expressway Public Corp.

Keiichi Tamura PWRI

Yuzo Yamaguchi NTT

1. Objectives and Scope of Work

To improve (a) the performance of lifeline systems during earthquakes and (b) engineering and

other seismic countermeasures including the capability in damage estimation techniques and

inspection procedures.

o Planning and conducting workshops on lifeline systems,

o Facilitating exchange of technical information.

o Promoting development of design guidelines and standards for lifeline systems.

2. Accomplishments

a. Conducted the Sixth Joint Workshop on Disaster Prevention for Lifeline Systems in

Osaka, Japan, July 18-19, 1995. The workshop addressed research and development

topics in the lifeline area and gave special emphasis to lessons learned from the 1995

Hyogo-ken Nanbu (Kobe) earthquake and the 1994 Northridge earthquake. The

proceedings has been printed by the Public Works Research Institute and distributed to

all the workshop participants and others in the lifeline earthquake engineering

community.

b. Members of this T/C have been actively involved in the planning of two major efforts

between the two countries in earthquake disaster reduction. The first one is led by the

STA of Japan and the OSTP of U.S. to develop a list of scientific and engineering

projects for future collaboration. The second is the planning for two earthquake policy

symposia, led by the State Department (through FEMA) and the Japanese Foreign

Ministry. In both cases, this UJNR Panel on Wind and Seismic Effects is recognized for

its excellent work in the past and recommended as the vehicle for further enhancement

of the two countries' joint activities.

563
3. Future Plans

a. Encourage collaborative research and development in the following areas:

• performance prediction and post-measurement through damage assessment of

lifeline systems.

• systems approach to lifeline performance

• soil-structure interaction

• geographic information systems (GIS)

• cost-effective ground improvement techniques and other countermeasures for

lifeline facilities

b. Encourage and strengthen current efforts in both countries to establish seismic design

guidelines and standards for lifeline systems. Existing UJNR channels should be fully

used to facilitate the exchange of relevant information concerning guidelines and

standards development. Possible collaboration of joint development of guidelines and

standards for lifeline systems should be pursued.

c. Hold the Seventh Joint Workshop on Disaster Prevention for Lifeline Systems in

June/July 1997 in Seattle, Washington (Seattle is a sister city of Kobe). The workshop

will focus on the presentations and discussions of the results of the studies completed

and to be completed after the Northridge and Kobe earthquakes.

564
Report of Task Committee G

STRUCTURAL CONTROL AND INTELLIGENT MATERIAL SYSTEMS

Date: 15 May 1996

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. side - Shih-Chi Liu (Chairman) NSF

Riley M. Chung NIST

H.S. Lew NIST

Ian Buckle NCEER

James Cooper FHWA

Glen Smith FHWA

Hamid Ghasemi FHWA

W. Phillip Yen FHWA

Japan side - Hisanori Otsuka (Chairman) PWRI

Kazuhiko Kawashima Tokyo Institute of Technology

Hisashi Okada BRI

Kenji Kosa Hanshin Expressway Public Corp.

Hisahiro Hiraishi BRI

Hiroshi Sato PWRI

1. Objective and Scope of Work

o develop research plans in control of equipment and structures and in high performance

structural and material systems;

o implement control techniques for motion reduction or modification;

o implement use of advanced materials in actual design and construction of buildings and

other infrastructure systems under seismic or wind environments;

o promote U.S.-Japan cooperation in structural control and intelligent material systems

research;

o bring together governmental, academic, and industrial participants in joint pursuit of

these efforts; and

o contribute to IDNDR by organizing joint research and other technical activities in

structural control and intelligent material systems research based on international

cooperation.

T/C "G" works closely with other organizations to provide the leadership in this emerging

research by facilitating the exchange of technical data and information through UJNR

mechanisms. The scope of work includes:

o providing technical assistance, consultation and coordination of UJNR affiliated

research organizations in the initiation, development, and execute their programs in

research areas. o promoting Joint government-university-industry collaborative efforts to facilitate

technology transfer and practical implementation,

o sponsoring and conducting interdisciplinary workshops and meetings to identify key

565
area of research and opportunities for cooperation, and to exchange new knowledge and

experience in practice.

o developing promotional and demonstrative activities to stimulate public awareness and

interest in this field of research.

o providing information useful for the establishment of performance standards, design and

retrofit/rehabilitation of existing structures.

o promoting research in intelligent material systems, sensors, actuators, optimal control

system design, and encourage laboratory and field experiments of prototype and fullscale structures.

2. Accomplishments

a. At BRI, performed research on structural control and intelligent material systems using

self- diagnosis and self-repairable materials; work initiated April 1994.

b. At PWRJ, initiated a research program on intelligent seismic structures; April 1995.

This research will develop new technology using self-diagnosis materials and

self-repairable materials.

c. Joint research programs for active control of bridges are being carried out through T/C

"G" and "J" by EPRI, NSF, FHWA, NIST, NCEER, and UC Irvine.

d. NSF worked with STA( Japan Science and Technology Agency) on "Smart Materials"

research program planning by STA.

e. A new NSF research initiative on sensors and sensoring systems for power systems and

other dispersed civil infrastructure systems was issued for requesting final proposals by

April 1995. EPRI and NSF are co-funding this initiative.

f. The proceedings of the 3rd U.S.-Japan Workshop on Protection Systems for Bridges

were published and distributed.

3. Future Plans

a. Under NSF sponsorship, an U.S. -Japan workshop on "Structural Control and Intelligent

Materials systems" is planned to be conducted in November 1996 in cooperation with

T/C "C." The University of Maryland at College Park will host the workshop.

b. Joint research projects on innovative technologies for active and hybrid control of

structures, such as bridges and buildings, will be developed by PWRI, BRI, NCEER,

CUREE, and other institutions.

c. In cooperation with T/C "J" the 4th U.S.-Japan Workshop on Earthquake Protective

Systems for Bridges is being scheduled for December 1996 in Japan. This workshop will

be combined with the 3rd U.S.-Japan Workshop on Seismic Retrofit of Bridges, which

will be held in Japan at the same time, also with T/C "J". Proposed topics for the

protective systems workshop include:

• innovative protection systems and high performance material

• design methods for seismic isolated/menshin bridges

• full-scale verification of component and system performance

• comparative design studies

Proposed topics for the retrofit workshop include:

• performance criteria and risk assessment

• improved evaluation methods

566
• innovative repair and retrofit methods

d. NIST will conduct a state-of-the-art review of active control systems and evaluate those

available technologies for use by designers.

e. In collaboration with T/C "C", the next U.S.-Japan Cooperative Research Programs on

Smart Structural Systems (CSSS) will be developed. A joint technical coordination

committee (JTCC) on SSS will be established to develop a technical planning report for

the joint research and to develop a joint initiative within two years. The first meeting

will be held in conjunction with the U.S.-Japan Workshop on Smart Structures and

Materials in November 1996 in D.C.

4. Other Activities

a. A joint ASME, JSME, and PVP conference on seismic base isolation was held at

Honolulu, Hawaii, July 1995.

b. Under NSF sponsorship, an U.S.-Japan Workshop on Smart Materials and Structures was

conducted at the University of Washington, 4-5 December 1995.

c. An U.S.-Japan Workshop on Structural Control was held in Kyoto, March 1996. The

objective of this workshop, which was jointly organized by the U.S. Panel on Structural

Control Research and the Japan Panel on Seismic Response Control, is to develop

cooperative research in various structural control subjects and to establish a broad

implementing mechanism involving academic, industrial, and governmental researchers

from both sides.

d. The second International Workshop on Structural Control will be held in December

1996 in Hong Kong. The workshop will be hosted by the Hong Kong University of

Science and Technology.

567
Report of Task Committee H

SOIL BEHAVIOR AND STABILITY DURING EARTHQUAKES

Date: 15 May 1996

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. side - A.G. Franklin (Chairman) WES

F.G. McLean U.S. Bureau of Reclamation

Ronald Andrus NIST

R.S. Olsen WES

Takeo Moriwaki NIST (guest researcher from

Hiroshima University)

Japan side - Keiichi Tamura (Chairman) PWRI

Yuzo Yamaguchi NTT

1. Objective and Scope of Work

Government agencies responsible for public works must assure seismic safety and provide

economical protection against earthquake hazards.

The objective of the Task Committee (H) is to assist in meeting these needs by enhancing the

availability of technology for predicting the dynamic behavior of soils, analyzing dynamic soilstructure interaction, and modifying the earthquake behavior of foundations and earth

structures to assure their safe performance during earthquakes.

In accordance with the objective, the scope of work includes:

o Exchange information on technological developments and on state-of-the-art and

practice related to soil behavior and stability during earthquakes.

o Exchange information and technical data relating to field performance, research, and

methods of practice.

o Plan and conduct programs of cooperative research and/or workshops in coordination

with the proposed or ongoing programs.

o Make other needed efforts including exchange of researchers between U.S. and Japanese

research institutions, and publication of research results and recommended practice.

2. Accomplishments

a. The Japanese Society of Soil Mechanics and Foundation Engineering (JSSMFE) has

published in Japanese a book on Liquefaction Remediation, and has produced a draft

English translation. Editing of the draft has been completed with the assistance of

Panel members and it is being prepared for publication by the JSSMFE.

b. The Port and Harbour Research Institute (PHRI) has published in Japanese Handbook

568
on Remedial Measures for Reclaimed Land Liquefaction, and a draft English translation

has been produced by the Corps of Engineers. Review and editing of the draft has been

completed by both the Japanese and U.S. sides, and it is being prepared for publication

by the PHRI by the end of 1996.

c. Dr. Takeo Moriwaki and Dr. Yasushi Sasaki, of Hiroshima University, visited the

Waterways Experiment Station in December, 1995, and March 1996, respectively, to

exchange technical information on earthquake engineering and to view the WES

research facilities, including the newly installed U.S. Army centrifuge. Dr. Sasaki also

visited the sites of Sardis and Enid Dams, in Mississippi.

d. The National Research Institute for Earth Science and Disaster Prevention (NIED), the

Building Research Institute (BRI), and Wayne State University (WSU), continued work

in their cooperative research program "Physical and Numerical Simulation of Structural

Damages Due to Liquefaction and Development of Countermeasure Techniques,"

planned for 1994-1998.

e. The PHRI, the Sandia National Laboratory, and the Lovelace Institutes (Albuquerque,

NM) have initiated joint research on application of the Nuclear Magnetic Resonance

Imaging (NMRI) method to the study of soil behavior and stability during earthquakes.

f. Dr. Takeo Moriwaki, of Hiroshima University, is a guest researcher at NIST from

March 1996 to September 1996.

3. Future Plans

a. The JSSMFE will publish the English translation of its book on Liquefaction

remediation.

b. The PHRI will publish the English translation of its Handbook on Remedial Measure

for Reclaimed Land Soil Liquefaction.

c. T/C (H) tentatively plans to hold a workshop on the Use of Centrifuge Experiments for

Research on Soil Behavior and Stability During Earthquakes in the U.S. in 1998.

569
Report of Task Committee I STORM SURGE AND TSUNAMIS

Date: 15 May 1996

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. side - Michael Blackford (Chairman) NOAA

Wilson Shaffer NOAA

Jye Chen NOAA

Japan side - Michio Okahara (Chairman) PWRI

Shigetoshi Kobayashi PWRC

1. Objective and Scope of Work

The objective of this Task Committee is to mitigate damage from storm surge and tsunami

through cooperative research and shared technology and information. Storm surges are considered tropical cyclones (hurricanes, typhoons). The primary cause of a tsunami is

considered to be sudden sea floor deformation due to earthquakes, volcanic activity or

landslides. Both hazards may cause disasters along coastal regions.

The scope of work of this T/C is as follows:

o Exchange results of research on storm surge and tsunami occurrence, generation,

propagation, and coastal effects. This includes observations on historical, current, and

theoretical tsunamis. Of particular interest is the effort by U.S. and JAPAN to acquire

deep ocean tsunami measurements.

o Exchange results and status of storm surge and tsunami mitigation activities including

analysis of the problem, planning, warning, and engineering approaches.

o Exchange information on development of technologies such as computer programs to

predict travel times, land-fall locations, inundation and run-up heights, and wave

characteristics, improved instrumentation, and use of satellite communication for

detection and warning.

o Facilitate dissemination through exchanges of literature, technical reports at joint

meetings, special workshops, joint projects, and direct interaction among participants.

2. Accomplishments

a. Task Committee (I) conducted its 4th Workshop on Storm Surge and Tsunamis, Two

Great Tsunamis: U.S.-Japan Anniversary Symposium,during 1-3 April 1996, in Hawaii. The

Workshop was a memorial of the centennial anniversary of the June 15, 1896 great

Sanriku Earthquake and tsunami in Japan and the 50th anniversary of the April 1,

1946, Aleutian earthquake and tsunami. The Proceedings of this Workshop is being

prepared by the Tsunami Society for publication in their journal. The proceedings will

be distributed to Task Committee Members and other interested parties when they

become available

570
3. Future Plans

The task committee will discuss, through mutual communication, plans for a 5th Storm Surge

and Tsunami Workshop in the near future. At the Task Committee meeting during the 28th

Joint Meeting of U.S. -Japan Panel on Wind and Seismic Effects, a tentative date and place,

Honolulu, Hawaii in 1998, was suggested by the U.S. side.

4. Others

a. The storm surge research community is small and incorporated into related societies and

international organizations which define and stimulate work in the field. Task

Committee "I" encourage these efforts.

b. Since tsunamis have been less severe in the U.S. than in Japan, the U.S. has fewer

researchers in this specialty. The Task Committee, through its meetings and

workshops, encourages exchanges of ideas and joint study by U.S. and Japanese

investigators of tsunami events throughout the world.

571
Report of Task Committee on J WIND AND EARTHQUAKE ENGINEERING FOR TRANSPORTATION SYSTEMS

Date: 15 May 1996

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. side - Mr. James Cooper (Chairman) FHWA

Dr. Philip Yen FHWA

Japan side - Dr. Hiroshi Sato (Acting Chairman) PWRI

1. Objectives and Scope of Work

To: (1) plan, promote, and foster research on the behavior of highway bridges when subjected

to wind and seismic forces and (2) disseminate research results and provide specifications and

guidelines based on the Task Committee's findings. Surface transportation systems play a vital

role in the movement of goods and people. Highway bridges are especially influenced by the

forces of wind and earthquakes because of their open exposure to those forces.

The scope of work includes:

o Focus research on highway bridges without limitation on their size and function,

o Investigate existing and new bridge designs and the behavior of whole bridge systems

and/or single components of a bridge.

2. Accomplishments

a. Conducted the 1 1th U.S.-Japan Bridge Engineering workshop during 30-31 May 1995,

Tsukuba, Japan.

b. Exchanged researchers and practitioners in Japan and the U.S. Dr. Hamid Ghasemi of

FHWA was hosted by PWRI for one month investigating seismic isolation. Mr. Sentaro

Takagi, Tokyo Metropolitan Government, visited FHWA and attended the FHWA

Bridge Inspection course. Mr. Tomofumi Nozaki of PWRI visited Turner Fairbank

Highway Research Center.

3. Future Plans

a. Conduct the 12th U.S. -Japan Bridge Engineering Workshop, October 26 through

November 3, 1996, USA.

b. Conduct the 3rd U.S.-Japan Workshop on Seismic Retrofit of Bridges in December

1996, in Japan. The workshop will focus on practical repair methods of damaged

bridges and seismic retrofit measures.

c. In cooperation with Task Committee (G), the 4th U.S.-Japan workshop on Earthquake

Protective Systems for Bridges is being scheduled in December 1996, in Japan. Topics

are expected to include (1) innovative protective systems; (2) design methods for

seismic isolated/menshin bridges; (3) full-scale verification of component and system

performance; (4) comparative design studies, and (5) smart/high performance structural

materials and systems.

d. Continue to investigate and exchange technical information on improved seismic retrofit

572
and strengthening procedures for highway bridges based on experimental, analytical,

and field studies. This exchange should include information on maintenance of existing

bridges.

Continue experimental research on the seismic performance of bridge piers and

columns, and encourage research on seismic isolation and hybrid control of bridges in

cooperation with Task Committee (G).

Encourage a research study on seismic, aeroelastic, and aerodynamic response of cablesupported bridges with emphasis on behavior of composite materials, cable inspection,

vibration control, and corrosion protection.

Continue the coordinated research study to compare the seismic design criteria for

bridges in Japan and the U.S. and discuss the method and analysis procedures for bridge

column design. Continue information exchange on the application of the limit state

design method.

Encourage a coordinated research study on seismic response and control, system

identification techniques, nondestructive evaluation of bridge structures, use and

performance of structural materials including new materials, and performance of

jointless bridges.

573
Report of Task Committee K

WIND AND EARTHQUAKE ENGINEERING FOR OFFSHORE AND

COASTAL FACILITIES

Data: 14 May 1996

Place: National Institute of Standards and Technology, Gaithersburg, MD, USA

Attendees: U.S. side -- Dr. Charles Smith (Chairman) MMS

Mr. Michael Blackford NOAA

Dr. Wilson Shaffer NOAA

Japan side -Dr. Sugano (Acting Chairman) PHRI

1. Objective and Scope of Work

To develop technical insights necessary to mitigate damage to offshore and coastal facilities due

to extreme wind and seismic effects. The Task Committee will plan, promote, and develop

research initiatives to meet this objective, and will disseminate the results of their research for

incorporation into future specification or design guidelines. Criteria for the design of offshore

and coastal facilities may differ greatly from their on shore counterparts. These differences can

arise due to their unique design or mass distribution, to the fluid/structure or wind/structure

interaction, to the placement of foundation elements in or on soft, fully saturated soils that can

be subject to large hydrodynamic pressures, and to the lack of specific environmental data or

the engineering experiences developed for most onshore sites.

The scope of work includes:

o Sponsoring and conducting workshops and meetings to identify key area of research,

opportunities for cooperation, and the exchange of knowledge,

o Predicting strong ground motions for offshore and coastal sites including assessing the

effects of basin geometry, linear, and nonlinear local geological effects using actual

seafloor response measurements,

o Determining the dynamic response and the interaction of structure/foundation/soil

systems to seabed motions and/or extreme wind forces,

o Assessing the dynamic response and behavior of various operational facilities mounted

on offshore and coastal structures,

o Developing assessment methodologies for seismicity and other characteristics of

potential seismic sources (e.g., faults) for offshore and coastal sites about how these

conditions relate to structural design criteria,

o Promoting the implementation of new research results into current design and

construction processes.

o Developing research efforts to include laboratory and field programs to obtain data on

the response of offshore and coastal facilities to extreme wind and seismic forces.

o Creating performance standards, design specifications, guidelines, and code

recommendations for application to new construction and remedial action for existing

574
facilities.

2. Accomplishments

a. Conducted two workshops on Wind and Earthquake engineering for offshore and

coastal facilities.

b. Published and distributed to Task Committee members the proceedings for the second

workshop.

3. Future Plans

a. All workshop proceedings will be published and made available to the U.S. side and

Japan side T/C members; copies will be provided to both sides Secretaries-General.

b. Develop workshop agenda of the third U.S.-Japan Workshop on Wind and Earthquake

Engineering for Offshore and Coastal Facilities in Japan in July 1997.

c. Coordinate, where possible, ongoing research on wind and earthquake engineering for

offshore and coastal facilities of interest to the members of the Task Committee.

•U.S. GOVERNMENT PRINTING OFFICE: 1996-404-525/67901

575

Technical Publications

Periodical

Journal of Research of the National Institute of Standards and Technology—Reports NIST research

and development in those disciplines of the physical and engineering sciences in which the Institute is active. These include physics, chemistry, engineering, mathematics, and computer sciences. Papers cover a broad range of subjects, with major emphasis on measurement methodology and the basic technology

underlying standardization. Also included from time to time are survey articles on topics closely related to the Institute's technical and scientific programs. Issued six times a year.

Nonperiodicals

Monographs—Major contributions to the technical literature on various subjects related to the

Institute's scientific and technical activities. Handbooks—Recommended codes of engineering and industrial practice (including safety codes) developed in cooperation with interested industries, professional organizations, and regulatory bodies.

Special Publications—Include proceedings of conferences sponsored by NIST, NIST annual reports, and

other special publications appropriate to this grouping such as wall charts, pocket cards, and bibliographies.

National Standard Reference Data Series—Provides quantitative data on the physical and chemical

properties of materials, compiled from the world's literature and critically evaluated. Developed under a worldwide program coordinated by NIST under the authority of the National Standard Data Act (Public

Law 90-396). NOTE: The Journal of Physical and Chemical Reference Data (JPCRD) is published

bimonthly for NIST by the American Chemical Society (ACS) and the American Institute of Physics (AIP).

Subscriptions, reprints, and supplements are available from ACS, 1155 Sixteenth St., NW, Washington, DC

20056.

Building Science Series—Disseminates technical information developed at the Institute on building

materials, components, systems, and whole structures. The series presents research results, test methods, and

performance criteria related to the structural and environmental functions and the durability and safety

characteristics of building elements and systems.

Technical Notes—Studies or reports which are complete in themselves but restrictive in their treatment of

a subject. Analogous to monographs but not so comprehensive in scope or definitive in treatment of the

subject area. Often serve as a vehicle for final reports of work performed at NIST under the sponsorship of

other government agencies.

Voluntary Product Standards—Developed under procedures published by the Department of Commerce

in Part 10, Title 15, of the Code of Federal Regulations. The standards establish nationally recognized

requirements for products, and provide all concerned interests with a basis for common understanding of

the characteristics of the products. NIST administers this program in support of the efforts of private-sector

standardizing organizations.

Order the following NIST publications—F1PS and NISTIRs—from the National Technical Information

Service, Springfield, VA 22161.

Federal Information Processing Standards Publications (FIPS PUB)—Publications in this series

collectively constitute the Federal Information Processing Standards Register. The Register serves as the

official source of information in the Federal Government regarding standards issued by NIST pursuant to the Federal Property and Administrative Services Act of 1949 as amended, Public Law 89-306 (79 Stat. 1127), and as implemented by Executive Order 11717 (38 FR 12315, dated May 11, 1973) and Part 6 of

Title 15 CFR (Code of Federal Regulations).

NIST Interagency Reports (NISTIR)—A special series of interim or final reports on work performed by

NIST for outside sponsors (both government and nongovernment). In general, initial distribution is handled

by the sponsor; public distribution is by the National Technical Information Service, Springfield, VA 22161,

in paper copy or microfiche form.
0>sa©U

T3

-o

© bO C O & 3

BU >—

o>

_

Q

g . o

E3

Z 8?ON

ON

00oQ

O (50 1 3 £ -s 8CIuD & I 1 £ « s _. 

.2 >> o — £ g o £