Wind and Seismic Effects

Proceedings of the 26th Joint Meeting

NIST SP 871

.U57

NO. 871

1994

U. S. DEPARTMENT OF COMMERCE

Technology Administration

National Institute of Standards and Technology
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J At Boulder, CO 80303.

2 Some elements at Boulder, CO 80303.
Wind and

Seismic

Effects

NIST SP 871

PROCEEDINGS OF

THE 26TH JOINT

MEETING OF

THE U.S.-JAPAN

COOPERATIVE PROGRAM

IN NATURAL RESOURCES

PANEL ON WIND AND

SEISMIC EFFECTS

Building and Fire Research Laboratory

National Institute of Standards and Technology

Gaithersburg, MD 20899

U.S. DEPARTMENT OF COMMERCE

Ronald H. Brown, Secretary

TECHNOLOGY ADMINISTRATION

Mary L. Good, Under Secretary for Technology

Issued September 1994

Noel J. Raufaste,

EDITOR

National Institute of Standards and Technology

Arati Prabhakar, Director
National Institute of Standards and Technology Special Publication 871

Natl. Inst. Stand. Technol. Spec. Publ. 871, 695 pages (Sept. 1994)

CODEN: NSPUE2

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1994

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, DC 20402-9325
PREFACE

This publication is Proceedings of the 26th Joint Meeting of the U.S. -Japan Panel on Wind

and Seismic Effects. The meeting was held at the National Institute of Standards and

Technology, Gaithersburg, Maryland during 17-20 May 1994. Forty-five papers were

authored~23 by U.S. members and 22 by Japanese. Thirty-four papers were presented

orally. The papers were organized into six themes: Wind Engineering; Earthquake

Engineering; Storm Surge and Tsunamis; Northridge Southern California and Hokkaido

Nansei-Oki Japan Earthquakes; Summary of Joint Cooperative Research Programs; and

Report of Task Committee Workshops conducted during the past year.

BACKGROUND

Responding to the need for improved engineering and scientific practices through exchange

of technical data and information, research personnel, and research equipment, the United

States and Japan in 1961 created the U.S. -Japan Cooperative Science Program. Three

collateral programs comprise the Cooperative Science Program. The U.S. -Japan Cooperative

Program in Natural Resources (UJNR), one of the three, was created in January 1964. The

objective of UJNR is to exchange information on research results and exchange scientists and

engineers in the area of natural resources for the benefit of both countries. UJNR is composed of 16 Panels each responsible for specific technical subjects.

The Panel on Wind and Seismic Effects was established in 1969. Seventeen U.S. and six

Japanese agencies participate with representatives of private sector organizations to develop

and exchange technologies aimed at reducing damages from high winds, earthquakes, storm

surge, and tsunamis. This work is produced through collaboration between U.S. and

Japanese member researchers working in 10 task committees. Each committee focuses on

specific technical issues, e.g., earthquake strong motion data. The Panel provides the

vehicle to exchange technical data and information on design and construction of civil

engineering lifelines, buildings, and waterfront structures, and to exchange high wind and

seismic measurement records. Annual meetings alternate between Japan and the United

States (odd numbered years in Japan; even numbered years in the United States). These one- week technical meetings provide the forum to discuss ongoing research and research results;

one-week technical study tours follow the meetings.

The National Institute of Standards and Technology (NIST) provides the U.S. -side chair and

secretariat. The Public Works Research Institute (PWRI), Japan, provides the Japan side

chair and secretariat.

Cooperative research is performed through formal Panel Programs. In 1981, cooperative

research in Large-Scale Testing was started under the auspices of the Panel. Also in 1981,

joint research on Reinforced Concrete Structures was initiated. Full-scale testing was

performed at the Building Research Institute (BRI), one of the six Japanese member

organizations, with supporting tests in Japan and in the United States. Two years later, a

joint research program on Steel Structures was initiated. Full-scale testing again was led by

BRI with supporting tests in the United States and Japan. The U.S. -Japan coordinated

iii
program for Masonry Building Research was started in 1985. A U.S. -Japan coordinated

program on Precast Seismic Structural Systems was initiated in 1991. A joint program on

Seismic Performance of Composite and Hybrid Structures was initiated in 1993. In 1994, a

joint program was initiated on Physical and Numerical Simulation of Structural Damages Due

to Liquefaction and Development of Countermeasure Techniques.

Task Committee meetings, exchanges of data and information through technical presentations

at annual Panel meetings, exchanges of guest researchers, visits to respective research

laboratories and informal interactions between Panel meetings, joint workshops and seminars,

and joint cooperative research programs all contribute to the development and effective

delivery of knowledge that has influenced design and construction practices in both countries.

Guest research exchanges have advanced the state of technology in areas of steel, concrete,

and masonry structures under seismic forces; developed techniques to analyze risks from

liquefaction; modeled water seepage in dam foundations; performed comparative analyses of

seismic design of U.S. and Japanese bridges.

Direct communication between counterpart country organizations is the cornerstone of the

Panel. Effective information exchanges and exchanges of personnel and equipment have

strengthened domestic programs of both countries. There are opportunities for experts in various technical fields to get to know their foreign counterparts, conduct informal

exchanges, bring their respective views to the frontiers of knowledge, and advance

knowledge of their specialties.

The Panel's activities resulted in improved building and bridge standards and codes and

design and construction practices in hydraulic structures in both countries, for example:

created and exchanged digitized earthquake records used as the basis of design and

research for Japan and the United States; transferred earthquake engineering information and strong-motion measurement

techniques for use by seismically active countries, e.g., Australia, Canada, Italy,

Mexico, Peru, Taiwan, Turkey, and North Africa;

produced data that advanced retrofit techniques for bridge structures; developed field test data for use in aerodynamic retrofit of bridge structures; produced full-scale test data that advanced seismic design standards for buildings;

advanced technology for repairing and strengthening reinforced concrete, steel, and

masonry structures; improved in-situ measurement methods for soil liquefaction and stability under

seismic loads;

created a database comparing Japanese and U.S. standard penetration tests to improve

prediction of soil liquefaction;

created database on storm surge and tsunamis and verified mathematical models of

tsunami and storm surge warning systems;

established a library resource of current research on wind and earthquake engineering

and on storm surge and tsunamis;

published proceedings of Panel meetings, Task Committee Workshops, and special

publications such as List of Panel Publications and translated two-volume series on

iv
earthquake resistant construction using base isolation systems;

gained better knowledge of both countries' research, design, and construction

capabilities from indepth visits to host country's laboratories and building and public

works projects. Results of such visits contribute to creation of new Task Committees,

agendas for Joint Panel meetings and task committee workshops, special visits of

U.S. -Japan researches, and joint collaborative research.

HIGHLIGHTS OF THE TECHNICAL SITE VISITS

Eight technical sites were visited at five locations. A summary of the visits follow.

GAITHERSBURG. MP

1. National Institute of Standards and Technology (NIST) 1

. The delegation was

provided an overview of the Building and Fire Research Laboratory (BFRL); tours of

BFRL's structures, materials, environmental, and fire facilities; NIST's Research Reactor

and its Cold Neutron Source of the NIST Materials and Engineering Laboratory; and an overview of the NIST Advanced Technology Program.

The mission of NIST is to promote U.S. economic growth by working with industry to develop and apply technology, measurements, and standards. BFRL, one of eight

Laboratories making up NIST, increases the competitiveness of industry and public safety

through performance prediction and measurement technologies and technical advances that

improve the life cycle quality of constructed facilities. BFRL's efforts are closely

coordinated with complementary activities of industry, professional and trade organizations,

academe, and other agencies of government. The vision for BFRL, the structure of its technical programs, and the determination and timing of its technical products are based on

analyses of industry needs and BFRL's own unique resources and capabilities.

Annually, BFRL publishes over 200 reports describing research findings from its over 150

research projects. Its projects, list of publications, impacts of its research, and examples of

how BFRL serves its customers are summarized in four companion reports Projects '94; Publications '93; Impacts; and NIST Building and Fire Research Laboratory - Collaborating

With Our Customers.

BFRL's laboratory facilities include: six-degree-of-freedom structural testing facility; largescale structural testing facility with the 53 MN (12-million pound) universal structural testing

machine; environmental chambers; guarded hot-plate; calibrated hot-box; plumbing tower;

building materials imaging and modeling laboratory; large burn facility for conducting

experimental fires in full-scale and related combustion toxicity facility, large industrial fire

test facilities, fire suppression test facilities; and a fire simulation laboratory.

For further information about NIST's work, contact Noel Raufaste, Secretary-General UJNR Panel on Wind and

Seismic Effects and Head, Cooperative Research Programs, NIST on facsimile 301-975-4032 or e-mail,

raufaste@micf.nist.gov.

V
In the Structures Division, the delegation was provided a summary of BFRL's research on

earthquake hazards reduction. Analytical and experimental studies are performed on seismic

behavior of masonry structures, seismic resistance of precast structures, residual strength and

energy absorbing resistance of precast concrete structures, strengthening methodologies for

buildings and bridge substructures, and performance requirements for passive and active

energy dissipation systems for buildings and lifeline structures. BFRL is mandated to perform post-disaster investigations; the findings are used to develop the most probable

technical causes of failures and to improve seismic design and construction of buildings and

structures against future earthquakes.

In the Building Materials Division, the delegation was provided a summary of work on the

development of methods to determine the quality and prediction of service lives of organic

building materials. Research includes identifying degradation mechanisms, improving

characterizing methods, and developing mathematical models of the degradation processes.

Research was discussed on developing computer models that simulate the development of the

microstructure of concrete during the setting process. Such models are used to predict

concrete performance, strength, and durability. The delegation discussed BFRL work in developing prototype expert systems such as HWYCON, an interactive expert system

designed to help highway inspectors and engineers diagnose problems, select materials for

construction, and repair highways and highway structures.

In the Building Environment Division, green technologies research was highlighted. BFRL

is exploring the use of refrigerant mixtures to improve the efficiency of refrigeration cycles

and to replace harmful chlorofluorocarbon refrigerants that damage the ozone layer of the

upper atmosphere. In a related area, the delegation toured BFRL's controls laboratory.

With the aid of a computerized energy monitoring and control system, this laboratory

performs fundamental research on heating, ventilating, and air conditioning (HVAC) control

systems, on control dynamics, and on adaptive optimization techniques. It uses a computerized energy monitoring and control system with BFRL software, which controls an

air handler, a building, a heating/cooling plant, and a laboratory test facility for evaluating

the performance of conventional and advanced HVAC/control systems. BFRL is fostering

the development of more intelligent, integrated, and optimized building mechanical systems.

A dynamic building/heating, ventilating, and air-conditioning control system simulation

program is used to study HVAC/control system dynamics and interactions. Research

addresses the development, evaluation, and testing of communication protocol standards for

the open exchange of information between equipment from different vendors and between

different control levels in hierarchical and distributed building management systems. Results

from this research serve as a basis for HVAC standards to assist the control system

manufacturers to develop interoperable systems and methods for testing conformance to the

standards.

In the Fire Safety Engineering Division, the delegation was exposed to its experimental

research on pool burning. This research addresses the characterization of the physical and

chemical properties of pool-burning flames as a function of pool diameter and liquid-fuel

molecular structure. The experimental program consists of measurements of flame

temperature, velocities, chemical-species composition, particulate characteristics, radiativevi
transfer characteristics, and energy feedback from the flame to the liquid fuel. A

demonstration was provided featuring a turbulent spray burner used to determine the fire

suppression capabilities of different gaseous agents, considered for possible halon

replacement (known to deplete stratospheric ozone).

The delegation visited NIST's 20-MW research reactor -- a national center for the application

of reactor radiation to a variety of problems. Each year, over 700 people from industry,

universities, and other government agencies use the experimental facilities at the reactor to perform collaborative and independent research projects. For example, researchers are using

neutrons produced by the reactor to study the structural properties of the new hightemperature superconductors and to determine how these properties are changed and related

to material processing. The Cold Neutron Facility makes a world-class, fully instrumented

laboratory for cold neutron research easily available for the first time to U.S. scientists

working in advanced materials science, chemistry, and biology. The measurement

capabilities of this facility are being used in a NIST Advanced Technology Program project

which aims to improve the properties while reducing the cost of finished ceramics. These

parts have a range of applications in a variety of industries, from high heat load engines to electronic components.

At the end of the day, NIST's Advanced Technology Program (ATP) was presented to the

delegation. ATP promotes rapid commercialization of new scientific discoveries. ATP

provides technology development funding for high risk technologies through cooperative

research agreements to single businesses or industry joint ventures. ATP supports

development of laboratory prototypes and proof of technical feasibility but not project

development or proof of commercial feasibility. ATP provides matching funds annually for

up to 5 years, not to exceed 50% of the total research.

BOSTON. MA

2. Massachusetts Institute of Technology. Center for Construction Research and

Education (CCRE)~. CCRE performs nonclassical civil engineering classwork and

contracted research directed at helping construction firms identify barriers to improving their

construction practices and recommending methods to resolve barriers. Examples of work

include developing procedures that forecast size of construction markets, identify pitfalls to performing successful construction practices, predict liability issues, develop construction

management techniques and environmental technologies such as hazardous waste reduction

including management of solid waste reduction, waste water treatments and airborne

pollution. Work in geotechnical engineering addresses fluid transport of toxic materials in water, earth, and rock, and work in structures and materials include composites, fiber

reinforced ceramics, NDEs for remote sensing of bridge decks and leaks in piping. CCRE's

work in intelligent highway vehicle systems (IHVS) Traffic Management Systems centers on

integrating those advanced technologies, developed for other industry applications, that have

potential to improve design and construction practices.

Dr. Fred Moavenzadeh, Director, Center for Construction Rcsearcli and Education hosted the delegation.

vii
CCRE is developing a Consortium for Infrastructure Development composed of 10 U.S.

firms to understand how different technologies interact within the infrastructure. CCRE

predicts infrastructure will be the next major U.S. design and construction challenge.

CCRE is part of the Pierce Laboratory, one of two Laboratories of the Department of Civil

and Environmental Engineering. It annually perform $10 million of contract research mostly

from the Federal Government. Their links with design and construction experts domestically

and internationally are strong and effective. Their outreach includes planning seminars and

conferences with domestic and foreign public and private construction organizations, hosting

foreign students and faculty and industry professionals to work on specific projects, sending

MIT students and faculty to U.S. construction firms and to other country laboratories, and

publishing a quarterly newsletter, Construction.

3. Central Artery (1-93)/ Tunnel (1-90) Project (CA/T) . This construction project will

replace Boston's 40 year old elevated Central Artery with a widened and mostly underground

8-10 lane interstate highway through downtown Boston and an east/ west Seaport Access Road

(tunnel) through South Boston to Logan Airport in the Third Boston Harbor Tunnel. CA/T

is under the direction of the Massachusetts Highway Department. Bechtel/Parsons

Brinckeroff are prime contractors managing CA/T's design and construction. Boston

designed their expressways in the 1950s to serve 75 K vehicles per day. Today it is overloaded and inefficiently serves 190 K vehicles daily. The new expressway is expected to accommodate 240 K vehicles daily. CA/T is the last link in the U.S. interstate highway

system; 12 km of urban highway, about 1/2 (5.6 km) in cut-and-cover tunnels. The overall

project is expected to cost $7.7 billion; 85% paid by Federal funds and 15% by State monies

derived from gasoline taxes and bonds.

The delegation visited the site of the 3rd Boston Harbor Tunnel. The actual construction for

this part of the CA/T Project began in September 1991 with the arrival of the huge Harbor

dredge "superscoop." The Boston Harbor crossing consisted of placing 12 double-boxed

immersed tubes in a 15 m deep by 30 m wide by 1.2 km long trench under Boston Harbor.

The associated east and south Boston land based cut and cover tunnel approaches to these

tubes will be completed at the end of 1995 in time for the Project's first phase opening in December 1995.

Modern Continental/Obayashi4

, the subcontractors to the Logan Airport end of the Third

Boston Harbor Tunnel, hosted this portion of the visit. The exit ramp required over 370 000 m

3 of concrete and 900 000 m

3 of material were excavated. The cut walls were

stabilized using a Japanese technique of soil/concrete mix of compressive strength about

300 Pa. Tie-backs range from 24 to 35 m in length. Soil from this site and other

construction sites are transported to several islands in the Boston Harbor for future use as parks. Boston citizens hope the largest Island, Spectacle Island, when completed will be

designated a National Park. The construction cost for this section is about $246 million.

Richard A. Jarvis, Public Outreach Coordinator hosted the delegation.

John Pastore, Assistant Project Manager hosted the delegation.

viii
The South Boston tunnel approaches, constructed by the joint venture of

Kiewit/Perini/Atkinson/Cashman, is 1.2 km in length; construction costs are about $180

million. A parking garage will be constructed on top of the south approach. While at the

Logan Airport approach, the covered space will revert back to the airport as tarmac and for

sound barrier buffer space. Caissons were driven to bed rock to prevent the roadway from

heaving. Slurry walls were placed to bedrock 37 m down. At each end of the approach, a 77 m diameter circular cofferdam was excavated to bedrock, 37 m down and concrete was

placed in eight lifts. The circular cofferdam is the meeting place for the East/South tubes

and will also serve as the foundation for Vent Building #6 in South Boston. Work is performed in 2 shifts of 10 hours each, 6-day s per week.

The highway approaches at Logan Airport and South Boston enter into an immersed

binocular tube tunnel made up of 12 concrete-lined steel tubes each measuring 98 m x 24 m

x 12 m high and weighing 2.3xl06 kg. A third tube, northbound from Boston to Logan

Airport, will serve as an HOV lane. A 30 m wide by 15 m deep trench was dredged to accommodate the submerged tunnel. The deepest section is 20 m below the harbor bottom.

Over 680,000 m

3 of rock and sediment were removed. Bethlehem Steel in Baltimore

fabricated the 12-section 2-tube tunnels, floated them to Boston and placed them using an

electronic positioning system by gantry cranes installed on two barges. The tunnel rests on

0.9 m bed of coarse gravel with 2 m ballast on top. Six meters of backfill line the sides of

the tunnel. Each tube was joined to its adjoining one by the pressure of the water

compressing its end gaskets. Workers inside the tube opened valves to drain water between

the end of the tubes. Watertight doors were removed and concrete cast around the inside of

the tubes. The outer steel walls include a 1.5 m by 2.1 m thick layer of reinforced concrete.

Tunnel concrete strength is 41 MPa. The road slab is 3 m thick with air ducts imbedded in

the slab. Fire detection systems are installed and emergency stairs are constructed into the

barrier walls separating the tunnels for driver/passenger egress into passageways.

A four-year environmental analysis was performed to consider appropriate methods for

transporting excavated materials to locations to dispose of the construction wastes. Over

9.2M m

3 of earth will be moved; some temporarily stored for later use as backfill; the

majority for permanent placement. About half of the 1.5M m

3 dredged material will be

dumped in the ocean 40 km offshore; excavated earth will be used as fill on several islands as noted before. CAD designs were used to help designers view the approaches and tunnel

through the perspective of the drivers.

The city center's new artery construction will be mostly underground (ranging from 24 m to 37 m deep) replacing the elevated 1-93 central artery. Work is expected to start in mid-1994.

Reinforced concrete walls of the cut and cover highway tunnel with a temporary steel structural system will support the elevated highway during construction of the underground

artery. Green space will cover the construction and help visually link neighborhoods that were separated by the elevated highway.

CAD designs simulated views of the project and showed relationships to adjoining

neighborhoods. Co-integration of the design and construction was performed with

professional societies such as AASHTO, ACI, ASCE, AISC. As there are no

ix
design/construction standards for underground highways, a modified AASHTO Standard was

used. Utility lines are now being relocated in Downtown Boston.

Working with the Federal Highway Administration and MIT, CA/T planners developed an

intelligent highway vehicle system (IHVS) to manage traffic by using techniques such as closed circuit television of traffic flow, CO0 monitoring, loop detection, electronic toll collection, randomly monitoring vehicles every 60 m to predict traffic delays, and ramp

signal traffic information. Traffic information will be displayed electronically at selected

shopping centers and broadcast on radio and television. An advanced public transportation

system will be installed at bus stops to display location of buses and a central control system

will manage transportation on the central artery.

CA/T management meets with citizen activist groups informing them about proposed design

and construction schemes and requesting their comments. CA/T's present plan reflects

reductions from potential vehicle pollution and improvements in highway/bridge architectural

appearances from such meetings. For example, the delegation was told about Scheme Z, a 37 m high six loop $350 million highway interchange. Citizens voiced concern about

expected air and noise pollution from the many vehicles and about the unsightly 12 story

highway ramps. After much public discussion, the interchange design was modified to a

12 m high two loop highway interchange above ground with other loops underground. The

revised costs, however, were three times the original cost, now totaling $1.2 billion.

The Massachusetts Highway Department highlighted some "innovated" design and

construction practices used on this project:

• a soil/concrete mix was used to stabilize earth serving as retaining walls to the 3rd

Harbor Tunnel resulting in economy over the traditional steel shoring procedures,

• caissons stabilizing entry roads permitted use of thinner slab thicknesses which

reduced project costs,

• micro tunnel jacking to install utility systems permitted installations at shallower

depths resulting in less excavation and a lower cost,

• cofferdams at tunnel ends facilitated highway connections at the tunnel,

• temporary beam structural system used to support existing elevated highways over new cut and cover tunnel construction permits normal traffic flow during period of the

construction,

• adaptive reuse by capping 27 landfills and islands in Boston Harbor with excavated

soil will result in new public parks,

rodent control officials are educating citizens about controlling displaced rodents

during construction,

discussions with citizen interest groups resulted in revised CA/T design and

construction plans by incorporating citizens' recommendations.

x
4. The Boston Harbor Project Deer Island Treatment Plant5

.

In the 1980's Boston Harbor was one of the most polluted in the United States. Covering

nearly 130xl06 m

2

, the harbor has been a dumping ground for all sorts of waste. In 1985,

the Massachusetts Legislature created the Massachusetts Water Resource Authority (MWRA)

to supply water and sewerage services to 61 Boston-area cities and towns (more than 2

million people) and to improve the water quality of Boston Harbor. Interim improvements

have included the cessation of scum discharges (1989) and sludge discharges (1991) from the

existing treatment plant at Deer Island. MWRA is in the process of upgrading its sewage

treatment operations. This facility on Deer Island, at the mouth of Boston Harbor, will be

replaced by a 4.81 million m

3 per day plant, the 2nd largest facility in the United States. A

new secondary sewerage treatment facility will be completed in 1994. Eight sludge digesters

have been constructed to come on-line with the primary plant; one will be used for methane

gas. Eight additional digesters will be installed from 1994-1999 to service the secondary

plant. Each egg-shaped digester will control 1 1 400 m

3 of sludge.

Instead of discharging effluent into the Harbor, it will be transported through a 7.5 m wide

outfall tunnel drilled in rock 133 m below the sea bed and connected to 55 diffusers at the

end of the tunnel 15.2 km out into the Massachusetts Bay. Tunneling proceeds at about

37 m per day performed in three, 8-hour shifts. About 6.5 km of tunneling have been

completed towards a planned completion in July 1996. The $200-million tunnel is a joint

venture of Kiewit, Atkinson, and Kenny Construction Companies.

BLACKSBURG. VA

5. Virginia Polytechnic Institute and State University, Center for Intelligent Material

Systems and Structures, Fiber and Electro-Optics Center, and Center for Construction

Engineering .

In the Intelligent Material Systems and Structures Laboratory6 the delegation discussed

several projects:

• Determining phase transfer characteristics of Shape Memory Alloys (SMA) to improve impact tolerance of SMA hybrid composites. By embedding small amounts

of fibrous SMA materials into brittle composite materials, greater composite impact

damage toughness can be achieved. Impact strain energy is more readily absorbed by

their high-strain-to-failure two-phase SMA materials than by the brittle host

composites and is not available to initiate damage in the host composite material.

• Detecting damage in structures using direct and converse electromechanical properties

of piezoelectric materials. This approach provides opportunities for the simultaneous

actuation and sensing of structures (e.g., collocated actuator/sensor). The variation in the electrical impedance of a piezoelectric patch bonded to a structure, while driven

Tim Watkins, Public Information Coordinator, hosted the delegation.

Dr. Zaffir A. Choudhry, Center for Intelligent Material Systems and Structures, hosted the delegation.

xi
by a fixed alternating electric field over a frequency range, provides valuable

information about the structural dynamic behavior. This information, which is analogous to the frequency response and easily obtainable, can use vibration

signatures-based non-destructive evaluation. This work is expected to have utility in damping antennae and controlling vibrations and noise in robotic systems, space

platforms, and machinery.

• Developing vibration signature based damage detection techniques which allow for the

localization of the actuation and sensing area. In experiments with a truss structure,

and using a single piezoelectric patch for actuation and sensing, localization was shown to be possible at higher frequencies typically from 20 to 100 kHz.

Developing low-cost variable friction electrostatic damper for active/passive vibration

control.

• Developing active and passive material tagging concepts for health monitoring of

structures. Tagging involves adding a small amount of micron-sized "tagging" sensor

particles to the host material and using them to determine the state-of-the-material.

Active tagging involves exciting the sensing particles and measuring the response of

the host material. Passive tagging involves sensing the distribution of the particles.

An example of a passive technique is adding magnetic particles to an adhesive and

using an eddy current probe to detect voids in a tagged adhesive bond. Applications

of passive and active tagging techniques include characterization of adhesive bonds,

cure monitoring, intelligent processing, nondestructive materials evaluation, damage

detection, and in-service health monitoring.

At the Fiber and Optical Research Center7 the delegation discussed work in:

• Monitoring building system's temperature and state-of-health using embedded optical

fiber sensors. Impact monitoring using time of flight signal processing was

demonstrated to determine when and where a structure is impacted. Acoustic waves

are measured using the optical fiber sensors embedded in the structure. Measurements also are being performed to determine a structure's plastic

deformation, elongation, etc. This work involves post processing of the data;

research is underway toward real time processing.

• Detecting corrosion on aluminum substrates under coatings using optical fiber sensing.

This work focuses on monitoring corrosion on aircraft parts to reduce failure

probability and downtime for inspection and repair.

0 Performing experiments using sapphire fiber with coatings such as titanium oxide as a

tool for high temperatures (i.e. greater than 1000 °C).

At the Center for Construction Engineering8

, the delegation was shown a demonstration of

their Laser Based Digital Positioning System, a metrology based technology. The system

makes 3-D measurements digitally in real time with accuracy of 1 mm at 100 m. The

system must use at least two fixed laser transmitters whose signals are recorded by a survey

Mr. Mark Miller, Research Associate hosted the delegation.

Dr. Yvan Beliveau, hosted the delegation.

xii
rod equipped with a receiver and processor. The information is processed and analyzed to provide the dimensional location of the desired point. The digital positioning data is entered

to CAD systems to create real time as-built models. The reverse permits locating or laying

out points in space. The system can be configured to self-calibrate.

The positioning system is being tested to measure pavement leveling, curb lanes, fine grading

for slopes, and equipment automation. It is claimed to have applications to model full vibrations of bridges or other structures such as space stations, controlling automated work

systems such as heavy equipment operation for earthwork operation, and used for just-in-time

fabrication.

This technology is being coordinated with the Civil Engineering Research Foundation and the

CORPS through a Cooperative Research Development Agreement (CRADA) program. A

consortium has been formed to bring together this CRADA; members include Spatial

Positioning Systems (a spin-off business), Bechtel Corporation, Jacobus Technologies,

National Science Foundation, Intergraph, and DuPont.

VICKSBURG. MS

6. Waterways Experiment Station

The delegation was hosted by the U.S. Army Corps of Engineers' Waterways Experiment

Station9 (WES). Located in Vicksburg, Mississippi, WES is an Army research and

development facility, comprised of five laboratories and a center. WES was established by

the U.S. Army in 1929. It's annual budget is over $360 million. It's staff includes about

1500 civilian employees and several members of the armed forces.

The program at WES included a mini-symposium and visits to the Geotechnical Laboratory,

the Structures Laboratory, and the Information Technology Laboratory (ITL). The

symposium included presentations by Mr. Vincent Chiarito on Seismic Evaluation of

Reinforced Concrete Intake Structures; by Dr. Takashi Iijima on Overview of Research

Performed by the Public Works Research Institute, and by Dr. Kazuhiko Kawashima on Ground Motion Characteristics Based on Analysis of Strong Motion Records. WES staff

learned that a number of PWRI's research programs are similar to theirs, and showed strong

interest in collaborative research with PWRI.

The world's largest centrifuge is under construction at the Geotechnical Laboratory. It has a

radius of 6.5 m, a maximum payload at 350 g of 2000 kg, and a maximum g-force of

1,144,000 kg. It is scheduled to be in operation in early 1995. ITL is designated as DoD

High Performance Computing Center. One of the highlights of the visit was a demonstration

of a virtual reality walk through the ITL. In the Structures Laboratory, computer-based

analytical modeling is extensively used to solve many of the solution-oriented tasks. It also

provided technical expertise in the Corps-wide effort to evaluate, rehabilitate, and maintain

Dr. A.G. Franklin, Chief Earthquake Engineering and Geophysics Division, WES scheduled the visit.

xiii
many aging hydraulic structures operated by the Corps.

LOS ANGELES. CA

7. Damage sites from Northridge Earthquake.

At the California Department of Transportation's (CALTRANS) District 7 Office, 10 the

delegation was briefed on the status of CALTRANS repairs to damaged bridges and

highways from the Northridge Earthquake. Following the briefing, CALTRANS hosted a

visit to three bridge sites. Within days after the earthquake, CALTRANS estimated highway

and bridge repairs would cost more than $670 million. In May 1994, these estimates were

revised to $308 million because many bridges and highways only required repairs, not full demolition and rebuilding, and there was less roadway damage than originally estimated.

The Governor of California enacted emergency procurement permitting CALTRANS to award contracts after the lowest bid was accepted for repairs, demolition and construction of

highways and bridges to minimize down time. Generally, a short list of five bidders having

proved experience with CALTRANS and the ability to finance portions of the job (would be

later reimbursed) were accepted. CALTRANS incorporated bonuses and penalties in many

of these contracts; some bonuses and penalties were as large as $200 thousand per day. One

contractor saved many days of proposed construction schedule by renting a train to haul

construction equipment and materials from the source to Los Angeles.

Immediately after the earthquake, CALTRANS began identifying alternative traffic routes

around damaged highways and bridges. CALTRANS daily adjusted traffic routings as debris was cleared and repairs completed. Citizens also offered recommendations to speed traffic

flow, which CALTRANS implemented.

The status of the three bridges the delegation visited are:

• Interchange I-5/Route 14 (Golden State/Antelope Freeway) now accommodates all vehicle traffic on its truck bypass. Two contracts were let for the demolition of the

three most seriously damaged overhead ramps. Work is expected to be completed in

July 1994. The five other damaged bridge superstructures (less severely damaged

than the first three) will be destroyed and rebuilt one at a time permitting vehicle

traffic use of the others.

• Eastbound spans of Route 118 (Simi Valley Freeway) at Mission Boulevard and

Gothic Avenue and at Bull Creek Canyon have been replaced. The eastbound

structure is in the process of being replaced. These bridges were designed to a site specific earthquake spectra based on a maximum magnitude of 7.5. The maximum

credible earthquake for this specific area has not been adjusted due to the Northridge

Earthquake; however, the new structures are designed to meet present seismic

criteria. The bridge decks were widened to accommodate a future HOV lane.

• I- 1 0 and LaCienega bridge is open. The bridge was designed with shear keys to

Mr. Vincent Moreno, Public Affairs Office, District 7, hosted the delegation.

xiv
reduce lateral movement; lack of adequate confinement was a major factor in its extensive damage.

Other bridge repairs are based on their original design specifications; however, footings are being strengthened by increasing their size and adding additional reinforcement.

8. Seven Oaks Dam Project .

The U.S. Army Corps of Engineers 11 hosted the delegation to this earth/rock fill dam site that will control the flood plain downstream of the Santa Ana River. Flood water will be

diverted, as needed, into an open space below the dam to provide a natural flood way and

into a wildlife habitat corridor. Fourteen floods occurred in this region during the 1900s. In the past, they flooded over 44,500 ha of land to an average depth of 1 m. The dam when

completed will be 192 m high (167 m above and 24 m feet below ground), its crest length is 914 m and 13 m wide at the top. Its detached spillway will be located about 0.5 km east of

the dam; water will drain into a neighboring valley. The dam's reservoir is designed for

about 60,000 ha. The inlet structure will screen sediment before traveling through the 61 m

high, 500 m outlet tunnel, the intake tower, air shaft, exit tunnel, and finally the plunge

pool. The project's cost is $1.4 billion; $460 million for the dam and $940 million to improve the channel to the Pacific Ocean and raise the Pravo Dam and enlarge its Reservoir

located downstream of the Seven Oaks Dam.

The delegation visited the dam construction site. Some construction has started in preparation for the actual dam. The coffer dam is under construction, pilot holes were

drilled for the air shaft, the spillway's foundation was grouted, the intake structure is nearing

completion.

Forty workers are now involved in the construction; they will increase to 300 as work on the

dam itself is started. Estimated date of project completion is between 1998 and 2000.

Noel J. Raufaste, Secretary-General

U.S. -Side Panel on Wind and Seismic Effects

Dr. William Roper, Assistant Director R&D (Civil Works) Corps of Engineers scheduled the visit.

XV
ABSTRACT

This publication is the Proceedings of the 26th Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects. The meeting was held at the National Institute of Standards and

Technology, Gaithersburg, Maryland, during May 17-20, 1994. The Proceedings include the

program, list of members, panel resolutions, task committee reports, and 45 technical papers.

The papers were presented under six themes: (I) - Wind Engineering, (II) - Earthquake

Engineering, (III) - Storm Surge and Tsunamis, (IV) - Joint Cooperative Research Program,

(V) - Northridge Southern California and Hokkaido Nansei-Oki Earthquakes and (VI) - Summaries of Task Committee Workshop Reports (oral presentations only).

KEYWORDS: accelerograph; bridges; building technology; disaster reduction; earthquakes;

geotechnical engineering; ground failures; lifelines; liquefaction; masonry; repair and retrofit;

risk assessment; seismic; soils; standards; storm surge; structural engineering; tsunamis; and

wind loads.

xvi
CONTENTS

PAGE

PREFACE in ABSTRACT xvi

AGENDA FOR 26TH JOINT UJNR MEETING xxm LIST OF PANEL MEMBERS xxxm LIST OF TASK COMMITTEE MEMBERS xlix

RESOLUTIONS liii

THEME: WIND ENGINEERING

Wind Tunnel Experiment for a Super Long-Span

Cable-Stayed Bridge in Smooth Flow

Hiroshi Sato, Ryuichi Toriumi,

Makoto Kitagawa, and Hiroshi Katsuchi 3

Wind Tunnel Tests on Tapered Cylinders for

Highway Support Structures

Harold R. Bosch 17

Standards to Resist Hurricane Wind Forces

G. Robert Fuller and R. D. Marshall 31

Inauguration of International Wind Engineering Forum and

New U.S.-Japan Cooperative Research Program

Hisashi Okada and Hiroyuki Yamanouchi 39

THEME: EARTHQUAKE ENGINEERING

A Seismic Retrofitting Manual for Highway

Bridges

Ian G. Buckle, Ian M. Friedland, and James D. Cooper 47

Development of Innovative Seismic Response

Control of Bridges

Kazuhiko Kawashima, Shigeki Unjoh, and Hidetoshi Mukai 59

xvii
Innovative Seismic Testing Procedures

Stephen A. Mahin and S. C. Liu 81

Standard Test Procedures for Seismic

Isolation Systems

Harry W. Shenton, III 97

i .-:!!.' "-i'ii IM Arts in Japan Concerned With

Application of Carbon Fiber for Rehabilitation

and Strengthening for Concrete Bridges

Nishikawa Kazuhiro, and Koga Masajiro 109

Const ruction A utomation Research

K. P. Chong and S. C. Liu 123

Design Guidelines for Precast Reinforced

Concrete Buildings

Shinsuke Nakata, Masaomi Teshigawara

Takashi Kaminosono, and Hitoshi Shiohara 133

A Post-Earthquake System for Estimating Seismic

Damage to Road Facilities by a Strong Motion

Observation Network

Hisanori Otsuka, Ryoji Hagiwara and Yoshio Ninomiya 139

Use of Probabilistic Earthquake Models for

Design and Evaluation

Jeffrey K. Kimball and Ann Bieniawski 151

Structural Model of Functional Damage Due

to an Eai thquake in Urban Area

Kazuhiko Kawashima, Hideki Sugita,

Tomoru Nakajima, and Takashi Sato 175

Dynamic Behavior of Pile Foundation in

I iquefaction Process

T. Kagawa, C. Minowa, and H. Mizuno 187

Suitability of SASW Method in Complex Geosystems

D. W. Sykora, J. M. Roesset, and K. H. Stokoe II 199

Effectiveness of Measures Against Liquefaction

Confirmed at a Recent Earthquake—A Case History

During 1993 Kushiro-Oki Earthquake

S. Iai, Y. Matsunaga, T. Morita, and H. Sakurai 213

Review of Progress in Dynamic Geotechnical

Centrifuge Research

Ronald F. Scott 227

Shaking Table Failure Tests of Actual Size

RC Structures

Chikahiro Minowams, Keiichi Ohtani, Nobuyuki Ogawa,

Tadashi Mikoshiba, Toshihiro Hayashida, and Tsuneo Okada 247

xviii
Response of a Mid-Water Depth Platform Subject

to Earthquakes

Charles E. Smith 267

Uplift of Sewage Manholes During 1993 Kushiro-Oki

Earthquake

Osamu Matsuo, Hisanori Otsuka, Yoshio Ninomiya, and Junichi Koseki 285

Survey of Research in the Area of Seismic Response

of Concrete Dams

Robert L. Hall, and William E. Roper, 297

Damage and Behavior of Dams During the 1993 Kushiro

Oki Earthquake and the 1993 Hokkaido Nansei Oki

Earthquake

Akira Nakamura, Nario Yasuda, and Tomoya Iwashita 305

THEME: STORM SURGE AND TSUNAMIS

Evaluation of Seismic and Flood Hazards Potential

for a Large Earth-Filled Dam

N. Simon, M. Reich, P. C. Wang, R. Khanbilvardi, and T. McSpadden 329

Tsunami Runup Distribution Generated by the July 12, 1993,

Hokkaido-Nansei-Oki Earthquake

E. N. Bernard and F. I. Gonzalez ' 335

Propagation of Hokkaido-Nansei-Oki Earthquake

Tsunami Around Cape Aonae

Shigenobu Tanaka, Shinji Sato, and Kenji Noguchi 345

Numerical Modeling Tsunamis

Philip L.-F. Liu 357

THEME: SUMMARY OF JOINT COOPERATIVE RESEARCH PROGRAMS

The Status of the U.S. Precast Seismic Structural

Systems (PRESSS) Program

M. J. N. Priestly and H. S. Lew 365

The Present and Future of the United States-Japan

Cooperative Earthquake Research Program Utilizing

Large-Scale Testing Facilities

Hiroyuki Yamanouchi 369

xix
THEME: NORTHRIDGE SOUTHERN CALIFORNIA AND HOKKAIDO

NANSEI-OKI EARTHQUAKES

The January 17, 1994, Northridge Earthquake, California

H. S. Lew, Jim Cooper, Sam Hacopian, Walter Hays,

and Mike Mahoney 375

Summary Report on the Northridge, South California,

U.S.A., Earthquake of January, 1994

Yukihiko Sumiyoshi, Tomomitsu Fujii, Kazuhiko Kawashima,

Tatsuo Uwabe, Kouichi Koshiumi, Eiichi Itoigawa, Yoshiaki Nakamura,

Kimihiko Izumi, and Hideki Sugita 427

The M8.1 Guam Earthquake of August 8, 1993

A. Gerald Brady 461

Infrastructure Damage by 1993 Hokkaido-Nansei-Oki

Earthquake

Takashi Iijima, Yutaka Iida, Kazuhiko Kawashima,

Hisanori Otsuka, Osamu Matsuo, Shigenobu Tanaka,

Akira Nakamura, Masanori Nakano, and Ryosuke Tsunaki 467

12 July 1993 Hokkaido-Nansei-Oki Earthquake

Riley M. Chung 499

MANUSCRIPTS AUTHORED FOR PANEL MEETING BUT NOT PRESENTED ORALLY

Technological Development for Automatic Construction

Methods of Caissons

Masanori Nakano, Jiro Fukui, and Shigeru Takagi 529

Strong-Motion Records on the Hokkaido-Nansei-Oki

Earthquake in 1993

Masashi Satoh, Yuji Ono, Toshio Yamauchi, and Hiroaki Nishi 541

Development of Integrated Numerical Research System

for Prevention and Estimation of Coastal Disaster

Tomotsuka Takayama, Chiaki Goto, Hidenori Shibaki, and Toshio Aono 555

Structural Testing of Confined Masonry for the

Third World Housing

Hatsukazu Mizuno, Tetsuro Goto, Masanori Iiba, Hiroto Kato,

and Yutaka Yamazaki 569

Guideline for Safety Evaluation of Structural

Response-Control Building (Tentative)

Kitagawa Yoshikazu, Midorikawa Mitsumasa,

Iiba Masanori, Fujitani Hideo, and Kashima Toshihide 581

Tsunami Disaster Caused by Hokkaido-Nansei-Oki

Earthquake in 1993

Shunji Niwa, Takekazu Akagiri, Masakatsu Horino,

and Masaharu Tsuzawa 591

xx
The Dam Safety Program

William S. Bivins and Harold W. Andress 601

Multi-Hazard Mitigation for Non-Engineered Structures

Michael Mahoney 607

Use of Structural Response Data from Small

Earthquakes and Aftershocks

Erdal §afak 613

Three Recent Tsunamis—Warning System Response

Michael E. Blackford 625

Hurricane Storm Surge Analysis for the Coast of Delaware

David J. Mark, Norman W. Scheffner, and Keith D. Watson 629

APPENDIX: TASK COMMITTEE REPORTS 649

xxi

Agenda

26th Technical Panel Meeting on Wind and Seismic Effects

16-20 May 1994

Monday 16 May

0930 Welcome and Overview of BFRL

Richard N. Wright, Director, BFRL

1005 Overview Structures Division - H.S. Lew, Chief

Laboratory B-135

1020 Overview Building Materials Division - Geoffrey Frohnsdorff, Chief

Quality Control/Quality Assurance (B-213)

Expert Systems and Materials Reference Laboratory

Image Processing and Modeling (B-337)

1115 Overview Fire Research - Ken Steckler, Physicist

Discuss Pool Fires, Other Topics

1135 Discuss Fire Suppression - Anthony Hamins, Mechanical Engineer

1150 Lunch

1330 Cold Neutron Research Facility - Michael Rowe, Chief, Reactor Radiation Division

Neutron Nondestructive Evaluation

Polymer Measurements. Other Topics

1410 NIST's Advanced Technology Program - James Hill, Head, ATP's Construction Program

1435 Green Technologies - James Hill, Chief, Building Environment Division

Refrigerant Laboratory (B-129)

Building Controls Laboratory (B-121)

1500 Wrap-Up - Noel Raufaste, Head, Cooperative Research Programs

BFRL Conference Room (B-22 1

)

xxiii
Tuesday 17 May

1000 OPENING CEREMONIES

(Lecture Room B, Administration Building)

Call to order by Noel RAUFASTE, Secretary-General US Side, UJNR Panel

Opening remarks by Raymond KAMMER, Deputy Director, National Institute of

Standards and Technology

Remarks by Yukihide HAYASHI, Counsellor for Science and Technology, Embassy of

Japan

Remarks by Richard N. WRIGHT, Chairman US-Side, Panel on Wind and Seismic

Effects, Director, Building and Fire Research Laboratory

Remarks by Takashi IIJIMA, Chairman Japan-Side, Panel on Wind and Seismic Effects,

Director-General, Public Works Research Institute

Introduction of U.S. Members by U.S. Panel Chairman

Introduction of Japan Members by Japan Panel Chairman

Elect Joint Meeting Chairman

Adopt Agenda

Adjourn

1115 Group Photograph

1130 Lunch: Hosted by the National Institute of Standards and Technology, Raymond

KAMMER, Deputy Director

xxiv
THEME - WIND ENGINEERING

1230-1410 Technical Session - Wind Engineering

Chairman: Dr. Takashi IIJIMA

1230 Wind Tunnel Experiment for a Super Long-Span Cable-Stayed Bridge in Smooth Flow,

Hiroshi SATO*, Ryuichi TORIUMI, Makoto KITAGAWA, PWRI, and Hiroshi

KATSUCHI, HSBA

1250 Wind Tunnel Tests on Tapered Cylinders for Highway Support Structures, Harold R. BOSCH, FHWA

1310 Standards to Better Resist the Effects of Hurricanes and Tornadoes, Richard

MARSHALL*, NIST and Robert FULLER, HUD

1330 Inauguration of International Wind Engineering Forum and New U.S. -Japan Cooperative

Research Program, Hisashi OKADA* and Hiroyuki YAMANOUCHI, BRI

1350 Discussion

1410 Break

TASK COMMITTEE MEETINGS

1430-1650 Task Committee Meetings

T/C A Strong Motion Data and Applications (G. Brady/S. Iai, Chairmen)

Room Bl 11 Administration Building

T/C C Evaluation and Improvement of Structures (K. Chong/Okada, Chairmen)

Lecture Room B Administration Building

T/C D Dam Earthquake Engineering (W. Roper/K. Nishikawa, Chairmen)

Room B 1 13 Administration Building

T/C H Soil Behavior and Stability During Earthquakes (G. Franklin/K. Kawashima, Chairmen)

Room A369/Building 224

T/C I Storm Surge and Tsunami (H. Meyers/H. Sato, Chairmen)

Room B221/Building 226

1650 Adjourn for day

*Presenter

XXV
Wednesday 18 May

THEME - EARTHQUAKE ENGINEERING

0830-1130 Technical Session - Earthquake Engineering Part 1

Chairman: Dr. Richard WRIGHT

0830 A Seismic Retrofitting Manual for Highway Bridges, Ian G. BUCKLE, Ian M.

FRIEDLAND (NCEER), and James D. COOPER, FHWA

0850 Development of Innovative Response Control of Bridges, Kazuhiko KAWASHIMA*,

Shigeki UNJOH, and Hidetoshi MUKAI, PWRI

0910 Innovative Seismic Experimental Methods, Steve MAHIN*, UC Berkeley and S.C. LIU,

NSF

0930 Discussion

0950 Break

1010 Standard Test Procedures for Seismic Isolation Systems, Harry SHENTON, III, NIST

1030 State-of-the-Art in Japan Concerning the Application of Carbon Fiber for the Rehabilitation and Strengthening of Concrete Bridges, Kazuhiro NISHIKAWA*, PWRI and

Masajiro KOGA, Obayashi Corporation

1050 Construction Automation Research, Ken P. CHONG* and S.C. LIU, NSF

1110 Discussion

1130 Adjourn for Lunch

1130 Lunch: Hosted by Federal Emergency Management Agency, Richard Moore, Associate

Director for Mitigation

1230-1410 Technical Session - Earthquake Engineering

Chairman: Dr. Richard WRIGHT

1230 Design Guidelines for Precast Reinforced Concrete Buildings, Shinsuke NAKATA,

Masaomi TESHIGAWARA, Takashi KAMINOSONO, and Hitoshi SHIOBARA*, BRI

1250 A Predicting System for the Seismic Damage of Road Facilities by a Strong Motion

Observation Network, Hisanori OTSUKA and Yoshio NINOMIYA, PWRI (presented by

Koichi YOKOYAMA*)

1310 Use of Probabilistic Earthquake Models for Design and Evaluation, Jeffrey K.

KIMBALL* and Ann BIENIAWSKI, DOE/DP

1330 Structural Model on Functional Damage Due to an Earthquake in Urban Area, Kazuhiko

KAWASHIMA, Hideki SUGITA*, Tomoru NAKAJIMA, and Takashi SATO, PWRI

1350 Discussion

1410 Break

xxvi
TASK COMMITTEE MEETINGS

1430-1700 Task Committee Meetings

T/C B Testing and Evaluation Procedures for Building Systems (H.S. Lew/K. Ohtani,

Chairmen)

Room Bill Administration Building

T/C F Disaster Prevention Methods for Lifeline Systems (R. Chung/K. Kawashima, Chairmen)

Room B 1 13 Administration Building

T/C G Structural Control and Intelligent Material Systems (S.C. Liu/Shiohara, Chairmen)

Room A369/Building 224

T/C J Wind and Earthquake Engineering for Transportation Systems (J. Cooper/K. Yokoyama,

Chairmen)

Lecture Room B, Administration Building

T/C K Wind and Earthquake Engineering for Offshore and Coastal Facilities (C. Smith/S. Iai, Chairmen)

Room B221/Building 226

1700 Adjourn for day

xxvii
Thursday 19 May

THEME - EARTHQUAKE ENGINEERING - PART II

0830-1130 Technical Session - Earthquake Engineering-Part II Chairman: Dr. Takashi IIJIMA

0830 Dynamic Behavior of Pile Foundation in Liquefaction Process, Chikahiro MINOWA*,

Hatsukazu MIZUNO, and Takaaki KAGAWA, NIED

0850 Suitability of SASW Method in Complex Geosystems, David W. SYKORA, WoodwardClyde Consultants, Jose M. ROESSET, and Kenneth H. STOKOE II, WES

0910 Effectiveness of Measures Against Liquefaction Confirmed at a Recent Earthquake - A

Case History During 1993 Kushiro-Oki Earthquake, Susumu IAI*, Yasuo

MATSUNAGA, Toshikazu MORITA, and Hirotaka SAKURAI, PHRI

0930 Discussion

0950 Break

1010 Review of Progress in Dynamic Geotechnical Centrifuge Research, Ronald F. SCOTT,

California Institute of Technology

1030 Failure Tests of Actual Size RC Structures, Keiichi OHTANI*, Nobuyuki OGAWA,

Tadashi MIKOSHIBA, Chikahiro MINOWA, Toshihiro HAYASHIDA, and Tsuneo

OKADA, NIED

1050 Response of Mid-Water Platforms Subjected to Earthquakes, Charles SMITH, MMS

1110 Discussion

1130 Adjourn for lunch

1130 Lunch: Hosted by U.S. Geological Survey, Robert L. Wesson, Chief, Office of

Earthquakes, Volcanoes, and Engineering

1230-1410 Earthquake Engineering (Continued)

Chairman: Dr. Takashi IIJIMA

1230 Uplift of Sewage Manholes During 1993 Kushiro-oki Earthquake, Osamu MATSUO,

Junichi KOSEKI, Hisanori OTSUKA, and Yoshio Ninomiya, PWRI (presented by

Masayuki KANDA)

1250 Survey of Research in the Area of Seismic Response of Concrete Dams, Robert L. HALL, WES and William E. ROPER*, CORPS

1310 Behaviours and Damages of Dams During 1993 Kushiro Oki Earthquake and 1993

Hokkaido Nansei Oki Earthquake, Akira NAKAMURA, Nario YASUDA, and Tomoya

IWASHITA, PWRI (presented by Shigetoshi KOBAYASHI)

1330 Evaluation of Seismic and Flood Hazards Potential for a Large Earth-Filled Dam,

Nick SIMOS, BNL, Morris REICH, BNL, Ping WENG, BNL, Reza KHANBILVARD,

BNL, and Tom MCSPADDEN, DOE/EH

xxviii
1350 Discussion

1410 Break

THEME - STORM SURGE AND TSUNAMIS

Chairman: Dr. Takashi IIJIMA

1430-1550 Technical Session - Storm Surge and Tsunamis

1430 Tsunami Runup Distribution Generated by the July 12, 1993, Hokkaido-Nansei-Oki

Earthquake, Eddie N. BERNARD and Frank I. GONZALEZ, NOAA (presented by

Michael BLACKFORD)

1450 Propagation of Hokkaido-Nansei-Oki Earthquake Tsunami Around Cape Aonae,

Shigenobu TANAKA and Kenji NOGUCHI, PWRI (presented by Takashi IIJIMA)

1510 Numerical Modeling of Tsunamis, Philip LIU, Cornell University

1530 Discussion

1550 Break

THEME - SUMMARY OF JOINT COOPERATIVE RESEARCH PROGRAMS

Chairman: Dr. Takashi IIJIMA

1610-1710 Technical Session - Joint Cooperative Research Programs

1610 U.S. -Japan Cooperative Research in Precast Seismic Structural Systems (PRESSS)

Program-Status Report, Nigel PRIESTLY, UCSD and H.S. LEW*, NIST

1630 The Present and Future of the U.S. -Japan Cooperative Earthquake Research Program

Utilizing Large-Scale Testing Facilities, Hiroyuki YAMANOUCHI, BRI (presented by

Hisashi OKADA)

1650 Discussions

1710 Adjourn for day

xxix
Friday 20 May

THEME - NORTHRIDGE SOUTHERN CALIFORNIA AND HOKKAIDO

NANSEI-OKI EARTHQUAKES

Chairman: Dr. Richard WRIGHT

0815-1030 Technical Session - Northridge Southern California and Hokkaido NanseiOki Earthquakes

0815 Summary U.S. -Side Investigation of Damages from Northridge Southern California

Earthquake, H.S. LEW, NIST

0835 Summary Japan-Side Investigation of Damages from Northridge Southern California

Earthquake, Yukihiko SUMIYOSHI, Tomomitsu FUJII, Kazuhiko KAWASHIMA*,

Tatsuo UWABE, Kouichi TORIUMI, Eiichi ITOIGAWA, Yoshiaki NAKAMURA,

Kimihiko IZUMI, and Hideki SUGITA, PWRI

0855 Discussion

0910 The M8. 1 Guam Earthquake of 8 August 1993, Gerald BRADY, USGS

0930 Infrastructures Damage by 1993 Hokkaido-Nansei-Oki Earthquake, Takashi IIJIMA*,

Yutaka IIDA, Kazuhiko KAWASHIMA, Hisanori OTSUKA, Osamu MATSUO,

Shigenobu TANAKA, Akira NAKAMURA, and Masanori NAKANO, PWRI

0950 12 July 1993, Hokkaido Earthquake, Riley CHUNG, NIST

1010 Discussion

1030 Break

THEME - REPORT OF TASK COMMITTEE WORKSHOPS

Chairman: Dr. Richard WRIGHT

1045-1145 Technical Session - Report of* Workshops

1045 T/C "A" Strong Motion Data (Presented by S. Iai)

1055 T/C "G&J" 3rd Workshop on Earthquake Protective Systems of Bridges

(Presented by K. Kawashima)

1105 T/C "I" 3rd Workshop on Tsunamis (Presented by H. Meyers)

1115 T/C "J" 2nd Workshop on Seismic Retrofit of Bridges (Presented by S. Mahin)

1125 T/C "J" 10th Workshop on Bridge Engineering (Presented by J. Cooper)

1135 Discussion

1145 Adjourn for Lunch

1145 Lunch: Hosted by the National Science Foundation, William W. HAKALA, Director for Mechanical and Structural Systems Division

XXX
1230-1415 TASK COMMITTEE REPORTS AND RESOLUTIONS

Report of Task Committees

Chairman: Dr. Richard WRIGHT

T/C A Strong-Motion Data and Applications

T/C B Testing and Evaluation Procedures for Building Systems

T/C C Evaluation and Improvement of Structures T/C D Dam Earthquake Engineering

T/C F Disaster Prevention Methods for Lifeline Systems

T/C G Structural Control and Intelligent Materials Systems

T/C H Soil Behavior and Stability During Earthquakes

T/C I Storm Surge and Tsunami

T/C J Wind and Earthquake Engineering for Transportation Systems

T/C K Wind and Earthquake Engineering for Offshore and Coastal Facilities

1415-1430 Break

1430-1530 Adoption of Final Resolutions

1530-1540 Break

CLOSING CEREMONIES

1540 Call to Order by Noel J. RAUFASTE, Secretary-General, US-Side Panel

Closing Remarks by Takashi IIJIMA, Chairman Japan-Side Panel

Closing Remarks by Richard N. WRIGHT, Chairman US-Side Panel

1600 Conclusion of Joint Meeting

xxxi

United States-Side Panel on Wind and Seismic Effects

Membership List

May 1994

Dr. Richard N. Wright

Chairman

Director, Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301-975-5900 FAX 301-975-4032 wright@micf.nist.gov

Mr. Noel J. Raufaste

Secretary-General

Head, Cooperative Research Programs

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301-975-5905 FAX 301-975-4032 raufaste@micf.nist.gov

Dr. Kharaiti L. Abrol

Acting Director

Structural Engineering Service (088475)

Department of Veterans Affairs

810 Vermont Avenue, NW

Washington, DC 20420

202-233-7379 FAX 202-233-2115

Dr. John Ake

Geophysicist

Seismotectonics and Geophysics Section

P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-236-4195 FAX 303-236-6763

Dr. S. T. Algermissen

Deputy for Geologic Hazard & Risk Assessment

Office of International Geology

U.S. Geological Survey

917 National Center

12201 Sunrise Valley Drive

U.S. Department of the Interior

Reston, VA 22092

703-648-6051 FAX 703-648-4227

xxxiii
Mr. Charles E. Anderson

Seismic Safety Coordinator

P.O. Box 25007 (D-3130)

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-236-9114 FAX 303-236-6763

Dr. Celso S. Barrientos

Supervisory Physical Scientist

National Environmental Satelite Data

Information Service - Code E/RA28

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

5200 Auth Road

Camp Springs, MD 20746

301-763-8102 FAX 301-763-8108

Dr. Eddie N. Bernard

Director, Pacific Marine Environmental Laboratory

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

7600 Sand Point Way, NE

BIN C15700/Building 3

Seattle, WA 98115-0070

206-526-6800 FAX 206-526-6815

Dr. Roger D. Borcherdt

Branch of Seismology

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415-329-5619 FAX 415-329-5163

Dr. A. Gerald Brady

Research Civil Engineer

Branch of Earthquake and Geomagnetic Information

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415-329-5664 FAX 415-329-5163

Mr. H. Lee Butler

Chief, Research Division

Coastal Engineering Research Center

U.S. Army Engineer Waterways Experiment Station

Office CEWES-CR

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2405 FAX 601-634-4314

xxxiv
Dr. Ken P. Chong

Program Director, Mechanical & Structural Systems

National Science Foundation

4201 Wilson Boulevard - Room 545

Arlington, VA 22230

703-306-1361 FAX 703-306-0291

Dr. Riley Chung

Group Leader, Earthquake Engineering

Structures Division

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301-975-6056 FAX 301-869-6275

Mr. James D. Cooper

Deputy Chief, Structures Division, HNR-10

Federal Highway Administration

U.S. Department of Transportation

6300 Georgetown Pike

McLean, VA 22101

703-285-2060 FAX 703-285-2766

Dr. A. G. Franklin

Chief, Earthquake Engineering & Geophysics Division

CEWES-GH Geotechnical Laboratory

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2658 FAX 601-634-4134

Mr. G. Robert Fuller

Special Assistant

Office of Single Family Housing

Room 9184 HS

U.S. Department of Housing and Urban Development

Washington, DC 20410-8000

202-708-4549 FAX 202-708-5966

Mr. James H. Gates

Caltrans - Structures

P. O. Box 942874

Sacramento, CA 94274

916-227-8773 FAX 916-227-8174

XXXV
Mr. Peter E. Gurvin

Director, Building Design and Engineering Division

Foreign Building Operations

Building SA-6, Room 335

U.S. Department of State

Washington, DC 20520

703-875-6117 FAX 703-875-6204

Dr. Walter W. Hays

Deputy for Research Applications

Office of Earthquake, Volcanoes and Engineering

905 National Center

U.S. Geological Survey

U.S. Department of the Interior

Reston, VA 22092

703-648-6711 FAX 703-648-6717

Mr. James R. Hill

Manager, Natural Hazards Mitigation Programs

Office of Safety, Health and Quality Assurance

U.S. Department of Energy

Washington, DC 20545

301-903-4508 FAX 301-903-8585 Jim . Hi 1 1@HQ . DOE . gov

Mr. Larry C. Hultengren

Senior Structural Engineer

Office of Civil/Structural Engineering

Foreign Building Operations

U.S. Department of State

Washington, DC 20520

703-875-6194 FAX 703-875-6204

Dr. William B. Joyner

Geophysicist

Branch of Seismology

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415-329-5640 FAX 415-329-5163

Mr. Roger M. Kenneally

Structural Engineer

Structural and Seismic Engineering Branch

Mail Stop NL/S - 217A

U.S. Nuclear Regulatory Commission

Washington, DC 20555

301-492-3893 FAX 301-492-3696

xxxvi
Dr. H. S. Lew

Chief, Structures Division

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301-975-6061 FAX 301-869-6275

Dr. Shih-Chi Liu

Program Director, Structural Systems

Division of Biological and Critical Systems

National Science Foundation

4201 Wilson Boulevard - Room 545

Arlington, VA 222 3 0

703-306-1362 FAX 703-306-0291

Dr. Stephen Markwell

Chief, Geotechnical and Geology Division

Code D-3600 - P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-236-9329 FAX 303-236-6763

Dr. Francis G. McLean

Chief, Geotechnical Engineering and Embankment

Dams Branch

P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303-236-3854 FAX 303-236-6763

Dr. Raymond E. Meyer

Senior Program Manager

Office of Foreign Disaster Assistance

N/S Room 1262A

Department of State

320 21st Street, NW

Washington, DC 20523-0008

202-647-7530 FAX 202-647-4036

Mr. Herbert Meyers

Chief, Earth Geophysics

National Geophysical Data Center

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

325 Broadway

Boulder, CO 80303-3328

303-497-6521 FAX 303-497-6513

xxxvii
Mr. Ugo Morelli

Policy Manager for Seismology

Federal Emergency Management Agency

500 C Street, SW

Washington, DC 20472

202-646-2810 FAX 202-646-3104

Mr. Howard D. Nickerson

Earthquake Engineering and Weapons Specialist

Naval Facilities Engineering Command

Hoffman Building #2, Room 12S63, Code 04B2

200 Stovall Street

Alexandria, VA 22332

202-433-8599 FAX 202-433-8777

Dr. William E. Roper

Assistant Director

Research and Development (Civil Works)

U.S. Army Corps of Engineers

2 0 Massachusetts Avenue, N.W

Washington, DC 20314-1000

202-272-0257 FAX 202-272-0907

Dr. John B. Scalzi

Program Director, Structures and Building Systems

National Science Foundation

4201 Wilson Boulevard - Room 545

Arlinton, VA 22230

703-306-1361 FAX 703-306-0291

Dr. Charles E. Smith

Research Program Manager

Offshore Minerals Management

Technology Assessment and Research Branch

Minerals Management Service

U.S. Department of the Interior

381 Elden Street, MS 4800

Herndon, VA 22070-4817

703-787-1559 FAX 703-787-1010

Dr. T. T. Soong

Samuel P. Capen Professor

Department of Civil Engineering

National Center for Earthquake Engineering Research

State University of New York at Buffalo

212 Ketter Hall

Buffalo, NY 14260

716-645-2469 FAX 716-645-3733

xxxviii
Mr. Stan Strickland

Chief, Air Base Systems Branch

Stop 37, WL/FIVC/OL, Bldg. 1120

139 Barnes Drive, STE 2

Tyndall AFB , FL 32403-6001

904-283-3707 FAX 904-283-3722

xxxix
ALTERNATE MEMBERS

Dr. Clifford J. Astill

Program Director

Division of Biological and Critical Systems

National Science Foundation

4201 Wilson Boulevard - Room 545

Arlington, VA 22230

703- 306-1361 FAX 703-306-0291

Dr. Mehmet K. Celebi

Research Civil Engineer

Branch of Earthquake and Geomagnetic Information

U.S. Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415-329-5623 FAX 415-329-5163

Mr. Michael Changery

Chief, Global Analysis Branch

National Climatic Data Center

National Oceanic and Atmospheric Administration

U. S. Department of Commerce

Federal Building

Ashville, NC 28801

704- 271-4765 FAX 704-271-4246

Mr. C. Y. Chen

Senior Civil/Geotechnical Engineer

Office of Foreign Buildings

Department of State

Code SA-6, Rm. 327

Washington, DC 20520

703-875-6207 FAX 703-875-6204

Mr. Vincent P. Chiarito

Research Structural Engineer

Structural Mechanics Division

Structures Laboratory

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2714 FAX 601-634-3412

Dr. James F. Costello

Senior Structural Engineer

Structural and Seismic Engineering Branch

Structural Section

Office of Nuclear Regulatory Research

U.S. Nuclear Regulatory Commission

Mail Stop 007NL

Washington, DC 20555

301-492-3818 FAX 301-492-3696

XL
Mr. Lucian G. Guthrie

Structural Engineer

Office of Chief of Engineers

Headquarters U.S. Army Corps of Engineers

Corps of Engineers Civil Works Engineering Div.

U.S. Department of the Army

20 Massachusetts Avenue, N.W.

Washington, DC 20314-1000

202-272-8673 FAX 202-272-1485

Dr. Robert L. Hall

Chief, Structural Analysis Group

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2567 FAX 601-634-3412

Dr. James R. Houston

Chief, Coastal Engineering Research Center

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601-634-2000 FAX 601-634-2055

Mr. James Lander

Geophysicist

Cooperative Institute for Research in

Environmental Sciences

University of Colorado

Campus Box 449, Room 152 RL3

3100 Marine Street

Boulder, CO 80309

303-497-6446 FAX 303-497-6513

Mr. Robert R. Ledzian

Senior Staff Assistant for Research

Office of Liaison - Engineering Research, Code 3020

Bureau of Reclamation

U.S. Department of the Interior

1849 C Street, N.W.

Washington, DC 20240

202-208-3432 FAX 202-208-6252

Mr. Tingley K. Lew

Research Structural Engineer

Structures Division

Naval Facilities Engineering Service Center - ESC 62

560 Center Drive

Port Hueneme, CA 93043-4328

805-982-1234 FAX 805-982-1418

XLi
Mr. Michael Mahoney

Physical Scientist

Federal Emergency Management Agency

500 C Street, SW

Washington, DC 20472

202-646-2810 FAX 202-646-3104

Dr. J. Eleonora Sabadell

Director, Natural and Manmade Hazards Mitigation Program Directorate of Engineering

National Science Foundation

4201 Wilson Boulevard - Room 545-03

Arlington, VA 22230

703-306-1361 FAX 703-306-0291

Dr. Erdal Safak

Research Structural Engineer

U.S. Geological Survey

U.S. Department of the Interior

922 National Center

Reston, VA 22092

703-648-6534 FAX 703-648-6717

XLii
Japan-Side Panel on Wind and Seismic Effects

Membership List

May 1994

Dr. Takashi Iijima

Chairman, Japan-side Panel on Wind & Seismic Effects

Director-General

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2821 Fax. 298-64-2148

Dr. Kazuhiko Kawashima

Secretary, Japan-side Panel on Wind & Seismic Effects

Research Coordinator for Underground Development

Department of Planning and Administration

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2473 Fax. 298-64-1527

Dr. Hiroshi Sato

Secretary, Japan-side Panel on Wind & Seismic Effects

Head, Structures Division

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2874 Fax. 298-64-0565

Dr. Tokunosuke Fujitani

Chief, The Second Research Laboratory

Applied Meteorology Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1 Nagamine, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-53-8620 Fax. 298-55-7240

Mr. Ryoji Hagiwara

Head, Ground Vibration Division

Director, Earthquake Disaster

Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2926 Fax. 298-64-0598

xLiii
Dr. Hisahiro Hiraishi

Associate Director for Composite

Structures Research

Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi

,

Ibaraki-ken 305

Tel. 298-64-6639 Fax. 298-64-6773

Dr. Kiichi Horai

Head, Seismology and Volcanology

Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-53-8675 Fax. 298-51-3730

Dr. Susumu Iai

Chief, Geotechnical Earthquake

Engineering Laboratory

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka

Kanagawa-ken 239

Tel. 298-53-8675 Fax. 298-51-3730

Dr. Takamasa Inatomi

Director, Structural Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka

Kanagawa-ken, 239

Tel. 468-44-5027 Fax. 468-44-4095

Dr. Yoshikazu Kitagawa

Director, International Institute of

Seismology and Earthquake Engineering

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-6677 Fax. 298-64-6777

Dr. Yasuyuki Koga

Director, Earthquake Disaster

Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2829 Fax. 298-64-0598

xLiv
Mr. Osamu Matsuo

Head, Soil Dynamics Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2933 Fax. 298-64-2576

Mr. Kiyoshi Mimura

Director, Geographic Department

Geographical Survey Institute

Ministry of Construction

1, Kitazato, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2667 Fax. 298-64-1804

Mr. Chikahiro Minowa

Cooperative Research Officer

National Research Institute for

Earth Science and Disaster Prevention

3-1, Tennodai, Tsukuba-shi,

Ibaraki-ken, 305

Tel. 298-51-1611 Fax. 298-51-1622

Mr. Hatsukazu Mizuno

Special Researcher for International

Standards, Planning Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-6760 Fax. 298-64-2989

Mr. Akira Nakamura

Head, Fill Type Dam Division

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2413 Fax. 298-64-0164

Mr. Masanori Nakano

Head, Foundation Engineering Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2873 Fax. 298-64-0565

XLV
Dr. Shinsuke Nakata

Director, Production Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-6644 Fax. 298-64-6774

Mr. Kazuhiro Nishikawa

Head, Bridge Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2905 Fax. 298-64-0565

Dr. Setuo Noda

Deputy Director General

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka

Kanagawa-ken 239

Tel. 468-44-5000 Fax. 468-42-9265

Dr. Shigeyuki Noto

Director, Structures Division

Civil Engineering Research Institute

Hokkaido Development Agency

1-3, Hiragisi, Toyohira-ku

Sapporo, Hokkaido, 062

Tel. 11-841-1111 Fax. 11-824-1226

Dr. Nobuyuki Ogawa

Head, Earthquake Engineering Laboratory

National Research Institute for Earth

Science and Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-51-1611 Fax. 298-51-5658

Mr. Keiichi Ohtani

Director, Disaster Prevention Research Division

National Research Institute for Earth

Science and Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-51-1611 Fax. 298-51-1622

xLvi
Dr. Hisashi Okada

Head, Aerodynamics Division

Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-6641 Fax. 298-64-6773

Dr. Izuru Okawa

Head, Geotechnical Engineering Division

Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-6642 Fax. 298-64-6773

Mr. Shin Ootsuka

Head, Typhoon Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-53-8663 Fax. 298-55-2683

Dr. Hisanori Otsuka

Head, Earthguake Engineering Division

Earthguake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2932 Fax. 298-64-0598

Mr. Masashi Satou

Head, Structures Section

Civil Engineering Research Institute

Hokkaido Development Bureau

Hokkaido Development Agency

1-3, Hiragishi, Toyohira-ku

Sapporo, Hokkaido, 062

Tel. 11-841-1111 Fax. 11-824-1226

Dr. Tomotsuka Takayama

Director, Hydraulic Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka

Kanagawa-ken 239

Tel. 468-44-5009 Fax. 468-44-3888

xLvii
Mr. Shigenobu Tanaka

Head, Coastal Engineering Division

River Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2327 Fax. 298-64-1168

Dr. Tatsuo Uwabe

Chief, Earthquake Disaster Prevention Laboratory

Structural Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka

Kanagawa-ken 239

Tel. 468-44-5030 Fax. 468-44-0839

Dr. Hiroyuki Yamanouchi

Director, Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-6633 Fax. 298-64-6773

Dr. Koichi Yokoyama

Director, Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 298-64-2835 Fax. 298-64-0565

XLviii
List of Task Committee Members

May 1994

Task Committee

A. Strong-Motion Data and

Applications

US Side

A.G. Brady*

C.J. Astill

R.D. Borcherdt

M.K. Celebi

A.G. Franklin

W.B. Joyner

R.B. MacDonald

F. G. McLean

H. Meyers

S.T. Algermissen

C.S. Barrientos

G. R. Fuller

T . K . Lew

E. Safak

Japanese Side

S,

Y , R.

K

K

S

H

K

Noda*

Kitagawa

Hagiwara

Horai

Mimura

Iai

Mizuno

Ohtani

B. Testing and Evaluation

Procedures for Building

Systems

H.S. Lew*

V.P. Chiarito

J.E. Sabadell

C.E. Smith

K. Abrol

S. Sweeney

A. Bieniawski

K. Ohtani*

0. Matsuo

H. Mizuno

S. Noda

1 . Okawa

H. Sato

C. Evaluation and Improvement

of Structures

K.P. Chong*

P. A. Brady

C. Mutreja

J.O. Jirsa

H.S. Lew

T.K. Lew

H.D. Nickerson

D. H. Oh

M.A. Phipps

J.B. Scalzi

S. Woodson

S.A. Asar

M.K. Celebi

G.R. Fuller

K.L. Abrol

R.L. Hall

D. Ellsworth

H. Yamanouchi*

H. Hiraishi

K. Yokoyama

Y. Koga

M. Nakano

T. Fujitani

S. Nakata

S. Otsuka

I . Okawa

H. Sato

XLIX
D. Earthquake Engineering for

Dams

E. OPEN

F. Disaster Prevention Methods

for Lifeline Systems

G. Structural Control and

Intelligent Material

Systems

H. Soil Behavior and Stability

During Earthquakes

I. Storm Surge and Tsunami

w. E - Roper* A. Nakamura*

R. L. Hall R. Hagiwara

L. G. Guthrie K. Kawashima

R. Davidson H. Otsuka

M. E. Hynes

A. G. Franklin

R. Navidi

L. Von Thun

W. Allerton

F. G. McLean

J. R. Hill

R. M. Chung* Y. Koga*

C. J. Astill K. Kawashima

M. K. Celebi R. Hagiwara

J. D. Cooper N. Ogawa

G. Al-Chaar H. Otsuka

T . K. Lew T. Uwabe

J . B. Scalzi S. Noto

J. S . Spencer M. Sato

S. Wu

J. R. Hill

s. Sommer

s. C. Liu* Y. Kitagawa*

V

.

P. Chiarito K

.

Kawashima

K. P - Chong N. Ogawa

J. R. Hayes H. Yamanouchi

L. C. Hultengren H. Sato

T. T. Soong C. Minowa

H. Otsuka

A. G. Franklin* R. Hagiwara

C. J. Astill A. Nakamura

R. D. Borcherdt 0. Matsuo

C

.

Y. Chen H

.

Mizuno

F. G. McLean S. Iai

C. E. Smith Y. Koga

R. M. Chung S. Noto

T. Inatomi

M. Nakano

H. Meyers* S. Tanaka*

C. J. Astill S. Otsuka

C. S. Barrientos T. Takayama

E. N. Bernard S . Otsuka

L. Butler

J. Lander

W.E. Roper

L
J. Wind and Earthquake

Engineering for

Transportation Systems

K. Wind and Earthquake

Engineering for Offshore

and Coastal Facilities

*Chairman

J. D. Cooper* K. Yokoyama*

A. G. Franklin T. Fuj itani

J. H. Gates R. Hagiwara

J . B. Scalzi jrJ\ . Nishikawa

H. S. Lew K. Kazuhiko

P. Yen H 0 Otsuka

H. R. Bosch H. Sato

M. Nakano

Y. Koga

M. Sato

C. E

.

Smith* T. Uwabe*

C . S. Barrientos T. Fuj itani

w

.

E. Roper M. Nakano

J. E. Sabadell S. Noda

M. C. Miller T. Inatomi

M

.

K. Celebi S. Iai

Li

RESOLUTIONS

Resolutions of the Twenty-Sixth Joint Meeting

U.S.-Japan Panel on Wind and Seismic Effects (UJNR)

National Institute of Standards and Technology

Gaithersburg, Maryland

May 17-20, 1994

The following resolutions are hereby adopted:

1. The Twenty-Sixth Joint Panel Meeting provided the forum to exchange valuable technical information

which was beneficial to both countries. In view of the importance of cooperative programs on the subject

of wind and seismic effects, the continuation of Joint Panel Meetings is considered essential.

2. The following activities have been conducted since the Twenty-Fifth Joint Meeting:

a. Seven Japan researchers performed earthquake engineering assignments as guest researchers in the U.S.: C. Minowa and N. Ogawa from NIED worked at Wayne State University; H. Sugita

and M. Kanda from PWRI worked at the University of California (Berkeley) and Virginia

Polytechnic Institute; M. Iiba and H. Fujitani from BRI worked at the National Center for Earthquake Engineering Research (SUNY) and H. Shiohara also from BRI worked at Virginia

Polytechnic Institute. N. Yasuda from PWRI visited the Headquarters U. S. Army Corps of

Engineers for exchange of technical information.

Four U.S. researchers performed earthquake engineering assignments as guest researchers in Japan: A. Taylor, NIST, D. Norman, WES, and M. Yashinsky, Caltrans worked at PWRI; and

M.A. Riley from NCEER worked at BRI.

b. Technical documents, research reports, and proceedings of workshops have been exchanged.

These contributed to the development of research programs and enhanced on-going research in both countries.

c. Seven workshops and conferences were held:

(1) Workshop on Strong Motion Data, Task Committee (A), Menlo Park, CA, 10-17 December

1993.

(2) The 1st U.S. -Japan Joint Technical Coordinating Committee on Composite and Hybrid

Structures Program, Task Committee (B), Tsukuba, Japan, 8-9 November 1993.

(3) The Fourth U.S. -Japan Joint Technical Coordinating Committee on Precast Seismic Structural

Systems Program, Task Committee (B), Tsukuba, Japan, 16-19 May 1994.

(4) The Third Workshop on Earthquake Protective Systems of Bridges, Task Committees (G&J),

Berkeley, CA, 24-27 January 1994.

Liii
(5) The Third Workshop on Tsunamis, Task Committee (I), Osaka, Japan, 28 August 1993.

(6) The Tenth Bridge Workshop, Task Committee (J), Lake Tahoe, NV, 10-14 May 1994.

(7) The 2nd Workshop on Seismic Retrofit of Bridges, Task Committee (J), Berkeley, CA,

January 20-21, 1994.

3. The Panel, in celebration of the 25th Anniversary of its creation and in remembrance of the 70 year

anniversary of the Kanto Earthquake, jointly conducted with the International Decade for Natural Disaster

Reduction (IDNDR) Promotion Section of the Ministry of Construction, a one-day Symposium on the

(IDNDR), in Tokyo on May 21, 1993. The Symposium increased the awareness of Japanese disaster

relief and mitigation decision-makers about the disastrous effects from strong winds and earthquakes.

MOC's IDNDR Promotion Section and PWRI published the Proceedings of the Symposium and

distributed copies to the U.S. -side and to the Japan earthquake disaster prevention community.

4. The Panel will continue to seek methods to contribute to the IDNDR such as exchanging Proceedings

of Joint Panel Meetings and of Task Committee Workshops with their respective country's National

Committees for the IDNDR.

5. Japanese and U.S. members performed joint investigations of damage of the 12 July 1993 Hokkaidonansei-oki Earthquake and the 17 January 1994 Northridge California Earthquake. Experiences from

these disaster investigations were reported at the 26th Joint Panel Meeting. The Panel recognizes the

importance of using these findings in its on-going work, and of the need to continue performing future

joint post-disaster investigations to improve engineering and construction practices. Information from

these natural disasters will be transferred to Panel members and to others in the engineering and

construction communities through Panel reports and presentations. Special Panel Theme Sessions, such

as featured at the 26th Joint Panel Meeting, are encouraged for dissemination of timely information.

6. The Panel approved the Task Committee reports developed and presented during the Twenty-Sixth

Joint Panel Meeting. Each report includes objectives and scope of work, accomplishments, future plans,

and other information.

7. The Panel endorsed the following proposed Task Committee Workshops during the coming year.

a. Task Committee (B) plans its 2nd Joint Technical Coordinating Committee Meeting on Composite

and Hybrid Structural Systems, July 1994, Chicago, IL, USA.

b. Task Committees (C&G) plan their 2nd Workshop on Structural Control and Intelligent Material

Systems, May 1995, Tsukuba, Japan.

c. Task Committee (F) plans its 6th Joint Workshop on Disaster Prevention for Lifeline Systems

December 1994, Los Angeles, CA, USA.

d. Task Committee (H) plans its 4th Joint Workshop on Remedial Treatment of Potentially

Liquefiable Soils, July 4-8, 1994, Tsukuba, Japan.

e. Task Committee (J) plans its 11th Bridge Workshop, May 1995, Tsukuba, Japan.

f. Task Committees (K&I) plan a Second Workshop on Wind and Earthquake Engineering for Offshore and Coastal Facilities, January 1995, Berkeley, CA, USA.

Liv
Scheduling for the Workshops shall be performed by the U.S. and Japan Chairmen of the respective Task

Committees with concurrence of the Joint Panel Chairmen. Results of each activity conducted before the

27th Joint Meeting shall be presented at the 27th Joint Panel Meeting.

8. The Panel recognizes the importance of collaborative joint research programs. The Panel endorses

the initiation of joint research programs on Composite and Hybrid Structures and on Soil Liquefaction

and Countermeasures.

9. The Panel, in December 1993, published its first issue of Wind and Seismic Effects Newsletter. The

newsletter summarized the Panel's activities and accomplishments. The second newsletter (to include a report on the 26th Joint Panel Meeting) will be published during the summer of 1994.

10. The Panel recognizes the importance of continued exchange of personnel, technical information,

research results, and recorded data that lead to mitigating losses from strong winds and earthquakes. The

Panel also recognizes the importance of using available large-scale testing facilities in both countries.

Thus, these activities should be continued, strengthened, and expanded. The Panel will provide official endorsement to facilitate these exchanges.

11. The Twenty-Seventh Joint Panel Meeting of the UJNR Panel on Wind and Seismic Effects will be

held at PWRI, Tsukuba, Japan, May 1995. Specific dates, program, and itinerary will be proposed by

the Japan-Side with concurrence of the U.S. -Side Panel.

LV

WIND ENGINEERING

Wind Tunnel Experiment for a

Super Long-Span Cable-Stayed Bridge in Smooth Flow

Hiroshi Sato", Ryuichi Toriumi2

', Makoto Kitagawa3

' and Hiroshi Katsuchi4)

ABSTRACT

Aerodynamic stability of the

Tatara bridge, which is a super

long-span cable-stayed bridge

whose center span length is 890m, was investigated through wind

tunnel test using full aeroelastic

model in smooth flow. The

geometric scale of the model was

1/70. The stiffness, mass and

geometric shape of girder and tower

were simulated by stiffening bars,

additional mass and exterior wooden

forms. Each cable was designed and

adjusted so that they had proper

mass distribution and tensile force.

The drag force acting on each cable

was also simulated. From the wind

tunnel experiment, it was found

that the flutter speed of the

bridge was high enough. The wind

induced vibration observed was

vortex-induced vibration with small

amplitude only. The vortex-induced

vibration will be evaluated

considering the results of wind

tunnel test in turbulent flow. As

for the aerodynamic stability of

the bridge in turbulent flow and at

erection stage, wind tunnel tests are in progress. The topographical

effects are also planned to be

studied.

KEYWORDS

Aerodynamic stability, Super

long-span cable-stayed bridge, Full

aeroelastic model test, Wind tunnel

test

1. INTRODUCTION

The Tatara Bridge, which is

under construction in Japan, is a super long-span cable-stayed

bridge (Figure-1, Photo-1). The

center span length of the bridge

will be 890m, which is almost 300m

longer than that of the Yan-pu

Bridge in China, the world's longest

cable-stayed bridge at present.

For such a long span bridge, it is

very important to evaluate the

aerodynamic stability of the bridge

before its construction. Usually,

the aerodynamic stability of

bridges in Japan is evaluated by

spring mounted rigid model wind

tunnel tests. However, as the

Tatara bridge has such a long

span, it was thought that wind

tunnel test using a full aeroelastic

model should be conducted in the

Large Boundary Layer Wind Tunnel

1) Head, Structure Div., Structure

and Bridge Dept., PWRI

2) Senior Research Engineer, ditto

3) Manager, First Design Div., Design

Dept., Honshu-Shikoku Bridge

Authority

4) First Design Div., ditto

3
in PWRI. The Tatara bridge will be

surrounded by islands as shown in

Figure-2. Some of the islands have

high mountains up to 400m above the

sea level. Therefore, it was

considered that topographical

effect should be studied using

topographical model.

This paper reports the design

and characteristics of full

aeroelastic model of the Tatara

bridge, and the wind tunnel test

result in smooth flow. The test

results of the wind tunnel test in

turbulent flow, and the

topographical effects on

aerodynamic response of the bridge

will be reported in other

opportunity.

2. PURPOSE OF THE STUDY

The purpose of this study is

to make clear the aerodynamic

characteristics of a super

long-span cable-stayed bridge, the

Tatara bridge. The wind tunnel

test in smooth flow gives us fundamental data to evaluate the

aerodynamic stability of the Tatara

bridge.

3. AEROELASTIC FULL MODEL OF THE

TATARA BRIDGE

The model was fabricated with

a geometric scale of 1/70. Except

for the cables, the model was

designed by well known methods,

i.e. steel frame was designed to

achieve proper stiffness,

additional mass was installed to

simulate mass and polar moment of

inertia, and the frame and

additional mass were covered with

exterior wooden forms to simulate

geometrical shape.

Froude number similitude was adopted to the design of the model.

This means wind speed ratio was 8.4

Table-1 shows the required

and designed mass and stiffness of

girder, tower and cable.

Attention was paid to the

design of the cables. As well as the number of the cable, mass and

tensile force of each cable was

simulated. The drag force acting

on each cable was also similar to

the prototype bridge. Piano wires

were used to simulate axial

stiffness. In order to give proper

mass to each cable, cylindrical mass

pieces were placed at a certain

interval. The mass pieces were so designed that the cables might have

proper mass and drag force. For

the similarity of drag force, the

difference in Reynolds number were

taken into account. Similar method

was used for the design of cables in

the model of Akashi Kaikyo

Bridge 13. Tensile force in each

cable was measured from its natural

frequency, and adjusted by turning

turnbuckle installed at the end of

cable. Figure-3 shows the final

error of the tensile force in each

cable.

During the preliminary wind

tunnel test, it was found that the

stay cables of the model vibrated

in wind. At certain wind speed, the

vibrational amplitude became so large that stay cables might touch

each other. In the real bridge, the

wind-induced vibration of stay

cables will be eliminated by

installing dampers or by some other

4
methods. Therefore the vibration

of each cable in the model was

eliminated by installing thin extra

cables as shown in Figure-4.

Through preliminary wind tunnel

study, it was found that the thin

extra cables would not affect

aerodynamic characteristics of the

model so significantly.

Photo-2 shows the model of

the Tatara bridge in the Large

Boundary Layer Wind Tunnel. The

structural characteristis of the

model were as follows.

(1) STATIC CHARACTERISTICS

Static loading tests were

carried out in order to check the

static characteristic of the model.

Figure-5 show the results. Either

in holizontal, vertical and

torsional loading, the results

shows good agreement with

calculated results.

(2) DYNAMIC CHARACTERISTICS

Free vibration tests in

windless condition were carried out

to check the dynamic

characteristics of the model.

Table-2 shows the calculated and

measured natural frequency of the

model. Figure-6 shows the

vibrational mode shape. Because of

the good agreement between

calculated and measured results, it was confirmed that the model

represented the prototype bridge

well. Table-2 also shows the

logarithmic damping decrement

observed. The damping observed

was a little larger than the target

value of 0.02. The difference

between target damping and

observed one should be considered.

4. EXPERIMENTS IN SMOOTH FLOW

The followings are results of

the wind tunnel tests in smooth

flow with the angle of attack 0 deg.

(1) Vibrational Amplitude

Figure-7 shows the

relationship between vibrational

amplitude and wind speed. No

divergent vibration such as flutter was observed up to the reference

wind speed of flutter. When the

reduced wind speed was about 1.0,

small vertical bending

vortex-induced vibration was

observed. The vibrational mode was

1st symmetric one. Usually, when

1st symmetric vortex-induced

vibration is observed,

vortex-induced vibration of higher

modes are expected in higher wind

speeds. However, in this case, no

higher modes vibration was

observed.

(2) Change of Damping

Relationship between wind

speed and logarithmic decrement is

shown in Figure-8. The decrease of

damping at the reduced wind speed

of 1.0 is coincident with the

appearance of vortex-induced

vibration. At higher wind speed,

the damping increased with the wind

speed.

(3) Change of Frequency

Change of the frequency with

the wind speed is shown in Figure-9.

The frequency of torsion, vertical

and horizontal bending scarcely

5
changed.

5. DISCUSSION

Through the full model wind

tunnel test, it was verified that

the critical wind speed for flutter

of the Tatara bridge is high

enough.

Relationship between wind

speed and logarithmic decrement

from the flutter analysis of the

Tatara bridge is shown in Figure-10.

This result shows the reduced onset

wind speed of flutter is about 7.5.

In the section model tests

conducted in advance, no flutter

was observed up to the reduced

wind speed of 5. In the present full

model wind tunnel test, no flutter was observed up to the reduced

wind speed of 4. As far as the wind

speed range where full model test

was conducted is concerned, all the

three results show stable

characteristics of the bridge to

flutter. Although the Tatara bridge

is a super long-span cable-stayed

bridge, no unfamiliar flutter

characteristics was observed so

far.

The only wind induced

vibration observed was

vortex-induced vibration with small

amplitude. In the section model

tests, no vortex-induced vibration

was observed as shown in Figure-7.

The vortex-induced vibration will

be evaluated considering its

characteristics in turbulent flow.

6. CONCLUSION

Through the full aeroelastic

model wind tunnel test, the

aerodynamic stability of the Tatara

bridge was investigated. No

divergent vibration such as flutter

was observed. This means the

Tatara bridge is aerodynamically

stable. The only wind induced

vibration observed was vortex-induced vibration with small

amplitude. The vibration will be

evaluated considering its

characteristics in turbulent flow.

7. FUTURE STUDY

The wind tunnel test for

erection stage of the Tatara bridge

and wind tunnel test in turbulent

flow are in progress. Wind tunnel

study for the topographical effect

on the Tatara bridge is scheduled.

The results will be reported in

other opportunity.

REFERENCES

DYokoyama K., Kanazaki T. and

Yasuda M. , Boundary Layer Wind

Tunnel Study on Full Model

Aeroelastic Long Span Bridge

Model, Proc. of 23rd. Joint Panel

Meeting on Wind and Seismic Effects,

UJNR, 1991.

2)Sumiyoshi Y., Endo T., Miyata T.,

Sato H. and Kitagawa M.,

Experiments for the Akashi Kaikyo

Bridge in a Large Boundary Layer

Wind Tunnel, Proc. of International

Seminar on Utilization of Large

Boundary Layer Wind Tunnel,

Tsukuba, Japan, 1993.

6
Table- 1 Similarity of the Model

Unit Requi red Des igned

G 1 rder( Cen ter of Center Span)

Axial Stiffness (E*A)

Bending Stiffness

Vertical (E*Iv)

Horizontal (E*Ih)

Torsional Stiffness (G*J)

Mass

Polar Moment of Inertia

N

N • C ID 3 N • c m

3 N * cm

2 g / c ffl g

• c m

3 / c m

456. 5

17. 368

730. 1 68

14. 638

38. 4

5. 543. 5

475, 808

17. 375

729. 654

16. 352

40. 3 5. 421. 6 Tower( Average)

Stiffness

Mass( Average : each shaft) g/cm

See below

25. 8 See below

25. 8 Cable(Uppermos t Cable of Center Span)

Axial Stiffness (E*A)

Mass

N g 'cm 63. 28

0. 22815

64. 1 1 0. 22794

Deformation of Tower by specified load

In plane Out of Plane Torsional

Table-2 Natural Frequency and Damping

Vibrational Mode

Natural Frequency (Hz) Logar i thro ic Damping

Req u l red Measured Decremen t

Vertical

1 s t symme trie

1st antisymmetric

2nd symme trie

1. 883

2. 252

2. 757

2.014

2. 407

2. 959

0. 0429

0. 0408

( 0. 0238 )*

Torsional

1st symmetric 3. 987 4. 326 (0. 0180)*

Horizontal

1 s t symme trie 0. 684 0. 725 0.0130

*:Measured without extra cables

7
8 i
OnomichHmabafi

Kojima-Sakaide

Kobe-Naruto

Tatara Bridge

Onomichi I.C.

I.C. Onomichi Bridge (1968)

Mukaihigashi I.C.

/.TMukaishima I.C. Ohhama P.A.^Hnnoshima

^

Bridge (1983) Innoshima I.C. m A. / ^ Jlkuchi Bridge

/ / ^/

*l(ojcnijima I.C.

Hiroshima Pref.

Ehime Pref.

Kurushima Bridges ^tm*++

j

Ohmishima I.C. Kamiura P.A>|«

Ohmishima Bridge

'""

(1979) Hakatajima I.C.

' Hakata-Ohshima Bridge (1988)

Ohshima I.C. Seto Inland Sea

$ Imabari-kita I.C.

Ii Imabari-Minami I.C.

n Onomichi-imabari Route

i n hi mi i i Completed

A mmm Construction

Figure-2 Location of the Tatara Bridge
1A 2P 3P 4P

Figure-4 Extra Cable Arrangement

10
Vertical Loading

30

1 5

0

-

1 5

- 30

40

20

0 -20

-40

I . 0

0. 5

0

0. 5

1 . 0

4 0 4

J

4 0 4

(Uni t : mm)

Horizontal Loading

4 0 4 —^ T

2P 3

P

(Uni t : mm

)

Torsional Loading

0-400.4 -0.400.4

1

F

II

2 P 3 P (Unit

- : A n a 1 y t i c a • : Measured

Figure-5 Static Loading Test Results

1st symmetric mode( vertical)

: i

i .

|'a

' 2P ^nL^* 3P Vp

- : A n a 1 y t i c a 1 O : Ueasu red

1st symmetric mode( torsional)

3P 4P

- : A n a 1 y t i c a 1 O : Measured

1st anti-symmetric mode( vertical)

in-"

I A 2P 4P

: A n a 1 y t i c a 3 : Measured

1st svmmetric mode(horizontal)

3P 4P

- : A n a 1 y t i c a 1 O : Measured

2nd svmmetric mode( vertical)

A 2P 3P o 4P - : A n a 1 y t i c a 1 O : Heasu red

r i e u r c 6 Natural Vibrational Mode Shape

11
0 .010

-0.008

\

Full Model Test Result

O : 1st symmetric mode(vertical)

Section Model Test Result • : Vertical

006

Ua

Q.S

<

004

^0 002

=3

S6S

o

- Q,

o

1— 2.0

O -Cu

4.0 6.0 8.0

Reduced Wind Speed

1.0

0.8

0.6

* 0.4

0.2

Full Model Test Result

O : 1st symmetric m o d e ( t o r s i o n a 1

)

Section Model Test Result • : Torsional

2.0 4.0 6.0 8.0

Reduced Wind Speed

Figure-7 Reduced Wind Speed - Vibrational Amplitude

12
0.6

c

CD

Escd

s-. *o

CD

COc

•iHQ.

ca

cdO

-J

0. 55

0.5

0.45

0.4

0. 35

0.3

0. 25

0.2

0. 15

0. 1

0. 05

ST

IT

2 4 6 Reduced Wind Speed

10

m Vertical

Horizontal

A Torsional

(1st symmetric mode)

Figure-8 Reduced Wind Speed Logarithmic Damping Decremen

>, 3

CJc

CDcr

CD

To r s i o n a 1 (

* A A

A

1st symi

A

e t r i c motie )

Vc r t i c a 1 ( 1 s t s y m m e trie mod 2)

S ii m m

Y

Ho

>\.

O

r i z o n t a 1 ( 1st s y mit O

e t r i c mo O

de )

2 4 6 8 Reduced Wind Speed

10

Figure 9 Reduced Wind Speed Frequency

13
c

CD

EH

CD

(hU

CD Q

5£C

CO Q

0. 15

0. 1

0.05

a -o.o5 o

i—

3

0.

1

4 6 Reduced Wind Speed10

Figure-10 Reduced Wind Speed Damping (Analytical)

14
15

Wind Tunnel Tests on Tapered Cylinders for Highway Support Structures

by

Harold R. Bosch*

ABSTRACT

Highway support structures are an important

component of the nation's highway system.

Over the years, wind-induced vibration of these

structures has been observed throughout the

country under a variety of conditions. A

number of recent major failures, some resulting

in deaths, has sparked interest in the wind loads on and fatigue design of highway support

structures. Design is currently guided by the

AASHTO specifications. Although the design

codes received major revision in 1985, there is concern that they still do not adequately address

the effects of wind on these structures. To gain

better understanding of the problem and to develop recommendations for code revisions,

several projects have been initiated at the State and Federal levels. One of these studies is being

conducted by the Federal Highway

Administration's research staff located in McLean, Virginia. A portion of this project

involves wind tunnel testing to establish the

effects of wind on tapered cylinders. The test program and some of its early results will be the

topic of this paper.

KEYWORDS: cylinders; drag; highway signs;

luminaires; traffic signals; wind loads.

1. INTRODUCTION

Along the nation's highways and within its cities, there is a vast array of highway support

structures such as overhead or roadside signs,

high light towers, light poles, and signal

standards. These structures and the hardware

they support are essential for the safe and

efficient movement of vehicles, passengers, and

goods throughout the highway system. They

provide the means for illuminating the right-ofway, directing the traveller through signing, and

regulating traffic flow at intersections. Highway

appurtenances, as they are sometimes called,

consist mainly of circular or multi-sided

cylindrical components which are interconnected

through welded and/or bolted joints.

Frequently, the components are uniformly

tapered along their length to optimize the design,

reduce cost, and improve aesthetics. Most

structures rest on concrete footings or foundations and are attached using an anchor

bolt arrangement. For safety reasons, the bolt or base assembly is often designed to breakaway

easily under impact loading.

Over the years, wind-induced vibration of

highway support structures has been observed

throughout the country under a variety of

conditions. While random, small amplitude

motions are seldom of concern, some types of

vibration can and have lead to failures due to fatigue. These failures may be of the nuisance

type requiring maintenance and repairs or, in some cases, catastrophic resulting in collapse

and deaths. Structural vibrations can result from

slow steady winds, strong turbulent winds,

and/or gust loads from trucks. The nature and

significance of the vibrations depends upon wind

conditions, structural dynamic properties, and

stress levels within critical components. The

design of these structures to resist wind and

other loads is currently guided by the American

Association of State Highway and Transportation

Officials (AASHTO) Standard Specifications for Structural Supports for Highway Signs,

Luminaires, and Traffic Signals [1].

Although the design codes received a major

revision in 1985, there is concern that they still do not adequately address the effects of wind on

Research Structural Engineer, FHWA Turner- Fairbank Highway Research Center, 6300

Georgetown Pike, Mclean, Virginia 22101.

17
modern highway support structures. Since

1985, significant changes have occurred in design philosophies, material choices, and

manufacturing processes. Several recent

structural failures underscore the need for

improved design procedures and revised

specifications to ensure more fatigue-resistant structures. To gain a better understanding of

how wind affects these structures and to develop

recommendations for code revisions, several

projects have been initiated at both the State and

Federal levels.

The Federal Highway Administration's

(FHWA's) Aerodynamics Laboratory at the

Turner-Fairbank Highway Research Center

(TFHRC) has long been involved in the study of

wind effects on transportation structures. To

address the problem of wind-induced vibration

of highway support structures, the Aerodynamics

Laboratory staff have conducted a comprehensive literature search. Although

significant data exists regarding basic wind

forces on uniform (non-tapered) structural sections, little or no detailed information could

be found for the wide variety of tapered, crosssectional shapes employed in these structures. Several studies have been conducted to investigate local problems, specific structures, or

structural details; however, no comprehensive

investigations were identified.

Recently, the FHWA has initiated research to improve our understanding of this problem and

to establish detailed wind load data which is not

presently available. The first phase of this work

will involve wind tunnel testing of scaled section

models to establish wind forces on and response

behavior of structural components as well as attached hardware such as signal heads,

luminaires, and sign panels. For the second

phase, wind tunnel tests will be performed using

full aeroelastic models to characterize the

dynamic response of highway support structures under a variety of wind conditions. These

laboratory tests will be complemented by full scale wind and response measurements on

existing structures wherever possible. The third

and final phase will concentrate on developing new or improved methods for predicting the

response of these structures to wind.

Phase one of the research is well underway and

progress on wind tunnel testing of tapered

cylinders will reported in what follows. This

paper will describe the laboratory, model design,

and test methods. The results of force

measurements on a representative sample of

tapered and non-tapered cylinders will be

presented. These results will be discussed and

conclusions will be drawn.

2. AERODYNAMICS LABORATORY

The research is being conducted in the

Aerodynamics Laboratory at TFHRC.

Established in the late 1950's, the Laboratory

has been used primarily to ensure the

aerodynamic stability of long span bridges. The

facility includes a large lab area containing a wind tunnel, several unique test fixtures,

instrumentation systems, and lab test equipment

as well as an adjacent office area housing staff, computer systems, reference materials, and

records. The wind tunnel is a low velocity,

open-circuit, laminar flow design. Air enters a double inlet section, passes through a diffusing

section containing a series of 8 screens, and

exits a 6 by 6-ft nozzle into the test area (see

Figure 1). The air flow is quite laminar and

well controlled at speeds up to 50-ft/s. For

experiments where turbulent flow is desired, an

active turbulence generator is inserted into the

circuit between the wind tunnel nozzle and the

test area. With this system, large scale

turbulence can be simulated in the vertical and

along-wind directions, independently. The

vertical angle of the wind can be changed by

rotating the wind tunnel nozzle up and down or by rotating the model around its longitudinal

axis as needed.

Section models are used for most experiments.

These models are designed to relatively large

scale ratios (1:25 to 1:100) to ensure adequate

reproduction of geometric details. To measure wind forces, the section model in mounted in a force-balance assembly and placed in the test area. For dynamic response measurements, the

models are mounted on a spring suspension

18
system in a rigid frame surrounding the test area. Air flow studies are accomplished by

installing velocity probes on a 3-D, motorized

traverse system and surveying the flow field around the section model

.

3. MODEL DESIGN

A series of rigid, lightweight section models has

been designed for the static and dynamic wind

tunnel tests. These models comprise a range of

cross-sections and taper ratios which are intended to be representative of those commonly

found in highway support structures. Model

diameters were selected to provide a suitable

Reynolds number (Re) range for testing (see

Figure 2). So far, 53 models have been

fabricated with either circular, octagonal, or hexagonal cross-sections. All sections were

carefully machined from wood, aluminum, or

steel to ensure crisp details, proper surface

roughness, adequate rigidity, and low total weight. Initially, sharp corners have been

specified for the octagonal and hexagonal

sections. Later in the research program, models

with a range of corner radii will be fabricated

and tested. All models are 60 inches long with

rigid force-balance adaptor plates mounted on

each end. For each of the 3 shapes, there is a

set of 5 uniform (non-tapered) section models

with diameters of 2, 3, 4, 5, and 6 inches.

Tapered sections have 3 equivalent (central)

diameters of 3, 4, and 5 inches each with taper

ratios (in/ft) of 5, 10, 14, 20, 25, and 30

percent. In addition to the manufactured shapes,

two tapered models were fabricated using

sections of actual poles. The first is a circular

aluminum specimen with a central diameter of 5

inches and a taper ratio of 26.6 percent. The

other is an octagonal steel specimen with a

central diameter of 4.5 inches, taper ratio of

13.4 percent, and corner radius of 5.6 percent.

4. TEST PROCEDURE

To measure wind forces on the cylindrical

sections, each model was prepared and carefully

mounted in the force-balance. This forcebalance system supports the section model in a horizontal orientation and measures the 3 force

components of lift, drag, and pitching moment

(see Figure 3). Although not used during this testing, it is capable of rotating the model about

its longitudinal axis to vary the angle of attack. Since the force-balance is actually two separate

balance mechanisms, it is capable of measuring

unbalanced loads on models. This feature has

proved to be extremely useful for this study.

All models were thoroughly inspected for surface defects and weighed prior to testing.

For tapered models, a tuning weight was

installed inside the small end to establish a perfectly balanced section prior to testing. Each

section model was tested at 31 different wind

speeds and all tests were performed twice to ensure repeatability of the results. Wind speeds were carefully chosen to provide the maximum

range of Reynolds number possible and to avoid

regions where undesirable vibrations would

develop in the model. In general, speeds ranged

from 5 to 30 ft/s and the same speed settings were not necessarily selected for each model.

Multi-sided sections were tested in two primary

orientations - corner and face (see Figure 4).

At a later time, tests will be conducted over a range of wind angles. During each test, readings from the balance, pitot tubes,

barometer, and temperature sensors were

continuously sampled at a rate of 100 Hz for a period of at least 90 seconds. Each of the 6

force components were low-pass filtered at 30

Hz prior to recording.

5. TEST RESULTS

At this point in the test program, force

measurements have been completed in laminar

flow on all of the 53 models available. This

represents more than 82 separate setups and 164

total tests. Mean and root-mean-square (RMS)

values have been computed for all 6 of the

model end forces at each wind speed tested. These "end" forces have been combined into 3 "model" forces (lift, drag, and moment) for

computing MEAN and RMS force coefficients.

Only the results obtained for mean axial (drag)

forces will be presented here.

5.1 Circular Cylinders

19
For the uniform (non-tapered) circular cylinders

tested, the drag forces are very consistent.

When model end forces are compared, it is clear

that the wind load on the cylinders is perfectly

balanced as expected (see Figure 5). The

measured drag coefficients range from 1.21 to

1 .23. These measurements correspond well with

values found in published data [3].

The influence of taper was evaluated using

"equivalent" cylinder diameters of 3, 4, and 5

inches. For the tapered circular cylinders tested,

the total drag on the model generally decreases

with increasing taper ratio. It is interesting to note that most of the reduction in force results from the first 5 percent of tapering (see Figures

6-7). The difference (unbalance) between model

end forces becomes larger with increasing taper

ratio as expected. The magnitude of the

unbalance seems to decrease, however, as the

equivalent cylinder diameter increases (see

Figures 8-9). For the 3-inch cylinders, the

measured drag coefficients range from 1.22 to 1.17 over the 30 percent range of tapers. The 4

and 5-inch cylinders produce drag coefficients

which range from 1.23 to 1.16 and 1.22 to 1.12,

respectively. Generally, results indicate that tapering of circular cylinders has a positive

influence on drag by reducing forces as much as 8 percent in some cases.

To determine the adequacy of the fabricated

section models, tests were also performed on a section removed from an actual luminaire pole.

The section selected was aluminum with an

equivalent diameter of 5.065 inches and a taper

ratio of 26.6 percent. The member end forces

and drag coefficients obtained for the prototype

specimen correspond very well with those

obtained for Model CC5.25 (see Figures 10-11).

This indicates that model details faithfully

reproduce the prototype conditions.

5.2 Octagonal Cylinders

For the uniform (non-tapered) octagonal

cylinders, the drag forces were also quite

consistent. The end drag forces for each model were equal, indicating balanced load conditions.

Model drag forces were higher for the face

orientation than the corner orientation (see

Figure 12). Measured drag coefficients range

from 1.60 to 1.67 and 1.30 to 1.35 for the face

and corner orientations, respectively. These

values compare very well with those obtained in other laminar flow wind tunnel and tow tank

tests on octagonal cylinders [4-5].

As before, the influence of taper was evaluated

using 3 equivalent diameters of 3, 4, and 5

inches. The diameter referenced here is that of

the inscribed circle. For the tapered octagonal

cylinders tested, the total drag force also tends

to decrease with increasing taper ratio. The

amount of force reduction, however, does not appear to be as great as for the circular cylinders

(see Figure 13). The unbalance between model

end forces tends to increase with increasing

taper ratio and decrease with increasing

equivalent diameter.

For 3-inch octagonal cylinders tested in the corner orientation, the measured drag

coefficients range from 1.43 to 1.34 over the 30

percent range of tapers tested. The 4 and 5-inch

cylinders produce drag coefficients which range

from 1.41 to 1.29 and 1.36 to 1.25,

respectively.

As found in the uniform cylinder tests, significantly higher drag forces and coefficients were measured for tapered octagonal cylinders in the face orientation. For the 3 and 4-inch

cylinders, measured drag coefficients range from

1.72 to 1.58 and 1.65 to 1.55, respectively.

Problems were encountered when processing

data from the 5-inch cylinders in face orientation

and the results of those tests will not be reported

here.

To assess the adequacy of the fabricated

octagonal models, a comparison was also made

using a prototype specimen. The section

selected was galvanized steel with an equivalent

diameter of 4.468 inches and a taper ratio of

13.4 percent. The model end forces and drag

coefficients obtained for the prototype specimen

correspond reasonably well with those obtained

for Model OC4.14 (see Figures 14-15). The

small differences found are likely due to the

20
larger diameter and corner radius which are not

replicated in this particular model.

5.3 Hexagonal Cylinders

For the uniform (non-tapered) hexagonal

cylinders, only 3 of the 5 section models have

been tested. Tests on cylinders with diameters

of 2, 3, and 4 inches have been completed for

both face and corner orientations. As for the

other uniform cylinders, the end drag forces were found to be equal, indicating balanced

loading on the section. In contrast with the

octagonal cylinders, drag forces on the

hexagonal sections were larger for models in corner orientation than in face orientation (see

Figure 16). Measured drag coefficients range

from 1.99 to 1.89 and 1.66 to 1.51 for the corner and face orientations, respectively. Since no tapered sections have been tested yet, results for these sections cannot be presented here.

6. CONCLUSIONS

Although the research reported herein is continuing and is only part of a larger

investigation, some interim conclusions may be

drawn. For all of the circular cylinders tested,

including the tapered ones, drag coefficients

ranging from 1.12 to 1.23 have been obtained,

with an average value of about 1.17. While

these values are in excellent agreement with

published data, they are 2 to 12 percent higher

than the 1 . 1 recommended in the AASHTO code

specifications. Generally, tapering is beneficial

and reduces the drag force on these cylinders.

Design calculations based upon code values

would seem to be more applicable to tapered

sections than uniform ones.

For octagonal cylinders, the AASHTO

specifications recommend a drag coefficient of

1.2 for design. Tests on cylinders with a corner

orientation, produced drag coefficients ranging

from 1.25 to 1.43 with an average of about

1.34. A comparable range of 1.55 to 1.72 and

average of 1.62 were obtained for face

orientation. Again, these values compare very

well with other research findings; however, they

are 4 to 44 percent higher than code values.

Test results indicate that tapering may not be as beneficial for the octagonal cylinders.

For the 3 non-tapered hexagonal cylinders

tested, average drag coefficients of 1.60 and

1.94 have been obtained for the face and corner

orientations, respectively. It is interesting to note here that the corner orientation results in the larger forces on the section. These limited

results appear to be reasonable. The AASHTO

specifications do not specifically address

hexagonal members.

The reader is reminded that the tests reported on

here were conducted using laminar approach

flows and were confined to the subcritical flow

regime with Reynolds numbers ranging from

about 3 x 103 to 1 x 105

. Additionally, all multi-sided models were designed and fabricated

with sharp corners. The results presented are applicable only to these conditions.

As mentioned earlier, this is one of several

studies underway to improve our understanding

of how wind affects highway support structures. In the months and years ahead, these on-going

investigations should provide the detailed

information necessary for recommending code

revisions.

7. ACKNOWLEDGEMENTS

A portion of the work reported here was

accomplished by Ms. Anna M. Chesnul under a Research Fellowship sponsored by the National

Highway Institute. The author would like to express his appreciation to Ms. Chesnul as well

as her academic advisor, Dr. Ahmad H. Namini,

for their significant contributions.

8. REFERENCES

[1] AASHTO, Standard Specifications for Structural Supports for Highway Signs.

Luminaires, and Traffic Signals , 1985.

[2] ASCE Task Committee on Wind Forces,

"Wind Forces on Structures,"

Transactions . ASCE, Volume 126,

1961.

21
[3] Hoerner, Sighard F., Fluid-Dynamic

Drag, published by the author, 1965.

[4] James, W.D., Effects of Reynolds

Number and Corner Radius on TwoDimensional Flow Around Octagonal.

Dodecagonal, and Hexdecagonal

Cylinders . PhD Dissertation, University

of Iowa, 1976.

[5] Mehta, Kishor C, et al, Wind Drag

Coefficients for Octagonal Cylinders , report to Texas Department of Highways

and Public Transportation, Texas Tech

University, 1990.

22
23
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24
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30
Standards to Resist Hurricane Wind Forces

by

G. Robert Fuller 1 and R. D. Marshall2

ABSTRACT

On August 24, 1992, Hurricane Andrew struck

the coast of south Florida, and then proceeded

across the Florida peninsula and the Gulf of

Mexico before hitting Louisiana on August 26.

Damages of over $30 billion, deaths exceeding

55, and over 200,000 people left homeless

prompted the Department of Housing and Urban

Development (HUD) to examine its wind design

standards for manufactured housing and other

residential construction. To review the adequacy

of HUD's wind standards, the Department

contracted for a study by the National Institute of Standards and Technology (NIST). Based on

this study, a wind design and construction

standard for manufactured housing, referencing

the American Society of Civil Engineers (ASCE)

Standard 7-88 was developed and published on

January 14, 1994. Research is now being

conducted by NIST to provide information so

that HUD can further develop standards for wind and tornado resistant construction in other

areas of the U.S.

KEYWORDS: anchorage, building codes,

damage assessment, design standards, hazard

mitigation, hurricane, manufactured housing,

residential construction, structural engineering,

tornado, wind, wind speed.

1. INTRODUCTION

The purpose of the National Manufactured

Housing Construction and Safety Standards Act

of 1974 is to reduce the number of personal

injuries and deaths, and the insurance costs and

property damage resulting from manufactured

home failures, and to improve the quality and

durability of manufactured homes. In 1993,

manufactured housing accounted for

approximately 25 percent of all new singlefamily homes sold in the U.S. Design and

construction of this type of housing is covered

by the Manufactured Home Construction and

Safety Standards (MHCSS), administered by

HUD. The MHCSS, which became effective in 1976, prescribes minimum loads for design

against wind and snow.

Based on the performance of manufactured

housing in Hurricane Hugo in 1989 and in Hurricane Andrew in 1992, it became apparent

that the current provisions of the MHCSS were

inadequate to ensure satisfactory performance

under extreme wind conditions, and the

Department undertook a comprehensive review

of the requirements in 1992. This paper

describes the current wind load requirements,

the general types of failures experienced in south

Florida during Hurricane Andrew, and

amendments to the wind load provisions which

are to become effective in July of 1994.

Additional problems with ensuring the

satisfactory performance of this type of

construction in extreme winds are described in

this paper.

Following Hurricane Andrew, HUD conducted

field investigations of the damage experienced in the hurricane by manufactured housing, as well

as other residences constructed under HUD

subsidized or mortgage insurance programs.

The primary goal of these investigations was to ensure that Federal standards provide adequate

protection to manufactured home occupants

during high wind conditions. In addition,

damage surveys and assessments of single- and

1 U.S. Department of Housing and Urban

Development, Washington, DC 20410-8000

2 National Institute of Standards and

Technology, Gaithersburg, MD 20899

31
multi-family housing construction were made to evaluate the adequacy of building codes and

HUD's Minimum Property Standards.

2. WIND SPEEDS IN HURRICANE

ANDREW

The number of reliable wind speed records in a

landfalling hurricane usually is quite limited, and

Hurricane Andrew was no exception. In fact, no anemometers exposed to winds in the eyewall

of Andrew survived long enough or had the

recording capability to register the maximum

wind speeds. In the months following Andrew,

considerable effort was expended in locating

records of wind speed and/or barometric

pressure from which to reconstruct the wind

field over the area of heaviest damage. The

highest recorded gust speed, corrected for

anemometer error, was 79 m/s (177 mph) at Perrine. There were no confirmed sightings of

tornadoes in Dade County and subsequent aerial surveys did not show any evidence of tornado

damage.

Reinhold et al. (1993) used a computer-based

model that makes use of information on the

storm track, barometric pressure data, and the

radius of maximum winds. The transition from

over-water to over-land exposure and reductions

in speed as the storm moves inland are accounted for by an empirical decay model.

The model was calibrated against partial records

obtained inside the eyewall and complete records

obtained outside the eyewall. Results of the

analysis are shown in Figure 1 as fastest-mile

isotachs. Estimates of fastest-mile wind speeds

for standard conditions (10 m (32.8 ft) height in

flat, open country) in the area of heaviest

damage range from 55 m/s (122 mph) at Florida

City to 65 m/s (145 mph) north of Perrine

where the northern sector of the eyewall crossed

the coastline. The upper-bound estimate of the

fastest-mile wind speed in the Tamiami

Airport/Country Walk area is 60 m/s (135 mph).

A map of the affected area is shown in Figure 2

(Marshall 1993). The maximum recorded gust

speeds at selected locations are given in the

original units of knots.

3. PERFORMANCE OF MANUFACTURED

HOMES IN HURRICANE ANDREW

Damage to manufactured homes in Hurricane

Andrew ranged from minor loss of roofing

materials to total destruction. In general, those

units manufactured after the adoption of the

MHCSS in 1976 suffered less damage than did

those units manufactured prior to 1976.

However, conventional residential construction

located adjacent to manufactured home parks

performed better, in some instances significantly

better, than did manufactured homes, including

HUD-labeled units. Based on damage surveys

and a reconstruction of wind speeds over the

affected area, it appears that HUD-labeled units

began to experience damage to roof and wall

coverings at fastest-mile speeds of up to 42 m/s

(95 mph) and significant structural damage at wind speeds of from 45 to 54 m/s (100 to 120

mph). At wind speeds ranging from 54 to 60

m/s (120 to 135 mph), there were numerous

instances of HUD-labeled units suffering total destruction.

Commonly observed failures included loss of

roof membranes and failure of roof sheathing,

failure of uplift straps at truss-to-wall

connections where staple crowns pulled through

the strap material, loss of cladding on endwalls

and near corners where large negative (suction)

pressures develop, loss of add-on construction

such as expanded rooms or porches with

resulting missile damage and damage to the

parent unit at points of attachment, complete

separation of superstructure from floor and

underframe, and loss of the complete unit due to

failure of tiedown straps or withdrawal of soil anchors. In Florida, some form of anchorage

system had been installed in almost every case. The anchor failures that were observed involved

helical anchors or rock anchors installed in the

local coral which is covered by 150 to 300 mm

(6 to 12 inches) of sand.

4. CURRENT WIND LOAD

REQUIREMENTS OF THE MHCSS

The MHCSS wind load requirements in effect at the time of Hurricane Andrew identified two

32
wind zones for the United States. Zone II (hurricane) included the Gulf and Atlantic Coast

region, the coastal regions of Alaska, and Puerto

Rico. All other regions of the United States were designated as Zone I (non-hurricane).

Approximately, the boundary between the two

zones followed the 40 m/s (90 mph) wind-speed

contour of an early version of American

National Standard A58.1 (Now ASCE 7 - Minimum Design Loads for Buildings and Other

Structures). The basic wind load requirements

of the MHCSS were as follows:

Horizontal Load Uplift

Zone I 718Pa(15psf) 431Pa(9psf)

Zone II 1,197 Pa (25 psf) 718 Pa (15 psf)

For the design of components/elements such as eaves and cornices, the MHCSS required uplift

loads equal to 2.5 times the uplift loads listed above. For wind exposures in coastal and other

areas where wind records indicated significantly

higher wind speeds than are implied by the loads

listed above, the Department could establish more stringent requirements for homes known to be destined for such areas.

There were no requirement in the MHCSS for permanent foundations, but it was recommended

that windstorm protection in the form of soil anchors and tiedown hardware be designed to

resist horizontal and uplift loads equal to 1.5

times the values listed above. However, specific

requirements for windstorm protection devices

and the enforcement of these requirements are

left to the individual States. Recommendations

for the design and installation of such devices

are addressed in American National Standard

A225.1 - Manufactured Home Installations.

5. REVIEW OF BUILDING CODES AND

STANDARDS

Following Hurricane Andrew, the wind load

provisions of selected codes and standards were

compared. For Dade County, Florida, the basic

wind speed (fastest-mile speed at 10 m (32.8 ft) in exposure category C) specified by ASCE 7-88

is 49 m/s (110 mph), and the corresponding

design speed is 52 m/s (116 mph). In general,

the wind load requirements of ASCE 7-88

exceed those of the other codes and standards

included in this comparison.

Based on the requirements for structural stability

(sliding and uplift), the design loads required by

the MHCSS at the time of Hurricane Andrew

corresponded to a basic wind speed of from 36

to 38 m/s (80 to 85 mph). A similar analysis of

the South Florida Building Code (SFBC-88)

provisions indicates that the specified design

speed of 54 m/s (120 mph) was, in effect, a gust

speed. The drag and uplift loads required by the

SFBC-88 corresponded to basic wind speeds of

41 and 44 m/s (91 and 98 mph), respectively.

Although ASCE 7-88 and the Standard Building

Code (SBC-91) reference the same basic wind

speed of 49 m/s (110 mph) and the same source of pressure coefficient data, SBC-91 requires

average drag and uplift loads that are approximately 75 percent of the values required

by ASCE 7-88. The major reasons for this difference in design loads are a reduction factor

of 0.8 applied to the pressure coefficients used

in SBC-91 and disregard for the fact that the

wind speed probability distributions for

hurricanes and for extra-tropical storms are

different.

6. PROBABILITIES OF FAILURE

Design wind speeds should reflect the local wind

climate (distribution of extremes) and the

consequences of structural failure. For ordinary

buildings and structures it is generally accepted

that the design wind loads should have as their

basis the wind speeds associated with a mean

recurrence interval (MRI) of about 50 years. Although the probability that these speeds will

be exceeded in a 50-yr period is relatively high

(64 percent), the use of load factors or allowable

stresses in the design process reduces the risk of

a structural failure to about 5 percent over the same interval (Gupta and Moss 1993).

Generally, the available information from which

to estimate the probability of failure of a building or other structure is very limited.

33
Nevertheless, it is useful to examine the relative

risk associated with various design requirements

and extreme wind environments. In the

following, probabilities of failure implicit in the

MHCSS wind load requirements in effect at the

time of Hurricane Andrew are compared with

those of ASCE 7-88 (Minimum Design Loads

for Buildings and Other Structures) for Dade

County, Florida; Omaha, Nebraska; and Tucson,

Arizona. Extreme wind speeds for Dade County

are due to hurricanes while the extremes for the

other two locations are governed by extratropical wind events.

Implicit in the MHCSS wind load requirements

in effect at the time of Hurricane Andrew are

basic wind speeds (50-yr MR!) of 29 and 36 m/s

(65 and 80 mph) for Zone I (non-hurricane) and

Zone II (hurricane), respectively. According to ASCE 7-88, the basic wind speeds for Dade

County, Omaha and Tucson are 49, 37 and 34

m/s (110, 83 and 75 mph), respectively. The

ultimate load capacity or strength of the

structure can be estimated from the design

equation

<*»R ^ YL (1)

in which R is the nominal resistance of the

material or component under consideration, <j> is the resistance factor, L is the nominal (codespecified) load or load combination, and y is the

load factor. If typical values of 0.8 for the

resistance factor and 1.3 for the wind load factor are selected, the reference wind speed at or near

the ultimate strength of the structure (ultimate

limit state) will be (1. 3/0. 8)

1 ' 2 = 1.275 times the

basic wind speed. In the case of Dade County,

Florida, for example, the wind speed at failure

should correspond to (49)(1.05)(1.275) = 66

m/s (147 mph). Note that the factor of 1.05 is a structure importance factor required by ASCE

7-88 for design in hurricane-prone regions. If it

is further assumed that the coefficient of

variation (COV) of structural strength is 0.1,

then the distribution functions for load and

ultimate strength can be plotted against a reference dynamic pressure, q, as is shown in Figure 3 for Dade County, Florida. By

integrating the product of the load function,

QL(q), and resistance function, pR(q), it is possible to estimate the probabilities of failure

with the results as shown in Table 1. The load

function, QL(q), for Dade County is based on the

wind speed distributions developed by Georgiou

et al. (1983), and for Omaha and Tucson on the

distributions developed by Simiu et al. (1979).

Although the absolute values of the probabilities

listed in Table 1 are reliable only to the extent

that the simplifying assumptions represent the

true relationship between load and resistance, it

is their relative values (MHCSS/ASCE 7) that are of primary interest. It can be seen from

Table 1 that, at the time of Hurricane Andrew,

manufactured homes sited in Dade County had

a risk of failure during a 10-yr exposure that was of the order of 10 times the risk of ordinary

buildings designed in accordance with the wind

load provisions of ASCE 7-88. This same ratio

is seen to apply for Omaha, even though the

corresponding probabilities are only about half

those for Dade County. For Tucson, the

probability of failure for manufactured homes

during a 10-yr exposure is about 5 times that of

ordinary buildings and structures designed in accordance with ASCE 7-88 and about 1/5 that of manufactured homes sited in Dade County.

Not shown in Table 1 are the effects of

variability of resistance on the calculated

probabilities. In general, increasing the

coefficient of variation of resistance will increase

the probability of failure, particularly for lower mean values of resistance as is the case for the

MHCSS wind load criteria. For example, the

probability of failure during a 10-yr exposure at Tucson ranges from 0.27 to 0.42 for a corresponding range in COV of 0.05 to 0.20 in the case of the MHCSS design criteria.

7. REVISED STANDARDS

Based on the experience gained from Hurricanes

Hugo and Andrew, on a comprehensive review

of the wind load provisions of the model codes

and of ASCE 7-88, and on the results of

economic impact studies, an amended rule was

published in the Federal Register in January,

1994. This amended rule is to become effective

34
on July 13, 1994. The new wind zones are shown in Figure 4 with Zones U and III corresponding to basic wind speeds of 45 and 49

m/s (100 and 110 mph), respectively. Zone I retains the current MHCSS wind load

requirements for non-hurricane regions. The new design wind load requirements for Zones U

and III are summarized in Table 2.

8. FUTURE WORK

The MHCSS rule change is considered to be an

interim measure as it is limited to the design and

construction of manufactured housing units

destined for hurricane-prone areas for which the

basic wind speed is 45 m/s (100 mph) or higher.

The rule change does not address construction in non-hurricane areas for which the current requirements still apply. As is indicated by the

failure probabilities in Table 1, there are areas

in the non-hurricane region for which the

probabilities of failure remain unacceptably high.

Specifically, improvements in the provisions for

the design of cladding are required for wall corner zones, roof edge zones, and roof

overhangs. In addition, it is doubtful that

traditional approaches to windstorm protection

(shallow soil anchors, tiedown straps and drystacked concrete block piers) will prove to be

adequate or cost-effective in hurricane-prone

regions. Alternative approaches using

permanent foundation systems are being

investigated. Finally, the significance of

tornadoes as a design consideration is being

examined.

9. REFERENCES

ASCE (1990). Minimum Design Loads for

Buildings and Other Structures, ASCE 7-88,

American Society of Civil Engineers, New

York, NY, 94 pp.

Georgiou, P.N., Davenport, A.G. and Vickery,

B.J. (1983). "Design Wind Speeds for Regions

Dominated by Tropical Cyclones." Journal of

Wind Engineering and Industrial Aerodynamics,

Vol. 13, 139-152.

Gupta, A.K. and Moss, PJ. (Eds.) (1993).

Design of Low-Rise Buildings Subjected to Lateral Forces. CRC Press, Boca Raton, FL,

286 pp. HUD (1994). Final Rule - Manufactured Home

Construction and Safety Standards on Wind

Standards, Federal Register, Vol. 59, No. 10,

January 14, Office of the Assistant Secretary for

Housing, Dept. of Housing and Urban

Development, Washington, DC, pp 2456-2474.

Marshall, R.D. (1993). "Wind Load Provisions

of the Manufactured Home Construction and

Safety Standards - A Review and

Recommendations for Improvement." NISTIR

5189, National Institute of Standards and

Technology, Gaithersburg, MD, 90 pp.

MHCSS (1992). Federal Manufactured Home

Construction and Safety Standards, 24 CFR,

Part 3280, U.S. Dept. of Housing and Urban

Development, Washington, DC, pp 196-223.

NCSBCS (1987). "NCSBCS Standard for

Manufactured Home Installations (Manufactured

Home Sites, Communities, and Set-ups) - ANSI

A225. 1-1992." National Conference of States on Building Codes and Standards, Herndon, VA,

49 pp.

Reinhold, T.A., Vickery, P.J. and Powell,

M.D. (1993). "Wind Speeds in Hurricane

Andrew: Myths and Reality." Proceedings, 7th

U.S. National Conference on Wind Engineering,

UCLA, Los Angeles, CA, Vol. 2, pp 563-577.

Simiu, E., Changery, M.J. and Filliben, J.J. (1979). "Extreme Wind Speeds at 139 Stations

in the Contiguous United States. " NBS Building

Science Series 142, National Bureau of

Standards, Washington, DC, 147 pp.

35
Table 1. Probability of Attaining Ultimate Limit State

<t> = 0.8 7 = 1.3 COV (R) = 0.1

Location Design Probability

Criteria

Exposure Period 1 vr 10 yrs 25 yrs 50 yrs

Dade County, Florida MHCSS 0.037 0.313 0.609 0.847

Ujo = 49 m/s (110 mph) ASCE 7 0.003 0.033 0.081 0.155

Omaha, Nebraska MHCSS 0.018 0.165 0.362 0.594

Us, = 37 m/s (83 mph) ASCE 7 0.002 0.016 0.039 0.077

Tucson, Arizona MHCSS 0.007 0.067 0.160 0.295

U-so = 34 m/s (75 mph) ASCE 7 0.001 0.014 0.034 0.066

Table 2. MHCSS Revised Design Wind Pressures, January 14, 1994

ELEMENT ZONE II ZONE III (psf) (psf)

Anchorage for Lateral and Vertical Stability:

Net horizontal drag ±39 ±47

Uplift -27 -32

Main Wind Force Resisting System:

Shearwal Is, diaphragms and their fastening and anchorage systems ±39 ±47

Ridge beams and other main roof support beams -30 —36

Components and Cladding:

Roof trusses in all areas; trusses shall be doubled within 3'-0" -39 -47

from each end of roof

Exterior roof coverings, sheathing & fastenings; all areas except following —39 —47

Within 3"-0" of gable end or endwall if no overhang is provided —73 —89

Within 3'-0" of ridge or sidewall if no eave is provided -51 -62

Eaves (Overhangs at sidewalls) -51 -62

Gables (Overhangs at endwalls) -73 -89

Wall studs in sidewalls and endwalls, exterior windows & sliding glass doors,

exterior coverings, sheathing & fastenings:

Within 3'-0" from each corner of sidewall and endwall ±48 ±58

All other areas ±38 ±46

Note: 1 ft = 0.3048 m 1 psf = 47.88 Pa

36
81 .OW 80.5W

Figure 1. Isotachs of estimated maximum fastest-mile wind speeds (mph) in Hurricane Andrew.

(Reinhold et al. 1993). Note: 1 mph = 0.447 m/s.

1000

LOCATION Pk. Gust

(knots)

1 Perrine* 154

2 Fowey Rocks C-MAN 147

3 Tamiami Airport* 108

4 Miami WSFO/NHC 142

5 Miami International 81

6 Virginia Key 98

7 Miami Beach DARDC 92

8 HauloverNOS 85

9 Turkey Point (10 m)* 85

10 Turkey Point (60 m)' 83

1 1 Fort Lauderdale 76

* Anemometer failed

1 kt = 0.514 m/s = 1.151 mph

Figure 2. Area map showing approximate storm track and key locations, Hurricane Andrew.

37
DYNAMIC PRESSURE (psf)

0 10 20 30 40 50 60 70 80

Dade County, Florida

0 0.5 1.0 1.5 2.0 2.5 3.0 3.5 DYNAMIC PRESSURE (kPa)

Figure 3. Probability functions for load and resistance, Dade County, Florida.

(Based on MHCSS wind load provisions in effect at time of Hurricane Andrew)

Figure 4. Basic wind zone map for manufactured housing, based on ASCE 7-88.

(Revised Manufactured Home Construction and Safety Standard, January 14, 1994)

38
Inauguration of International Wind Engineering Forum

and

New U.S.-Japan Cooperative Research Program

by

Hisashi Okada(1) and Hiroyuki Yamanouchi12 '

Abstract

Interest in wind engineering are recently

growing. Many researchers and engineers have

become to seek much information and to wish

information exchange on wind engineering in the wide of the world. But it is rather difficult because of language barrier and etc. Responding to the request, the International

Wind Engineering Forum (IWEF) was

established in Japan and the US, where

information exchange can be carried out without strict limitations but somewhat

systematically. The inaugural meeting of the

forum was held in Tokyo on March 17, 1994.

Tentative activities of the Forum in Japan side

will be establishment of data banks of research

information on wind engineering, despatching

the information in English and Japanese,

publishing news letters and etc.

Key Words:

wind engineering, information exchange,

cooperative activities between the US and

Japan

1 . INTRODUCTION

Japan is well known as one of the most active

seismic countries. Attentions of the most

people gather fairly less to wind engineering

than to earthquake engineering. But a little change in this situation is occurring these days.

Topics on wind engineering are more often

heard than before. Domestic and International

symposiums in wind engineering are called more frequently. The number of researchers is increasing. Figure 1 shows the number of

papers in World Conference of Wind

Engineering [1]. The Number of papers has

been increasing steadily. It is remarkable thing

that the papers of Japanese has occupied about

25% of the total since 6th conference.

The reasons why the change mentioned above

is occurring are considered to come from the

constructions of high-rise buildings of which

design wind loads are greater than the design

seismic loads and etc. Figure 2 shows the

comparison of seismic and wind design loads

in Japan.

Under these situations, wishes for the

international exchange of information have

been raised up. Remarkable increase of

Japanese papers in international conference is one of the evidences. Many researchers want to exchange information more actively but some

of them cannot find the ways. There are also

language barrier and etc. There is the background which we could not miss on the way to esablishrnent of the IWEF.

New research cooperation in construction

technology is required under the New USJapan Economic Partnership Framework.

Disaster prevention/ Building safety is one of

the important items in the research cooperation.

Wind engineering researches on mitigation of

disaster coused by strong winds, control of

response of tall buildings due to strong winds

and etc. are expected to have much weight in the item.

Responding to the wishes under the background, we made the International Wind

Engineering Forum (= IWEF) where

international exchange of information can be

carried out without strict limitation but somewhat systematically.

Authors have participated in the action for the establishment as originators and made a great

role on the way. Having meetings with

professors of TTU and CSU and program

directors of NSF were some of the actions to

( 1 ) Head, Aerodynamics Division, Structural Engineering Department, Building

Research Institute, Ministry of

Construction

(2) Director of Structural Engineering

Department, ditto

39
make up the Forum.

The inaugural meeting of the IWEF was held at Toranomon Pastoral Hotel , Tokyo Japan, on March 17, 1994. The number of participants

including person from Korea were more than

100. Four person including Prof. Bienkiewicz, one of the originator of the IWEF, CSU, were

invited from the US and one person from

Taiwan as the speakers. They and Japanese

originators presented proposal of the IWEF,

present states of wind engineering in Taiwan,

US and Japan, cooperative research activities on wind engineering which had produced

fruitful results, and etc. in the meeting. The

program is shown in Figure 2. Enthusiastic

discussion on the gist, management and goal of

the IWEF followed the presentations.

In this paper, the gist, tentative activities,

management and etc. of the IWEF which were

presented in the meeting are described [2].

2. PROPOSAL OF THE IWEF

Inaugural Meeting of the IWEF was opened

with opening address and proposal of the

IWEF by Dr. Fujii, president of Wind

Engineering Research Institute Co., Ltd. In the

proposal, he explained motive and gist for the

establishment, intended tentative activities management and etc. of the IWEF. The details

will be introduced in the following sections.

3. MOTIVE FOR THE ESTABLISHMENT

Things relevant to wind such as wind resistant design, wind response evaluation and wind

environmental evaluation is recently raising the

practical importance. Many related domestic

and international symposiums has been held. The international information exchange has

been extensively carried out. State of the

international information exchange and

collaboration has been improved. But the

present state is not sufficient. One of the reasons is a language barrier between Japan

and English language countries. Some

exchanges and collaborations have been

carried out on personal relationship. Many

things cannot be expected from such

exchanges and collaborations. The idea

regarding the forum, where international

information exchange can be carried Dut somewhat systematically without severe

restriction, has come from the background.

4. ACTIVITIES OF THE IWEF'

Tentative activittes which are carried out in Japan side are as follows:

(1) Collection of publications, information of

research facilities on wind engineering in Japan, establishment of these data base arid despatching the information in English to foreign countries.

(2) Collection of information relative to wind

engineering in overseas and preparation of the

data base.

(3) Collection of information on strong wind

disaster, establishment of the data base and

giving support to surveys in damage sites. (4) Planning and holding lecture meetings and

seminars regarding to wind resistance design

of building construction, offer of the useful

information and promotion of the spread of

wind engineering information.

(5) Planning and promotion of advanced

research projects relative to wind engineering

(6) Support to the human exchange both in Japan and between foreign couniri.es and Japan.

(7) Publishing news letters (2 times a year) on which activities of the IWEF, information on

the data base established in the IWEF and etc. are informed.

5. MANAGEMENT OF THE! IWEF

The organization of the IWEF will be

tentatively be placed in Japan and the US. on

equal terms. Fig. 1 shows the organization.

The headquarters are positioned in the fallowings.

Japan side Kenchiku Kenkyu Shinko

Kyokai (Japan Association for Building

Research Promotion)

US. side : Colorado State University

The organization is expected to expand to the other countries. The organization in the both

countries intends to organize respective

branches. The branches are expected to support the IWEF activities from various

40
aspects. The constituent members are composed of the following originators.

Co. - Chairman : Japan side

:

T. Ohkuma

(Kanagawa University)

US. side : B. Bienkiewicz

(Colorado State University)

Advisory Board : Japan side : H. Okada

(Building Research Institute,

Ministry of Construction)

K. Fujii

(Wind Engineering Research

Institute Co., Ltd.) Y. Tamura

(Tokyo Polytechnic Institute) H. Yamanouchi

(Building Research Institute,

Ministry of Construction)

US. side : Four People are supposed to be the

Advisory Board members.

One of the most important problem is how to secure funds for the management of the IWEF.

In the meantime, the management will be

carried out by the donations and compensation

for the information given by the IWEF. The

volunteer activities by the people who

understand the gist of the IWEF are expected

and welcomed. It is required for the IWEF to make much effort on getting understanding of

many people. The efforts surely produce

cooperation of many people with founds and/or

supporting activities for the IWEF.

Applications for the public financial supports

are also very important to secure the founds.

6. FUTURE PROSPECTIVE

After the proposal of the IWEF, states of

wind engineering research in Taiwan, the US.

and Japan were introduced. The issues of wind

engineering we face are presented.

Cooperative research activities between the US

and Japan and the Universities in the US

which have been carried out and produced

fruitful results were also introduced. These

presentation are very references on

consideration of what and how to do in the

IWEF.

Following the presentations., enthusiastic

discussions were exchanged on the activities,

perspective, management, goal and etc. of the IWEF among participants in the meeting.

In the concluding remarks of the meeting,

Prof. Ohkuma, chairman of the Japan side,

presented history of wind engineering in Japan,

contribution which Japan should render in the

wind engineering area and etc. And he said he

wished that many countries join into the

circles for information exchange and

cooperative research activities which the I WET

was going to promote and many people

support the activities of IWEF.

Actual activities will be decided with

reference to questionnaires which were offered

from participants after the meeting. The first action will be publication of the first news

letter.

LATEST NEWS

(1) BRI will invite Prof. Bienkiewicz for the

promotion of the IWEF activities in this summer.

(2) List of research papers in domestic

conferences in 1993 and 1994 will be made in English and delivered people who want to receive this fall. (3) Seminar on the wind resistance design on

roof will be held in this summer.

Reference

[1] T. Ohkuma," Concluding/ Philosophical

Remarks - Japanese wind engineering is expected to contribute to the world", Inaugural

Meeting of the International Wind Engineering

Forum, 1994

[2] K. Fujii , "Proposal of the International

Wind Engineering Forum", Inaugural Meeting

of the International Wind Engineering Forum,

1994

41
300

250

£ 200

D. t 150

o

£ 100

50

0

H Japan

Total

—Lljsmss: 1st 2nd

SN8«

3rd 4th 5th 6th

Conferrence

7th 8th

Fig. 1 The Number of Papers presented at the

International Conference of Wind Engineering

OCD—

o

? 8 aj ooO

—

O Seismic L. (RC)

O Seismic L. (Steel

)

Wind L. (RC) • Wind L. (Steel

)

o--o-#-$—

o

9 o-c^o o

0 100 200 300 400

Height of Building ( m )

(a) 30 tn x 30 m square section

0 100 200 300 400

Height of Building ( m )

(b) 60 m x 60 m square section

Fig.2 Comparison of seismic and wind design loads in Japan

42
INAUGURAL MEETING

OF

THE INTERNATIONALWIND ENGINEERING FORUM

Toranomon Pastoral Hotel, Tokyo, Japan

March 17, 1994

Presiding: Prof. J.Kanda (University of Tokyo)

9:00- 9 30 Opening Address / Proposal of the International Wind Engineer ing Forum

Dr. Fujii (Wind Engineering Institute Co. Ltd.)

8:30- 10:15 Wind Engineering Research in USA

Prof. A. Kareem ( University of Notre Dame)

10:15-10:30 Coffee Break

Presiding: Prof. J.R. McDonald (Texas Tech University)

1 0:30- 11:15 Wind Engineering Activities in Taiwan

Prof. CM. Cheng (Tamkang University)

11:15-1 2:00 W ind Engineering Activities in Japan

Prof. Y. Tamura(Tokyo Institute of Polytechnics)

12:00-13:30 Lunch

Presiding: Prof H.Kawai ( Tokyo Denki University)

13:30-14:15 US-Japan Activities in Wind Engineering

Prof. B.Bienkiewicz (Colorado State University)

14:15-15:00 US-Japan Governmental Organizations Cooperative Activities

Dr. H. Okada (Building Research Institute)

15:0-15:15 Coffee Break

Presiding: Prof.A.Kareem(University of Notre Dame)

15:1 5 -1 6:00 Wind Engineering Research: Status and Issues

Dr. J. E. Sabadell (National Science Foundation)

16:00-16:45 CSU/TTU Cooperation Program in Wind Engineering

Prof. J. R. McDonald (Texas Tech University)

16:45-17:00 Concluding/Philosophical Remarks

Prof. T. Ohkuma (Kanagawa University)

18:00-20:30 Party

Fig. 3 Program of the Inaugural Meeting of the I WEF

43
Organization of IWEF

Headquaters

US side

Colorado

State University

Japan side

Japan Association lor

Building Research Promotion

(K.K.S.K.)

Co-chairman Prof.B.Bienkiewicz Prof.T.Ohkuuna

Advisory Board IWEF has commenced with four members

in the US and Japan respectively.

B nineties

US side

Japan side

Other Contrv

Fig. 4 View of Organization of the IWEF

44
EARTHQUAKE

ENGINEERING

A Seismic Retrofitting Manual for Highway Bridges

by

Ian G. Buckle 1 , Ian M. Friedland 1 , and James D. Cooper 2

ABSTRACT

In 1983, the Federal Highway Administration

(FHWA) published a set of guidelines for seismic retrofitting of highway bridges. These

guidelines presented what was then considered to be the state-of-the-art for screening, evaluating,

and retrofitting of seismically deficient bridges. In the 10 years since publication of the FHWA

guidelines, there has been significant progress in understanding the response of bridges and in the

development of new and improved retrofit technologies. As a consequence, the 1983

guidelines have recently been updated and

reissued as a retrofit manual. This paper

describes this revision, which was performed by

the National Center for Earthquake Engineering

Research under contract to the FHWA, and

discusses a number of significant changes made

in the new manual.

KEYWORDS: bridges; earthquake; seismic

design; seismic retrofit.

1. BACKGROUND

Of all the components of a typical highway

system, the highway bridge has been the most

closely studied for seismic vulnerability. As a consequence, standards have been developed and

adopted nationwide for the seismic design of

new bridges. By comparison, the seismic

retrofitting of existing bridges is a relatively new

endeavor. Only a few retrofitting schemes have

been used in practice and, given the present state of knowledge, retrofitting is still somewhat of an

art requiring considerable engineering judgement.

The San Fernando, California, earthquake of

1971 caused a significant amount of bridge

damage and, as a result, initiated a large research

effort into seismic bridge design and behavior in

the mid- to late 1970s. The majority of this work was jointly sponsored by the California

Department of Transportation (CALTRANS), the

FHWA, and the National Science Foundation.

The primary goal of this effort was to minimize

the risk of unacceptable damage during a design

earthquake. One result of this effort was the

publication by the American Association of State Highway and Transportation Officials

(AASHTO) of the first comprehensive national

highway bridge seismic design guide, the

AASHTO "Guide Specification for Seismic

Design of Highway Bridges," in 1983.

The San Fernando earthquake also provided the

impetus for the FHWA and CALTRANS to address the seismic retrofitting of existing

bridges to withstand earthquake forces and

movements. The culmination of this effort was

the publication of the 1983 FHWA report

"Seismic Retrofitting Guidelines for Highway

Bridges" (Report No. FHWA/RD-83/007) which was based primarily on research results and

information available in the late 1970s. At the

time the guidelines were issued, the existing

technology for highway bridge retrofitting was

limited and many of the proposed techniques had

not been field demonstrated. In the 10 or so years since, new and improved technologies for

retrofitting bridge columns and footings have

been developed and implemented, together with

methods to stabilize soils to prevent liquefaction,

and to ensure adequate connections between the

'National Center for Earthquake Engineering

Research, State University of New York at Buffalo, Buffalo, NY 14261

2 Federal Highway Administration, TurnerFairbank Highway Research Center, 6300

Georgetown Pike, McLean, VA 22101

47
bridge superstructure and substructure. Many of

these advances in the state-of-the-art are the

result of an aggressive research program which was begun by CALTRANS following the Loma

Prieta earthquake which occurred near San

Francisco in October 1989.

2. THE SEISMIC RETROFITTING MANUAL

FOR HIGHWAY BRIDGES

In order to capture these advances in seismic

retrofitting and to make the current state-of-theart available to bridge owners and engineers, the

FHWA initiated a project to update the 1983

guidelines. This effort has resulted in a new

document titled the "Seismic Retrofitting Manual

for Highway Bridges" (hereafter referred to as the Retrofitting Manual) which was completed

near the end of 1993 by the National Center for Earthquake Engineering Research (NCEER),

under contract to the FHWA.

The new Retrofitting Manual offers procedures

for evaluating and upgrading the seismic

resistance of existing highway bridges.

Specifically it contains:

A preliminary screening process to identify and prioritize bridges that need

to be evaluated for seismic retrofitting;

A methodology for quantitatively evaluating the seismic capacity of an existing

bridge and determining the overall

effectiveness of alternative seismic

retrofitting measures; and

Retrofit measures and design requirements for increasing the seismic resist- ance of existing bridges.

The Retrofitting Manual does not prescribe

requirements dictating when and how bridges are to be retrofitted. The decision to retrofit a bridge depends on a number of factors, several

of which are outside the scope of the Manual.

These include, but are not limited to, the

availability of funding, as well as political,

social, and economic considerations. The

primary focus of the Retrofitting Manual is directed towards the engineering factors.

3. MAJOR CHANGES

The Retrofitting Manual is based in large part on

the 1983 guidelines. In some cases, existing

material was updated as appropriate. In other

cases, completely new material has been added.

One notable editorial change is the format of the

new document: the 1983 guidelines used a specification-language format followed by a commentary; the new Retrofitting Manual

presents guidelines, recommendations, and

commentary in one combined section. This shift in style, from a specification to a technical

report, led to the change in title from

"Guidelines" to "Manual," and it is believed that

this improves overall readability. In addition,

the Manual format reflects the fact that the state- of-practice in bridge seismic retrofitting is changing rapidly at this time, and that it is premature to prepare specifications at a time

when new insight and experience is an almost

daily occurrence.

Major technical changes in the Retrofitting

Manual include the following:

An increased emphasis is placed on

bridge importance, and the minimum

requirements for retrofitting "essential"

bridges have been increased, even in low

seismic zones. This has been achieved

by redefining the Seismic Performance

Categories (SPC) so that, for example,

an essential bridge that was previously in SPC A is now classified as SPC B with more rigorous retrofitting requirements

as a consequence.

A new preliminary screening algorithm

has been introduced which separates the

quantitative evaluation of structure vulnerability and seismic hazard from

the subjective assessment of importance

and other societal issues. A two-part

48
procedure is thus proposed which first involves the ordering of all bridges

based on a numerical rating of structure vulnerability and seismic hazard. A

prioritized list is then obtained by

reordering this first list to include such

judgmental issues as importance, nonseismic deficiencies, remaining useful

life, network redundancy, and economic

and political factors. A new Priority

Index is then defined which is a mix of

engineering and societal factors.

In addition to the capacity/demand (C/D)

ratio method for detailed bridge

evaluation, a second method based on

equivalent lateral strength is now

included. This method avoids some of

the conservatism in the C/D method

which can lead to unnecessarily

expensive retrofit schemes. The new

equivalent lateral strength method does

require more effort to understand and

apply; however, it's use is expected to result in lower retrofit costs.

Expanded sections on retrofit measures

are included, reflecting recent progress

in the development of practical field techniques. These sections cover column strengthening using jackets,

footing upgrades using overlays, and

seismic isolation using elastomeric

bearings. An expanded section on cable

restrainers is also included. It is interesting to note that at the time the

1983 guidelines were issued, most of the

retrofitting techniques described therein were identified as potential retrofit measures since few, if any, had actually

been implemented in the field at that time. In the intervening ten years, this situation has changed and most of the measures described in the new Manual

are based on field experience.

Some of these technical changes are discussed in more detail below.

4. RETROFITTING PHILOSOPHY

The underlying philosophy in the new

Retrofitting Manual is that, whenever practical,

deficient components be strengthened to new

design standards. At first sight this may appear

to be inconsistent with the overall goals of

retrofitting, and not economically justifiable if the structure as a whole will perform below the

standards for new construction. There are two

reasons that the Retrofitting Manual makes this recommendation. One is that the cost to strengthen a component to new design standards

is usually not that much greater than the cost of

partial strengthening. The second reason is that

it is possible that retrofitting will be a phased

operation that takes place over the life of the

structure. Changes in construction technologies

and economic situations may make it feasible to strengthen some components in the future even though it is not economical to do so now. If component retrofitting were performed to standards below those for new construction, it could become necessary to restrengthen these

components during a second phase of retrofitting,

resulting in a higher total cost.

There may be cases, however, where it is not

feasible to strengthen components to new

standards. In these cases, the Retrofitting

Manual recommends a preference to at least strengthen such components to lower standards

rather than to reject retrofitting altogether. The

Retrofitting Manual provides guidance on the

selection of acceptable levels of strengthening,

but notes that this still requires the judgement of

the engineer, taking into consideration the

performance of the remainder of the structure.

The Retrofitting Manual also discusses a number

of secondary factors that must be considered

when retrofitting. One of these is the

repairability of the structure following an

earthquake. If possible, component

strengthening should not be done at the risk of

forcing damage to other components that are more difficult to inspect and repair. For

example, it is undesirable to strengthen a ductile

component if load would then be transferred to

49
a nonductile or brittle component. This may be

the case even if calculations indicated an overall

increase in seismic capacity.

Maintenance and inspection of retrofitted components are also discussed in the Retrofitting

Manual, as they should also be considered during

the retrofit design stage. Many years may pass before a structure is subjected to an earthquake.

The retrofit measure must be designed so that it can be maintained to function as planned when

and if an earthquake does occur.

5. BRIDGE CLASSIFICATION

Before seismic retrofitting can be undertaken for a group of bridges, they must first be classified

according to their Seismic Performance Category

(SPC). The SPC is determined by a combination

of seismic hazard and structure importance.

Seismic hazard is reflected in the Acceleration

Coefficient, A, which is assigned to all locations

covered by Division I-A of the AASHTO

Standard Specifications for Highway Bridges.

When multiplied by the acceleration due to gravity, g, the product, A-g, represents the likely

peak horizontal ground acceleration that will occur due to an earthquake sometime within a 475 year period. More rigorously, this acceleration has a 10 percent probability of being

exceeded within a 50 year time frame.

Bridge importance is not so readily quantified.

Two Importance Classifications (I) are specified

in the Retrofitting Manual: essential and

standard. "Essential" bridges are those which

must continue to function after an earthquake or which cross routes that must continue to operate

immediately following an earthquake. All other

bridges are classified as "standard." The

determination of the Importance Classification of

a bridge is necessarily subjective and

consideration should be given to societal/survival

and security/defense requirements.

The societal/survival evaluation addresses a number of socio-economic needs and includes,

for example, the need for access for emergency

relief and recovery operations immediately

following an earthquake.

Security/defense requirements may be evaluated

using the 1973 Federal-Aid Highway Act, which

required that a plan for defense highways be

developed by each state. The defense highway

network provides connecting routes to military

installations, industries, and resources not covered by the Federal-Aid primary routes.

An "essential" bridge is therefore one that

satisfies one or more of the following conditions:

a bridge that is required to provide

secondary life safety; e.g., a bridge that provides access to local emergency

services such as hospitals. This category

also includes those bridges that cross routes which provide secondary life safety, and bridges that carry lifelines such as electric power and water supply

pipelines;

a bridge whose loss would create a major economic impact; e.g., a bridge

that serves as a major link in a transportation system;

a bridge that is formally defined by a

local emergency plan as critical; e.g., a bridge that enables civil defense, fire departments, and public health agencies

to respond immediately to disaster

situations. This category also includes

those bridges that cross routes which are defined as critical in a local emergency

response plan and those that are located on identified evacuation routes; or

a bridge that serves as a critical link in the security/defense roadway network.

Based on the above considerations for seismic

hazard and importance, four Seismic

Performance Categories are defined in the

Retrofitting Manual as shown in Table 1

.

50
These SPC's are assigned differently from those

in the AASHTO specifications for new design,

where no allowance for structure importance is made in seismic zones with acceleration

coefficients less than 0.29. In view of the high

cost of retrofitting, it is important to be able to distinguish between "essential" and "standard"

structures and especially so in low to moderate

seismic zones. Such a distinction also enables a more rational allowance to be made for the

nature of the seismic hazard in the central and

eastern U.S. where the maximum credible

earthquake is expected to be significantly larger

than the "design" earthquake (475 year-event).

This implies that if an essential bridge in the east

is to remain fully operational following a large

earthquake, it will need to be retrofitted to a standard higher than that required by the current

specification for new construction. This

observation is reflected in the assignment of

SPC's for essential bridges in Table 1.

6. THE RETROFITTING PROCESS

Seismic retrofitting is one solution for minimizing the hazard of existing bridges that are vulnerable to serious damage during an

earthquake. Because not all bridges in the

highway system can be retrofitted simultaneously, the most critical bridges should be

retrofitted first. The selection of bridges for

retrofitting requires an appreciation for the

economic, social, administrative, and practical

aspects of the problem, as well as the engineering aspects. Seismic retrofitting is only one of

several possible courses of action; others include

bridge closure, bridge replacement, or acceptance

of the risk of seismic damage. Bridge closure or replacement is usually not justified by seismic

deficiency alone and will generally only be

considered when other deficiencies exist. Therefore, for all practical purposes, a choice

must be made between retrofitting or accepting

the seismic risk. This choice will depend on the

importance of the bridge and on the cost and

effectiveness of retrofitting.

The process of retrofitting bridges involves an assessment of a multitude of variables and

requires the use of considerable judgement. It is therefore helpful to divide the process into three

major stages. These are:

preliminary screening;

detailed evaluation; and

design of retrofit measures.

Each of these stages is outlined below and

described in further detail in the Retrofitting

Manual. Figure 1 is a flow chart which

illustrates the retrofitting process for each SPC.

7. PRELIMINARY SCREENING

Preliminary screening of an inventory of bridges

is recommended to identify those bridges which

are seismically deficient and those in the greatest

need of retrofitting. This is particularly useful

when a comprehensive retrofitting program is to be implemented.

The Retrofitting Manual describes a method for developing a Seismic Rating System which may

be used to prioritize bridges on a highway

system according to their need for seismic hazard

reduction. Factors considered in the seismic

rating process include structural vulnerabilities,

seismic and geotechnical hazards, and bridge

importance or criticality. In this way, the most

hazardous bridges are identified. Bridges high on the list should be investigated further to determine the benefits of retrofitting. Because

the decision to retrofit depends on political,

social, and economic factors as well as engineering issues, high priority bridges may not

necessarily be retrofitted. On the other hand,

bridges with a lower priority may need to be

retrofitted immediately.

One very important consideration that is not adequately reflected in the Seismic Rating

System is the relationship of the bridge to other

bridges on the system that may also be damaged

during an earthquake. These types of

considerations should be made prior to making a

51
detailed evaluation of the seismic capacity of the

bridge.

A further consideration when deciding if retrofit- ting is warranted is the age and condition of the

bridge. It would not be rational to spend a large

amount to retrofit a bridge with only five years

of service life remaining. An unusually high

seismic vulnerability may, however, be a justification to accelerate closure or replacement of

such a bridge.

A bridge in poor physical condition that is scheduled for nonseismic rehabilitation should be

given a higher priority for seismic retrofitting,

since construction savings can be realized by

performing both the nonseismic and seismic

work simultaneously.

The above examples do not represent all possible

cases, but they do illustrate some of the principles involved in a retrofitting decision. In most cases, the Seismic Rating System is used as a guide to making retrofitting decisions, but not

as the final word. Common sense and engineering judgement will be necessary in weighing the

actual costs and benefits of retrofitting, against

the risks of doing nothing. Also, the effect on

the entire highway system must be kept in mind.

The preliminary screening process recommended

in the new Retrofitting Manual is demonstrated

in Figure 2, where the terms are defined as follows: A is the acceleration coefficient for the

bridge site; / is the bridge importance; V is the

vulnerability rating of the bridge; S is the soil

site coefficient; and E is the seismic hazard

rating, which is based on the acceleration and

site coefficients.

8. DETAILED EVALUATION

Two alternative methods for the detailed

evaluation of existing bridges are currently

available. One is based on a quantitative

assessment of the "capacities" and "demands" of

individual components of a bridge structure. The

other evaluates the lateral strength of the bridge

as a new structure.

8.1 Capacity/Demand Method

The first method was proposed in the 1983

FHWA Retrofit Guide and has been used in a modified form by CALTRANS since the early

1980's. In this method, the results from an

elastic spectral analysis are used to calculate the

force and displacement "demands" which are then compared with the "capacities" of each of

the components to resist these forces and

displacements. In the case of reinforced concrete columns, ultimate capacities are modified to reflect the ability of the column to resist postelastic deformations. Capacity/ Demand (C/D)

ratios are intended to represent the decimal

fraction of the design earthquake at which a local failure of the components is likely to occur. Therefore, a C/D ratio less than 1 .0 indicates that component failure may occur during the design

earthquake and retrofitting may be appropriate.

An overall assessment of the consequences of

local component failure is necessary to determine

the need for retrofitting. Retrofitting should be

considered when an assessment indicates that

local component failure will result in unacceptable overall performance. The effect of

potential retrofitting may be assessed by

performing a detailed re-evaluation of the

retrofitted bridge.

The determination of what constitutes a serious

consequence of component failure will depend

on the importance of the bridge. Collapse of the

structure is serious in almost all cases since there

is always a potential for loss of life in such an

occurrence. In other cases, severe distortions or

critical loss of strength will impair the ability of

the bridge to carry emergency traffic which is unacceptable for certain important bridges.

Repairability of seismic damage is also a consideration. If repairs can be made quickly

without serious delays to traffic, damage may be

acceptable. This is another area in which

engineering judgement is required.

Once it has been determined to consider

retrofitting, acceptable methods may be selected

from among a number of those suggested in the

52
Retrofitting Manual. If the seismic response of

the structure is affected, then a reanalysis should

be performed and new set of component C/D

ratios calculated. The new C/D ratios will

reflect a change in the size of the earthquake that will cause serious damage. A decision to use any retrofitting method will be based on a

relative benefit-to-cost analysis. Hypothetically,

this benefit-to-cost analysis may be objective and

rigorous, but it is more likely that it will be

subjective and based, in large part, on

judgement.

8.2 Lateral Strength Method

The Retrofitting Manual also provides an

alternative analysis approach which examines the

lateral strength of the bridge as a system, or at least individual segments of the bridge as a system, and determines, through an incremental

collapse analysis, the load-deformation

characteristics of the bridge up to collapse. The

fraction of the design earthquake that can be

resisted without collapse is then an indicator of

the need for retrofitting and the extent of

strengthening required. This procedure therefore

determines the strength and ductility of the

critical collapse mechanism but it can also be

used to identify the onset of damage when

serviceability criteria may be important. The

method emphasizes deformation capacity rather

than strength since, although strength is important, it is less important than the ability to

sustain substantial deformations without collapse.

It is believed that fewer bridges assessed under

this procedure will be found in need of retrofit than by the C/D ratio method. When retrofit is required, it should be less extensive. The

increased level of effort required of the designer

will then be offset by reduced retrofit costs in the field.

9. RETROFIT MEASURES

There are two alternative strategies that a designer may adopt when faced with retrofitting a bridge. One is based on conventional

strengthening techniques which increase the

capacity of the structure to meet the likely

demand. This is the most common approach

used in the United States at this time. The

second strategy is based on reducing the demand

on the structure such that its existing capacity is sufficient to withstand the given earthquake.

This latter approach involved the use of an

earthquake protective system, such as seismic

isolation or the addition of a mechanical energy

dissipation device. Both strategies are described

and detailed in the Retrofitting Manual.

9.1 Conventional Retrofit Measures

The new Retrofit Manual describes conventional

retrofit measures for bearings, seats, and

expansion joints, including joint and bearing

restrainers, bearing seat extensions, and overall

bearing replacement. The Manual also discusses

techniques for strengthening reinforced concrete

substructures through column jacketing and

wrapping, cap/column and column/footing joint

strengthening, and cap beam retrofitting.

Retrofit measures for foundations include

strengthening footings for flexural and shear

strength, ensuring adequate reinforcement

anchorage, and sufficient strength in the

pile/footing connection to resist overturning or

uplift, along with problems related to abutments

and approach slabs. Finally, the Manual also

discusses problems associated with hazardous

sites, including sites with liquefiable soils,

bridges on or near unstable slopes, and bridges

crossing or near active faults. These techniques

represent the current state-of-the-art; however,

the art is changing rapidly at this time.

9.2 Earthquake Protective Systems

The term "earthquake protective system" includes

passive and active devices which can be installed

in a bridge to minimize the seismic demand on

the members of the structure. Active systems

are considered outside the scope of the Manual

but passive devices are being used in several

states as a cost-effective retrofit measure for many bridge types. Passive systems discussed in the Manual include mechanical systems, which

simply dissipate energy and thus reduce

response, and seismic isolation systems, which

53
change the natural period of a bridge so that earthquake loads are significantly reduced. The

Manual provides a discussion on seismic

isolation concepts and on some of the options

available at this time for design and

implementation.

10. SUMMARY

The new "Seismic Retrofitting Manual for Highway Bridges" contains detailed information

about each of the major steps in the bridge

seismic retrofit process, as shown in Figure 1. It describes procedures for preliminary screening of

bridges along with two alternate procedures for the detailed evaluation of them. These

evaluation methods include a quantitative

evaluation of the C/D ratios for individual bridge

components, and an alternative method based on

assessment of a structure's lateral strength.

The procedures for evaluating bridges for

retrofitting also include the identification and

assessment of retrofit measures. A number of

potential retrofitting measures and retrofit design

requirements are discussed in the Manual.

Specifically, retrofit measures for the types of

bridge components which have performed poorly

during past earthquakes are discussed in detail. Retrofitting by these or other equivalent methods

should be considered when components are identified by a detailed evaluation as being

deficient. The decision to use a retrofitting

scheme will be based on an assessment of its effectiveness in preventing unacceptable overall

performance, the cost of retrofitting, and the

remaining service life of the bridge. The

Retrofitting Manual also includes several worked

example problems intended to help illustrate the

process of planning the retrofitting of a typical

highway bridge.

Detailed design of retrofit measures should be

performed using the guidelines contained in the

Retrofitting Manual in conjunction with the current AASHTO bridge design specifications.

If possible, components which are selected for

retrofitting should generally be strengthened to conform to the specifications for new

construction, even though the structure may

otherwise be seismically deficient.

11. ACKNOWLEDGEMENTS

The "Seismic Retrofitting Manual for Highway

Bridges" was developed under the FHWAsponsored project titled "Seismic Vulnerability of

Existing Highway Construction, FHWA Contract

DTFH61-92-C-00106, by the National Center for Earthquake Engineering Research. The authors

acknowledge the assistance of Dr. John Kulicki

of Modjeski and Masters Inc., and Dr. Roy A.

Imbsen of Imbsen & Associates Inc., in the

review phases of the Manual. Any opinions,

findings, and conclusions or recommendations

expressed in the Retrofitting Manual or this paper are those of the authors and do not

necessarily reflect the view of NCEER, the

FHWA, or the U.S. Department of

Transportation.

54
Table 1. Seismic Performance Category

Acceleration

Coefficient

Importance Classification

Essential Standard

A < 0.09

0.09< A<0.19

0.19 < A < 0.29

0.29 < A

B

C

C

D

A

B

C

C

55
DETERMINE

SEISMIC PERFORMANCE

CATEGORY

SEISMIC

PERFORMANCE

CATEGORIES B, C & D

SEiSMIC

PERFORMANCE

CATEGORY A

PRELIMINARY

SCREENING

(Chapter 2)

t

DETAILED EVALUATION

OF AN EXISTING BRIDGE

(Chapter 3 & Appendices A&B)

DESIGN RETROFIT

SCHEMES

(Chaoters 4. - 8)

NO

RETROFITTING

NOT

REQUIRED

Figure 1. The Seismic Retrofitting Process

56
DETERMINE ACCELERATION

COEFFICIENT (A) AND

IMPORTANCE (

I

)

I

DETERMINE

SEISMIC PERFORMANCE

CATEGORY

*

"\ r SEISMIC SEISMIC

PERFORMANCE PERFORMANCE

CATEGORY A CATEGORIES 8, C & D

T

COMPILE

INVENTORY

(V&S)

r BRIDGE

RANK

OTHER

ISSUES (0)

T T

PRIORITY INDEX

P=r"(R, 1.0)

RETROFITTING

NOT

RECUIRED

Figure 2. The Preliminary Screening Process

57

Development of Innovative Seismic Response Control of Bridges

by

Kazuhiko Kawashima", Shigeki Unjoh 2

' and Hidetoshi Mukai3'

ABSTRACT

This paper presents variable dampers in which the damping force can be changed

depending on the structural response, and the

variable stiffness control in which the

instantaneous natural period of structures can be

changed so that the resonance with the

predominant ground motion be avoided. A

cylinder/piston type model of the variable

damper was developed, and the effectiveness of

the variable dampers was verified by a shaking

table test. An experimental model of the

electro-Theological (ER) fluid damper is also being developed, and its performance is being

studied by a series of dynamic loading test. Key Words .'Variable Damper,

Variable Stiffness Control,

Electro Rheological Fluid,

ER Damper, Damper, Bridges,

Seismic Response Control

1. INTRODUCTION

Variable damper, in which damping

coefficient can be variable dependent on

structural response, for structural response

control against an earthquake was proposed by

the authors in 1989 (Kawashima and Unjoh

1989, Taguchi et al 1990), and various studies

have been made at the Public Works Research

Institute. This investigation was initiated to develop an appropriate seismic control

algorithm for bridges. Because the active

control requires a lot of energy to control the

bridge seismic response in a satisfactory level, an alternative approach was required.

The concept of variable dampers is very

unique. The variable damper works as a load

transmitting member as well as a usual energy

dissipator. The variable damper has been

effectively adopted in a field of mechanical

engineering, in particular in a suspension

system for an aircraft and an automobile. An

active suspension for improving driving

comfortableness and stability of automobiles is one of such examples (Sugawara 1986,

Millikan 1987). However, it is a new idea to apply the variable damper for structural response control against an earthquake. Two

variable dampers were developed at the Public

Works research Institute. One adopts a

cylinder/piston type, and the other uses the

electro-rheological (ER) fluid. Variable stiffness is an interesting field of

extension of the variable damper. If one shifts the natural period of structures, the structural response may be significantly reduced by

avoiding the resonance with the ground motion.

This paper presents the current studies for the variable dampers and variable stiffness for response control of bridges against an

earthquake.

2. VARIABLE DAMPER

2.1 Concept of a Variable Damper

A unique technic in application of

dampers to bridges is to distribute the seismic

lateral force to as many piers as possible. By

1) Research Coordinator for Underground Development,

Planning and Research Administration Department,

Public Works research Institute, Ministry of

Construction, Tsukuba Science City, 305 Japan

2) Senior Research Engineer, Earthquake Engineering

Division, Earthquake Disaster Prevention Department,

ditto 3) Assistant Research Engineer, ditto

59
using the dampers in an over-damped zone, the dampers work as force transmitting members

from deck to piers. The dampers used for such

purpose are called as the damper stoppers. It should be noted that by adopting the damper

stoppers, thermal force induced in substructures

due to elongation/shrinkage of deck associated

with the temperature change can be reduced

because the damper stoppers do not resist for the low-rate movement. The damper stoppers

have been successfully implemented in many

bridges (Izeki 1980).

Because few energy dissipation is made

in the damper stopper, it is effective to decrease

the damping ratio so that energy dissipation can be maximum when the bridge is subjected to a ground motion caused by a large earthquake.

Because the damper stopper is effective for preventing a small deck vibration associated

with breaking loads and wind effects, the

function as the damper stopper is effective at small deck displacement. On the other hand

when the deck response becomes excessive

during an earthquake, the stopper is required to present further build up of the deck response. Thus, the variable dampers with the

following requirements may be superior to the

existing dampers and damper stoppers:

1) Damping coefficient is very large during a small deck vibration for preventing the deck

vibration due to breaking loads of vehicles and

wind effects. It is movable against the low rate motions such as the elongation of deck by

temperature change.

2) Against the seismic response with the

deck response over a certain value, the damping

coefficient is set so as to make energy

dissipation maximum.

3) Against excessive deck response during an

earthquake, high damping coefficient is required

to function as a stopper. Smooth and gradual

increase of the damping force may be effective

for preventing shocks.

Although various types of the variable

damper can be made, the simplest structure may

be an piston-cylinder viscous damper as shown

in Fig. 1. A by-pass is installed between the

cylinder-cells divided by the piston. Damping

coefficient of the damper can be controlled by

varying an amount of viscous flow of the

by-pass. Various technologies for controlling

such flow are available. External energy

required for such control is generally much

smaller than that required for the active control.

2.2 Effectiveness of Variable Dampers

Fig. 2 shows the deck displacement

dependent damping ratio of the variable damper

proposed for bridges (Kawashima, Unjoh and

Shimizu 1991, Kawashima, Unjoh, Nagashima

and Shimizu 1991, Kawashima, Unjoh and

Shimizu 1992). Various control algorithm may

be used such as to vary hi, h2 and h3 in time

depending on bridge response for optimizing

the deck response. The damping ratio can be

varied depending on both/either the deck

velocity and/or the deck displacement

(Kawashima, Unjoh and Shimizu 1991). Partial control in which the control is made only

partially in time can be made (Kawashima,

Unjoh and Shimizu 1991). For simplicity, the

damping ratios hi, h2 and h3 are assumed

here to be constant during an earthquake. It should be however noted that any relation

between the damping ratio vs. the relative displacement and/or relative velocity can be

incorporated in the analysis. In fact, various

control algorithm is being developed. Ban-Ban

control algorithm (Feng and Shinozuka 1990)

and more sophisticated time dependent

algorithm is also being developed (Yang, Li

and Wu 1994, Sun and Goto 1994).

For studying the effectiveness of the

variable damper, a simple span girder bridge, as shown in Fig. 3, with a span length of 30m was

analyzed. The response in longitudinal direction was considered. Although the variable damper

has better implementation for multi-span

continuous bridges, a simple span bridge was

considered for analysis for simplicity. The deck was assumed to be supported by elastic bearings, and two variable dampers were

installed between the deck and the piers.

The weight of superstructure and stiffness

60
of a pier were assumed as 2368 KN and 17350

KN/m so that the fundamental natural period of

the bridge is 0.5 second if the both bearings are hinged. The lateral stiffness of rubber bearings

per pier is assumed as 6276 KN/m so that the fundamental natural period of the bridge

supported by the rubber bearings be 1.0 second.

Damping ratio of the bridge was assumed as

0.02. The bridge response with and without

control by the variable dampers was computed.

The damping ratios in Fig. 2 were assumed as

hi = 0.5 and hi = h3 = 3. As shown in Table 1, the peak deck displacement and the peak deck

acceleration decrease to 26% and 44% of those

without control, respectively.

The damping force and stroke required

for the variable dampers are 632 KNX2 =

1264 KN and 5.78 cm, respectively. Because

the weight of deck is 2368 KN, the damping

force is 53 % of the deck weight. The variable

dampers with this specification can be designed

and fabricated within the current scope of

dampers technology.

23 Verification of Variable Dampers

through Shake Table Tests

For verifying the feasibility of the

variable dampers, a prototype variable damper

was developed as shown in Photo 1 (Kawashima, Unjoh and Mukai 1993). It is a piston-cylinder type viscous damper with a by-pass installed between two cylinder-cells

divided by the piston. The model was designed

so that the maximum damping force is 200 KN

and the stroke is ±13 cm. Damping force can be controlled by a personal computer depending on a relative displacement and/or a velocity

between a deck and a pier. The electricity

required for control is only 50 W. This is extremely small as compared with the energy

required for the active control. A series of dynamic loading tests was made for the model. Fig. 4 shows a hysteresis

loop of the damping force vs. stroke. Both

command signal and actual data are presented.

The hysteresis loop obtained by the loading test

is very close with the command.

For verifying the effectiveness of the

variable damper, a series of shake table tests was made. A simple span bridge model was

fabricated on a shake table as shown in Fig. 5. The deck is 7.6 m long and 390 KN in weight.

The piers are 1.6 m long. The deck is supported

by four rubber bearings, and the prototype

variable damper was installed between the deck

and the pier as shown in Photo 2. The model bridge was excited by a ground acceleration. The peak table acceleration was set as about 180 cm/sec

2 and 413 cm/sec2

. The excitation by the table motion with peak

acceleration of 413 cm/sec

2 was not made for the model without control, because the deck

response became excessive.

Figs. 6 and 7 compares the deck

displacement of the model bridge, with and

without control by the variable damper,

subjected to the table acceleration with peak

acceleration of about 180 cm/sec

2

. Predicted

response by the analysis which will be

described later is also presented. The deck

responses are 376 cm/sec and 9.2 cm for no

control, and they are decreased to 203 cm/sec

2

and 1.3 cm by the variable damper. The

maximum damping force required for this control is 71 KN, which corresponds to 18% of

the deck weight.

A simulation of the experiments by the

analytical method presented in the previous

chapter was made. The deck, piers and rubber

bearings were idealized by a discrete linear

frame. Figs. 6 and 7 compare the deck

displacement and the damping force developed

in the variable damper between the experiments

and analyses. The predicted responses are quite

close with the measured responses.

3. VARIABLE STIFFNESS CONTROL

3.1 Concept

Most of bridges with usual size have the natural period ranging from 0.3 to 1.5 second,

and this range corresponds to the predominant

period of earthquake-induced ground motions.

61
Thus the resonance of the bridges response with

the ground motion tends to be developed during an earthquake. The bridge response could be

reduced provided the resonance of bridge

response with the ground motion could be

avoided.

Although it is hard to predict the

predominant period of ground motion prior to an earthquake, it can be evaluated time by time

basis during an earthquake. The moving

window analysis may be used for evaluating the

instantaneous predominant period of ground

motion T*. The resonance of bridge response

with the g ^und motion may be avoided if the

instantaneous natural period of bridges T be

shifted from the instantaneous predominant

period of ground motion T* during an

earthquake.

There are various methods to shift the

instantaneous natural period of bridges during an earthquake. Flexibility of bearings may be

changed, and active tendons/bracings may be

adopted.

If one provides the variable dampers for

a bridge which is elastically supported by

rubber bearings, the natural period of the bridge can be varied by changing the damping

coefficient of the variable dampers. If one makes the damping ratio of the variable damper

very large, the variable dampers would acts as the damper stoppers. In another word, they

work as force transmitting members. On the

other hand, if one makes the damping ratio small, the natural period of the bridge takes the

value computed from the stiffness of the rubber

bearings and piers. Thus although the variable

dampers are essentially dampers, they can be

used in an over-damped range as force

transmitting members to change the natural

period of structures. The variable dampers have

various advantage for controlling the

instantaneous natural period of bridges because

it is more stable for long term use than the

active tendons/bracing systems. The energy

required is also much less. Fig. 8 shows the algorithm of the variable

stiffness control. The instantaneous predominant

period of ground motion T* at time U is computed by the moving window Fourier

spectrum. The damping coefficient of variable

dampers is varied so that the instantaneous

natural period of the bridge T be shifted to prevent the resonance with the ground motion.

Although several algorithm can be

considered, the period shift is assumed here as T = aT (1)

where,

T : instantaneous natural period of a bridge at time u T*: instantaneous predominant period of an

earthquake ground motion at time u a : control coefficient

The coefficient a has to be determined

depending on the undamped natural period of

the bridge To and the instantaneous predominant

period T* of an ground motion.

3.2 Effectiveness of

Variable Stiffness Control

To study the effectiveness of the variable

stiffness control, a three-span continuous girder

bridge, as shown in Fig. 9, with a span length

of 90m is analyzed. The response in longitudinal direction is considered. The

superstructure of the model is supported by

elastic bearings, and four variable dampers with

the same characteristics are installed between

the deck and the piers. The weight of the superstructure and spring stiffness of the rubber

bearings are assumed as 4719 KN and 6247

KN/m, respectively. The stiffness of piers is assumed so that the fundamental natural period

of the bridge is 0.5 sec when the deck is supported by hinges. The stiffness of rubber

bearings is set so that the fundamental natural period of bridge is 1 sec when the deck is supported by the elastic bearings. The damping

ratio of the bridge without control is assumed

as 0.02.An acceleration ground motion as shown

in Fig. 10 is used as an input acceleration. The

instantaneous predominant frequency (1/T*) of

the input acceleration is also shown in Fig. 10.

62
Because the predominant period of ground

motion is about 1 sec, a in Eq. (1) was

assumed as 1/2. Fig 11 shows how the natural period of

the bridge change depending on the damping

coefficient of the variable dampers. At the damping coefficient C of about 4200 KN»

sec/m, the fundamental natural period of bridge

becomes very large, because this corresponds to the critical damping of the structure. Although

the period shift may be made in the damping

coefficient range of 0 to over 8000 KN^sec/m,

it is assumed here that the period shift be made

in the over-damped range(damping coefficient

C ^ 4200 KN'sec/m).

Fig. 12 compares the deck acceleration

and displacement of the bridge with and

without control. The peak displacement and

peak acceleration of deck decreases to 17% and

50% of those without control. The bending

moment of pier bottom decreases to 52% of the one without control. The peak damping force

required for the four variable dampers is 761

KNX4 = 3040 KN and peak stroke is 3.1cm.

33 Development of a Variable Damper

for Variable Stiffness Control

For verifying the variable stiffness control, a variable damper is being developed.

Although Eq. (1) was originally proposed, some

modifications are required for making the control simple and effective. Firstly it is modified so that the control is made only when

the instantaneous natural period T of the

structure at time t is between fiiT* and feT*

.

The coefficients fii and $2 are tentatively

proposed to set as 3/4 and 5/4, respectively.

Secondly, it is modified so that T increases to

1.5 times as large as 7*.

In this modification, Eq. (1) is written as 7 = 1.57* ^ TF (#7* ^ 7 ^ 02T*)

(2)

With use of the variable damper being

developed, a shake table test similar with the one presented in 2.3 is planned to verify the

effectiveness of variable stiffness control

algorithm.

At this moment, an algorithm to detect

the predominant period T* is being studied.

Some time delay associated with the time

required for AD conversion, a computation of

running Fourier spectrum and DA conversion at each time step requires to be shorten.

5. DEVELOPMENT OF A VARIABLE

DAMPER USING ER FLUID

5.1 Design and Development of

a ER Damper

Electro-rheological (ER) fluid is an

innovative material which has a potential

application for a wide range of engineering

fields. The viscosity of ER fluid can be

changed depending on the electric field applied.

Because reliable ER fluid with high application

to practical use was recently developed, various

application has been initiated especially in mechanical systems (Sasada et al 1975A,

Sasada et al 1975B, Ushijima et al 1988,

Duclos 1988, Ehrgott and Masri 1992).

In a cylinder-type damper system with an orifice and a piston as shown in Fig. 13, the

damping force of the ER damper may be given

as F = Fv + FE (3)

12-L-A 2 -v Fv = B^JP

71 (4)

Fe = ~~~Tf(F^) (5)

where,

F : damping force developed by an ER

damper

Fv :damping force developed by fluid viscosity

through an orifice Fe rdamping force developed by ER effects A : pressure area of a piston v : velocity of a piston

L,B,H : length, width and distance of electrodes n : viscosity of ER fluid Ty(E) : shear stress developed by ER effects E : applied electric field strength

63
Shear stress developed by the ER effects Ty(E) is controlled by the applied electric field strength E between the electrodes, and is written as

r,(E) = a-& (6)

where

a , & : coefficients dependent on the material

According to the dynamic loading tests for an ER fluid used in this study, which will

be described later, they are a = 2.06X10" 4

>> » = 1.7 > W

It should be noted in Eq. (3) that only FE

can be changed by controlling the applied

electric field strength. Fig. 14 shows an

illustration of Eq. (3) for a steady state constant displacement loading. The damping force due

to ER effect is frequency independent while the

damping force due to fluid viscosity is frequency dependent. Therefore, it is required

that the damping force developed by the ER

effect be much larger than the damping force

developed by the fluid viscosity to control

structural response in a wide range by varying

the applied electric field strength.

5.2 Development of ER Damper and

a Dynamic Loading Test

An experimental model of ER damper is being developed. It was designed so that the

maximum damping force and the maximum

stroke of the piston are 20 kgf and ±40 mm,

respectively. Fig. 15 and Photo 3 show the

model. The damper is 521 mm long, and has a

orifice (electrodes) of 134 mm long and 3 mm

wide. Pressure area of the piston is 2,089 mm2

. Rubber sleeve is used to enclose the ER fluid. Colloidal solution which includes silicon oil with mixing carbon powder is used as the ER

fluid. The initial viscosity of the ER fluid is 1.3 X10~ 8 kgf# s/mm 2

, and the shear stress z> at 2 kV/mm is 68.0 x 10

6 kgf/mm 2

. The voltage

of the electrodes is controlled between DC 0

and 6kV.

The performance of the model ER

damper is being studied through a series of

dynamic loading tests. Tests were made for the

displacement, velocity and electric field dependency of the damping force.

Fig. 16 shows the force vs. displacement

hysteresis loops. With the loading displacement

being ±20 mm, the loading frequency was

varied from 0.5 to 10 Hz. It is seen in Fig. 16

that because there is a static friction of about 10

kgf between the piston and the cylinder,

accuracy of control with small damping force is poor. It is also seen in Fig. 16 that there is a phase lag. This is caused by the compressibility

of ER fluid and deformation of the rubber

sleeve.

Fig. 17 shows the voltage vs. the

damping force relation. The damping force at zero voltage represents the damping force due

to fluid viscosity. The damping force increases

as the voltage increases due to the ER effect. However, the damping force due to ER effect is smaller than the damping force due to fluid viscosity. An improvement for making ER

effect larger is required.

6. CONCLUDING REMARKS

The variable dampers and variable

stiffness control which are being developed at the Public Works Research Institute was

described. Because the variable damper is a simple extension of the damper stopper

technology which has been used in Japan for

long time, it is expected that the variable

dampers be implemented in the near future. Various research on the variable stiffness including development of the prototype model

and shake table tests are currently being made.

An experimental model of ER damper is being developed, and a series of dynamic

loading tests are being made. Although it was

developed as designed, it was found that the damping force developed by the ER effect is smaller that the damping force developed by

the fluid viscosity. It is required to modify the damping force due to ER effect large enough so

that the control by ER effect can be made in a

64
wide range of damping force. ACKNOWLEDGMENTS

The research presented herein is being

made as a part of the U.S. -Japan Cooperative

Research on Hybrid Control of Bridges under

auspices of the Panel on Wind and Seismic

Effects, UJNR. The authors extend their sincere

appreciation to the strong support of Dr. S. C.

Liu, NSF, Dr. H. S. Lew, NIST, Mr. J. D. Cooper, FHWA, Professor M. Shinozuka,

Princeton University, Professor J. N. Yang and

Professor M. Feng, University of California at Irvine.

REFERENCES

1) Duclos, T., G. (1988) "Design of devices

using electro rheological fluids," SAE

Technical Paper No. 881134

2) Ehrgott, R., C. and Masri, S., F. (1992)

"Modeling the oscillatory dynamic

behavior of electro rheological materials in shear," Smart, Mater. Struct. 1

3) Feng, Q. and Shinozuka, M. (1990) "Use

of a variable damper for hybrid control of

bridge response under earthquake, U.S.

National Workshop on Structural Control

Research, University of Southern

California, CA, U.S.A.

4) For example, Izeki, H. (1980) " Viscous

shear stopper, proposal to the

multi-continuous girder bridge," Bridges

and Foundations, pp. 30-33 (in Japanese)

5) Kawashima, K. and Unjoh, S. (1989).

"Development of Hybrid Control

Technology." Civil Engineering Journal,

Tokyo, Japan, pp. 2-5, Vol. 32-6 (in

Japanese).

6) Kawashima, K, Unjoh, S., Nagashima, H.

and Shimizu, H. (1991). "Current research

efforts in Japan for passive and active

control of highway bridges against

earthquakes." Proc. 23rd Joint Meeting,

U.S.-Japan Panel on Wind and Seismic

Effects, UJNR, Tsukuba, Japan, 187-209.

7) Kawashima, K, Unjoh, S. and Shimizu, H.

(1991). "Earthquake response control of

highway bridges by variable damper."

Proc. of Colloquium on Control of

Structures, Japan Society of Civil

Engineers, Part. B, pp.221-224, July, 1991

(in Japanese)

8) Kawashima, K, Unjoh, S. and Shimizu, H.

(1992). "Seismic response control of

highway bridges by variable damper." 24th

Joint Meeting, U.S.-Japan Panel on Wind

and Seismic Effects, UJNR, Gaithersburg,

May, 1992, NIST Special Publication 843

9) Kawashima, K, Unjoh, S., Iida, H. and

Mukai, H. (1992). "Effectiveness of the variable damper for reducing seismic

response of highway bridges, Proc. of 2nd

U.S.-Japan Workshop on Earthquake

Protective Systems for Bridges, Tsukuba,

Dec. 7-8, 1992, Technical Note of the

Public Works Research Institute, No. 3196.

10) Kawashima, K, Unjoh, S. and Mukai, H.

(1993) . "Seismic response control of

highway bridges by variable dampers."

25th Joint Meeting, Panel on Wind and

Seismic Effects, UJNR, Tsukuba, Japan,

Technical Note of Public Works Research

Institute. 11) Kawashima, K, Unjoh, S. and Mukai, H.

(1994) "Seismic response control of

highway bridges by variable stiffness control," The 3rd US-Japan Workshop on

Earthquake Protective Systems fro Bridges,

Berkeley, CA, U.S.A.

12) Kawashima, K, Unjoh, S., Suzuki, S. and

Endoh, S. (1994) "Application of

electro-rheological fluid for variable

damper", The 3rd US-Japan Workshop on

Earthquake Protective Systems for Bridges,

Berkeley, CA, U.S.A.

13) Milliken, Jr., W., F. (1987). "Lotus active

suspension system." Proc. 11th

International Conference on Experimental

Safety Vehicles, Washington D.C, U.S.A.,

467-475.

14) Sasada, T., Kishi, T. and Kamijo, K.

(1975A) "Electro viscous effects of

65
polymer and solid particle solutions," Proc. Japan Society of Mechanical Engineers,

Vol.41, No.342

15) Sasada, T., Kamijo, K. and Kishi, T. (1975B) "Electro viscous effects of

uni-phase liquid," Proc. Japan Society of

Mechanical Engineers, Vol.41, No.343

16) Edited by Scott, D. and Yamagichi, J. (1983) "Solidifying fluid transforms

clutches and flow valves," Automotive

Engineering, Vol.91, No. 11

17) Sugawara, F. (1986). "Electronically

controlled shock absorber system used as a road sensor which utilizes super sonic

waves." Proc. Society of Automotive

Engineers, Milwaukee, Wisconsin, U.S.A.,

6.15-6.25.

18) Sun, L. and Goto, Y. (1994) "Fuzzy

control of bridge vibration by using

variable dampers, " The 3rd US-Japan

Workshop on Earthquake Protective

Systems for Bridges, Berkeley, CA, U.S.A.

19) Taguchi, J., Iwasaki, T., Adachi, Y.,

Sasaki, Y. and Kawashima, K. (1990).

" U.S.-Japan cooperative research program on hybrid control of seismic response of

bridge structures." 22nd Joint Meeting,

U.S.-Japan Panel on Wind and Seismic

Effects, UJNR, Gaithersburg, Maryland,

USA, May, 1990, NIST Special

Publication.

20) Ushijima, T, Takano, K. and Noguchi,

T.(1988) "Rheological characteristics of

ER fluids and their application to anti-vibration devices with control

mechanism for automobiles," SAE paper

#881787

21) Yang, J. N., Li, Z. and Wu, J., C. (1994)

"Hybrid protective systems for seismic-excited bridges," The 3rd

US-Japan Workshop on Earthquake

Protective Systems for Bridges, Berkeley,

CA, U.S.A.

66
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67
CONTROLLER

Fig. 1 Variable Damper

DAMPING COEFFICIENT C

RELATIVE DISPLACEMENT (xdu)

Fig. 2 Analytical Idealization of Damping Force vs. Relative Displacement Relation

68
ISOLATOR

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Fig. 3 Bridge Analyzed and Analytical Idealization

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Fig. 4 Damping Force vs. Relative Displacement Relation from Dynamic Loading Test

of Variable Damper

69
7.6

WEIGHT 390 KN

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Fig. 5 Shake Table Test of a Model Bridge with a Variable Damper

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Fig. 6 Measured and Predicted Response of a Model Bridge without Control,

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70
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Fig. 7 Measured and Predicted Response of a Model Bridge with Control,

subjected to a Table Motion with Peak Acceleration of 185 cm/sec

71
TSTART

Time ti

CALCULATION OF PREDOMINANT

FREQUENCY OF GROUND MOTION

UJo

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Time t

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PREDOMINANT

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CRITICAL

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Fig. 8 Control Algorithm of Variable Stiffness

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Fig. 9 Model Bridge Analysed

72
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200 400 600 800

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Fig. 11 Dependence of Natural Period of the Bridge on Damping

Ratio of Variable Dampers

73
LU^ E

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Fig. 12 Effectiveness of Variable Stiffness Control

74
7T 2 2

Pressure Area of Piston : A= — (D - Dc ) 4

Width of Electrode : B = 7T (D + H)

Fig. 13 Cylinder Type ER Damper

(a) Damping Force developed by (b) Damping Force developed by

the Fluid Viscosity the ER Effects

Fig. 14 Illustration of the Damping Force developed by the ER Damper

75
ER Fluids Electrode

Rubber Sleeve

C Unit : mm]

Fig. 15 Experimental ER Damper Developed

120 ~ 100

o

80 -

LOADING DISPLACEMENT : 20 mm

2 4

Voltage (KV)

Fig. 17 Relation between Damping Force and Applied Voltage

76
Displacement (cm)

(3) Applied Voltage = 6kV

Fig. 16 Hysteresis Loops of Force-Displacement Relation

77
Photo 1 Prototype Model of Variable Damper

Photo 2 Model Bridge and Prototype Variable Damper Model

78
Photo 3 Experimental Model of ER Damper

79

Innovative Seismic Testing Procedures

by

Stephen A. Mahin* and S. C. Liu*

ABSTRACT

Experimental methods remain a key

factor in developing and verifying seismicresistant design and analysis methods. The

need for experimental research is likely to increase in the coming years as engineers

develop innovative structural systems utilizing

high performance materials, active and passive

control systems and special structural systems.

Moreover, experimental results are needed to develop and validate numerical models that can

take advantage of high performance

workstations and other advanced computer

capabilities becoming available within the

design industry. Recent experiences with

earthquakes also suggests that greater emphasis

will be placed in the years to come on more

quantitative prediction and control of damage,

leading to increased needs for verified design

and analysis methods.

As attention is turned to more complex

and larger structural systems, especially those

comprising the civil infra-structure, researchers

are challenged to develop increasingly complex

experimental test methods. In this paper, a brief overview of available experimental

methods is presented and future directions are examined. Several recent examples of special

testing arrangements used for infrastructure

elements will used to highlight the paper. The

first of these relate to simple tests of large

single degree-of-systems requiring the cyclic

application of large lateral loads. Other tests requiring refined three-dimensional control of

the forces and displacements applied to multiple

degree of freedom specimens are then

presented. Extension of on-line computer

control techniques to the realistic simulation of

dynamic aspects of seismic structural response

are then briefly described.

KEYWORDS: experimental methods,

earthquakes; seismic-resistant design; infra- structure; buildings.

INTRODUCTION

The ease with which experimental

research may be performed has improved

markedly in recent years with the increased

availability of electro-hydraulic actuator

systems to impose forces and displacements,

the greater reliability and accuracy of

instrumentation and, most significantly, the

development of high speed, relatively low cost

digital data acquisition and control systems. In most instances, experiments used to assess the

quasi-static inelastic cyclic behavior of

structures and structural components remain

relatively simple. In fact, the increasing need to carry out field tests on very large components

of infrastructure systems results for practical

and economical reasons in the use of very

simple, low tech approaches to loading.

However, complications can arise due to the

magnitude of the required loads and the need to investigate the three dimensional response

characteristics of specimens. In other cases,

quasi-static loading is not sufficient and loading

histories more representative of the response of

the specimen to actual earthquake ground

motions are required.

Shaking tables enable the researcher to simulate within the laboratory the dynamic

motion of the ground supporting a structure

during an earthquake. These simulators remain

the most realistic seismic test procedures. None

the less, available shaking tables significant

place significant limitations on the weight,

strength and size of specimens that can be

tested. In the U.S. several moderate sizedables

are available with sizes up to 20 ft (6.1 m)

square. The largest of these earthquake

Nishkian Professor of Structural

Engineering, Dept. of Civil Engineering,

University of California, Berkeley, CA

** Program Manager, Earthquake Hazards

Mitigation, National Science Foundation,

Arlington, VA

81
t simulators, located at the University of

California at Berkeley, is being upgraded with

funding from the National Science Foundation

to subject specimens weighing up to 70 tons to

full three-dimensional seismic excitations.

Several groups in the U.S. are currently

considering the technical and economic

feasibility of large shaking tables.

In addition, many infrastructure

elements, like bridges and pipelines, are supported at points not expected to have the

same ground motion. Several laboratories in

the U.S. (e.g., the University of Southern

California, Energy Technology Research

Center, University of California at Berkeley,

etc.) have developed several relatively small

shaking tables capable of subjecting such

spatially distributed systems to multiple support

excitations.

Dynamic loading can be also imposed

on a structure by means of underground

explosions, and by a variety of mechanical

vibration generators. Often vibration

generators utilize reciprocating masses and can

only develop simple harmonic excitations in the

structure. Recent efforts have been under taken

to utilize controllable gas jets or mechanical

extruding devices to impose variable force

excitations of short duration.

Because of the need to test large

structures under dynamic excitations several

efforts have been carried out to develop on-line

computer control procedures. These

experiments outwardly resemble typical quasistatic tests, but a computer is used during the

test to monitor the specimen and determine its deformation response. Ideally, the specimen

undergoes the same displacement history as if it were subjected to a particular ground motion.

These procedures are similar to typical

computer programs except that the restoring

force characteristics of the test structure are measured experimentally, rather than computed

analytically. The procedures have been shown

to produce accurate results on large structures and have a great deal of flexibility to carry out

innovative tests. Typically, tests are carried out

slowly, but recent efforts have focused on

developing real time loading capabilities. A full scale, five story masonry structure was recently

tested using such techniques at the University

of California at San Diego.

CYCLIC PLANAR LOADING OF

LARGE STRUCTURAL SYSTEMS

In most cases, planar tests of structural

components and systems can be carried out in a simple manner. The specimen and actuator can

be attached to a suitable reaction floor or frame,

and the actuator can be used to impose the

appropriate load or displacement history.

However, in some cases the loads required

becomes so large that the size (and cost) of the

required reaction frame and actuators become

excessive. While it is possible to build larger

reaction and loading facilities, structural test demands can always exceed capacity.

A special case requiring the test of a

large and strong structure arose following the

1989 Loma Prieta earthquake. It was desirable

to test a portion of the 1-880 Cypress viaduct

that partially collapsed in Oakland. This two

level reinforced concrete structure utilized

multi-celled box girders to support the upper

and lower level roadways, and a two level,

single span beam-column framework to support

the box girders and to provide lateral resistance

in the transverse direction. Tests were desired

to assess the precise causes of the observed

collapse of the upper level of the structure and

to determine the efficacy of various retrofit

schemes suggested for other double deck

viaducts damaged in the San Francisco bay

area. Some of the methods considered for

loading the viaduct are schematically illustrated

in Fig. 1. These included the attachment of

post-tensioning tendons to the deck levels and

using post-tensioning jacks to apply the lateral

loads. Because cyclic loading was desired,

tendons would have to be installed on both

sides of the viaduct. Problems associated with

this scheme relate to the need for piles or other

types of (costly) foundations to transmit the

tendon force to the ground, and the need to anchor the tendons relatively far from the

structure in order to minimize the vertical

component of the tendon force on the structure. The congested urban setting of this structure precluded this simple approach.

82
Another approach considered the

construction of a reaction frame as shown in the

center portion of Fig. 1. This would allow the use of double acting electro-hydraulic actuators. However, the tests were estimated to require

upwards of 4 million pounds of lateral force.

The design of the foundations for the external

reaction frame in this case became very

difficult. In particular, uplift forces were very

large because of the desire to minimize the

width of the reaction frame. Because of the

poor soil at the site, the reaction frame scheme

was considered prohibitively expensive.

The strategy eventually selected for

these tests is shown on the right hand side of

Fig. 1. Because the failure occurred in the

upper level, it was not viewed as essential to apply lateral loads directly at the lower level.

Similarly, it was not necessary to assess the

contribution of the soil-foundation system to the lateral response. As a result, it was

possible to devise a self-equilibrating steel

reaction frame within the structure. Large steel collars were attached around the base of the

lower columns. Diagonal steel braces extended

upwards from these collars (through holes

punched in the lower deck) and joined just

below the upper deck at a reaction block. Pairs

of simple, single acting (compression) oil jacks were installed on both sides of the reaction

block. These transmitted force to loading

blocks that were post-tensioned to the upper

deck of the viaduct. Braces and oil jacks were

installed in pairs on each side of a frame to preserve symmetry of loading. Three sets of

transverse frames were introduced into the test specimen, resulting in six diagonal braced bents

and a total of twelve hydraulic jacks.

Simple oil jacks were utilized instead of

double acting actuators in this case because of

the time (and expense) required to special order

the large number of relatively high capacity

actuators needed to develop the loads

anticipated in this test. The oil jacks were

easily available from local prestressing supply

companies.

The oil jack arrangement could produce

the cyclic loading required. For specimen

motion to the left, the left jacks would be

extended (with the right jacks retracted). For

motion to the right, the jacks on the right were

extended. Relatively simple, manually operated

hydraulic equipment was effectively employed

to carry out the tests.

In another similar situation, a retrofit

strategy for a telecommunications center was to be evaluated. The existing structure contained a composite concrete/structural steel shear wall.

A structural steel frame was provided in the

existing structure to help carry gravity loads

and seismic overturning forces, and the

concrete infill and encasement was utilized to

resist shear. Retrofit of the structure included

thickening the concrete walls, filling in the

numerous openings located in the existing

walls, adding shear transfer studs to the

existing steel framing to enhance shear transfer

between the steel frame and the new concrete,

and the addition of partial transverse

confinement. Because of the complex stress

transfer mechanism between the steel frame and

the concrete and between the new and old

concrete, it was desired to employ a large scale

test specimen to verify the retrofit concept.

Similarly, it was desirable to include as many

bays and stories in the model as possible to minimize boundary condition effects where the

specimen were attached to actuators and the

reaction floor. It was, of course, not

economically possible to test a model of the

entire 15 story building. Thus, a portion of the

structure was selected incorporating as many of

the salient features as possible.

Several alternatives for specimen

configuration and loading were considered.

Ideally, a model would have been constructed

and attached with double acting actuators to an

existing strong floor and reaction wall. This

approach is shown in the left hand illustration

of Fig. 2. The minimum one-third scale

considered acceptable for these tests resulted in a very large lateral load resistance. This load

(in excess of 3,000 kips) could be developed

by available laboratory actuators; however, the

overturning moment developed resulted in

excessively complex and expensive details for

attaching the specimen to the reaction floor.

Another option considered for loading

the specimen was to utilize a large capacity

(4,000 kips) universal testing machine. In this case, the problem of resisting wall overturning

83
moments could be resolved by doubling the

size (height) of the specimen. In this fashion,

the wall would be tested as a simply supported

beam, with its midspan representing the base of

the wall in the prototype. While this option

solved most of the problems encountered with

the previous approach, it resulted in a specimen

with twice the size, but without any apparent

gain in useful information. In addition, for

cyclic loading, it would be necessary to remove

the specimen from the testing machine and

invert it. Because instrumentation would have

to be repeatedly removed and re-installed each

time the wall was flipped, the time and effort

required for testing made this option

impracticable..

The option eventually employed for this

test was adapted from the testing approach used

for the Cypress viaduct (Fig. 1). A selfequilibrating frame was attached directly to the

wall test specimen. The specimen selected was

three bays wide and two stories tall, and

constructed at one-third scale. Large composite

steel and reinforced concrete load transfer

girders were attached monolithically to the top

and bottom of the test panels. Pairs of steel

diagonal braces were attached externally to the

lower transfer girder on each face of the test specimen. Pairs of single acting oil jacks were

attached to the reaction block at the top of the

braces . These jacks transferred load to each

side of the specimen by means of a reaction

block attached to the upper distribution beam.

Operation of this test under cyclic loading was

similar to that described previously for the

Cypress viaduct.

Another interesting example involves

field testing of a seismically isolated bridge

located in Walnut Creek, California. Since this

was the first totally new bridge to be isolated in

the U.S., it was desirable to assess the ability

of design and analysis assumptions to capture

its dynamic response characteristics, especially

in the nonlinear range. In this case a series of

tests were undertaken at different times during

construction. First, the individual columns

were tested to determine their dynamic

characteristics. This was done by pulling pairs

of columns together using wire rope and a long

stroke hydraulic jack. An explosive bolt, used

to connect segments of the wire rope, was

detonated to release the applied loads. This

procedure proved to be extremely effective in

achieving the quick release vibratory response

of the columns.

Once the isolators and deck were

installed on top of the columns, ambient and

forced vibration tests were carried out. A pair

of reciprocating mass vibration generators were

used to induce forces along the longitudinal and

transverse axes of the deck, as well as torsion

action about a vertical axis. These test produced information regarding the period,

mode shapes and damping characteristics of the

bridge in the elastic range. Because of the

nonperfect symmetry of the bridge, modes

were highly cross coupled. The resulting

modal beating phenomenon complicated

interpretation of results.

While of great importance, the forced

vibration tests provide little insight into the

inelastic response of the system. Because the

base isolation system would permit the

structure to be deformed into the inelastic range

with little permanent damage, a series of quick

release tests were carried out with laterally

imposed displacements approaching the

maximum displacements anticipated during the

design earthquake. In this case, time and

financial constraints prevented the use of

loading apparatus shown in Figure 1. It was

decided to load the bridge along its longitudinal

axis and utilize the abutment or adjacent frame

as the reaction frame. The seismic isolation

system resulted in relatively low loads being

transferred to the abutment. Consequently, the

capacity of the abutment was insufficient to

resist the loads required to load the entire

bridge. As a result, hydraulic actuators were

placed between the isolated section of the

bridge and an adjacent non-isolated cast-inplace reinforced concrete bridge. This is schematically illustrated in Fig. 4.

It was estimated that it would be

necessary to displace the bridge about 6 inches

(15 cm) to allow the deck to spring back to zero permanent displacement following the test. This displacement also corresponded to that

anticipated during the design level earthquake.

Loading was again achieved using hydraulic

actuators. Initially, it was anticipated that the

rapid release of load would be achieved by

installing gas springs on the compression side

84
of the jacks. When the oil pressure used to load the jacks was quickly released, the

pressurized gas on the opposite side of the jack

piston would force the jack to retract quickly.

Such tests have been successfully carried out in

Japan by the Public Works Research Institute.

Unfortunately, the available construction time

schedule did not permit construction and

verification of this system. In stead a simpler

system was developed. This is schematically

shown in Fig. 5. In this case, simple oil jacks were used to push the seismically isolated

portion of the bridge away from the adjacent

cast in place bridge. When the desired

displacement of the isolated bridge was

reached, a displacement restraining wedge was

inserted in the gap between the two bridges and

the oil jacks were removed. The displacement

restraints were diamond shaped with loading

applied along one of the diagonals. Stability was maintained by a large tension bolt installed

along the other diagonal. An explosive charge

was placed in the bolt and detonated to achieve

the quick release. Four sets of these oil jacks

and displacement restrain devices were used at the interface between the two bridges. These

simple tests worked well and provided useful

information on the nonlinear dynamic response

of the isolated bridge. Primary difficulties were

associated with coordinating test and

construction schedules, practical issues of

instrumenting a 700 ft long structure (more than

5 miles (8.3 km) of wire was needed for the

instrumentation), and weather.

CYCLIC LOAD TESTS IN MULTIPLE

DEGREE OF FREEDOM SPECIMENS

In multiple degree of freedom systems,

additional complexity is introduced because of

the need to determine the distribution of forces

or displacements applied at each actuator. As a means of simplifying this problem, it is common to make the imposed loads or displacements a function of parameters

measured at other locations during the test; thus, an invariant relative distribution of

displacements or forces is often imposed.. In the case of inelastic response, it is usually

preferable to control the force distribution, as localized damage to the structure could

substantially alter the structure's deformed

shape (e.g., by the formation of weak stories).

This test procedure can be implemented easily

with conventional electro-hydraulic actuators by

identifying one actuator as the master and

setting the command signal for the other (slave)

actuators to be a proportion of the force

measured in the master actuator. Because of

the possible instability of a structure tested into

the inelastic range using force control only,

most tests of this type would specify the

displacement at the master actuator. The force

in the master actuator would still be used to control the forces in the remaining actuators even though the master actuator is under

displacement control. This arrangement is shown schematically in Fig. 6. When specimens are loaded in two or

three dimensions, other complications arise.

For example, consider the simple floor

diaphragm in Fig. 7. Three actuators are attached to control the in-plane transverse,

longitudinal and rotational motion of the

diaphragm. If only longitudinal motion is required, elongation of only the longitudinal

actuator will not produce the desired result.

This is because of the arching action of the

typically short length actuators. The in-plane

rotation of the transverse actuators due to longitudinal actuator motion causes transverse

motion of the specimen as shown in Fig. 7,

along with transverse forces in the specimen.

In addition, the displacement and forces

measured in the longitudinal actuator will differ

from those developed in the specimen in the

undistorted longitudinal direction. These

undesirable effects must be corrected for by

lengthening of the transverse actuators.

Under displacement control, the

command displacement of an actuator in the

above test setup should thus depend on the

command displacement of the other actuators. The geometric corrections will depend on the

test setup, and are usually straight foreword to derive. However, they have in general a nonlinear form which is difficult to implement

by simple analog manipulation of the control

signals.

This nonlinearity does not introduce any

special complication when the specimen will be

subjected to predetermined displacement

histories. Appropriate geometric manipulation

of the displacement command signals can be

85
made by digital computer (e.g., using a simple

spreadsheet program) prior to the test.

Force readings in actuator load cells,

must be processed following a test to correct

for the rotation of the actuators relative to the

initial coordinate system of the specimen (see

Fig. 8). The measured (or commanded)

displacements are used (along with information on the actuator and structural configuration) to compute the forces along the structure's

principal axes. At large displacements (and/or

short actuator lengths) an actuator can induce

significant forces along both of the principal axes of the specimen (Px and Py in the figure)

and correction of the measured load cell force

becomes essential.

An example of such a test is that shown

in Fig. 9. The prototype bridge structure

supports an elevated roadway and an extended

"outrigger" beam is required to contribute to the

transverse load resisting system through

bending, shear and axial load. In the

longitudinal direction of the roadway, the beam

is also required to develop torsion when

contributing to the lateral load resisting system.

Testing of this outrigger was carried out

considering horizontal motions of the deck in

the transverse and longitudinal directions. The

simplified test specimen used is shown in Fig.

9. A "cloverleaf" displacement pattern was

specified at the tip of the column in the test specimen which results in simultaneous

displacements in both principal directions. The

ideal displacements were modified prior to the

test to obtain command displacements corrected

for geometric effects. A computer based

function generator is used to send the corrected

command signals (as voltages) to conventional

analog servo-controllers.

An interesting feature of this test is that,

in the transverse direction, the axial load in the

column depends not only on initial gravity

loads, but also varies due to overturning

moment effects as a function of the transverse

load applied. Simple considerations of statics were used to determine the appropriate axial

load as a fixed fraction of the applied transverse

lateral load (plus the initial gravity loading).

This was implemented in the tests using simple

analog summing amplifiers and analog

circuitry.

This situation becomes more complex,

if mixed force and displacement modes of

actuator control are used. While the

displacements in master actuators can still be

pre-computed based on the desired specimen

displacements at these locations, the corrections

for the forces must be performed on-line during

the test so that they can be used to specify the

forces to be imposed at other levels. The

transformation functions relating forces

measured in the load cell and along the

specimen's principal axes can be derived prior

to the test, but they will change throughout the

test as the displacements vary. More

significantly, the displacements at the degrees

of freedom under force control will not be

known prior to the test, so measured

displacements must be used during the test to determine the appropriate transformations

between the desired force in a particular

direction and the force measured in the load

cell.

For reasonably small displacements and

simple test configurations, it may be possible to

utilize analog techniques to modify the loads to be applied to actuators in three dimensional

tests. In other cases, the required

transformations may be come sufficiently

complex, and the displacements sufficiently

large that digital procedures are necessary (or

preferred). Relatively simple and low cost

micro-computer based, control processors are available for single and multiple axis control.

This may be implemented in such a manner that

the digital controller makes the various

transformations and carries out the appropriate

electronic control functions for the servocontrol system. It can also be implemented in a more conventional analog control situation. In

this case, the micro-processor repeatedly reads

the displacements and forces at the actuators,

makes the appropriate transformations and

sends the command displacements or forces to conventional analog controllers.

PROOF TEST OF A DOUBLE LEVEL

VIADUCT

An example of a test in which such

digital control is used is shown if Fig. 10. In
this case the lateral load resistance of a double

deck viaduct was investigated. Earthquake-like

motion in each principal plan direction was

imposed quasi-statically using hydraulic

actuators attached to rigid reaction frames posttensioned to the laboratory strong floor. The

columns in the prototype structure were assumed in design to be pinned at the base of

the lower column and at the tip of the upper

column. Considerations of asymmetry under

lateral loading were used to isolate the test specimen from the remainder of the structure along the midspan of the deck in each direction.

Roller supports were desired at these locations;

they were implemented using vertical pin ended

steel struts because of the significant uplift

forces that could be developed.

A total of 10 hydraulic actuators were employed in the test. Lateral loads were

applied to the top of the upper column and to the lower deck. Gravity loads were simulated

in this one-third scale model using a combination of concrete and lead weights as well as vertically oriented hydraulic actuators. Because the finite length of the vertical support

struts resulted in small vertical motions of the

support points under lateral displacements, a specially designed vertical actuator was

provided under the column to maintain the axial

load in the column at a proper level. This

actuator also compensated for inelastic

elongation of the plastic hinge region in the

lower column, and overturning moment effects

due to transverse loading. A post-tensioning

rod also extended up through the center of the

column and was loaded with a hydraulic

actuator at its top in order to maintain the

column axial load at the proper level.

Because of the desire to simulate

realistically the distribution of gravity moment

and shear in the transverse bent cap, two

vertical actuators were used to load the bent cap

at the interior webs of the box girder used for

the bridge deck. In addition, to replicate the

positive moment at the midspan of the bent cap

due to dead load, a steel outrigger was

cantilevered from the edge of the specimen and

a constant vertical upward force was applied

during the test.

At the lower deck level two actuators were used to impose specified displacements in

the transverse direction and to limit torsion. A

single displacement controlled actuator was

used in the longitudinal direction at the deck

level. Two actuators were connected in a horizontal plane to the top of the column. The

forces in these actuators were controlled rather

than displacement as done at the lower level. If displacements were controlled at both levels,

changes in deflected shape (that might occur as a result of concentrated yielding in one of the

columns) could not be detected. In this case,

the force in the upper level actuators was based on an assumed triangular mode shape, the ideal

mass distribution at the two levels, and the

column heights. For the test specimen, the

force at the top level was 1.25 times the force

applied at the lower level in the same direction.

This actual force in the actuators were

computed on-line during the test.

Because the actuators at the upper level were rotated with respect to the principal axes

of the structure, a geometric correction was also

applied. Since the top of the column displaces

during the test, the geometric correction

depended on the displaced configuration of the

specimen at each step in the test. Furthermore,

the forces measured in the load cells of the

actuators attached to the lower deck were not

oriented in the principal direction of the deck

once the structure was subjected to large bidirectional motions. Thus, these forces were

also corrected prior to being used to determine

the lateral forces to be imposed at the top of the

columns. The horizontal components of the

forces imposed by the initially vertical actuators were also computed and used in the derivation

of the forces to be applied at the upper level. A dedicated high speed PC-based

microprocessor was used to make these

coordinate transformations and to control the

specimen during the test. The microprocessor

required continual input of the measured deck

level actuator loads, the measured column top

and deck level displacements. The control

process was repeated approximately 7000 times

per second. The microprocessor provided the

actuator control apparatus with analog output

signals for controlling the force at the top of the

column.

87
A wide variety of loading histories were

considered for the test. These included simple

diagonal load paths, cloverleaf patterns and

other sequences that would help identify the

mechanical characteristics of the structure. The

loading history selected is shown in Fig. 11.

Each step of loading was divided into

three phases. In the first phase, the structure was subjected to a simple displacement cycle in

the transverse direction (longitudinal

displacement at the lower deck is restrained

during this excursion) followed by a simple

cyclic excursion in the longitudinal direction.

The second phase involved two cycles of bidirectional loading in a square pattern; one

clockwise and one counter clockwise. The

maximum amplitude of the bi-directional

excursion was the same as that for the unidirectional excursions; i.e., the projected

displacement in the transverse direction was

reduced to about 70% of that imposed during

the unidirectional excursions. The third phase

repeated the first phase so that deterioration of

mechanical properties may be easily detected.

This loading history was found to be very

effective in assessing the effect of bidirectional

loading histories. In more recent tests, the

square pattern has been replaced with a circular

pattern. This appears to be more realistic and

not to have as severe effects on the specimen as

the square pattern.

The displacement history shown relates

only to the lower level. The forces applied to the top of the column corresponded to those

associated with the triangular mode shape and

the forces required at the lower level to develop

the specified displacements. Because of the

complexity of bi-directional response, the upper

column moved bi-directionally even under unidirectional displacements at the lower level. PSEUDO-DYNAMIC TESTING

These on-line computer controlled test procedures can be extended to include hybrid

analysis. The most common form of this type

of testing in earthquake engineering is the so- called pseudo-dynamic test. In most current

applications, the actuators are under

displacement control. The control computer, in

addition to making the various required

geometric and other corrections, uses data

obtained regarding the restoring force

characteristics of the specimen at a particular

instant in time to solve the governing

differential equations of motion of the test specimen for a numerically specified ground

motion. In this manner the dynamic response

of the specimen can be simulated quasistatically using conventional electro-hydraulic

loading systems. Dynamic effects are simulated in the computer by the governing

differential equations, and the nonlinear

restoring forces are determined 'continuously'

from the specimen during the test.

Detailed discussion of these procedures,

and their advantages and limitations, are beyond the scope of this paper. More complete

description of these procedures may be found

citations listed in references by Thewait (1991a

and b).

Important extensions to the pseudodynamic test method include:

O Real-time or near real-time testing of

structures exhibiting rate sensitivity, such

as the seismic response of structures

containing visco-elastic dampers.

O Development of new "implicit" algorithms

that allow the use of time integration

intervals that do not depend on the shortest

natural period of the entire system. These

algorithms will substantially accelerate

testing of large complex systems with

numerous degrees of freedom, and will

improve accuracy by reducing numerical

error propagation effects.

O Development of reliable methods for online diagnosis of the accuracy of pseudodynamic test results. Assessing the

accuracy of pseudo-dynamic test results is difficult, and reliable methods to identify

and eliminate spurious results are required.

Issues related to identifying deleterious

effects associated with stress relaxation

and strain rate effects for yielding

structures need especially to be addressed.

Methods for certifying pseudo-dynamic

test results or facilities should be

developed and adopted.

88
O Development of new algorithms are needed

that can exploit the abilities of digital

(rather than analog) controllers. For

instance, it is theoretically possible to develop test procedures in which the

specimen's motion is continuous. That is, the need to repeatedly stop for the

measurements and calculations to take

place at each step would be eliminated.

This feature would make the tests more

realistic, and significantly reduce error propagation effects due to stress relaxation

in the inelastic range.

O A key feature of pseudo-dynamic testing is the ability to test a part of the structural

system while analytically representing the

remaining portions. Substantial additional

work needs to be carried out to explore

fully the limitations and capabilities of this

substructuring approach.

The deformations at rotational

degrees of freedom at the intersection

of the analytical and experimental

substructures are difficult to predict

and control. Such difficulties should

be explored and refined solutions

developed.

Better methods for assessing the error propagation characteristics of

substructuring approaches needs to be

developed.

1/ Methods for mixed displacement and

force control need to be developed.

For example, if a column (or base

isolator) at the base of a structure is to be tested, the axial load in the

specimen is likely to be an important

parameter. However, it may be very

difficult to control this axial force

using the axial deformation degree of

freedom of the test specimen because

the required accuracy of displacement

control may exceed the ability of most

available hydraulic equipment. In addition, the high stiffness (and

therefore frequencies) associated with

this vertical response mode may likely

result in significant numerical

difficulties. Direct computation and

control of the axial force in the

specimen would be preferable. For

practical and safety reasons, it is desirable to continue to control the

more flexible lateral motion of the

specimen using displacement

command signals. Additional

information on numerical procedures

to implement such mixed

force/displacement algorithms is presented by Thewalt (1991 b).

In large structures, it may be desirable

to test several major components

simultaneously. Such "distributed"

pseudo-dynamic tests may be

desirable in bridge structures where

the bridge deck is expected to remain

essentially elastic, whereas columns,

piers, abutments and restraining or energy dissipation devices may

undergo significant yielding.

Substructuring techniques may be

effectively used in such situations to obviate the need to construct those

large portions of the structure which

under-go well understood behavior.

Details on methods to implement such

procedures are discussed by Thewalt

and Mahin (1994).

O Pseudo-dynamic tests also may have

difficulties where the specimen to be tested

is especially stiff. In such situations, very

small displacements may have to be

controlled. Because of imprecision in

displacement transducers, friction and

other mechanical errors associated with

electro-hydraulic actuators, and so on, such tests are likely to result in spurious

results. In some cases, the use of very

precise instrumentation has produced

satisfactory results. However, in other

cases, dynamic interaction between the

electronic control system and the test specimen can occur. This can lead to unstable tests. Various approaches to this

have been attempted (e.g., detuning the

control system, artificially softening the

system through use of elastomeric bearings

between the actuators and the specimen,

adding virtual mass to the specimen, etc.)

and in some instances these have been

89
successful. However, the ability to specify force as the control parameter,

rather than displacement, would seem to be

desirable in such situations. Algorithms to implement such tests are described by

Thewalt and Mahin (1994).

O While some significant testing has been

done to account for soil-structure

interaction effects, more effort should be

carried out on this topic for different types

of situations and foundations. The ability

of testing structures in the field using

pseudo-dynamic methods would be a significant advance. This would be

particularly useful in the assessment of as- built seismic characteristics of existing

structures and in evaluating the efficacy of

seismic retrofits. The theoretical and

practical issues related to performing

pseudo-dynamic tests where the structure

is founded on soil, rather than on a rigid

laboratory floor, need to be resolved,

however.

Additional experience with pseudodynamic tests is required, especially with

structures subjected to multiple components of

excitation. Special emphasis is required to validate the method through matched shaking

table and pseudo-dynamic tests.

Extension of the pseudo-dynamic test method to non-seismic problems should also be

explored. For example, collision problems (ice

impact on offshore structures, for instance) or dynamic aspects of moving loads (vehicles

moving over a bridge) appear to be logical and

useful extensions of the method.

CONCLUDING REMARKS

A wide variety of loading procedures

are available for assessing the cyclic inelastic

behavior of structures. Usual concerns relate to instrumentation, data acquisition and

load/stroke capacity. In some tests special

loading systems are required to develop and

react the large forces required. In other cases, accurate measurement and control of structural

response requires a variety of geometric and

other corrections. Modern microprocessors

make the execution of such tests relatively

simple, and allow for even more complex

hybrid experimental analysis procedures, such

as the pseudo-dynamic test method.

Often practical considerations place the

main restrictions on laboratory testing. As

testing is applied to larger and more complex

systems, costs of testing increases accordingly.

Efforts are needed to upgrade experimental

facilities and equipment to facilitate the types of

tests described in this paper. Field testing is often difficult, but provides important real

world conditions not possible to simulate in the

laboratory.

ACKNOWLEDGMENTS

The authors are especially indebted to Professors Christopher Thewalt and Jack

Moehle at the University of California at Berkeley who participated significantly in some

of the test programs discussed in this paper.

The tests described in this paper were funded

by the California Department of Transportation,

the National Science Foundation, and EQE

International. The authors gratefully

acknowledge this support. It must be

recognized that the views expressed in this

paper are those of the author alone. They do

not necessarily reflect the policies or opinions

of the of the research sponsors, the National

Science Foundation, or the University of

California at Berkeley.

REFERENCES

1. Thewalt, C. and Mahin, S. " The Pseudodynamic Test Method: Numerical

Aspects," Experimental and

Numerical Methods in Earthquake

Engineering, Ed. Donea, J. and Jones,

P., Kluwer Academic Publishers, Boston,

1991.

2. Thewalt, C. and Mahin, S. " The Pseudodynamic Test Method: Verification and

Extensions," Experimental and

Numerical Methods in Earthquake

Engineering, Ed. Donea, J. and Jones,

P., Kluwer Academic Publishers, Boston,

1991.

90
3. Thewalt, C. and Mahin, S., "An

Unconditionally Stable Hybrid

Pseudodynamic Algorithm," submitted to International Journal of Earthquake

Engineering and Structural

Dynamics, 1994.

Actuators

Tendons

Jack Pairs

Reaction

Frame

Collar

Fig. 1 Alternative Methods Considered for Field Testing of Viaduct

Reaction Frame Specimen

Actuators

V'vW-.

r:x":l:x:'

1:::":t'::':- i

Jack Pairs

Loading

Block

fvV

:•: :•:(•: :•::

Loading

Block

Load

Transfer

Girders

Fig2 Alternative Loading Arrangements Considered for Composite Wall Tests

Wire Rope and Harness

Hydraulic Jack—7 LSwivel

Explosive Bolt

4

ent Cap

Column Extending

into Cast in Drilled

Hole Pile

Fig. 3 Mechanism Used to Achieve Snap Back Vibration Experiment for Bridge Columns

91
Fig. 4 Schematic Drawing of Walnut Creek Seismically Isolated Bridge Showing Placement of

Hydraulic Jacks Used in Quick Release Tests

Hydraulic

Jack

Displacement Restraint Device

Fig 5. Schematic Detail "A" Showing Hydraulic Jack and Displacement Restraint Device Used to Achieve Quick Release Test of Walnut Creek Isolated Bridge

92
xxxxxxxiP30GII1 30a

Hydraulic

Actuators

3na

—y Command Displacement

J j——Measured

Force

Fraction of

Measured

Force

Specimen Displacement

Fig. 6 Typical Planar Test of Multi-degree of Freedom System

<xxxxxxx

TXX

:x xx Change in Actuator Length = L

Plan

Deformed Position

Initial Position

T

L - AL : Actuator Lengthening Differs from Structural Displacement

Transverse Movement

-without Change in / Length of Transverse \& Actuators

Fig. 7 Interaction of Actuator Displacements in Three Dimensional Tests

Fig. 8 Differences Between Forces Measured by Load Cell and

Those Applied Along Specimen's Principal Axes

93
Fig. 9 Bi-directional testing of an Outrigger Bent Cap

1.0 8t

5l

i

->

r Pattern A1 ,C 1 Pattern A2, C2 Pattern B Pattern B2

1

Fig. 10 Displacement History for Proof Test Specimen

94
72 k

Simulation of dead loads

Fig. 1 1 Test Setup for Proof Test of Double Deck Viaduct

95

Standard Test Procedures for Seismic Isolation Systems

by

Harry W. Shenton, III*

ABSTRACT

The Building and Fire Research Laboratory of

the National Institute of Standards and Technology is currently engaged in an effort to develop

guidelines for testing of seismic isolation sys- tems. A comprehensive set of draft guidelines

for testing has been developed and are available

for use. The guidelines were developed to be

independent of the type of superstructure, and

generally independent of the type of isolation

system. Three classes of tests are addressed in the guidelines: pre-qualification, prototype and

quality control testing. The final guidelines for

testing will be developed from the draft guidelines, based on industry feedback of the draft guidelines, and a test program to assess and

evaluate the draft guidelines.

KEYWORDS: base isolation; building code;

experimental; guidelines; seismic isolation; standard; testing.

1 . INTRODUCTION

Seismic isolation is now generally accepted as a proven technology in earthquake hazard mitigation. A testimony to this fact is the increasing

number of structures to be isolated in the United

States and around the world (Kelly, 1993). One

factor that has most certainly fostered confidence

in the technology is the heavy reliance placed on

testing to verify the performance and quality of

manufacture of the isolation system.

Testing is currently required by the two major

building codes that have adopted guidelines for design of isolated structures: for buildings, the

1991 Uniform Building Code (Uniform, 1991),

and for bridges, the 1991 AASHTO Guide

Specifications for Seismic Isolation Design

{Guide, 1991). And although each requires an

extensive series of tests, there currently do not

exist standards for conducting these tests. As a

result, test requirements are freely interpreted,

leading to variability in the test procedures and

subsequent test results from one supplier to the

next. Furthermore, the lack of standard test procedures precludes an easy comparison of the

performance of different isolation systems.

The Building and Fire Research Laboratory of

the National Institute of Standards and Technology (NIST) is currently engaged in an effort to develop guidelines (i.e., a pre-standard) for

testing of seismic isolation systems. The first phase of this effort is complete, as signified by

the publication of draft guidelines for testing of

seismic isolation systems (Shenton, 1994a;

Shenton, 1994b; Shenton, 1994c). The final guidelines for testing will be developed from the

draft guidelines, based on industry feedback and

a testing program. Presented in the paper is a

brief summary and overview of the draft guidelines and a discussion of the plans for developing the final guidelines for testing.

2. OVERVIEW OF THE DRAFT

GUIDELINES

2.1 Scope

Rather than focus on a particular isolation sys- tem and application, e.g., seismic isolation of

bridges using elastomeric bearings, the guidelines were developed to be comprehensive and

broad in scope. The guidelines were developed

to be independent of the type of superstructure,

and therefore, are applicable to a broad range of

projects, e.g., buildings, bridges, water towers,

Research Structural Engineer

National Institute of Standards and Technology, Gaithersburg, Maryland 20899

97
etc. The guidelines are also generally independent of the type of isolation system, i.e., the test procedures are applicable to elastomeric, sliding, or hybrid isolation systems.

Three basic classes of tests are covered by the

guidelines: pre-qualification, prototype and

quality control testing. These are defined as follows:

Pre-Qualification tests need not be project

specific and are conducted in order to establish the fundamental properties and characteristics of the isolation system, and to determine the extent to which these properties are dependent on load and environmental factors.

Prototype tests are project specific and are conducted to verify the design properties of

the isolation system prior to construction.

Quality Control (QC) tests are project specific and are conducted to verify the quality

of manufacture and as-built properties of the

isolation system prior to installation.

The guidelines are intended for passive systems

that isolate in the horizontal plane only. Although some of the test procedures may be

applicable to active systems, or systems that isolate in the vertical plane, the procedures were

not developed with the latter in mind.

2.2 Rated Capacity

Fundamental to the guidelines is the concept of

rated capacity. The onus is on the supplier of the

isolation system to report certain fundamental

properties of the isolation system, prior to testing. The properties to be reported include

such things as seismic design displacement,

thermal design displacement, effective horizontal

stiffness at the design displacement, and energy

dissipated per cycle at the design displacement.

The complete list of properties to be rated is presented in Table 1.

Conditions under which the tests are conducted

as outlined in the guidelines are based on the

rated capacity of the system. For example, in a

typical cyclic lateral load test, the vertical load

and lateral displacement of the test are based on

the suppliers rated capacity of the isolation sys- tem. In this way the test procedures are independent of the type of superstructure and autono- mous of existing design guidelines and building

codes for isolated structures. The standard list of

rated properties should also facilitate the comparison of isolation systems from different

suppliers and assist the designer in selecting the

most appropriate system for the application.

2.3 General Requirements

The majority of the tests found in the guidelines

are of a compression/shear type, i.e., a constant

vertical load is maintained while the specimen is deformed laterally in shear. Many of the tests are, therefore, very similar in terms of the test set-up and test requirements. Consequently,

general requirements are outlined in the guidelines for a typical compression/shear test, and

exceptions or special requirements are noted in the test description.

General requirements address most of the issues

related to test setup and instrumentation. Minimum requirements of the test facility, instrumentation, instrumentation calibration, data acquisition, and data analysis are discussed. This

includes, for example, the number, type and

positioning of displacement transducers; load

capacity and calibration of the test facility, and

a detailed description of the procedure for

determining the effective stiffness and energy

dissipation from a typical compression/shear

test.

2.4 Pre-qualification Tests

Pre-qualification tests are formally not required

by the codes today; nevertheless, tests of this type are usually conducted in one form or another as a new isolation system is developed.

The purpose of the test series is to establish the
fundamental properties of the system and to characterize the response. The tests are designed

to determine such things as the dependence on

frequency of loading, vertical load, load direction and temperature, to name just a few. Tests

are also outlined for establishing the ultimate or

reserve capacity under load conditions likely to be encountered during an earthquake, and under

normal operating conditions. The pre-qualification series is the most extensive and comprehensive of the three class of tests: pre-qualification

tests are to be conducted only once, for any

given system of a particular design, material and

construction. Conceivably, in the future, a supplier may be required to submit the results of

their pre-qualification test series as part of the

bid package in order to be considered for a job.

Briefly, a complete series of pre-qualification

tests is to include all tests listed in Table 2, a

full series of prototype tests (Table 3) and

applicable quality control tests (Table 4). Each

pre-qualification test is to be performed separately on two specimens. Whenever possible, tests are to be conducted on full scale specimens;

however, recognizing the limitations of existing

test facilities, scale model specimens are acceptable, provided they are not less than 1/4 full scale and are representative of the full scale

prototype.

Performance criteria, or a framework for specifying the criteria in the final guidelines, have

been established for all test in the draft guidelines. Some of the criterion for the pre-qualification series are simply "benchmarks" for classifying the response of the isolation system. Others

are measures of performance and the quality of

manufacture: systems that do not meet or exceed

these criteria may not perform adequately in service.

An example test description from the pre-qualification series is presented in Figure 1

.

2.5 Prototype Tests

The prototype test series is intended to verify the

principal design properties of the isolation

system, namely, effective stiffness and energy

dissipation. Other tests are included to verify reserve capacity or satisfactory performance

under simulated seismic and non-seismic loads.

The prototype series outlined in the guidelines is similar to, and in part based on, the test requirements given in the 1991 UBC {Uniform, 1991)

and the 1991 AASHTO Guide {Guide, 1991).

A complete series of prototype tests is to include

all tests listed in Table 3, and applicable quality

control tests listed in Table 4. Each prototype

test is to be performed on two full scale speci- mens. Properly documented prototype tests previously conducted on a unit of similar size

may be used to satisfy the requirements of

prototype testing, provided:

• the unit to be tested is similar in design,

materials and construction;

• the largest dimension of the unit to be tested

is within ±10% of the same dimension of

the unit previously tested;

• most other critical dimensions are within

+ 15% of the size previously tested.

The latter is intended to provide relief from

testing units that are identical in all respects,

from one project to the next.

The most important test of the prototype series

is the test to determine effective stiffness and

energy dissipation. The basic test requires three

fully reversed cycles at +0.25D, +0.50D,

+ 0.75Z) and + 1.0D, at the design temperature

and under the design vertical load, where D is the design lateral displacement. The full extent of testing, however, is in part based on the out- come of the pre-qualification test series: systems

that are found to be dependent on vertical load,

frequency of load, bilateral load or temperature

are required to undergo additional tests, over and above the basic sequence described above.

2.6 Quality Control Tests

Quality control is an essential and integral

element of any manufacturing process. It in- cludes, among other things, proper documenta99
tion, material traceability, manufacturing instruction, inspection, part identification and testing.

The quality control tests described in the guidelines are intended to be included as part of the

supplier's overall quality control/quality assur- ance program, and are not to be construed as a comprehensive quality control program.

Two types of QC tests are discussed in the

guidelines, production tests and completed unit

tests. Production tests are conducted during

fabrication, on the materials or parts that go into fabrication of a unit or device. Tests are then

conducted on completed units, generally to verify the properties and manufacturing consistencies of the production units.

Pre-qualification and prototype tests can and

should be independent of the type of isolation

system; however, because of the nature of

production tests and some of the completed unit

tests, QC tests tend to be system specific. For

this reason, the QC tests are contained in separate reports (Shenton, 1994b; Shenton, 1994c).

Presented below is a brief overview of recom- mended QC tests for elastomeric and sliding

isolation systems.

2.6.1 Elastomeric Systems

The guidelines outline a number of production

tests for elastomeric systems. These tests are to be conducted on representative samples of the

elastomer used in fabrication of the isolation

unit. The recommended tests include hardness,

tensile strength and elongation at break, bond

strength, compression set, low temperature

properties, high temperature aging and ozone

resistance. A performance criterion is established

for each production test that is based on a design

specified value. Materials that do not meet or exceed these criteria should not be used in fabrication of isolation units.

Three completed unit tests are outlined for

elastomeric systems: sustained compression,

compression stiffness, and effective stiffness and

energy dissipation (Table 4a). The sustained

compression test was developed years ago as a means of testing the strength of the elastomer

and steel bond. In this test the unit is subject to a vertical load equal to 1.5 times the nominal

vertical load capacity, for a duration of not less than twelve hours. An exception is provided in the guidelines to reduce the time to three hours

if a minimum number of consecutive tests have

been conducted without failure.

It should be noted here that there is considerable

debate within the community regarding the scope

of quality control testing required. Since QC

tests can be very expensive and time consuming,

the debate centers around the number of completed unit tests to be conducted as part of the

QC program. One position advocates testing

every unit that comes out of the plant, while

another would permit testing a percentage of the

units produced, provided the results proved

satisfactory based on a specified performance or acceptance criterion. Both options are communicated in the draft guidelines in a special "draft option" box. This is one example of an issue

that must be, and will be resolved in the process

of developing the final "Guidelines for Testing

2.6.2 Sliding Systems

A series of production tests are outlined in the

guidelines for a generic sliding device. The

recommended tests include surface roughness,

trueness of surface, interface material properties,

backing material properties, bearing pad attach- ment and sliding interface attachment. Again,

performance criteria are outlined that are based on a design specified value: materials that do not meet or exceed the minimum values should not be used in the fabrication of units or devices.

Two completed unit tests are included for sliding

systems: sustained compression and effective

stiffness and energy dissipation (Table 4b). The

sustained compression test is similar to the

sustained compression test for elastomeric sys- tems, but is intended only for systems that are susceptible to creep.

100
3. PLANS FOR DEVELOPING THE FINAL

GUIDELINES FOR TESTING

The draft guidelines are extremely comprehensive and fairly complete, nevertheless, they are

still considered draft and are likely to undergo

modest revision before the final "Guidelines for

Testing..." are published. Over the next twelve

to eighteen months efforts will be focused on

developing the final guidelines for testing. The

final guidelines are to be based on industry

review of the draft guidelines, and a testing

program. These steps will provide important

feedback for developing the final guidelines.

The industry review of the draft guidelines will be carried out on several fronts. The draft guidelines are to be distributed to various indi- viduals and organizations over the next several

months. Individuals will be encouraged to provide written comments on the guidelines, and to provide reference to evidence or documentation

that supports any suggested changes. The draft guidelines will also be submitted to various

working groups or committees of selected code

writing organizations for formal review and

comment. Finally, a one day workshop is scheduled to be held that will provide a forum for review and discussion of the draft guidelines.

The workshop will bring together individuals

from private industry, research and government,

with experience and interest in the design,

fabrication and construction of seismic isolation

systems and isolated structures. Participants will be charged with addressing some of the more

important and unresolved issues related to testing. Feedback from these different avenues

of review will be taken under consideration as the draft guidelines are revised and brought into

final form.

A limited test program is to be undertaken at NIST over the next year, for the purpose of

assessing and evaluating the procedures outlined

in the draft guidelines. A selected number of

isolation units will be tested that are representative of hardware available today. Tests will be

conducted in strict accordance with the draft guidelines, with the objective of uncovering

problems or inconsistencies in the procedures.

Note that although different isolation systems

will be tested, the purpose here is not to conduct

a rigorous study to compare the relative perfor- mance of various isolation systems.

In a related effort, HITEC, the Highway Innovative Technology Evaluation Center of the Civil

Engineering Research Foundation (CERF) is presently collaborating with the California

Department of Transportation, and the Federal

Highway Administration, to test and evaluate

market ready seismic isolation systems for

bridges. Organizers of the HITEC program are currently reviewing the NIST draft guidelines; in

all probability the draft guidelines will be adopted for use in the HITEC test program. The

results of the HITEC program should be ex- tremely valuable, and will provide additional

input in the effort to develop the final guidelines

for testing.

4. SUMMARY

The Building and Fire Research Laboratory of

the National Institute of Standards and Technology is presently engaged in an effort to develop

guidelines for testing seismic isolation systems,

i.e., a precursor to a standard for testing. The

first phase of the effort is complete, as signified

by the publishing of draft guidelines for testing.

The draft guidelines are extremely comprehensive and broad in scope. The guidelines are independent of the type of superstructure and

generally independent of the type of isolation

system. Therefore, the guidelines are applicable

to seismic isolation of buildings, bridges, etc. Three classes of tests are covered by the guidelines, pre-qualification, prototype and quality

control testing. The pre-qualification series is the

first and most extensive of the three, but would

only be required once for any system of a given

design, material and construction.

The effort to develop the final guidelines will be

carried out over the next year; however, the

draft guidelines are sufficiently complete that they can be used effectively in their present

form. The effort to develop the final guidelines

101
Uniform Building Code (1991), International

Conference of Building Officials, Whittier,

Calif.

5. ACKNOWLEDGEMENTS

The draft guidelines were developed with the

assistance of an Oversight Committee of experts

from the field of seismic isolation, this included:

Dr. Ian Buckle, Dr. Ron Mayes, Dr. Charles

Kircher, Dr. Victor Zayas and Professor James

M. Kelly. The author gratefully acknowledges

the committees assistance and support in this effort.

6. REFERENCES

Guide Specifications for Seismic Isolation Design

(1991), American Association of State Highway

and Transportation Officials, Washington, D.C.

Kelly, J.M., (1993), "State-of-the-Art and State- of-the Practice in Base Isolation", Proceedings

of ATC 17-1 Seminar on Seismic Isolation,

Passive Energy Dissipation, and Active Control,

Applied Technology Council, Redwood City,

California.

Shenton III, H.W. (1994a), Draft Guidelines

for Pre-Qualification and Prototype Testing of

Seismic Isolation Systems, NISTIR 5359,

National Institute of Standards and Technology,

Gaithersburg, Maryland.

Shenton III, H.W. (1994b), Draft Guidelines

for Quality Control Testing of Elastomeric

Seismic Isolation Systems, NISTIR 5345,

National Institute of Standards and Technology,

Gaithersburg, Maryland.

Shenton III, H.W. (1994c), Draft Guidelines

for Quality Control Testing of Sliding Seismic

Isolation Systems, NISTIR 5371, National

Institute of Standards and Technology,

Gaithersburg, Maryland.

is to be based on industry review of the draft guidelines and a testing program. The adoption

of these guidelines for testing should facilitate and further promote the use of seismic isolation as a viable means for earthquake hazard mitigation.

102
Table 1. Standard Rated Capacity List

Parameter Notation Description

Stiffness:

xiorizunidi IS diccuvc nunzunidi silliness di uic ucsign

Displacement and Design Vertical Load.

nUllZUUldl UIIUCI V> 111U CIlCLllVC I1UI IZUIlldl J>lllLIlCdi> dl UIC L/CMgll W 1I1U Load and Design Vertical Load.

Vertical Effective vertical stiffness at the Design Vertical

Load.

Energy Dissipation EH Energy dissipated per cycle at the Design Displacement and Design Vertical Load.

Lateral Deformation:

Design Displacement D Nominal displacement capacity, including that re- sulting from torsion.

Draft Option

corresponding to a level of ground motion that has

a 10 percent probability of being exceeded in a 50

year period.

Maximum Displacement Dm Total maximum displacement capacity, including

that resulting from torsion,

Draft Option

corresponding to a level of ground motion that has

a 10 percent probability of being exceeded in a 100

year period.

Thermal Displacement D, Nominal thermal displacement capacity.

Vertical Deformation:

Design Displacement Dy Nominal vertical displacement under the Design

Vertical Load.

Crppn V— 1 Di^nlarpmpnt L^S 1 . 1 1 ' 1 CI C* 1 1 1 ^- 1 1

1

n Creep displacement under the Design Vertical

Load.

Rotation B Nominal rotation capacity about an axis in the horizontal plane, and perpendicular to the direction of

lateral loading under the Design Vertical load.

Compression:

Low Lower limit of load range of satisfactory seismic

performance, includes the effect of vertical ground

motion and overturning.

103
Table 1. (cont'd) Standard Rated Capacity List

Parameter Notation Description

Design Vertical Load PnD Nominal capacity in compression for dead and live load.

High Pu Upper limit of load range of satisfactory seismic

performance, includes the effect of vertical ground

motion and overturning.

Tension Pr Nominal capacity in tension.

Lateral Load:

Wind Fw Nominal wind load capacity.

Braking/Centrifugal load Nominal braking/centrifugal load capacity.

Degradation Cycle Limit K Number of cycles to ±D with a vertical load of PD corresponding to a +15% change in Effective Stiff- ness, or a ±30% change in Energy Dissipation

relative to the first complete cycle Effective Stiff- ness or Energy Dissipation, respectively.

Thermal Cycle Limit N, Number of cycles to ±D, with a vertical load of PD corresponding to a +15% change in Effective Stiff- ness, or a +30% change in Energy Dissipation

relative to the first complete cycle Effective Stiff- ness or Energy Dissipation, respectively.

Temperature:

Low TL Lower limit of operating temperature.

Design TD Nominal design temperature.

High Ty Upper limit of operating temperature.

104
Table 2. Schedule of Pre-Qualification Tests

1

Test Titlp/Purnnsp

1.1 Fstahlish dfnpndpnop on virgin loading

T 2 L/oLdUl F^tahli^h loll Hpnpndpnrp UvUtllUClltC onUll frennpnfv llCUUCIILy ofVJ1 load IVJdVJ

T 3 i_*oiciL7i P^t^hli^h loii rlpnpnrlpnpp ucijtiiuviiuw onwii load iuau rvrle L-y^it/ hi^torv liioivji y

1.4 F<\tahl Kh dependence on load cvclincr

1.5 Establish dependence on vertical load

1.6 Establish dependence on load direction

I 1.7 Establish dependence on load plane rotation

1.8 Establish dependence on bilateral load

1.9 Establish dependence on temperature

1. 10 Establish dependence on creep

1. 11 Establish dependence on aging

II. 1 Ultimate compression under zero lateral load

II.

2

Compression in displaced position

II II.

3

Ultimate Tension under zero lateral load

II.

4

Tension in displaced position

II.

5

Lateral load and displacement capacity under design vertical load

'Pre-qualification shall also include a complete series of prototype tests (Table 3) and quality control

tests (Table 4).

105
Table 3. Schedule of Prototype Tests

1

Category Test Title/Purpose

III

III. 1 Effective Stiffness and Energy Dissipation

III.2 Stability against degradation

III.

3

Stability at Maximum Lateral Displacement

IV

IV. 1 Wind load

IV.

2

Thermal displacement

IV.

3

Stability with thermal cycling

IV.4 Braking/Centrifugal force

'Prototype shall also include a complete series of quality control tests (Table 4).

Table 4. Schedule of Quality Control Tests

(a.) QC Tests for Elastomeric Systems

Test Title/Purpose

1 Sustained Compression

2 Compression Stiffness

3 Effective Stiffness and Energy Dissipation

(b.) QC Tests for Sliding Systems

Test Title/Purpose

1 Sustained Compression

2 Effective Stiffness and Energy Dissipation

106
Test Designation:

Purpose:

Sequence:

Procedure:

Criteria:

1.5 Establish dependence on vertical load.

Three fully reversed cycles to peak displacements of ±D. Tests shall be

conducted for vertical loads corresponding to PL , PD, Pv

. The frequency of

loading shall be not less than fL or 0.004 cyc/sec.

Place the specimen in the test machine and secure to the supports and loading

plate. Apply the full vertical load to the specimen and allow the load to stabilize. Apply the cyclic lateral load to the specimen for the required 3 fully reversed

cycles of the test. Remove the vertical load. The test shall be run continuously

without pause between cycles. The test shall be conducted at the vertical loads

specified in the order PL, PD and Pv. Sufficient time shall be allowed between

tests at the different vertical loads to dissipate any heat developed during the

previous test. The System, Unit or Component response is considered to be independent of

vertical load if:

(1.) the Average Effective Stiffnesses measured at vertical loads corresponding

to PL and Pv are within ±a% of the Average Effective Stiffness measured at the

vertical load corresponding to PD , i.e.,

< 0.01a

where KHis the reference Average Effective Stiffness measured at a vertical

load corresponding to PD, and Kf, denotes the Average Effective Stiffness

measured at vertical loads corresponding to PL and Pv

.

(2.) the Average Energy Dissipation measured at vertical loads corresponding

to PL and Pv are within ±/3% of the Average Energy Dissipation measured at the vertical load corresponding to PD, i.e.,

< 0.01/5

where Eu

is the reference Average Energy Dissipation measured at a vertical

load corresponding to PD, and E

1

,, denotes the Average Energy Dissipation

measured at vertical loads corresponding to PL and Pv

.

Figure 1. Example Test Description (Shenton, 1994a; see original reference for details and symbol

definitions)

107

State of the Arts in Japan Concerned With Application

of Carbon Fiber for Rehabilitation and Strengthening

for Concrete Bridges

by

Nishikawa Kazuhiro", and Koga Masajiro2'

ABSTRACT

Currently, some rehabilitation

and strengthening methods (retrofitting methods) are employed for con- crete structures. But these methods

have many problems such as reliability, weight increase and existing

member deterioration. Recently,

Carbon Fiber Sheet Glueing Method or Carbon Fiber Strand Winding Method

are employed, because of easy execution, no weight increase and no deteriorate influence to existing members. In this report, fundamental

properties of carbon fiber products

and verification tests for application to concrete structures are briefly described.

KEYWORDS: Retrofitting,

Seismic retrofitting,

Carbon fiber, CFRP Strand,

CFRP Sheet

REQUIREMENT FOR REHABILITATION AND

STRENGTHENING OF CONCRETE STRUCTURES

1. Current Situations On Concrete

Structures In JAPAN

Current situations on concrete

structures, which were constructed in

1950 's as infrastructures in Japan

with careful attensions toward the

concrete material selection and the

construction management, is faced to

rehabilitate oneself because of a shortage of serviceability instead

of a life-span of used materials.

Concrete structures, constructed in 1970 's, are heavily damaged

like surface cracks of concrete and

rusts of reinforcing steel by the reason that we could not help from

employment of undesireble materials

such as alkali-active aggregate and

sea-sand, because of the shortage of

good materials for concrete under

construction fevers in Japan's drastically expanded economy.

Concrete structures came after

expansion of basic infrastructure,

which were constructed in severe cir

cumstances, for example under salty

atmosphere in seaside or under attack of de-icing agent in cold district, are now facing with the area- decreases or breakes of reinforcing

steels because of own corrosion.

1) Head, Bridge Division, Structure

and Bridge Department, Public Works

Research Institute, Ministry of Construction,

Tsukuba Science City, JAPAN

2) Secretary of Carbon Fiber Retrofitting System's Technical Research

Association,

Chiyodaku, Tokyo, JAPAN

109
2. Increase Of Design Load

Concrete structures, constructed in 1980's, are not allowed to in- crease working load additionally,

because these were desired to design

a smart structures by the economic

reason, and which was reallized by

the development of design technique.

From end of 1980's to beginning of 1990's, it was realized that

the bearing forces of members in the

structures were not enough against

the increase of traffic capacity and

load, and sometimes occured the damages of it. In 1993, the design

load for bridges has changed from 20

Tons to 25 Tons.

By the reason of above mentioned, the rehabilitation and strengthening of concrete structures is ur- gently required, today.

3. Outline Of Rehabilitation And

Strengthening Methods

For Concrete Structures

Currently, some rehabilitation

and strengthening methods (shortly

named in this report; retrofitting

methods) are employed for concrete

structures, for example;

(D Repair of members by resin based

cohesive-aid injection

(2) Retrofitting of members by con- crete placing with reinforcement;

namely, concrete thickness increasing method

® Retrofitting of members by steel

plate glued; namely, steel jacketing method.

But these methods have many

problems such as executive reliabilities, increase of member weight and

deterioration to existing member.

Recently,

® Carbon-Fiber-Sheet Glueing Method

or Carbon-Fiber-Strand Winding Method; namely, Carbon-Fiber method,

has been employed, because that this

method has easy executive process, no weight increase, no deteriorate influence to existing members and high

retrofitting effect.

The comparison between method

(2), (D, (g) is shown in Table 1.

Table 1 Comparison between Retrofitting methods

Items (D Carbon Fiber (3) Steel Jacket (2) Concrete

Method Method Placing Meth.

I Structure

1. Weight Increase non increase heavy

2 . Influence to non drilling for surfaceexist. Structure Anchor Chipping

3. Durability long term short term long term

II Construction

1. Applicable Place narrow & wide flat place wide

2. Workability simple & fast complex long curing

3. Work Circumstance noiseless loud loud

III Liability Problems

surfacing. injection- combinedresin cohesion aid effect ability

crush resist.

IV Economy; Cost

1 . Material high high low

2 . Construction low high high

3 . Maintenance low high low

110
CARBON FIBERS FOR RETROFITTING

1. Carbon Fiber Production

High Performance Carbon Fiber

(shortly named; H.P.C.F.) is manu- factured by following methods;

CD Chemical made from continued

organic compound "Molecule-Fiber"

(2) Chemical made from "Pitch"

(3) Chemical made by "Chemical

Vapour Deposition Method"

Highly continued Organic moleculefiber, which is mainly used as a raw materials for HPCF. under industrial production, is Rayon or PolyAcrylic-Nitrile (PAN) . The raw ma- terial is carbonized under stretched

and stabilized condition to make a carbon fiber structure having selectional fiber alignment. Through

these procedure, HPCF. is manufactured, which has anisotropy, high density, high tensile strength and high

modulus characteristics.

Pitch, which is used as a raw materials for HPCF. under industrial

production, is petroleum pitch or liquified coke pitch. Manufacturing

procedure for HPCF. made from pitch

is same as that from PAN, but "Mesophase-Pitch" is used to make easily

a carbon fiber structure having se- lectional fiber alignment.

HPCF. , as above mentioned, has

superior mechanical and chemical

characteristics than that of steel

and in same time has lightweight, no magnetic, no rust, no corrosion and

high durability.

Toray Industries Inc. developed carbonizing technology for Filament-Yarn produced from PAN in 1969.

Tonen Corporation, petroleum

refining company, have been developing to product a Carbon-Fiber from

oil type Mesophese-Pitch from 1970'.

Mitsubishi-Kasei Corporation

is continueing to accumulate the

coal chemical technology for half a century and succeeded to industrialize a coal pitch based Carbon-Fiber

production in 1978.

2. Carbon Fiber Strand

Carbon Fiber Strand, consisted

of about 12,000 filaments of HPCF., can be employed to retrofit by winding onto the surface of concrete and

steel structures because of superior

mechanical and chemical characteristics and flexibility carried from HP

CF. . External view of carbon fiber

strand is shown in Photo 1.

3. Carbon Fiber Sheet

(1) Torayca-Cloth is a textile

material fabricated from HPCF. "Toray

ca" produced by Toray. Torayca-Cloth

can retrofit the concrete structure

by glueing onto concrete surface

with epoxy resin, which can be hardened under atmospheric temperature.

(2) Forca-Tow-Sheet is a dry

uni-directional fiber fabric made up

of HPCF. . The fibers are attached

to a netted thin glass fabric with a layer of resin adhesive which is

easily dissolved into resin.

(3) Replark is a sheet of HPCF

. "Dialead" produced by Mitsubishi-Ka

sei, which is impregnated with epoxy

resin under stretching to uni-direction. Replark can be applicable

onto surface of various structures

because of high realization of full

tensile strength of HPCF. and superior workability.

External view of these carbon

fiber sheets are shown in Photo 2 to

4, respectively.

Ill
Photo 1 Carbon Fiber Strand Photo 2 Torayca-Cloth

Photo 3 Forca-Tow-Sheet

(4) Design Standards employed

by CRS Research Association:

Now, under the activity of "Carbon

Fiber Retrofitting System's (CRS)

Technical Research Association",

which was established and has been

managed by the co-operation with Oba

yashi, Toray, Tonen and MitsubishiKasei and has been researching the

application of HPCF. to the concrete

structure, following data is used as specific values on design work;

Tensile Strength : 250 kgf/mm2

Elastic Modulus : 25,000 kgf/mm2

Elongation : over 1.0 %

RETROFITTING METHOD WITH CARBON

FIBER STRAND FOR EXISTING REINFORCED

CONCRETE COLUMNS AGAINST SEISMIC

LOAD

Photo 4 Replark

1. Application Research On Carbon

Fiber

It is noted that the research

activities on Carbon Fiber need to

fully employ the merits of that, such

as high strength and durability, be- cause unit-weight price of Carbon

Fiber is 100 to 200 times of that of

steel. It is also necessary to develop the applications under special

conditions, in which the current con- struction procedures are not available nor effective. From these views,

the employment of carbon fiber for

retrofitting of existing reinforced

concrete structures is considered to

be technically and economically

promising or not.

112
For retrofitting of existing

reinforced concrete structures, two

types of carbon fiber products;

Carbon-Fiber Strand and Carbon-Fiber

Sheet, are employed. The aimes of

retrofitting techniques employed

these products are as follows;

1) Shear retrofitting by winding with Carbon Fiber Strand

or Sheet.

2) Flexural retrofitting by

glueing with Carbon Fiber

Sheet.

3) Combination of the above

technique 1) and 2)

.

Winding Spirally Carbon Fiber Strand

Fig. 2 Shear Retrofitting for Columns

2. Fundamental Properties Of Carbon

Fiber Strand

The high strength and high

modulus type of Carbon-Fiber (HPCF.

)

can be employed for the structural

engineering applications, and to re- alize full strength of Carbon-Fiber,

Carbon-Fiber Reinforced Plastics (CF

RP) , in which Epoxy Resin is impregnated and hardened, is usually car- ried out.

Stress-strain relationship of

CFRP Strand is as shown in Fig. 4,

comparing other materials, and fundamental properties are as follows;

Bundle of 1000 to 12000 A Piece of Fiber Monofilaments

I Monofilament! 1 Strand! | UP Sheet]

Fig. 1 Products of Carbon Fiber

Fig. 3 Flexural Retrofitting for Chimneys

Strain (10

" 2

) Fig. 4 Comparison of Stress

Strain Relationship

113
Tensile strength: 300 kgf/mm2,

10 times of that of steel

Elastic modulus : 2,4000 kgf/mm2,

almost equal to that of steel

Elongation : 1.2%,

much smaller than that of steel,

Yielding does not appear

Weight density : 1.8 gw/cm3,

one-forth of that of steel

Durability : Excellent,

rust does not appear

The tensile test results of

the CFRP Strands manufactured by Mit

subishi-Kasei , which is the only one currentry employed for this retrofitting method in Japan, are shown

in Fig. 5

.

In usual, because of the nominal tensile strength of CFRP Strand

is defined to 95% of passing rate

within test pieces through the material test, the value of specified

tensile strength of CFRP Strand is

adopted to be 270 kgf/mm2. If 99%

of passing rate is necessary, then

the value of that is adopted to be

2 50 kgf/mm2.

Carbon Fiber Strand has superior properties than steel, such as high strength, high durability, and

light weight, however, brittleness

should be noted for application of

Carbon Fiber.

For retrofitting material , Carbon Fiber, which has higher elastic

modulus than that of Aramid or Glass

Fibers, is chosen by the reason that

confinement of concrete is increased

by employing higher modulus materials.

A

10

o

c

<L>

<D 5

10 Average Value 30.5 tonf/cm2 (434x1 Opsi)

Standard Deviation 1.55tonf/cm2 (22x1 Opsi)

27.0 28.0 29.0 30.0 31.0 32.0 33.0

27.9

/ 28.9 ////// 29.9 30.9 31.9 32.9 33.9

Fig. .5 Tensile Test Result of CFRP strands

Tensile Strength (tonf/cm2

)

3. Seismic Retrofitting Method For

Existing Concrete Columns

(1) Outline of CFRP Strand Winding

Method

There are some existing reinforced concrete columns, which have

not enough shear strength and ductility against a severe earthquake

shock. For these occasion, a new seismic retrofitting method for such

columns has developed by Obayashi

Corp. and Mitsubishi-Kasei Corp. . In this method, Carbon Fiber Strand is

wound onto the surface of the existing column. This technique can improve easily the earthquake resistant capacity of columns as follows:

1) Increase of shear strength

2) Improvement of ductility

3) Increase of compressive

capacity

114
Various tests, described in

the following chapters, were carried

out to discuss the above mentioned

retrofitting performance.

From these result, it is recognized

that this retrofitting method is effective for improving shear strength

and ductility, and quantitative evaluations of retrofitting performances

were established, and design procedures of this technique were proposed.

The winding work of Carbon Fibers in this retrofitting method is

carried out by an automatic winding

machine, by the reason of expensive

Carbon-Fiber is employed, therefore,

machine-based construction procedure

is necessary to save labors and

costs. The automatic winding ma- chine for chimney is shown in photo

5 and for building in Photo 6.

This machine is also applicable to

retrofitting for bridge columns.

(2) Advantages of CFRP Strand

Winding Method

This method has the following

advantages, comparing with others.

1) easy realization for required

shear and ductile capacities.

2) no deteriorate influence to

stiffness of the existing columns.

3) minimum weight increase by

retrofitting.

4) no skillful workers for con- struction.

5) easy quality control for

construction.

Winding Spirally Carbon Fiber Strand

Fig.Q Seismic Retrofitting for Columns

Photo 5 Winding Machine for Chimneys

Photo 6 Winding Machine for Columns

115
4. Structural Performance Of Columns

Retrofitted By CFRP Strand

Winding Method

(1) Lateral loading test against

retrofitted columns with circular

cross section

The aim of this test was to

verify the influence on strength and

ductility of retrofitted columns

with CFRP Strand. The employed

specimens, as shown in Fig. 7, were 5 reinforced concrete columns, which were 1/4 scaled models with circular

cross section.

Test variables were ® retrofitting

quantity of CFRP Strand and (2) substrate treatment (Bond treatment and

Unbond treatment) between CFRP

Strand and concrete surface as shown

in Fig. 8.

800

A-A' Section

Fig. 7 Column Specimen

The findings from this test are as follows;

1) The CFRP Strand Winding Method improves strength and ductility

of columns.

2) The CFRP retrofitting can change the failure mode of columns

from bond splitting type to flexural

type.

3) Heavy CFRP retrofitting is

quite effective for strength in- crease.

4) The Bond treatment is effective for strength increase and the

Unbond treatment is effective for

ductility improvement.

(1) Bond Treatment (2) Unbond Treatment (3) Non-Treatment

Carbon Fiber Strand (Impregnated with Resin) Carbon Fiber Strand (Impregnated with Resin)

Resio

Crack

Concrete

. Stress Concentration , Strain and Stress

"Crack

Strain and Stress Distribution of CFRP Strand

Fig. 8 Substrate Treatment

of Concrete Surface

H 100

20 40

Displacement (mm)

a) Retrofitted

-20

(a) Retrofitted (b) Unretrofitted

20 40

Displacement (mm)

(b) Unretrofitted

Photo 7 Columns after Testing Fig. 9 Load displacement Relationship

116
(2) Lateral loading test against

retrofitted columns with square cross section

The aim of this test was to

discuss the performance of the CFRP

retrofitting method for columns with

square cross section. The employed

specimens, as shown in Fig. 10, were

10 reinforced concrete columns, which

were 1/4 scaled models with square

cross section.

Test variables were CD retrofitting

quantity of CFRP Strand and (2) substrate treatment (Unbond treatment

and Non-treatment) between CFRP

Strand and the concrete surface.

The failure modes are shown in

photo 8, comparing the retrofitted

specimens with the non-retrofitted

ones. The findings from this test are as follows;

1) For the columns with square

cross section, the CFRP retrofitting

is as effective as for the circular

cross section.

2) Diffrence between the Unbond

treatment and the Non-treatment is

negligibly small. The Non-treatment is preferred, because it saves

one process of construction.

3) The reinforcing performance

of CFRP Strand can be evaluate as that of ordinary steel bars.

In this procedure, the strength ratio

between CFRP and steel can convert

the CFRP quantity into the steel

reinforcement quantity.

(3) Flexural shear loading test

against retrofitted beam

The aim of this test was to

propose a quantitative evaluation

for shear strength of columns retrofitted by CFRP Strand Winding Method.

The employed specimens, as shown in

750

D13 Hoop

Ml

3 C o o t— CN r-1

140

200

—or1

30

A-A' Section

Fig. 10 Column Specimen

(a) Retrofitted (b) Unretrofitted

Photo 8 Failure mode of Tested Columns

Fig. 11, were 15 reinforced concrete

beams with rectangular cross section*

Test variables were CD shear

reinforcing quantity of CFRP Strand,

(2) shear span ratio (= Depth/Shear

span) and (3) compressive strength

of concrete.

The failure patterns and shear

& deflection relationship are shown

in Fig. 12 and Fig. 13, respectively.

The findings from this test are a

s follows;

1) CFRP Strand Winding is quite

effective for shear reinforcement.

2) Increase in CFRP Strand improves maximam shear bearing force

of the specimen.

117
3) The shear strength of the

members retrofitted by CFRP Strand

can be calculated by the application

of the current design equation for

ordinary steel reinforcing bars.

In the equation, the design tensile

strength of CFRP Strand is defined

to be 2/3 of the full of that.

400

BS

BM

Series 800

Series 1,2 00

BMW Series 1,200

BL Series 1,600

Carbon Fiber Strand

400

180 4d3o3oko

" ,D22 R=30

Unretrofitted Retrofitted

Fig. 1 1 Specimen of Shear Strength Test

(a) pcf =0.24 % (c) pc{ =0.06 %

(b) pc{ =0.12% (d) pc{ =0%

Fig.ll Failure Patterns (BMW Series)

o o —°- o o o —'Q

Carbon Fiber Strand

Shear Force(tonf)

30 Maximum Strength

\7

i°cf =°- 24%

pcl =0.12%

40 50

Deflection (mm)

Fig. 13 Shear Force - Deflection Relationship

118
o

fl

R=30

ISO

400

200

] 1 1 1 X : Fracturing of CFRP Strand CR40

^""^N^CROO

CFRP Strand

CB Series CU Series CR Series CQ Series

FigAi Specimens

0.0 0.4 0.8 1.2 Longitudinal Strain (10-2

)

Fig. 15 Stress - Strain Relationship (CR Series)

o

54-H

400 Pcf =

Act 4

19 V

i

V

[]

200

0

o

Ac - S

/O

f

; Ratio of CFRP Strand

Acf : Area of CFRP Strand

Ac

: Area of Concrete Section

t : Perimeter of Specimen

S : Spacing of CFRP Strand

o

o

o

• : CB Series

V : CU Series

o : CR Series

CQ Series

r

0.1 0.2 0.3 0.4

/°cf<*>

Fig. 1 6 Compressive Strength

of Confined Concrete

(4) Compression test of concrete

cylinders confined by CFRP Strand

Winding

The aim of this test was to

evaluate confinement effect on com- pressive strength and ductility of

concrete confined by CFRP Strand

Winding. Test variables were CD

quantity of CFRP Strand, (2) type of cross section of the specimen, and

so on

.

The examples of Stress-Strain

relationship of concrete is shown in

Fig. 15, and the relationship between

compressive strength and CFRP quantity is shown in Fig. 16. The findings from this test are as follows;

1) The CFRP Strand retrofitting

improves the compressive strength

and ductility of the concrete specimen. This effect grows when the

quantity of CFRP Strands is increased.

2) The confinement effect is more evident for the cylinders than

for the rectangulars.

(5) Lateral loading test against

retrofitted bridge columns with

rectangular cross section

The aim of this test was to

119
discuss the performance of bridge

columns with rectangular cross sec- tion retrofitted by CFRP Strand

Winding method against shear failure

and buckling of existing longitudinal reinforcing bars. The employed

specimens were 11 reinforced cncrete

bridge columns, which were 1/3 scaled

models with rectangular cross section. Test variable were (T) quantity of CFRP Strand, (2) shear span

ratio, and so on. The failure pattern and load & displacement relationship for shear

failure type of the specimen are

shown in Photo 9 and Fig. 17, respectively. Others for Re-bar buckling

type of the specimen are shown in Ph

oto 10 and Fig. 18, respectively.

These test results lead the

following conclusions for the appropriate retrofitting with CFRP Strand

; 1) Shear failure of the bridge

columns can be avoided, so that a required level of ductility can be

provided.

2) Buckling of longitudinal

reinforcement in flexural failure

can be avoided, so that a required

level of ductility can be provided.

(a) Retrofitted (b) Unretrofitted ( a) Retrofitted (b) Unretrofitted

Photo 9 Failure Mode Photo 10 Failure Mode

Fig. 18 Load - Displacement Relationship (Re-bar Buckling Type)

120
5. Durability of CFRP Strand

To verify long-term effects of

the retrofitting by CFRP Strand

Winding Method, the following tests

have been carrying out.

CD An accelerated weathering test

of CFRP Strand.

® An out-door exposure test of

CFRP Strand.

® A chemical resistance test of

CFRP Strand.

The result of the accelerated

exposure test is shown in Fig. 19.

In usual, this retrofitting method

employs some finishings over the

wound CFRP Strands around the column

surface, therefore, durability problem of CFRP Strand does not seem to

rise. Even if there is no finishing, durability of the CFRP Strand

is almost perfect.

RETROFITTING WITH CARBON-FIBER

SHEET GLUEING METHOD

FOR CONCRETE STRUCTURE

1. Application To Concrete Beam

Retrofitting

Carbon-Fiber Sheet has a wide

application area by the reason of

its speciality; mechanical and chemical superior characteristics, and

easy employment in the narrow and

broad space or at the small and wide area of structure without machinery

but with only man-power because of

light weight of that.

In the application to beam

retrofitting, as shown in Fig. 20,

Fig. 20 Glueing of carbon fiber

I i ! ii ; ; ! i l_

0 2000 4000 6000 8000 10000 Exposure Time(hrs)

Fig. 19 Tensile Strength - Exposure Time

Relationship

Carbon Fiber Sheet is glued onto the

surface along to the member axis direction of existing beam, with the

epoxy resin. After resin impregnated into Carbon Fiber Sheet and

hardened, Sheet type Carbon Fiber

Reinforced Pasties (CFRP Sheet) works

for flexural retrofitting.

On this retrofitting method,

as shown in Fig. 21, CFRP Sheet, which

is glued onto face of concrete, can work as same as Re-Bar and co-operated with Re-Bar.

CFRP Sheet, consisted of Carbon

Fiber volume 175~300 gw/m2, has a- bout 0. 1 mm converted thickness and

has about 250 kgf/cm permissible

force per each ply, so that equals to

13 mm diameter Re-Ber arranged about

150 mm spacing. The number of glueing plies of CFRP Sheet would be determined by the condition of working

load. LOAD

Carbon fiber UD-prepreg

Fig. 2 1 Retrofitting for bending moment

121
Fig. 22 Pier's Column Retrofitting by Carbor Fiber Sheet

2. Application To Pier's Column

Retrofitting

As shown in Fig. 22, Carbon Fiber Sheet is glued along to the member axis direction of existing pier

with epoxy resin. CFRP Sheet can resist against working moment, as same as beam. And in the case that

CFRP Sheet is added to horizontal

direction, that Sheet can work for

shear force resistance and also have

the confinement effect for the ver- tical CFRP Sheet.

REFERENCES

1) Y.KOBATAKE: TA RESEARCH FOR SEISMIC

RETROFITTED REINFORCED CONCRETE COLUMN BY NEW MATERIALS (No. 1 ~3)j,0bayashi Corp. Technical Research Institute's Report No. 33 —38, 1986—89

2) Y.KOBATAKE: TA STUDY ON THE STRENGTHENING WITH CARBON FIBER FOR EARTHQUAKE RESISTANT CAPACITY OF EXISTING

REINFORCED CONCRETE COLUMNS.!, Workshop on Repair and Retrofit of Existing Structures UJNR Tukuba, Japan

May 8-9, 1987

3) Y.KOBATAKE: TAPPLICATIONS OF RETROFIT METHOD WITH CARBON FIBER FOR EXISTING REINFORCED CONCRETE STRUCTURE

J, The 22nd Joint UJNR Panel Meeting,

Japan, May 12-14, 1990

4) Moto. 0KAN0: TThe Strengthening

Methods of Existing Road Bridge with

Carbon Fiber Reinforced Plastics],

Fire Resistant Materials, No. 45 [5],

p332— 339, 1993

5) Masaj. KOGA: TRehabilitation and

Strengthening for Structures with

Carbon Fibers], No. 5, Technical Research Report of N.S.I. and T. Association, p37— 43, 1993.11

6) T.H0SHIJIMA: TStrengthening of Rein

forced Concrete Chimneys, Columns

and Beams with Carbon Fiber Reinforced Plastics], FRP Reinforcement for

Concrete Structure -International

Symposium- ACI.SP-138, 1993

7) Y. KOBATAKE: TSeismic Retrofitting

of Existing Reinforced Concrete Bridge Columns by Carbon Fiber Strand

Winding Method], Obayashi Technical

Presentation Report, March 1994

122
Construction Automation Research

by

K. P. Chong and S. C. Liu

National Science Foundation

ABSTRACT

Construction processes are behind the state- of-the-art of manufacturing as well as computer sciences and control. Successful

and efficient construction automation require

robust, effective and flexible material

handling, system integration, sensing and

control, etc. A survey of mainly NSF

supported research in construction is presented in this paper.

KEYWORDS: Construction automation;

material handling; computer- integrated

methods; robotics; sensing and control.

1. INTRODUCTION

Construction is one of the largest industry in

the U.S.(Chong, et al, 1993; NCR, 1988),

amounting to about 8% of the GNP and

employing six million people ( Rossow and

Moavenqadeh, 1974). A bottleneck of

construction research, development, and

competitiveness is automation. Constructed

systems include the traditional civil

engineering structures (such as, buildings,

bridges, tunnels, railways, airports and

harbors), demolished constructed systems

due to earthquakes, hurricanes, blasting and

other natural/man-made hazards. In both

cases there are mismatches between

manufacturing know-how and construction;

between computer science/control know-how

and construction. Successful and efficient

construction automation require robust and

effective: material handling; data capture;

artificial intelligence; system integration;

operation research; computational mechanics

in structured and unstructured (demolished)

structures; sensing, pattern recognition, and

control; qualitative vision analysis;

prefabrication; intelligent and flexible

robots, etc. The end product should be

safe, durable, environmentally friendly and

intelligent civil infrastructure systems

(Chong, et al 1992; 1993). These

techniques and research findings can also be

applied to mining, ship construction and airplane manufacturing.

Major objectives of the research in

construction automation are:

1. To investigate construction dynamics of

both structured and unstructured facilities.

2. Continue to develop a better

understanding of analysis, design and

construction processes, leading to basic

principles and methods for robust

automation.

3. Continue to support research efforts in

computer integrated construction for the

competitiveness of the construction industry

and to build up the technology base.

4. Encourage multidisciplinary research

collaborations among groups of academic,

industry and government agencies.

5. Encourage knowledge and technology

transfer to industry and users.

In a recent paper (Lewis, 1993 + ) the

inclusion of various structures to automate was addressed. In buildings, the primary

needs for automation occur in structure,

123
enclosure skin, interior finishes. In light

industrial the primary needs are structure,

piping, electrical. In heavy industrial the

needs are piping, electrical, mechanical

equipment. Finally, in power, chief

requirements are in piping, mechanical

equipment and electrical.

The Construction Industry Institute (CII) has

studied what makes sense to automate

(Tucker et al. 1990); the answers depend on

the type of construction application^ usiness

Roundtable Report 1982).

According to Hendrickson and Au (1989)

construction automation makes sense in the

following situations:

* in hazardous environments - e.g. outer

space, nuclear reactors and undersea

construction

* robotics permit broader use of non- human senses and constrained spaces

* robots are tireless

* to achieve improved quality, greater

consistency and precision

* cost savings

Examples of existing construction robots

have been documented by various

researchers (e.g. Oppenheim and

Skibniewski, 1988; Paulson, 1985; and

Whittaker, 1986).

2. WORKSHOPS AND SYMPOSIA

Numerous workshop related to construction

automation have been held. The following

is a summary of three of the workshops

sponsored by NSF.

o NSF workshop on "Construction

Automation: Computer-Integrated

Construction", April, 1987; John L. Wilson

(Lehigh University).

The commonalty of directions is discussed

under six headings (Wilson, 1987):

(a) System Architecture and Organizational

Structure - an architecture is needed for

open, interactive environment for computer

applications in engineering, design,

construction and facilities management.

(b) Structure, Formalization and

Classification of Knowledge - research

includes data requirements in each discipline

and sharing between disciplines.

(c) New Languages and Representation

Techniques - to provide unified, timely and

accurate two-way communication among all parties.

(d) Intelligent Interfaces - to interface

computer applications and databases

intelligently with knowledge-based

components.

(e) Designing for Automation - new design

that recognizes the characteristics of

construction machinery, including material

handling, maintenance.

(f) Sensing and Monitoring - for

dimensional/configuration control,

materials/components identification, location

and control, etc.

o "Technological Innovation in

Construction", April, 1991; Dan Halpin

(Purdue University)/R. Fechtig (ETH,

Zurich).

124
Subjects considered as major activities are:

(a) Organizational structure and its impact on innovation and competitiveness

(b) Management of technological

innovation in construction

(c) The influence of contracting forms on

technological innovation and transfer

(d) Means of evaluating risks and

potential returns on investments in advance

technology

(e) Identification of major needs for

advanced technology in individual countries

(f) Analysis of the process by which

technologies have been successfully

transferred

o "Research Needs in Automated

Excavation and Material handling", April,

1993; B. Dendrou (ZEi/Mi Inc.) / R.Sterling

(University of Minnesota) / M. Gaus

(University at Buffalo).

Excavation technology is a key component

in many publicworks activities and services

which affect the national health and safety,

standard of living and economic

competitiveness. Furthermore, the relative

importance of underground construction in

infrastructure will increase drastically in the near future, as environmental and living

standards rise. However, the U.S.

underground construction industry at present

is weak. The Directorate of Engineering at NSF recognizing the importance of this

industrial basis has recently supported a review of excavation and underground

technology in Europe and Japan. The results

of this review clearly demonstrate the both

the Japanese and Europeans are moving

ahead of the U.S. in Research an

Development, and innovation in automated

excavation technology, which will position

them to capture a potential world market

worth 100 billion dollars over the next 10

years (this estimate reflects the findings of

two recent surveys conducted for the

directorate for Engineering at NSF).

Topics that are being addressed are the

following:

* New developments of the Underground

and Surface Space

* Automated Excavation and Materials

Handling

* Site characterization and Modeling

Other Major symposia include the annual

"International Symposium on Automation

and Robotics in Construction(ISARC), the

most recent one (10th Symposium) was held

in Houston, Texas in May, 1993 (Watson, et

al, 1993)

3. EXAMPLES OF RESEARCH AREAS

AND PROJECTS

In 1987, NSF issued a "Construction

Automation Research" Initiative, which was

well received by the community. The

following examples listed in the initiative are purely illustrative. The computer is expected to play a central role in all of these

areas, in terms of modeling, analysis,

representation, and control. The close

coupling of the design process with

construction via computerization will

enhance constructability, functionality,

flexibility, and maintainability of the end

product.

125
* coupling of design and construction

through project databases and knowledge

bases

* design integration with construction

process planning and control, end use, and

maintenance

* computer models of the whole

construction process (e.g., incorporating

geometric and functional relationships in

large spatial assemblies)

* methodologies for planning and

scheduling of the whole construction process

* knowledge management (computerbased deployment and use of knowledge

among interacting processes)

* adaptable construction systems

(analogous to flexible manufacturing

systems)

* logistics of parts, materials, and

information

While robotics, sensing, metrology, materials

handling, and other specific technologies

may be included in the scope of proposed

projects, the work supported by this initiative

will focus on computer integration, and will

be expected to produce new engineering

knowledge relating to the systems integration

of construction rather than to any individual

operation or component of the process.

Proposals may involve real systems and

processes as well as more theoretical or simulation-based investigations. The

research emphasis is on technical rather then

management issues.

Strong involvement of industry in the

proposed research is particularly encouraged.

This may take several forms, such as establishment of a research advisory board

including members from industry, or actual

research collaboration between academic and

industry engineers.

In the U.S., most of the basic physical

systems research is carried out by materials

and equipment manufacturers (e.g. W.R.

Grace, Caterpillar Tractor, etc.). This is strictly product oriented research (Halpin, et

al, 1991). Recent studies indicate that the

U.S. has not been aggressive in transforming

basic research in the construction area into

marketable products (Gerwick, 1990). The

Japanese excel in taking existing research

and coverting it into market ready

construction technology. In 1990, the senior

author organized a panel of experts to visit

strongholds and sites in Japan to assess its construction technologies. According to this

report(Tucker, et al, 1991) Japanese

companies have invested heavily in

developing automated equipment, although

they have produced very few practical pieces

to date. Much of their motivation to automate stems from their desire to improve

the unsafe, dirty, and unsophisticated

reputation construction has among workers.

A poor safety record can be cause for an owner to exclude a contractor from bidding

his projects for a period ranging from 30

days to several months. Nevertheless, very

few young workers are entering construction,

leaving companies with an aging work force.

The companies believe that automating

construction equipment will make

construction safer and give it a more

technical aura.

Additionally, technological innovation can be

used to help sell both existing customers

and new prospects. Despite their push

toward automation of construction

equipment, Japanese companies do not use computers as widely as U.S. companies for

schedule or cost control, relying instead on manual methods for the most part. However,

126
Table 1. Relative Status of Construction R&D in the U.S. and Japan.

R&D IMPLEMENTATIO

N

STATUS TREND STATUS TREND

Structural Steel + T + T

High-Performance Concrete + T T

High-Performance Composites + T + T

Geotechnical Materials + T 0 T

Construction Contractors + T

Construction Suppliers 0 T

Universities i

Government 0 —

Coding System

Japan Compared to the U.S.

Status Trend

< Far Behind

Behind

O Even

+ Ahead

> Far Ahead

ft Pulling Ahead Strongly

T Gaining Ground

—> Holding Constant

i Falling Behind

•U Losing Quickly

127
Japanese companies are actively exploring

ways to transfer information form CAD

models to field equipment, including

manipulation of that information and data

from the surrounding environment using

artificial intelligence. The reduced incidence

of claims in Japanese construction may

partially explain the less frequent use of

computers in project controls. The following

Table summarizes the relative status of

construction R & D in the U.S. and Japan

(Tucker, et al, 1991):

Initiatives such as the Civil Infrastructure

Systems Research of NSF (Chong, et al, 1993; Lepkowski, 1993) and the

Construction Productivity Advancement

Research (CPAR) program of the U.S. Army

Corps of Engineers are attempting to rectify

this shortcoming by funding and/or

encouraging technology transfer of

construction technology.

In order to focus resources and accelerate

technology development in architectural,

engineering and construction related areas, research centers have grown up within the

past decade to link industry and universities

in cooperative efforts (Halpin, et al. 1991).

Among these centers are:

( 1 ) The Construction Industry Institute (CII)

at University of Texas, Austin, TX

(2) NSF-ATLSS Engineering Research

Center (ERC) on Large Structural Systems,

Lehigh University, Bethlehem, Pennsylvania.

(3) Center for Integrated Facilities

Engineering (CIFE), Stanford University,

Stanford, California

(4) NSF-National Center for Earthquake

Engineering Research, Sunny-Buffalo, NY

(5) NSF - ERC for Offshore Technology

Research Center, Texas A&M University

and University of Texas at Austin

(6) NSF - Science & Technology Center

(STC) for Advanced Cement-Based

Materials

(7) NSF - STC for Southern California

earthquake Center, University of Southern

California

(8) National Institute of Standards and

Technology (NIST), Robot Systems Division

(see Albus, et al 1992)

(9) U.S. Army Corps of Engineers

Construction Engineering Research

Laboratory (CERL)

(10) Factory-Built Housing Research Center

Tennesse Technological University,

Cookville, TN

The following are examples of NSF

sponsored construction automation research

projects related to civil infrastructure

systems.

*Bob Tatum, Stanford University:

"Mechanism and Strategies for

Technological Innovation in Construction."

For fundamental understanding of

construction technology and to provide

mechanisms/strategies for more rapid

technical advancement.

*Chris Hendrickson/Dan Rehak, Carnegie

Mellon University: "KBES Construction

Planning Process." For the selection of

construction technology, work tasks

definition, costs/time/resources estimation

and scheduling.

128
*Jeffrey Ullman, et al, Stanford University:

"Integrated Data Exchange and Concurrent

Design for Engineered Facilities." An

integrating compute environment to support

concurrent design with fast data exchange

and powerful change management

capabilities, and improve life-cycle facility

management.

*Owen R. Mitchell, et al, University of

Texas, Arlington: "Integrated Modeling and

Control for Intelligent Material Handling."

Intelligent Material handling systems

approach for rapid deployment and

automation of new material handling

requirements in environments with

uncertainties and minimal structure.

*Gerhard Fischer et al, University of

Colorado at Boulder: "Supporting

Collaborative Design with Integrated

Knowledge-based Design Environments."

Conceptual framework and prototype system

for collaboration among members of design

teams.

Raymond Levitt, Stanford University: "The

virtual Design Team: Simulation, Decisionmaking, and Information Flow in

Concurrent, Multidisciplinary Design."

Integration of Al and coordination theory to develop a virtual design team, which is an

object-oriented, discrete event computer

simulation model of the flow of decisions

and information among participants in

complex, multidisciplinary design projects.

* J.M. De La Garza, VPI: "Methods for

Expediting the Transfer of Construction

Technology." KBES and CAD are used in

design/construction integration. Life-cycle

applications and technology transfer are also

investigated.

Yvan J. Beliveau, VPI: "Dynamic Damping

of Payload Motion for Cranes." Develop a

strategy for damping of payload motion for

various kinds of cranes used in construction.

*J. Vanegas, Georgia Tech: "Integrated

Design/Construction." For conceptual

framework and model for infrastructure

rehiblitation.

*J.S. Russell, University of WisconsinMadision: "Computer-Integrated Concrete

Placement Systems." Built with existing

pumped-concrete equipment.

*O.R. Mitchell, F.L. Lewis, University of

Texas at Arlington: "Integrated Modeling

and Control for Intelligent Material

Handling." Allows rapid automation of any

material handling requirements including

uncertainties in the environment.

*M. J. Skibniewski, Purdue University:

"Computerized Robotics Management for

Construction Sites." A system to manage

diverse robots on construction sites

(Skibniewski, 1992; Russell, et al, 1990).

William L. Whittaker, et al, Carnegie

Mellon University: "Autonomous Retrieval

of Buried Objects." To develop and integrate

the perception, planning, and manipulation

technologies needed for autonomously

detecting, uncovering, and retrieving fragile

buried objects, e.g. waste containers,

munitions, and cables exposure.

*Boyd C. Paulson, Stanford University: "An

Object-Oriented Approach for integrated

Project Management Software." To construct

systems that offer greater flexibility, higher

intelligence and better integration, using

"building block"

129
*Victor E. Sanvido, Pennsylvania State

Univeristy: "Applying Computer-Integrated

Manufacturing Concepts to Construction."

Identifies activites in construciton (in

relationship to integrated Identifies activities

in construction(in relationship to integrated

manufacturing), techniques nad tools for

construction automation.

*Carl Haas, et al, University of TexasAustin:"Large Scale Manipulator Technology

for Construction." Automation in piping

construction, e.g. in power plants, petrochemical and chilling plants

4. SUMMARY AND CONCLUSION

An overview of the state of the art and NSF

projects in construction automation are presented. The authors hope that this paper

will act as a catalyst, generating interest and

further research in these areas. This paper

reflects the personal views of the authors,

not necessarily those of the National Science

Foundation.

BIBLIOGRAPHY

Albus, J.S., Bostelman, R.V. and Dagalakis,

N.G., "The NIST SPIDER, a Robot Crane,"

J. of Robotic Systems, October 1992.

Business Roundtable Report (BRR) B-3,

Construction Technology Needs and

Priorities, August 1982.

Chong, K.P., Liu, S.C., et al Civil

Infrastructure Systems Research: Strategic

Issues, NSF 93-5, Jan. 1993.

Chong, K.P., and Liu, S.C., "Engineering

Research on Smart Materials," NSF

Directions, Vol. 5, No. 5, 1992.

Halpin, D.W., et al, Technological

Innovation in Construction , Purdue

University, April 1991.

Halpin, D.W., "Automated Construction

Technology: Background and Barriers,"

IABSE Proceedings P- 147/90, August 1990.

Hendrickson, C, and Au, T., Project

Management of Construction , Prentice Hall,

1989.

Lepkowski, W., "NSF Proposes New

Infrastructure Research Unit," Chemical and

Engineering News, P.27, June 7, 1993.

Lewis, F.L., "Robotics, Automation, and

Machine Intelligence in Construction," L

American Institute of Constructors, to appear

(1993+).

NRC (National Research Council), Building

for Tomorrow - Global Enterprise and the

U.S. Construction Industry National

Academy Press, 1988.

Oppenheim, I.J., and Skibniewski, M.J.,

"Robots in Construction," Encyclopedia of

Robotics, John Wiley & Sons, 1988.

Paulson, B.C., "Automation and Robotics for

Construction," ASCE J. of Construction

Engineering & Management, Vol. Ill, No,

3, pp. 190-207, 1985.

Rossow, J.K., and Moavenzadeh, F., The

Construction Industry, Department of Civil

Engineering, MIT, 1974.

130
Russell, J. Skibniewski, M., and Vanegas, J., "Framework for Construction Robot Fleet

Management System," ASCE Journal of

Construction Engineering and Management,

Vol. 116, No. 3, September, pp. 448-462,

1990.

Skibniewski, M. "Robot Implementation

Issues for the Construction Industry," in

Human Robot Interaction , Rahimi, M. and

Karwowski, W., eds., Taylor & Francis

Publishers, London, U.K., -Washington,

D.C., ISBN 0-85066-809-3, pp.347-366,

1992

Tucker, R.L., et al, Construction

Technologies in Japan Japanese Technology

Evaluation Center (JTEC), Loyola College,

Baltimore, MD, June, 1991.

Tucker, R.L., et al, CII Task Force, "Needs

Assessment for Construction Automation,"

Construction Industry Institute, University of

Texas-Austin, September 1990.

Watson, G.H., Tucker, R.L. and Walters,

J.K. (eds). Automation and Robotics in

Construction X, Elsevier, Amsterdam, 1993.

Whittaker, W.L., "Cognitive Robots for

Construction," Annual Research Review, The

Robotics Institute, Carnegie Mellon

University, Pittsburgh, pp. 6-17, 1986.

Wilson, J.L., NSF Workshop on

Construction Automation: ComputerIntegrated Construction, Lehigh University,

April, 1987.

Design Guidelines for

Precast Reinforced Concrete Buildings

by

Shinsuke Nakata*. Masaomi Teshigawara*

Takashi Kaminosono* and Hitoshi Shiohara*

ABSTRACT

This paper introduces "Design

Guidelines for Reinforced Concrete

Buildings (draft)," under development

as a part of Japanese PRESS S project.

The guidelines has been drafted by

Guidelines Drafting Working Group,

and discussed in Design Guidelines

Committee of Japanese PRESSS.

Values of some design coefficients have

not been finalized. Extensive

commentary will accompany the

guidelines to explain the concept behind

requirements and to suggest methods of

calculation.

1. INTRODUCTION

Japanese side PRESSS project team has

already reported three research results; "Design Guidelines", "Construction

Guidelines" and "Design Manual for Precast Connections' . This introduces

only seismic design concept from

"Design Guidelines". This guidelines were completed by PRESSS Guidelines

Drafting Group ; S.Otani, Chairman,

University of Tokyo, M.Teshigawara,

Secretary, Building Research Institute .Y.Inoue, Housing and Urban

Development Corporation , M. Hayashi,

P.S. Corporation , T. Kaminosono,

Building Research Institute, I.Kawabata

Taisei Corporation , M.Kimizuka,

Tokyu Corporation ,S. Nakata, Building

Research Institute .K.Yagishita, Toda

Corporation ,K.Yoshioka, Ohbayashi

Coporation .

2. SCOPE

The total height of a building shall be

not more than 60 m. A structure, in each principal direction, shall consists of

ductile moment-resisting frames, or

ductile moment-resisting frames with

continuous structural walls, or independent structural walls. The

structural wall shall be continuous from

the foundation to the roof level. Mass,

stiffness and strength in a structure shall be arranged uniformly in plan and along

elevation. Eccentricity ratio and rigidity

ratio of a structure, specified in Building

Standard Law Enforcement OrderJapan

shall be not more than 0.15 and not less than 0.6, respectively. Height-to-width

ratio of a building shall be not more than

4. The structure of a building shall be

clearly planned to form a specified total yield mechanism, in which flexural yield

hinges shall develop, as a general rule,

at the ends of all floor beams and at the base of the first story columns and

walls. A building, designed and constructed

under the guidelines, shall resist gravity

loads and medium intensity ground

motions to ensure serviceability, and

shall not collapse during an intense

ground motion. The building shall maintain serviceability during a strong

wind and snow loading. Precast connection (denoted as PCa

connection) of PCa members shall be

provided with strength sufficient to transfer member actions caused by

specified design loads. The strength

deterioration and slippage deformation at PCa connections shall meet "Evaluation

Criteria for PCa Connection

Performance".

PCa members and connections shall

satisfy required serviceability, durability

and fire resistance.

3. MATERIALS

Quality, mix, production, materials

of concrete and precast concrete

specified for RC members shall meet the provisions of Japan Architectural

Standard Specification JASS-5

Building Research Institute Ministry of Construction Japan

133
"Reinforced Concrete Work".

Quality and types of concrete for PCa

connection shall be as good as or

superior to those of concrete specified

for RC members. Quality, mix,

production and materials shall meet the

provisions of "PRESSS PCa

Construction Guidelines." Grade and

quality of reinforcing bars shall meet

Japan Industrial Standards (JIS) G-3112

"Steel Bars for Concrete Reinforced."

Compressive strength of mortar used in

PCa connection shall be not less than

that of concrete specified for RC and

PCa members. Types, quality, mix,

production and materials shall meet the

provisions of "PRESSS PCa

Construction Guidelines."

Reinforcing bars can be jointed by

gas pressured welding, flare welding,

lap splicing and other joints specially

approved by Minister of Construction.

Work of gas pressured welding shall

meet "Guidelines for Gas Pressured

Welding for Reinforcing Bars" issued

by Japan Gas Pressured Welding

Institute.

Steel plates can be jointed by welding

or high tension bolts. Work of welding

and high tension bolt friction joint of

steel plates and work of flare welding of

reinforcing bars shall meet the

provisions of JASS-10 "Precast

Concrete Work" and JASS-6 "Steel

Work".

Table 1 : Types and nominal strength

of concrete specified for

RC and PCa members

Type of Specified Concrete

Concrete Strength (kgf/cm2)

Normal Concrete 210-360

Light-weight 210-270

concrete

4 .EARTHQUAKE RESISTANT

DESIGN (NONLINEAR

ANALYSIS PROCEDURE)

4.1 Outline of Structural Design

This section defines the structural

design method for structural members.

Structural design of PCa members may

follow the methods for RC members.

Design for Gravity Loads

Stress in every part of structural

members under dead load, live load,

and snow load, specified in Building

Standard Law Enforcement Order, shall

not exceed allowable stresses for longterm loading . Design of structural

members in the super-structure under

earthquake loads shall conform to two

cases provisions ; nonlinear analysis or

linear analysis . However,design of a building of more than 31 meters in

height shall conform to the provisions of

nonlinear analysis

.

The performance of super-structure,

basement and foundation structure of a building shall be examined for

serviceability limit state under medium

intensity earthquake motions and for

ultimate limit state under an intense

earthquake motion.

4.2 Method of Earthquake Resistant

Design

(1) Earthquake resistant design shall be

based on a static nonlinear analysis of a building under monotonically increasing

lateral loading taking into account

realistic elastic and inelastic

characteristics of constituent structural

members.

(2) The analysis may be carried out in

the longitudinal and transverse

directions, separately.

(3) Lateral load shall be increased

monotonically in the analysis under dead

loads, specified in Article 84 of Building

Standard Law Enforcement Order, and

live load for earthquake load calculation.

4.2.1 Serviceability Limit State

Design story shear Qi at story i for the

serviceability limit state design under the

action of an earthquake motion shall be

Qi =CiWi (4.1)

Ci = Z Rt Ai CB (4.2)

where,Ci: story shear coefficient, Wi:

sum of dead and live loads above the i-th story, Z: seismic zone coefficient, Rt: vibration characteristic coefficient, Ai:

coefficient for story shear distribution,

CB: standard base shear coefficient of

0.2. Coefficients Z, Rt, and Ai are defined in Notification No. 1793 of

Ministry of Construction, 1980.

134
4.2.2 Ultimate Limit State Required lateral load resisting capacity

Quni of story i for the ultimate limit state design shall be

Quni= Cuni Wi (4.3)

Cuni = Z Rt Ai Cunfi (4.4)

where,Cuni: story shear coefficient for ultimate limit state design, Z, Rt, Wi,

and Ai are defined in Section 4.2.1,

CunB: standard base shear coefficient of

0.3 for moment-resisting frame

structures, and 0.3, 0.35, and 0.4 for

wall-frame structures in which structural walls resist up to 30%, 30 to 70%, and over 70% of the base overturning

moment at design limit deflection Rul,

respectively. Fig.l shows the outline of

this design criteria.

4.3 Nonlinear Incremental Lateral Load

Analysis

4.3.1 Modeling of Building

(1) A building structure may be idealized

as a series of plane frames in a principal

direction if the effect of torsion and

transverse frames can be neglected.

(2) If the effect of torsion and transverse frames cannot be neglected, a building

must be analyzed as a three dimensional

structure. (3) The lateral displacement of the

parallel frames at each floor level may be

assumed to be equal.

(4) A structure shall be analyzed

including super-structure and substructure.

4.3.2 Lateral Force Distribution

The distribution of lateral loads shall be

the same as the one assumed in the

serviceability limit state design

4.3.3 Lateral Loading Analysis

(1) Horizontal loads shall be assumed to act at the floor level of each floor. (2) The analysis may terminated when

the maximum story drift angle of a story

reaches a design proof drift angle of

Ru2.

4.3.4 Modeling of Structural Members

(1) A column and beam shall be

represented by a line member

considering the following deformations:

Column: Bending, shear and elastic axial deformations,

Beam: Bending and shear deformations.

(2) A beam-column connection may be

assumed to deform in shear, or to be

rigid in a region specified in Commentary of Article 8.2 in "Standard

of Structural Calculation for Reinforced

Concrete Structures" of Architectural

Institute of Japan.

(3) Inelastic deflection of a column and

beam may be assumed to concentrate at the member end, represented by rotation

of a rigid-plastic rotational spring.

(4) Shear, flexural and axial

deformations at section centroid of a

structural wall shall be included in the

model.

(5) The deformation of soil and piles

shall be considered in the analysis if appropriate.

4.4 Stiffness and Strength of Structural

Members

4.4.1 Elastic Stiffness

(1) The effect of reinforcing bars and

orthogonal members shall be included in the flexural stiffness of a member.

(2) Shape factor for shear deformation

may be taken to be 1.2 as a rectangular

section.

4.4.2 Hysteresis Model

(1) The stiffness characteristics of a column and beam under monotonically

increasing load may be assumed to be of

trilinear type with stiffness changes at flexural cracking of concrete and tensile yielding of longitudinal reinforcement.

(2) Axial stiffness of a column may be

assumed to be proportional to axial load. (3) Stiffness characteristics of a structural wall shall be evaluated

separately for shear and flexure.

Stiffness in shear shall be assumed to change at shear cracking and shear

strength, represented by a trilinear

relation. Stiffness in flexure shall be

assumed to change at flexural cracking

and yielding, represented by a trilinear relation.

4.4.3 Yield Resistance of Structural Members

Yield resistance of structural members

shall be evaluate using material strength

specified in Chapter 3.

4.5 Performance Criteria at Serviceability Limit State

135
The building shall satisfy the fallowings

at design earthquake load for

serviceability limit state specified in

Section 4.2.1;

(1) No flexural yielding shall occur in

structural members, and

(2) Story drift angle at each story shall

be less than 1/200 rad.

4.6 Performance Criteria at Ultimate

Limit State

4.6.1 Performance Criteria at Design

Limit Deflection

At the maximum story drift angle of a

story reaching design limit deflection of

Rul rad, the story shear at each story

shall exceed 0.9 times the required

lateral load resisting capacity. The

design limit deflection, Rul, shall be as follows:

Frame structure: 1/100 rad.,

Wall-frame structure: 1/100. 1/120, or 1/150 rad.,

in which structural walls resist up to 30%, 30 to 70%, and over 70% of the

base overturning moment at design limit

deflection Rul, respectively, (refer to Fig.l)

4.6.2 Performance Criteria at Design

Proof Deflection

At the maximum story drift angle of a

story reaching a design proof deflection

of Ru2 rad., the structure and structural

members shall satisfy the fallowings;

where, the design proof deflection in

terms of story drift angle shall be given

below:

Frame structure: 1/50 rad.,

Wall-frame structure: 1/50, 1/60, and

1/75 rad.,

in which structural walls resist up to 30%, 30 to 70%, and over 70% of the

base overturning moment at design

proof deflection Ru2, respectively.

(1) The story shear at each story shall

exceed the required lateral load carrying

capacity specified in Section 4.2.2,

(2) The deformation at a specified yield

hinge of each structural member shall

not exceed the deformation capacity, and

(3) The resistance other than the

specified yield hinges in each structural

member shall be greater than the design

force amplification factors a times the

stress calculated at the design proof

deflection. The design force

amplification factors shall be as follows;

a) ai for shear and bond stress of

girders,

b) a2 for flexural moment of

columns,

c) a3 for shear and bond stress of

columns,

d)M for flexural moment of walls,

e) a5 for shear of walls,

f)a6 for shear in beam-column

connections,

(4) Axial force in columns shall be less

than 2/3 Nu and 3/4 Nt, where Nu:

ultimate compressive strength of the

column, and Nt: ultimate tensile

strength.

4.6.3 Limit of Axial Stress in Exterior

Columns

Axial force in corner columns under bidirection earthquake loading shall be less

than 2/3 Nu and 3/4 Nt.

4.6.4 Design of PCa Connections

Design stresses for PCa connections

shall be equal to the stress developed at the connections calculated at the design

proof deflection Ru2 multiplied by

amplification coefficients specified in

4.6.2. Design of PCa connections shall

conform to the provisions of "PRESSS

Design Manuals for PCa Connections."

5 DEMAND TO CONNECTION

PERFORMANCE

The equivalent monolithic connection

has three subcategories including

(1) equivalent monolithic strong

connection (EMSC),

(2) equivalent monolithic ductile

connection (EMDC),

(3) equivalent monolithic extensible

connection (EMEC).

These are sumarized as shown in Table

2. The object of this chapter is to give

quantitative definition to the performance

of good connection designed by

emulation approach.

The demand of equivalent monolithic

connection is summarized as follows;

a) Design moment and/or forces for

connections and precast elements can be

predicted by the structural analysis

generally used for monolithic system.

136
b) Connections are strong enough to transfer the design moment and/or

forces. If the connection failure or yielding occurred in early stage of

loading, unexpected failure or yielding

mode of precast system will take place.

To avoid such change of structural

performance of precast system,

connections should have at least comparable strength to equivalent

monolithic system.

c) Serviceability, durability and fire resistance are as same as in monolithic

system.

Specified conditions for monolithic

system are also applied to precast

system. Items to be considered are deflection, crack width, creep and

shrinkage of concrete and mortar,

concrete cover thickness, corrosion

protection, fire protection, mechanical

properties of connection materials under

high temperature

d) Structural performance of precast

system under earthquake loading is comparable to that of monolithic system. Precast system should have at least comparable deformation capability and same collapse mechanism to equivalent

monolithic system for lateral loading.

Maximum inter story drift angle of

precast system during an earthquake

should be within the limit specified for monolithic system even if some slip

deformation take place at connection

interface. Allowable limit of ship

deformation is given in terms of the ratio of the additional inter story drift angle

due to slip to the total inter story drift angle .

6. CONCLUSION

As a Japanese side for U.S. - Japan

Cooperative Research Program on

Precast Seismic Structural System

(PRESSS) , the research results were completed ; "Design Guidelines for Monolithic Reinforced Concrete

Buildings" "Construction Guidelines"

and "Design Manual for Precast

Connections". Since 1989, we , U.S.- Japan PRESSS Project Team aimed

most advanced methodology for the

targets of this project. That lead to really

fruitful discussion from the viewpoint of

research methodology.

Table 2 : Classification of Connection Performance for equivalent monolithic connection

Construction Method Type of Connection Description

strong connection rigid connection for non- (EMSC) yield part

equivalent monolithic ductile connection rigid connection for yield

(EMDC) part

extensible connection connection for non-yield

(EMEC) part

137
Storey Shear Coefficient

0.3

0.3x0.9

0.2

Load -Deflection Relationship of

ding

1/200

Reliable Curve

imate Limit Desii

oo

Ftui 1/50

*u2 R (Story Drift Angle)

Serviceabilty Limit Design

; any member does not yield

(a) Frame Structure

Story Shear Coefficient

0.40x0.9

PA>70% Ru1=1/150,Ru2=1/75

1/100 1/"'5 1/50 1/50

70%*PA>30%

Ru1=1/120,Ru2=1/60

PA*30%

Ruf=1/100

Ru2=1/50

p . wall shear capacity

total shear capacity

R(Story Drift Angle)

(b)Frame+Wall Structure

Fig.1 OUTLINE OF PRESSS SEISMIC DESIGN

138
A Post-Earthquake System for Estimating Seismic

Damage to Road Facilities by a Strong Motion Observation Network

by

Hisanori Otsuka", Ryoji Hagiwara2

' and Yoshio Ninomiya3)

ABSTRACT

The Ground Vibration Division of the

Public Works Research Institute studies

procedures for efficiently inspecting road

facilities immediately after an earthquake.

This paper presents a system for estimating

seismic damage to bridges using data from

real-time-transmitting accelerographs

stationed in a strong motion observation

network.

KEYWORDS: Strong Motion Observation

Network, Seismic Damage, Spectrum

Intensity, Earthquake Hazard Information

1. INTRODUCTION

Japan largely depends on road traffic for

transportation of people and commodities.

Roads also play an important role in providing lifelines and supporting urban

activity.

Thus, after an earthquake, immediate

repair and restoration are required. In the

aftermath of an earthquake, many road

administrators such as the Ministry of

Construction, local governments and so forth,

require employees and contractors to inspect

road facilities, direct traffic, establish detours

and restore damaged facilities. Earthquakes in large cities pose special

problems since roads are congested and

roadsides are densely built-up. Therefore,

immediate traffic recovery and restoration of

road facilities are of special concern there. This requires a reasonable number of

accelerographs to be taken. Estimating

seismic damage to road facilities immediately

after an earthquake would allow inspection of

road facilities to be conducted efficiently.

The Work Office of the Regional

Construction Bureau, Ministry of

Construction plans to install about 10

accelerographs and establish the first phase of

a Strong Motion Observation Network in this

fiscal year. This paper will discuss a system for

estimating both ground vibration

characteristics at random points and seismic

damage to reinforced concrete piers.

Both recorded accelerograms and

observed damage to road bridge piers, which

occurred during the 1993 Kushiro-Oki

Earthquake are used to check the validity of

this system.

2. SYSTEM FOR ESTIMATING SEISMIC

DAMAGE

Figure 1 shows flow of information.

When an earthquake occurs,

real-time-transmitting accelerographs record

strong motion and transmit the data to a Work Office immediately. The Work Office measures ground vibration characteristics

such as maximum acceleration and Spectrum

Intensity from all administered

accelerographs, and estimates seismic

1) Head, Earthquake Engineering Division,

Public Works Research Institute,

Ministry of Construction, Tsukuba City,

Japan

2) Head, Ground Vibration Division, ditto

3) Senior Research Engineer, ditto

139
damage with an Engineering Work Station

(EWS). The estimated results are shown in the EWS at the Work Office and used for the

inspection and so forth. The allowable

timelag from occurrence of an earthquake to display of estimated seismic damage is considered to be only several minutes.

Figures 2, 3 and 4 show examples of

display in the EWS, i.e., epicenter, locations

of seismic-damaged road facilities and a liquefaction hazard map, respectively.

To establish a system based on

strong-motion data, map data, road facility

data and ground data, we designed a program

to calculate the following and display the

results at the EWS:

(1) Epicenter location

(2) Horizontal maximum acceleration and

Spectrum Intensity at each mesh in a map

and important road facilities

(3) Liquefaction potential of meshes in a map

(4) Seismic damage to road facilities This paper presents methods to calculate

both Spectrum Intensity at random points,

and the degree of seismic damage to reinforced concrete piers of bridges.

3. ESTIMATION OF GROUND

VIBRATION CHARACTERISTICS AT

RANDOM POINTS

Strong motion data in the estimation

system is transmitted immediately from

accelerographs to an EWS at a Work Office.

To produce useful information from these

data, we need to estimate ground vibration

values which can estimate the seismic damage

to each type of road facility. Since it is well

known that Spectrum Intensity (SI) has a high correlation to damage to road bridge

piers, the authors adopted SI to estimate such

damage. Three methods to calculate SI are presented and observed SI values are compared. Strong motion data observed at 12 points in the 1993 Kushiro-Oki

Earthquake are used to calculate the observed

SI values.

Table 1 and Figure 5 show observed SI and strong motion observation points

respectively.

Three different methods to calculate SI

are as follows.

3.1 A method based on attenuation

formulae using a strong motion data base

SI can be represented by the following

formula according to the definition proposed

by Housner

3)

:

2.5

J0.1 f

SI= f Sv(T, h=0.2)dT- (1)

where SI : spectrum intensity (cm)

Sv : velocity spectrum (cm/sec)

T : natural period (sec)

h : damping factor

SI Attenuation formulae are obtained

using 394 components of horizontal strong

motion shown in the Seismic Design of

Specification for Highway Bridges

4)

. When

ground classification and formula patterns

are the same as those used in the

specifications

55

, the following formula can be

obtained:

{SI(M, A, GCi)}=

65.056 X 10

0.2557M

7.181 X 10

0.4398M

11.341X10

0.4241M

-1.224 X A+30 <2)

where M : magnitude of an earthquake

A : distance from the epicenter

GCi : ground type (i=l,2,3) indicated in Specifications for Highway

Bridges

SI at a random point is calculated by the

formula (2) for the ground type at the point.

3.2 A method based on attenuation

formulae using observed strong motion data

Since the formula in 3.1 are based on

data from all of Japan, the dispersion of data

140
is large. Therefore, SI should be calculated

by a method considering characteristics of

both the earthquake and the local ground

conditions. Thus SI attenuation relationships

are formulated based on the data obtained in the Strong Motion Observation Network. The

procedure to derive the new formulae is the same as 3.1.

3.3 A method for calculating directly from

observed strong motion data

When the epicenter and the magnitude

of an earthquake are known, the methods

presented in 3.1 and 3.2 can be used. But SI

at any point can be directly calculated

without those methods from strong motion

data from observation points. SI is calculated

considering the distances from the calculation

point to observation points, based on the following formula :

Sip =SI + ^2 ^(Sli - Sl)jrj y

• (3)

where Sip : SI at the point concerned (cm)

SI : The average SI of all observation points (cm)

n : The number of observation

points used in the calculation

Sl i : SI at observation point i L : The average distance from all observation points to each

calculation point

Li : Distances from all observation

points to one particular point

In this paper, n equals 11, SI and L are the average value of 11 points excluding the

point concerned. Since the number of

observation points is relatively small, the

values obtained using formula (3) are multiplied by 0.76 for ground types D and III

.

Here, 0.76 was derived from the relation

between the attenuation formulae in (2). Thus

in this paper, the third method had to use information concerning the epicenter and the

magnitude of the earthquake. If we have

enough observation points for each ground

type, we do not need such information.

3.4 Comparison of 3 methods

Figure 6 shows observed SI and the

results calculated by the methods proposed in

3.1, 3.2 and 3.3. Where (a): observed values and attenuation curves

(b): points plotted in coordinates

of observed values and

calculated values, and lines drawn

as observed value = calculated

value.

Figure 6 (a) indicates that the

attenuation curve of (1) in 3.1 for the ground

type III is the largest, following by II and I (I < n < HI ). However the difference

between n and HI is small. Major observation

values are plotted above those curves. Because the curves of 3.2 consider

characteristics of the earthquake and the local

ground conditions, it is clear that the

attenuation curves coincide with observation

values.In Fig. 6(b), almost all points due to 3.1

are below the solid line: i.e. the calculated

value is smaller than the observed value

shown in Fig. 6(a). The solid line is located in roughly the center of the points derived in

methods of 3.2 and 3.3. Calculated

systematic error and dispersion are small. In method 3.3 only distance from an

observation point to a point concerned is considered, but in method 3.2, the direction

to the epicenter is also considered, thus it can be said that the latter is better than the

former. However, it is difficult to evaluate

characteristics of the local seismic

amplification at any ground in the method of

3.2. On the other hand, the method of 3.3 can calculate automatically the ground

vibration characteristics of the observation

point; accordingly, SI can be accurately

calculated when the point concerned is

141
similar to neighboring observation points in ground seismic amplification. This method

can be improved to reflect more local

characteristics based on the accumulation of

strong motion data and ground seismic

amplification data. In addition, the fact that

this method does nc f require the information

of the epicenter and magnitude of the

earthquake is a great advantage, because this system must automatically calculate SI values

immediately after an earthquake.

4.CALCULATION OF SEISMIC

DAMAGE TO ROAD BRIDGE PIERS

The authors estimated seismic damage

to 10 road bridge piers using strong motion

characteristics (SI) calculated by the method

proposed in 3.3. Road bridges damaged by

the 1993 Kushiro-Oki earthquake are used.

The difference between the degree of actual

damage to piers due to the Kushiro-Oki

Earthquake and the degree of estimated

damage is considered.

4.1 Seismic damage and characteristics of

objective bridges

10 bridges were selected considering

degree of damage (3 class) and ground type

(3 types) among damaged bridges located on

national highways and prefectural roads.

Figure 7 and Table 2 represent structural

characteristics and degree of seismic damage

of these bridge piers, respectively.

As shown in Figure 7, severe damage

was caused by shear cracking at the ends of

reinforced bars in reinforced concrete

circular-column piers.

4.2 A method for earthquake response

analysis

Table 2 also shows estimated SI and

estimated ductility factor [i. Ductility factor [i

is defined as the ratio (S/oy) of response

displacement at the top of the pier 6 to the

yield displacement at the same point Sy. An

Analytical model of a reinforced concrete

pier is assumed to be a 1-mass system (see

Fig. 8). Subjecting this model to earthquake

motions of 3 different SI values (30, 60 and

150 cm), provided the ductility factors are shown in Fig. 9. Input motions for earthquake

response analysis were adjusted in amplitude

to attain the target values of SI. Estimated [l values in Table 2 were thus obtained for each

pier.

4.3 Comparison of actual damage and

estimated damage for study bridges

Figure 10 shows analytical results as components of both ductility factors and

allowable ductility factors. This figure

indicates that wall-type piers were not

damaged when \i was less than 4, but

circular-column piers were damaged even at a

(x value of 1.5. Theoretically, circular column

piers are supposed to be more resistant to earthquake-induced damage, but this not

actually the case. According to this figure,

circular-column piers were damaged at p4ia

^0.7 (solid line), in which \ia is the allowable

ductility factor.

5.CONCLUSION

We have examined system for estimating

seismic damage to road facilities by a strong

motion observation network. In this paper,

methods for both calculating spectrum

intensity at any point and evaluating the

degree of seismic damage to reinforced

concrete piers were considered. The

following results were obtained:

(1)The methods of estimating SI at random

points using attenuation formulae based on observed strong motion data and

considering the weighted distances show a good coincidence of observed values.

(2) For damage to piers, earthquake

response analysis using Si-adjusting input

motion was conducted. As a result of

these calculations, the relationship between

142
SI and the ductility factor is clarified. Therefore, if estimated SI values are known

after the earthquake, ductility factors can be obtained without complicated response

analysis.

(3)Wall-type piers are damaged when [x is more than around 4 and circular-column

piers are damaged when m/M^— 0.7.

6. ACKNOWLEDGEMENTS

This study was conducted by PWRI and

Kanto Regional Construction Bureau. The

authors wish to express their thanks to the

Hokkaido Development Bureau and the

Hokkaido prefectural government for

collecting the material.

Ninomiya Y. Tokida K, Otsuka H:

Response Analysis for Ground Vibration

Characteristics and Damage to Structures The 22nd JSCE Earthquake

Engineering Symposium, 1993.6

(in Japanese)

Otsuka H, Yoshimi S: Chap.l Earthquake and Ground

Vibration, Report on the Disaster

Caused by the 1993 Kushiro-Oki

Earthquake, PWRI, 1994.6 (in Japanese)

G.W. Housner: Behavior of Structures

During Earthquakes ASCE, 1959.10

Japan Road Assosiation: Specifications

for Highway Bridges part V : Seismic Design, 1990.2

Kawashima.K, Aizawa.K,

Takahashi.K: Attenuation of Peak

Ground Motions and Absolute

Acceleration Response Spectra,

Report of The Public Works Research

Institute, Ministry of Construction,

Vol.166, 1985.9 (in Japanese)

143
Table-1 Observation SI

No. Observat ion Point

ftpnnnd Ul \j U II U Type

Uni<;tanrp I olalilC frnm 1 l U III the epicenter O I Observation

ii T

1 01 R9 Q

9 nui Uttta u or luge T

1 111 141 1 9 ^

9o flfarmchiVo RriH cro ULdllUbnirtC 01 1 Ugc TTU on 09 R

A4 Tchtlrsri-lriVn Rri Hero

ill 1 Q 9 iy. c

0 OUdTl oriugc TT 113 7A A

i 4. 4

R0 un uc 1 on or 1 ugc TT11 milul yi. 6

7 T1 1U0 ac\ r 4U. 0

Q0 on l Es^Sd 1 so ZdWd or 1 us" TT

11 9T1 £01 ID. 0

Q

a Qinnnrn TP V/larhirf TT11 64i AO 7

flhKn Ull l 0 U Rank OdllK nrILL uo 100, 0

11 Sarukawa Bank n 197 66.5

12 Mukawa Bank 1 200 62,4

Table-2 Specification and damages of the study bridges

Degree of Damage High Kay Route Tow

Bridge Foundation TypeObjective pierGround

Type Distance fron

the epicenter Atta)

Estimated S I (cm)

Estimated U

Damage

1 Prefectaral 8osd

Taiki-flHdil Yoda Bridge Pile P, Clrcalar

colore n 82 911 466 Fractnre of Reinforcement(Cut-off Point)

Severe Damage 2 National High fey 240 Atan-machl

Uatsrcnoe Bridge Caisson Pz Circular

column

1 38 70.2 L40 Buckling of Reinforcement

(Cut-off Point)

3 Prefectural Joad

Onbetso-Bacai

Hatsnne Bridge Pile P, Circular

colore n 42 9L7 L83 Shear Cracking

(Cat-off Point)

Uinor Damage 4 Ualn Prefectarsl load Shibectu-nadil

Shintasa

Bridge Spread Pi fell

type

1 56 62.5 4.38 Cracking

5 National High fey 391 Shibecha-machl Goilkkoku

Bridge Caisson P. Clrcalar coIm

n 46 84.2 211 No- Damage

6 Prefectaril 2o»d

Akan-oachl Assaaugna

Bridge Spread P. Clrcalar

co lure

i 36 721 L44 No-Damage

No-Oaange

7 national High toy 38 ShiranuJa-aachl

Xnltol Bridge Spread P. fell

type

1 23 85.5 142 No-Damge

8 .National High fey 44 Neauro-cltyCn-netoh

Bridge Caisson P. fell

type n 101 sai L38 | No-Damage

9 National High fey 44 Knshiro-machlTxlodai Bridge Pile P, fell

type i 18 95.2 133 ; No-Damage

10 .National High fey 38 Knshlro-cityQtanoshlke

Bridge Pile P, fell

type n 20 75.5 3.40 No-Damage

144
Table-3 Input motion used for analysis

Ground

Type

Observation Values Input Motion

Name

Distance from the epicenter

Amax

(gal)

CT (cm)

INO. Amax

(gal) (cm)

I Chiyoda Bridge LG

A 01

1

I - 1 159.7 30

345 62.9 I - 2 493.4 90

I -3 822.3 150

n Otanoshike Bridge LG

l\ =zUKm

n -

1

147.8 30

456 92.6 n -2 443.4 90 n -3 739.0 150

m Ohtsu Bridge TR

A =83km

m - i 63.3 30

314 135.8 m - 2 207.9 90

in - 3 346.5 150

Strong Motion Observation Point

Real-Time— Transmitting Accelerograph

Branch Work Office

Engineering Workstation

5> Neighboring Work Offices

Engineering Workstation Engineering Workstation

Fig-1 Flow of Information

145

Fig-4 Liquefaction hazard map

Legend

Structure, Ground Type, SI (Degree of Damage, Pier Type, Ground type, Estimated SI)

© Bridge, IB, 19.2

Viaduct, II ,42.7

Bridge, n,15.6

Bank, 111,62.4

Bank, 11,66.5

!^^jTSS8® Bndg, 0,91.2

'©Bridge, I , 62.9

Bank, 111,135.8

1 (Severe Damage, Circular column, II , 91.7)

©Bridge, I ,40.6 Bridge, I , 12.5

Fig-5 Strong motion observation points

147
0 Grouod Type I

"j O Ground Type II > (Observation) Ground Type E

10

Ground Type II , M

(Attenuation Curve in 3.2) ^/Ground Type 1

Legend

Ground Type I Ground Type II Ground Type ITJ

(Attenuation Curve in 3.1)

I

i I

I

I I

I

3 4 5 6 7 8 9100

Diatance from the epicenter A (km)

(a) The comparison between observed SI and predicted SI

Fig-6 The comparison of observed SI to estimated SI

148
Epicenter

+

Fig-7 Road bridge points for seismic response analysis

777777

P : Horizontal Force 6 : Horizontal Displacement

C : Crack point

Y : Yield point

U : Ultimate point

Fig-8 Analysis model

149
} 0 50 100 150 200

Estimated SI

Fig-9 Relation between ductility factor and Spectrum Intensity

(Hatsune bridge)

•1 /

ah/

1 J \ Legend

Circular Column

Not Damaged

e Damaged

*

Wa 1 1 Type

Not Damaged

a Damaged

i D If! i ;/ i * / D \

1

0 2.4 6 8 10 12 14 16 18 20 M a=H(5u-5y)// 1.55y

Fig-10 The relation between real and analytical damage

150
Use of Probabilistic Earthquake Models

for Design and Evaluation

by

Jeffrey K. Kimball* and Aiin Bieniawski*

ABSTRACT

This paper outlines the use of

probabilistic seismic hazard models

to establish seismic design and

evaluation requirements consistent

with the earthquakes which control

the seismic hazard. Current

probabilistic seismic hazard curves

for the Eastern United States

indicate that the size of the

controlling earthquake(s) is

generally less than magnitude 6 at a distance within 50 kilometers of the

site. This information has

implications regarding the response

spectral shape which should be used

for the design of new facilities and

the evaluation of existing

facilities. For the majority of

Eastern United States sites with

rock site conditions the use of the

NEHRP (1991) design spectral shape

will overpredict ground motion for

frequencies less than about 5 Hertz.

This paper suggests that a separate

design earthquake spectral shape

should be developed for the Eastern

United States that is consistent

with the magnitudes and distances

which control the seismic hazard.

KEYWORDS: seismic hazard; spectral

shape; response spectra; controlling

earthquakes; seismic design.

1. INTRODUCTION

The purpose of this paper is to

outline how probabilistic seismic

hazard models can be used to

establish seismic design and

evaluation requirements consistent

with the earthquakes which control

the seismic hazard. The Department

of Energy (DOE) has established

seismic design requirements based on the use of target performance goals

which are expressed as the mean

annual probability of exceedance of

acceptable behavior limits of

structures and equipment due to the

effects of earthquakes. The overall

DOE approach is described in the

proceedings for the 25th Joint

Meeting on Wind and Seismic Effects

in the paper of Hill and Murray

(1993). The seismic criteria

utilize probabilistic seismic hazard

curves for specification of

earthquake loading combined with

deterministic response evaluation

methods and permissible behavior

limits. As discussed in Hill and

Murray (1993), the Design Basis

Earthquake (DBE) is established by

selecting a probabilistic mean peak

ground acceleration used to scale a median deterministic response

spectrum shape. One reason for this

choice is the controversy that

exists regarding the shape and

amplitude of the probabilistic

Uniform Hazard Spectra for all

spectral frequencies.

The median deterministic DBE

response spectrum shape should be

consistent with the expected

earthquake magnitudes, distances and

the site conditions for the site in

question. Current DOE seismic

design requirements, as discussed in

Hill and Murray (1993), specify two

available choices in the selection

of the response spectrum: The first

choice involves the use of standard

spectral shapes such as that

U.S. Department of Energy

Washington, DC 20585

151
developed by Newmark and Hall

(1978); the second and more preferred choice involves the

development of a site-specific

spectral shape. The impact of the

choice of spectral shape on design

and evaluation is not trivial. The

spectral shape, along with the peak

acceleration, can determine how a structure is designed and can impact

decisions regarding which existing

structures require strengthening to

meet modern seismic design

requirements. The probabilistic

seismic hazard curve can be used to

aid in making the selection of the

spectral shape to be used for design

and evaluation. The discussion

below describes one approach for

using probabilistic seismic hazard

information to assist in selecting

the spectral shape for Eastern

United States rock sites.

2. PROBABILISTIC SEISMIC HAZARD:

CONTROLLING EARTHQUAKES

As discussed in Kimball (1992), the

general purpose of a probabilistic

seismic hazard analysis is to

evaluate the vibratory ground

shaking at a particular site by

considering the viable sources of

earthquakes, the likelihood of

earthquake occurrences for each

source, an estimate of the

particular ground motion parameter

of interest, and an analytical model

for estimating the probability of

exceeding the ground motion

parameter. State-of-the-art

probabilistic seismic hazard

analysis allows the use of multivalued models and alternative

hypotheses, with the results

displaying both the central value

and the full distribution of hazard

results. The probabilistic seismic

hazard analysis can be used to

perform sensitivity studies which

determine the significance of the

various input parameters.

Additionally, the identification of

the significant seismic sources in

terms of magnitude and distance can

be quantified.

Comprehensive probabilistic seismic

hazard studies have been completed

by Lawrence Livermore National

Laboratory (LLNL) for the Department

of Energy and the United States

Nuclear Regulatory Commission.

Figure 1 displays the peak ground

acceleration mean seismic hazard

curve for a DOE site near Kansas

City, Missouri, using the LLNL

methodology. Figure 1 shows the

total seismic hazard, and the

contribution to the seismic hazard

for earthquakes between magnitude 5 and 6, and above magnitude 6.

Figure 1 shows that the peak

acceleration seismic hazard is based

on about equal contributions from

earthquakes above and below

magnitude 6.

The probabilistic seismic hazard

result can also be displayed by

magnitude and distance bins. Figure

2A shows the seismic hazard results

for peak ground acceleration using

the Electric Power Research

Institute (EPRI) seismic hazard

methodology for a site in South

Carolina, for five magnitude and six

distance bins with the percent of

seismic hazard contribution shown

for each bin. The total for all of

the bins would sum to one. Figure

2A shows that the peak acceleration

is dominated by magnitudes less than

6 at distance less than about 50

kilometers. Figure 2B shows a similar plot in terms of 1 Hertz

spectral acceleration. In this case

the seismic hazard is dominated by

magnitudes 6 to 7 at distances

between 50 and 200 kilometers.

The seismic hazard results can also

be used to derive the average

magnitude(s) and distance(s) which

controls the seismic hazard curve. The average magnitude(s) and

152
distance(s) can be considered as defining the probabilistic

controlling earthquakes; those

earthquakes that have the greatest

influence on the ground motion at

the site in question. Given the

fact that different magnitude(s) and

distance(s) may control the high and

low frequency content of the

probabilistic seismic hazard

estimate (as shown by Figures 1 and

2), the assessment of controlling

earthquakes should be completed for

both high and low ground motion

frequencies. The discussion below

shows how the probabilistic derived

average magnitude(s) and

distances(s) can be used to estimate

design spectra for frequencies

greater than 1 Hertz.

A compilation of average magnitudes

and distances is available from LLNL

for the location of 69 commercial

nuclear power plants in the Eastern

United States. These estimates were completed by determining the average

magnitude and distance which

controls the probabilistic seismic

hazard curve for the average of 5 and 10 Hertz spectral acceleration

(the high frequency portion of the

spectra), and the average of 1 and

2.5 Hertz spectral acceleration (the

low frequency portion of the

spectra). Summary results from this

compilation are shown on Tables 1 and 2, which display the number of

reactor sites which fall within each

magnitude/distance bin, with the

first number being associated with

the average of 5 and 10 Hertz and

the second number being associated

with the average of 1 and 2.5 Hertz.

Table 1 shows the dominant

magnitudes and distances for the

LLNL mean seismic hazard data while

Table 2 is based on the LLNL median

seismic hazard data. Table 1 shows

that 15 of 69 reactor sites had a dominant magnitude of 5.5 with 7 of

the 15 being in the distance range

of 35 to 39.9 kilometers for the

high frequency (average of 5 and 10

Hertz) portion of the hazard curve.

The information contained on these

tables can be visualized using the

magnitude/distance bin graphs shown

on Figures 3A through 3D.

Tables 1 and 2 and Figures 3A

through 3D can be used to reach the

following conclusions. Generally

magnitudes less than about 6

dominate the seismic hazard at the

reactor sites for frequencies

greater than 1 Hertz. There is a slight shift towards larger

magnitudes and greater distances

controlling the lower frequency

portion of the ground motion but

this shift does not appear to be

very significant. In reviewing the

individual reactor sites the shift

towards larger magnitudes at greater

distance dominating the lower

frequencies is most significant when

the site is within about 100 to 200

kilometers of larger (M>6.5)

historic earthquakes such as the

1886 Charleston, South Carolina

event or the 1811-1812 New Madrid,

Missouri events. This trend is

evident on Figure 2 with the EPRI mean seismic hazard curve as well.

The result shown on Figure 2 is for

a site that is about 145 kilometers

from the 1886 Charleston South

Carolina earthquake. In terms of

determining the dominant distances,

the seismic hazard appears to be

controlled by earthquakes within 45

kilometers for the mean hazard and

about 25 kilometers for the median

hazard.

3. CONTROLLING EARTHQUAKES:

IMPLICATIONS FOR THE DESIGN

As indicated above for the Eastern

United States the controlling

earthquakes are generally less than

magnitude 6, for distances less than

about 25 to 45 kilometers. With

respect to seismic design

implications it is instructive to

153
compare the response spectra from

the controlling earthquakes to the

response spectra typically used for

design. Figure 4 displays two

typical design earthquake response

spectra based on the 1991 National

Earthquake Hazards Reduction Program

(NEHRP) Recommended Provisions

(NEHRP, 1991) and Newmark and Hall

(1978) for rock site conditions.

The median and one standard

deviation spectral shapes are shown

for the Newmark and Hall (1978)

response spectra. As Figure 4

shows, the NEHRP (1991) rock

spectral shape, which is also the

spectral shape utilized in the 1991

Uniform Building Code (UBC, 1991),

is between the median and one standard deviation Newmark and Hall

(1978) rock spectra for frequencies

above 2 Hertz and about equal to the

median Newmark and Hall spectra for

lower frequencies. Figure 4 also

shows that the 1 Hertz spectral

acceleration is equal to the peak

ground acceleration for the NEHRP

spectral shape and the median

Newmark and Hall spectral shape, and

about 1.35 times the peak

acceleration for the one standard

deviation Newmark and Hall spectral

shape.

How do the NEHRP (1991) and Newmark

and Hall (1978) spectral shapes

compare to the spectral shapes from

the controlling earthquakes? Tables

1 and 2 can be used to select

magnitude/distance pairs to compare

to standard design spectral shapes.

Table 3 displays eight sets of

magnitude/distance pairs that

represent the information displayed

in Tables 1 and 2. Table 3 also

lists three Eastern United States

and three Western United States

ground motion attenuation models to

derive the spectral shapes for the 8

magnitude/distance pairs. The

primary reason to use both Eastern

and Western United States

attenuation models is to quantify

the relative difference in spectral

shape between the two geographic

regions, recognizing that the data

base on which the NEHRP (1991) and

Newmark and Hall (1978) spectral

shapes were based is exclusively

Western United States data. The

spectral shape for each of the

magnitude/distance pairs will be

quantified using each of the

attenuation models, with the results

being normalized to a peak ground

acceleration of lg to compare with

the NEHRP (1991) and Newmark and

Hall (1978) spectral shapes. The

normalized spectral shapes for the

eight magnitude/distance pairs for

the Eastern and Western United

States attenuation models are shown

on Table 4 in terms of relative

acceleration amplification factors

at seven ground motion frequencies.

The results shown on Table 4 can be

used to derive a spectral shape for

both the East and West. Figure 5

shows the envelope of the data from

Table 4 for the two geographic

regions. From a relative spectral

shape perspective, the Eastern

United States spectral shape is

enriched in the high frequency

region when compared to the Western

United States. Considering the

uncertainties in characterizing

Eastern United States earthquake

source spectra and ground motion, an approach would be to develop a "composite" spectra based on the two

individual spectra shown on Figure

5. In developing the "composite"

spectra the following considerations

were utilized. First, an attempt

was made to maintain the broadness

of the spectral shapes shown on Figure 4, with the maximum spectral

acceleration occurring over about a factor of four in frequency.

Second, an attempt was made to

maintain the general spectral shapes

shown on Figure 5 recognizing that

to match the Eastern United States

spectral shape a shift in frequency

154
would be necessary. Finally, an attempt was made to conservatively

reflect the values for the maximum

spectral acceleration based on the

data shown on Table 4.

The "composite" spectral shape

derived from Table 4 is shown on Figure 6 compared to the NEHRP

(1991) spectral shape and the median

Newmark and Hall (1978) spectral

shape. The "composite" spectral

shape has a maximum spectral

acceleration of 2.4 between

frequencies of 5 to 20 Hertz,

retaining the broadness shown in

Figure 5, shifted to higher

frequencies to reflect Eastern

United States earthquake source characteristics. For frequencies

less than 4 to 5 Hertz the

"composite" spectra is significantly

less than the NEHRP (1991) and

Newmark and Hall (1978) spectral

shapes. The primary reason for this

discrepancy pertains to the

magnitudes of the earthquakes used

in this analysis compared to those

that were used to derive the

standard spectral shapes shown on Figure 4.

In an attempt to explore the

differences between the "composite"

spectra and the Newmark and Hall

(1978) spectra, the actual data base

that was used to develop the Newmark

and Hall spectral shape was reviewed

(see Newmark 1973, Newmark et al

1973 and Hall et al 1976). Newmark

and Hall (1978) compiled a set of

representative strong motion data

which was normalized to lg peak

ground acceleration, and statistical

analyses were performed to derive

the spectral amplification

functions. Retrospectively looking

at the data base used, the majority

of data pertains to soil sites which

recorded strong motion from

earthquakes in the magnitude range

5.3 to 7.7, with about two thirds of

the data from earthquakes larger

than magnitude 6. Data was from

soil sites because Newmark and Hall

recognized that the rock data was

very limited and could represent

biased statistical results.

Newmark and Hall (1978) also

developed recommendations for the

ratio of velocity to acceleration

(V/A) for both rock and soil. In general the peak velocity region of

the response spectra is associated

with frequencies of about .5 Hertz

to 5 Hertz, while the peak

acceleration is associated with the

high frequency portion of the

response spectra.

The velocity to acceleration ratio

for the strong motion data from soil

sites from Newmark (1973) are shown

on Figure 7A as a function of

earthquake magnitude, along with the

best fit correlation between

magnitude and V/A. Figure 7A also

lists the recommended value for V/A

of 121.9 cm/sec/g from Newmark and

Hall (1978) which intersects the

best fit line at about magnitude

6.5. Figure 7A indicates that the

ratio of V/A is a function of

earthquake magnitude, with larger

earthquakes providing more lower

frequency energy relative to smaller

earthquakes

.

While the Newmark (1973) data base

was not sufficient to make a similar

plot for rock sites, the strong

motion data base for pre-1972 has

been searched to identify all rock

records within 100 kilometers of

earthquakes above magnitude 5. The

rationale for selecting pre-1972

data was to keep the data as close

to that available in defining the

original Newmark and Hall (1978)

spectra as possible. Figure 7B

shows the V/A for all rock records

above magnitude 5 plotted as a function of magnitude. Figure 7B

also plots the V/A derived from the

Campbell (1990) attenuation model

155
for rock like sites for distances of

5 and 100 kilometers. The Campbell

V/A lines tend to bracket the pre1972 recorded rock data and show a similar magnitude dependent trend

when compared to Figure 7A.

However, the V/A for rock sites is

substantially lower than soil sites

reflecting the general trend that

soil sites amplify ground motion in

the peak velocity region compared to

the peak acceleration region (see

Borcherdt 1994).

Given the fact that the controlling

earthquakes for the Eastern United

States are generally lower than

magnitude 6, Figure 7B can be used

to justify a lower spectral shape

compared to either the NEHRP (1991) or Newmark and Hall (1978) response

spectra for frequencies less than

about 5 Hertz for use in the Eastern

United States. It is beneficial to

point out that the relative

underconservatism of the NEHRP

(1991) spectral shape in relation to

the composite spectral shape for

frequencies greater than 10 Hertz

does not actually exist in design.

The NEHRP Recommended Provisions

uses a normalized lateral design

force coefficient which utilizes the

maximum spectral acceleration value

for the very low period portion of

the spectra instead of the actual

value. For controlling earthquakes

between magnitude 5.3 and 6 from

Tables 1 and 2, the rock V/A from

Figure 7B ranges from about 20

cm/sec/g to 80 cm/sec/g,

substantially less than the 91

cm/sec/g implied by the NEHRP (1991) or Newmark and Hall (1978) response

spectra. The composite spectra

shown on Figure 5 has an implied V/A

of about 60 cm/sec/g which is

consistent with the empirical trend

shown on Figure 7B. A similar

conclusion would be applicable for

soil sites for earthquakes less than

magnitude 6 based on the data shown

on Figure 7A.

4. DISCUSSION

The above discussion indicates that

the design earthquake response

spectra shape referenced in the 1991

version of the NEHRP Recommended

Provisions may not be appropriate

for the Eastern United States

between frequencies of 1 and 5

Hertz, primarily because the

magnitudes of the earthquakes which

dominate the seismic hazard in the

east are smaller compared to the

magnitudes used to develop the NEHRP

(1991) design spectra.

The choice of design earthquake

spectral shape has implications for

the seismic design of both new and

existing facilities. If the goal of

the NEHRP Recommended Provisions is

to develop a consistent seismic

design process for the entire United

States, the basic seismic design

input must recognize differences in

the size of the earthquakes which

dominate the seismic hazard. For new buildings which have fundamental

frequencies between 1 and 5 Hertz,

the design of a building in the

East, on rock, will be more

conservative than the design of a building in the West, on comparable

rock, due to the inherent

conservatism in the NEHRP (1991)

spectral shape. As the NEHRP

Recommended Provisions are revised,

an attempt should be made to

determine if this current practice

should be continued, or as this

paper suggests, a separate design

earthquake response spectra be

developed for the east. Additional

work should also be completed to

determine if the trend identified in

this paper extends to frequencies

less than 1 Hertz.

The shape of the design earthquake

response spectra also has

implications for the seismic

assessment of existing facilities.

The United States Federal Government

156
will soon be assessing the seismic

safety of all existing Federally

owned and leased buildings. Because

seismic retrofit can be expensive

and because the Federal inventory is

quite large, this process must

include some type of prioritization

process for both the evaluation and

potential seismic strengthening for

existing Federal facilities. If the

design earthquake spectral shape

provides an inconsistent measure of

earthquake likelihood, then the

prioritization and mitigation

process may not identify or retrofit

the facilities with the highest

seismic risk.

5. CONCLUSIONS

The assessment of the probabilistic

seismic hazard can be used to

determine the magnitudes and

distances of the earthquakes which

dominate the probabilistic seismic

hazard. This information has

implications regarding the response

spectral shape which should be used

for the design of new facilities and

the evaluation of existing

facilities. Current probabilistic

seismic hazard curves for the

Eastern United States indicate that

the size of the controlling

earthquake is generally less than

magnitude 6 at a distance within 50

kilometers of the site. The

spectral shape from magnitudes less

than about 6 will not be the same as the spectral shape included in the

1991 NEHRP Recommended Provisions.

A separate design earthquake

spectral shape should be developed

for the Eastern United States that

is consistent with the magnitudes

and distances which control the

seismic hazard. Figure 6 shows one possible spectral shape developed

from both Eastern and Western

attenuation models for magnitudes

between 5 and 6 at distances less

than about 50 kilometers for

frequencies greater than 1 Hertz.

For the majority of Eastern United

States sites with rock site

conditions the use of the NEHRP

(1991) design spectral shape will

overpredict ground motion for

frequencies less than about 5 Hertz.

Additional work should also be

completed to determine if the trend

identified in this paper extends to

frequencies less than 1 Hertz.

6. REFERENCES

1. Atkinson, G.M., and Boore,

D.M. (1990). "Recent Trends

in Ground Motion and Spectral

Response Relations for North

America," Earthquake Spectra,

Vol. 6, #1, pg. 15.

2. Boore, D.M., et al (1993).

"Estimation of Response

Spectra and Peak Acceleration

from Western North American

Earthquakes: An Interim

Report," U.S. Geological

Survey Open File Report 93-

509, 72 pgs.

3. Campbell, K.W., (1990).

"Empirical Prediction of NearSource Soil and Soft-Rock

Ground Motion for the Diablo

Canyon Power Plant Site, San

Luis Obispo County,

California," Dames and Moore

letter report to Lawrence

Livermore National Laboratory.

4. EPRI, (1993). "Guidelines for

Determining Design Basis

Ground Motions," Electric

Power Research Institute, TR102293, 4 Volumes.

5. Geomatrix, (1992). "Seismic

Ground Motion Study for West

San Francisco Bay Bridge,"

Final report to Caltrans,

Division of Structures,

Sacramento, California.

157
6. Hall, W.J., et al (1976).

"Statistical Studies of

Vertical and Horizontal

Earthquake Spectra," Report for

U.S. Nuclear Regulatory

Commission, NUREG-0003, 140

pages

.

7. Hill, J.R., and Murray, R.C.

= ^ • "DOE NPH Mit igation

Policy and Standards Program,"

Proceedings of the 25th Joint

Meeting, UJNR, Wind and Seismic

Effects, ISSN 0386-5878, pg.

655

Report for U.S. Nuclear

Regulatory Commission,

NUREG/CR-0098, 49 pages.

14. Uniform Building Code (UBCh

(1991 Edition), International

Conference of Building

Officials, Whittier

Cal ifornia.

8. Kimball, J.K. (1992).

"Resolving Uncertainties in

Seismic Hazard Models,"

Proceedings of the 24th Joint

Meeting, UJNR, Wind and Seismic

Effects, NIST SP843, pg. 115.

9. McGuire, R.K., et al (1988).

"Engineering Model of

Earthquake Ground Motion for

Eastern North America,"

Electric Power Research

Institute Report, NP-6074.

10 NEHRP (1991 )

. "NEHRP

Recommended Provisions for the

Development of Seismic

Regulations for New Buildings,"

FEMA 222.

11. Newmark, N.M., (1973). "A

Study of Vertical and

Horizontal Earthquake Spectra,"

Report for U.S. Atomic Energy

Commission, WASH-1255, 151

pages

.

12. Newmark, N.M., et al (1973).

"Seismic Design Spectra for

Nuclear Power Plants," ASCE

Journal of the Power Division,

Vol. 99, No. P02, pg. 287.

13. Newmark, N.M., and Hall, W.J.,

(1978). "Development of

Criteria for Seismic Review of

Selected Nuclear Power Plants,"

158
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174
Structural Model of Functional Damage Due to an

Earthquake in Urban Area

by

Kazuhiko Kawashima", Hideki Sugita2

', Tomoru Nakajima31 and Takashi Sato

ABSTRACT

In urban area, functions of various facilities are closely related. Once a physical damage is developed to a facility due to an earthquake, this would cause loss or decrease of functions of the other facilities which are related with the facility

directly damaged by the earthquake. The propagation of functional damage is developed in this way, and this would be considered as a new type

of earthquake disaster especially in urban area.

This paper presents the structural model of functional damage due to an earthquake in urban

area. The structural model is formulated for ana- lyzing how the interaction between functions of

various facilities can be developed. Nineteen

functions including road, railway, water, electricity and telephone are considered in the analysis.

Based on the analysis, it is pointed out that the

facility most influential to the propagation of

functional damage is road.

KEY WORDS: Earthquake Disaster, Functional

Damage, Physical Damage, Propagation Path,

Structural Model

1. INTRODUCTION

In urban area of Japan, large scale earthquake

has not been occurred since 1923 Kanto Earthquake of magnitude 7.9. The metropolis in urban

area has been developed based on the highly

growth of economy after this event. Although the metropolis has not experienced large scale earth- quake, once it occurs, new type of earthquake

disaster which is completely different from the

disaster shown in Kanto Earthquake would be

developed. For upgrading the seismic safety of

urban area, it is quite important to figure out the

outline of new type earthquake disaster considering with the configuration of metropolis.

Following facts can be pointed out as the fea- tures of metropolis.

(1) Transportation systems and mass-communication systems have been highly developed.

Based on these developments, urban area has

been much enlarged. Breakdown of these sys- tems will make it impossible to maintain the activity of whole urban area.

(2) Computer systems for transmitting and analyzing information have been highly developed.

Dependency on computer systems has been in- creased in almost all industries. Breakdown of

these systems will make it impossible to maintain the industrial activities. (3) Lifeline systems such as road, railway, electricity, gas, water and telephone have been highly developed. Breakdown of these systems will make it impossible to maintain the life activities. (4) Those systems described above have highly

dependency on each other. Due to the interaction

of systems, function of facilities becomes possible to be affected by the earthquake damage of

the other facilities.

(5) Propagation of functional damage can be

developed simultaneously in whole urban area. Because of the complexity of propagation paths,

it becomes difficult to evaluate the total loss or

1) Planning Division, Public Works Research

Institute (Head, Earthquake Engineering Division)

2) Research Engineer, Earthquake Engineering

Division, Public Works Research Institute 3) Kita-sorachi Construction Office, Hokkaido

Development Bureau (Earthquake Engineering

Division, Public Works Research Institute)

4) Earthquake Engineering Division, Public

Works Research Institute

175
decrease of functions due to an earthquake.

Structural model of functional damage which

represents the propagation of functional damage

among fundamental facilities in urban area was

studied. Because the structural model can be var- ied with the scale of earthquake, the scale of

metropolis and the configuration of industries, it was assumed as a premise that Tokyo metropolis was attacked by Kanto Earthquake. Because emphasis of the study was placed on the propagation paths of earthquake disaster, brainstorming

method was adopted to figure out the interaction

among fundamental facilities qualitatively.

2. FUNCTIONAL DAMAGE DUE TO AN

EARTHQUAKE IN URBAN AREA

Fig. 1 shows the classification of fundamental

facilities of which industries of a metropolis are

consists. Fundamental facilities were classified

by modifying the industrial classification which

has been used for the input-output analysis of

industries. Because the damage of [Electricity,

Gas and Water], [Transportation and Communication] and [Public Service] facilities were sup- posed to have much influence on the propagation

of functional damage, they were classified into more precise facilities so that propagation paths

from each facilities could be shown in the structural model. On the contrary, [Agriculture], [Forestry], [Marine Products], [Mining], [Manufacture] and [Construction] facilities were summarized into [Production] because they were not supposed to have much influence on propagation

of functional damage. And because the facilities associated with life activities were not shown in the industrial classification, [Residence] facilities were newly added to the fundamental facilities for taking account the damage of residential

activity into the structural model.

As a result, nineteen facilities were determined

as the fundamental facilities in urban area. And

for making the expression easier, nineteen facili- ties were classified into five groups, such as [Transportation Facilities], [Lifeline Facilities],

[wholesale/ Financial Facilities], [Service/ Public

Facilities] and [Others].

Concerned with each nineteen fundamental facilities, seismic damage which can be expected to be developed by an earthquake was studied by

brainstorming method. Bramstorming was carried out by 4-6 persons who were specialists on

earthquake engineering. Associated with [Lifeline

Facilities] and [Wholesale/ Financial Facilities],

interviews to administrators were carried out in addition to the results of brainstorming.

On the brainstorming," seismic damage was classified into physical damage, direct functional

damage, primary functional damage and secondary functional damage in accordance with the

propagation stages of earthquake disaster. Physical damage means the damage of structures and

equipments. Direct functional damage means the

loss or decrease of functions developed in the

facility due to physical damage. Primary functional damage means the loss or decrease of

functions developed in the facility as a result of

the primary propagation from direct functional

damage. Secondary functional damage means the

loss or decrease of functions developed in the

facility as a result of the secondary propagation

from primary functional damage.

Based on the brainstorming and interview, 527

items of seismic damage which include 148

items of physical damage, 121 items of direct

functional damage, 138 items of primary functional damage and 120 items of secondary functional damage were extracted. Table 1 shows the seismic damage of road facilities and financial

facilities which are expected to be developed by

an earthquake. Associated with road facilities,

typical propagation paths are estimated as fol- lows;

(1) [Subsidence on Road] and [Slope Failure] as physical damage will cause [Passing Prohibition]

and [Passing restriction] as direct functional

damage.

(2) [Passing Prohibition] and [Passing Restriction] as direct functional damage will cause

176
[Passing Trouble of Vehicles] as primary functional damage.

(3) Passing trouble of emergency vehicles, such

as fire engines and ambulances, cause [Spread of

Fire] and [Casualties] as secondary functional

damage.

Associated with financial facilities, typical propagation paths of functional damage are estimated

as follows;

(1) [Collapse of Bank Building] as physical

damage cause [Trouble of Service at Counter] as

direct functional damage.

(2) [Trouble of Service at Counter] as direct

functional damage cause [Restriction of Payment] and [Psychological Confusion] as primary

functional damage.

(3) [Restriction of Payment] as primary functional damage cause [Bankruptcy of Companies]

as secondary functional damage.

As was mentioned before, emphasis of this study was placed on clarifying the outline of disaster

propagation rather than the detail of functional

interaction. Because it was difficult to express

527 items of seismic damage in a structural model, before the formulation of structural model, it was required to summarize all these items

into fewer representative seismic damage. Therefore 527 items of seismic damage were summarized by VE (Value Engineering) method. Fundamental policy of summarization was as fol- lows;

(1) Physical damage and direct functional damage are closely related each other. Because direct

functional damage could be represented by physical damage, direct functional damage was neglected in the structural model.

(2) Physical damage was represented by nine- teen seismic damage for each nineteen fundamental facilities. For example, all physical damage of [Road] facilities such as [Subsidence in Road] and [Slope Failure] were represented by

[Physical Damage of Road Facilities].

(3) Primary functional damage was also repre- sented by seismic damage for each nineteen fundamental facilities. Primary functional damage of

[Road] facilities such as [Passing Trouble of Fire

Engines] and [Passing Trouble of Ambulances]

were represented by [Trouble of Road Transportation]. Associated with [Wholesale/ Financial],

[Service/ Public] and [Others] facilities, functional damage ought to be represented by [Restriction of Commerce], [Restriction of Finance]

and [Restriction of Service] alternatively. But

they were defined as secondary functional damage because they could be also considered as the

damage due to primary functional damage of

[Transportation] and [Lifeline] facilities. (4) Secondary functional damage was classified

based on the fundamental functions of metropolis, i.e. economy, residence and public service.

Functional damage associated with economy was

represented by [Restriction of Production], [Restriction of Commerce], [Restriction of Finance]

and [Restriction of Service]. Functional damage

associated with residence was represented by

[Restriction of Residence], [Restriction of Materials], [Restriction of Behavior] and [Restriction

of Information]. And functional damage associated with public service was represented by [Restriction of Medication], [Restriction of Education], [Restriction of Administration] and [Restriction of Mass Communication].

Although seismic damage was summarized into

fewer representative damage to make the structural model clear, 527 items of seismic damage

was reflected in evaluating the interaction between representative damage.

3. STRUCTURAL MODEL OF

FUNCTIONAL DAMAGE DUE TO AN

EARTHQUAKE IN URBAN AREA

Fig. 2 shows the structural model formulated by

brainstorming method. It represents the basic

interaction between functional damage developed

by an earthquake. Based on this model, several

features of the earthquake disaster propagation can be deduced.

(1) As shown in Fig. 3, two types of disaster

propagation can be observed. Associated with the

first type of disaster propagation (designated

177
hereinafter as forward propagation), seismic

damage propagate from physical damage to sec- ondary functional damage via primary functional

damage. Associated with the second type of di- saster propagation (designated hereinafter as same stage propagation), several seismic damage

which belong to the same stage cause each other. For example, [Damage of Road Facilities] cause [Damage of Gas Facilities] in the stage of physical damage. Also, [Trouble of Electricity Supply] cause [Trouble of Road Transportation] in the

stage of primary functional damage.

(2) In the stage of physical damage, [Damage of

Road Facilities], [Damage of Harbor Facilities],

[Damage of River Facilities] and [Damage of

Water Supply Facilities] cause the same stage

propagation. [Damage of Road Facilities] causes

physical damage of lifeline facilities such as

electricity, gas, water, sewerage and communication which are embedded in the road embankment. Associated with [Damage of Harbor Facilities], [Damage of River Facilities] and [Damage

of Water Supply Facilities], flood developed by

these damage cause physical damage of road

facilities and railroad facilities. (3) In the stage of primary functional damage,

[Trouble of Electricity Supply] causes the same

stage propagation. [Trouble of Electricity Supply] causes the primary functional damage of

transportation facilities and lifeline facilities due

to their highly dependency on electricity.

Because the structural model represents the outline of disaster propagation qualitatively, quantitative effects such as the strength of interaction

and the rapidity of propagation are not considered at this time. Therefore, It is required to improve the structural model for taking account

these quantitative effects into analysis in the future.

4. FUNDAMENTAL FACILITIES MOST

INFLUENTIAL TO EARTHQUAKE

DISASTER PROPAGATION

Fig. 4 shows the comparison of total number of

propagation paths starting from each physical

damage of nineteen fundamental facilities. Because the quantitative effects was not expressed

in the structural model, the influence of physical

damage was evaluated by counting the number

of propagation paths for each nineteen fundamental facilities. From this study, following

points can be deduced.

(1) Most of all the propagation paths are developed by physical damage of [Transportation Facilities] and [Lifeline Facilities]. Propagation

paths developed by physical damage of [Wholesale/ Finance Facilities], [Service/ Public Facilities] and [Others] are quite few. This is because

[Transportation Facilities] and [Lifeline Facilities] are used for various industries depending on

the characteristics of users. On the contrary,

[Wholesale/ Finance Facilities], [Service/ Public

Facilities] and [Others] are used by limited users

in specific industry. Therefore it is pointed out

that physical damage of facilities with highly

publicity, such as [Transportation Facilities] and

[Lifeline Facilities], is more influential to the

earthquake disaster propagation.

(2) Among [Transportation Facilities] and [Lifeline Facilities], [Road], [Electricity Supply] and

[River] facilities produce more propagation paths

than the other facilities. The number of propagation paths is increasing in the order of [Road],

[Electricity Supply], [River] facilities. If physical

damage of [Road] facilities can be prevented by

conducting appropriate seismic countermeasures,

about 36 % of all the propagation paths in the

structural model can be interrupted. And if physical damage of these three facilities can be pre- vented, about 62 % of all the propagation paths can be interrupted.

For studying the difference of disaster propagation characteristics among [Transportation Facilities], Fig.5 and Fig.6 show the number of propagation paths starting from [Road] facilities and

[Airport] facilities alternatively. Although both of

these facilities have highly publicity, the number

of propagation paths starting from [Airport]

Facilities is much less than that of [Road] facilities. This is because physical damage of [Road]

facilities develop the same stage propagation,

178
and physical damage of [Airport] facilities does

not develop the same stage propagation. Therefore it is pointed out that fundamental facilities which develop the same stage propagation are more influential to the earthquake disaster propagation.

5. FUNCTIONS OF METROPOLIS MOST

VULNERABLE AGAINST EARTHQUAKE

DISASTER

Fig.7 shows the comparison of total number of

propagation paths getting to each secondary

functional damage of metropolis. Because propagation paths getting to the secondary functional

damage represent the factors causing each damage, the vulnerability of metropolitan functions can be evaluated by counting the number of

propagation paths for each metropolitan functions. Based on this study, the number of propagation paths is largest in [Restriction of Production], [Restriction of Commerce], and [Restriction of Service] among all functional damage. [ Restriction of Education], [Restriction of Medication] and [Restriction of Administration] are following above-mentioned functional damage.

Therefore, for increasing the seismic safety of

metropolis, appropriate seismic countermeasures

which interrupt the propagation paths getting to these vulnerable functions are required to be

conducted.

For studying the factors which develop [Restriction of Production], Fig.8 shows the number of

propagation paths from each physical damage of

nineteen fundamental facilities. Associated with

[Restriction of Commerce] and [Restriction of

Service], the tendency is almost same with [Restriction of Production]. From this figure, it is pointed out to be effective to prevent the physical damage of [Road] facilities for reducing [Restriction of Production].

CONCLUDING REMARKS

For studying the interaction of functional damage

due to earthquake in urban area, structural model

of earthquake disaster propagation was formulated by brainstorming method. Associated with the

disaster propagation characteristics, fundamental

facilities most influential to disaster propagation

and metropolitan functions most vulnerable

against earthquake disaster, following conclusions were deduced with use of the structural model.

(1) Two types of disaster propagation were observed. In the forward propagation, seismic damage propagate from physical damage to secondary functional damage via primary functional

damage. In the same stage propagation, several

seismic damage which belongs to the same stage cause each other. Physical damage of road facilities, harbor facilities, river facilities, water supply facilities and primary functional damage of

electricity supply facilities cause the same stage

propagation.

(2) Physical damage of road facilities, electricity

supply facilities and river facilities are most in- fluential to the disaster propagation. If physical

damage of road facilities can be prevented, about

36% of all the propagation paths can be inter- rupted. If physical damage of three facilities can

be prevented, about 62% of all propagation paths can be interrupted.

(3) Among the functions of metropolis, production, commerce and service are most vulnerable

against earthquake disaster. Because the major

factor causing restrictions of these functions is the physical damage of road facilities, it is important to prevent the physical damage of road

facilities for increasing the seismic safety in ur- ban area. REFERENCES

(1) Kazuhiko KAWASHIMA and Takashi

KANOH: Evaluation of Indirect Economic Effects Caused by The 1983 Nihonkai-chubu, Ja- pan, Earthquake, Earthquake Spectra, Vol.6,

No.4, November 1990

(2) Public Works Research Institute: Evaluation

of Economical Indirect Damage Caused by

Earthquakes, Report of PWRI, Vol. 186-1, Dec.

1991

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Fig. 1 Classification of Fundamental Facilities in Urban Area

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185

Dynamic Behavior of Pile Foundation in

Liquefaction Process

by

T. Kagawa 1 , C. Minowa2 and H. Mizuno3

ABSTRACT

This paper highlights results and a

preliminary analysis of a series of

shaking-table tests on model pile-foundation

systems in loose saturated sand. The

ultimate objective of this research is to

develop numerical methods and design

guidelines for pile foundations in potentially

liquefiable sand deposits. This study,

although preliminary in nature, demonstrated

the significance of liquefaction on pile

responses and provided an improved

understanding of the soil-pile interaction in

liquefying sand.

INTRODUCTION

Pile foundations are common at potentially liquefiable sites, and they have

experienced damage and failures during

major earthquakes. A number of studies

have been made in recent years on dynamic

and seismic behavior of isolated piles and of

groups of piles, to increase our

understanding of seismic soil-pile-structure

interaction. The state of the art in this field,

however, has not advanced to a point where

we can predict the performance of pile

foundations when the surrounding soils

undergo liquefaction during an earthquake.

A joint research was initiated in 1992

between the National Research Institute for

Earth Science and Disaster Prevention

(NIED), Science and Technology Agency of

Japan and Wayne State University to study

the response of pile foundations in liquefying

sand. As an initial step toward establishing

rational analysis procedures and design

guidelines for pile foundations at potentially

liquefiable soil deposits, this joint research

attempts to quantify the lateral soil-reaction

response of piles when the surrounding soils

undergo liquefaction. To accomplish this

objective, the first and the second series of

shaking-table tests were already conducted,

respectively, in April, 1992 and July, 1993 at NIED. Also model pile-section tests are being conducted at Wayne State University

to obtain an improved insight into the results

from the shaking-table tests.

This paper highlights the first series of

the shaking-table tests on model piles in

loose saturated sand and the major findings

from the tests. This is followed by a summary of results from our preliminary

1) Professor, Department of Civil Engineering, College of Engineering, Wayne State

University, Detroit, Michigan, U.S.A.

2) Cooperative Research Officer, National

Research Institute for Earth Science and

Disaster Prevention. Science and Technology Agency, Tsukuba-City, Japan

3) International Codes and Standards

Research, Building Research Institute,

Ministry of Construction, Tsukuba-City,

Japan

187
numerical analyses using existing nonlinear

soil-pile-structure interaction methods.

DESCRIPTION

OF SHAKING-TABLE TESTS

Background

No field observations on the response

of pile foundations in liquefying sand layers are available at present. Therefore, one of

the major tasks of this research is to obtain

reliable experimental data on the lateral soil

reaction responses of piles when the

surrounding soils undergo liquefaction and

flow through the piles. Such data may be

obtained by two experimental methods,

centrifuge tests and 1-g model tests. The

similitude requirements for these types of

tests were summarized by several

investigators; i.e., Kagawa (1978), Iai

(1989), etc. Centrifuge tests involve large

geometrical scaling, typically on the order of

50 to 100. Therefore, it becomes difficult to

satisfy the geometrical similarity on soil

particle size. This may result in incorrect

representation of the failure strength of the

sand mass around the model pile. Also,

small-scale tests require large reduction in

the rate of pore pressure redistribution. For

these reasons, the authors prefer 1-g model

tests with reasonably large models.

However, the size of the model

pile-foundation system in this research is rather small to reduce experimental cost and

to serve as the preliminary tests to the

large-scale tests, which are scheduled to be

performed in 1996 through 1998. A

large-scale laminar shear box with a height

exceeding five meters is being designed at NIED. Pile foundations with nearly

prototype scales will be installed in this shear

box, and the effects of liquefaction and

lateral flow of liquefied soils on piles will be

studied.

Sand Layer

The saturated sand layers for the tests were prepared in a laminated shear box with

a plan dimension of 2 by 2 meters and a

height of 1 meter. The shear box consisted

of twelve independent rectangular steel frames with a height of 10 cm, stacked on

top of each other with bearings between the

frames to force the sand layer to deform in one-dimensional shear deformation. The

laminated shear box with sand and the model

pile foundation is shown in Fig. 1.

The test sand was the Kasumigaura

sand taken from the bed of the Kasumigaura

Lake near the city of Tsuchiura, Japan. The

physical characteristics of the Kasumigaura

sand have been studied reasonably well by

various researchers in Japan. The sand is well graded with a moderate amount of

fines. The specific gravity of the sand was

2.71; the percent weight of fines (silts and

clays) was 3.4 %; the mean particle diameter

(D50) was 0.0885 cm; the uniformity

coefficient was 5.62; the maximum and

minimum dry densities were 1.76 and 1.36

g/cm3

; and the coefficient of permeability

was 3.1 x 10'2 cm/sec.

For each shaking-table test, a fresh

sand layer was prepared by pouring the wet

test sand into the shear box, which was

partially filled with water. The density,

water content, and the shear-wave velocity

of the sand layer thus prepared were

measured before and after shaking. Results

are summarized in Table. 1.

Model Pile Foundation

The model pile foundation system

consisted of two steel strips, placed in line

with the direction of shaking with a clear

distance of 18 cm, and a steel box filled with

188
lead shots. The length, width, and thickness

of the steel strips (model piles) were 91.7

cm, 5 cm, and 0.6 cm. The plan dimension

of the steel box was 27.8 by 17.3 cm with a height of 8.3 cm. The steel box with lead

shots represented a structure with a

translational degree-of-freedom only. The

total weight of the steel box and the lead

shots was 12.4 kgf. The steel strips were

rigidly bolted to the steel box, to reproduce

the fixed-head condition for the model piles.

The tips of the model piles were pinned at the base of the shear box. Therefore, they

were free to rotate.

Instrumentation

Figure 2 shows a schematic view of

the cross section of the sand layer-model pile

foundation system and of the layout of

various instrumentation devices. For the

sand layer, acceleration responses were

measured at four levels, A-l, A-2, A-3 and

A-5; and the excess pore-water pressures

were measured also at four levels, P-l

through P-4, and P-5. For the model pile

foundation system, six pairs of strain gages

were attached to the model pile, 1-1 through

6-1 and 1-2 through 6-2; the soil stresses on

the model pile were measured at three levels,

F-l through F-3; and the acceleration and

the displacement responses of the structural mass were measured, A-5 and D-4. In

addition, the displacement responses of the

frames were monitored at three levels, D-l

through D-3 to estimate free-field soil movements.

Shaking-Table Input

The large-scale shaking table at the

National Research Institute for Earth

Science and Disaster Prevention has a table

dimension of 15 by 14.5 meters with a maximum model weight of 500 tons. The

table is excited by four synchronized

actuators in a fixed horizontal direction with

a maximum displacement amplitude of 22

cm.

In the first series of the shaking-table

tests conducted in 1992, a random wave

with uniform power amplitudes over the

frequency range up to about 25 Hz was used

as input to the shaking table. The peak table

acceleration was about 100, 140, 140, 160,

and 160 gals respectively for Test Nos. 2

through 6. These acceleration levels were

considered to be consistent with the typical

acceleration levels associated with historical

liquefaction events and sufficiently large to

produce liquefaction over the entire depth of

the sand layers on the shaking table.

SUMMARY OF TEST RESULTS

Pore Pressure Responses

Figure 3 shows the time histories of

measured responses from Test 3. Results

from the other tests are similar to those in this figure. Several important observations

may be made. For example, acceleration at shallow depths decreases dramatically as pore pressures build up, and it tends to increase again with time. Excess pore

pressures at deeper depths increase first and

then decrease steadily with time, while those

at shallower depths tend to be constant or even increase during shaking. These

responses are due to the upward

redistribution of excess pore pressures. The

excess pore pressures at P-5 are difficult to

dissipate, due mainly to its long drainage

path.

Table 2 summarizes the peak excess pore-water pressure ratios within the sand

layers. Except Test 2, the sand layers

liquefied almost throughout their entire

189
depths. It should be noted that larger pore

pressure ratios were recorded at shallower

depths and that excess pore pressure ratios

exceeded unity by large margins at locations

P-4 and P-5. The latter observation is due to

the upward flow of the excess pore

pressures developed at deeper depths. Also

it should be noted that the excess pore

pressures measured at P-5 (at 10-cm depth

and between the model piles) are different

from the readings at P-4, which was located

at the same depth but in the free field.

These seemingly inconsistent and complex

observations are related closely to the

relative speed of pore pressure buildup and

of pore pressure redistribution and

dissipation.

Acceleration Responses

The peak acceleration within the sand

layers are summarized in Table 3. Except

Test 4, the peak acceleration is larger at shallower depths. The peak accelerations of

the structural mass are higher than those at A-3, except Test 5. These observations

could be explained by carefully examining

how excess pore pressures developed and

redistributed within the sand layers, and

they demonstrate the importance of the

effect of liquefaction on the responses of the

sand layer and the model pile.

p-y Relations

A preliminary analysis has been

completed on the data obtained from the

strain-gage pairs attached to the model pile.

Figure 4 shows the typical results from this

analysis.

The analysis yielded the following

tentative conclusions on the lateral soil

reaction versus pile deflection (p-y) relation

of the model pile. The p-y relation for the

initial stage of shaking appears to be a

softening type; the p-y relation becomes

mainly of a viscous type when the

surrounding sands undergo liquefaction; and

the p-y relation gradually transforms back to the initial softening type as excess pore

pressures dissipate from the sand layer.

Also the dilatancy tendency of liquefied

sands is seen in the p-y relations.

Another analysis yielded the simple

results on the change of lateral soil reaction

coefficient versus pile, and reduction of the

shear velocity of sand layer during the

excitation. Fig. 5 shows the reductions

which were brought by excess pore

pressure increase. In the estimation of

lateral soil reaction coefficients, the

deformations of pile were assumed the fifth

polynomials of height positions.

NUMERICAL ANALYSIS METHODS

Response of the model pilefoundation system to shaking-table

excitation was computed in two steps;

computation of the response of the sand

layer and computation of the response of the

model piles. The sand layer was assumed to deform in one-dimensional shear, and the

response of the sand layer was obtained by a

nonlinear site-response method SRANG2.

The response of the model piles was

computed by a nonlinear soil-pile-structure

interaction method NONSPS2. These

programs were documented by Kagawa

(1992a and 1992b). Therefore, the key

features of these methods are highlighted

below.

Response of Sand Layer

It was reasonable to assume that the

sand layer in the laminated shear box

deformed in one-dimensional shear.

Therefore, the response of the sand layer

190
was obtained by the one-dimensional

wave-propagation theory that has been the

most common tool for site-response

analyses. Computed responses of the sand

layer were then used in a subsequent

soil-pile-structure interaction analysis as multiple soil-support input motions. The

stress-strain relation of the sand was

represented by a multiple, elastic-plastic

spring model, which may be considered a

simplified version of the multi-surface

plasticity model. Excess pore pressures

developed in the sand layers during shaking

and they redistributed during and after

shaking. The simple pore pressure model

by Kagawa and Kraft (1981) was used to

evaluate the pore pressure development

aspect, and a standard one-dimensional

consolidation model was used to compute

redistribution and dissipation of excess pore

pressures.

In the present analysis, the sand layer

was divided into eighteen horizontal layers

with an equal thickness of 5 cm. The

small-strain shear moduli at the mid heights

of the layers were computed from the

experimental equation proposed by Iwasaki,

Tatsuoka and Takagi (1978). The backbone

curve of the shear stress versus shear strain

relation was determined from an average

trend of the shear-modulus reduction curves

obtained by various investigators for

cohesionless soils. The computed

shear-wave velocity was about 60 m/sec. at 5-cm depth and 150 m/sec. at 85-cm depth.

These agree well with the measured

shear-wave velocities in Table 1, which

range from 80 to 118 m/sec.

Response of Model Pile-Foundation

System

The numerical method used to obtain

the response of the model pile-foundation

system was based on a beam-on-Winkler

foundation model, which is similar to the

one reported by Kagawa and Kraft (1981).

In this method, a pile-supported structure is represented by a single-pile system; the

superstructure is represented by a series of

lumped masses, connected by beams; the

pile is divided into segments; and each pile

segment is connected to free-field soil

through soil-pile elements. The numerical

method is capable of representing complex

soil-pile conditions such as; 1) nonlinear

stress-strain behavior of soils around a pile,

2) excess pore pressure buildup in the

free-field and around a pile, and 3) excess pore pressure redistribution and dissipation

in the radial direction away from the pile

and in the vertical direction. The EI of the

model pile was estimated to be 1.9 x 105 kgf-cm2

. SUMMARY OF RESULTS FROM

NUMERICAL RESPONSE ANALYSIS

Figure 6 shows typical comparisons

between measured and computed responses

of the model pile-foundation system.

Acceleration responses at A-4 (structural

mass), A-3 (10-cm depth), A-2 (3 5-cm

depth), and P-3 (2 5-cm depth) are shown in

Fig. 6. Meassured accelerations and pore

pressures agree reasonably well with

measured responses. However, computed

acceleration responses tend to include larger

amplitudes than those from computations.

Also computed sand displacements were

generally less than those from the tests. These results appear to suggest the

importance of dilatancy tendency of

liquefied sand in controlling the responses

of the sand layer and the model pile. In

addition, dissipation of excess pore

pressures had a significant impact on computed pore pressure responses. Bending

moment in the model pile was controlled

mainly by the deformation of the sand layer

except near the structural model, and

liquefaction of the sand layer increased

191
IKI1IIU considerably the magnitude of the be„j;^g

. . : the n i >del p?ie.

: COMMENTS

This study yielded several important

conclusions and observations itemized

below:

1. Excess pore pressure responses in

the sand layer are affected

significantly by their redistribution

i id dis: ipation.

2. The excess pore pressures in sand

beneath the structural model

appear to be "trapped," and they are slow to dissipate compared to those

in the free field at the same

elevation.

3. The sand layers in liquefied

conditions exhibit dilatancy. This

was indicated in the responses of

the sand layers and in the lateral

soil reaction versus pile

displacement (p-y) relations of the

model piles.

4. The lateral soil reaction versus pile

displacement relations exhibit

complex behavior during shaking.

_ lateral soil reaction to the

model pile does not vanish even

after the liquefaction of

surrounding sand.

5. Bending moment in the model pile

increased significantly due to

liquefaction of the sand layer.

The experiment and the analysis

presented in this paper could not answer all the questions that we wanted to resolve. We

expect that these questions will be answered

from ongoing experimental and analytical

efforts.

ACKNOWLEDGMENT

The authors appreciate encourage

-ment provided by Dr. N. Ogawa, the

National Research Institute for Earth

Science and Disaster Prevention and

technical supports provided by Mr. A. Abe,

Tokyo Soil Research Co., Ltd., and Dr. M.

Yoshikawa, Tsukuba Research Institute,

Okumura Corporation.

REFERENCES

Iai, S. "Similitude for Shaking Table Tests

on Soil-Structure-Fluid Model in lg

Gravitational Field." Soils and Foundations,

Vol.29, No. 1 (1989): 105-118.

Iwasaki, T., Tatsuoka, F., and Takagi, Y.

"Shear Moduli of Sands under Cyclic

Torsional Shear Loading." Soils and

Foundations, Vol.20, No.l (1978): 45-59.

Kagawa, T. "On the Similitude in Model

Vibration Tests of Earth Structure."

Proceedings, Japan Society of Civil

Engineers, No.275 (1978): 57-67 (in

Japanese).

Kagawa, T. and Kraft, L.M., Jr. "Lateral Pile

Response during Earthquake." Journal,

Geotechnical Engineering Division, ASCE,

Vol.107, No.12 (1981): 1713-1731.

Kagawa, T. SRANG2 and NONSPS2 - Numerical Methods for Nonlinear

Earthquake Response of Soil-Pile System, Geotechnical Engineering Report,

Department of Civil Engineering, Wayne

State University, Detroit, Michigan, 1992a.

Kagawa, T. "Effects of Liquefaction on

Lateral Pile Responses." Geotechnical

Special Publication, No. 34, ASCE (1992b):

207-223.

192
Fig. 1 Photo of Shaking-Table Model

Shear Box

Shaking Table

A-1 - A-5 ... Accelerometers D-1 ~ D-4 .... LVDT P-1 ~ P-5 .... Pressure Transducers

1-1 - 6-1 & 1-2 ~ 6-2 .... Strain Gages F-1 - F-3 .... Soil Pressure Transducers

Fig. 2 Layout of Instrumentation

193
position

Hf/Y^ .-f —- ' ~

A2

A3

A4

A5

p-1

P-2

P-3

P-4

P-5

1-2

2-2 ^^-v:—j-rr^^r^y^^ 3-2

4-2

5-2

._ 6-2

l-l

1 1 ' T » l r

' ' 1 -—r-"-n «[" i«; ' r

" r r" _ — v- » —v — " m

2- 1

3- 1

4- 1

5- 1

6- 1

Psl

Ps2

Ps3

D 1

D2

D3

D4

1 10.00 SEC.

-I I I I I I I I I I I LU 1 III, III, I 1 1 I 1 1 II, III I_l 1—1—1 ' ' I ' ' ' ' ill III I I I I II I i

Fig. 3 Time Histories of Measured Responses (Test 3)

194
0.4

-1 An 1 1 1 1 1 -1.5 -1 -0.5 0 0.5 1 1.5 Deflection, cm

Fig. 4 Soil Reaction versus Pile Deflection Relation of the Model Pile

195
1.2

1.0E3— B - -

0.8 —

0.6 -

0.4 g

- TM ~~

0.2 - ~ ~- a—

QQ] , , , , L, 0.0 0.2 0.4 0.6 0.8 1.0 1.2

Excess Pore Pressure Ratio

K/Ko S.V/S.Vo

Fig. 5 The change of lateral soil reaction coefficient versus pile and the shear velocity

196
CO 900

03CD

C 1 UU

/—i

> 03

i_ 0

(ii 0

O -100

< -900 -zuu

CO ZUU 900

03O)

c

•1 on l UU

o

2 0

ft -100

o

<

-200

9 11 13 15 17 19 21 23 25

A-3

J I I L

1 7 11 13 15 17 19 21 23 25

_co 03CD

o

-I—'

0)

<DO

o

<

150

100

50

0

-50

-100

-150

2 2.0

1.5

1.0

0.5

A-2

J L J L

1 3 5 7 9 11 13 15 17 19 21 23 25

03 Od

COCO0

a.o

CL

0.0

n 1

Computed

Measured

i r

P-3 -

3 5 7 9 11 13 15 17 19 21 23 25

Time, sec.

Fig. 6 Comparison between Computed and Measured Soil-Pile-Responses

197
Table 1. Physical Properties of the Sand Layers for the Shaking-Table Tests.

Physical Properties Test 2 Test 3 Test 4 Test 5 Test 6

Dry unit wt. bef. shaking (g/cm3) 1.488 1.495 1.490 1.489 1.492

Dry unit wt. aft. shaking (g/cm3) 1.495 1.508 1.495 1.498 1.500

Water content bef. shaking (%) 26.4 28.7 27.2 27.8 26.9

Relative density bef. shaking (%) 37.8 40.1 38.5 38.3 39.2

Relative density aft. shaking (%) 39.7 42.2 40.1 40.9 41.2

Shear-wave velocity bef. shaking (m/sec.) 86 118 80 96 92

Shear-wave velocity aft. shaking (m/sec.) 132 132 128 128 135

Table 2. Peak Pore Pressure Ratios within the Sand Layer

Location Test 2 Test 3 Test 4 Test 5 Test 6

P-5 (bet. Piles) 1.40 1.11 3.86 3.52 3.68

P-4 1.86 2.56 1.82 3.18 1.26

P-3 1.13 1.29 1.27 1.32 1.14

P-2 0.66 1.05 1.03 1.00 1.03

P-l 0.49 0.98 0.94 0.93 0.91

Table 3. Peak Accelerations (in gals) within the Sand Layer - Model Pile System

Location Test 2 Test 3 Test 4 Test 5 Test 6

Structural Mass 376 413 398 368 442

A-3 329 294 175 450 363

A-2 102 215 667 318 333

A-l 100 149 154 164 172

Shaking Table 98 143 143 156 165

198
Suitability of SASW Method in Complex Geosystems

by

D. W. Sykora*, J. M. Roesset**, and K. H. Stokoe II**

ABSTRACT

Surface wave methods were developed as a nonintrusive and non-destructive means to determine

the variation of elastic moduli with depth. These

methods have been widely applied to soil, rock,

and pavement systems. Initial simplifying

assumptions included that the ground is level, the sub-surface layers are horizontal and extend

laterally to infinity, and the effective sampling

depth can be related to some fraction of the wavelength. Recent modifications and the findings of field and analytical studies indicate that surface wave methods, specifically the SpectralAnaJysis-of-Surface-Waves (SASW) method, can be useful for conditions that do not meet these assumptions.

This paper addresses the application of the SASW

method to complex systems of soil and rock

(geosystems). The results of numerical analyses

and field measurements are used to show that the SASW method can reliably and accurately

determine the variation of elastic moduli in sloping ground or a system where discontinuous

layers exists. The SASW method has also been

used successfully to:

1) profile tunnel and shaft walls; 2) locate significant fractures within structural beams or behind tunnel liners; and

3) detect cavities or other irregularities at shallow depths below an exposed surface

which are also briefly described.

The SASW method may someday be used to characterize material properties in even more

complex systems such as zoned, earth-fill embankments. The SASW method has been

found to not work well to detect and locate

cavities or tunnels at significant depths below the ground surface.

KEYWORDS: cavities; elastic moduli; rock; seismic methods; site characterization; soil; surface waves.

1. INTRODUCTION

The analysis of surface waves has been used for several decades to obtain the variation of soil stiffness with depth. The steady-state surface wave technique was developed in the late 1950's (Heukelom 1958; Jones 1958; Heukelom and

Foster 1960) for the analysis of pavement systems.

This technology was adapted by the U.S. Army

Engineer Waterways Experiment Station (WES)

to be an exploration tool at soil sites (Fry 1963; Ballard 1964). This technique involved

interactively determining the wavelength at several frequencies of excitation by measuring the distances from the source to successive standing wave (displacement) peaks. The Rayleigh wave

velocity, VR

, is calculated knowing the wavelength,

X, and frequency of excitation, / : /A (1)

and was assumed to correspond to an average depth between X/3 and X/2. The SASW method was established to obtain a more accurate

variation of elastic moduli with depth.

The SASW method was developed nearly two

decades later (Heisey 1982;Heisey, Stokoe, and

Meyer 1982; Nazarian 1984; and Nazarian and

Stokoe 1985a, 1985b) at the University of Texas at Austin (UT). Since that time, research,

development, and application activities have

continued at UT (e.g., Coll-Calderon 1985; Nogueira 1986; Sanchez-Sal inero 1987; Sheu 1987; Rix 1988; Lopez 1989;Gauer 1990; Kang 1990; Madianos 1991; Sedighi-Manesh 1991;Bowen

Woodward-Clyde Consultants, Santa Ana,

California, 92705; formerly, Geotechnical Lab.,

U.S. Army WES, Vicksburg, Mississippi 39180

Dept. Civil Engr., University of Texas,

Austin, Texas 78712

199
1992;Sykora 1993; Kalinski 1994; Luke 1994), at the University of Michigan (e.g.,Hiltunen 1988;

and Gucunski 1991), at WES (Alexander 1992; Sykora, Alexander, and Roesset 1994), as well as

at other universities and laboratories.

2. SASW METHOD

The determination of stiffness profile using the SASW method consists of three phases: data measurement, signal analysis, and forward

modeling/inversion, each of which is described

briefly below.

2.1 Data measurement

The SASW method involves measuring seismic

signals at the ground surface (thus it is nonintrusive) produced by some type of vertical excitation, shown as either harmonic or impulsive

in Figures 1 and 2, respectively. Signals typically are recorded with two receivers and stored

digitally. The receivers may be velocity or

acceleration transducers; acceleration transducers

are normally reserved for cases when a significant

amount of high frequency energy is being

recorded.

2.2 Signal analysis

Signal analysis is normally performed using a

spectral analyzer. If an impulsive source is used, a time window with the full length of the transient response is usually collected. The cross-power spectrum phase difference and coherence between

the two receivers is calculated, averaged with a number of other responses using the same

configuration and source, and stored. Data for harmonic sources are averaged and converted to cross power phase and coherence. Then the data

are analyzed to determine the variation of signal

phase as a function of source frequency and

develop a dispersion curve (variation of phase

velocity versus wavelength).

The phase velocity and wavelength are calculated

directly from the real and imaginary parts of the dynamic displacements (Nazarian and Stokoe

1985a). At a given frequency, /, the travel time,

t, between two receivers is:

where <j>j and <j>2 are the phases at points 1 and 2

(the location of the receivers) in degrees. Then,

the phase velocity, CA , is:

(cL-d.) (<L-d,)

r = — 1 = 360 / — 1 (3) * t (4>2-<l>i)

where the wavelength, k, is:

A = -± (4) f

2.3 Forward Modeling/Inversion

This phase is an iterative process of forward

modeling which is independent of whether the data were collected using an impulsive or a harmonic source. The site is assumed to be

ideally layered (horizontal and extending to

infinity). Different algorithms allow for the evaluation of plane waves or cylindrical-front surface waves. Typically an initial profile of layer

thicknesses and moduli is selected and the dispersion curve calculated using matrix solutions (Haskell 1953 and Thompson 1950), comparing

the measured and matrix-calculated dispersion

curves, modifying the profile if necessary, then

repeating the process until a close match exists. Automatic matrix inversion using least squares is also used at UT and elsewhere (Nazarian and

Desai 1993) and neural networks (Rix and Leipski

1991) have been proposed as an alternative to matrix inversion. The inverse problem does not

necessarily produce a unique solution, however.

2.4 Validation

Numerous validation studies of the SASW method

have occurred over the years, mainly through

comparisons with in-situ shear wave velocity

profiles determined by independent seismic measurements. The independent measurements

have generally been performed using either the crosshole or downhole method. Typical

comparisons of this type are shown in Figures 3 through 5 for some of the geotechnical settings

where the SASW is especially well suited.

The results of tests performed on liquefiable

gravelly soils are shown in Figure 3. These

gravels are difficult to sample because of their looseness as well as their large aggregate size. Figure 4 shows a comparison of shear wave

200
velocities measured at the site of an asphaltic

(AC) pavement. This site represents a difficult

site for testing by non-intrusive techniques

because of the very stiff layer (asphaltic concrete)

overlying the much softer base and subgrade

layers. Finally, the results of a recent application offshore are shown in Figure 5. In this case, the

cost effectiveness of the SASW make its application offshore very desirable.

In the examples shown in Figures 3 through 5, a good comparison between the shear wave velocity

profiles determined by the SASW and

conventional intrusive methods is seen. This close comparison has occurred because of the lateral uniformity which existed in the test areas. Hence,

these comparisons could be used to validate the SASW method. However, it must be kept in mind that intrusive techniques, like the downhole

and crosshole tests, result in measurements over

short travel paths. Therefore, when lateral variability exists, one should expect to see differences in the shear wave velocity profiles

determined by the test methods because of the

different materials sample by each test method.

3. DISCONTINUOUS LAYER

3.1 Numerical formulation and implementation

The computer code vib3 is a research tool used to evaluate the effects of complex geosystems on

surface wave propagation (Sykora and Roesset

1992; Sykora 1993). This semi-analytical

computer code was created using principles of the

Finite Element Method (FEM) and validated with

closed-form solutions for axi-symmetric problems

(Kausel 1981). Parametric analysis have also been

conducted to define appropriate ranges for system

and problem variables.

The primary assumptions used to develop vib3 are

that the geometry and boundary conditions of the system and the distribution of material properties

are planar (2-D) but the loads can be non-planar

(3-D). This set of conditions has a broader range

of applications than that for plane strain (commonly used in engineering analysis) while

circumventing expensive 3-D solution methods.

Other assumptions are that all media are

isotropic, the hysteretic behavior is represented by

complex moduli, and the source produces vertical,

steady-state excitation on surface at one discrete frequency.

The formulation involves two primary

components: the condensation of 3-D dynamic

stiffness matrices to equivalent 2-D matrices and

the representation of the distribution of loads in the out-of-plane direction using a Fourier

expansion. The condensed 2-D system of

equations are first solved in the frequency and

wave-number domain; inverse Fourier transforms

are then performed to obtain the solution as a function of out-of-plane distance and time, if so

desired.

Validation studies were conducted to prove that the formulation and computer implementation are sound for the limited class of problems

considered. The general accuracy of calculated displacements in terms of amplitude and phase was evaluated. Validation of vib3 through

comparisons with analytical results was possible

only for the simplest class of planar geometry—

a

horizontally layered system extending to infinity. Green's function solutions formulated for axi- symmetric problems (with disk loads) by Kausel

(1989) were used exclusively with point and

square loads (in plan). The same total area and a

total load of unity were used to adequately equate

problems with different load configurations.

Comparisons between the numerical

approximations and closed-form solutions at distances from the source between 0.5X and 5. OA. are excellent for nearly all of the cases considered.

The data confirm that the formulation and

computer implementation are accurate.

3.2 Results of numerical analysis

The results of computations suggest that surface wave measurements can be used to detect, locate,

orient, and characterize discontinuous buried

layers (within certain limits). The ability to detect and characterize a buried layer is strongly

dependent on the velocity contrast, I, and the thickness-to-depth ratio, t/zL

. It appears unlikely

that surface wave methods will work well when

the thickness of the layer is less than 20 percent

of the depth to the layer.

An example of how a buried layer can be detected

is shown in Figure 6. The dispersion curves corresponding to surface waves propagating away

from the end of the buried layer (+x direction)

and surface waves propagating over the buried

201
layer (-x direction) when the load is directly over

the discontinuity are considerably different.

The orientation (azimuth of the edge) of a discontinuous buried layer can be determined

from a point on the edge by making

measurements along lines in a number of

horizontal directions. The azimuth of the discontinuous edge is 90 degrees from the measurement line that produces the greatest contrast in C^.

The location of the edge of (point above) a buried

layer is more difficult to determine. This process

is iterative and would also work from the direction of maximum difference in C^. Once

that direction was determined, dispersion curves would be developed along lines perpendicular to

that direction. A significant change in the dispersion curve would mark a transition from a homogeneous system to a buried layer system, or

vice versa.

4. SLOPING GROUND

4.1 Results of numerical analysis

The variations of dynamic vertical displacements

in the x-direction (direction of maximum

topographic gradient) and the y-direction

(direction of zero topographic gradient) were

calculated and analyzed using the computer code

vib3 described in section 3.1. The variables defining the problem are shown in Figure 7 with

the slope angle, a, and the load set-back distance, xLC , being the primary variables. Square loads were used throughout and can be related to circular load areas by equating total loads.

Calculations were made using six common

engineering slope ratios (horizontal: vertical)—1:1, 1.5:1, 2:1, 2.5:1, 3:1, and 6:1. The load was

placed at the top and base of the slope with these

ratios to produce a = 135, 146, 153, 158, 162,

171, 189, 198, 202, 207, 214, and 225 degrees.

(An a of 1 80 degrees corresponds to the level ground conditions.) The center of the load was

used as the reference point for distance measurements. Unit-less dimensions are used for convenience. The shear wave velocity was fixed at 1000 with 5 percent damping. The mass density

and Poisson's ratio were 4.00 and 0.40,

respectively, producing a VR

equal to 940.

The variations of vertical displacement and phase

at a slope, <(> s

, for a slope ratio of 1.5:1 (a = 146 degrees when the load is at the crest of

the slope and a = 214 degrees when the load is at the base of the slope) are shown in Figures 8 and

9, respectively. The relative location of the

discontinuity along the abscissa is shown as a

vertical dashed line. The level-ground solution is also shown for comparison.

The primary effect of the slope discontinuity is a phase shift for waves traveling in the direction of

maximum topographic gradient that occurs between the discontinuity and a distance of 3X/4

beyond (when xLC = 1.0 A). The phase difference

becomes essentially constant at greater distances. The phase has a marked lag from the level ground solution at distances beyond the discontinuity when a = 146 degrees and a marked

phase lead when a = 214 degrees. The calculated

solutions in the y-direction along the crest and

base correspond well to the level-ground solution,

regardless of a.

The amount of phase shift at these two slope

angles was examined by comparing the total (unwrapped) phase with the total phase assuming

a uniform increase of 360 degrees per wavelength

and passing through the origin, <j> 360 (which is used to calculate dispersion curves for the SASW

method). The difference in phase is shown in Figure 10. The variation of phase difference after the shift, beyond distances of about 3A./4 from the discontinuity, is relatively constant when

a = 146 degrees. The small fluctuations that exist are similar to those for the solution along the

crest. The overall variation of phase in the x- direction when a = 214 degrees is much larger

and less consistent. Along the base and crest there is no noticeable phase difference and the magnitude of fluctuations is small.

These data strongly suggest that measurements

can be reliably used both down the slope and

perpendicular to the slope to determine soil stiffness using the SASW method, especially if the

receivers are placed at a distance greater than

3X/4 beyond the discontinuity to avoid the phase

shift. The results for measurements up the slope

(a = 214 degrees) are not as consistent. The

phase difference appears to vary considerably at surface distances beyond the initial phase shift. The general trend at distances greater than 2.5 X

is for the phase to decrease uniformly (lag) as a

202
ratio of distance until a sudden shift occurs (e.g.,

at 3.4 A).

The variation of phase from a larger perspective

indicates that the differences shown in Figures 9

and 10 may not be of importance from a practical

standpoint. The variation of unwrapped phase

with distance for most slope angles (minor

discontinuities not included) are shown in Figures 11 and 12. Also shown in Figure 11 are

lines corresponding to a least squares linear fit of

data between surface distances of 1.0 and 5.0 A, for each slope angle.

The data in Figure 1 1 shows that the distribution of unwrapped phase is fairly consistent and unique

when a < 180 degrees. The best-fit segments at these angles effectively represent a variation of

phase between 360 and 350 degrees per wavelength. The increased variation in phase

when a > 180 degrees is also seen in Figure 11 although the best-fit linear segments are very

similar and are close to <t> 360 . The error associated with this variation is small and can be

minimized by using large spacings. The data in Figure 12 clearly shows the independence of

phase change along the crest and base to a and

how close the phase is to <t>360 (shown as a line segment in Figure 12).

Although it appears possible to predict the phase

shift caused by a surface discontinuity, calculated phase velocities are likely to be more accurate if this phase shift is avoided in the analysis. The

difference in phase as a function of a was found

to be independent of xLC as long as xLC > 0.5 X.

The combined results suggest that measurements made along the crest or base of a slope can safely

be processed using existing SASW inversion

schemes to obtain stiffness profiles. Phase

velocities representing measurements down a slope can also be reliably obtained from vertical measurements. The best method would be to place receivers beyond 3A./4 thus avoiding any

interpretation. Much more variation is expected

for measurements made up a slope. For homogeneous systems, the phase velocity will closely approximate VR

.

4.2 Field measurements

Field measurements of vertical dynamic

displacements on level ground and sloping ground

were made on the WES Reservation in Vicksburg,

Mississippi, during the month of August, 1993.

The site consists of a road cut with a maximum

height of 25 ft and slope angles between about

155 and 165 degrees and a concrete bridge with a

steel support system. Natural soil consists of

Pleistocene-age loess indigenous to the Vicksburg

area. This loess is uniformly graded and typically

has a low unit weight. Groundwater does not

exist within the depths tested.

A field testing program was designed to measure dynamic vertical displacements along the crest and

down the slope at two locations. A harmonic

source was used to produce surface waves. The

vibrator was programmed to run a swept-sine test between frequencies of 25 and 325 Hz at increments of 0.375 Hz. Five cross-correlations between the source and each receiver were

averaged at each frequency, the coherence

calculated, and results stored.

The slope angles were 160.5 and 163 degrees at two sites where measurements were made

(Locations A and B). The testing configuration was selected to avoid measuring phase shifts near

the discontinuity and allow the use of a wide

range of frequencies. Velocity transducers were

positioned vertically at locations down the slope

and along the crest (toward the other array

location) at 1 -ft spacings between distances of 10

and 21 ft from the center of the source platen.

The measured displacements were examined at a number of discrete frequencies between 50 and

200 Hz although aliasing of phase relationships

typically began to occur at wavelengths less than

3 ft (frequencies greater than about 125 Hz).

Typical results of slope measurements are shown

in Figure 13. The real and imaginary parts of

displacement are shown along with the wrapped

and unwrapped phase for waves propagating down

the slope and along the crest.

Very little difference exists between the change in phase with distance for wave traveling down the slope as compared to waves traveling along the

crest as shown in Figure 14. This finding is consistent with the results of numerical

approximations at distances greater than 3A./4 beyond the discontinuity.

203
5. TUNNELS AND SHAFTS

The SASW method has been used successfully

underground to evaluate conditions around

unlined and lined openings (Lopez 1989; Nelson

et al. 1989;Madianos 1991; Olson et al. 1993).

For underlined openings in rock, in-situ evaluation

of rock mass stiffness and changes associated with

the excavation and stress redistribution processes have been studies. A typical profile of shear wave

velocity determined at the Waste Isolation Pilot Project (WIPP) is shown in Figure 15. In this case, testing was performed along the exposed

rock wall of one of the underground rooms. Very

little change in the rock mass occurred due to the excavation and stress redistribution processes as shown by the nearly uniform velocity with depth

behind the exposed wall.

Lined openings of tunnels and shafts have also been tested with the SASW method. Testing has

been successfully used for purposes such as:

1) evaluation of liner integrity,

2) determination of liner thickness,

3) detection of voids immediately behind the

liner, and

4) evaluation of the rock stiffness behind the

liner.

A typical result for an intact concrete liner with

sound, but softer rock behind the liner, is shown

in Figure 16.

6. CAVITY AND TUNNEL DETECTION

6.1 Numerical formulation and implementation

Computer codes were created at UT to solve the problem of a harmonic line load imposed on the surface of a half space containing a void. Coll- Calderon (1985) used a horizontal line load and

Nogueira (1986) used a vertical line load. Assumptions of plane strain were used (making

the void extend to infinity—a tunnel) and the line load was oriented into the plane. The model was

formulated using the Boundary Element Method

(BEM) which allowed discretization of only the 2-D cavity outline. The codes also computed the cross power spectrum and displacements at target

points along the ground surface.

6.2 Results of numerical analyses

The results of these studies indicated that there should be some clear and distinguishable

variations in the dispersion curve if the source and

receivers are moved along the surface, as they

pass over a cavity. Figure 17 shows the variation of the phase velocity, normalized by the free field value, between closely-spaced points on the

surface. The arrows extending from abscissas

(x/d) of 0.5 to 1.5 represent the position of a square cavity with side d and its center at a depth

also equal to d. Similar variations are obtained

for the displacement amplitudes rather than the phases. One must have, however, some idea of

the expected size and depth of the cavity to select an appropriate range of frequencies, or wavelengths.

6.3 Prototype testing

Rosenblad et al. (1994) conducted small-scale

tests using a prototype model to test the ability of

the SASW method to detect near-surface cavities. The model was 1.2 m by 3 m by 0.5 m deep

concrete slab which was composed of

unreinforced Portland cement concrete with a maximum aggregate size of 10 mm. Two square

"tunnels" (elongated cavities), each 50 mm by

50 mm, were cast in the upper portion of the slab. The cavities had crown depths of 25 mm and

70 mm.

SASW testing of the prototype slab was

performed using a "marching" technique such that the source and receivers were moved

incrementally along an array line on the slab. The

array line was oriented perpendicularly to the

longitudinal axis of the cavity. Constant source and receiver spacings were maintained during the marching process. A spacing ratio, d2/d^, of 2 was used (see Figure 1), with values of dx of

76 mm and 152 mm. Measurements from each

test were performed over a frequency range of 0

to 100 kHz and an exponential window was

applied to each record in order to minimize the

effects of reflections from the edges of the slab.

The results in terms of dispersion curves from two

sets of measurements are shown in Figure 18. A

source-receiver spacing of 76 mm was used in each measurement set. The dispersion curve for sound concrete was determined well away from

the cavity. This curve is shown by the dotted line

204
and is typical of tests on intact concrete slabs,

beams, and columns (Bowen and Stokoe 1992;

Kalinski et al. 1994). This dispersion curve

exhibits some fluctuations but shows the general

trend of a constant velocity with wavelength. The

dispersion curve determined when the receivers were centered directly over the shallower cavity is shown by the solid line in Figure 18. For very

short wavelengths, wavelengths less than about

25 mm, the dispersion curve overlaps the one

determined without a cavity. However, for longer

wavelengths, the cavity has the effect of

significantly reducing wave velocities and can be

readily identified by this change. Similar results were also found with the deeper cavity, with the difference being that longer wavelengths were

required before the divergence in dispersion curves could be recognized. Bowen and Stokoe

(1992) and Kalinski et al. (1994) have shown

similar comparisons for the identification of

cracked versus sound concrete in beams and

columns.

7. SUMMARY

Although the SASW method was developed under

the assumptions that the ground surface and

subsurface layers are horizontal and extend to

infinity, the results of numerical approximations

and field measurements show that it is a versatile

tool that can be applied to more complex

geosystems. Examples of validation studies and

the results of numerical approximations and field measurements show that the SASW method can be widely used to accurately determine the

variation of stiffness wherever a free surface

exists. The primary limitation is the depth

sampled which is controlled by the range of

frequencies applied.

8. ACKNOWLEDGEMENTS

Research on the application of surface waves to sloping ground, through a discontinous layer, and

for tunnel and cavity detection was sponsored by

the U.S. Army Engineer Waterways Experiment

Station. Permission was granted by the Chief of

Engineers to publish this information.

9. REFERENCES

Alexander, D. (1992). "In Situ Material

Characterization for Pavement Evaluation by the Spectral-Analysis-of-Surface Waves (SASW)

Method," Tech. Rpt. GL-92-10, U.S. Army WES,

Vicksburg, MS.

Andrus, R., Stokoe, K., II, Bay, J. and Youd, T.

(1992). "In Situ V$ of Gravelly Soils Which

Liquefied, " Proc, 10th World Conf. Earthq. Engr.,

Madrid.

Ballard, R.,Jr. (1964). "Determination of Soil Shear Moduli at Depth by In-Situ Vibratory

Techniques," Misc. Pap. 4-691, U.S. Army WES,

Vicksburg, MS.

Bowen, B. (1992). "Damaged Detection in Concrete Elements with Surface Wave Measurements," PhD diss., Univ. Texas, Austin, TX.

Bowen, B. and Stokoe, K.,II (1992). "Evalua-tion

of Cracked and Repaired Beams and Columns

Using Surface Stress Waves," Proc, 10th World

Conf. Earthq. Engr., Madrid.

Coll-Calderon, J. (1985). "Cavity Detection by the SASW Method: Preliminary Studies," MS thesis,

Univ. Texas, Austin, TX.

Fry, Z. (1963). "Development and Evaluation of

Soil Bearing Capacity, Foundations of Structures," U.S. Army WES, Tech. Rpt. 3-362, No. 1, Vicksburg, MS.

Gauer R. (1990). "Experimental Study of Applying

the Spectral-Analysis-of-Surface_waves Method

Offshore," MS thesis, Univ. Texas, Austin, TX.

Gucunski, N. (1991). "Generation of Low

Frequency Rayleigh Waves for the SpectralAnalysis-of-Surface-Waves. Method," PhD diss.,

Univ. Michigan, Ann Arbor, MI.

Haskell, N. A. (1953). "The Dispersion of Surface

Waves in Multilayered Media, " Bull. Seism. Soc,

Vol 43, No. l,pp 17-34.

Heisey, J. (1982). "Determination of In Situ Shear

Wave Velocities from Spectral Analysis of Surface

Waves, "MS thesis, Univ. Texas, Austin, TX.

205
Heisey, J., Stokoe, K., II, and Meyer, A. (1982).

"Moduli of Pavement Systems from Spectral

Analysis of Surface Waves,Trans. Res. Rec. 852,

Washington, DC, pp 22-31.

Heukelom, W. (1958). "Investigation Into the Dynamic Mechanical Properties of Some Runways

in the U.S.A. Carried Out in 1958,"Rpt. M-35220,

Shell Oil Labs, Amsterdam.

Heukelom, W., Foster, C. (1960). "Dynamic

Testing of Pavements," J. Soil Mech. and Found.

Div.. ASCE, Vol 86, No. SMI, pp. 1-28.

Hiltunen, D. (1988). "Experimental Evaluation of

Variables Affecting the Testing of Pavements by

the Spectral-Analysis-of-Surface-Waves Method,"

U.S. Army WES, Tech. Rpt. GL-88-12, Vicksburg,

MS.

Jones, R. (1958). "In-Situ Measurement of the Dynamic Properties of Soils by Vibration

Methods," Geotechnique, Vol 8,pp 1-21.

Kang, Y. (1990). "Effect of Finite Width on Dynamic Deflections of Pavements," PhD diss.,

Univ. Texas, Austin, TX.

Kalinski, M. (1994). "Non-destructive Testing of Concrete Structural Elements Using the SASW

Method," MS thesis, Univ. Texas, Austin, TX.

Kalinski, M., Stokoe, K., II, Jirsa, J. and Roesset,

J. (1994). "Nondestructive Identification of

Internally Damaged Areas of a Concrete Beam

Using the SASW Method," Mtg..Transp. Res. Board, Washington, DC.

Kausel, E. (1981). "Explicit Solution for the Green

Functions for Dynamic Loads in Layered Media,"

MIT Res. Rpt. R81-13, Cambridge, MA.

,(1989). "PUNCH: Program for the Dynamic Analysis of Layered Soils," ver. 3.0,

Massachusetts Inst. Technology, Cambridge, MA.

Lopez, M. (1989). "Investigation of Variations in Rock Stiffness by the SASW Method Following

Excavation by Blasting and Mining, "MS thesis,

Univ. Texas, Austin, TX.

Luke, B. (1994). "In Situ Measurement of Stiffness Profiles in Seafloor Using Spectral-Analysis-ofSurface-Waves Method," PhD diss., Univ. Texas,

Austin, TX, (in progress).

Madianos, M. (1991). "Field and Laboratory

Investigation of Rock Masses Using Surface Wave

Testing (SASW)," MS thesis, Univ. Texas, Austin,

TX.

Nazarian, S. (1984). "In Situ Determination of

Elastic Moduli of Soil Deposits and Pavement

Systems by Spectral-Analysis-of-Surface-Waves

Method," PhD diss., Univ. Texas, Austin, TX.

Nazarian, S. and Desai, M. (1993). "Automated

Surface Wave Method: Field Testing,

J. Geotech.

Engr., ASCE, Vol 119, No. 7, pp 1790-1804.

Nazarian, S. and Stokoe, K.,II, (1985a). "In Situ Determination of Elastic Moduli of Pavement

Systems by Spectral-Analysis-of-Surface-Waves

Method (Practical Aspects)," Res. Rpt. No. 368-

1F, Cntr. Transp. Res., Austin, TX.

, (1985b). "In Situ Determination of

Elastic Moduli of Pavement Systems by SpectralAnalysis-of-Surface-Waves Method (Theoretical

Aspects)," Res. Rpt. No. 437-1, Cntr. Transp.

Res., Austin, TX.

Nelson, P., Lopez, M., Stokoe, K.,II, and Roblee, C. (1989). "In Situ Evaluation of Rock Mass

Stiffness Variations for Underground Openings

Using Surface Waves and Boreholes Seismic

Methods," Proc.,7th Workshop, Mines Systems

Design and Ground Control, Blacksburg, VA.

Nogueira, E. (1986). "Effects of Cavities on the Propagation of Surface Waves," MS thesis, Univ.

Texas, Austin, TX.

Olson, L.,Sack, D., Stokoe, K.,II, and Buchinski, K. (1993). "Stress Wave NDT of Tunnels/Shafts,"

Mtg., Transp. Res. Board, Washington, DC.

Rix, G. (1988). "Experimental Study of Factors

Affecting the Spectral-Analysis-of-Surface-Waves

Method," PhD diss., Univ. Texas, Austin, TX.

206
Rix, G. and Leipski, E. (1991). "Accuracy and

Resolution of Surface Wave Inversion." Recent Advances in Instr., Data Acq. & Test, in Soil Dvn.. ASCE. Geot. Pub. No. 29, ed. Bhatia and

Blaney, pp 17-32.

Rosenblad, B.,Stokoe, K.,II, Chiang, C.,and

Roesset, J. (1994). "Near-Surface Cavity Detection

Using the SASW Method," Geot. Engr. Rpt.

GR-94-2, Univ. Texas, Austin, TX.

Sanchez-Salinero, I. (1987). "Analytical Investi- gation of Seismic Methods Used for Engineering

Applications," PhD diss., Univ. Texas, Austin, TX.

Sedighi-Manesh, M. (1991). "Theoretical Investi- gation of Spectral-Analysis-of-Surface-Waves

(SASW) Technique for Application Offshore,"

PhD diss., Univ. Texas, Austin, TX.

Sheu, J. (1987). "Applications and Limitations of

the Spectral-Analysis-of-Surface-Waves Method,"

PhD diss., Univ. Texas, Austin, TX.

Sykora, D. (1993). "Dynamic Vertical Displace- ments in Planar Geosystems," PhD diss., Univ.

Texas, Austin, TX.

Figure 1 . Schematic of SASW method using

harmonic source

Sykora, D., Alexander, D.,and Roesset, J. (1994).

"Determination of Soil Moduli in Soil-Structure Systems on Highways; Report 1: Surface Waves in Sloping Ground," Tech. Rpt. (in publ.), U.S. Army WES, Vicksburg, MS.

Sykora, D. and Roesset, J. (1992). "2-D Planar

Geosystems Subjected to 3-D Dynamic Loads,"

Tech. Rpt. GL-92-16, U.S. Army WES, Vicksburg,

MS.

Thomson, W. (1950). "Transmission of Elastic Waves Through a Stratified Solid Medium, " L

Sound Vib.. Vol 2, No. 3, pp 210-226.

Figure 2. Schematic of SASW method using

impulsive source

6a. Q

Sandy

Gravel with few thin sandy silt layers

Extent of sampling

4

-

water

table, 8/91

SASW

Crosshole

Soil

Profile

100 200 300 Shear Wave Velocity, Vs , m/sec

Figure 3 . Shear wave velocity profile determined

at site where gravelly soils liquefied during the 1983 Borah Peak Earthquake (Andrus et al. 1992)

207
10

15

AC

BASE

0.5

I 5

2 0

2.5

r

'

I

<

c

Test Method

r —SASW

• Crosshole

C

•I

.J i i

Assumed

Layering

Material

Profile

300 600 900 1200

Shear Wave Velocity Profile, m/s

Figure 4. Comparison of shear wave velocity

profiles at a pavement site (Nazarian and Stokoe

1985b)

B

8

U

GO

m

-5cx a

0 100 200 300 400

Shear Wave Velocity, m/s

Figure 5. Comparison of shear wave profiles

from SASW and seismic cone penetrometer

(SCPT) measurements offshore Venice, Italy

(Luke 1994)

PHASE VELOCITY, C m 700 900 U00 1300

Figure 6. Dispersion curves for surface waves

travelling over and away from high-velocity buried

layer (Sykora 1993)

XLOAD e —

< r

-Ay

l

w

Figure 7. Variables used to define sloping ground

problem (Sykora 1993)

Level Ground

2.0 3.0 4.0 SURFACE DISTANCE (in X )

Figure 8. Displacements and phase for a — 146 degrees (Sykora 1993)

208
— a. o.o

1 ' : r- "

:

* V

*- vt

i i

1

A v 1

0.0 1.0 2.0 3.0 4.0 5

a = 135 to 225

|9= 0.05

x lc /X = 100 / = 3.0 Hz XLOAD/X = ± 0.01

0)p

<

SURFACE DISTANCE (in X )

X-DIRECTION

Figure 9. Displacements and phase for a = 214 degrees (Sykora 1993)

<x

a

x "

Ql 5? _ a> Z— x> u I z e

Q

1.0 2.0 3.0 4.0 SURFACE DISTANCE (in X )

a = 214' * Lt A = 100 XLOAD/X = ± 0.01

/3 = 0.05 / = 3.0 Hz

O j-di-recti.

SURFACE DISTANCE FROM DISCONTINUE i (in X)

Figure 11. Comparisons among the variations of

phase with distance for different slope angles less than 180 degrees (Sykora 1993)

(/i <X

a

-2500

Y-DIRECTION

0.0 1.0 2.0 3.0 4.0 5.0 DISTANCE FROM CENTER OF LOAD (in X)

1.0 2.0 3.0 4.0 SURFACE DISTANCE (in X )

Figure 10. Difference in phase between sloping

ground and uniform increase of 360 degrees per wavelength (Sykora 1993)

Figure 12. Comparisons among the variations of

phase with distance for different slope angles

greater than 180 degrees (Sykora 1993)

209
•

• 5.0 10.0 15.0 SURFACE DISTANCE (ft)

Figure 13. Displacements for measurements down slope (Sykora 1993)

"J Q

1500

Shear Wave Velocity, m/s

Figure 15. Shear wave velocity profile determined

by SASW testing along an exposed rock wall at the underground WIPP facility (Nelson et al. 1989)

0.0 5.0 10 0 15.0 SURFACE DISTANCE (ft)

Figure 14. Phases for measurements down slope

(Sykora 1993)

210
0

0.5 _

1.0

1.5

2.0

Concrete Liner T

Rock

I

1 900 1800 2700

Shear Wave Velocity, m/s

3600

Figure 16. Velocity profile determined from

SASW tests performed on a concrete shaft liner (Olson et al. 1993)

Figure 17. Variation of phase velocity as a function of normalized distance for a half-space

with a square cavity of width d at a depth d

(Noguiera 1986)

2500

S 2000

o

_c>

>

I 1500 -

1000

Sound Concrete

Cavity (Crown depth at 25 mm)

JL 0.0 0.1 0.2 Wavelength, m

0.3

Figure 18. Comparison of experimental

dispersion curves determined by SASW testing on

sound concrete and over a shallow cavity in concrete (Rosenblad et al. 1994)

211

Effectiveness of Measures Against Liquefaction

Confirmed at a Recent Earthquake—A Case History

During 1993 Kushiro-Oki Earthquake

by

S. Iai*, Y. Matsunaga*, T. Morita* and H. Sakurai*

ABSTRACT

During the 1993 Kushiro-Oki

Earthquake of magnitude 7.8,

Kushiro Port being located at

15 km from the epicenter was

shaken with a peak horizontal

acceleration of 0.47g. Many of

the quay walls at the Kushiro

Port suffered damage due to

liquefaction of backfill sand.

However, the quay walls having

backfill sand compacted as measures against liquefaction survived the earthquake without

damage. Since the contrast in

the performance of those quay

walls are of significance in

the engineering practice, the

present paper is devoted to

report the details including

the recorded earthquake motions

and the ground conditions.

Key words : liquefaction, sand,

earthquake, quay walls

INTRODUCTION

In recent Japanese practice,

measures against liquefaction

have often been put into practice in order to mitigate damage to structures sitting on

potentially liquef iable foundations. While laboratory data

are abundant to evaluate the

effectiveness of those measures

against liquefaction, in-situ

data during strong earthquake

shaking are very scarce.

On January 15, 1993, an earthquake of magnitude 7.8 in the

scale of Richter- Japan meteorological Agency (JMA) hit a northern island of Japan called

Hokkaido. The epicenter was

located 15 km south of Kushiro

city as shown in Fig. 1; the

earthquake was named the 1993

Kushiro-Oki earthquake. The

focal depth was 107 km, being

much deeper than the ordinary

depth of about 40 km for the

earthquakes occurring around

Japan.

Kushiro Port was shaken with a

peak horizontal acceleration of

0.47g. The earthquake caused

liquefaction of untreated backfill sand at the port, resulting in damage to many of the

quay walls. The extent of the

damage was generally moderate;

the displacements of the quay

walls were less than about 5

percent of the water depth in

front of the quay walls. In

contrast, the quay walls having

the backfill sand compacted as

measures against liquefaction

survived the earthquake without

damage

.

In the present paper, the performance of these quay walls

will be reported in detail including the recorded earthquake

motions and the ground conditions .

* Geotechnical Earthquake Engineering Laboratory, Port and

Harbour Research Institute

213
KUSHIRO PORT

Kushiro Port is developed at

the estuary of the Kushiro River as shown in Fig. 2. As understood by the former beach

line shown by the broken line

in Fig. 2, most of the port

area has been reclaimed from

the sea by filling the sand

dredged from the nearby sea.

The port has been constructed

step by step from a quay wall

located at the eastern most end

toward west direction along the

former beach line; the quay

wall at the western most end,

i.e. Pier No. 3 in the West Port

District, has been just constructed in 1992.

The natural ground along the

former beach line is of sand

dune origin, being dense with

SPT N-values of about 50. The

fill above the original ground

gradually increases its thickness from none at the former

beach line to about ten meters

at the southern end of the

piers in the West Port District . Though Kushiro City is located

at the latitude 43 degrees

north, at about the same as

that of Rome in Italy, the average temperature falls below

the water freezing point during

winter. When the earthquake

hit, the ground surface was

said to be frozen with a thickness of about 0.5 to 1.0 meters

in Kushiro City. However, the

port engineers reported that

the materials used for the

ground surface layer in the

port areas were chosen to serve

as anti-freezing layer to prevent the damage to the pavement

due to the soil freezing. The

anti-freezing layer is made of

coarse sand, its thickness

about 0.8 meters. At present,

the authors could not obtain

the definite evidence to determine whether the ground surface

in the port area was frozen at

the time of the earthquake.

When the earthquake occurred,

the ground at Kushiro City was

not covered with snow. It began

to snow the next day and the

ground was covered with snow

with a thickness of about 0.2

meters

.

EARTHQUAKE MOTION

AT KUSHIRO PORT

Kushiro Port has a strong motion recording station at its

central location. This is one

of the station deployed under

the network of the Port and

Harbour Research Institute in

cooperation with the Kushiro

Port Construction Office,

Hokkaido Development Bureau

(Kurata and Iai, 1992).

The strong motion recording

station is located on the former beach line at the estuary

of the Kushiro River as shown

in Fig. 2. The ground condition

is shown in Fig. 3. In this

figure and the figures to follow, all the borings and the

Standard Penetration Tests

( SPT ) were conducted before the

earthquake. The ground at this

site consists of a fill with a

thickness of about three

meters, underlain by a medium

to dense sand layer of SPT re- values about 20 with a thickness of about seven meters,

underlain by a dense sand layer

of SPT N-values of about 50.

The instrumentation at this

site consists of two accelerographs; one is installed at the

ground surface, the other installed at the base of the bore

214
hole 77 meters below the ground

surface. The accelerographs are

of force balance servo type,

having the flat gain in the

freguency range from 0.01 to 30

Hz. The acceleration data are

recorded in the digital format

in the integrated circuit (IC)

cartridge memory with 16

bits/data at the time intervals

of 0.01 seconds. The digital

recorder is eguipped with a pre-event memory to record

earthguake motions from ten

seconds before triggering.

The recorded earthguake motions

during the 1993 Kushiro-Oki

Earthguake are shown in Fig. 4.

In this figure, main portion of

the earthguake motion is plotted out of the original record

obtained for a duration of 180

seconds; the origin of the time

is ten seconds before the triggering of the recorder. Loci of

integrated velocities and displacements computed for the

accelerations at the ground

surface and the base, shown in

Fig. 5, clearly indicate that

the predominant motion is in

the north-south direction in

this earthguake.

By comparing the ground surface

motion in the north direction

shown in the third row in Fig.

4 with the corresponding base

motion at a depth of 77 meters

shown in the bottom row, it is

easy to notice the sudden

change in the response of

ground after about 30 seconds;

most of the high freguency motions are filtered out; instead, a long period motion

with a period of about 1.5 seconds becomes predominant, overlain by a spiky wave form at

each peak.

This type of ground response

has been often observed in the

laboratory (e.g. Lee and

Schofield, 1988); it is considered as manifestation of the

effect of cyclic mobility of

dense sand. The effect of the

dilatancy of sand plays a major

role when the ground is saturated and shaken with a strong

motion as in this earthguake.

In in-situ condition during

earthguakes, similar spiky acceleration was observed at

loose silty sand deposit at the

Wildlife site during the 1987

Superstition Hills Earthguake

(Holzer et al, 1989 ). In the

dense clean sand deposit, the

present record represents the

very evidence that the effect

of dilatancy of sand plays a significant role in the ground

response during strong earthguake shaking.

In the current practice of

ground response analysis such

as often done with the eguivalent linear model, the effect

of the dilatancy of sand has

been often neglected. The definite evidence presented here

cautions us against the current

practice of ground response

analysis at strong earthguake

motions

.

DAMAGE TO GRAVITY QUAY WALLS

The most serious damage to the

gravity guay walls in Kushiro

Port was found at the North

Pier in the East Port District.

The location is designated in

Fig. 2 as the site A. As shown

in Fig. 6, displacement toward

sea registered 2.0 meters at

maximum, settlement 0.4 meters

at maximum. The ground conditions shown in Fig. 7 indicates

possibility of liguefaction

down to an elevation of about -

6 . 0 meters below the water level

.

215
Among the gravity quay walls

affected by liquefaction of

back fill sand, quay walls designed with a seismic coefficient of 0.2 suffered relatively minor displacement. As an

example is shown damage to a

quay wall with a water depth of

12 meters located at the south

end of the Pier No. 2 in the

West Port District. The location is shown in Fig. 2 designated as the site B. As understood from the height to width

ratio of the caisson being

close to unity, the quay wall

was designed to resist relatively strong earthquake motions . The backfill sand at this pier

was loosely deposited as shown

in Fig. 8 and was of a coarse

sand as shown in Fig. 9. It was

planned to be compacted as measures against liquefaction;

unfortunately the earthquake

hit the Kushiro area before the

work was actually done. The

back fill sand liquefied due to

the earthquake. The caisson

suffered deformation as shown

in Fig. 8 but the displacement

toward sea was only about 0.3

meters

.

DAMAGE TO A SHEET PILE

QUAY WALL

The most serious damage to the

sheet pile quay walls in

Kushiro Port was found at the

Fishery Pier in the East Port

District. The location is designated in Fig. 2 as the site

C. The structure of this quay

wall is of a steel sheet pile

type anchored by battered steel

piles. The cross section is

shown in Fig. 10. As shown in

Fig. 11, the ground at this

pier consists of a loosely deposited back fill sand with a

thickness of about ten meters,

underlain by medium to dense

sand deposit which forms the

original ground.

The original ground level becomes deeper to the level of

11.5 meters at the eastern end

of the quay wall so that this

part of the quay wall slightly

differs from that shown in Fig.

10; the quay wall at the eastern end uses longer sheet piles

with an embedment depth of - 20.5 meters and longer anchor

piles installed to a depth of - 14.9 meters on the sea side and

-20.6 meters on the land side.

The earthquake caused liquefaction in the back fill sand,

resulting in the serious deformation of the sheet pile wall

as shown by the solid line in

Fig. 10. In accordance with the

deformation of the sheet pile

wall, the apron exhibited serious settlement between the

sheet pile wall and the anchor

piles. Detailed investigation

of the sheet pile wall by diving in the sea revealed that

cracks opened in the steel

sheet pile wall at the level of

four meters below the water

level

.

PERFORMANCE OF A SHEET PILE

QUAY WALL WITH

COMPACTED BACK FILL SAND

In contrast to the damaged

sheet pile quay wall described

earlier, quay walls with compacted back fill sand survived

the earthquake without damage.

An example is the quay wall at

the south end of the Pier No.l

at the West Port District. The

location is designated as the

site D in Fig. 2.

The structure of this quay wall

216
is of a steel pipe pile wall

anchored by a steel sheet pile

wall as shown in Fig. 12. The

water depth of this wall is 12

meters. The ground at this wall

had initially consisted of fill

sand with a thickness of about

ten meters, underlain by the

original ground of medium to

dense sand. The back fill sand

was later treated by a sand

compaction pile method as shown

in Fig. 12. The SPT N-values of

the ground before and after the

compaction are shown in Fig.

13. In order to avoid the adverse effect upon the existing

steel pipe pile wall during the

ground compaction work, gravel

drains were installed near the

wall as shown in Fig. 12.

Despite the strong shaking of

the earthguake, the earthguake

caused no effect on this guay

wall. The measures against liguefaction as well as the overall design of the guay wall

according to the current design

practice ( Ports and Harbours

Bureau, 1991) was proven to be

good enough to provide resistance the strong earthguake

shaking

.

LIQUEFACTION AND DAMAGE

TO STRUCTURES

In order to grasp an overall

picture of the effects of liguefaction on damage to guay

walls at Kushiro Port, Table 1

summarizes the relevant data

for guay walls with water

depths deeper than 7.5 meters.

In this table, " liguefaction"

is identified with respect to

the soils within "the area of

a guay wall cross section,

shown in Fig. 14, which includes the back fill area treated as measures against liguef action. This similarly applies

to guay walls without measures

against liguefaction, including

the back fill area which is

supposed to be treated as measures against liguefaction.

As shown in Table 1, the guay

walls designed with a seismic

coefficient of 0.15 suffered

serious sea ward displacement

of more than 20% of the water

depth. Both the excessive

strong shaking and the liguefaction might be the cause for

this. In comparison to these

guay walls, the guay walls designed with a seismic coefficient of 0.20 suffered sea ward

displacement of less than 10%

even when liguefaction occurred .

A closer look at the displacements of the guay walls with a

seismic coefficient of 0.20

reveals that those without measures against liguefaction suffered displacement ranging from

0.5 to 10% whereas those with

measures against liguefaction

suffered displacement less than

3%, suggesting the measures

against liguefaction reduces

displacement of guay walls.

Among these features observed

above, the effectiveness of

measures against liguefaction

to reduce displacement of guay

walls is what was expected

through laboratory studies; the

in-situ evidence observed here

is the very confirmation of

this fact. The effects of seismic coefficient on displacement

of guay walls are also what was

expected from the balance of

inertial forces and resistance

force of guay walls. However,

the effects of seismic coefficient also include the resistance against the effects of

liguefaction. The larger resis217
tance of gravity body, which is

massive and large if designed

with seismic coefficient of

0.20, may contribute to resist

the increase in earth pressures

due to liquefaction as well.

From the point of view of measures against liquefaction, the

back fill part of ground consisting of rubble behind a wall

shown in Fig. 14 might be regarded as treated area as measures against liquefaction.

Though the area of this part is

not wide enough to cover whole

area of active earth failure

zone, this part still contributes to reduce the effects of

liquefaction. Indeed, past case

histories of performance of

quay walls during earthquakes

suggests that the effects of

rubble can be regarded to reduce liquefaction induced displacement of quay walls

(Tsuchida, 1981).

CONCLUSIONS

The present paper was devoted

to compile the case history

data on the performance of quay

walls during the 1993 KushiroOki Earthquake. The analyses

are currently underlay and will

be reported elsewhere. Implications from the present case

history data can be summarized

as follows.

( 1 ) The strong motion record

with a peak acceleration

of 0.47g at Kushiro Port

indicates that the effect

of dilatancy of sand

plays an important role

in the ground response

during strong shaking.

( 2 ) Performance of the quay

walls at Kushiro Port

manifest the importance

and the effectiveness of

the measures against soil

liquefaction in the earthquake resistant design

of waterfront structures.

REFERENCES

Holzer, T.L., Youd, T.L. and

Hanks, T.C. (1989) : "Dynamics of liquefaction

during the 1987 Superstition Hills, California

Earthquake, Science 244,

pp . 56

Kurata, E. and Iai, S. (1992)

: "Annual report on

strong motion earthquake

records in Japanese ports

(1991)," Technical Note

of the Port and Harbour

Research Institute,

No. 727

Lee, F.H. and Schofield, A.N.

(1988) : "Centrifuge modelling of sand embankments and islands in earthquakes, " Geotechnique , Vol.38,

No.l, pp. 45-58

Ports and Harbours Bureau (ed.

)

(1991) : "Technical standards for port and

harbour facilities in

Japan, " Overseas Coastal

Area Development Institute of Japan

Tsuchida, H. (1981) : "Damage

to quay walls affected by

liquefaction, " 16th Research Conference of Earthquake Engineering,

JSCE, pp. 201-204 (in Japanese )

218
Table 1 Liquefaction and deformation of quaywalls

quaywalls structure

water- depth

(m)

seis

mic

coef

f ici ent

displa

cement toward

sea

(cm)

settle

ment

of

apron

(cm)

settle

ment behind

apron

(cm)

liq- ue- faction

Pe»ntf q1 WTi;ar"f Pact Giinvwal 1 V/Clllldl nlldl 1 Ddol yUdjndll steel sheet pile -7 5 0. 20 n 0 —1 0

^oJIlldl nlldl 1 DdoL yUdj Wai 1 steel frame -7. 5 0. 20 n 0—10

P^nti-al Viharf Pact Qnavu/al 1 ^/CllLlal nlldl 1 Laol yuaj Wall steel sheet pile -7. 5 0. 15 n 0—10 - cell (under renovation)

W/*»t + Vi WViot"F Poof OiiqvujgI 1 l\Ui III filial I Ldo I WUdyWdll gravity type -8.

1

0. 15 a. 200 North Wharf South Quaywall gravity type -8.0 0. 15 50- 20- e

1 fiO 1 00 Wnrth liUlLll Wharf Wp^t Gnavu/al 1 it lldl 1 lieu L v^Uaj'Ttaii gravity type -9.0 0. 15 1 00~ 10—30

160 Fi^hprv Wharf South Quavwall steel sheet pile -7. 5 0. 20 0—30 30 •

Wharf No 1 Fast Quavwal

1

jointed steel -9.0 0. 20 o o 0 o

pipe wai l Wharf No. 1 South Quaywall jointed steel -12.0 0. 20 0 0 0 O

m no mill pipe wai

l

Wharf No. 1 West Quaywall jointed steel -10.0 0. 20 -25—0 20-30 0 o

pipe Wdl

1

Wharf No. 1 West Quaywall gravity type -9. 0 0. 20 5-70 20-30 10-35 •

Wharf No. 1 West Quaywall gravity type -9. 0 0. 20 20-75 10-35 10-35

Wharf No. 2 East Quaywall gravity type -7. 5 0. 20 10 10-35 0-10 •

Wharf No. 2 East Quaywall gravity type -10.0 0. 20 5 25-30 0-10 m

Wharf No. 2 South Quaywall gravity type -12.0 0. 20 20-30 30-45 20-70 •

Wharf No. 2 West Quaywall gravity type -9.0 0. 20 20-30 30-40 10-40 #

Wharf No. 2 West Quaywall gravity type -7. 5 0. 20 10-25 10-30 0-30 •

Wharf No. 3 East Quaywall gravity type -7. 5 0. 20 0-15 10-15 20-30 ©

Wharf No. 3 South Quaywall gravity type -12.0 0. 20 -7—6 10-20 5-35 ©

Wharf No. 3 West Quaywall gravity type -12.0 0. 20 5-25 10 10—40 ©

Wharf No. 3 West Quaywall gravity type -10.0 0. 20 15-25 10-15 5 ©

* O : Non-liquefaction (no need for measures against liquefaction)

© : Non-liquefaction (with measures against liquefaction)

# : Liquefaction (without measures against liquefaction)

219
(a) Location of Kushiro (b) Closer View

Fig. 1 Location of Kushiro City and the epicenter

Fig. 2 Plan of Kushiro Port

220
Fig. 3 Ground condition at the strong motion recording station

221
500 , (Gal)

250 \- 0

-250

-500

500

Ground Surface

(Gal)

: Ground Surface

I Ground Surface

(Gal)

250

0

-250

-500

300

200 |lBase(-77m)

100

i 0

-100

-200

-300

300

200 - Base(-77m)

100 - 0 [—^^4^^

-100

-200

-300

300

200

100

0

-100

-200

-300

(Gal)

East (Max. 344.2)

Up (Max. -381.5)

North (Max. -469.3)

East (Max. -262.1)

Up (Max. -122.1)

(Gal)

LBase(-77m) North (Max. 204.2)

mm

10 20 30 40

Time (s)

50 60

Fig. 4 Strong earthguake motions at Kushiro Port on Jan. 15, 1993

222
a

EoO

opz

500

250 -

0

-

Acceleration (Gal)

-250

-500

-500 -250 0 250 E-W Component

500

80

'. D

-40

Velocity (klne)

-80

-80

Surfac

•

1 • j —

—.—i—i—

i

. i

. . . . . . . . . -40 0 40 E-W Component

80

Displacement (cm) T

-10 0 10 E-W Component

20

300

150

-150

Acceleration (Gal)

-300

. Base

, ,

|

.

,

-300 -150 0 150 300 E-W Component

20

i Z

•10

Velocity (klne)

-20

II''' Base :

, . .

. . , i . . . i . . . . . . -20 -10 0 10 E-W Component

20

10 Displacement (cm)

10

....

Base ....

.... . . .... . . . . -10 -5 0 5 E-W Component

10

Fig. 5 Loci of accelerations, velocities and displacements

7 270

V+Z5Q

y H.W.L + 1.50 tl.5Q

[

Elevation

Before Earthquake ( m)

After Earthquake ^ Coarse

Sand

SPT N-value

0 10 20 30 40 30

10-

Fig.6 Damage to a quay wall at Site A

-20-

Gravelly\

Sand |

Fine

Sand

Coarse

Sand

SandyGravel

Coarse

Sand

SandyGravel

p t

• 4

Fig. 7 Ground Condition

at Site A

223
20.0

14 0

+ 3.0 0.407

I 6.0

0.860 __L__

0.723 Unit (m)

V_ H.W.L + 1.5 V L.W.L t 0.0 —

-12.0 -12.1 fn0.26

Caisson

L B H

I5.0xl4.0x 13.6

15.6, Rubble Stones

Rubble

Stones

before earthquake

after earthquake

X

Filled Sand

0 rigina I Ground

Depth

(m) 0

-10

-15

Soil Type | SPT- N Value 0 10 20 30 40 50

\Q rove I

Coarse Sand

mixed

with Grovel

J Si It Fine

\ Sand

Fine Sand

Fig. 8 Damage to a guay wall at Site B Fig. 9 Ground Condition

at Site B

200 Unit (m)

V H.W.L + 1.5 V LW.LiO.O

10

-15

before earthquake

after earthquake

Depth

( m)

i

°4

Soil Type I SPT-N Value 0 10 20 304050

Elevation 2.34 m

Sand

/Gravelly

Sand

Fine Sand

Gravelly Sand

Fine Sand

Coarse Sand

Sand

and

Grovelt

"*l.25

Fig. 10 Damage to a guay wall at Site C Fig. 11 Ground condition

at Site C

224
^_ H.WLfl.5

T L.W.L±0.0

300

20 0

.-v .-v

.- - I 2 0

-22.7

rovel mot d = 0.3Q

3 4

Tie-Rodctcl.98

Grovel Droin

•2.0

<t> = 400mnr

_____-_75

Sond Compoction Pile

£

>

<£=700mm

- I 2.0

Unit(m)

Fig. 12 Cross section of a quay wall

at Site D

Depth (m) ;0

- 5

-10

-15

20

Soil Type | SPT- N Value 01020 30 40 50

Elevation"! 2.94m J Coarse Sand

Coarse Sand

mixed with

Gravel

Fine Sand

Fine Sand

ond Silt Fine Sand

mixed with

Shells

Shelly Sond

Coarse Sand

mixed with Gravel

Coorse Sand before compaction :

Fig. 13 Ground condition

at Site D

Area of Quay Wall

Area of Quay Wall

Caisson

- HWL*I50 "L.WLtOOO

10_LH

Area of Ground

Treatment

Area of Ground Treatment

188 —3 ? - HWLH50

- LWL;000

Drain Sand

Compaction

Fig. 14 Schematic figures of the area" of quay walls

225

Review of Progress in Dynamic Geotechnical Centrifuge Research

by

Ronald F. Scott

ABSTRACT

A great deal of research has been performed in the last ten years in the area of dynamic

geotechnical centrifuge testing. The progress

made is reviewed, with particular reference to papers presented at this Symposium, problem

areas are identified, and suggestions are made to

clarify some of the difficulties. KEYWORDS: centrifuge; dynamic testing;

earthquake; soil.

1. INTRODUCTION

In the past decade, activity in dynamic geotechnical centrifuge research has increased very

substantially; new apparatus has been developed, instrumentation and data acquisition has

become more complex, input excitation has

been tailored to represent more realistic earth- quake strong ground motions, and a great vari- ety of model geometries has been explored.

This review paper attempts to give a summary

of this effort, the progress made, and improvements which, in the author's view, still need to be made.

2. PROGRESS TO DATE

Dynamic centrifuge testing has become virtually

routine at a number of research institutions. The

progress that has been made may be classified

into a number of different areas.

2.1 Input Motions

In the early stages of dynamic centrifuge testing

it was only possible to submit model structures

in the centrifuge environment to very simple

dynamic motions such as impact or shock tests or simple vibratory motion caused by compressing and releasing a spring to shake the specimen

container. The frequency and ring-down duration depended on the mass and spring constants

involved. There were also a number of investigations of explosion-generated shock waves in the centrifuge for the purpose either of shaking a

structure, or to study the explosive tests themselves. In some tests of this character the explosive charge was used to form a crater in the soil medium during centrifuge rotation, for the

purpose of studying the crater depth and di- ameter for different locations and sizes of explosive charge (Schmidt and Holsapple 1980).

The possibility of dynamic testing using elec- tromagnetic means has been examined (Fuji

1991); such equipment had been used in laboratory and centrifuge testing for material other

than soil, for example, structures, telemetry

systems, or components of missile systems.

Electromagnetic shaking has a number of desirable features including high-frequency activation of the shaking table, and the ready control

of frequency content and amplitude. However,

the equipment is subject to a number of drawbacks such as its substantial mass. This is generally unsuitable for testing soil specimens,

which form payloads inherently large in mass. Another system of shaking developed was the

so-called "bumpy road" at Cambridge

University to subject samples to approximately

10 cycles of a rather noisy sine wave motion.

The advantage of the device is its simplicity and

in its original form it was eventually intended to include replaceable bumpy road tracks on the

wall of the centrifuge containment. With these,

different frequency contents could be included

in the motion such as, for example, earthquakelike motions, but only two different tracks were

ever produced. The disadvantage of the bumpy

* California Institute of Technology

Pasadena, California 9 1 1 25

227
road motion is that the developed frequency

input depends on the rotation rate of the centri- fuge, since the frequency is controlled by a cam

riding on a bumpy track of fixed geometry on

the wall of the centrifuge. Thus, the faster the

centrifuge goes, the higher the frequency that is obtained. However in most seismic centrifuge

testing, the desirable frequency is set by the

scale of the model and a natural event which is to be simulated. In the centrifuge, the model

scale is set by the g-level, which depends on the

square of the centrifuge rotation rate whereas, as pointed out above, the excitation frequency of

the bumpy road depends on the rotation rate

directly. This does not permit the necessary

flexibility in the test input.

Finally, an electrohydraulics shaker was developed for the centrifuge (Aboim et al. 1983);

these are commonly used for operation of laboratory shake tables of all sizes. The electrohydraulic apparatus had to be adapted for centri- fuge tests because of the relatively small duration (0.3 to 0.5 seconds) of earthquakes in centrifuge testing at high g-levels. Thus, the actuat- ing servovalve has to supply a surge of hydraulic fluid at a high flow rate to the hydraulic

pistons that actuate the shaking table to get the

high frequencies demanded by the compressed

acceleration histories required. This is accomplished by mounting hydraulic accumulators on

the centrifuge in which hydraulic fluid is stored

at a pressure of about 3,000 psi in sufficient

quantity to conduct the test without additional

flow of hydraulic fluid from outside the centri- fuge arm. These accumulators have generally

been mounted on the arm, which is probably the

most efficient location for them, but in some

centrifuges with a lack of space available for arm mounting, the accumulators have been

placed on the centrifuge bucket itself; this considerably reduces the space available for the

shake table. An advantage of the electrohydraulic apparatus is that the nature of the motion

imparted to the specimen is readily controllable.

It is possible to supply the shake table with a

variety of motions of selectable durations

ranging from sine, square or triangular waves, to random excitation, and earthquake-like excitation, which is essentially random motion

shaped in terms of amplitude as a function of

time. There are difficulties in obtaining suitably

high scaled frequencies on the centrifuge to simulate the complete range of earthquake

motions. A number of these electrohydraulic

shakers are now in operation on various centri- fuges.

In addition to this equipment, the shake tables

themselves have developed in a variety of types.

Some have been suspended by tension members

from overhead beams, others are mounted on

roller bearings or elastomeric pads which permit

greater visibility of a cross section of the specimen if the side of the sample container is made

of transparent material.

2.2 Tests Performed

A number of numerical codes has been developed with the object of predicting the pore

pressure increases in a saturated fine sand as a

result of dynamic or earthquake shaking. These

numerical codes have been developed in a

virtual absence of measured field data on the behavior of such soils in earthquakes. They are lacking because, of course, the occurrence of

earthquakes is essentially random and there

have been very few sites at which pore pressure

transducers have been installed in ground, which

has subsequently been shaken by an earthquake.

It was proposed to remedy this deficiency by

supplying the necessary data by centrifuge tests

in the "VELACS" project (Verification of

Earthquake Liquefaction Analyses by

Centrifuge Studies) (Arulanandan and Scott

1993). As described by Arulanandan et al (1994), the required experiments have been

conducted by eight universities on a variety of

centrifuges. The testing arrangement among the

universities was to shake saturated soil speci- mens in models of controlled geometric configuration, arranged so that they would either

undergo high pore pressure increases or liquefaction during earthquake-like events. The

results from these tests could be used to verify

numerical calculations performed with the same

input applied to models with the same simulated

numerical properties as the material used in the

centrifuge tests. During preliminary studies for

228
the performance of the tests to be carried out, the question arose as to the degree of difference

between test results carried out on different

centrifuges, of different sizes, by different in- vestigators at the various universities. If substantial differences in test data from "identical"

tests were actually discovered, these results would mislead comparisons with numerical in- vestigations. Consequently, it was decided to perform one uniform centrifuge test at all universities and to compare the results from that

test to see what sort of differences could be

expected. The paper reports the results of that

test and compares and contrasts the information

received from the different investigators.

There have been a number of tests done to in- vestigate the behavior of soils subjected to explosions, which have been generated on a

centrifuge by means of small plastic explosive

charges set on the surface or below the surface

of the soil specimen Schmidt and Holsapple

1980; Gill and Kuennen, 1991). Some of these

tests have been performed in order to examine

the stresses generated on structures as a result of

the pressure waves generated; these have

required strain gauges, load cells, and pressure

transducers attached to the structures. The great

advantage of the centrifuge apparatus for

explosion tests consists in the scaling of energy

implicit in the scaling relations. The energy

scaling varies as the cube of the ratio of the

centrifuge g levels to lg. In general, for most

static testing, this is in the range 50 to 100 so

that if, for example, a scale of 100 is chosen,

then a 1 gram explosive charge on the centrifuge

at lOOg represents a 10^ gram explosive charge

or 1 ton of explosive in the prototype. This is a

realistic amount for real tests. In practice, however, it would be of considerable advantage to have a centrifuge running at a much higher rota- tion rate generating, perhaps, l,000g at the

specimen, in which case the magnification of

the explosive charge in the centrifuge in terms

of energy is a factor of 109 so that a few grams

of high explosive in the centrifuge can represent an explosion in the prototype at the low kiloton

level.

In cratering tests there are some difficulties such

as, the possible distortion of the trajectories of

the particles emerging from the crater by

Coriolis forces compared to those of the prototype. In addition, the limited size of the test bucket means that not all of the material thrown

out by the explosion is collected in the centri- fuge bucket. Therefore, the pattern of fallback

of material that typically occurs around a crater

to form some of its rim will not be correctly rep- resented in the centrifuge. It must be borne in mind for these, and other tests where particles are generated that move freely through the space

adjacent to the centrifuge bucket, that the centri- fuge does not represent an ng planet, but is a

highly specialized dynamic environment for test purposes. Only when a particle is connected to the centrifuge container is it subjected to ng

radially. Released into the fluid above or in the

container, which follows its prescribed circular

path, the particle experiences the one g vertical gravity field of earth as well as, of course, fluid

movements (Tan and Scott 1985).

There has been a great increase in the number of

structures that have been subjected to dynamic

loads and tested in the centrifuge, from early

tests on simple structures like single piles, or model footings, to complicated evaluations of

earth dams with internal structure and layering,

retaining walls, a variety of foundations, embedded structures, and some structures on the

surface of the soil. These structures have been

subjected to shocks and earthquake-like vibrations in order to generate solutions for comparison with numerical codes, and also to compare

with full-size prototype tests on similar geometries of test specimen. Full understanding of

model behavior requires comprehensive pro- grams of instrumentation.

2.3 Instrumentation and Data Acquisition

In the past few years there has been a wider use

of, and more ambitious programs in, instrumentation for centrifuge testing. It is not uncommon

now to have 20 to 30 channels of information

returned from a specimen in-flight. The equipment involved includes pressure transducers,

accelerometers, strain gauges, and displacement

transducers. The signals from these pass

229
through data acquisition equipment in which

various signal conditioners are employed to amplify, filter, or otherwise treat the signal

before it is transmitted out of the centrifuge

through slip rings. The nature of the equipment

has not changed particularly during this period.

It is common practice to mount the signal conditioning equipment and, as described above,

portions of a computer on the centrifuge arm in order to process the signal and, in some cases, perform the analog to digital conversion before

passing it out of the centrifuge to the storage

medium, usually in another computer outside

the centrifuge. The signal may be stored in a memory device on the arm for downloading

later. Some increase in the rate at which signals can be sampled and acquired has taken place,

and it is now routine to sample signals from

dynamic tests at 10 to several hundred kilohertz

sampling rates, with the occasional use of a

single channel sampled at 1 megahertz sampling

rate for certain specialized tests, such as wave

propagation in structures.

2.4 Test Containers

There has been an increasing use of specialized

containers particularly for earthquake tests. Formerly, tests were generally done in rigid

containers with stiff walls constructed for static testing purposes. Here, the primary consideration was that the walls would not flex and cause

lateral strains, or stress reductions in the test specimen, that were not accountable for by the

analytical developments. These were replaced

in the early 1980's by test containers deliberately built to be flexible so that the motion of

the soil in the container was dictated by the soil

itself rather than by the container. When a rigid

container is used and the box is excited by, for

example, an earthquake, then there are unrealistic (in terms of prototype scale) vibrations that pass back and forward between the ends of the

box. These reflections interfere with the

"natural" seismically-generated motion of a test specimen that might be embedded in or mounted

on the soil surface. On occasion an essentially

rigid box is used, lined with an

energy-absorbing material called "duxseal"

applied at the ends perpendicular to the axis of

one-dimensional excitation. The "duxseal" is intended to damp out the reflected motions.

Although the study of such tests is still of value

and they can be simulated with numerical codes

involving the same boundary conditions, these

reflection effects are undesirable for tests in which simulation of a specific prototype is required. In consequence, many tests in recent years involving earthquake-like excitation of a container on the centrifuge have employed a device called, variously, a stacked ring or lami- nar box, (Hushmand et al, 1988) which is designed to permit the soil to deform with

minimal interference by the box in one-dimensional shear tests. The device consists of a number of rings of metal, usually in rectangular

form and separated from one another by either

teflon or by roller bearings to reduce the lateral resistance of the rings to movement. The idea is that an ideal box would be essentially massless

and frictionless; the response of the soil to input

shaking at its base would be controlled by the

soil rather than the box properties. Laminar

boxes eliminate many of the end reflections and

permit a container full of soil to deflect in the

shear beam mode that is expected in a layer of

soil of infinite lateral extent during excitation by

an earthquake.

The principal criterion with respect to the rings

of a laminar box is their mass compared to the

overall mass of the soil specimen. If the mass of

the confining rings can be made essentially

negligible (less than 15%, say) of the total mass

of the specimen then the contribution of the

rings to the dynamics of the soil motion is very

small. Because of the increased ring thickness

required to preserve box stiffness these systems

are less efficient the larger the container size

becomes and therefore, the system works better for smaller centrifuges. In addition, such lami- nar boxes permit the lateral motion of the soil specimen at all depths to be measured by the use

of displacement transducers mounted outside the

container and bearing on the ends of the con- tainment rings. When the displacement data are assembled, snapshots can be obtained of the

vertical profile of soil at any time during the

excitation. The disadvantage of laminar boxes

230
is that the specimen preparation is rendered more complicated, since the box requires a

plastic or rubber liner, and special preparations

have to be made in order to apply a vacuum on

the box when tests are required on saturated

sands.

Since, in all centrifuge shake tables, the input

force is applied to the base of the box, well

below the center of mass of the box and soil specimen, some test boxes exhibit rocking or

pitching motion on being shaken. This generates accelerations or stresses in the vertical

direction 180° out-of-phase at each end of the

box.

After this brief review of the state-of-the-art in dynamic testing to the present, a discussion

follows of improvements or advancements that are needed in order to develop more sophisticated tests and to enhance our knowledge of the

behavior of soil under conditions more complex

than those which have been attempted in the

centrifuge so far. This will be considered in the next section.

3. NEEDED IMPROVEMENTS

3.1 Equipment

3.1.1 Two and Three-Dimensional Shaking

So far in earthquake studies the shaking that has

been imparted to the soil model has been

primarily one-dimensional, the equivalent of

horizontal ground motion in the prototype in one

direction only. Although some other components of motion are normally represented in the data (Figure 1), they are accidental byproducts of the dynamics of the centrifuge and

shaking system and vary from centrifuge to centrifuge. Depending on the design of the

shaking apparatus, vertical motions (referring to the full-scale prototype) of an intensity that depends on the design are usually generated.

These can be as high as 30 to 50% of the horizontal peak earthquake accelerations. To a

lesser extent, transverse horizontal motions with

peak accelerations of the order of 2 or 3% of the

primary horizontal input accelerations are also

developed. In all cases the basic intention is to eliminate other components of motion to simplify analysis and interpretation of the

results. When two different centrifuges are used

to study the same scaled event, with the same

model and with intentionally similar one-dimensional horizontal earthquake input, it can be the case that differences in the results are obtained

due to different intensities of vertical shaking

accompanying the same (approximately) horizontal motion (Figure 2). No studies have yet

been made of the effect of the additional accidental vertical motion on the phenomena being

examined. This would require an apparatus in which the vertical motion was capable of being

controlled and could be added to the horizontal

motion in any intensity desired.

For this reason and for other investigations, it would be very desirable to place on a centrifuge a shake table with at least two- and preferably

three-dimensional capability of shaking the

specimen to simulate the major components of

measured input earthquake motion. It appears

feasible to produce two horizontal components

but it may be difficult to obtain controllable

vertical earthquake motion by design.

3.1.2 Other Input Requirements

An examination of the spectra such as are shown in Figure 3 reveals that the typical centri- fuge motion is limited in the frequencies that can be attained by means of the electrohydraulic

devices in current use. Although there are some

differences in the spectra from different

machines, in general, the frequencies are too low to properly represent the rather broad fre- quency spectra of real earthquakes. There is therefore, a necessity to improve the characteristic motion represented in centrifuges in order

to make the spectra more realistic and more rep- resentative of the real earthquake motion. Very

little is known at present on the effect of differ- ent frequencies on the performance of soils in earthquakes. It seems likely that, in particular,

pore pressure increases might be substantially

affected by the frequencies developed during the

231
earthquake, but it is not known whether a given

intensity of high frequency will be more or less likely to liquefy soil than the same intensity at a lower frequency. In some centrifuges (see

Figure 3) there are very distinct gaps between

the peaks on the spectra; these appear to be

caused by various harmonics of the system and

the apparatus used to generate the earthquake

motion. Because of other design constraints it may be difficult to eliminate equipment frequencies in the design range. There are differences

in the performance of large and small centri- fuges in this respect.

In addition, it is desirable that higher accelerations be achieved. The Northridge, California

earthquake of January 17, 1994 developed at one site (Tarzana) sustained peak ground horizontal accelerations of about 1.8g, and it must

be possible in centrifuge tests to achieve this level of acceleration so that the behavior of

structures in such areas can be adequately

followed. It may be noted that, with time, the

increasing density of strong motion accelerographs produces greater peak accelerations as earthquake motions are recorded at a greater

number of sites. The Northridge earthquake was the most recent to demonstrate this.

Increased accelerations in shake tables appears

to be a fairly difficult goal to achieve. Peak

accelerations, at the present time, of about 35g

are attainable on a few centrifuge shaking apparatuses. If the tests are carried out at 50g this represents about 0.7g peak horizontal acceleration in the prototype; if at lOOg then, of

course, only 0.35g prototype acceleration is obtained. The peak acceleration generated is a

result of the combination of the mass of the

shaking table, the size of the servovalve, the

accumulator volume and pressure, and the

design of the piping and tubing which admits

hydraulic fluid to the servovalve. To some

extent the peak acceleration achieved is a chance event which depends on the natural fre- quencies of the equipment as described above,

but it should be possible to design systems

which would achieve higher acceleration fre- quencies than those commonly experienced.

3.1.3 Transducers

A variety of instrumentation is employed in the

soil in centrifuge testing, including strain gauges, pore pressure transducers, accelerometers and pressure transducers in general.

External to the soil mass, other transducers such

as displacement transducers (LVDTs are normally used) can be applied to measure settle- ment or displacements of the soil mass or

container. The sensing components that are normally imbedded in the soil have dimensions

of the order of 3 to 7 millimeters in diameter

and length depending on the manufacturer and

the type of instrumentation involved. When

employed in centrifuge tests at 50 to lOOg range

these then represent prototype dimensions of

transducers of approximately 0.15 to 0.7 meters

in size.

In any normal prototype application, such

instruments would be considered to be large;

their presence would interfere with the response

being measured. It must therefore be assumed

that there is a similar uncertainty in the interpretation of the output of these transducers in the

centrifuge test where, in addition to their absolute size, they are also relatively large compared

to the soil specimen dimensions. It is past time

to develop equipment that is substantially

smaller than these dimensions for centrifuge

testing; advantage should be taken of the great

advances in fabrication of micromechanical

devices developed for other purposes (Barth et

al, 1988; Schlichting et al, 1993). Smaller

devices, of the order of a sand grain in size,

would give more reliable readings of the vari- ables in soil, and more of them could be placed

in or around model structures to determine the

stress fields, pore pressures, etc. on a wider

basis. Local inhomogeneities in the stress field could be detected by such small devices. Lower

practical limits to the size are imposed by the

connecting wires, and the packaging of the

device.

Not much attention is paid at present to the mass

density of transducers; this is an important

feature in dynamic tests, where a mismatch in

232
soil-transducer density will result in spurious

signals. The transducer and its attached cable can be viewed as a small structure with mass,

stiffness and damping (cable) which are differ- ent from those properties in the soil which it is replacing. Consequently, when vibrations pass

through the soil, they interact with the trans- ducer and cable, to cause motions in that system

which are not the same as those in the causative

vibrations.

One difficulty with all such equipment is, of

course, the necessity of attaching it to signal

conditioning equipment by means of wire,

which, with a larger number of sensors, raises

questions about the effect of wires on the

response of the material. An accompanying

issue that needs to be addressed is: If more sensors are to be introduced into the soil then more

data channels need to be available in order to process the signals and make them available for

subsequent analysis. If 50 to 100 transducers

could be installed in a soil profile a corresponding increase in the amount of the signal conditioning equipment would be required. At the

moment about 20 to 30 channels are typically

used for centrifuge testing at the upper end of

the range (frequently, fewer are used), and this already strains most signal conditioning equipment, recording devices, and the slip ring

capabilities.

The development of signal conditioning equipment has been leading towards the installation

of computers on the centrifuge arm where the

data is not only processed, but stored during the

centrifuge test without passing through slip

rings to outside data storage facilities. If a substantially greater number of channels is envisioned, then this form of data acquisition might

be necessary. There are advantages, particularly

in noise level suppression, to storing the data on

the centrifuge arm, but the disadvantage is that the performance of the test would not be known

(without using slip rings) until the centrifuge

had been brought down to lg, and the signals

downloaded to a laboratory computer for

examination. In the present arrangement of

passing signals through the slip rings there is the

great advantage in knowing what is happening

while the test is in progress. There is then a

capability of repeating the test if, in fact, something seems to be wrong that is curable. An

expansion in the number of transducers installed

also complicates the preparation of the soil model.

3.1.4 Data Acquisition

This subject has already been taken up in the

previous section, but there are other considerations with respect to data acquisition which need

to be addressed in the next generation of this equipment. One of these is the transmission of

the signals within and without the centrifuge. It

is desirable to avoid the wiring harnesses, which

are now such a common and obvious feature of

centrifuges, by introducing alternative

photo-optical or infrared transducer devices to transmit the signal. This could lead to a greater

compactness of the equipment, and to isolation

of the instrumentation in the soil from the signal

acquisition equipment mounted on the centri- fuge arm, with the consequent elimination of

noise which is always picked up by the current wiring arrangements. The further possibility of

using radio transmission from transmitters

closely associated with the instrumentation to receiving stations outside the centrifuge also

needs to be examined. Some preliminary

investigations were made on this by the author

15 years ago, but, at that time, the signal

transmitters were very expensive and the apparatus was limited in the number of channels that could be handled. With the micro-miniaturization of equipment, it is now possible to include the radio in the package with the trans- ducer. This does not appear to have been done

in centrifuge testing as yet.

It would be desirable to try to match instrumentation development, which needs to be done by

manufacturers, with such requirements as these. This raises the difficult question that, although

the centrifuge business is proliferating, there are

still, at a commercial level, rather few

centrifuges in the world, and they do not represent a very large market for a manufacturer.

How is it possible to get electronic manufacturing companies to design and build

233
micro-miniature transducers for a market which

might only purchase several hundred of these

devices a year? The same question applies to the development of new signal transfer devices

and, to a lesser extent, to data acquisition. The

latter is used in more fields of experiment, and

is therefore common to a variety of equipment,

including centrifuges. An examination of the

advertisements in electronics and physics

magazines, for example, will show a wide range

of data acquisition equipment made available

for general purpose use.

In this subject area there are certain other needs

which require to be met. At present the state-of

the-art of data acquisition seems to be limited

(at a reasonable price) to approximately 1 million sample acquisitions per second (1 megahertz sampling rate), and this can be per- formed on only one channel. If two channels

are used, half the sampling rate is obtained, etc. Such a rate is applicable for tests such as wave

transmission in parts of structures embedded in

soil such as, for example, in a pile struck by a

pile driver. The pile, if it is made of steel (both

model and prototype), has a wave speed through

the material of approximately 5,000 meters per

second; however, in the model the model pile is perhaps only 30 centimeters long, and therefore,

the wave makes one trip down the pile and back

up again in only 120 microseconds. The total duration of the signal from a strain gauge

attached to the model pile surface is only a few

milliseconds. The short duration of reverberation at a point on the pile requires a very high

sampling rate. However, since in such an experiment there might be 6 to 10 channels of

information from strain gauges or accelerometers mounted on the pile as well as from a number of pressure transducers in the adjacent

soil, a megahertz sampling rate would need to be available for perhaps 15 or 20 channels

instead of 1 channel. At present, this does not seem to be available at a reasonable price. The

second aspect of such data acquisition is, of

course, the volume of data that will be received

from 20 transducers, say, all sending in data at a

1 megahertz or greater rate. This is a very substantial amount of information, coming in at a very rapid rate so that storage needs to be available in the order of 20 or more megabytes per

second.

One other problem requires attention. In the

past, at Caltech, we have made use of a high

speed movie camera, mounted on the centrifuge arm, to record earthquake-induced displacements of the test structure (say a model dam) at speeds of up to 6,000 frames per second. Some

of these models were tested at lOOg, and, if a conventional movie camera filming the prototype has a framing rate of 20 frames per second,

the movie of the model on the centrifuge, to be

shown at a normal viewing rate, should be taken

at 2,000 frames per second. Since "slow

motion" permits a better analysis of the data, we

chose to use 6,000 frames per second. Special

high speed film is required. Normally, at Caltech we would take the 16mm film to a company in Hollywood for same day processing (a

"rush"), view the movie, and decide on the next

test to be done. The last time we attempted this the film company said "nobody here processes

16mm film anymore; you have to send it to San

Francisco". We did, and received it back 10

days later, which had an impact on our testing

operations. Clearly, soon, if not already, no

commercial concern at all will process 16mm

film, and we will have to do it ourselves.

Processing color movie film is not a simple task.

The 16mm film is vanishing because of video cameras and videotape. But their framing rate is only about 30 frames per second, and an

increase to 6,000 is not in the foreseeable future. Even were the video camera available, with one

frame generating say, 700,000 bytes of data,

then at 6,000 frames per second, a storage rate (tape or other) of about 4,200 megabytes per

second would be required. Since a model earth- quake might last 0.5 seconds, a storage capacity

of about 2,000 megabytes is needed. For some

tests, such as those involving explosion or impact, much higher framing rates (to 100,000

frames per second) would be desirable. It is possible, with special equipment, to capture a few frames (Gill and Kuennen 1991) with video

techniques, but the technology is at an early

stage. There seems to be an impossible gap

here. What is to be done?

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3.2 Physical Problems

3.2.1 Scaling Relations

Geotechnical centrifuge tests were devised so

that properly derived scale relations could be

applied to scale models involving soil. In a

substantial number of static and dynamic situ- ations these, by now familiar, relations are appropriate and have been used. There are, however, conditions for which conflicting scale

relations apply, and where the proper construction of models, and interpretation of their test behavior is not clear. Perhaps the best-known of

these circumstances is the conflicting time

scales encountered when the generation and

dissipation of pore pressures occur as a result of

dynamic excitation. Time for dynamic effects in the model is n times the prototype time, where n

is the ratio of prototype to model lengths,

whereas in the diffusion process the model time

is rP- times full-scale time. Thus the rate of pore

pressure dissipation in, for example, a seismically excited saturated sand may exceed the rate of generation.

It is possible, if the prototype sand is used in a

centrifuge test, that little or no pore pressure

increase will be observed during excitation,

because of the rapidity of diffusion. The expedients adopted to resolve this dilemma so far have drawbacks. One technique is to use finer

soil in the model than prototype, but then the

mechanical properties of the material, on which

the scaling assumption depends, are no longer

the same. The other is to employ a liquid more

viscous by the factor n, in the model than the

water usually found as the pore fluid in the

prototype. Silicone oil, or glycerin/water mixtures are commonly used. In this case a question arises as to the effect of a more viscous

fluid on the mechanical behavior of the soil,

specifically, in the seismic case, on those properties which affect pore pressure generation.

A reductio ad absurdum argument would indi- cate that in the limiting case of an extremely

viscous pore liquid, say asphalt, no grain separation, and thus no pore pressure generation

would occur on seismic input to the soil. This

would indicate that viscosity has a significant

effect, but does a higher viscosity by a factor in the typical centrifuge range of 50 to 100 substantially modify the pore pressure generation,

or not? This needs to be investigated, probably

for sands of different grain sizes.

The other discrepancy in scaling laws occurs if viscous flow is important in the model during

the time of the experiment - that is to say in a

quasi-static test. If the soil (model and prototype the same) can be represented as a linearly

viscous material, then, since strain is dimensionless, and the shearing stress is the same at homologous points in model and prototype, it follows that the time scale is unity (model and

prototype times are the same). A certain amount

of strain would take the same time to develop in model and prototype.

Slope creep would be a condition where this consideration holds. Another is volumetric

change in a soil over a period of time, where a component of the change during and after primary (diffusion) consolidation, referred to as secondary consolidation, continues as a function

of time. A way around the difficulty of modeling this phenomenon has been suggested

(Sathialingam and Kutter, 1994).

There are still a substantial number of problems

involved if very large structures require modeling on the centrifuge since, in general, the

scaling ratio cannot be made high enough with

existing machines to bring the model dimensions down to a size suitable for the model test chamber. Under these circumstances, on occasion, individual components rather than the

whole of the structure can be tested. It is common to use centrifuge testing as a verification or validation of numerical models, and

many of the above considerations become rela- tively less important if the point of the experiment is simply to test the correct physical phe- nomena at the correct stress levels in the centri- fuge for code verification purposes. If a given

model can be simulated correctly by a numerical

program, then confidence is given that a

235
numerical calculation with, say, different

geometry or material properties, can handle the

real prototype question at issue.

Scaling relations also need to be examined in problems involving liquid density-driventhermal convection in the centrifuge, although

dynamic conditions related to this mechanism

have not yet appeared. The relevant dimensionless parameters are the Rayleigh number, where

the gravitational acceleration appears, the

Prandtl number, which expresses the importance

of mechanical to thermal diffusion, and the

Taylor number, which brings in the rotation of

the centrifuge. For liquid flowing through soil,

the kinematic viscosity of the liquid should

probably be replaced by the consolidation coefficient of the soil.

3.2.2 System Dynamics

Much of the work that has been done so far on

the development of rapid motion-generating

systems on centrifuges has been done on a purely ad hoc basis. The shaking tables, ser- vovalves, and accumulators which accompany

them have been assembled more or less instinctively, based on past experience. It is time for

analysis of the performance of such systems to be based on proper numerical models of the

entire system structure. It is necessary to form a model (Figure 4) showing the system including

centrifuge arm, drive system, shaft, boundary

conditions, and so on, and then to insert in this model the desired shaking system performance

at the appropriate intermediate position on the

swinging bucket, to examine the effect on the

system. Only by such studies can an understanding be obtained of the source of the natural

frequencies that are observed in the shake table response, and of what can be done in order to make the motion more representative of real-life or other desired input motions. When even the

simplified Figure 4 is viewed it can be seen how

many inputs come from the mechanical structure of the centrifuge. The results of such numerical

studies combined with modal data from tests with shaking machines mounted on the centri- fuge will give the necessary system or transfer

functions.

One component in particular in this system

needs to be described rather precisely and that is the actual hydraulic shaking system. It is sug- gested that a model of the hydraulic system also

needs to be constructed, including the accumulators, piping system, the compliance of the

various components, the compressibility of the

hydraulic fluid, and the system function of the

servovalve that is employed. With a suitable

and realistic numerical model of this kind it would be possible, as a separate issue, to examine the effect of increasing the hydraulic flow or changing the design or capacity of the

servovalve.

At the same time, the design of the shaking sys- tem itself needs to be evaluated more closely

than has been the case to date. The motion of

the shake table is intimately connected with the

mechanics of its design. In particular, for

example, an important consideration is the location of the actuating pistons with respect to the center of mass of the soil container being

shaken. Such a study might make it possible to construct a better shaking table to eliminate some of the vertical and transverse motions

which are observed in most such machines

(Figures 1 and 2). An undesirable component

observable in some centrifuges is the presence

of pitching motions of the centrifuge test box

caused by the location of the actuating pistons

far below the center of mass, as discussed

earlier. This shaking motion (see Figure 5)

gives rise to vertical accelerations at the ends of

the soil container which are as large as the horizontal values but out-of-phase with respect to one another.

The analysis of the entire centrifuge system and

the hydraulic actuators involved should, of

course, be carried out under in-flight conditions

with payload so that the stresses in the arm are

realistic.

3.2.3 Model Construction in Flight

A question which persists in centrifuge model

testing and analysis refers to the difference in the stress field in models which are constructed

at lg in the test laboratory before being brought

236
to 50 or lOOg in the centrifuge, and the stress

field in the prototype, where everything is constructed or formed at lg. Ideally all models in the centrifuge should be constructed in flight,

including the installation of the appropriate

structure. To explain in more detail: the empty

container should be brought up to speed in the

centrifuge, and the soil placed in it by raining or by centrifugation of the material in water so that consolidation and/or stress-induced settlement

takes place as the soil profile is formed under

the artificial gravity field of the centrifuge. This

ensures that the region to be tested is under the

correct stress conditions (depending on the

boundary conditions of the specimen). The next

stage is to insert, for example, a pile or other

required structure into the soil to generate the

correct stress conditions appropriate to the

prototype. Although it would be easier to insert the pile by pushing it into the soil in flight, this would not represent the conditions which would

be generated by a driven prototype pile and thus,

model piles ought to be driven by a model

hammer. During this operation in flight the

stress field caused by the pile or other foundation insertion would be duplicated in the

pressure transducers, strain gauges, accelerometers, and so on, installed in the soil.

For this example (pile) the installation of the

system, although complicated, is perhaps more

simple than many other realistic conditions in practice. For example, there have been studies

(Kimura 1984) of the construction of sand

embankments on top of clay layers in-flight on

the centrifuge by raining or pouring the sand

from a hopper. When the embankment reaches

a certain height failure may occur, and the

observation of a failure mechanism can be

made. However, because of Coriolis effects it is very difficult to construct an embankment with

proper shape. The sand passing through the air from the hopper to the soil surface follows a curved path, as shown in the paper cited. Tests

including buried structures subjected to explosions and/or dynamic shaking are even more

difficult to consider. A particular difficulty of

course involves installing the necessary struc- ture with attached instrumentation.

It is suggested that it might be necessary to develop much more complicated systems than

are presently visualized for the construction of

such detailed and correctly-dimensioned models. The process might be facilitated by an

excursion into the problem of designing and

constructing suitable robots or robotic arms which could deposit soil or pick up and place

transducers in a soil while the centrifuge is in

flight. Such transducers or other probes could

be moved at various times during the experiment

in order to examine different aspects of the

material behavior. Alternatively, the robotic arms could introduce testing devices into the

soil to measure soil properties during flight. A

conceptual sketch of such an experiment with

suitable robotic arms to extract piles from a

quiver and install them in a group is shown in Figure 6. One difficulty in design is that the

robot is also required to function at 50g, lOOg,

or higher.

It is necessary at this stage to make a plea for more full-scale field tests of a variety of problems and boundary conditions with which centrifuge tests can be compared, so that the

necessity for conducting more realistic sample

preparations can be documented, and those

which are necessary, as compared with those

which cause only minor differences in the

behavior of a sample can be identified.

An additional consideration, which follows the same train of reasoning, is that there is a need

for more in-flight studies of the material behavior and material properties. In general, centri- fuge test properties are determined largely in the

lg condition of the material. In the case of

sands, the material properties are identified from

the known stress field and the initial in-place

density of the sand. There have been a few tests in which material properties have been deter- mined from in-flight tests by means of miniature cone penetration tests or miniature vane tests (Corte et al, 1991; Stewart and Randolph 1991);

these tools have been inserted into the soil in

flight. Seismic analysis techniques are useful in

this regard. For example, it is possible to place

seismic signal generating elements in the walls

237
of a container, and receiver elements on other

walls. Signals are passed through the soil at dif- ferent depths in order to determine P and S wave

velocities in the material and thus, the profile

and properties at least in the elastic range of the

soil (Gohl and Finn 1987; Shibata et al, 1991).

Care must be taken that the signal does not take a different route through the container walls. This can be done before and after performing a

test. It is possible, for example, to perform such

measurements while a test is in progress, and, by means of mathematical transform and deconvolution techniques, to examine the change in

soil properties graphically as the soil is stressed

by static or dynamic loads. Soil profiles have

been examined by means of the generation and

measurement of Rayleigh waves which have

also been adapted for centrifuge experiments.

The advantage of the centrifuge over the prototype material is that in all centrifuge tests the

soil is placed and therefore the experiment

carries with it the possibility of including

experimental micro-apparatus in the soil profile as it is being constructed. These pieces of

equipment could then be actuated subsequently

during the actual test.

3.2.4 Test Containers

In all centrifuge tests so far, great use has been

made of rigid-walled containers, some with a

transparent single wall so that the experiment

cross section can be visualized in flight by

movie or television camera. For dynamic tests the stacked ring or laminar box equipment has

been developed; this has been of value in one- dimensional shaking tests but presents difficulties, of course, in its present form, for multidimensional tests. For many dynamic tests it is desirable to design and construct different types

of boxes which will allow for two- and three- dimensional motion of the specimen. The

majority of centrifuge tests performed to date

has consisted of plane strain tests, in which the

influence of the side-wall boundaries is always

open to question, and a variety of methods of

treating these boundaries has been suggested.

More work needs to be done on the possibility

of substituting various materials, which modify

the reaction of the box on the soil or permit

wave reflections to be dampened or modified

sufficiently so that they are no longer a consideration in the behavior of the model. These new

boxes should have built-in drainage conditions

at the boundary to permit soils to consolidate more rapidly or to admit the introduction of

water into a soil sample to saturate it. The

application of a vacuum on the soil specimen in order to facilitate deairing the specimen and to ensure appropriate performance of pore pressure

transducers embedded in it is also necessary. A

test container manufactured so that pressure and

other transducers can be more readily installed

in the soil with their leads and terminals plugged

into the side of the test box is also useful in the

soil testing process.

4. CONCLUSIONS

Up to this point it has been relatively easy to develop new dynamic centrifuge tests, models,

inputs, and instrumentation. It will get progressively more difficult as the rigor of experimental

proof becomes more demanding. Not enough

work has yet been done to demonstrate

conclusively that even the most straightforward

of static centrifuge model tests exhibit correctlyscaled prototype behavior. It may require the

effort and expense of constructing a prototype

test to represent, complete with respect to soil,

boundary conditions and input, a centrifuge

model. In dynamic tests, the problem is more

difficult, when different scaling factors,

dynamic material properties, and material substitutions are employed. It is not clear that numerical simulations assist in the elucidation

of the mechanisms at work.

There are further difficulties, alluded to earlier, in obtaining quantitatively similar data from the same dynamic centrifuge test carried out on

different centrifuges at different institutions, even when the same sample preparation procedure is used for all tests. Small differences in

soil density persist, the input motion is very

difficult to reproduce, especially if it is intended

to represent the motion of an earthquake, and

there may be differences caused by the centrifuge radius, direction of excitation

(circumferentially or axially), and the vibration

238
field of the rotating centrifuge arm. The conse- quences of these variables need to be explored

in more detail, and their effects minimized, in order to produce uniform test results to provide

a basis for extrapolations to prototype scale and

for verification of numerical predictions.

Undoubtedly, because of the attractiveness of

the method, the centrifuge method will continue

to be used, but a pessimistic view indicates,

from past experience, that the fundamental

question of the validity of the technique will

continue to go unexamined to the depth it deserves.

5. REFERENCES

1. Aboim, C. A., R. F. Scott, and W. Roth,

"Centrifuge Earth Dam Studies: Earthquake

Tests and Analyses," Report to the U.S.

National Science Foundation, Grant CEE7926691, Dames & Moore, Los Angeles,

California, 1983.

2. Arulanandan, K., and R. F. Scott, "Project

VELACS-Control Test Results," paper

accepted for publication in Journal of

Geotechnical Engineering, ASCE, 1993.

3. Arulanandan, K., R. Dobry, A. Elgamal, H.

Y. Ko, B. L. Kutter, J. Prevost, A. N.

Schofield, R. F. Scott, "Interlaboratory

Studies to Evaluate the Repeatability of

Dynamic Centrifuge Model Tests,"

Dynamic Geotechnical Testing II , R. J. Ebelhar, V. P. Drnevich, and B. L. Kutter,

Eds., ASTM STP 1213, January 1994.

4. Barth, P. W., F. Pourahmadi, R. Mayer, J. Poydock and K. Petersen, "A Monolithic

Silicon Accelerometer with Integral Air

Damping and Overrange Protection,"

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Actuator" Workshop. 35-38, Hilton Head

Island, S.C., June 1988.

5. Corte, J.-F., J. Gamier, L. M. Cottineau, and

G. Rault, "Determination of Model Soil

Properties in the Centrifuge, Centrifuge 91 , Balkema Rotterdam, 607-614, 1991.

6. Fuji, N., "Development of an

Electromagnetic Centrifuge Earthquake

Simulator," in Ko, H. Y., and F. G. McLean,

eds., Centrifuge 91 , Balkema, Rotterdam

351-354, 1991.

7. Gill, J. J., and S.T. Kuennen, "Half-Space

Modeling of Explosively-Formed Craters,"

in Centrifuge 91 , Balkema, Rotterdam, 465-

472, 1991.

8. Gohl, W. B., and W. D. L. Finn,

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9. Hushmand, B., R. F. Scott, and C. B. Crouse, "Centrifuge Liquefaction Tests in a Laminar Box," Geotechnique. 48 (2) 253-

262, 1988.

10. Kimura, T., "Geotechnical Centrifuge

Model Tests at the Tokyo Institute of

Technology," Proceedings of International

Symposium on Geotechnical Centrifuge

Model Testing. 59-79, Tokyo, Japan, 1984.

11. Sathialingam, N. and B. L. Kutter, "Scaling

Laws for Rate Dependent Shear and

Consolidation of Clay," Dynamic

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STP 1213, January 1994.

12. Schlichting, V., G. Pollak-Diener, E. Obermeier, D. Hammer-Schmidt, F. V.

Schatz, and B. J. Hosticka, "Digital

Programable Pressure Sensor with On-Clip

CMOS Signal Processing and Data

Storage," Proceedings of the Seventh

International Conference on Solid-State

Sensors and Actuators . Transducers '93 , Digest of Technical Papers, 988-991,

Yokohama, Japan, June 1993.

13. Schmidt, R. M., and K. A. Holsapple,

"Theory and Experiments on Centrifuge

Cratering," Journal of Geophysical

Research. 85 (Bl), 235-252, 1980.

239
14. Scott, R. F., "Centrifuge and Modeling

Technology: A Survey," Revue Francaise de

Geotechnique. 48, 15-34, 1989 (in French).

15. Shibata, T., K. Kita, S. Kobayashi, and A.

Yashimia, "Performance of Shaking Table

Tests and Measurement of Shear Wave

Velocities in a Centrifuge," in Centrifuge

21, Balkema, Rotterdam, 391-398, 1991.

16. Stewart, D. P., and M. F. Randolph, "A

Newsite Investigation Tool for the

Centrifuge," in Centrifuge 91 . Balkema,

Rotterdam, 531-538, 1991.

17. Stokoe, K. H., and S. Nazarian, "In Situ

Shear Wave Velocity from Spectral

Analysis of Surface Waves," Proceedings of

the Eighth World Conference on Earthquake

Engineering. 3, 31-38, 1984.

18. Tan, T.-S., and R. F. Scott, "Centrifuge

Scaling Considerations for Fluid Particle

Systems," Geotechnique 35 (4) 461-470,

1985.

240
Time (sec)

Long. Accel Vert.Accel Transv. Accel

FIG. 1 Accelerations in longitudinal (principal horizontal), vertical and

transverse directions on a centrifuge shake table.

241
0.15

-0.2-

-Q.35"t— 1 1 1 1 1 1

1 0 2 4 6 8 10 12 14 16

Time, sees

Centrifuge A Centrifuge B

FIG. 2 Vertical acceleration in difference centrifuges for (approximately) same

horizontal earthquake motion.

242
Frequency, Hz

0.2g Peak C/f 0.024g Peak C/f N-S El Centra

FIG. 3 Fourier Spectra of horizontal input motions on a centrifuge compared to NS El Centro, 1940.

243
COUNTERWEIGHT

-:->:-:-foundation •»>*

PHYSICAL CONFIGURATION

DYNAMIC REPRESENTATION

FIG. 4 Centrifuge and representation for dynamic numerical simulation.

244
1 1.5 2 2.5 3 3.5 4

Time, sees

Right Left Center

FIG. 5 Pitching motion of centrifuge shaking table, as indicated by out-of-phase

vertical accelerations at ends.

245
FIG. 6 Possible use of robotic arms to drive piles in flight and assemble with pile

cap for centrifuge test.

246
Shaking Table Failure Tests of Actual Size RC Structures

by

Chikahiro Minowams", Keiichi Ohtani 2

', Nobuyuki Ogawa31

, Tadashi Mikoshiba4

', Toshihiro Hayashida5

', Tsuneo Okada6'

ABSTRACT

We conducted the failure tests for actual

size reinforced concrete frame structures (1-

story and 3- stories models) by using the

large-scale shaking table of the National

Research Institute for Earth Science and

Disaster Prevention (NIED). 2 specimens of

1- story model were tested for making clear

the effects of hoop reinforcement of column.

The specimen, which has a dense hoop

reinforcement, was shown a ductile behavior,

but the other specimen, which has a rough

hoop reinforcement, has no ductile behavior.

The damping coefficient was increased depend

upon the plastic properties. The actual size 3-

stories RC moment frame specimen was

designed to week beam and strong column

properties. The test result was shown the

failure mechanism as same as predicted.

Keywords: Reinforced concrete structure,

Dynamic failure mechanism, Shaking table

test, Effects of hoop reinforcement.

INTRODUCTION

We supposed that the buildings, which were constructed in the recent year, have a good ductile properties and enough aseismic

performance, but the buildings, which were

constructed until more than ten years ago, did not always have a sufficient aseismicity. The

study of the failure mechanism for the old

buildings is very important to establish the

earthquake countermeasures. And it is to say

nothing that the study of the failure procedures of buildings is also important to develop the advanced earthquake resistant

design criteria.

The earthquake resistant design criteria was constructed mainly by using the study

results of static tests. It was scarce the using

of the dynamic test results for the design

criteria. We pointed out for this reason that

we do not have enough data of shaking table tests and we are difficul to obtain the

remarkable fruits. There are few large-scale

shaking table, which has a capacity of failure test for the actual size specimen, such as RC

frames. The shaking table test has the

demerits as follows; (1) the difficulty to obtain

the response of specimen, (2) the need of long

preparing time and big cost for the 1 minute

testing duration. But, the shaking table test has the merit that we can clearly reproduced

the actual earthquake response such as a

1) Cooperative Research Officer, NIED

2) Director, Disaster Prevention Research

Division, NIED

3) Head, Earthquake Engineering Laboratory,

NIED

4) Researcher, Earthquake Engineering

Laboratory, NIED

5) Head, Vibration Laboratory, Tokyu Co. Ltd.

6) Professor, Institute of Industrial Science,

University of Tokyo

247
dynamic behavior, whcih is more real testing

method than that of static test such as pseudodynamic testing.

We, National Research Institute for Earth

and Science and Disaster Prevention,

conducted the failure tests of actual size

reinforced concrete frame structures by using

the large-scale shaking table of NIED, for the

specific joint research program between the

government and private sectors under the

sponsorship of the Science and Technology

Agency. This research program is the 4-year

project, which is including the actual size 1- story tests, the actual size 3-stories tests for

the reinforced concrete frame. This report

describes the test results of the actual size 1- story specimens and the actual size 3-stories

specimen.

The research objects for 1-story

specimen tests, which those have different

hoop intervals, were the realization of

mechanism of the column failure during

earthquake motion. And, for the 3-stories

specimen, we want to conduct a beam failure

of the frame by earthquake motion, which was

designed a weak-beam and strong-column

frame.

ACTUAL SIZE 1-STORY SPECIMENS

1) Specimens

We conducted the test for two specimens.

These specimens were constructed 1-story, one bay RC frame with 4 columns. The

section and height of columns are 35 cm x 35

cm and 2 m, respectively. The roof slab was made by H shaped steel and steel block for the

simulation of dead and live weights. The

weight of roof slab is 103 ton, and height of

gravity of roof slab is 453 cm from the surface

of shaking table. The specimens were

arranged to the symmetry for both directions

with the center of table. Photo. 1 shows the

total view of specimen, which was assembled

on the shaking table. The elevation and plan

of specimen arrangements are shown in Figs.

1 and 2. The sections of each specimen are shown in Figs. 3 and 4. At the rearrangement

of specimen, the part of roof weight was

removed. We constructed the steel frames for mitigation of drop shocks at the column failure. The weight of roof was fixed prior day of

testing. Until them, the weight was supported

by the oil jacks, which was set on the steel

frames.

The column of A-type specimen (Fig. 3) was designed the early occurrence of bending

yield, which has the shear capacity larger than

the bending capacity. The column of B-type

specimen (Fig. 4) was designed the early occurrence of shear yield, which has the

bending capacity larger than the shear capacity.

The main reinforcing bars of each column were same arranged to 8-D22. The

reinforcing bars of hoop were used D6 for both

specimens. The intervals of hoop were 10 cm

for A-type and 40 cm for B-type specimens.

The values obtained by the material tests are as follows;

A-type B-type

Design Target Bending failure Shear failure Section 35cm x 35cm ditto

Pw 0.0018 0.0005

Nc 21 kg/cm2 ditto

Fc1 c 358 kg/cm2 357 kg/cm2

cFt 32.2 kg/cm2 29.9 kg/cm2

CE 249 kg/cm2 264 kg/cm2

sFt 3730kg/cm2 ditto

SE 2.1xl06kg/cm2 ditto

Ke 30.1 t/cm 31.9 t/cm

Pb 504 t*cm 476 t*cm

Db 0.117 cm 0.104 cm

p

y 1868 t*cm 1868 t*cm

Dy 2.17 cm 2.11cm

Qy 18.7

1

18.7

1

248
Qu 20.6 1 17.3

1

Qy/Qu 0.91 1.08

QB

0.725 0.670

Where : Ratio of shear reinforcement,

Nc: Axial stress, CFC

: Compressive strength

of concrete, cFt

: Tensile strength of concrete,

CE : Young modules of concrete, sFt

: Yield

strength of steel bar, SE : Young modules of

steel bar, Ke: Rigidity of elastic range, Pb, Db

: Strength and displacement at bending

crack, P

y, Dy

: Strength and displacement at bending yield, Qy

: Shearing force at bending

yield, Qu: Ultimate ahear strength, Qy/Qu

: Ratio between Qy and Qu , QB

: Base shear

coefficient.

2) Testing and Measuring Methods

The one dimensional shaking table of

NIED was used for this test program. The

dimension of shaking table is 14.5 m x 15 m. The maximum velocity and maximum

amplitude of table are 75 cm/sec and ± 22 cm,

respectively.

In this test program, the shaking table

have been dried by the shock wave for measurement of damping properties of

specimen (rectangular wave of displacement,

period : 20 sec, double amplitude : 0.5 cm)

and the strong earthquake motion recorded of

E-W component at Hachinohe harbor during

1968 Tokachi-oki earthquake. The shock wave was generated by the synthesizer. The

strong motion was generated by the personal

computer through the high-pass filter of 0.4

Hz. Fig. 5 shows the time history and velocity

response spectra of shaking table response for the A-type specimen.

We used several types of sensor; the

strain gauge type accelerometer and the

potential type, the inductance type and the

laser type displacement sensors. The strains

of steel bars were measured by the 2 gauge

method. The data from sensors were

recorded to the hard disk of the computers by

the 1,000 Hz sampling through the 16 bits A/D

converter. And we used 8 of Video recorders.

The data collection system ad the video

recording systems were synchronized by the

time controller and wave duplicating system.

3) Test Results

3-1 A-Type Specimen

The specimen was collapsed by the two

times excitations of strong motion. The

reason of two times excitations was that the

specimen was not collapsed by the occurrence

of emergency stop of shaking table during the

first exception. Therefore, we conducted the

second excitation to make a complete collapse

of specimen. The maximum displacement,

velocity and acceleration of table response at the first excitation were 144 mm, 73 cm/sec

and 800 GAL, respectively. The response

wave forms of relative displacement and

acceleration of roof slab, and the hysteresis

loop between acceleration and relative

displacement of roof slab are shown in Fig. 6. We conducted the shock motion excitations to make clear the change of rigidity and damping

characteristics between before and after the

strong motion excitation. The response of

free vibrations, which was measured by the

shock motion, are shown in Fig. 7. The

predominant frequencies and damping

coefficients, determined by the AR model

analysis for this free vibrations, are follows;

Before the test

4.2 Hz, 1.6%, c=4?cmx 0.0672

After the test 3.2 Hz, 2.9% c=47cmx 0.0928

The two columns faced north side within

total 4 columns were collapsed by the second

strong motion excitation. The two columns

faced south side were not collapsed, but

severely cracked. The maximum displace249
ment, velocity and acceleration of table

response at the second exception were 165

mm, 68 cm/sec and 730 GAL, respectively.

The response wave forms of relative

displacement and acceleration of roof slab, and

hysteresis loop between acceleration and

relative displacement of roof slab are shown in Fig. 8. The predominant frequency was

changed from 4.2 Hz to 3.2 Hz during the first exception and to 2 Hz at just before the

collapse by the second excitation. The

hysteresis loop in Figs. 6 and 8 are recorded at C4 column located in north side. Photos. 2 to 5 show the column collapse procedures of the

north side during second excitation. From

these photographs, the shear crack was

occurred at the top of C3 column, the cracks were progressed after that, and the shear

crack was occurred at the bottom of C4 column,

which is the main cause of collapse by this

excitation. The reason, why the south side

columns didn't collapsed, is estimated to the

effect of slips of specimen from shaking table. We installed the displacement sensors between the foundation of column and table

surface for the purpose of watching the occurrence of slip between them. These

records are shown in Fig. 9. We can find the occurrence of slip at the south side columns

during both strong motion excitations. The

north side columns were occurred little amount of slip during the first excitation, but

the bottom of C3 column, which was happened

the first shear crack, didn't slipped during the

second excitation, we supposed.

3-2 B-Type Specimen

This specimen was failed at once by the

strong motion excitation. The maximum

displacement, velocity and acceleration of table

response were 109 mm, 56 cm/sec and 630

GAL, respectively.

The vibration characteristics of specimen

at the initial state are as follows;

4.4 Hz, 1.7% c=47rmx 0.0794

The wave forms of relative displacement and

acceleration of roof slab, and the hysteresis

loop between them are shown in Fig. 10. The

running gain spectrum between roof and table was shown in Fig. 11. From this spectrum, we

can find the predominant frequency was

changed from 4.5 Hz at the initial state to 2.5

Hz at just before the collapse. The hysteresis

loop (Fig. 10) is drawn the features of the C4

column at the north side. Photos. 6 to 9 show

the procedures of column collapse. From

these photographs, the shear cracks at the top

of C3 and C4 columns in the north side and the

bottom of CI column in the south side were

occurred at the nearly simultaneous timing

(within 0.01 second time differences), the

bottom of C2 column in the south side was

cracked at 1.5 second after the former crack

occurring, these cracks were propagated to the upward direction of column, and then the

specimen was totally collapsed. The slip

between the foundation of column and the

table surface didn't occurred by the effects of

installation of L-shaped steels between the

foundation of column and the steel frames for the mitigation of drop shocks.

4) Consideration for Hysteresis Loop

Properties

4-1 A-Type Specimen

The third gradient for both excitations were estimated to the negative slopes as the

peak oriented rule for reinforced concrete

structures. The maximum strength of this

specimen was estimated to 103 ton x 0.9 G =

92.7 ton. The residual displacements were

observed at every columns for west direction

after the first excitation. The amount of

residual displacements of each column were as

follows;

Column DR Column DR

CI 7.4 mm C2 7.6 mm

250
C3 13.1mm C4 5.4 mm

The second elastic limit displacements were

estimated to 1.5 - 2 cm for the first excitation

and 2 - 3 cm for the second excitation. These

phenomena were corrsponded to the rcord of

video tapes sech as the opning of cracks.

4-2 B-Type Specimen

The hysteresis loop between the load and

displacement can not estimated, because the

specimen was destroyed before the drawing of

loop. The maximum strength of this specimen was estimated to 103 ton x 0.8 G = 82.4 ton. From the correspondence of video tape

records, the specimen was collapsed by the occurrence of cracks just after the arrival to maximum strength. The second elastic limit was estimated to 1.8 - 2 cm.

4-3 Comparison of Specimens

The both A-type and B-type specimens were recorded to the two different values of

the second elastic limit. The both specimens

had the third elastic limit and the negative

slope after this limit. If they had the third

elastic limit, the displacement between the

second and third elastic limits was estimated

to about 5 mm.

ACTUAL SIZE 3-STORIES SPECIMEN

1) Specimen

This specimen had a plan of 6 m x 6 m

with 3 stories and one bay. The section of

each column was 45 cm x 45 cm (8-D22, - D10-(&100) for each story. The height of each

story was 3 m. The sections of girder were 50

cm x 30 cm (2-D22+1-D19, 2-D16, -D10-

@200) for the second and third floors and 40

cm x 25 cm (2-D22, 2-D16, -D10-@200) for the roof. The thickness of slabs was 15 cm for

each floor. The concrete foundations were 2.6

m x 2.6 m x 0.9 m. The total height of

specimen was 10.2 m. This specimen was

designed to have a property of weak- beam

and strong-column. The beams were

assembled to the transverse direction of

excitation for each slab, which size was same

as girder of each floor (2-D19, 2-D16, D-D10-

@200). The reinforcement arrangements of

slab were D10-@300 for the excitation

direction and D10-@200 for the transverse

direction. The 18 of one ton steel block were

arranged on the slab for the simulation of live load. Therefore, the total weight of each floor was 30 ton. For the test safety, we were

constructed the steel frame. Fig. 12 shows

the schematic view of the specimen.

2)Testing Method

The shaking table was driven by the

strong motion as same as 1-story test. The

strong motion was generated by the personal

computer through the 0.4 Hz high pass filter from FFT method. Fig. 13 shows the preanalytical results of the displacement response

for each floor during the strong motion. We

conducted this test on March 14 of this year. And now, we are processing the test data and

analyzing the test results.

CONCLUSION

During the first excitation for the 1-story

A-type specimen (hoop interval : -D6-

@100), the emergency stop, just after the first peak response, was occurred depend on the

safety-control logic system of shaking table. At the second excitation for this specimen, the

shear crack was generated at the C3 column

after the second peak response, and the

specimen was failed by the opening of cracks

after the third peak response with 6 cm

displacement. The B-type specimen (hoop

interval : n-D6-@400) was collapsed just after the occurrence of crack at the first peak with 2

cm displacement. From these facts, we can

confirm that the dense reinforcing against

251
shear force is very effective to maintain the

earthquake resistant capacity of building.

We conducted the test for the actual size

3-stories specimen at 2 month ago. The

dynamic and failure behaviors of specimen were shown almost same as the pre-analyzed

results. We hope that we will present a more

detailed report at the next joint meeting.

ACKNOWLEDGMENTS

These test projects were conducted by

the specific research program between the

government and private sectors under the

sponsorship of Science and Technology

Agency. For the success of this project, we

established the technical coordinating

committee (chairman : Prof. Tsuneo Okada),

which the committee members are consisted

from NIED, University of Tokyo, Tokyu Co.,

Obayashi Co. and Aoki Co.. The authors

acknowledge the assistance of the members of

the committee.

Photo. 1. Actual Size 1-stoty Reinfoced Concrete Specimen

252
Phpto. 2. Initial Crack at the Top of C3 column (A-TYpe, North Side)

Photo. 3. Crack at the Bottom of C4 Column (A-Type, North Side)

253
i

Photo. 6. Shear Cracks at C3 nad C4 Columns (B-Type, North Side)

Photo. 7. Cracks at the Bottom of Cl Column (B-Type, South Side)

154
Photo. 4. Crack Opning at C3 and C4 Columns (A-Type, North Side)

Photo. 5. Damage Process of North Side (A-Type)

255
Photo. 8. Separation of the Top of C3 and C4 Columns (B-Type, North Side)

Photo. 9. Damage Process at Soutn Side (B-Type)

256
6.000

Supprot Frame

I

I— J i

103 ton

I _ I

' r

Test Column N

l

Roof

1

' 1

,

1 11 III II I ll

1,250 1,250 3,500 1,250 1,250

Fig. 1. Setup Elevation of 1-Story Test Specimen

6,000

North Side

C3 C4

-o

to

CI C2

South Side

Driving Direction

Fig. 2. Plan of 1-Story Test Specimen on Shaking Table

257
Fig. 3 Column of A-Type Specimen Fig. 4. Column of B-Type Specimen

(Hoop : C\-m-@lQ0) (Hoop : LJ-D6-@400)

MEASURE WAVE MAX.= 728.8 GAL WAVE LENGTH* 40.0 SEC

700.0 VELOCITY RESPONSE SPECTRA

h= 0.000 h= 0.005 h= 0.010

0 ! J I I

I I I '. I 1.0 natural frequency in hz

100

Fig. 5. Velocity Response Spectra of Table Response

(A-Type, Second Excitation)

258
76 CH DUX 15.1 mm -J4.I mm Relative Displacement (C4)

28 CH A.MCX 906.0 CAL -US 2 GAL Roof Acceleration

_j i u

Fig. 6. Hysteresis Loop of A-Type C4 Column at the First Excitation

259
Fig. 8. Hysteresis Loop of B-Type C4 Column at the Second Excitation

260
No Crack

41.9

After First Damage Excitation

GAL

r.o sec

Fig. 7. Free Vibrations of A-Type Specimen

IS position

(1) Column Head - Support Frame

45.1 nun

37.3 nun 6 mm -1-1-5 mm

(2) Column Head - Column Foot *

Differeoe (l)-(2)

3.7 nun -7.1 Base - Shaking Table

13. '2 nun -9.9 nun !).5 nun

mm 1.0 mm

20.00 SEC,

Fig. 9. Slip of Column Base (CI) of A-Type Specimen at Second Excitation

261
1X-3X 203.6 Relative Displacement (C4)

amcx stu.i gal .629.9 cal Roof Acceleration

Roof Acceleration

T 1000.0GAL

V A

•50.0 mm

Relative Displacement

Fig. 10. Hysteresis Loop of B-Type C4 Column

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265

Response of a Mid-Water Depth Platform

Subject to Earthquakes

by

Charles E. Smith*

ABSTRACT

Seismic analyses of offshore

platforms may be accomplished using

classical finite element techniques

with the input forcing function

modeled for earthquake loadings. The

significant difference between an offshore structure and an onshore

structure is that the analysis

procedure must be modified to

consider the fluid-structure and the

soil-pile interaction effectsCurrent practices use earthquake

forcing functions obtained from

ground accelerations measured in the

vertical and horizontal directions

recorded from onshore sites.

However, recent research has

suggested that the seafloor may not

respond in the same manner as onshore

sites during an earthquake. This

paper will compare the response of a typical mid-water depth offshore

platform subject to both onshore and

offshore recorded ground motions from

the same earthquake. Recommendations

are put forth for the type of seismic

forcing function to be in the design

or requalif icat ion of offshore

structures, as well as relevant

research for the future.

KEYWORDS: Earthquake ground motions,

offshore platforms, offshore

research, platform seismic response,

seismic design criteria.

1. INTRODUCTION

Offshore platforms have several key

structural and environmental

differences as compared to typical

onshore buildings [ 1 ] . The

differences can be grouped into the

following areas: (1) geology and

response of subsea soils (softsaturated soils and lack of offshore

seismic data), (2) a water column

that provides a source of loading and

dampening to the structure (fluidstructure interactions), (3) special

structural characteristics (deep pile

foundation, large mass, large number

of redundant members, larger

overturning moments, and the fact

that the mass is concentrated at the

top of the platform as opposed to

being distributed over its height as

in most building structures), and (4)

the unusual engineering properties of

offshore soil deposits (often

normally consolidated and extremely

soft and compressible near the

seafloor) . Thus, the dynamic

behavior of an offshore structure

during an earthquake may be

significantly different from its

onshore counterpart.

The current (1994) methodology for

the design and/or reassessment of

offshore structures makes use of data

obtained from onshore seismic events.

However, based on the vital

differences between onshore and

offshore structures cited above, a joint program was started in 1976

between the Department of Energy and

the Conservation Division of the

U.S. Geological Survey (now the

Minerals Management Service) to

develop advanced instrumentation

systems to assist government and the

petroleum industry in characterizing

environmental and engineering

conditions offshore. As part of this

program, an instrumentation system

was developed and installed offshore

southern California to collect strong

motion seismic data from remote ocean

floor sites [2]. The program was called the Seafloor Earthquake

Measurement System Program and was given the name SEMS for brevity.

Sleefe[3] studied data obtained from

these units, as well as data obtained

from a second series of SEMS

instruments installed offshore Long

Beach in 1985. Of particular

*Research Program Manager

Technology Assessment and Research

Branch

Minerals Management Service

Herndon, Virginia 22070-4817

267
interest was the data obtained from

the North Palm Springs Earthquake

(6.0 magnitude, 91 mile epicentral

distance) of July 8, 1986. Figure 1 shows a plot of the peak horizontal

accelerations from the North Palm

Springs Earthquake measured from a number of land-based sites, the SEMS

unit, and data recorded at the base

of Shell's Platform Eureka as a function of their epicentral

distances form the earthquake.

Figure 2 shows the relationship

between peak vertical accelerations

and peak horizontal accelerations for

the land-based and SEMS sites. It is

obvious from the data that the

offshore site exhibited a lower peak

vertical acceleration than onshore

sites with comparable peak horizontal

accelerations. Based on this, as well as similar data from the July

13, 1986, Oceanside Earthquake (5.8

magnitude, 46 mile epicenter

distance) , Sleefe makes the following

observations: "The vertical

component of acceleration observed at

both the SEMS site (i.e., free-field

seafloor motion) and on the platform

legs at mud-line level is

significantly different than the

corresponding vertical components of

acceleration observed on-shore. . . . It appears that the peak horizontal

components of acceleration observed

at the SEMS site and by the platform

legs at mud-line level are comparable

to the corresponding onshore

measurements.

The above considerations have

important implications for the design

and requalif ication of offshore

structures located in seismic

regions. With the marked reduction

in the vertical accelerations, as well as any possible changes in the

horizontal accelerations, it is clear

that the design process using such

data would result in a more

economical structure. The design of

topside production facilities, as well as the analysis of a structure

for requalif ication, would greatly

benefit from seismic data obtained

from realistic offshore sites.

Therefore, due to the uncertainties

which exist concerning the use of

onshore seismic data in the analysis

of an offshore platform, a study was undertaken to investigate the dynamic

response of a typical offshore

structure subject to actual seafloorinduced ground motions. The

structure will be analyzed using

strong motion data obtained from an onshore and an offshore site located

at approximately equal epicentral

distances from the actual earthquake.

The information from these analyses

was used to formulate conclusions and

recommendations for future guidelines

to be used in the design of new platforms or in the requalif ication

of existing structures.

2. PLATFORM SEISMIC RESPONSE

ANALYSES

The subject platform is a typical

structure used by the oil and gas

industry to drill for and produce

hydrocarbons in seismic regions of

the world. As shown in Figures 3 and 4, the facilities at the top of

the structure are quite large to

include the drilling, production, and

processing operations, the crew quarters with helideck, the drilling

derricks, and miscellaneous items

associated with working offshore.

Also visible from the rendering is

the massive steel-jacket structure

which supports the topside

facilities. The platform was detailed in accordance to current

U.S. industry standards [4] for

platforms located in seismic areas. Both onshore and offshore scaled

earthquake records (time series of

ground accelerations) were used in

the platform analyses. The platform

foundations were modeled using three

different site conditions to include

a soft clay, a stiff clay, and a dense sand that are representative of

offshore areas as shown in Figure 5.

Time histories of data were obtained

on deck displacements, on member

forces in the steel- j acket , on pile

loads, and on the global base shears

and overturning moments and used to assess the platform response.

3. GROUND MOTION DATA

The 1981 Santa Barbara Island

Earthquake of September 4, 1981,

offers the best opportunity to

compare the effects of offshore and

onshore recordings due to the

availability of compatible data sets.

As shown in Figure 6, the SBVictor

(onshore) and SBHenry (offshore)

sites are located in the same general

direction from the earthquake

268
epicenter and are within 10 percent

of the same epicentral distance.

These recordings were obtained from

similar SEMS instruments, both

onshore and offshore, thus avoiding

another source of uncertainty. Also,

additional recordings from other

onshore stations (SB38 and SB51) were available for comparison. Another

incentive to use this data set was

the fact that Anderson[5] completed

an extensive study of onshore

recordings for the same earthquake

(did not include the SBVictor site)

.

An analysis of the recording for

SBVictor in terms of peak horizontal

and vertical accelerations

corresponds to Anderson's data for

other onshore sites, thus providing

confidence in the SBVictor results.

The SBVictor and SBHenry earthquake

were scaled to match the API response

spectrum using an effective

acceleration of 0.25 g. This

resulted in a scaling factor of 18

for SBVictor and 12 for SBHenry. An

additional analysis was conducted

with a scaling factor of 18 for

SBHenry in order to have a direct

corresponding earthquake for

SBVictor. Table 1 lists the

earthquakes and soil profiles used in

the analyses. Figures 7-9 show the

actual ground motion records used to

conduct the platform analyses.

4. PLATFORM RESPONSE ANALYSES

The platform was analyzed using the

Seastar and Cap computer codes on a SUN IPC Workstation with 24 megabytes

of ram and a 643 megabyte hard disk

drive. The SUN operating system,

WINDOWS, Seastar and Cap Computer

Codes, the platform model, and other

general support programs took over

half of the available memory space,

thus leaving approximately 240

megabytes remaining for computed

results. Therefore, due to the

limited available memory, only

selected node and member responses;

i.e., pile load, base shear,

overturning moment, and deck node

displacements, etc., could be saved.

A typical analysis took upwards of

120 hours of run time and created

approximately 228 megabytes of data.

The data for each run were downloaded

onto one quarter inch magnetic tape

and were re-recorded as necessary to

obtain results.

Considerable attention was given as to what data would be best suited to assess the differences in the

response of the platform when

subjected to the designated

earthquakes. A corner node at the

top of the steel jacket was selected

to judge the deck responses, and a brace in both the longitudinal (Xaxis, stiff) and transverse (Y-axis,

soft) directions was selected to assess member forces. Base shears were obtained in the longitudinal and

transverse directions, as well as the

effective total vertical force

exerted by the platform. The base

overturning and torsional moments

were also calculated, as well as pile

capacities for a corner pile in both

the longitudinal and transverse

faces. Table 2 provides a listing of

the time series response plots

obtained for each run listed in Table

1.

A frequency response analysis of the

platform was conducted for each of

the three soil profiles considered;

i.e., soft clay, stiff clay, and a dense sand. Table 3 shows a comparison of the platform responses

in the first 10 modes for each soil

condition.

Table 4 lists the ratios of the

results from the platform analyses

using onshore and offshore records of

earthquakes as forcing functions.

The SBVictor (onshore - Runs 1, 2,

and 3) and SBHenry (offshore - Runs

4, 5, 6, and 7) earthquake records

show a consistent relationship for

the soft clay site (Runs 1 and 4),

the stiff clay site (Runs 2 and 5),

and the dense sand site (Runs 3, 6,

and 7) in that the offshore site

produced smaller ratios of results

comparing offshore to onshore data.

The ratios of data for the soft clay

site are somewhat smaller than for

the stiff clay or dense sand sites as was previously stated. The peak

values for the acceleration response

spectra for the horizontal directions

of the deck Node 56 are of the same magnitude; however, the periods at

which the peak values occurred are consistently larger for the offshore

records. This could have an impact

on the design of deck equipment

components depending on their period

of response. The periods of the peak

values of the response spectra for

the vertical direction for Node 56

269
are consistent for both the onshore

and offshore data. This would imply

that there is no difference in the

response as a result of the frequency

content of the offshore and onshore

records and that the differences in

the vertical component of deflection

are due only to the differences in

the intensity of the offshore and

onshore earthquakes' vertical

component of motion.

5. CONCLUSIONS

A total of seven computer runs and

three site conditions were used to assess the differences which might

exist in analyzing a typical offshore

platform using either onshore or offshore earthquake records as the

forcing function. Reference 6

presents a more detailed analysis of

earthquake ground motions from

onshore and offshore sites and

contains the results of additional

computer runs that support the

conclusions stated here.

Additional offshore records are needed from different offshore

regions in order to fully generalize

the results. However, the data

obtained from this investigation are still quite significant for the

design or requalif ication of

platforms offshore California.

The following are the main results

from the study:

o The type of foundation

conditions (soft versus stiff)

has a major impact on the

periods of vibration of the

platform. As the foundation

stiffness increases, the

periods of vibration decrease

(frequencies of vibration

increase) . The effect is more

pronounced for the higher

periods of vibration

(fundamental modes) than for

the lower periods of vibration.

The major changes occur in the

first six modes of the platform

used in this investigation.

o The offshore earthquake

recordings produced, for all

site conditions, smaller

displacements and loads within

the platform as compared to

compatible onshore earthquake

records. The horizontal

components of the deck

displacement was of the same magnitude; however, there was a marked difference in the

vertical component of motion by

a factor of 2 to 3; i.e., less

than that produced by an onshore recording. The same trend was also present for the

base shears and overturning

moment with the total.

o The periods of peak horizontal

responses of the deck node for

the offshore records were greater than the periods of

peak horizontal responses for

the onshore records. However,

the periods of peak vertical

responses were equal for both

types of earthquake records.

With the records (both onshore

and offshore) used in the

analyses, it appears that there

are no differences in the

frequency content of the

vertical motions of the deck,

but there are slight

differences in the frequency

content of the horizontal

motions. The differences are more pronounced for lower

periods of motion than for

higher periods of motion. The

major differences in the

vertical components of motion

are due to ground motion

intensities and not frequency

content

.

o Additional offshore earthquake

records are needed from

different regions with varying

soil conditions and water

depths to study and develop

more specific recommendations

concerning the relationship

between the types of records.

A corresponding need exists to

determine the influence of the

various environmental factors;

i.e., soil type, water depth,

etc. on the properties of the

offshore records.

6. FUTURE RESEARCH NEEDS

The following recommended research

needs are not listed in any order of

priority but represent the

uncertainties identified as part of

this paper that would improve the

assessment of the seismic response of

offshore steel- jacket platforms:

270
A need exists to instrument and

create a data base of recorded

responses of steel-jacket

platforms subject to seismic

events. The validity of

computer codes and models

should be tested and calibrated

through the use of field

measurements of actual offshore

structures subject to strongmotion earthquakes. This could

be further enhanced by having

an instrumented free-field

seafloor site near the platform

so that the structure could be

analyzed using actual seafloor

seismic response as a forcing

function. This is extremely

important for strong

earthquakes that would cause nonlinear behavior in the

structure, as well as in the

surrounding soils.

There is a critical need for more earthquake seafloor motion

data, especially in areas of

different geology and local

soil conditions. There is an abundance of data for onshore

sites consisting of soft and

firm soils and for rock.

However, the data for varying

offshore sites are very

limited. Before specific

recommendations can be put

forth for using offshore

recordings, as opposed to

onshore recordings, for the

design of offshore structures,

it is important that additional

offshore recordings be obtained

and analyzed. Seafloor

response data for earthquakes,

especially strong earthquakes,

would be of tremendous

assistance to the offshore

industry

.

Additional information should

be developed on methodologies

for scaling existing earthquake

records. Large earthquakes are extremely rare, and the ability

to obtain such data is

compounded by the small

probability of having an instrument in place at the

desired site to measure the

event. This situation is even

more pronounced for offshore

sites. Recordings from smaller

or more distant events can be

obtained rather easily; yet to

make them useful, they need to

be increased in intensity.

Analytical and experimental

studies should be conducted to

examine the effect of using

different scaling techniques on the response of offshore

structures

.

7 . REFERENCES

1. Poulos, H.G. (1988). Marine

Geotechnics , Allen & Unwin,

Ltd., Wellington, New Zealand.

2. Ryerson, D.E. and Reece, E.W.

(178). "The Development of a Seafloor Earthquake Measurement

System, " Proceedings , Oceans

'78 Conference, Washington,

D.C.

3. Sleefe, G. E. (1990). "The

Long-Term Measurement of

Strong-Motion Earthquakes

Offshore Southern California,"

OTC 6336, Proceedings , Offshore

Technology Conference, Houston,

Texas

.

Recommended Practice for

Planning

,

Designing, and

Constructing Fixed Offshore

Platforms , (1993). API P2A,

20th Edition, American

Petroleum Institute,

Washington, D. C.

Anderson, J. G. (1984). "The

4 September 1981 Santa Barbara

Island, California, Earthquake:

Interpretation of Strong Motion

Data," Bulletin of the

Seismological Society of

America, Volume 74.

Smith, C. E. (1994). "Dynamic

Response of Offshore SteelJacket Platforms Subject to

Measured Seafloor Seismic

Ground Motions," D . Sc

.

Dissertation , The George

Washington University,

Washington, D. C.

271
hik liritMtal Acet lent ions - nC

1000

100

10

© SEMS

Shall Eur*k«

• Land-Basad

©

A

0

100

Epietntnl listwie* - kn

i i i i I 10

Figure 1. Peak Horizontal Accelerations vs. Epicentral Distances

for Onshore and Offshore Sites (North Palm Springs

Earthquake of July 8, 1986)

100

PEAK VERTICLE ACCELERATION- -mG

10

1

• • • •

w

10

LAND-BASED

MEASUREMENTS

SEMS MEASUREMENT

PEAK HORIZONTAL ACCELERATION- -mG

100

Figure 2. Peak Vertical Accelerations vs. Peak Horizontal

Accelerations for Land-Based and SEMS Sites (North Palm

Springs Earthquake of July 8, 1986)

272
Figure 3. Artist's Rendering of the Offshore Platform Selected

for Analysis (Rendering Courtesy of PMB Engineering)

273
® © © ©

Figure 4. Platform Elevations Showing General Arrangement of

Members (Figure Courtesy of PMB Engineering)

274
I. SOFT CLAY

8 lb. LIKE MUDLINE % - 476 ft.

- 200 p«f 7

EFFECTIVE SOIL WT. - 40 l/CF.

HARD SOIL Cu - 2600 psf

II. STIFF CLAY

12 lb. LINE MUDLINE # - 476 ft.

J r - 800 PSF—^

EFFECTIVE SOIL WT. - 50 «/CF.

)IL^ cu " 4400 PSF

III. DENSE SAND

API TABLE 6.4.3-1

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j 7 "i v * jij

DLINE ? - 4"£ ft.

/A , vV.A \ J i d = 3 0 degrees

§ EFFECTIVE SOIL WT . * 551/CF

HARD SOIL \

Figure 5. Soil Profiles and Properties

275
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Figure 6. Location of SBHenry (offshore) and SBVictor (onshore) Recording Sites with Respect to the Earthquake Epicenter

276
Ground Motion Record - Primary Direction

c

o

-H

4-J C\)

<rs < u u

r-\ CO

<D \

U U

U Du < —

8.364

8.114

0.000 Time (Sec) 27.290

Ground Motion Record - Secondary Direction

c

o

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ft! <

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0 0/ —i X

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Ground Motion Record - Vertical Direction

ccH .—

.

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Figure 7. SBVictor; Scaled (x!8) Components of Acceleration Time Histories

277
Ground Motion Record - Primary Direction

8.400

03 < U U

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Ground Motion Record - Primary Direction

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Figure 9. SBHenry; Scaled (xl8) Component of Acceleration Time Histories

279
Table 1 - Earthquake Time Series and Soil Profiles Used in

Assessing Platform Responses

Run No. Earthquake Structure Soil Profile

1 SBVictor x18 Linear Soft Clay

2 SBVictor x18 Linear Stiff Clay

3 SBVictor x18 Linear Dense Sand

4 SBHenry x12 Linear Soft Clay

5 SBHenry x12 Linear Stiff Clay

6 SBHenry x12 Linear Dense Sand

7 SBHenry x18 Linear Dense Sand

^Primary earthquake direction along transverse (Y-Axis) of the platform

Note: Each earthquake includes 3 components (primary, secondary, and vertical

directions) applied to the structure simultaneously.

280
Table 2 - Time History Response Plots Obtained from a Computer Run

Data Plot Item

*1 Node 56 - Deflection Along X-Axis; dx

*2 Node 56 - Deflection Along Y-Axis; dy

*3 Node 56 - Deflection Along Z-Axis; dz

*4 Node 56 - Acceleration Response Spectrum at 5% Damping; dx

*5 Node 56 - Acceleration Response Spectrum at 5% Damping; dy

*6 Node 56 - Acceleration Response Spectrum at 2% Damping; dz

7 Mnrjo IMUUc ROvJ^ *- \/olr^r*it\/ VtJUJOIly Rocnnnco Rt;o|JLMIoc QOfJcUllUIII r\o^ + ri i m ClatL C)\J Q/£\ /O L/Cla m1 1 ipir^i n1 ly, r% r\Uav 8 Node 56 - Velocity Response Spectrum at 0% Damping; dz

9 Node 56 - Fourier Response Spectrum; dx

10 Node 56 - Fourier Response Spectrum; dz

*11 Base Shear; Fx

*12 Base Shear; Fy

1 o 1 ULdl V tSI Lltdl LUdU, l L.

*14 Base Overturning Moment; Mx

*15 Base Overturning Moment; My

*16 Base Torsional Moment; Mz

17 Pile Load; Fx (Pile 1 07 1 -82, Longitudinal Face)

18 Pile Capacity Chart (Pile 1071-82, Longitudinal Face)

19 Pile Capacity Chart (Pile 1 1 41 -93, Transverse Face)

*20 Brace Longitudinal Face (49); Fx

•21 Brace Transverse Face (1715); Fx

22 Node 327 - Deflection Along X-Axis; dx

23 Node 327 - Deflection Along Y-Axis; dy

24 Node 327 - Deflection Along Z-Axis; dz

25 Node 327 - Acceleration Response Spectrum at 5% Damping; dx

26 Node 327 - Acceleration Response Spectrum at 2% Damping; dz

Note: a) Node 56 at Elevation ( + ) 51 ft. b) Note 327 at Elevation (-) 1 96 ft. c) Base Shear and Moments at Elevation (-) 480 ft. d) ^Indicates Primary Data Set

281
Table 3 - Platform Frequency Analysis Summary

Mode No. Soft Clay

fnprinH cop \

Stiff Clav

\ L/C 1 1 WU , OCL/ . / Dense Sand

(nprioH cor \ Description

i

i ? 1 QO 9 fi1 "3 z . JtU I SI I raMSVBiSe

z9 9 QQR 9 4.06 z . tzVJ i si Longiiuainai

o Z9 .UZ.O 098 1 1 1I . O C? "+ i si i orsioriai

At 1 RfiO 0 951 0 Qfid ZIIU 1 idllSVelbe

5 1.415 0.941 0.935 2nd Longitudinal

6 0.91 1 0.808 0.808 2nd Torsional

7 0.809 0.771 0.772 1 st Vertical

8 0.774 0.765 0.765 3rd Transverse

9 0.766 0.761 0.761 3rd Longitudinal

10 0.762 0.741 0.741 Complex Mode

Note: The Longitudinal, X-Axis, corresponds to the platform North-South Direction (West

Elevation); and the Transverse, Y-Axis, corresponds to the platform East-West Direction

(North Elevation) as shown in Figure 63.

o Frequency (Cycles/Sec.) = 1 /period o Circular Frequency (Rad./Sec.) = 2/7 x Frequency

282
Table 4 - Ratios of Primary Data from Platform Analyses

Based on Offshore to Onshore Earthquake Records

Data Plot Number Run 7/Run 3 Run 6/Run 3 Run 4/Run 1 Run 5/Run 2

1. Node 56 - Deflection X-Axis 0 .950 0.659 0 .554 0 .483

2. Node 56 - Deflection Y-Axis 0 .676 0.610 0, cor* .529 0 .593

3. Node 56 - Vertical Deflection 0 .61 1 0.458 0 .248 0 .315

4. Node 56 - Horizontal Accel. Resp.

Spectrum at 5% Damp, (dx) 1,.207 0.802 0 .813 0 .969

Period 1,.601 1.601 1,.149 1 .599

5. Node 56 - Horizontal Accel. Resp.

Spectrum at 5% Damp, (dy) 0,,887 0.588 0,,726 0 .669

Period 1.,315 1.315 1,,640 1 .369

6. Node 56 - Vertical Accel. Resp.

Spectrum at 2% Damp, (dz) 0,,565 0.387 0 329 0 ,357

Period 1.,000 1.000 0,,953 1,.009

11. Base Shear (Fx) 1.,060 0.798 0, 412 0,,698

12. Base bnear (ry) u, boo U.4U1 u, boy (J oyu

13. Total Vertical Load (Fz) 0. 462 0.323 0, 258 0, 369

14. Base Overturning Moment (Mx) 0, 586 0.401 0. 469 0,,591

15. Base Overturning Moment (My) 0. 877 0.603 0. 576 0.,442

16. Base Torsional Moment (Mz) 0, 718 0.533 0. 598 0, 916

20. Brace Longitudinal Face (Fx) 0,,691 0.461 0. 449 0. 496

21. Brace Transverse Face (Fy) 0,,565 0.419 0. 459 0.,464

Note: a) Run Numbers correspond to Computer Runs listed in Table 1

.

b) Runs 4( 5, 6, and 7 are from offshore recordings, Runs 1, 2, and 3 are from onshore recordings.

283

Uplift of Sewage Manholes During 1993 Kushiro-Oki Earthquake

by

Osamu Matsuo", Hisanori Otsuka2

', Yoshio Ninomiya3

' and Junichi Koseki 4'

.ABSTRACT

Boring and cut-off investigations were

carried out to survey both the original

ground and back-filled soil surrounding

sewage manholes damaged by the 1993

Kushiro-oki earthquake. Based on these

results, a liquefaction resistance ratio

Fu and a safety factor against uplift Fs were estimated, and those factors were

compared with the observed uplift

displacement of the manholes.

KEYWORDS: Sewage Manhole, Uplift, 1993

Kushiro-oki Earthquake, Liquefaction

1. INTRODUCTION

The Kushiro-oki Earthquake on January

15th, 1993 caused severe damage to

roads, railways, port and harbor

facilities, lifeline infrastructures and

fill-up ground for housing lots 1523

. Among these, several wastewater

manholes in Kushiro Town along National

Highway No. 44 suffered uplift amounting

up to 1.5m due to liquefaction of

surrounding ground. This paper deals

with the results of intensive survey

which was conducted after the

earthquake to reveal the mechanism of

uplift.

2. SURVEY METHODS

Location of surveyed manholes is

shown in Fig. 1. All 18 manholes in the

section requiring restoration work were

investigated. They consist of 5 manholes

on Nichii route, 6 on Ishiguro Homer

route and 7 on Satsuru Veneer route.

The total length of installed sewer

pipes in the investigated section is

723m.

Contents of the survey are summarized

in Table 1. Before the restration work,

level survey and boring survey in

conjunction with standard penetration

tests for both the original ground and

back-filled soil surrounding the

manholes were conducted. During the

restoration work, which was executed by

open-cut method, the soil profile was observed to find any sign of ground

alteration which is suggestive of

liquefaction effect. Cyclic triaxial

tests were performed using undisturbed

tube samples for the original ground

and block samples for the back-filled

soil.

Based on these results, a liquefaction

resistance ratio Fu35 and a safety

factor against uplift Fs*)ca were

evaluated as follows.

R= R, + Ra + R3 (2)

1) Head, Soil Dynamics Div., Public Works

Research Institute, Ministry of

Construction, Tsukuba Science City,

Japan

2) Head, Earthquake Engineering Div.,

ditto (formerly head, Ground Vibration

Div.)

3) Senior Research Engineer, Ground

Vibration Div., ditto

4) Research Engineer, Soil Dynamics

Div., ditto

285
L=r d -k8

- -^r (3)

where R: dynamic shear strength ratio,

L: shear stress ratio during an earthquake,

Rt

: first terra of R represented by a function of SPT N-value and effective

overburden pressure a v

'

,

R2: second term of R represented by a function of mean grain diameter DBq, R3: third term of R represented by a function of fines content FC,

r d : reduction coefficient along depth,

kQ : horizontal seismic coefficient on the

ground surface defined by kQ=amax /g

(amax is the maximum horizontal

acceleration on ground surface and g is a gravitational acceleration), and

a v : total overburden pressure.

Ud=A-Lu-o vb ' (5)

Lu= FZ-p (if FZ<1.0 then Lu=l) (6)

where W: weight of manhole,

Q: frictional resistance force between

the side wall of manhole and

surrounding ground (a section which is

shallower than the water level was

considered),

Us: buoyancy force due to static

hydraulic pressure,

Ud: uplifting force due to excess pore

water pressure caused by liquefaction,

A: cross sectional area of manhole on a horizontal plane,

Lu: excess pore water pressure ratio,

ovt>': effective overburden pressure in

surrounding ground at the same depth as the bottom of manhole,

FT: liquefaction resistance ratio

averaged over the depth of interests

(refer to 3.4 for layers considered in

the calculation), and

p: parameter representing the

characteristics of excess pore water

pressure generation (in ref.4, it is set

to be 7).

3. RESULTS AND DISCUSSIONS

3.1 Ground Profile

The general ground profile is

summarized in Table 2. The original

surface of the surveyed ground had

been peat (denoted as Ap), and it was

filled with sand (F) dredged from the

Old Kushiro River around 1962. The

thickness of alluvial deposit is between

30 to 40m. Its shallower part mainly

consists of sandy soil which can be

classified into two groups. The upper

sandy layers (Acsl, Acs2 and Asl) are

with mean SPT N-value 3 to 17 and with

fines content 20 to 30% approximately.

The lower sand layer (As2) is with mean

SPT N-value more than 30 and fines

content 10% approximately. Alluvial

clay (Ac) is occasionally sandwiched

between Acsl and Acs2. The grain size

distribution is shown in Fig.2.

The estimated ground profile along

each route is shown in Fig.3 as well as SPT N-value for the original ground

(denoted as N) and the back-filled soil

(B). The boundary between the upper and

the lower sand layer is slightly

inclined to downstream side along each

route. The ground water level is 2m

below the ground surface on the whole.

The back-filled soil is with mean SPT

N-value 3 and with fines content 8%

approximately.

3.2 Condition of Manholes

The standard cross section of

facilities buried under the road is

shown in Fig.4. After excavating the

original ground by 1.3 to 1.6m in width

and 2 to 5m in depth, the wastewater

manholes and sewer pipes were installed

and were back-filled by dredged sand

and commercial sand brought from

mountainous area. They were put into

286
operation in 1978.

Adjoining manholes for storm-water were not damaged, which had shallower

foundation than the uplifted manholes

for wastewater. Some part of connecting

sewer pipes for storm-water are reported to have been bent slightly.

The composition of the wastewater

manholes is shown in Fig. 5. The main

body is a precast-concrete pipe with a diameter of 1,050mm. Underneath its

bottom plate, a latice-shaped wooden

foundation was placed with crushed

stones.

Observed height of the steel cap from

the ground surface after the earthquake

(denoted as uplift displacement),

estimated settlement of the ground

surface due to the earthquake and

original depth of the bottom plate of

each manhole are shown in Fig.6. The

maximum uplift displacement was 1.46m at

No. 65 on Nichii route, 0.56m at No. 92 on

Ishiguro Homer route and 1.24m at No.

9

on Satsuru Veneer route. Manholes which

did not uplift had relatively shallow

foundation, and the uplift displacement

increased as the settlement of the

ground surface developed. However the

site at which uplift displacement was

the maximum did not coincide with one at

which ground settlement was the

largest.

3.3 Condition of Surrounding Ground and

Connecting Pipes

The cut-off investigation revealed

that the manhole bodies were not

damaged except some cracks at several

sites. Underneath the bottom of the

uplifted manholes, a void was not

observed and the occupying sand was

similar to the back-filled soil. Although

sign of ground alteration indicating

liquefaction effect such as penetration

of liquefied sand into neighboring layer

was not observed, the back-filled soil

with low SPT N-value as shown in Fig. 3 is

judged to have liquefied and moved

laterally toward the bottom of the

manholes.

The latice-shaped wooden foundations were found almost at the original depth.

Although this may imply that no ground

movement occurred under the

foundation, further investigations are required on its effect.

At some manholes where uplift

displacement was large, the grain size

distribution of the sand underneath the

manhole was similar to that of

adjoining original ground such as Acsl.

This suggests that lateral movement of

liquefied soil in the original ground

may have occurred at these manholes.

Pipes connecting the uplifted manholes

also moved upward without any breakage

at the connection with the manholes. At some sections the vertical displacement

of the sewer pipes was larger than the

uplift displacement of the manholes.

This supports the judgement that the

back-filled soil was liquefied, which was

described in the first paragraph of this

section.

3.4 Liquefaction Assessment of Original

Ground

The liquefaction resistance ratio FZ

averaged over the layers Acsl, Acs2 and

Asl is shown in Fig.3. The layer As2 was excluded from the assessment, because

its SPT N-value was more than 30. When

the seismic coefficient kB is 0.2, VZ was

not more than 1.0 except at manholes

No. 63 and No. 67. It was less than 1.0 at

all manholes when kQ is 0.25.

For reference of kB , the recorded

maximum horizontal acceleration on ground surface was 320gal at the foot of

a levee in the Kushiro River which was

located at an epicentral distance of

23km. Because the uplifted manholes

were located at an epicentral distance

287
of 19km, actual value of kQ is estimated

to be at least 0.25. This implies that

liquefaction may have occurred in the

original ground surrounding the

manholes.

Comparison of the dynamic shear

strength ratio R estimated from the

results of boring survey and R2q from

cyclic triaxial tests using undisturbed

tube samples of the original ground is

shown in Fig. 7, where R20 is defined as the ratio of cyclic shear stress causing

5% double amplitude axial strain in 20

cycles to initial effective confining

pressure. For the layers Acsl and Acs2,

R2D was almost equal to or slightly

larger than R. For the layer Asl, R20 was smaller than R. Reduction of R2D due

to sample disturbance may have affected

the results for Asl, because it was

relatively clean sand with fines content

of 13%.

Relationship between uplift

displacement of the manholes and the

estimated total thickness of liquefied

layer in the original ground is shown in

Fig.8, where the liquefied portion was

defined as a layer with FL less than 1.0. When kQ is 0.15, the uplift displacement

increased as the liquefied layer got

thicker. Although kQ may have been

larger than 0.25, as mentioned above,

the liquefied layer estimated using

kQ=0. 15 can be regarded as the one which

liquefied extensively during the

earthquake. This suggests that

extensively liquefied layer in the

original ground as well as the

back-filled soil may have affected the

uplift of manholes.

Relationship between settlement of the

ground surface and the estimated total

thickness of liquefied layer in the

original ground is shown in Fig.9. As

well as uplift of the manholes, ground

settlement increased as the liquefied

layer estimated using kQ =0. 15 got

thicker.

3.5 Stability of Manholes against Uplift

Relationship between uplift

displacement of manholes and a safety

factor against uplift Fs based on FZ in

the original ground is shown in Fig. 10. The parameter p, which was used in the

prediction of excess pore water

pressure, was evaluated to be 2.1 on the

average by anal i zing the result of

cyclic triaxial tests.

When the seismic coefficient kB is 0.2,

which is lower than the value estimated

based on observed strong motion, Fs was

less than 1.0 at all manholes

irrespective of the uplift displacement.

This indicates that the evaluation of

the stability against uplift based on Fs

may have some safety margin.

However when kQ is reduced to 0.15, Fs was still less than 1.0 at manholes with

large uplift displacement. On the

contrary, it was more than 1.0 at other

manholes except one. This indicates that

the uplift displacement agreed with Fs

qualitatively by reducing kQ and

supports that original ground may have

affected uplift of manholes as mentioned

in 3.4. Further investigations are needed on the mechanism of uplift of

manholes including the wooden

foundation in a lattice shape which did

not move during the earthquake.

4. SUMMARY AND CONCLUSIONS

Based on the cut-off investigation and

SPT N-value, the back-filled soil of the

manholes is judged to have liquefied and

moved laterally toward the bottom of

the uplifted manholes. An analysis of

the average liquefaction resistance

ratio FZ, which was evaluated using SPT

N-value, grain size distribution and

horizontal seismic coefficient kQ , implies that liquefaction may have

288
occurred in the original ground

surrounding the manholes.

Uplift displacement of the manholes

and ground settlement increased as the

extensively liquefied layer, which was

estimated based on FL, got thicker. The

uplift displacement also agreed with the

difference of safety factor against

uplift Fs qualitatively by reducing kB . Further investigations are needed on

the mechanism of uplift considering the

lattice-shaped wooden foundation which

did not move during the earthquake.

ACKNOWLEDGEMENTS

This investigation was carried out by

PWRI and Japan Sewage Works Agency in

cooperation with Kushiro Town. The

authors wish to thank the concerned

persons for their thoughtful and

helpful assistance.

REFERENCES

1) Iijima, T. et al.: Damage to

Infrastructures Caused by the 1993

Kushiro-oki Earthquake, 25th UJNR Joint

Meeting on Wind and Seismic Effects,

1993.

2) Murota, T. et al.: Damage to Buildings

and Housing Lots Caused by the

Kushiro-oki Earthquake, 1993, 25th UJNR

Joint Meeting on Wind and Seismic

Effects, 1993.

3) Japan Road Association:

Specifications for Highway Bridges, Part

V: Seismic Design, 1990.

4) Japan Road Association: Design Manual

for Common Utility Ducts, 1986 (in

Japanese).

5) Koseki J. and Koga, Y.: Uplift of

Semi -buried Structures in Liquefiable

Sands during Earthquake, Proc. of 8th

Japan Earthquake Symposium, pp. 933-938,

1990.

Table 1 Contents of Survey

Route Nichii Route Ishiguro Homer Route Satsuru Veneer Route

Level

Survey

No. 63, 64 , 65, 66, 67 No. 91, 92, 93, 94, 95, 96 No.7, 8, 9, 10, 11, 12, 13

Boring

Survey

No. 63, 65, 67 No. 91, 92, 94, 95 No.8, 9, 11

Cut-off

Investigation

No. 64, 65, 66 No.91, 92 , 93, 94, 95 No.8, 9, 10, 11, 12, 13

Cyclic

Triaxial Tests

No.65 (Acsr\ Acs2),

No. 66 (B)

No. 92 (Acsl, Acs2, B) No.9 (Acs2, B), No. 11 (Acs2, Asl, B)

1) Refer to Table 2 for notations.

Table 2 General Ground Profile

Profile Notation Major Soil SPT N- value

Back-filled Sand B Fine sand with some silt 0-11 (3)"

Original

Ground

Filled Sand F Fine sand (with some silt) 1-4 (2) Peat Ap Peat, Organic soil 1-3 (2) Upper Sandy Soil Acsl Silty fine sand, Medium Sand 1-21 (7) Clayey Soil Ac Silt, Sandy clay 1-5 (3) Lower Sandy Soil Acs2 Silty fine sand 1-13 (5) Upper Sand Asl Fine sand with some silt 8-38 (17)

Lower Sand As2 Fine sand, Medium sand 16-50 (43)

1) Minimum— Maximum (Mean value)

289
Surveyed Area

Kushiro City

downstream

side

Katsuragi

1-chome (Nichii route)

Restoration

64 65 66 67

Repairment

69 70 71 72 73

4' /

upstream side

91 92 93 94 95 Restoration §96 2 4 5 Repairment

(Ishiguro Homer route)

Kiba 1-chome

»

= »

<= o

8- Kiba 2-chome 5= 3

S=| a sr. s —

3 =

14

1312*

11

10

9

50 1J0 () ; Lumberyard

Legend

. Location of

Manhole

Restoration:

Section Requiring

Restoration Work

Fig. 1 Location of Surveyed Manholes

Repai rment: Section Requiring

Repairment

downstream

side

290
291
upstream side downstream si

(3) Satsuru Veneer route

Fig. 3 Estimated Ground Profile (continued)

(along Nichii route)

Wastewater

Catch Basin e

Wastewater Manhole -*

Replaced Sand

\ Storm-water

Sewer Pipe

LfjTLXjT Wastewater Sewer Pipe

Back-filled Area 1-800-PioO-S * 300

Unitrmm

Fig.4 Standard Cross Section of Buried Facilities

292
Block for Height Adjustment

Top Plate

Side Wall

(Precast-concrete Pipe)

Wastewater Sewer Pipe

/ Invert

Bottom Plate

Crushed Stone

Lattice-shaped

Wooden Foundation

Unit:mm

Fig. 5 Composition of Wastewater Manhole

.. 0 0.1 0.2 0.3 0.L

° Dynamic Shear Strength Ratio R

Estimated from Results of Boring Survey

(Grain size distribution was adjusted to that

of samples used for cyclic triaxial tests)

Fig.7 Comparison of Dynamic Shear Strength Ratio

293
Uplift

Displacement

of Manhole

(cm)

150

100

50

n

downstream side upstream side

Settlement

of Ground

Surface (cm)

010

2030 Factor of 1.5 Safety against 1 -° Uplift Fs °- 5

Averaged

Liquefaction

Resistant

Ratio FT

Uplift

Displacement

of Manhole

(cm)

0 1.5

L. 0

0. 5

0

73 — Manhole No.

0 . 2 Original Depth

1 of Bottom Plate

6 from Ground

Surface (m)

(1) Nichii route

downstream side upstream side

Settlement

of Ground

Surface (cm)

o1020 Factor of 1^5 Safety against 1.0 Uplift Fs 0.5

0 Averaged 1.

5

Liquefaction 1-0 Resistant 0. Ratio VZ 0

5 — Manhole No.

^ t

-t

-x

II 1=I 1—

,

|_ .

I Tj J 1=

' Original Depth

of Bottom Plate

from Ground

Surface (m) —

»-A

--a

—

0

-$

Uplift

Displacement mo

upstream side

150

(2) Ishiguro Homer route

downstream side

of Manhole

(cm)

/

1 1 »-—

Settlement

of Ground

Surface (cm)

n 13 12 11 10

0102030 Factor of j 5 Safety against 1.

0

Uplift Fs 0.5

=0" E

Averaged

Liquefaction

Resistant

Ratio FT

0 1.5

0 .5

0

7— Manhole No.

0

2 Original Depth

4 of Bottom Plate

6 from Ground /

Surface (m)

9—

5

(3) Satsuru Veneer route

- Legend

for Settlement

and Depth

"O - I Settlement of

Ground Surface

-jv- % Original Depth

U of Bottom Plate

for Fs and FZ

-O- : ks=0. 15 --A-- : ks=0. 20 -O- : k s=0.25

Fig. 6 Distribution of Uplift Displacement of Manhole, Ground Settlement, FZ and Fs

294
c

o> ^

01 | —. x

Cl o>

O : ks«o. 15 A : ks=0. 20

: k s=0. 25 0 &

c

> / /

/ O A

/

/ -JO-X. a /< -

5 10 if Estimated Total Thickness of

Liquefied Layer in Original Ground (m)

Fig. 8 Uplift Displacement of Manholes vs. Estimated

Total Thickness of Liquefied Layer

=

oCm~

3X—

aouo

O : ks=0. 15 A : k s=0. 20

: k s=0. 25

c A n

*

t' O A .

o,-'

a

0 5 10 15 Estimated Total Thickness of

Liquefied Layer in Original Ground (m)

Fig. 9 Settlement of Ground Surface vs. Estimated

Total Thickness of Liquefied Layer

c

= -s

0) E

w u

GS ^

a 8 a —

Q ©

— •—

=3 O

o

a o .

O : ks=0. 15 A : ks=0. 20

: ks=0. 25 p=2.1

i

i

i

i

(based c triaxia

>n cyclic

I tests)

CD

no * \

<

, o

1—2_Q

>

i 0 0.5 I 1. 5 2 2.5 Estimated Safety Factor against Uplift Fs

Fig. 10 Uplift Displacement of Manholes vs. Esimated Safety Factor against Uplift Fs

295

Survey of Research in the Area of Seismic Response of Concrete Dams

by

Robert L. Hall, PhD* and

William E. Roper, PhD, PE**

ABSTRACT

The Corps of Engineers (CE) has started an earthquake engineering research and

development effort called the Earthquake

Engineering Research Program (EERP). The

EERP will address needs in the general areas of

earthquake ground motions and site

characterization, material behavior, and the

seismic response of concrete dams. The

priorities of the EERP have been established by

both structural and geotechnical field review

groups consisting of the Corps' premier

engineers who are actively designing new

structures or the remediation of existing

structures. The EERP is closely coordinated with

representatives from the Office of the Chief of

Engineers in Washington, D.C., who solicited

the assistance of the Civil Engineering Research

Foundation (CERF), which is part of the

American Society of Civil Engineers (ASCE).

CERF's role was to ensure that the proposed

structural research was a well-developed

program which would: (1) develop the state-of- the-art design and analysis tools dealing with all aspects of concrete dams and appurtenant

structures, (2) identify any ongoing or planned

research which would either duplicate or complement the Corps' planned research, and

(3) identify any possible opportunities to develop

partnerships with others to meet the objectives of

the planned research program. This paper

summarizes the findings of CERF's activities. KEYWORDS: concrete dams; earthquake

engineering program; nonlinear response; intake towers.

1. INTRODUCTION

The CE has begun an earthquake engineering

research and development effort (Roper, et al, 1992) called the Earthquake Engineering

Research Program (EERP). The EERP will

address needs in the general areas of earthquake

ground motions and site characterization,

material behavior, and the seismic response of

concrete dams. The Corps has an inventory of

22 concrete dams subject to potentially

significant earthquake ground motions. The

failure of any of these critical structures would

typically result in more damage than that with

the 1989 Loma Prieta or 1994 Northridge

earthquakes. The concrete portion of the

Folsom Dam serves as an example of typical

damage which would occur as a result of a dam

failure. Studies from about 10 years ago

indicate that the damage would exceed

$6,000,000,000 and 6,000 lives would be lost.

It is planned that the EERP will coordinate

closely with other organizations conducting

research pertinent to the seismic response of

dams. In the development of these plans,

CERF, which is part of the ASCE, was

contacted to assist the Corps. CERF's role was

to ensure that the proposed structural research was a well-developed program which would: (1)

develop the state-of-the-art design and analysis

tools dealing with all aspects of concrete dams

and appurtenant structures, (2) perform literature searches on recent publications in the area of

earthquake engineering related to hydraulic

structures, (3) identify any ongoing or planned

research in the U.S. and worldwide which would

either duplicate or complement the Corps'

planned research, and (4) identify any possible

opportunities to develop partnerships with the

Corps to meet the objectives of the planned

* U.S. Army Engineer Waterways Experiment

Station, Vicksburg, MS 39180-6199

** U.S. Army Corps of Engineers, Washington,

D.C. 20314-1000

297
research program. This paper summarizes the

findings of these efforts.

2. BACKGROUND

The objective of the EERP will be to develop

the necessary validated analytical tools and

corresponding guidance to predict the dynamic

response of the Corps' critical structures from

elastic behavior to the development of failure

mechanisms. The understanding of the failure

mechanisms will allow the determination of the

ultimate capacity of these critical structures during and after structural response induced

from earthquake ground motions. The results of

past research have recently resulted in the saving

of millions of dollars in rehabilitation costs. However, there are no analytical tools or guidance for considering the nonlinear response

of these structures. The accomplishment of this objective will be through a program utilizing the

expertise in the U.S. and worldwide.

3. EARTHQUAKE ENGINEERING

PROGRAM

Details about the EERP have been described in two publications. The first paper (Roper et al, 1992) provides information about the entire

program which includes both geotechnical and

structural research. The second paper (Hall and

Roper, 1993) describes the structural aspects of

the program which were reviewed by CERF.

The current structural aspects of the EERP are

the following areas:

Concrete Dams (Area 1) - Focus on research to determine the dynamic ultimate capacity of

concrete gravity, roller-compacted, and concrete

arch dams. This research will address the

nonlinear behavior and failure mechanisms of

the different types of dams as well as component

tests, scale-model tests, and tests of existing

structures.

Intake Towers (Area 2) - Research to determine

dynamic ultimate capacity of intake towers. One

major area will be to determine the ductility of

existing lightly reinforced structures. Other

research will focus on other nonlinear behavior

and failure mechanisms including stability of

these reinforced concrete structures.

Earthquake Analysis Tools (Area 3) - Develop

analytical tools to support the research on

concrete dams and intake towers. This research

will produce postprocessing tools to aid in the

interpretation and understanding of output from

time-history and response spectrum analyses.

4. STUDY OVERVIEW

CERF surveyed 90 organizations from

government, academia, private industry, and

foreign countries for information related to the

EERP (CERF, 1993). The survey documentation which was sent to each organization

contained a letter explaining the survey, the

priorities of the EERP, and a list of the

organizations being contacted. Responses were

obtained from each organization by responding

to the initial mailing or by follow-up telephone

calls. All responses are given in Vol 1 of

CERF's report.

Thirty of these organizations indicated an

interest in partnering with the EERP on at least one of the three structural areas of research and

development. CERF considered 14 of these to be excellent candidates for partnering and 6

others as good candidates for partnering. The

selections were based on the responses to the

survey and the applicability of their R&D to the

EERP. Table 1 provides a summary of these

organizations and their research and

development areas of interest. The following are summaries of the capabilities of the 14

candidates and partnering opportunities.

4. 1 National Center for Earthquake Engineering

Research, Buffalo, NY

The National Center for Earthquake Engineering

Research (NCEER) is a National Science

Foundation-funded Research Center. NCEER

conducts research in the areas of buildings, nonstructural components, lifelines, and highways.

The NCEER utilizes a consortium of researchers

in universities and private industry throughout

the U.S. in performing their research. NCEER

does perform research on large hydraulic

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structures or large mass concrete structures. They are conducting some research on dams and

appurtenant structures in association with their

Lifeline Project. Their research on analytical

tools for time-history analysis of lightlyreinforced concrete structures, soil structure

interaction, and seismic-induced displacements on buried and partially buried structures complements milestones of the EERP.

4.2 Bureau of Reclamation, Denver, CO

The Bureau of Reclamation has designed and

constructed over 400 concrete and embankment

dams. Since many of these are in high seismic

zones, the Bureau has conducted R&D to support their seismic evaluations of their

extensive inventory of dams. These seismic

evaluations consist of seismotectonic studies,

ground motion studies, geotechnical site investigations, and structural analyses. These

evaluations are conducted with close

coordination to provide confidence in the

conclusions of the studies.

The Bureau provided the following five

examples of their R&D which are directly

applicable to the EERP:

• Modification of ADAP-88 to include the

effect of contraction joints and nonlinear

behavior of concrete with constitutive modeling.

• Proposed funding of centrifuge modeling of

concrete arch dams with contraction joints.

• Capability to perform laboratory vibration

experiments to determine dynamic characteristics

of concrete.

• Software development of seismic analysis

tools.

• Investigation of a combined analytical

approach to concrete dams and their foundations,

including abutments.

4.3 University of California, Berkeley, CA

Professors Anil K. Chopra and Gregory L. Fenves conduct extensive research in the area of

seismic response of concrete dams and

appurtenant structures at the University of

California, Berkeley. Some of their research

accomplishments and ongoing research are :

• Earthquake analysis of concrete dams

including soil-rock interaction, reservoirstructure interaction, and reservoir-sediment

interaction. The time-history procedure has

been implemented in the computer program,

EAGD-84.

• A simplified analysis procedure including all the important interaction effects and the

fundamental mode of vibration.

• Investigation of the nonlinear structural response in gravity dams from water cavitation

and the tensile cracking of the concrete.

• Development of nonlinear analysis

procedures to determine the effects of

contraction joint openings on arch dams.

• Development of a nonlinear procedure for

determining the earthquake sliding stability of

concrete gravity dams.

• Development and implementation of a unified constitutive model for mass concrete for

determining the earthquake behavior of

concrete in dams.

4.4 Ray W. Clough, Sc.D., Structures Consultant, Sunriver, OR

Dr. Clough is a world-renowned expert in the

field of earthquake engineering research for concrete dams and appurtenant structures. He

has worked closely with the CE on many

earthquake-related R&D programs and projects.

His contributions in this field are demonstrated

by the many research projects successfully

completed at the University of California,

Berkeley, and the current design guidance within

the CE.

4.5 Agbabian Associates, Pasadena, CA

Agbabian Associates is a private consulting firm

specializing in earthquake engineering and

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structural dynamics. They also perform applied

R&D, analysis, and designs related to hydraulic

structures. Recent applied R&D (internally and

externally funded) projects include:

• Retrofit design for large, complex highway

bridges.

• Seismic qualification of components and

systems.

• Seismic risk analysis.

• Seismic risk/mitigation development for

museum contents.

• Design, installation, and operation of a network of vertical strong-motion instrumentation arrays in California (funded by the

Japanese government) for study of soil site response.

• Retrofit design for an intake tower, based

upon a new nonlinear analysis technique.

• Geophysical measurements and analyses, some related to seismic analyses of dams.

• Seismic analyses of several concrete dams

and field studies including forced-vibration

studies to verify analytical models.

4.6 Quest Structures, Emeryville, CA

Quest Structures is a private engineering

consulting firm which specializes in earthquake

engineering and has been working with the CE

on many earthquake engineering projects.

Dr. Yusof Ghanaat, Principal Investigator, states that their research interests in earthquake

engineering are primarily focused on the

earthquake response prediction of concrete dams

validated by field measurements and dynamic

response during actual earthquakes. Details were given about experimental studies of damwater-foundation interaction conducted on the

Dongjiang Dam, the highest double-curvature

arch dam in China. Quest Structures has

assisted with the following CE projects:

• Portuguese Arch Dam, Puerto Rico.

• Seven Oaks Inclined Intake Tower, CA.

• Olmsted Locks and Dam, KY.

• Montgomery Point Dam, AR.

4.7 Professor Xia Songyou, Hohai Unversity,

China

Professor Songyou's research efforts have been

focused on the seismic response of concrete dams, navigation locks, and electric

powerhouses. The following is a brief summary

of his past and planned earthquake engineering

activities:

• Modeling the seismic response of concrete dams and powerhouses.

• Foundation-structure-reservoir interaction.

• Model ing earthquake-induced water pressure

in navigation locks.

• Modeling the seismic response of nuclear

power stations.

• Modeling the earthquake-induced excitation

for foundation-structure-reservoir interaction

effects.

• Modeling the response of the concrete face

of rock-filled dams.

4.8 Professor Zhang, Chuhan, Tsinghua

University, China

Professor Zhang, Chuhan has been engaged in earthquake engineering research on dams since

the Hsinfengjiang concrete dam in China was

severely damaged in a 1962 earthquake. A

summary of R&D activities at Tsinghua

University is:

• Preliminary design of the Laxiwa arch dam

(250 m high).

• Feasibility study of the Yunnan Province

arch dam (290 m high).

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• Numerical modeling of the infinite canyon

and dam-foundation-reservoir system.

• Effects of radiation damping and inertial mass of the canyon on arch dam response.

• Effects of nonuniform input and wave

propagation on arch dam response.

• Numerical modeling of reservoir

sedimentation based on Biot's saturated porous

theory and effects on arch dam response.

• Seismic cracking analysis of concrete gravity

dams using fracture mechanics.

• Numerical modeling of dynamic behaviors

of Roller-Compacted Concrete (RCC) gravity

dams including effects of layering.

• Dynamic tests of RCC specimen cut from

concrete layer from the field.

• Development of numerical model for static and dynamic stability analyses of RCC.

• Dynamic modeling of arch dam abutments,

3-D discrete element model for evaluation of

seismic safety of arch dam abutments.

• Field measurement and numerical

simulations of the rocky slope stability using

explosive techniques.

4.9 Public Works Research Institute, Ministry

of Construction, Japan

The Public Works Research Institute (PWRI) has

been active in the areas of earthquake

engineering R&D of concrete and steel hydraulic

structures. PWRI has been actively involved

with the Panel on Wind and Seismic Effects,

UJNR, for the exchange of technical information on wind and seismic effects between the U.S.

and Japan for the past 24 years. They are presently working with this Panel to organize a new Task Committee for Dam Structures. Some

R&D conducted by PWRI includes:

• Seismic design for hydraulic structures.

• Linear and nonlinear dynamic behavior of

concrete dams.

• Strong motion observation of dams.

• Dynamic interaction of dams including damfoundation interaction and dam-reservoir

interaction.

• Seismic design of steel gates.

4.10 Swiss Federal Office for Water Economy,

Division of Safety of Dams.

The Swiss government performs earthquake

engineering R&D through in-house resources and contracts to universities. Their current earthquake engineering R&D in hydraulic

structures is summarized:

• Development of a nonlinear analytical

method of analysis accounting for effects of

interactions between reservoir, foundation, and

concrete dams.

• Creation of a strong-motion instrumentation

network including a free-field network of 33

accelerographs and a network of 29

accelerographs on four different dams.

4.11 ANATECH Research Corp., Lajolla, CA

ANATECH has been performing earthquake

engineering R&D for the Electric Power

Research Institute and the California Department

of Transportation. The following is a brief list of ANATECH's earthquake engineering R&D:

• Modeling of the earthquake-induced excitation

accounting for foundation-structure-reservoir

interaction effects.

• Modeling the response of the dam-foundation

structure to determine the ultimate capacity of

the dam with respect to the concrete and damfoundation interface strengths and material

properties.

• Developing analytical models with field and

physical model verification.

301
• Developing extensive earthquake analysis

tools, including pre- and postprocessing design

aids.

4. 12 University of Texas at Austin, TX

The University of Texas at Austin and the

Ferguson Structural Engineering Laboratory

have been active in research related to earthquake engineering. Their work has not been directly related to hydraulic structures but

they have conducted research for concrete and

steel structures which could be applicable to hydraulic structures as well. The following is a summary list of their R&D activities:

• Behavior of slab-beam-column joints under

simulated seismic loads.

• Behavior of columns under uni- and bi- directional cyclic loadings.

• Development of technical data for design of

rehabilitation work on inadequate existing

buildings.

• Behavior and design studies of welded

connections in steel systems subject to simulated

seismic loading.

• Behavior and design of connections in composite frame systems under monotonic and

cyclic loading histories.

• Soil -structure and soil-dynamic studies.

4.13 Applied Technology Council, Redwood

City, CA

The Applied Technology Council (ATC) is a nonprofit corporation established in 1971

through the efforts of the Structural Engineers

Association of California. The purpose of ATC

is to assist the design practitioner in structural engineering and related fields in keeping abreast

of and effectively using technological

developments. ATC has worked with the CE by

giving seminars on post-earthquake building

evaluation.

4.14 Southeast Missouri State University, Cape

Girardeau, MO

The Center for Earthquake Studies at Southeast

Missouri State University has a partnership with

the Missouri State Emergency Management

Agency for dissemination of earthquake risk and

preparedness information to governments,

schools, businesses, health-care facilities, and

the general public.

5. COMMENTS ON EERP PRIORITIES

The survey indicated that the priorities

developed for EERP are important and

applicable to current dam safety issues. The

Bureau of Reclamation provided the following

comment: "We feel the proposed research noted

in your memorandum is vital to our industry and

applaud your efforts." Dr. Yusof Ghanaat of

Quest Structures states, "The research topics

identified by the CE include specification of

seismic input, modeling of nonlinear response

behavior, and verification of analytical

procedures with field and model tests. These

constitute essential research needed to perform a

realistic analysis for seismic safety evaluation of

the hydraulic structure."

6. CONCLUSION

The results of CERF's survey have identified

ongoing or planned research which will assist the CE in locating partners to conduct the

necessary R&D to develop state-of-the-art earthquake design and analysis tools for all aspects of concrete dams and appurtenant

structures. This will also result in the

elimination of the duplication of any earthquake

engineering R&D which is being performed by

other organizations. This worldwide review of

the EERP's priorities assures the CE of having

a well-developed earthquake engineering R&D

program.

7. ACKNOWLEDGMENT

We gratefully acknowledge permission from the

Chief of Engineers to publish this paper.

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8. REFERENCES

1. Roper, W.E., Hynes, M.E., and Davidson,

R.F. (1992). "U.S. Army Corps of Engineers

Earthquake Engineering Research Program,

" Wind and Seismic Effects Proceedings of the

24th Joint Meeting. "

2. Hall, R.L., and Roper, W.E. (1993). "Corps

of Engineers Research Program on Concrete

Dams,

" Wind and Seismic Effects Proceedings

of the 25th Joint Meeting. "

3. CERF (1993). "Civil Works Structural

Engineering Seismic Research Study," Vol 1.

303
Table 1. Organizations and Research Areas of Interest

Organization

Research Areas of Interest

Concrete

Dams

A

Outlet

Structures B

Analysis

Tools

C

Public

Information

D

National Center for Earthquake

Engineering Research X X X

Bureau of Reclamation X X X

University of California, Berkeley X X X

Dr. Ray W. Clough, Structures Consultant X X X

Agbabian Associates X X

Quest Structures X X

Hohai University, China X X

Tsinghua University, China X

Public Works Research Institute, Japan X X X

Swiss Division of Safety of Dams X

University of Texas at Austin X

ANATECH Research Corp. X X X

Applied Technology Council X

Southeast Missouri State University X

National Science Foundation X X X

National Institute of Standards of

Technology X

Naval Civil Engineering Laboratory X

North Carolina State University X

Washington University X X X

Tokyo Electric Power Services, Japan X X

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Damage and Behavior of Dams During the 1993 Kushiro Oki Earthquake

and the 1993 Hokkaido Nansei Oki Earthquake

by

Akira Nakamura", Nario Yasuda2

', and Tomoya Iwashita3)

ABSTRACT

In 1993, two large earthquakes with a magnitude of 7.8 occurred near Hokkaido in Japan, severely damaging public works

facilities in the area. The 1993 Kushiro Oki

Earthquake damaged the upstream slope

protection and the ground slope of the

Mombetsu Dam slightly. The 1993 Hokkaido

Nansei Oki Earthquake caused moderate

damage to the Niwa-Ikumine Dam. The

liquefaction of the lower part of the dam body

and the riverbed deposit under the dam caused

settlement, cracking, and bulging of the

upstream slope of the dam body. Slight damage

was caused at the Pirika Dam and the

Makomanai Dam. At some other dams, the

both earthquakes increased the amount of

leakage or the turbidity of the leakage water. This paper begins by describing the

damage and effects on these dams. Next, with

the results an analysis of the acceleration

records observed at the dams during the

earthquakes, we present the relationship

between the maximum acceleration or predominant period and the epicentral distance

and the acceleration response amplification

factor at the crest for each dam type. The

results show that, for a certain distance, the

attenuation of the maximum rock acceleration

conforms closely to the attenuation formula for

rock mass proposed after Tamura and

Okamoto.

KEYWORDS: Dam; Earthquake Damage;

Observed Earthquake Motions;

The 1993 Hokkaido Nansei Oki Earthquake;

The 1993 Kushiro Oki Earthquake

1. INTRODUCTION

Hokkaido in Japan was attacked by two

earthquakes in 1993. The first was the 1993

Kushiro Oki Earthquake of January 15, and the

second was the 1993 Hokkaido Nansei Oki

Earthquake which occurred about 6 months

later on July 12. The magnitude of both

earthquakes was 7.8. The 5th or 6th degree on

the seismic intensity scale was observed close

to the epicenters of the two earthquakes by the

Japanese Meteorological Agency (J.M.A.).

They resulted in severe damage: human

casualties, the destruction of dwellings, and

damage to public works facilities. The total damage exceeded 50 billion yen and 160

billion yen respectively. Some dams were either damaged or suffered slight effects. This paper

describes the dams damaged and effected by

the two earthquakes and the behavior of the

dams during the earthquakes.

2. OUTLINE OF THE EARTHQUAKES

2.1 The 1993 Kushiro Oki Earthquake

The 1993 Kushiro Oki Earthquake was a magnitude 7.8 earthquake with its hypocenter

located at a focal depth of 107 kilometers and

over 10 kilometers offshore from the City of

Kushiro. The Kushiro Meteorological

1) Head, Fill-type Dam Division,

Public Works Research Institute, Ministry of

Construction, Tsukuba City, Japan

2) Head, Planning Division, ditto (Former Senior Research Engineer,

Fill-type Dam Division)

3) Research Engineer, Fill-type Dam Division,

ditto

305
Observation Station observed it as a violent

earthquake with the 6th degree on the J.M.A.

seismic intensity scale. A total of 968 persons

in Hokkaido suffered from this earthquake, and

damage to public works facilities amounted to approximately 21 billion yen.

2.2 The 1993 Hokkaido Nansei Oki

Earthquake

The 1993 Hokkaido Nansei Oki

Earthquake was a magnitude 7.8 earthquake

with its hypocenter at a focal depth of 34

kilometers and about 80 kilometers offshore to the south-west of Shiribeshi-Province

Suttsu-cho. Several meteorological observation

stations facing the Japan Sea near the epicenter

observed it as a strong earthquake with the 5th

degree on the J.M.A. seismic intensity scale. Giant tsunami induced by this earthquake

caused a total of 552 casualties including 202

fatalities. The damage to public works facilities amounted to about 79 billion yen.

3. DAMAGE AND EFFECTS ON DAMS

3.1 Damage and Effects on Dams Caused by

the 1993 Kushiro Oki Earthquake

Emergency inspections of the dams were

conducted immediately after the earthquake, but

because it was the coldest period of the winter, snow was piled up on the dam bodies and the

surrounding ground, so detailed visual

inspections were not possible. After the spring

thaw, staff from the dam control office

conducted an inspection. The inspection

revealed that only the Mombetsu Dam suffered

slight damage. At the Mombetsu Dam, part of

the concrete facing blocks forming the slope

protection on the upstream side of the dam

body were raised, but no deformation was

found which would have any effect on the

stability of the dam.

After the earthquake, the amount of

leakage water increased at three dams (Kanoko

Dam, Tokachi Dam, and Samani Dam), and the

turbidity of the leakage water increased at three dams (Samani Dam, Takami Dam, Niikappu

Dam), but after a period ranging from a few

hours to a few months, the leakage water at the dams returned to its pre-earthquake state. Fig.l

shows the change over time of the amount of

leakage water at the Kanoko Dam. Generally,

temporary changes in the amount of leakage

water at dams after earthquakes have been

observed in the past, but it is assumed that this does present particular problems as long as the

change is not extremely large and increased

leakage does not persist. Fig.2 shows the

locations of dams where the above damage and

effects occurred.

No dam with a height of 15 meters or more was located within about 100 kilometers

of the epicenter. Therefore, damage and effects

to dams was small considering the strength of

the earthquake. The following describes the

damage to the Mombetsu Dam in detail.

3.1.1 Mombetsu Dam

The Mombetsu dam is a homogeneous

earthfill dam managed by the Mombetsu Land

Improvement District. Fig.3 and 4 are the plane

diagram and a typical cross-section of the dam.

Concrete facing blocks have been placed on the

upstream slope of the dam body to protect the

dam from wave action. The geology of the dam

site consists of mudstone in a Kasai Formation,

the Takigawa Group of the Pliocene Series, the

Neogene System, and includes fine sandstone

and siltstone. The dam body consists of

material excavated from the dam foundation

and soil from the weathered mudstone at downstream ground of the left bank. The dam

is 180 kilometers from the epicenter.

After the earthquake occurred, the dam

staff conducted an inspection, but the reservoir

surface was covered with a thick frozen layer,

preventing the inspectors from determining the

full extent of the damage. In February, they

confirmed that there was water leaking from

the dam body drain. Soon after the spring thaw

in April, they conducted a second inspection

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after removing all the water from the reservoir.

This confirmed that the upstream slope of the

dam body, and the ground slope on the left bank were deformed.

The deformation of the dam body was

found at protective facing blocks which

overlapped one another and rose up about 10

centimeters along the entire dam axis directly

below the elevation where the gradient of the

upstream slope changed from 1:2.5 to 1: 2.0

(EL. 87.50 meters and with a reservoir water

level of EL 89.0 meters at the time of the

earthquake) (Photo. 1). It is believed the

earthquake caused the top 1/3 of the dam body

to settle, applying compressive force in the

slope direction on the facing blocks causing

them to buckle. When the inspectors removed

the deformed facing blocks to check the surface

of the earthfill dam, they concluded that there was no deformation in the body, and that the

safety of the dam was unaffected. Generally, in a case where the dam body of a fill dam is surfaced with material such as concrete blocks,

because their stiffness is different from that of

the dam body, this kind of facing cannot follow

the deformation of the dam body during an

earthquake, with the result that the facing is deformed.

Undulating deformation was seen on the

facing blocks close to the intake pipe on the

upstream concrete part of the left bank

(Photo.2)

3.2 Damage and Effects on Dams Caused by

the 1993 Hokkaido Nansei Oki Earthquake

Immediately after the earthquake, staff of dam offices performed emergency

inspections of a total 52 dams: 6 dams owned

by the Ministry of Construction (M.O.C.), 11

dams subsidized by the M.O.C., and 35 other

dams used for water utilization. The inspection

revealed that among dams with a height of 15

meters or more, the Niwa-Ikumine Dam

suffered moderate damage. Minor damage was

discovered at the Pirika Dam and the

Makomanai Dam, but it was not severe enough

to effect the safety of these dams.

Immediately after the earthquake,

increase in leakage water was observed at six dams (Pirika Dam, Jozankei Dam, Okusawa

Dam, Sasanagare Dam, Kamiiso Dam, Shimoyu

Dam), and the amount of leakage water

decreased at 2 dams (Asari Dam, Subari Dam).

Fig.5 shows the change over time of the

amount of leakage water at the Jozankei Dam.

The turbidity of leakage water increased at the

Subari Dam. However, these changes in leakage water were not considered to indicate

serious problems, because the leakage water

returned to its pre-earthquake state within a period ranging from a few days to a few

weeks. The surface displacement of the

Kamiiso Dam changed a little, while at other

dams (Pirika Dam, Shiriuchi Dam) the width of

the cross joints was slightly altered, but in all cases the displacement was small, and the

deformation does not seem to cause any safety

problems. Fig.6 shows the locations of dams

where the above damage and effects were

discovered.

Furthermore, two small irrigation ponds

with dikes less than 15 meters high located

only 65 and 70 kilometers from the epicenter in Setana-cho in Hokkaido were destroyed. And

along the Japan Sea Coast near the epicenter, some dikes of this kind were damaged by

settlement or cracking.

The following is a detailed description of

damage and effects suffered by the Pirika Dam,

the Niwa-Ikumine Dam, and the Makomanai

Dam.

3.2.1 Pirika Dam

The Pirika Dam, which is managed by

the Construction Department of the Hokkaido

Development Agency, is a combined dam

consisting of a concrete gravity dam and a

rockfill dam. Fig.7 and 8 are a section diagram

of block 39 in the concrete dam section and a downstream surface diagram of the Pirika Dam.

This dam is located 89 kilometers from the

epicenter.

307
As Fig.9 and 10 show, in the structure between block 0 of the concrete dam body and

the connecting corridor leading to the control

office, the floor panel concrete of the

connecting corridor was pushed up and spalled over a length of about 1.5 meters due to compression in the axial direction of the dam.

The spalling is shown in Photographs 3 and 3.4

and in Fig.ll. Slight concrete spalling was also

discovered in other cross joints of the

connecting corridor. This connecting corridor

rides on the rock zone of the core wall

connected perpendicularly to the concrete dam

body, so it was subjected to severe compression

caused by differences in the earthquake

response of the dam body and the rock zone. Because the connecting corridor and the concrete dam body are completely separated

structurally and there was no deformation of

the inspection gallery in block 0, the

deformation did not effect the dam body itself. Above the previously mentioned

connecting section, the surface paving of the

crest road settled, and a gap was formed on the

connection to the upstream railing (Photo.4).

And at a nearby location in the park plaza

above the connection between block 0 of the concrete dam and the left bank core wall, part

of the block surface was lifted (Photo.5). The

locations of these deformations are shown in Fig.9 and 10. All these deformations were

caused by unequal behavior of the concrete dam body and the backfilled soil above the

connecting corridor or the rock zone of the core wall at the left bank. And when the block

pavement was restored by removing mortar and

base material, a slight unevenness was revealed on surface of the core wall on the left bank. So

as a precaution, rolling compaction was

performed. This unevenness, however, is considered to be a result of initial execution

work instead of an effect of the earthquake.

After the earthquake, water leaked from

the intersection of the inspection gallery and

the blocked section of the temporary diversion

channel inside the dam on the block 28. This

section was sealed after leakage was observed

from it during the previous test filling, but up

till the time of the earthquake, no leakage was

observed at the sealing, so it is considered that the earthquake opened a small gap in the

sealing, causing another leakage. Fig. 12 shows

changes over time in the amount of leakage.

The amount has been small, only between 80

and 100 milliliters per minute since 10 days

after the earthquake, and it was not turbid, so

the leakage is not considered to be a serious

problem.

The water leakage from both foundations

and cross-joints of the two zones of the

divided concrete dam body (AW-1: Blocks 1 to 34, AW-2: Blocks 35 to 47) is collected and

that from the fill dam section is also collected

by a drain pipe installed at the end of the

downstream filter. Fig. 13 shows changes over

time in the amount of water leakage before and

after the earthquake. At the time of the

earthquake, the amount of water leakage from

the concrete dam declined only about 0.8 liters per minute and that from the fill dam body rose by 4.7 liters per minute, but the change was

small, and it returned to its normal level after a few days, so it is not a problem.

The width of the joint aperture in the

inspection gallery ranged from 0.85 to 0.6

millimeters before and after the earthquake

(June 29 - July 13). After the earthquake,

many of the joints closed and traces of the

ejection of rusty water were found at many

joints in the inspection gallery of the concrete dam.

3.2.2 Niwa-Ikumine Dam

The Niwa-Ikumine Dam is a central core type earthfill dam managed by the

Kita-Hiyama Land Improvement District. It is an old dam, completed 66 years ago in 1927,

and rehabilitated in 1934 and 1961. Fig.14 are

a typical section diagram of the dam. This dam

was not provided with a seismograph or water

leakage observation equipment. However,

considering that it is located 70 kilometers

from the epicenter, that a maximum

308
acceleration of 114gal was recorded in the

foundation bedrock of the Pirika Dam, which is 89 kilometers from the epicenter, and that the

foundation rock of the both dams belongs to the same Setana Formation, the acceleration at the Niwa-Ikumine Dam was likely higher. By

substituting a magnitude 7.8 and an epicentral

distance of 70 kilometers into the distance

attenuation formula of maximum acceleration

after Tamura and Okamoto, we obtain the

maximum acceleration 188gal, and it is estimated that the rock at the Niwa-Ikumine

Dam site quaked at an acceleration of about

200gal.

Photo. 6 shows the damage to the dam

body three days after the earthquake. And

Fig. 15 is a plane diagram of the dam body

showing an outline of the damage. The

earthquake caused bulging and cracking of the

dam body at two elevations, and undulating

deformation of the upstream slope. The bulging

appeared in the middle and close to the toe of

the upstream slope (Elevation EL.90.90 meters,

or about 50 centimeters below the water level

at the time of the earthquake, EL.91.35 meters).

The maximum level difference of the bulge at the toe of the slope was about 1 meter over the

40 meter length (Photo.7). As shown in Fig.15,

the cracking consisted of five longitudinal

cracks at the crest of the dam. The longest

crack extended for a total length of

approximately 120 meters on the crest at a depth of two meters. It settled on the upstream

side, and its maximum level difference was

about 1 meter (Photo.8). Fig.16 is a 10-meter

interval section diagram of the dam body from

where cracks formed at a point 80 meters from

the right bank spillway, which corresponds

almost exactly to the center of the dam. And a 60-centimeter deep crack was formed at an

elevation of about 2 meters above the bulge

near the toe of the upstream slope. Also,

several cracks were formed with a sand boil

extending from the sand stratum in a sedimentary mound at the bottom of the

reservoir near the toe of the upstream slope.

The day after the earthquake, the District

used an underwater pump to drain the water

(water depth at the time of the earthquake was

about 4 meters). It was completely drained by

the morning of July 21. Believing that there was a degree of partial loosening on the

upstream side of the dam body, they faced a

safety control problem: if a flood occurred, it would be difficult to release the water to prevent the reservoir water level from rising by

using only the inclined and the bottom intake

drain pipe (drain capacity of 2.5m /s). Therefore, on July 26, the District began an emergency excavation at the left side of the

dam body to open a channel with a bottom

width of 2 meters and a section area of 3

square meters, to provide discharge capacity of

10.9 cubic meters per second (About 5,500

cubic meters of the dam body was excavated).

And at the same time, they carried out emergency repair work on the dam body by

pouring lime into the cracks on the crest, and

roller compacting it with a heavy compactor.

This emergency work was completed on August 6. The water channel that they

excavated is shown in Photographs 9. It was

difficult to visually confirm the sliding surface on the excavated slope surface. As a result of later consultation with the

river authorities, the dam body embankment

would be removed, and new headworks would

be installed. At a point shown in Fig.15, boring

inspections and Swedish sounding tests carried

out. The dam body is an embankment on the

riverbed deposit between 4 and 5 meters thick,

and a Setana Formation composed of sandstone

with a low degree of consolidation is found

under that. The dam body consists of core zones of clay mixed with pebbles and

silt-and-sand-zones mixed with pebbles, and

core zones separate into two zones of center core and upstream slope core. Silt-and-sand-zone is extremely loose with N

value between 2 and 4 and compaction degree

of a 81%. The riverbed deposit consist of loose

sand or sand mixed with pebbles with N value

between 3 and 8 and gravel with N value about

309
20.

It is believed that the earthquake caused

the silty sand zone of the upstream side part of

the dam embankment to occur liquefaction and

the dam body to settle, because the zone

consists of loose sand and is saturated with

water owing to the location between two core

zones.

3.2.3 Makomanai Dam

The Makomanai Dam is a central core

type rockfill dam with a height of 34 meters

managed by the Agricultural and Fisheries

Department of the Hokkaido Development

Agency. It was completed in 1987, and located

66 kilometers from the epicenter. This dam was

not provided with seismographs.

The earthquake caused settlement of the

dam body and bulge of the lower part of the

body. So, the concrete frame of the upstream

slope had been deformed ,and the joints of it had been slightly broken. Several cracks had

been caused at the sprayed concrete in front of

the overflowing wall of the spillway. The

facing concrete blocks on the right bank had

undergone undulating deformation. None of

these damages presented any particular

problems related to the safety of the dam body.

4. EARTHQUAKE MOTIONS OBSERVED

AT DAM SITES

Records of acceleration in the dam

bodies and at the dam sites were obtained at 15

dams in Hokkaido and at 26 dams in the

Tohoku region during the 1993 Kushiro Oki

Earthquake, and at 10 dams in Hokkaido and

19 dams in the Tohoku region during the 1993

Hokkaido Nansei Oki Earthquake. The

locations of these dams are shown in Fig.2 and

6. The attenuation of the maximum

accelerations at dam site rock with epicentral

distance during the two earthquakes, are shown

in Fig.17 and 18 respectively. Where dam site rock includes the dam foundation rock or the

rock in the ground of both banks, and in the case of a fill dam, because an inspection

gallery is excavated in the foundation bedrock,

the filldam inspection gallery is included.

Within the figures, a broken line represents a magnitude 7.8 earthquake for the attenuation of

the maximum accelerations formula after Tamura and Okamoto(1979). This attenuation

formula was based on seismic records observed

at underground rock within an epicentral

distance of 280 kilometers, but records of the

1993 Kushiro Oki Earthquake provide values

that tend to be smaller than those of the

attenuation formula at dams less than 300

kilometers from the epicenter and larger at dams further than 300 kilometers from the

epicenter. In the case of the 1993 Hokkaido

Nansei Oki Earthquake, up to an epicentral

distance of about 400 kilometers, the

attenuation is extremely close to that provided

by the attenuation formula after Tamura and

Okamoto.

Fig.19 and 20 show the relationship

between the predominant period of the seismic

vibration in the dam site rock and the

epicentral distance for both earthquakes. And

within the figures, a dotted line shows the

assessable line of a predominant period after Seed et a/.(1968) with a magnitude of 7.8. The maximum acceleration at the dam

foundation (the foundation or the inspection

gallery) and the ratio of the maximum

acceleration at a crest to that at a foundation, in other words its relationship with the

acceleration response amplification ratio for both earthquakes are shown in Fig.2 1 for fill dams, and in Fig.22 for concrete gravity dams

and concrete arch dams. Within the figures, cases of other earthquakes are also shown. In the case of any earthquakes, the response

amplification ratio ranged from 2 to 8 for filldams and from 1 to 9 for concrete gravity

dams. The response amplification ratio was

larger in the case of concrete arch dams,

ranging from 6 to 13, and in some cases, as high as 18. Overall, the response amplification

ratio for the upstream-downstream component

310
tended to be larger than that for the dam axial

component.

5. EARTHQUAKE ACCELERATION

RECORDS OF TYPICAL DAMS

5.1 Acceleration Records During The 1993

Kushiro Oki Earthquake

The following is a report on the results of analyses of acceleration records from three

typical dams where the maximum acceleration

observed at the crest exceeded 60gal.

5.1.1 Kanoko Dam

The Kanoko Dam is a concrete gravity

dam with a height of 55.5 meters, located 116

kilometers from the epicenter. Seismographs were installed at three locations, a low

elevation in the dam body (Al), a middle

elevation in the dam body (A2), and the crest (A3). These locations are shown on Fig.23 and

24: a typical section diagram and a downstream

surface diagram of the dam. The geology of the

dam foundation rock consists of pyroxene

andesite in the Pliocene Series, the Neogene

System. The maximum accelerations recorded

in the up-downstream direction were 23.1gal at Al and 64.9gal at A3. The predominant

frequency of the power spectrum of the

acceleration was in the low frequency zone of

2Hz or less for Al, while for A3, the frequency was close to 6Hz, except for the dam axial

direction. Fig.25 shows the Fourier spectrum

ratio or frequency response function (The

Parzen Window with a band width of 0.4Hz was executed.) of the up-downstream direction

component of Al and A3. From this figure, the

fundamental natural frequency is 6.4Hz, a value

smaller than 8.2Hz of the natural frequency

theoretically obtained from a bending vibration

of a cantilever element under storing in the

reservoir.

5.1.2 Tokachi Dam

The Tokachi Dam is a rockfill dam with

a height of 84.3 meters, and located 125

kilometers from the epicenter. Seismographs were installed at four locations, in the

inspection gallery (Al), inside the core (A2), on the crest (A3), and in the left-bank ground

(A4). These locations are shown on Fig.26 and

27: a typical section diagram and a longitudinal

section diagram. The geology of the dam

foundation rock consists of slate, sandstone,

and diabase in the Upper Cretaceous System,

and while on the right bank, includes

keratophyre. The maximum accelerations

recorded in the axial direction of the dam were

34.9gal at Al and 120gal at A3. The power

spectrum of the up-downstream and dam axial

direction component accelerations at A3 peaked

sharply near 2Hz. Fig.28 shows the frequency

response function of the up-downstream

direction component. From this figure, the

fundamental natural frequency is 2.0Hz, and

this is in the range of 1.8 ~ 3.4Hz obtained

from the empirical formula of natural frequency

for rockfill dams as the result of earthquake

observations and vibration tests after Okamoto(1984).

5.1.3 Takami Dam

The Takami Dam is a rockfill dam with

a height of 120 meters, and located 149

kilometers from the epicenter. Seismographs were installed at four locations, in the

inspection gallery (Al), inside the core (A2), on the crest (A3), and at the toe of the slope of

the downstream left-bank (A4). These

locations are shown on Fig.29 and 30: a typical

section diagram and a longitudinal section

diagram. The geology of the dam foundation

rock consists of schalstein, diabase schist, and

chert in the Iwashimizu Formation of the

Hidaka Group. The maximum accelerations

recorded in the up-downstream direction were

38.5gal at Al and 230.6gal at A3. The power

spectrum of A3 includes almost no high

311
frequency constituent of 5Hz or higher, and it peaked sharply. Fig.31 shows the frequency

response function of the up-downstream

direction. From this figure, the fundamental

natural frequency is 1.9Hz, and this is in the

range of 1.3 ~ 2.4Hz obtained from the

empirical formula of natural frequency for

rockfill dams as the result of earthquake

observations and vibration tests after Okamoto(1984).

5.2 Acceleration Records During the 1993

Hokkaido Nansei Oki Earthquake

The following is a report on the results of analyses of acceleration records from three

typical dams where the maximum acceleration

observed at the crest exceeded 60gal.

5.2.1 Pirika Dam

The Pirika Dam is located closer to the

epicenter (89 kilometers) than any of the other

dams where the time history of the acceleration was recorded. This Dam is a combined dam

with a height of 40 meters. At the dam, 3

seismographs were installed: in the inspection

gallery, (Al), at the crest (A2) in block 39 of

the concrete dam body section, and at the

foundation rock under block 14 (A3). These

locations are shown in Fig.7 and 8: a section

diagram of block 39 in the concrete section and

a diagram of the downstream surface of the

dam. The maximum accelerations recorded in the up-downstream at Al and A2 were

122.8gal and 142.2gal respectively. The

principal motion began about 12 seconds after the initial motion, and continued for about 60

seconds. The maximum acceleration of the

horizontal direction component appeared about

43 seconds after the initial motion. The

predominant frequency of the power spectrum

of the horizontal direction component ranged

between 1.2 and 1.8Hz at all points. Fig.32

shows the frequency response function of A2 to Al of the up-downstream direction component.

In this figure, the frequency response function

is lined without a peak that indicates the natural frequency, within the range of 12Hz or

less. The response was not amplified at the

crest of the Pirika Dam because, not only is this dam low, the ratio of the elastic modulus

of the foundation rock to that of the dam body

concrete (Ed), Ef/Ed, is small at approximately

1/40, and as Fig.8 shows, the alternative beds

of mudstone and shale (Shm) that form the

foundation rock for a concrete gravity dam

have an elastic modulus (Ef) of 700MPa, and

as foundation rock for a concrete gravity dam,

is comparatively soft rock.

5.2.2 Asari Dam

The Asari Dam is a concrete gravity

dam with a height of 73.9 meters, and is located 155 kilometers from the epicenter.

Seismographs were installed at two locations: on the inspection gallery (Al), and on the crest (A2). These locations are shown on Fig.33: a diagram of the downstream surface of the dam.

The maximum accelerations recorded in the

up-downstream direction were 15.1gal at Al

and 68.9gal at A2. The predominant frequency

of the power spectrum of the acceleration was,

in the horizontal component of Al, in the low

frequency range of 1Hz or lower, but was 6Hz

in the up-downstream direction component of

A2. Fig.34 shows the frequency response

function in the up-downstream direction

component. From this figure, the fundamental

natural frequency is 6.5Hz, and this approximates the value of 6.2Hz theoretically

obtained from a bending vibration of a cantilever element under storing in the

reservoir.

5.2.3 Hoheikyo Dam

The Hoheikyo Dam is a concrete arch

dam with a height of 102.5 meters, and is located 160 kilometers from the epicenter.

Seismographs were installed at three locations:

in the lower inspection gallery (Al), in the middle inspection gallery (A2), and on the crest

312
(A3). These locations are shown on Fig.35 and

36: a typical section diagram and a diagram of

the downstream surface of the dam. The dam

foundation rock consists of andesite. The

maximum accelerations recorded in the

up-downstream direction were 9.5gal at Al

and 119.6gal at A3. The power spectrum of the

acceleration at Al peaked in the low frequency zone of 1Hz or lower, but at A3, it peaked

sharply at about 3.6Hz. Fig.37 shows the

frequency response function of the up-downstream direction component. From this figure, the fundamental natural frequency is 3.8Hz, and this approximates the value of

3.3Hz obtained from the empirical formula of

natural frequency for rockfill dams as the result of earthquake observations and vibration tests after Okamoto(1984). Additionally, the

acceleration response amplification ratio at the

natural frequency was large at approximately

40, the same as the 1993 Kushiro Oki

Earthquake. This indicates that a thin-walled

structure such as an arch dam considerably

amplifies the response.

6. CONCLUSIONS

The study confirms that, aside from old

dams and irrigation pond dikes constructed

using experience-based construction

technology, those dams constructed on rock

foundations using modern design and execution

technology survived these powerful earthquakes

without any adverse effects to maintain their

safety, and that they are still in sound

condition. The behavior of the dams found by

analyzing acceleration records coincided with

the findings of empirical formulae.

ACKNOWLEDGEMENTS

The authors wish to express sincere

thanks to all those persons from the Tohoku

Regional Construction Bureau, the Hokkaido

Development Bureau, the Tohoku Regional

Agricultural Administration Office, Hokkaido

Prefecture, Aomori Prefecture, Akita Prefecture, Iwate Prefecture, Miyagi Prefecture, various

dam operation offices, land improvement

district offices who provided them with

valuable materials and cooperated fully with

them during the conduct of this study.

REFERENCES

Okamoto,S. (1984): Introduction to Earthquake

Engineering, pp.403-523, University of

Tokyo Press. Seed, H.B., Idriss, I.M. and Kiefer, F.W.

(1968): Characteristics of Rock Motions

during Earthquakes, p.8, EERC 68-5,

U.C. Berkeley.

Tamura,C, Okamoto, S. and Kato, K. (1979):

Maximum Rock Accelerations of Earthquake

Motions , Proceedings of the 13th JSCE on

Earthquake Engineering Symposium, Japan

Society of Civil Engineers, pp. 181-184.

(In Japanese)

313
B 4

1993 KUSHIRO OKI EARTHQUAKE

KANOKO DAM

Water Level

10 n

1992

12

450

440

430

120

111)

KID

1993

2 3 4 5 (month) —Earthquake

Fig.l Change Over Time in Amount of Leakage at Kanoko Dam

No

12

34

567

8

910

1112

1314

1516

1718

19 20

2122

23

24

252627

2829

30

3132

33

34

3536

37

38

3940

4142

DAM NAME

Mombetsu Dam

Kanoko Dam

Tokachi Dam

Taisetsu Dam

Kanayama Dam

Iwaonai Dam

Katsurazawa Dam

Izarigawa Dam

Jozankei Dam

Hoheikyo Dam

Pirika Dam

Samani Dam

Takami Dam

Aibelsu Dam

Asari Dam

Niikappu Dam

Shijyushida Dam

Gosno Dam

Tase Dam

Yuda Dam

Ishibuchi Dam

Narugo Dam

Kamafusa Dam

Shichigasyuku Dam

Kawauchi Dam

Taki Dam

Tsunatori Dam

Irihata Dam

Urushizawa Dam

Minamikawa Dam

Nanakita Dam

Tarumizu Dam

Yamase Dam

Minase Dam

Takashiba Dam

Shitoki Dam

Shiramizugawa Dam

Odagawa Dam

Namioka Dam

Hayaseno Dam

Shimizume Dam

Wada Dam

Fig.2 Location of Damaged or Effected Dams and Dams where Acceleration Records Were Observed

from The 1993 Kushiro Oki Earthquake

314
Fig.3 Plane Diagram of Damage Conditions at Mombetsu Dam

MOMBETSU DAM

,6.00.,

Fig.4 Typical Section Diagram of Damage Conditions at Mombetsu Dam

315
43°

42°

41'

40'-

39'

139*

EPICENTER

' (M. 7. 8) X

-100km

-200km

—300km

—400km

139'

No

12

34

567

8

9

10

1112

13

14

151617

18

1920

2122

2324

252627

28

2930

3132

33

DAM NAME

Makomanai Dam

Niwa-Ikumine Dam

Okusawa Dam

Sasanagare Dam

Pirika Dam

Hoheikyo Dam

Jozankei Dam

Izarigawa Dam

Asan Dam

Aibetsu Dam

Samani Dam

Uzura Dam

Kamiiso Dam

Shiriuchi Dam

Aseishigawa Dam

Yuda Dam

Tase Dam

Shimoyu Dam

Hayakuchi Dam

Subari Dam

Yamase Dam

Moriyoshi Dam

Haginari Dam

Asahikawa Dam

Iwarn i Dam

Taki Dam

Tsunatori Dam

Irihata Dam

Odagawa Dam

Namioka Dam

Hayaseno Dam

Shimizume Dam

Wada Dam

140'

Fig.6 Location of Damaged or Effected Dams and Dams where Acceleration Records Were Observed

from The 1993 Hokkaido Nansei Oki Earthquake

140

120

100

I 80

60

40

20

2

1993 HOKKAIDO NANSEI OKI EARTHQUAKE

JOZANKEI DAM

_\£/ater Level

4 5 6 Earthquake

400

8 9 10 11 12 (month)

1993

PIRIKA DAM

Fig.5 Change Over Time in Amount of Leakage at Jozankei Dam

Fig.7 Section Diagram of Block 39 of the Parika

Dam Concrete Body

316
Concrete Dam Section 755.000 PIRIKA DAM

Overflow Section 222.000 © © ©©

i

i i 1 Tixi ThT

» © © © © ©EL. 126.50 J I I I I I

Shm

EL.91.50

~_ Sill A3

• Seismograph

GEO]I 0 Q Y

The Setana Formatioi1 The Kuromalsunai ~ Yakumo Formation

Fine-grained Sandstone Ssf Siltstone Sil 2 Medium-grained

Sandstone Ssm

Alternative beds of

Siltstone. Mudstone

and Shale

Sil 1

Coarse-grained Sandstone Ssc Alternative beds of Mudstone and Shale Shm

Fig.8 Downstream Section Diagram of the Parika Dam Concrete Body

PIRIKA DAM

Block 0 of Ihe Concrete Dam Body

PIRIKA DAM —Connecting Corridor

Settlement of the Surface Paving and Gap to the Railing at the Crest

Core Wall Unit :m

Fig.9 Plane Diagram of Deformed Locations Between

Block 0 and the Connecting Corridor at Parika Dam

Fig.ll Spalled Cracks of the Floor Concrete in the Connecting Corridor (Parika Dam)

Settlement of the Surface Paving and Gap to the Railing

Backfilled Soil Zone

Connection Corridor —»-For the Control ~ office Rock Zone

Spalling in the Floor Concrete Panel

Fig. 10 Longitudinal Section of Deformed Locations

Between Block 0 and the Connecting Corridor at Parika Dam

PIRIKA DAM

Flow Conctoe Pind

Upnnam 9

°°5 DownJireim

View of (he Connecting Corridor from Block 0

Spilling

Upstream

Connecting Corridor

Block 0 of the Concrete Dam Body

Plane Unit on

317
180 , 1993 HOKKAIDO NANSEI OKI EARTHQLJAKF

1 60

90

2 3 (month) 9 10 11 12 Earthquake

Fig. 12 Change Over Time in the Amount of

Leakage in the Inspection Gallery at Block 28 of

Parika Dam

BO

40

20

0

PIRIKA DAM

-

% . Water Level " -

[Fill Body

1 \ \

1 \\M -

AW-2 /

'Vv

~~"aw1 L_

,.-< v . iConcrete Bodvi: J. 5 5 !!;?

'—, 1 L L J-. 1

120 —

j

110 ~

u

a

100 I

70

4 5 6 7 * 8 9 10 11 12(.onth) Earthquake —I 1993 Fig. 13 Change Over Time in Amount of Leakage at Pirika Dam

NIWA-IKUMINE DAM

, 4.00 EL. 100.65 D.W.L. EL. 98.60 N.W.L. EL.98.00

Slope Protection

Fig. 14 Typical Section Diagram of Niwa-Ikumine Dam

NIWA-IKUMINE DAM

_ / ®VjtaS-6

1 \ i \ 1

Na B Boring © NaS Sounding

Upstream Fig. 15 Plane Diagram of Damaged Niwa-Ikumine Dam Body

318
NIWA-IKUMINE DAM f-ijs.

Upstream

120

Section after the earthquake

Section at the time of completion

Fig.16 Cross Section Diagrams of the Cracked Niwa-Ikumine Dam Body

200 1993 KUSHIRO OKI EARTHQUAKE

2 h

1

<0a 100

uo

50 •i

O

GO : 4 go 20 a

CO Qk

c

10 a

o - m

CO •

CD

CD 5

formula after Okamoto

and Tamura (M. 7. 8)

Hor i2ontal Resul tant P

Up-Do«nstream •

Dam Axis ii

Vertical

_ _ _

100 200 300 400 500 600 700 800 900

Epicentral Distance (km)

200

100

50

20

10

5

1993 HOKKAIDO NANSEI OKI EARTHQUAKE

\ 4

8V

a

Formula after Okemoto

and Tamura (t 7. 8)

Up -Downstream •

Dam Axis

Vertical

"-••SI

\ A

i

•

200 300 400

Epicentral Distance (km)

500

Fig. 17 Attenuation of the Maximum Dam Site Rock

Acceleration with Epicentral Distance from The

1993 Kushiro Oki Earthquake

Fig. 18 Attenuation of the Maximum Dam Site Rock

Acceleration with Epicentral Distance from The

1993 Hokkaido Nansei Oki Earthquake

319
1993 KUSHIRO OKI EARTHQUAKE 1993 HOKKAIDO NANSEI OKI EARTHQUAKE

l 5

-

0 5

Ub-Dcnms*

'

Dan Aiit

Br Seed e <«7 f

I*

A*

am&

© Q

4i a

ir &

6

A

1 00 200 300 400 500 600

Epicentral Distance (km) Fig. 19 Relationship Between the Predominant Period

of Earthquake Motion at Dam Site Rock and

Epicentral Distance from The 1993 Kushiro Oki

Earthquake

up-Dorrnstrean o

HOKKAIDO NANSEI OKI

Ud -Downstream A

Other Earthquakes

Up-Downstreara

KUSHIRO OKI Dam Ans

HOKKAIDO NANSEI OKI

Dam Axis A

Othar Earthquakes

Dam Axis

1 3 10 30 100 300 Maximum acceleration at dam foundation (gal) Fig.21 Relationship Between the Maximum

Acceleration at the Dam Foundation and the Acceleration Response Amplification Ratio at the Crest (Fill Dam)

2 5

" 2

% 1-5

0 5

Up-Downstream • A

Dam Ai i

s

A

Vertical * By Seed et a

(M.7.8) •

T

•" A

i :> <

£ 1,1. 1

0 100 200 300 400

Epicentral Distance (km) Fig.20 Relationship Between the Predominant Period

of Earthquake Motion at Dam Site Rock and

Epicentral Distance from The 1993 Hokkaido Nansei

Oki Earthquake

CONCRETE GRAVITY DAfvf AND ARCH DAM

O

4 A A '

hi A A A %

KUSHIRO OKI O

HOKKAIDO NANSEI OKI A

Other Earthquake:

HOKKAIDO NANSEI OKI

Other Earthquake:

1 3 10 30 100 Maximum acceleration at dam foundation (gal) Fig.22 Relationship Between the Maximum

Acceleration at the Dam Foundation and the Acceleration Response Amplification Ratio at the Crest (Concrete Gravity Dam and Arch Dam)

KANOKO DAM

Fig.24 Dam Downstream Surface Diagram of Kanoko Dam

320
Fig.23 Typical Section Diagram of Kanoko Dam Fig.25 Frequency Response Function at Kanoko

Dam

S.W.L.EL.340,00 v

TOKACHI DAM

EL. 343.00

12.00 —

H

,6.00 hi • Seismograph

Fig.26 Typical Section Diagram of Tokachi Dam

443.00 TOKACHI DAM

Seismograph

Fig.27 Longitudinal Section Diagram of Tokachi Dam

321
25 1993 KUSHIRO OKI EARTHQUAKE

TOKACHI DAM

1993 KUSHIRO OKI EARTHQUAKE

TAKAMI DAM

6 8 10 12 14 16 18 20 Frequency ( Hz

)

Fig.28 Frequency Response Function at Tokachi

Dam

4 6 8 10 12 14 16 18 20 Frequency (Hz)

Fig.31 Frequency Response Function at Takami

Dam

,H.WL 221.000

TAKAMI DAM

,D.WL.226.300LjSWL.2 ,SWL 225.000

'NW.L. 218.000

11000 M EL 2 # Seismograph

Sediment Level £L 160. 000

Fig.29 Typical Section Diagram of Takami Dam

TAKAMI DAM A TM nn 25.728

A3•- EL 229.00

18.00 V

7 EL 206.239 •

EL 195.50 A2

EL. 109.00 ^

y^EL 106.70

A1 "A EL. 104.00

Fig.30 Longitudinal Section Diagram of Takami Dam

in-.

O 8-

Rat

ion 6-

ificat 4-

Ampl 2-

1993 HOKKAIDO NANSEI OKI EARTHQUAKE

PIRIKA DAM

25

.9 20 I

J

15

o§

io "Se< 5

0

1993 HOKKAIDO NANSEI OKI EARTHQUAKE

ASARI DAM

I)

Frequency [Hz]

Fig.32 Frequency Response Function at Pirika Dam

Frequency (Hz] ^

Fig.34 Frequency Response Function at Asari Dam

322
ASARI DAM 390.00 ®® @©©©®®®©®©©®©©®®®®®®®®®®

A2 EL. 207.900

Fig.33 Dam Downstream Surface Diagram of Asari Dam

HOHEIKYO DAM 305.00

K ® ® © © © © © © © © @ ® © © © © EL. 477.50

# Seismograph

Fig.36 Dam Downstream Surface Diagram of Hoheikyo Dam

HOHEIKYO DAM

EL. 4 77. 50

1993 HOKKAIDO NANSEI OKI EARTHQUAKE

HOHEIKYO DAM

Fig.35 Typical Section Diagram of Hoheikyo Dam

Frequency (Hz]

Fig.37 Frequency Response Function at Hoheikyo

Dam

323
Photo. 1 Raised Facing Concrete Blocks on the Upstream Slope Protection of Mombetsu Dam

Photo.4 Settlement of Asphalt Paving and Gap of

the Railing at the Crest around the Connection with

Block 0 at Pirika Dam

Photo. 2 Undulating Deformation of Facing Concrete Blocks on the Upstream Ground Slope of the Mombetsu Dam Left Bank

Photo. 3 Floor Concrete Spalled Cracks in the Structure between Block 0 and the Connecting

Corridor at Pirika Dam (Photograph in the Block 0

Direction from the Connecting Corridor)

Photo.5 Rising of the Surface Blocks in the Park

Plaza above the Connection Between Block 0 and

the Core Wall at Pirika Dam

324
Photo. 6 Damage to Niwa-Ikumine Dam

(Photographed from the spillway on the left bank on

July 15, By the Courtesy of the Kita-Hiyama Land

Improvement District)

Photo. 7 Bulging on the Upstream Slope of the Photo.8 Level Difference on the Crest Caused by the Niwa-Ikumine Dam (Photographed on July 22) Settlement of the Upstream Side of the Niwa-Ikumine Dam (Photographed on July 15, By

the Courtesy of the Kita-Hiyama Land Improvement

District)

Photo. 9 Channel Excavated as Emergency Work on the Niwa-Ikumine Dam (Photographed from the Reservoir Side on August 7, and by the Courtesy of the Kita-Hiyama Land Improvement District)

325

STORM SURGE

AND

TSUNAMIS

Evaluation of Seismic and Flood Hazards Potential

for a Large Earth-Filled Dam

by

N. Simon, M. Reich, P. C. Wang, R. Khanbilvardi* and T. McSpadden**

ABSTRACT

The impacts of seismic and flood events on th? PAR Pond Dam were evaluated by Brookhaven National

Laboratory (BNL) for the Department of Energy.

The seismic evaluation involved several steps, each

of which addressed a specific aspect of dam

stability. Results of these various steps are

linked and assessments of stability are made. Specifically, the analysis includes: (a) the determination of the basic physical parameters

(density, shear modulus, etc.) from available or extrapolated field data (such as standard penetration tests results and soil permeability),

(b) the evaluation of the steady state stress conditions corresponding to soil dead weight and reservoir water, (c) the calculation of the seepage

forces and consequently the initial effective

stresses, (d) the evaluation of the potential for

slope sliding failure for static conditions and for static conditions amplified to simulate a seismic

input using a simplified method to estimate slope

stability, (e) the calculation of the dynamic

response of the dam and its foundation to earthquake excitations and, (f) the evaluation of the failure potential based on the calculated

dynamic and steady-state results.

The flood evaluation used deterministic and probabalistic models together with stochastic and actual rainfall data to investigate the probability

of a flood overtopping the dam. The Probable

Maximum Flood (PMF) was determined using rainfall

data consistent with the Probable Maximum

Precipitation (PMP). This is compared to a flood frequency analysis that uses probabilistic methods

and actual rainfall data for the Savannah River

site and region.

KEYWORDS: dams; earthquake; flood; liquefaction;

probabilistic; rainfall.

1. INTRODUCTION

The PAR Pond Dam is an earthen dam constructed in 1958-1959 across the Lower Three Runs Creek located

on the U.S. Department of Energy's (DOE) Savannah

River Site in South Carolina. The dam has a 30 foot-wide crest at a design elevation of about 211 feet above mean sea level (MSL). It has a height

of 66 feet and a length of 4500 feet. At the spillway elevation (200.00 feet MSL) the reservoir

has a surface area of 2700 acres and an impoundment

of approximately 54000 acre-feet.

The embankment cross section is essentially

homogeneous consisting of si Ity and clayey sands

with relatively low permeability, flanked on the downstream side by a pervious zone of si Ity sand

with a width about one-fourth of the total embankment width. The foundation is relatively

pervious. A reinforced concrete outlet conduit,

with internal sectional dimensions of 8 feet by 8 feet, was constructed at the base of the dam near

its maximum section to serve as a sluiceway and

spi llway.

In 1991 a depression in the vicinity of the outlet

conduit was discovered, together with flow from beneath the conduit and associated downstream wet areas. Because of the adverse conditions, DOE contractors performed comprehensive geotechnical

and hydraulic investigations in 1991 and 1992. A series of soil tests conducted for both the embankment and the foundation layers indicated that

a layer of loose sand exists deep in the foundation

of the dam which had not been accounted for in the original design of the dam.

Further studies to determine the capacity of the dam to withstand flooding and earthquake events were performed in 1992 and early 1993. This paper

is a summary of the seismic and flood hazards

safety evaluations performed by BNL for DOE.

2. SEISMIC EVALUATION

Due to the complexity of the physical problem and uncertainties in the true values of key parameters

(such as soil damping, soil relative density,

etc.), assessments of the failure potential of the dam were made through parametric evaluations to provide estimates of the susceptibility to failure

due to either liquefaction or slope sliding.

The dynamic response is evaluated by treating the saturated soil as a two-phase medium. This

approach involves the determination of both the intergranular stresses and the pore pressures in the medium for the entire duration of the earthquake. Even though the formulation is a linear description of the constitutive response of the soil-water medium, important conclusions can drawn concerning the dam stability.

The failure potential of the dam for a Charlestonlike earthquake with a peak ground acceleration of 0.11g, considered to be representative of the site,

is estimated. The analysis indicated that, (1) for competent soil property assumptions the safety

factor against slope failure remains above 1.0 and,

(2) for the same dynamic event and under extremely

conservative assumptions of soil damping and

strength, liquefaction failure can be triggered at the bottom of the weak layer.

In further analyses the benefits of a berm placed

downstream of the dam were evaluated. The presence

of the berm reduced the potential for liquefaction

and shifted the critical zone downstream. Additional studies to assess the behavior of the weak sand under the increased effective stress field produced by the berm must be performed to confirm the positive effect.

Lastly, based on analysis results, stronger ground

motions combined with the conservative estimates of the key soil properties, reduce the safety factors

for both modes of failure below tolerable limits.

2.1 Technical Analysis

The analysis consists of a series of steps each of which addresses a specific aspect of dam integrity.

* Brookhaven National Laboratory, Upton, NY 11973

** US Department of Energy, Washington, DC 20585

329
Specif ically, the following important parameters

were assessed:

• Seepage forces generated by the flow of water through the dam and its foundation.

Present soil conditions within the embankment that dictate both the amount of seepage as well as the inherent strength of the structure.

• Driving forces along potential failure

surfaces that determine the level of safety

against slope sliding and failure during an earthquake.

• The influence of critical soil parameters

and their impact on the dynamic response

and structural integrity of the dam.

• The profile and the condition of the soil constituting the foundation of the dam and

its potential to liquify.

This complete matrix of critical system parameters

is addressed with emphasis given to the coupling between them. Subsequently, on the basis of the computed dam response the potential for failure is estimated. The slope failure potential in the embankment is assessed in terms of a safety factor

against sliding along a failure surface while the liquefaction potential is measured against the ability of the soil in the embankment and the foundation to withstand an anticipated level of shear stress during a seismic event. It should be emphasized that while basic procedures exist in assessing such failure potential, they fail to (a) represent the soil as a two-phase medium, and (b) implement seismic inputs in the form of acceleration time histories.

Initial conservative estimates of safety are obtained by utilizing the slope stability program

REAME. This program is based on the quasi-static

analysis of slopes and treats the soil as a singlephase medium. Subsequent analyses are performed

using the POROSLAM 2-0 finite element program which

allows two-phase soil representation and accepts a time history variation of the ground acceleration.

The stability of the dam and its susceptibility to failure was evaluated taking proper consideration

of the important coupling exists between the various steps. It is important to note that results from one phase of the evaluation provide

important information for the next phase. A profile of these analysis steps in their proper

order is provided below:

2.1.1 Screening

Screening of the susceptibility of the dam to slope

failure is first performed using the REAME (Rotational Equilibrium Analysis of Multi layered

Embankments) program. The code provides

conservative first estimates of the safety margins

regarding slope stability of the dam. Important

parameters that impact on the safety criteria are

the soil frictional angle and cohesion. Safety

factors against sliding along assumed failure

surfaces through the embankment and the foundation

layers are evaluated for a range of soil parameters

that seem reasonable. The analysis considers both

the steady-state (static) and a pseudo-static (to account for seismic effects through a seismic

coefficient) loadings. Based on the results, the

impact of each parameter is assessed.

2.1.2 Seepage Forces

Seepage forces are evaluated by considering the

flow of water through the dam. The steady state pore water pressure field is evaluated through a finite element seepage analysis using the ANSYS general purpose program. The pore pressure field

is an important input in the determination of the steady state stress distribution. The pore water

supports a fraction of the static stresses reducing

the intergranular soil stresses. The analysis also

provides the location of the phreatic surface.

2.1.3 Steady state stress

The steady state stress condition in the embankment

and the foundation layers is evaluated. This in- situ stress field, resulting from the action of the reservoir water and the soil overburden, is key to

the definition of stability of the dam. With

elastic soil properties extrapolated from test data

for the site and using finite element analysis, the stress field in the embankment and the foundation

caused by the reservoir load and the overburden is evaluated. The effective stress field that exists

prior to a dynamic event is evaluated by

incorporating the calculated pore pressure field. Effective stress is the difference between the soil overburden stress and the pore water pressure. Through the same analysis the steady state shear stresses are also evaluated. Each of these stress calculations is performed using the ANSYS code.

2.1.4- Dynamic response

The dynamic response of the embankment/foundation

system to earthquake loadings is evaluated next. This seismic analysis provides the stress time

histories throughout the structure caused by the design earthquake. This stress history combined

with the steady state effective stress field become

the basis of the safety assessment for both slope

stability and liquefaction. The design seismic

input is a Charleston- 1 ike earthquake with a ZPA of 0.11g. The dynamic response is evaluated using the POROSLAM code which is a linear analysis program

that treats the soil as a two-phase medium.

2.2 Failure Potential Evaluation

Based on the in-situ stress conditions and its dynamic response, the margins of safety against

failure in the following modes are evaluated to assess the capacity of the dame to withstand the design earthquake.

2.2.1 Slope Stability

Along various surfaces where sliding can occur, the dynamic factors of safety are computed and the most critical or potential failure surfaces are

identified. This process is performed using

extrapolated values of soil cohesion and frictional

angle. Since there are uncertainties in the exact values of these two parameters, the safety factors are calculated considering a parametric variation

of them.

330
2.2.2 Liquefaction

This mode of failure occurs when the embankment and

the foundation Liquify during a dynamic event. The driving force in the Liquefaction process is the dynamic (cycLic) shear stress generated in the soi

L

Layers. The resisting force is the shearing

strength of the soil. This strength can be addressed by bore hoLe testing and correLation with

Laboratory test results. The dynamic shear stress

is provided by the POROSLAM analysis while the standard penetration test results along with

predicted initial (steady state) effective stress provide the shear strength of the entire cross section of the embankment and the foundation

Layers.

2.2.3 Slope Failure/Liquefaction Connection

The possibility that a Localized liquefaction

condition can trigger a slope failure is also

examined. It is possible that in the Lower foundation Layers where dynamic shear stresses are

at a maximum, portions of the weak Layer will

Liquify. Localized Liquefaction does not necessarily mean loss of strength in the entire

Layer, but it could mean a reduction of the safety

factor over a potential failure surface transversing the liquified spot. Such a reduction

could precipitate sliding failure. In order to evaluate the coupling of the two failure modes, the safety factor against slope failure is re-evaluated

taking into account the change in cohesion and frictional angle that occurs within the Liquified

zone.

2.3 Results and Conclusions

2.3.1 Significant Seismic Results

Slope failure and Liquefaction were identified as the mechanisms that could lead to failure during a seismic event. The potential for failure in these

two modes taken separately and coupled was evaluated to assess the safety of the dam. The

results and the conclusions from the analysis are summarized below:

• The strength of the in-situ soil from which

the susceptibility of the section is judged

is related to the friction angle <J> ar|d cohesion c. The values of these soil

parameters are estimated from the available

penetration resistance test results performed at the site.

• Screening analyses that are based on the stability program (REAME, based on pseudostatic assumption), with the assumption of

a uniformly weak soil layer (critical zone)

in the lower foundation of the dam,

indicated that there was a high potential

for slope failure even under the action of the existing static loads. Since the dam

is able to support its static loads in spite of the apparent weak zone, the REAME

Code results are interpreted as very

conservative. It is more probable that the assumption that the weak zone is uniformly

weak is incorrect.

• A more sophisticated finite element analysis indicated that unless the critical

layer is uniform and very weak (very low frictional angle $), the dam is able to withstand an earthquake of intensity of 0.11g without exhibiting slope failure. An earthquake of this level is considered

appropriate for the Savannah River site. Further, based on the limited measurements

taken at the site the critical layer is not uniformly weak, and hence, the potential

for slope failure is even less. Assuming

the soil strengths chosen for the analysis

are correct, the 0.11g mentioned above is considered an appropriate assessment of the dam capacity.

• With conservative assumptions of soil properties, (i.e., assumed low values of hysteretic soil damping), the potential for liquefaction in the critical zone is high. The liquefaction, however, is expected to occur at the bottom of the critical zone directly under the downstream toe. This

zone is susceptible to liquefaction because

of its relatively high induced shear stresses and smaller confining stresses. However, since the location of the

liquefiable soil is about 100 feet below

the ground surface, is old geologically,

and is a small zone, it should not lead to liquefaction induced failure of the dam.

• Localized liquefaction alone does not lead

to failure. Re-evaluation accounting for slope failure coupled with localized

liquefaction effects was made. Even for this case, the safety factor was found not to fall below acceptable levels. This

result is based on a model incorporating

stiff and weak subzones in the critical

layer. Use of weak and stiff subzone

modeling is consistent with the standard

penetration resistance test results and is therefore more indicative of actual site

conditions.

• With a value of 2% for the hysteretic soil

damping assigned to the material in the critical zone (instead of the very conservative value of 0.5% used in the analysis), the susceptibility for liquefaction is eliminated for an earthquake with a peak level of 0.11g. Based on the computed effective strains of 0.13% obtained for the 0.5% damping case,

it is anticipated that the actual hysteretic damping for the soils will be significantly higher than 2%.

• Results show that when the most conservative assumptions for soil properties are used with the most conservative evaluation procedures,

liquefaction is possible under the upstream

pool. However, since the confining pressure is higher upstream, the safety

factor remains above 1.0 and liquefaction

induced failure of the dam is not indicated.

• An analysis that incorporates both vertical

and horizontal earthquake inputs,

correlated in time, indicates significant

increases in the embankment shear stresses

331
which reduce the safety factor against

slope sliding, but does not drive it below

1.0. Moreover, this combined earthquake

activity does not amplify the liquefaction

potential in the critical zone. It should

be noted that horizontal and vertical

earthquake inputs correlated in time is not considered to be realistic. It will

usually be uncorrelated in time. This case

can, however, be considered as an upper

bound load case.

• when a seismic input of 0.2g is considered

(instead of the 0.11g), the potential for both slope failure and liquefaction

increase to unsafe levels.

2.3.2 Remedial Benefits of a Berm

The analyses have shown the structure to have sufficient structural integrity to survive a credible seismic event. However, that conclusion

is strongly dependent on the soil properties

assigned to the critical zone. If the critical

zone is assigned the most conservative (poorest)

definition for soil properties, the dam seismic

integrity was found to be marginal. Given the uncertainties in the properties of the soil in the critical zone, a limited study of the placement of

a berm on the downstream surface of the dam, as a possible remedial action, was undertaken.

A berm improves the seismic integrity of the dam by

increasing the effective static stresses in the dam/foundation system. The berm induces greater

static effective stresses which reduce the likelihood of liquefaction, shifts the zone susceptible to liquefaction downstream, and increases slope stability. As berm size increases,

the susceptible zone shifts further downstream.

The predicted benefits of the berm are based on the assumption that the soil can support the additional

load without excessive consolidation. If consolidation is great, the dam foundation system

may not exhibit an effective stress increase in the critical locations and the benefits of the berm

will not be realized. Needless to say, if the placement of a berm is seriously considered the competency of the soil to support its additional

load must first be determined.

2.3.3 Results Requiring Further Definition and/or

Clarification

Although the P0R0SLAM code analysis is substantially advanced from those of the REAME Code

and other analytical techniques used in the

industry, there are still areas of the safety

evaluation, especially those pertaining to the development of a High Confidence of Low Probability

of Failure, that need further definition and/or

clarification as listed below:

• As has been stressed, the seismic integrity

of the dam is strongly dependent on the properties of the soil and the extent of the weak or critical foundation layer.

Recently, extensive additional soil testing

of the dam/foundation system were

performed. This information should be made available for review and correlation.

Using this new material information a comprehensive properties matrix for the dam/foundation system will be developed.

A detailed finite element model will be created consistent with the soil properties

matrix to correctly characterize the dam/foundation system.

• Following an accurate definition of soil

properties another important element affecting dam seismic integrity is the

exact character of the site seismic

excitation. To correctly account for this,

develop a design earthquake incorporating

site specific properties appropriate to PAR

Pond. Use this as input for the new finite

element model and perform dynamic analyses

considering simultaneous horizontal and vertical inputs. Assess dam/foundation

safety margins. Investigate the sensitivity of the safety margins to different families of earthquake inputs and evaluate the sensitivity of results to the

rigid base assumptions.

• Although the onset of failure is adequately

predicted with linear analysis methods,

only the application of non- linear methods

will allow a realistic simulation of dam response in a failure mode and an assessment of the corresponding

consequences of the failure. In order to address this deficiency, incorporate the capability to account for non-linear soil behavior into the P0R0SLAM code. Use the updated code to obtain more realistic

estimates of dam response, including pore pressure buildup and possible large

deformations in the embankment, that would

result from liquefaction and slope sliding.

3. HYDRAULIC AND HYDR0L0GIC EVALUATION

To investigate the probability of a flood overtopping the dam, BNL conducted a hydrological

and hydraulic study using deterministic and stochastic models. The study was first carried out using rainfall data consistent with the Probable

Maximum Precipitation (PMP). Using standard

procedures the 72-hour PMP, a "catastrophic" storm, was found to be 47.8 inches for the Savannah River

Site. This is the maximum amount of precipitation

expected in any period of 72 hours, based on a symmetrical "center-peaking" distribution for the precipitation intensity during the period of the

storm. The PMP distribution used throughout the

study was developed for a representative watershed

area of 31.7 square miles. The resulting Probable

Maximum Flood (PMF) was computed using the PMP as

input to the HEC-1 program.

The HEC-1 estimate of the PMF corresponding to the

PMP indicated that the level of water would be 3.56

feet above the top of the dam. The concept of the

PMP and the resulting PMF are often considered to

be rather low probability and arbitrary criteria

that cannot be substantiated for some applications.

The PAR Pond Dam is considered to be one such

appli cation.

Historical rainfall data have been collected at the Savannah River Site (11 rain gages) and in the regional area of the site (two raingage stations).

332
These data sets are available for periods of 30 to

46 years. This rainfall data is clearly helpful in the rainfall analyses for this site. A flood frequency analysis based on these 13 raingage

stations collecting data for 30 to 46 years is more

useful than the PMP value obtained from the National Weather Service's HMR-51 and HMR-52. The

PMP is a concept which is developed to simulate the ultimate "catastrophic" precipitation event that could ever be created under the most unfavorable

combination of the specific meteorological

conditions of a region involving parameters such as temperature, wind, and humidity. Although the

impact of the 72-hour PMP was evaluated in this

study, BNL contends that a 72-hour rainfall event

with a probability of occurrence of 10 developed

from actual site/region rainfall data is more

meaningful and appropriate. The 13 raingage

stations with 30 to 46 years of data produce a very

specific rain distribution for the Savannah River

Site. Applying probabilistic methods and using

this data, the 72-hour 100,000-year storm was found

to be approximately 1/2 of the PMP. DOE Standard

1020 permits use of a 100,000 year flood to evaluate structures with the potential for high

consequences.

The HEC-1 evaluation for this 72-hour rainfall

estimate indicated that the dam would not be overtopped by a 100,000-year storm. That is, a rainfall event with a probability of occurrence of 10" would not cause the PAR Pond Dam to be overtopped. However, this result is based on selected values of parameters related to the watershed characteristics for the site. The

importance of accurately estimating the Curve

Number, soil losses from sub-basins, and sedimentation potentials in the reservoir behind

the dam are discussed in the BNL report.

4. REFERENCES

4.1 Simos, N., and Reich, M., (1994). "Seismic

Analysis of the PAR Pond Dam." Brookhaven

National Laboratory, Upton, New York.

4.2 Reich, M., et. al. (1993). "Hydraulic and Hydro logic Evaluation of PAR Pond Dam." BNL-49681, Brookhaven National Laboratory,

Upton, New York.

4.3 Natural Phenomena Hazards Design and Evaluation Criteria , DOE-STD-1020.

333

Tsunami Runup Distribution Generated by

the July 12, 1993, Hokkaido-Nansei-Oki Earthquake

by

E. N. Bernard and F. L Gonzalez*

ABSTRACT

The Hokkaido-Nansei-Oki earthquake produced

one of the largest tsunamis in Japan's history.

Tsunami vertical runup measurements varied

between 15 and 30 m over a 20-km portion of

the southern part of Okushiri Island with several

10-m values on the northern portion of the

Island. Along the west coast of Hokkaido, no

survey values exceeded 10 m, but damage was

extensive at several coastal towns. Given the

sudden onset of the tsunami and its high energy,

it is amazing that more people were not killed. We also observed that the seawall at Aonae was

effective in reducing runup amplitudes.

KEY WORDS: Sea of Japan; TSUNAMI

RUNUP; TSUNAMI SURVEY

1. INTRODUCTION AND BACKGROUND

On July 12, 1993, at 2217 local time (1317

UTC), an Ms -7.8 earthquake rocked the west

coast of Hokkaido and the small, offshore island

of Okushiri in the Sea of Japan (Figure 1). A

major tsunami was generated and, within 2 to 5 min, extremely large waves engulfed the

Okushiri coastline and the central west coast of

Hokkaido. As of July 21, 185 fatalities were

confirmed, with 120 attributed to the tsunami;

this death toll is expected to rise, as missing

persons are included among the fatalities (Hokkaido Police Headquarters). Property losses

have been estimated at $600 million, due

principally to tsunami damage.

The Japanese immediately dispatched damage

assessment and survey teams. Most of these

Japanese teams were mobilized and began

surveying tsunami runup by July 13; on July 18,

three U.S. scientists, under the auspices of the

January 31, 1994, U.S. -Japan Cooperative

Program in Natural Resources (UJNR), joined

Japanese scientists to complement the tsunami

survey. The 1993 Hokkaido Tsunami Survey

Group included the following teams and

members:

Tohoku University team: N. Shuto Oeader),

H. Matsutomi

University of Tokyo team: Y. Tsuji (leader)

Public Works Research Institute, Ministry of

Construction, team: H. Ito (leader),

K. Yamamoto, Y. Iwasaki, K. Sasaki,

A. Hattori Port and Harbour Research Institute, Ministry of

Transportation, team: Y. Suzuki (leader),

S. Takahashi, C. Goto, N. Hashimoto,

T. Hosoyamada, T. Nagao

Civil Engineering Research Institute, Hokkaido

Development Bureau, team: Y. Mizuno

(leader), K. Yano

United States-Japan Natural Resources team:

E. Bernard (leader), F. Gonzalez, D. Sigrist,

H. Tsuruya, K. Kato

Japan Meteorological Agency team: O. Nagaoka

(leader), S. Honda, H. Tatehata

2. FIELD MEASUREMENTS

The primary measurements in the field surveys were the tsunami vertical and/or horizontal runup

values, which are the maximum vertical height

and/or horizontal extent of flooding. In this report, only vertical runup estimates are presented. Traces left by the tsunami include

water marks on buildings, debris lines along the

coast, or vegetation that is damaged or killed by

salt water. Measurements of vertical height

above sea level of the tsunami trace are obtained

National Oceanic and Atmospheric

Administration, Seattle, Washington 98115

335
by a series of measurements down to the

shoreline, using a surveyor's staff and level. The

effects of tides are normally removed from such

measurements, but during the period of the

survey, the predicted tidal range in the study area was only 2-30 cm, referred to mean sea level. Since this range is a relatively small fraction of

the observed runup, the data presented in this report are not corrected for tides. This report is a collaborative effort that pools 281

measurements acquired by seven of the survey

teams—188 around the island of Okushiri and 93

along the western coast of Hokkaido. These

measurements included 110 by Tohoku

University, 48 by the Public Works Research

Institute (PWRI), 44 through the combined

efforts of the Port and Harbour Research Institute (PHRI) and the Civil Engineering Research

Institute (CERI), 36 by the Japan Meteorological

Agency (JMA), 28 by UJNR, and 15 by the

University of Tokyo. Some sites were visited by more than one team as shown in Figures 2 and

3, which summarize the data.

Another very important source of information is the approximate time of tsunami arrival, as inferred by the time at which clocks have

stopped due to saltwater flooding. The UJNR

team conducted an active search for such clocks

in the tsunami debris and in the surrounding area of each of the 28 sites visited. Eight clocks were

found in various locations—on the beach, in

wrecked automobiles, and in homes and other

buildings—which are taken to be a rough

indication of tsunami arrival times.

3. DESCRIPTION OF THE EARTHQUAKE/

TSUNAMI

Figure 1 provides an overview of the region,

including the aftershock zone (courtesy Japan

Meteorological Agency) for a 40-hour period,

and its location relative to Hokkaido and

Okushiri Island. Note that this zone, frequently

taken as an indicator of the area of crustal deformation, includes the island of Okushiri.

The seismic moment of 5.6 x 1027 dyne/cm

(corresponding to Mw= 7.8) with a (fixed)

centroid depth (a point source representative of

the overall faulting) of 15 km was estimated by

Harvard University. Assuming a fault length of

about 150 km and a fault width of 50 km based on aftershock data and a shear modulus of

3 x 1011 dyne/cm2

, this moment and fault geometry suggests an average slip of 2.5 m on

the fault plane. The focal mechanism estimated

by Harvard has one nodal plane dipping eastward

at 24° from the horizontal. This fault plane

orientation is comparable with the aftershock

distribution and is consistent with subduction of

the floor of the Japan Sea beneath northern

Japan.

The tsunami was probably generated within the

deformation area as defined by the aftershock

pattern shown in Figure 1. Eye-witness accounts

collected by Y. Tsuji (University of Tokyo) and

F. Imamura (Tohoku University) indicate that the

tsunami hit the west coast of Okushiri Island

almost immediately after the main shock. The

UJNR tsunami survey team found the electrical clock at an electrical power station had stopped

at 2223, which was 6 min after the start of the

earthquake. Along the west coast, the tsunami

runup measurements were the highest from north

of Monai (see Figure 2). The village of Monai

was totally destroyed (10 persons killed and all 12 houses destroyed). Runup measurements

around the village were 20 m, and in a small

valley north of Monai, the runup was measured

at 31 m. South of Monai, tsunami runups

between 15 and 20 m were measured all along

the coast. Vegetation was stripped off the

hillside, and large boulders (up to 1 m in diameter) were deposited where the vegetation was flooded. These data are consistent with the

initial tsunami arriving from the west of Okushiri

Island very near the generation area.

The tsunami was refracted by the shoaling

bathymetry at both ends of the Island. Hardest

hit was the town of Aonae (population 1,600),

where the first tsunami wave flooded the

southern tip of the Island and the entire first row

of houses in the harbor area within 4-5 min after the main shock (see cover photos). Tsuji

reported (based on eye-witness interviews) that the tsunami arrived from the northeast, with

flooding of 3-7 m throughout the town. About

7 min after the first wave, a second, larger wave

336
hit from the east carrying boats into the main

town. The second wave completely flooded the

first three rows of houses, and runup was

measured around 5-10 m throughout the town.

The UJNR survey team found battery-operated

clocks in this area that had stopped at 2237 and

2238. Fires ignited at 2240, and the combination

of a strong, northeast wind and an ample supply

of propane and kerosene (used for heating)

spread the fire quickly, which burned through the

night and destroyed 340 homes. Autopsies

revealed that only 2 of the 114 deaths in Aonae

were caused by fire. This section of Aonae was

the hardest-hit developed area in spite of the fact

that a massive, 4.5-m breakwater and 10-m high

sand dunes were very effective in reducing the

runup to 5-10 m along the southeastern tip of

the peninsula (Figure 2). In this regard, note that runup values rise rapidly again to the 10-20-m

level a short distance northeast of the peninsula

(Figure 2); this is probably due to the absence of

breakwaters or sand dunes along this part of

exposed coast.

The extent of damage to Aonae is illustrated in figures 3 and 4. The breakwater in the

foreground of photograph 1 is 4.5 m. The

houses in the central part of the photograph were

flooded by tsunami waves that ran up to a height

of 5 m. To the right of these houses, one can

see the smoking remains of over 200 houses.

The area in the left portion of the photograph was a residential section before the tsunami.

This area was completely stripped of all houses

as seen in photograph 2, a view of the area from

the south. The tsunami destroyed a portion of

the seawall as seen in the lower right-hand

portion of photograph 2. Notice that the seawall was destroyed by wave forces pushing it seaward. Also note the scattered fishing boats in the center of the photograph. In both

photographs, the remnants of the wooden

structures are seen as the brown debris,

completely covering the harbor area and

outsidethe breakwater. These two photographs were taken by Y. Tsuji from a helicopter the

morning after the earthquake.

A similar refraction of the tsunami took place on

the northern point of Okushiri Island at Inaho

where 13 persons were killed and all houses

were destroyed by waves that ran up over 10 m. The eastern side of Okushiri Island was less affected where the tsunami runup was measured

between 2 and 5 m. Subsidence was observed

by Tsuji at Aonae, Okushiri, and Monai, while

uplift was observed at Inaho Point. These data

again suggest that Okushiri Island was part of the

deformation area, which is consistent with the

aftershock data.

The tsunami also hit the island of Hokkaido

(Figure 5), arriving at Ota Bay within 5 min of

the main shock and destroying five homes.

Runup in this area was 9 m. Damage was also

observed at Setana due to 6-m runup waves. The

coastline from Suttsu to south of Ota Bay was

hardest hit with runup values of 5-9 m. Outside

this area, the tsunami intensity tapered off

rapidly, and runup values fall below 5 m. The

closest tide gauge to the earthquake was Esashi,

which recorded a 2-m wave approximately

10 min after the main shock. Those data suggest

that the eastern portion of the source was

between Okushiri Island and Hokkaido.

The tsunami propagated to Russia within 30 min,

where 1-4-m tsunami runups were reported by

Valentin Fedorey (Hydromet). Damage estimates were over $6 million, with $4 million in damage

at Kamenka. After 90 min, the tsunami struck

the coast of South Korea, where B.H. Choi of

Sung Kyun Kwan University measured tsunami

runup of 1-2 m.

4. SUMMARY

In summary, the Hokkaido-Nansei-Oki

earthquake produced one of the largest tsunamis

in Japan's history. Tsunami vertical runup

measurements varied between 15 and 30 m over

a 20-km portion of the southern part of Okushiri

Island with several 10-m values on the northern

portion of the Island. Along the west coast of

Hokkaido, no survey values exceeded 10 m, but

damage was extensive at several coastal towns.

Given the sudden onset of the tsunami and its high energy, it is amazing that more people were

not killed. We also observed the seawall at Aonae was effective in reducing runup

amplitudes.

337
Figure Captions

Figure 1. Locator map of study area. Also shown in the inset is the survey area and the

aftershock zone for a 40-h period.

Figure 2. Distribution of vertical tsunami runup for Okushiri Island. The legend indicates

data source, and runup values are projected onto offshore scales. A projection from a data point

onto the coastline will identify the geographical position of the measurement.

Figure 3. A view of the tsunami damage from the east of Aonae, a small town on the island of

Okushiri, which is in the Sea of Japan east of Hokkaido. Photo by Y. Tsuji.

Figure 4. A view of the tsunami damage from the south of Aonae. Photo by Y. Tsuji.

Figure 5. Distribution of vertical tsunami runup for the west coast of Hokkaido. Legend and data

presentations are the same as in Figure 2.

338
Hokkaido Nansei Tsunami

130° 140°E

339
Hokkaido Nansei Tsunami

12 July 1993

Tohoku U. OPWRI

AUJNR

-j^-U. Tokyo

0 PHRI/CERI

+ JMA

340
341
342
Hokkaido Nansei Tsunami

12 July 1993

343

Propagation of Hokkaido-Nansei-Oki Earthquake

Tsunami Around Cape Aonae

by

Shigenobu Tanaka", Shinji Sato 2) and Kenji Noguchi 3)

ABSTRACT

Hokkaido-Nansei-Oki earthquake

rocked the west coast of Hokkaido

on July 12, 1993. It caused

devastating tsunami which

quickly hit Okushiri Island and the

west coast of Hokkaido. Because

the tsunami source was in the

west of the island, the tsunami

runup was generally large on the

west coast but small on the east

coast of the island. However, it was pointed out that the runup

height at Hamatsumae was

extraordinarily larger than that

near around. The location is on the

southeast edge of the island, but

it was sheltered from the tsunami

source area. So, it has been

remained as a problem what

mechanism had occurred during

the tsunami propagation. This

paper shows in what way the

tsunami propagates in shallow

water around Cape Aonae and

attacks Hamatsumae based on a

large scale hydraulic model test.

KEYWORDS

Hokkaido-Nansei-Oki earthquake;

tsunami runup; hydraulic model

test; soliton wave; non-linear

effect on tsunami propagation

INTRODUCTION

Hokkaido-Nansei-Oki earthquake

of magnitude 7.8 rocked the west

coast of Hokkaido at 22:17 on July

12, 1993. The earthquake caused

devastating tsunami which quickly

hit Okushiri Island and the west

coast of Hokkaido. The tsunami

also attacked whole edge of the

Japan Sea. The death toll was 202,

29 were still missing, and the

property losses were 145 billion

yen within Japan.

1) Head of Coastal Engineering Division,

River Department, Public Works

Research Institute, Ministry of

Construction

2) Dr., Senior Researcher, ditto

3) Researcher, ditto

345
Public Works Research Institute, Ministry

of Construction (PWRI), surveyed

vertical tsunami runup heights along the

southwest coast of Hokkaido and

Okushiri Island's coast. Also, many

survey teams measured vertical runup

heights. The highest runup 31 .7m, which

is the highest record in this century in Japan, was found at Monai, west coast of

Okushiri Island. The runup heights along

west coast were about 5 to 22m except

for Monai, and those around south tip,

Aonae Cape, and north tip, Inaho Cape,

were about 1 0m. Those for the east side

of the island were about 3 to 8m except

for around Hamatsumae. The highest

runup in Hamatsumae was about 20m.

Among the researchers, it was pointed

out that the runup height at

Hamatsumae was extraordinarily larger

than that near around. The location is on

the southeast edge of the island, but it was sheltered from the tsunami source

area. So, it has been remained to be

investigated what mechanism had

occurred during the tsunami

propagation.

In order to study this phenomenon,

PWRI carried out a large scale

hydraulic model test. The result

reveals the tsunami propagation

mechanism clearly. A refracting wave

and a diffracting wave generated by the

offshore shoal attack together in a big

synthetic wave against Hamatsumae

coast. In the nearshore, furthermore, the

second soliton catches up to the first one

and then runup onto the shore.

FIELD SURVEY

PWRI sent a survey team to survey

the damage due to the tsunami on

July 14. They selected local

maximum runup points and

measured vertical height above the

sea level of each point using a

handlevel and staffs. They also

wrote down the time of each

measurement and modified the

vertical height referring to the

tide table of Esashi Port. Figure 1 shows vertical runup heights for

the southwest coast of Hokkaido.

The runup heights are higher in

Shimamaki, Setana, Kitahiyama

and Taisei. Figure 2 shows the

runup height distribution for

Okushiri Island. PWRI survey team

could not survey some west part of

the island. PWRTs highest runup

19.8 m was found at Cape Kuki and

the second one 15.9 m was at

Hamatsumae.

Figure 3 shows distribution of vertical

tsunami runup for Okushiri Island

compiled by Bernard and

Gonzalez(1 993). It can be seen that the

346
Cape Lnaho-r4

Inaho Fiherv PortMatsue Fishery Pore

Hamatsumae

n.s

-The north of Aonae Fishery- Port

Fig. 1 Distribution of vertical runup for

the southwest coast of Hokkaido

Fig. 2 Distribution of vertical runup

for Okushiri Island

347
Hokkaido Nansei Tsunami

(after Bernard et. al.) Fig. 3 Distribution of vertical runup

for Okushiri lsland(after E.

Bernard et al.)

highest runup more than 30m was

measured at Monai on the west coast.

On the east coast, however, because it is sheltered from the tsunami source, runup

height was generally small except for the

southeast edge, Hamatsumae, where

runup was as high as about 20m. It remains to be investigated why so high

tsunami runup took place at

Hamatsumae only.

EXPERIMENT ON TSUNAMI

PROPAGATION

PWRI carried out a hydraulic model

experiment in order to find out the

above phenomenon because it was

pointed out that numerical simulations

could not give good explanation to the

phenomenon.

(1)Model and Tsunami Generator

Figure 4 shows seabed topography

around Cape Aonae. It can be seen that

there is a very shallow region stretching

to the south from the cape and there is Murotu Shoal about 4 km offshore. The

water depth between Cape Aonae and

Murotu Shoal is less than 20m. Figure 5

shows tsunami travel time and refraction

diagram around Cape Aonae by

numerical simulation. It can be seen that

tsunami source is very close to Okushiri

Island and tsunami of wide area refracts

through the shallow region and gathers

Aonae and Hamatsumae.

As the purpose of our experiment

was focused on finding out the

mechanism which caused the high

tsunami runup at Hamatsumae, we

made an extensive model to

include all composition in the

source affecting Hamatsumae. The

scale of the model is 1:1100

without distorsion. Figure 6

shows a plan view of our model.

Figure 7 shows sectional profile of

pneumatic type tsunami generator. It can generate a tsunami by sucking water

348
349
height in the model(cm) i -40

-35 mFig. 6 Plan view of experimental model>3 msuction valve

release valve

\

Fig. 7 Sectional profile of tsunami generator

3S0
into the air chamber and releasing it suddenly. Any reflected waves by the

wall of tank do not affect the first tsunami

attacking Hamatsumae in this model.

(2) Measurement

Tsunami profile was measured

with capacitance-type wave gages

at 30 points shown in Fig. 6 and 8.

Water level after the suction was

set to the mean water level.

Because the scale of the model is

very big, we do not strictly stick

to the similarity of the wave

height but we focused on the ratio

of wave height at each point to the

incident wave height. In order to

catch the movement of crest line,

we record wave with a VCR.

(3) Results

Figure 9 shows wave profiles with

incident wave height of 4mm. As

to the first wave, soliton fission

was observed in the east side of

the ridge of the spur. Especially, a

very clear soliton fission was

observed at pt. 10. But at pts. 25,

26 and 27, near Hamatsumae, soliton

fission was not so clear and wave profile

was sharp and peculiar. On the other

hand, at points near around Cape Aonae

and on the ridge or in the west side of the

X

4 6 8 10 y(m)

Fig. 8 Arrangement of wave gages

ridge of the spur, no soliton fission was

observed in the first wave. The first

soliton at pt. 10 had a longer period than

the successive ones and the one at

another point.

Concerning wave height, the

magnifi-cation ratio against the

incident wave height was about

two at pts. 14, 15, 17, 24 and 25.

The ratio was about three at pts.

10, 26 and 27. In this way, wave

height was magnified near around

Hamatsumae. The occurence of the

351
incident wave

pt.11 pt.7 —i

t—i r-4**-f i -t i -*-\-f 1- 4

1 0 15 0 TT

Fig. 9 Tsunami profiles measured with wave gages

(incident wave height : 4mm)

large wave in this area agrees well with

the field survey, although the maximum

wave height in the experiment was

attained at a little northern point.

It was also observed that a reflected

wave due to Cape Kuki propagated to the

south along the coast and diffracted at

Cape Aonae. This situation appeared as

the second wave in the record at pts. 1 1

,

12, 18 in Fig. 9. At the nearest point to

Cape Kuki, pt. 19, the reflected wave

arrived at 13sec after wave generation

with a clear soliton fission and it passed

pt. 18 and then pts. 11, 12. Point 19 is located near the west foot of Cape

Aonae. At this point, the reflected wave

was superimposed to the incident first wave, resulting a rather big wave height.

The magnification ratio at pt. 1 9 against

the incident wave height was about

352
(a) waves as soliton fission

4th 5th

Fig. 10 Snapshots of wave crest line

Fig. 11 Sketch of the propagation of the wave front

353
three. This value is almost same as that

near Hamatsumae. Furthermore, it agrees well with the distribution of runup

height shown in Fig. 3.

In order to study the mechanism of the

occurrence of the large wave around

Hamatsumae, the propagation of wave

crest lines was investigated in detail

through video frame analysis.

Figure 10 illustrates snapshots of

wave crest line in the vicinity of the

southeastern edge of the Okushiri Island.

As was explained in Fig. 9, the tsunami

was split into several solitons. Since the

area in the vicinity of the shoreline is shallower than the offshore area, the

direction of the propagation of

each soliton is changed through

refraction so that the crest lines

become parallel to the shoreline.

The celerity of the soliton is also

affected by the bottom topography

so that the celerity in the vicinity of the

shoreline is smaller than that in the

offshore. The celerity of the second

soliton is larger than the first one since

the second soliton propagates on a

relatively high water level as seen in Fig.

9. The bottom topography in Fig. 6

indicates that the first soliton lost its speed in the vicinity of the shoreline and

the crest line of the first soliton was

strongly curved at around pt. 27. It is

also noticed that the second soliton

catched up to the first soliton at this

location. It is therefore considered that

the wave height at pt. 27 becomes large

due to the combined effects of the

refraction and the merge of two solitons.

Figure 11 illustrates a sketch of the

propagation of the wave crest line of the

first soliton. As is noticed in Fig. 6, there

is a shallow shoal at the upperright edge

of this figure. In the behind of this shoal,

two crest lines were formed owing to the

wave diffraction. The two crest lines

intersect with a thin angle, forming a new

crest with large amplitude through the

nonlinear interaction. Since the newly

created wave propagates with large

celerity because of its large

amplitude, another nonlinear

interaction occurs at the both

sides of the newly created wave

crest. Figure 1 1 indicates that the large

wave crest, which was formed through

repeated nonlinear interaction, may hit

the area of pts. 26 and 27.

CONCLUSIONS

Laboratory experiments were performed in order to study the

mechanism of the occurence of

large waves along Okushiri Island

observed in the field survey of the

Hokkaido-Nansei-Oki Earthquake

354
Tsunami. The propagation and the

transformation of the tsunami

were measured in detail with wave

gages and a VCR. It is concluded that

the two mechanisms are the reason for

the occurrence of the large wave. The

one is the combined effects of the wave

refraction and the merge of the solitons

which was split due to the effect of wave

dispersion. The other is the nonlinear

interaction of two waves intersecting with a thin angle.

ACKNOWLEDGEMENTS

The authors wish to express their

appreciation to Professor N. Shuto,

Tohoku University, for his helpful

suggestions on the design of the

hydraulic model test.

REFERENCES

E. Bernard and F. I. Gonzalez:

Tsunami Devastates Japanese

Coastal Region, Eos, Transactions,

American Geophysical Union, Vol.

74, No. 37, pp. 417-432,

September, 1993.

355

Numerical Modeling Tsunamis

by

Philip L.-F. Liu*

ABSTRACT

A brief review of different aspects of

numerical modeling of tsunami generation, propagation and inundation is given. Several difficulties are identified.

They include the accurate description of

the initial conditions from the seismic

data, the appropriate modeling of wave

breaking in the inundation model.

KEYWORDS: tsunami; inundation;

propagation; numerical modeling.

1. INTRODUCTION

Tsunamis are impulsively generated

ocean waves. They may be caused by an earthquake, landslide or volcanic

eruption. In this paper we shall focus our discussions on earthquake-generated

tsunamis. In an open ocean tsunamis are best described as long waves: the

characteristic wavelength of tsunamis is usually one to two order of magnitude

larger than the typical water depth. In the

Pacific Ocean, tsunamis move at speeds

of over 500 miles an hour, but their wave

heights are usually low, perhaps only a

foot. As tsunamis move into the

continental shelf, they are refracted to conform to the shape of the bathymetry.

At the same time, the shoaling water

causes wave heights to increase greatly.

As the tsunami rushes landward, the

maximum run-up could reach as high as 30 m.

Historically, many tsunamis have caused

severe damages in Japan, U.S. and other

countries around the Pacific rim.

Researchers and engineers have been

working toward a goal which can not

only predict the arrival time of a tsunami,

but also calculate the detailed inundation

map. Various degrees of success have

been achieved towards this ambitious

goal. In this paper, we will review briefly

different elements of the tsunami

modeling. We shall stress particularly

the difficulties and unresolved issues.

2. SOURCE REGIONS

To model the propagation of a tsunami

across an ocean, one must first calculate

the initial water surface displacement,

which depends on the characteristics of

the earthquake. Most of large tsunamis

have been generated near plate boundaries

where an oceanic plate and a continental

plate meet. The oceanic plate is subducted under the continental plate,

while each plate moves slowly against

each other. Although the movement is slow, roughly 2 to 10 cm/hr, a significant

amount of stress and stain is accumulated

along the interplate boundary. When the

strain reaches the failure strain, a portion

of the subduction boundary breaks and

causes the rebound of the continental

plate. Hence the earthquake. For an

ordinary earthquake, the size of the source region is much greater than the

water depth and the rupture velocity is usually very short in comparison with the

tsunami propagation velocity. Therefore,

the initial water surface displacement is almost exactly the same as the vertical

displacement of the seafloor.

*School of Civil and Environmental

Engineering, Cornell University, Ithaca,

NY 14853

357
Using the elastic dislocation theory,

Mansinha and Smylie (1971) obtained

analytical expressions for the seafloor

displacement due to inclined finite, strikeslip and dip-slip faults. To utilize

Mansinha and Smylie's model, the

following parameters are required: the

length of the fault area (L), the width of

the fault area (W), orientation of the fault area with respect to the North (9) the dip

angle (8), the slip angle (X), the

dislocation on the fault (D), and the focal

depth (d). All these parameters can be

estimated from the seismological data.

However, the accuracy of these

parameters is often not certain. For

instance, the dip angle, the slip angle, the

orientation of the fault, and the seismic

moment (MQ ) can be obtained by

inverting the spectra of both Rayleigh and

Love waves over a specified period (e.g.

Kanamori and Given, 1981). The fault area is determined approximately from

the distribution of aftershocks. The

dislocation on the fault, D, can be

approximated as

where |i is the rigidity of the earth at the source region. Therefore, this simple

model ignores the heterogeneous fault

motion, which could become important

in some tsunamis (e.g. Satake and

Kanamori, 1991).

/ In some subduction zones, a layer of

subducted soft sediments could be

sandwiches between the oceanic and

continental plates. The rupture velocity

becomes smaller, and the high-frequency

part of the seismic spectrum is greatly

reduced for the same seismic moment

M Q . Therefore, a small earthquake

measured in terms of seismic waves of

high frequencies could produce a large

tsunami as long as the rupture velocity is

still much smaller than the tsunami

propagation velocity. This kind of

earthquake is called tsunami earthquake

(Kanamori, 1972). Further complication

could exist when an accretionary prism

appears near an oceanic trench. Because

of the large amount of soft sediments in

the accretionary prism, seismic slip

cannot reach the seafloor. Thus these

earthquakes do not excite tsunamis very

efficiently. As the deformation in the

accretionary prism proceeds as a result of

repeated occurrence of these events,

sediment instability may eventually

develop and finally a large-scale

slumping occurs. This slumping may be

the cause of a large tsunami.

While the elastic dislocation theory

(Mansinha and Smylie, 1971) has been

used widely to calculate the initial water

surface displacement, numerical

simulations of three recent tsunamis

indicate that the theory under-estimates

the resulting runup heights (Imamura et

al., 1992). The causes for the discrepancies are not clear. Furthermore, the

theory for determining the size and the

shape of slumping and the corresponding

water surface displacement has not been

developed.

3. TRANSOCEANIC TSUNAMI

PROPAGATION

To calculate the propagation of a tsunami

over an open ocean, one needs to select

appropriate governing equations. In the

vicinity of the source region, the wavelength of the leading wave is roughly the

same as the length of the source region,

L, which is usually much greater than the

water depth. Furthermore, the initial free

surface displacement is also small in

comparison with the depth. For instance,

in the source region of the 1960 Chilean

tsunami, the initial free surface displacement has a wave height about 8 m and a wavelength of 600 km, while the

averaged water depth in the Pacific Ocean

358
is 5 km. Therefore, the wave slope is 0(10"5

) and the depth to wavelength ratio

is 0(10~2

). It is obvious that the linear

long wave theory is quite adequate for

describing the transformation of the initial surface profile in the neighborhood of the source region.

As the tsunami propagates into an open

ocean, the wave amplitude decreases

because of the two-dimensional spreading. The wavelength increases due to the

dispersive effects. Away from the

continental shelf, the transoceanic

tsunami propagation can be suitably

described by the long-wave approximation. Although Boussinesq equations

are the most uniformly valid, solving

simpler equations is more preferable, if appropriate. Kajiura (1963) suggested

that within the following traveling

distance

R = — h (^)3 64 h

(2)

the linear and nondispersive shallowwater theory is adequate. Kajiura's

analysis was based on a two dimensional

theory. More recently, Hammack and

Segur (1978) re-examined the one- dimensional waves and reached the

similar conclusion that the dispersion

effects become important only if the

leading wave of the tsunami has traveled

a distance much greater than 6h (L/h)3

. Using (2) as a guide, we can estimate the

region within which the linear and

nondispersive theory is valid for the 1960

Chilean tsunami, i.e. R = (6/64)(5)(102

)3 = 106 km, which is much larger than the

dimension of the Pacific Ocean.

Therefore, as far as the modeling of the

leading wave of the 1960 Chilean

tsunami is concerned, the linear,

nondispersive wave theory is appropriate.

On the other hand, for a small size

earthquake such as the 1983 Hokkaido

earthquake in the Japan Sea, the length of

the source region is L = 20 km and the

averaged depth is h = 2 km. As the result

R = 200 km, while the length of the

Japan Sea is about 900 km. Therefore, if one needs to obtain the tsunami

information along the Korean coastal

region, the dispersive effects must be

included in the propagation model.

The Coriolis parameter of the earth is roughly 10~4 rad/s. For a transoceanic

tsunami across the Pacific Ocean the

Coriolis force becomes important

because the traveling time is about 23

hours or 8.28 x 104 sec. Similarly, for

the tsunamis in the Japan Sea, the

traveling time from the Japanese west

coast to the Korean east coast is approximately 9 hours or 3.24 x 104

sec. Therefore, the Coriolis force needs to be

included in the computations if such a long duration is desirable.

In summary, to model the tsunami

propagation over a long distance, the

linear Boussinesq equations with the

Coriolis force is the most adequate

governing equations:

dM

dt

^ + V-M = 0

at

+ gh Vrj + 2Q. x M = £

3

Ay3t

A

h

(3)

(4)

where n. is the free surface displacement,

M the depth-averaged volume flux, h the

depth and Q the Earth's angular velocity.

Equation (3) is the exact expression of

the conservation of mass, while equation

(4) represents the approximate equation

for the conservation of momentum

averaged over the depth. The right-hand

side term of equation (4) is the

contribution from the vertical acceler359
ation. Note that if this term is ignored,

equations (3) and (4) become the linear

shallow-water equations with the Coriolis

force.

Equations (3) and (4) can be rewritten in a scalar component form. The spherical

coordinate system coinciding with the

longitude and the latitude of the Earth

should be employed if the computational

domain is large, such as the entire Pacific

Ocean. On the other hand, for a smaller

region such as the Japan Sea, the local

Cartesian coordinate system can be used.

Many numerical schemes are available

for solving the Boussinesq equations.

Because the modeling of transoceanic

tsunami propagation involves a large

computational domain and a long

simulation time, both accuracy and

efficiency are essential. In this paper we

present and discuss an explicit, leap-frog,

finite difference scheme, which has been

used in many applications (Liu, et al. 1994).

9t

2

3 'a

4

r|

2„\

gh + 2

dy<

4~\

a

2 r] cTri

dx2 dy2 +

By

4 (8)

Liu, et al. (1994) show that the following

explicit, leap-frog, finite-difference

scheme satisfies (8) up to the third order

of the grid size and time step size. Thus,

-n+1/2

At

Qi,j+l/2

+

r i+l/2,j " r i-l/2.j

Q

n

i 12 Ay

Ax

= 0 (9)

pn+ l pn ^n+1/2 ^n+1/2 ri+l/ 2,j " r i +l/2,j + ^

_Mi+l,j

~

At Ax

gh

[(n

n+1/2

- 2nn+1/2 + n

n+1/2

)] l2Ax ,j J ,i+1 'j-1 /J ^„.i/2 . 2vS

*n + ^2)] = o (10)

For the ease of presentation, we only

demonstrate the case where the Coriolis

force is ignored and the depth is constant. The linear Boussinesq equations, (3) and

(4), can be written in the following

component forms:

*U^ + 3Q =0

3t dx dy —ap u ari + gh

- dx dx

—dQ , arj - + gh

- at * ay

3

a

3 p

atax

2 +

(5)

a

3 Q

dxdydt

(6)

a

3 p

axayat +

a

3 Q

atay

2

(7)

where P and Q are the volume fluxes in

the x- and y-direction, respectively. By

eliminating P and Q from (5), (6) and

(7), we obtain

Qn+l /-\n —n+1/2 ^n+1/2

Vi,j+_W2_ + ^

At Ay

+

g^1 [^"+ 1/2 9 T| n+l/2 , ^n+1/2 \~\ 12Ay - [«f - 2C1/2 + Cif)] = o

(11)

in which indices (i,j) denote spatial nodes

and n the time level, Ax and Ay represent

the grid sizes in the x- and y-direction,

respectively, and At the time step size.

The dispersion effects, represented by the

right-hand side terms in the Boussinesq

equation (8), are fully simulated by the

finite-difference scheme (9), (10), and

(1 1), up to the third order if the following

condition is satisfied

(Ax)2 = 4h2 + gh (At)2 (12)

360
The leap-frog scheme is stable if the

Courant number, Q = -N/gh At/Ax, is less

than or equal to 0.866.

In practical applications, water depths

vary and equation (12) cannot be satisfied

everywhere. The computational domain

is usually divided into several

subdomains according to the objectives

of the simulation. Within each subdomain an averaged water depth is used

in (12) to determine the appropriate At

and Ax. In general, in the region where h

is smaller, a finer grid system with a smaller time step is necessary. A

numerical algorithm has been developed

for exchanging information between two

regions of different grid sizes (Liu, et al. 1994).

4. INUNDATION MODEL

As the tsunami propagates over the

continental shelf and enters the coastal

zone, the amplitude grows and the

wavelength shrinks. The nonlinearity

gradually becomes important. During the

final flooding stage, a tsunami might

break and the energy dissipation features

significantly. Even if the tsunami does

not break the bottom frictional effects are definitely important because of the

shallowness of water.

In terms of numerical modeling, the

nonlinear shallow-water equations with

dissipation terms are the appropriate

governing equations. However, our knowledge on modeling wave breaking is

still in its infancy. At the present, only

the bottom frictional terms are included in

the governing equations.

dt dx dy

(13)

£ + A & + ± ( PQ)

9t dx H dy H

+ gH + x x H = 0 (14)

dx

^ + A (£Q) + ± (21)

3t dx H dy H

+ gH + x

y H = 0 (15)

dy

where H = n + h is the total depth and xx and T

y are the x- and y-component of

bottom frictional force, respectively.

Notice that the nonlinear convective

forces are included in the momentum

equations, (14) and (15). The frictional

force is expressed in terms of either the

Manning's n or the Chezy's coefficient Cf

.

Both Manning's and Chezy's formula are empirical formulas for steady state flows.

The direct application of these formulas

to a transient flow problem such as the

tsunami inundation is not obvious.

Moreover, during the run-up phase of the

leading wave of a tsunami, a mass of

water moves onto a dry land. The

bottom friction caused by a dry land is different from that by a wet surface.

Equations (13), (14) and (15) can be

integrated numerically by using the

similar leap-frog scheme presented in (9),

(10) and (11). The nonlinear terms are linearized by a upwind scheme. As a

result, the accuracy of the numerical

scheme is reduced to the first order in Ax

and At. Therefore, a fine grid mesh is usually required to calculate the tsunami

inundation heights. A moving boundary

algorithm is also necessary for computing the movement of the shoreline.

Liu et al. (1994) presented a simple

scheme, which satisfies both conservation of mass and momentum.

Many issues concerning the inundation zone are to be resolved. For instance, the

361
interaction of tsunami and large obstacles,

such as buildings and coastal facilities,

needs to be examined. On the other

hand, in an open coast, tsunami might

cause significant sediment transport,

which might be a good indicator for the

maximum run-up heights.

5. CONCLUDING REMARKS

The computing technology has been

improved significantly in the last decade.

This trend will continue in future.

Therefore, the accuracy and the capability

of the tsunami modeling have also been

increased. The weakest link of the

tsunami modeling is the accurate

description of the initial conditions from

the seismic data. Because the fault

parameters are obtained from the seismic

waves, it might be useful to examine the source region problem by using the

tsunami data directly. In the inundating

zone, the physics of the dissipative

mechanism, such as wave breaking, need

to be understood. Forces acting on the

coastal structures and buildings and

tsunami induced sediment movements

should also be examined in the future.

6. ACKNOWLEDGMENTS

This research was, in part, supported by a grant from the National Science

Foundation to Cornell University

(BCS9024965).

7. REFERENCES

1. Hammack, J.L. and Segur, H. 1978

"Modeling criteria for long water

waves", J. Fluid Mech., 84:359-373.

2. Imamura, F., S. Ide, Y. Yoshida, K.

Abe, and N. Shuto, 1992 "Estimate

of the Tsunami Source of the 1992

Nicaraguan Earthquake from

Tsunami Data", Geophys. Res. Let.

3. Kanamori, H. 1972 "Mechanism of

tsunami earthquakes", Phys. Earth

Planet Inter. 6:346-359.

4. Kanamori, H. and Given, J.W. 1981

"Use of long-period surface waves

for fast determination of earthquake

source parameters" Phys. Earth and

Planet.

5. Lewis, C.H. and Adams, W.M. 1983

"Development of a tsunami-flooding

model having versatile formulation of

moving boundary conditions",

Monograph series, The Tsunami

Society.

6. Liu, P.L.-F, Cho. Y.-S., Seo, S.N.

and Yoon, S.B. 1994, "Numerical

Simulation of Tsunami Propagation

and Inundation with Application to Hilo, Hawaii", Res. Rep. School of

Civil and Environmental

Engineering, Cornell University.

7. Mansinha, L. and Smylie, D.E. 1971

"The Displacement of the Earthquake

Fault Model", Bull. Seismol. Soc.

Amer. 61:1400-1433.

8. Mei, C.C. 1989 " The Applied

Dynamics of Ocean Surface Waves"

World Scientific Publishing.

9. Satake, K. and Kanamori, H. 1991

"Use of tsunami waveforms for

earthquake source study", Natural

Hazards, 4:193-208.

10. Shuto, N. 1991 "Numerical Simulation of Tsunamis - Its Present and

Near Future", Natural Hazards,

4:171-191.

362
SUMMARY OF JOINT

COOPERATIVE RESEARCH

PROGRAMS

The Status of the U.S. Precast Seismic Structural

Systems (PRESSS) Program

by

M. J. N. Priestly 1 and H. S. Lew2

ABSTRACT

The paper provides a brief overview of the

recently completed Phase I U.S. PRESSS

program, and the current Phase II program,

which emphasizes theoretical and experimental

studies of ductile connection systems for precast

frame and panel structures KEYWORDS:

buildings; concrete; earthquake; precast;

research; seismic; structures

1. INTRODUCTION

The U.S. PRESSS coordinated research program

on Precast Seismic Structural Systems has now

been underway for four years. The aims of the

program are to develop new economically and

technically viable precast systems for seismic

zones, and to develop design recommendations

to enable these to be used in practice.

A preliminary feasibility phase (Phase I) of

research was completed last year. This phase

concentrated on analytical and design studies to investigate feasible design concepts, and to provide frameworks for design recommendations

and for future analytical parameter studies

needed to calibrate and quantify the design

recommendations

.

A second phase of research, involving analytical

and experimental studies was initiated prior to completion of Phase I feasibility studies.

It is anticipated that several new programs will

be funded, with particular emphasis on panel

structures, analytical studies and cladding/frame

interaction. Applications have been submitted to the National Science Foundation for research

relative to the performance of precast structures

in the January 17, 1994 Northridge earthquake.

As a consequence of the post-earthquake

activities the scope and extent of the U.S.

PRESSS program is expected to expand

significantly in the near future. The following

notes briefly summarize the Phase I and II programs.

2. PHASE I RESULTS

As mentioned above, Phase 1 research is complete, and the following final reports are in preparation.

2.1 PRESSS Concept Development [1]

The report summarizes analyses and design

studies carried out to identify and investigate

feasibility of different structural concepts for

precast frame and panel structures.

2.2 PRESSS Connection Classification [2]

Different possible connection systems between

precast elements are collated and reviewed for

technical merit, constructability and versatility,

using a standardized review format.

2.3 PRESSS Analytical Platform [3]

The report consists of a user manual for a modified version of the well known DRAIN 2D

University of California, San Diego, La Jolla,

California 92093

National Institute of Standards and

Technology, Gaithersburg, MD 20899

365
inelastic time-history analysis program. Three new programs have been developed. Two of

these are three-dimensional programs (DRAIN3D and DRAIN-BUILDING), one with

simplified input and the ability to simply

constrain floor nodes in accordance with rigid

diaphragm behavior, and the other, an enhanced

version of DRAIN-2D, with features and

hysteretic models appropriate for precast

structures.

2.4 PRESSS Design Recommendations [4]

The principal effort described in the final report on this project was an investigation of the

advantages of use of reliability theory to seismic

design of ductile precast concrete structures. In

addition, a framework of options for seismic

design recommendations, using existing and

novel code formulations has been prepared [5].

3. PRESSS PHASE II PROJECTS

The following projects are currently funded in Phase II of the U.S. PRESSS effort.

3 . 1 Ductile Connections for Precast Concrete

Frame Systems [6]

Research is being jointly carried out at the

University of Minnesota and the University of

Texas at Austin, to provide rational design

recommendations for seismic design of precast

frame systems using ductile connections, based on experimented testing of a number of

options representing the four categories of

connectors being considered in the PRESSS

Phase II research program. The categories are:

1 . Nonlinear elastic connections

2. Connections relying on axial tension

and compression yield of connecting

elements

3. Connections involving shear yield

4. Connections involving added damping.

To date, tests have been carried out on the

first two categories.

3.2 Ductile Connections for Precast Concrete

Panel Systems [7]

Parallel to the above mentioned frame

program, a study being jointly carried out at the National Institute of Standards and

Technology (NIST), and the University of

Nebraska-Lincoln is investigating ductile

connections for panel structures, and seismic

systems for low-to medium-rise buildings in

regions of moderate seismicity.

Current testing is focusing on vertical

connections between panel units, where the

connectors are designed to be weaker than the

panels, and to exhibit some energy dissipating

characteristics.

3.3 Precast Frames with Unbonded Tendons

[8]

At the University of California, San Diego, the use of continuous prestressing with unbonded

tendons is being investigated. The focus of the

research is to structurally simplify the precast

elements as much as possible, and to minimize

the amount of mild steel reinforcement.

Particular emphasis is being placed on the

force transfer mechanism within the beamcolumn joint region. Preliminary testing has

been carried out on one exterior and one

interior beam/column units, and six more are planned. Results indicate much lower levels of

damage for a given drift level than for

equivalent monolithic joints of similar

dimensions and strength.

3.4 High Performance Fiber-ReinforcedConcrete (FRC) Energy Absorbing Joints

for Precast Concrete Frames [9]

Building on previous research at the University

of Michigan, Ann Arbor, experimental

research is being carried out to provide design

criteria for FRC plastic hinge regions forming

the connections between precast elements.

The specific objectives are (i) to determine the

bond-stress/slip relationship between FRC and

rebar in the plastic hinge region, (ii) to determine the shear strength of FRC under

366
monolithic and cyclic shear loading, and (iii) to develop a model for the joint to describe its hysteretic moment-rotation response.

3.5 Seismic Response Evaluation of Precast

Structural Systems for Various Seismic

Zones and Site Characteristics [9]

This analytical project is being carried out at Lehigh University. The aims are to carry out dynamic inelastic analyses of specific precast

frame and panel buildings with connection

characteristics appropriate for the four basic

connection categories described earlier. Based

on the results of these analyses, appropriate

simplified design rules will be generated. Of

particular importance is the quantification of

dynamic amplification effects for shear and

moment in columns, beams, and panels, so

that required strength of precast elements can be determined based on appropriate capacity

design principles.

3.6 Development of Seismic Design

Recommendations for PRESSS

The continued refinement of design

recommendation, begun in Phase 1 of the U.S.

PRESSS project, is being continued in Phase

II. This will be a joint effort with input from

the University of Illinois, the University of

California, Los Angeles, and the University of

California, San Diego. Current efforts are well

advanced to provide prescriptive requirements

for 'strong-connection' precast frames [9].

3.7 Dynamic Scale Model Tests

A program of small scale dynamic shake table model testing of precast frames using different

connector types is currently being planned at the University of Illinois, Champaign, Urbana.

These tests will enable comparisons to be

made between predicted results from time

history analyses, and experimental results on

complete structural systems.

4. CONCLUSIONS

The above represents a very brief summary of

the U.S. PRESSS program. It is expected that further projects will be added to the program

in the near future. A final stage (Phase III) is currently planned involving full-scale subassemblage testing of one or more five- story precast structures to test the final design

concepts and analytical models developed in the program. These will be accompanied by

rigorous analytical studies, and comparison of

theoretical predictions and experimental

results.

5. REFERENCES

[1] Dow Nakaki, Suzanne, and R.

Englekirk, "PRESSS Concept

Development, Project 1.1 - Final

Report," Report No. PRESSS 94/03

(In Press).

[2] Stanton, J., "PRESSS Connection

Classification, Project 1.2 - Final

Report," Report No. PRESSS 94/04

(In Press).

[3] Powell, G.H. and F. Filippou,

"PRESSS Analytical Platform, Project

1.4 - Final Report," PRESSS Report

94/06 (In Press).

[4] Hart, G., "PRESSS Design

Recommendations, Project 1.6 - Final

Report", Report No. PRESSS 94/06

(In Press).

[5] Priestley, M.J.N, and G. Hart, "A

Framework of Options for PRESSS

Seismic Design," Proceedings from the

"Fourth Meeting of the U.S. -Japan

Joint Technical Coordinating

Committee on Precast Seismic

Structural Systems," Tsukuba, Japan,

May 1994.

[6] French, C. and M. Kreger, "PRESSS

Phase II: Ductile Connections for Precast Concrete Frame Systems,"

367
Proceedings from the "Fourth Meeting

of the U.S. -Japan Joint Technical

Coordinating Committee on Precast Seismic Structural Systems," Tsukuba,

Japan, May 1994.

[7] Schultz, A., M. Tadros and R. Magana, "Ductile Connections for Precast Panel Structures - Concepts

and Experiments," Proceedings from

the "Fourth Meeting of the U.S. -Japan

Joint Technical Coordinating

Committee on Precast Seismic

Structural Systems," Tsukuba, Japan,

May 1994.

[8] Priestley, M.J.N, and G. MacRae,

"Testing of Two Precast Post- Tensioned Beam/Column Joint

Subassemblages with Unbonded

Tendons," Proceedings from the

"Fourth Meeting of the U.S. -Japan

Joint Technical Coordinating

Committee on Precast Seismic

Structural Systems," Tsukuba, Japan,

May 1994.

[9] Priestley, M.J.N. , Editor, "Report on

the Fourth U.S. PRESSS Coordinating

Meeting," Report No. PRESSS 94/02,

March 1994, 173 pp.

[10] Englekirk, R., "Damage to Ductile

Steel Frames in the Northridge

Earthquake," Proceedings from the

"Fourth Meeting of the U.S. -Japan

Joint Technical Coordinating

Committee on Precast Seismic

Structural Systems," Tsukuba, Japan,

May 1994.

[11] Cheok, G. and J. Stanton, "Beam-toColumn Connections for Precast Concrete Moment-Resisting Frames,"

Proceedings from the "Fourth Meeting

of the U.S. -Japan Joint Technical

Coordinating Committee on Precast Seismic Structural Systems," Tsukuba,

Japan, May 1994.

368
The Present and Future of the United States-Japan Cooperative Earthquake

Research Program Utilizing Large-Scale Testing Facilities

by

Hiroyuki Yamanouchi*

ABSTRACT

First, this paper reviews briefly the history and

significance of the U.S. -Japan Cooperative

Earthquake Research Program Utilizing Large

Scale Testing Facilities. Then, some proposals regarding a new stage of future

cooperation are presented.

KEYWORDS:

U.S.-Japan Cooperative Research Program,

future cooperation, framework

INTRODUCTION

The U.S.-Japan Cooperative Research

Program on Earthquake Engineering Utilizing

Large Scale Test Facilities under the auspices

of the U.S.-Japan Natural Resources Panel on Wind and Seismic Effects (UJNR) was

initiated in 1979 by an agreement between the

U.S. National Science Foundation (NSF), and

the Japanese Ministry of Construction (MOC)

and the Science and Technology Agency

(STA). The agreement was based on the

recommendations as outlined in the final report

of the U.S.-Japan Planning Group for the

program (Ref.l). The Cooperative Research

Program focused on the following four

building structure phases up to 1993, on the

line of the Planning Group report:

The first phase: reinforced concrete

structure (1979-1981)

The second phase: steel structure (1981-1984)

The third phase: masonry structure (1984-1989)

The fourth phase: precast concrete

structure (1989-1993)

Also, in the Planning Group report, research on Mixed Steel/Reinforced Concrete Structures was identified as an important phase of the

Program. After the recommendations, it has

become widely recognized that innovative uses

of two or more different materials in a structure

leads to more efficient system for resisting

seismic forces. In particular, during the past

ten years the use of composite and hybrid

structures has gradually increased both in the

U.S. and Japan. In spite of some research and

development work by Japanese construction

companies, not enough is known at present

regarding their seismic behavior and perfor- mance. Design procedures and codes for use

in typical design offices are, thus, currently

non-existent.

Looking at the importance of developing

design guidelines (a unified code development)

for typical composite and hybrid structures that are used in current practice, and of developing

new and innovative composite structural

elements and hybrid systems using advanced

new materials and/or devices, a five-year

research program on Composite and Hybrid

Structures was proposed by the Building

Research Institute (BRI) and the U.S.

interested professors in 1991 as the fifth Phase

of the U.S.-Japan Cooperative Earthquake

Research Program. Then, new recommendations on the new Phase were made on the

basis of a number of technical meetings of the

U.S. and Japan Planning Groups and a Joint

Planning Group Workshop held in Berkeley on September 10-12, 1992 (Ref.2). On the

basis of the Workshop, the first Joint

Technical Coordinating Committee (JTCC) on

the Composite/Hybrid Phase was held on 8-9

November 1993 at BRI, to discuss more

detailed research plans and summarize up-todate recommendations.

Further, considering the prospect of the

U.S./Japan Cooperative Research Program,

not only limited to seismic and wind effects,

some critical situations and needs raised both

in the U.S. and Japan should be taken into

Director of Structural Engineering

Department, Building Research Institute

(BRI), Ministry of Construction (MOC),

Tatehara-1, Tsukuba, Ibaraki 305, Japan.

369
account in planning the future cooperation.

Thus, in this paper, some thought and

proposals are given in this regard.

REVIEW OF THE PAST PHASES

A lot of technical knowledge and practical

information was obtained from the past

cooperative research phases. While elsewhere

we can refer to various papers and reports for

each research result and technical information,

the essence of significance and achievements

of the past Phases can be described as

follows:

(1) Through the first and second Phases, the

"Pseudo Dynamic Testing Method" applicable

to full-scale building structures completed.

After this success, the testing method become

one of major earthquake-resistance test methods; many research institutes of Japanese

construction companies, several U.S. universities and a few of developing countries have

introduced the pseudo dynamic testing

facilities.

(2) Since a full-scale test structure was

designed, tested and analyzed jointly by the

U.S. and Japanese sides in the first and second

Phases respectively, mutual understanding on

the difference between the U.S. and Japan in the design codes, practice and also even in the

way of thinking was improved.

(3) On the other hand, in the third and also in the fourth Phases, a common full-scale test structure between both sides was not reached,

because the discrepancy in the design codes

and practice between the two countries was so much large for determining a unified test building. However, from the viewpoint of

design and practice on each side, new

guidelines for design and construction have

been or are being developed, though the

practical results of the first two Phases were

referred to just modifying the then design

codes.

(4) In each Phase, active exchange of

personnels with short and long terms was

performed between both the countries. As the

effect of this activity, mutual understanding on

research and construction circumstances in both the countries was expanded and

deepened. Also through these exchangedpeople, associated man to man channels have

dramatically increased.

TOWARD A NEW STAGE

In one word, the past four Phases dealt with

examination and modification of existing

structural systems with conventional materials

such as ordinary steel and concrete. Also in these Phases, the technical validity of current codes and design practice was examined by

experimental and analytical manners, and

particularly in the latter two Phases, new

guidelines on design and construction

technology were proposed on each side.

Now, as previously mentioned, the current

Phase, Composite and Hybrid Structures, has

aspects somewhat different from the past

Phases. That is, the Phase will deal with new

structural members and systems by mixing two

or more different materials, and some research

topics will involve new materials and devices.

This slight but significant change of basic

characters of research contents means a

transition to a new stage of the cooperation.

Then, questions on the next step of the

cooperation would be raised as follows:

Based on the past and current achievements,

and experiences, what direction should the

future cooperation face to? And, what target

should the future program deal with?

Furthermore, what framework of cooperation

should we create for more efficient and

beneficial research and development? To

answer these questions, we should think about

the following items.

(1) Interactive Research between Structural

Engineering and Other Research Fields

The research on existing conventional

structural systems and materials, generally

speaking, has nowadays been less attractive as new topics of the future U.S. -Japan

Cooperative Research Program, since such the

research has almost reached to a "critical point"

where it becomes hard to attain significant

findings and innovative knowledge. This

means, in other words, that most significant

and principal problems have been solved.

Also, looking at various performance recently

required to building structures, such as

durability, fire-resistance, human comfort, life

370
Fig. 1 Interactive Cooperation with Other Research Fields

cycle cost and so on, existing structural

systems and materials cannot add function with

something new, or cannot give birth to high

performance based on today's diverse needs.

From this point of view, research on building

structures should be expanded into cooperative

research with other research fields (Fig. 1). In

this regard, the interactive research with the

field of material science and technology would

have the highest priority to stimulate and

improve structural engineering research.

(2) Increasing Roles of Private Firms

At least in Japan, during the past two decades

private companies have come to establish and

foster their own research laboratories or

institutes. Then, their ability of research and

development has much increased, and in many

areas, in particular related to construction or production, their ability and achievements

become superior to those of universities and

governmental research institutes. According to this circumstances, efficient research and

development can be done in cooperation with

private firms on equal terms.

In the Composite and Hybrid Phase, the above

thing is considerably taken into account, that is, research work and necessary funds are adequately shared among private sectors and

BRI. This type of cooperation will be much

more needed in a future new stage, not only in Japan but also in the U.S..

(3) Framework of Cooperation in Pursuit of

Mutual Benefits

The past Phases were performed under the

auspices of the UJNR Panel on Wind and

Seismic Effects . This is the formal framework

of the U.S.-Japan Cooperative Earthquake

Research Program; the UJNR Panel has been

useful and will continue to implement the

cooperative work. And, the Composite and

Hybrid Phase will be carried out also within

this framework. However, this framework is inflexible so far to incorporate private

companies and academic people into the Panel, on terms of equality. As previously

mentioned, considering the increasing

importance of roles of private sectors, a new

and expanded framework seems necessary to promote effectively the future new stage

cooperation where private firms can actively

participate on equal terms.

Regarding the framework, there is another

topic, that is, the U.S.-Japan Framework for A

New Economic Partnership / Positive

Cooperation in Global Perspective. The two

round-table talks at sub-cabinet level were held

in September and December 1993 to set up

Action Plans for the cooperation. In the broad

items for positive cooperation, "Construction

Technology" is included. To arrange Common

Agenda, BRI/MOC and NSF are working

together as the action offices. Here, our

371
Phase 1: RC

Phase 2: Steel

Phase 3: Masonry

Phase 4: Precast

Past

(1979-1993)

Phase 5

On Going Future

New Stage

Fig. 2 BRI Activities in US/Japan Cooperation

general agreement is that the cooperation could

be modeled after the UJNR except that it should be more broad-based under the new

cooperation program, minimizing any

duplication with the UJNR efforts. BRI ACTIVITIES

Now, considering the above situations and

NSF initiatives on recent programs, BRI will

move to the second stage of the U.S.-Japan

cooperation, with strong willingness, in the

future (Fig. 2). In this cooperation, the

contributions of BRI will be as follows:

(1) To fill the role as the host institution on Japanese side in the field of buildingrelated technology,

(2) To propose the evaluation system of

innovative technology, and

(3) To develop new comprehensive design/

evaluation methodology and system based on performance, which should be able to deal with innovative systems as well.

ment on building structures were discussed.

By these discussions, some proposals were

presented for the future new stage of

cooperation.

REFERENCES

1. "Recommendations for a U.S. -Japan

Cooperative Research Program utilizing

Large-Scale Testing Facilities," Report No.

UCB/EERC 79-26, September 1979.

2. "Recommendation for U.S.-Japan Cooperative Research Program -Phase 5

Composite and Hybrid Structures," Report

No. UMCEE 92-29, November 1992.

SUMMARY

The significance of the U.S.-Japan Earthquake

Research Program from 1979 to today was

briefly reviewed. Also, some recent change of

circumstances around research and develop372
NORTHRIDGE SOUTHERN

CALIFORNIA AND

HOKKAIDO NANSEI-OKI

EARTHQUAKES

The January 17, 1994, Northridge Earthquake, California

by

H. S. Lew 1 , Jim Cooper2

, Sam Hacopian3

, Walter Hays 4

, and Mike Mahoney5

ABSTRACT

The magnitude 6.8 earthquake occurred during the pre-dawn hours of January 17, 1994 provided a

crucial test for assessing our progress in earthquake

resistant design and construction over the past two

decades, following a similar magnitude event, the San

Fernando earthquake in 1971.

A reconnaissance team was organized by NIST

through the auspices of the National Earthquake

Hazards Reduction Program and the Interagency

Committee on Seismic Safety in Construction, to observe the damage, assess the performance of

various types of engineering structures, and document

the effects of the earthquake on the built environment: buildings, bridges, and lifeline systems.

This paper summarizes what we learned from the reconnaissance effort. More detailed documentation

has been presented in an NIST Special Publication,

862, "1994 Northridge Earthquake: Performance of

Structures, Lifelines, and Fire Protection Systems."

1. INTRODUCTION

A strong earthquake of magnitude 6.8 centered under

the community of Northridge in the San Fernando

Valley in California shook the entire Los Angles area

at 4:31 a.m. local time on Monday, January 17, 1994. January 17 was a federal holiday honoring Dr. Martin Luther King's birthday and, because of this and also the early morning hour, most non-residential buildings were empty and traffic was extremely light.

These fortuitous circumstances greatly helped limit the number of deaths and injuries.

As of February 14, there were a total of 58 deaths

attributed to the earthquake. About 1,500 people were admitted to hospitals with major injuries,

another 16,000 or so were treated an released.

Estimates of the number of people temporarily or permanently displaced because of damage to their houses or apartments ranged from 80,000 to 125,000.

Total loss from this earthquake was estimated to reach $30 billion, the costliest earthquake in U.S.

history.

Although the earthquake caused unprecedented

damage and disruption, it also created an unprecedented opportunity to learn about earthquake

mechanisms and effects. The earthquake triggered a

historically high number of strong ground motion

recordings. These recordings, coupled with

wide-reaching damage surveys and analyses of

specific structures, can provide a wealth of new

insight into and understanding of earthquakes and

their effects.

2. SEISMOLOGY AND GEOTECHNICAL

EFFECTS

2. 1 Seismology

The epicenter of the earthquake was located at 34°12'N, 118°32'W, about 30 km west-northwest of

Los Angeles in Northridge (Figure 1). The focal depth was estimated at about 15-20 km (USGS 1994

1. National Institute of Standards and Technology

2. Federal Highway Administration

3. Department of Housing & Urban Development

4. U.S. Geological Survey

5. Federal Emergency Management Agency

375
and EERJ 1994). The strong shaking lasted about 15 seconds in the epicenter area. The earthquake

occurred on an unidentified fault. The fault had no

distinct surface rupture.

Records from the main shock and aftershocks indicate

that the rupture had a thrust focal mechanism,

striking lOo north of west and dipping 30° to 45°

south (Hauksson 1994, EERC 1994, and EERI 1994).

Thousands of aftershocks were recorded in the following months; there were seven of them having

a magnitude 5 or greater as of March 20.

Over one hundred strong-motion instruments were

triggered by the event. In the epicenter area, peak

horizontal ground accelerations approached or exceeded lg in several locations (USGS 1994).

Figure 2 shows the locations of selected peak

horizontal and vertical ground acceleration recorded by the USGS's National Strong Motion

Program. The Los Angeles Building Code specifies a design acceleration of 0.4g. An accelerograph

recorded on the grounds of the Veterans Affairs Hospital in Sepulveda is shown in Figure 3. Note

that the transient peak accelerations shown were each

one-time events and not typical of the body of the shaking.

The high vertical accelerations recorded in numerous

locations have induced discussions among the engineering community over the need to consider

vertical accelerations in design. Currently, the model codes only require consideration of vertical accelerations in the design of cantilevered elements

and post-tensioned horizontal elements, and those

only in regions of high seismicity. In this event, a

free-field vertical acceleration of 1.1 8g was recorded

at Tarzana, about 6 km south of the epicenter, and

maximums of about 0.6g occurred at several other

locations.

2.2 Geotechnical Effects

The fracture of the thrust fault apparently did not reach the surface. No major surface faulting had

been identified as of late March. Landslides, rock

slides and slope failures were the most visible geotechnical effect caused by the earthquake. Soil liquefaction was widely reported, but apparently

caused little structural or agricultural damage.

Sand boils were reported along the Pacific coast from

Mugu Lagoon to the north to the Port of Los Angeles

to the south, and inland in the epicentral area, near

the junctions of interstate highways 5 and 210 and 5 and 405, and in areas around Simi Valley northwest

of the epicenter (EERI 1994 and EERC 1994).

Liquefaction-related surface cracking and vertical offsets were reported at all sites where sand boils were observed. At other sites, such as the Jensen

Filtration Plant in Sylmar and the nearby San

Fernando Juvenile Hall, no sand boils were reported

but lateral spreading and ground settlement indicate

that liquefaction occurred. Minor breaks in water, sewer, and other pipe systems were widespread in areas of lateral spread and liquefaction.

3. BUILDINGS

The earthquake provided the first full scale "test" of

modern (post 1970's) seismic building codes in this country. For the first time, a large and varied

population of buildings was subjected to ground

shaking equal to or exceeding that recognized in modern codes for design. The epicenter of the earthquake was located in a heavily populated

urban-suburban area, the San Fernando Valley

northwest of Los Angeles. In the epicentral region,

most buildings experienced ground accelerations equal to or greater than that upon which the code

design values are based.

The building damage caused by the 1971 magnitude

6.6 San Fernando earthquake in 1971 prompted

significant revisions to earthquake design

requirements. As a result, the 1976 Uniform

Building Code (UBC) is often specified as a "benchmark" code that ushered in "modern" seismic

design methods (FEMA 154, 1988). Comparisons of

building response in the 1971 and 1994 quakes give some insight into the efficacy of earthquake

mitigation practices that have been undertaken in the Los Angeles area in the intervening years.

The intent of building codes is to specify the minimum requirements needed for a structure to provide acceptable life-safety. Viewed in this

light, the earthquake demonstrated the success of

modern building codes. Of the 58 deaths attributed

to the earthquake, 22 were caused by structural

failures of buildings. The population of the three

affected counties (Los Angeles, Ventura, and Orange)

totals about nine million people. However, 80,000 to

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125,000 people were made temporarily or permanently homeless because of damage to their homes and apartments. Schools, hospitals, offices,

stores, and other commercial and industrial enterprises were forced to close due to damage, much

of it nonstructural. Viewed in this light, the earthquake demonstrated the limitation of modern

building codes. Because they do not include

postearthquake serviceability requirements for most buildings, they do not ensure preservation of

normal building function after an earthquake. Because they area not intended to prevent property

damage in a large earthquake, economic losses can be

high.

Building codes cover only new construction or changes to existing construction. The City of Los

Angeles established a requirement that all unreinforced masonry buildings (URMs) larger than

a specified size that were constructed prior to 1934

be assessed for seismic adequacy and, if necessary, be rehabilitated. This requirement, called "Division

88" led to the rehabilitation of all the older URMs,

certainly helped reduce the life loss from this earthquake.

Buildings of all ages and types were damaged in the quake. However, it also should be noted that many

buildings of all ages and types performed well. In

the epicentral area, from the California State University, Northridge campus, to the Northridge

Fashion Center mall, serious damage

occurred. Several pockets of severe damage

occurred at widely separated locations, such as downtown Santa Monica, along Ventura Boulevard

in Sherman Oaks, downtown Glendale, and

Hollywood Boulevard in Hollywood (Figure 4). The following sections present a few of the most

spectacular damages resulted from the earthquake.

3.1 Northridge Meadows Apartment Complex

Collapse in a single apartment complex caused

sixteen of the twenty-two deaths related to building

failures. The Northridge Meadows complex,

constructed in 1972, consisted of several three- story wood-frame buildings. About half of the units had open spaces for parking on the street level, typically around the perimeter of the complex.

Most of the interior courtyard facing units had

living space at the ground level rather than

parking. About half of the buildings in the complex

collapsed. The first story gave way while the second and third stories remained largely intact. The sixteen people killed were in the first story

apartments. All the buildings in the complex

suffered some damage (Figure 5).

The apartment buildings were predominantly

wood-frame,with wire mesh and stucco walls. Rows

of steel pipe columns supported the upper stories along the building perimeter between parking bays

(Figure 6). Transverse walls were typically spaced

about every three parking bays. Longitudinal walls

separated the parking areas from first story living spaces. The wire mesh and stucco coating on these walls provided insufficient resistance to the lateral earthquake forces generated by this earthquake.

The upper stories of the collapsed buildings were

displaced laterally and in some cases came to rest offset by as much as three meters from their original locations. The steel pipe columns, which

hinged at their bases, apparently controlled the distance of the offset as they bent to the ground

(Figure 7). The Northridge Meadows building

configuration, two stories of living space in a

long, narrow building above one level of parking,

is common in the San Fernando Valley. Many

buildings of this type sustained serious damage,

and some collapsed. These buildings were also

built in the late 1960's or early 1970's, when use of

stucco for seismic resistance in low-rise buildings was common. Current codes have essentially ban the use of this material.

3.2 Champagne Towers

Champagne Towers on Ocean Avenue in Santa Monica, a 16-story concrete apartment building,

suffered serious damage. The condition of this building illustrates two classic examples of

earthquake damage: damaged coupling beams

between linked shear walls and shear failure of

unintentionally short columns.

The transverse end walls are pierced by a single

vertical row of window openings. The wall segments between the windows act as coupling

beams linking the wall segments on either side. These beams suffered serious diagonal cracking

(Figure 8). In the longitudinal direction, the

lateral-force resistance is provided by concrete moment frames. The front of the building shows

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no sign of damage. At the back, the columns and

transverse wall ends are diagonally cracked (Figure

9) . The solid infill railings which rise to about

mid-height of the columns create an unintentional

"short-column" effect, causing the columns to fail in shear. At the front of the building, metal pipe

railings are provided at the balconies rather than solid

infill, as at the back. This may partially explain why

there is no apparent damage at the front of the building.

3.3 Hospitals

The original hospitals of Holy Cross and Olive

View were badly damaged in the 1971 San

Fernando earthquake. They were rebuilt following

the requirements of the Hospital Act introduced

by the State of California after the 1971 quake.

The structural systems of these two hospitals

performed well in the Northridge earthquake.

However, damage to the sprinkler and chilled water

systems at Olive View rendered the building

temporarily unusable. At Holy Cross, damage to the

air handling system forced closure of the hospital.

However, St. John's Hospital in Santa Monica was

closed due to structural damage. The 7-story

northwest wing, which was built in 1952, had severe diagonal cracks in the piers between windows (Figure

10) .

3.4 Bullock's Department Store

The collapse of large portions of the roof and

floors of the 3-story concrete Bullock's department

store at the Northridge Fashion Center could

have caused more deaths than any other single

building failure in this earthquake. Many multistory columns were left intact as the concrete slab

floors and roof fell on top of each to the basement, presumably due to punching shear

failure of the slabs. The remaining-standing

columns showed no evidence that slab reinforcement

had been continuous through the columns (Figure

11) .

3.5 Unreinforced Masonry Buildings

Reinforced masonry construction was introduced in 1934 after the 1933 Long Beach earthquake. Yet many pre-1934 unreinforced masonry buildings still exist today and all of them in the City of Los

Angeles have been retrofitted with through-bolts

and face plates to connect the masonry walls to the wood floor and wood diaphragms, as required by

the so called "Division 88 requirement (Figure 12). The City of Santa Monica doesn't have the same

rehabilitation requirements, nevertheless, most, if not all URMs have been retrofitted with abovementioned techniques.

Despite rehabilitation, many URMs in Santa Monica

and in Hollywood were badly damaged. As a

life-safety measure, rehabilitation was somewhat

successful in this earthquake. However,

rehabilitation was not successful at preventing

damage; portions of buildings disintegrated, top

stories and corners were particularly vulnerable

(Figure 13). Such damages could present a real and

serious threat to the safety of persons outside the

affected building.

3.6 Kaiser Permanente Office Building and

Barrington Building

The Kaiser Permanente office building on Balboa

Boulevard in Northridge was badly damaged. The

second story columns disintegrated, causing the upper

stories to collapse onto the first story, which

remained intact (Figure 14). The bare frame exposed

by the damage suggests that the buildings had a strong-beam/weak-column configuration (Figure 15). The damage appears to have initiated at the joints.

The L-shaped concrete frame Barrington Building

in West Los Angeles also had damage concentrated in the second story columns (Figure 16). The horizontal

architectural panels stiffened the lower half of the columns at each story, creating a short-column effect.

3.7 The American Savings Bank Building

A steel-frame office building on the west side of

Topanga Canyon Boulevard in Canoga Park, housing

the American Savings Bank, experienced significant

movement in its first story columns (Figure 17). A

nearly identical building on the same site, oriented at right angles to the first, had only minor damage to glass at the first story. The first story column

heights of the two buildings differed significantly: the ones in the undamaged building were considerably

shorter than the ones in the damaged structure.

In the weeks following the earthquake, reports surfaced about a surprisingly large number of

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damaged steel-frame buildings, which, as a class, have been viewed relatively earthquake-resistant. Failure modes reported include brittle fractures of

column flanges, base plates, and welds at connections; and buckling of diagonal tube bracing

(Los Angles Times, 2/27/94 and EERI 1994).

3.8 Parking Structures

Eight major public parking structures suffered

collapse or severe damage. Many other parking

structures suffered less spectacular, but nevertheless

serious, damage. The damaged parking structures were scattered throughout the affected area. A few

key damaged structures are described herein.

California State University at Northridge Campus

The four-level garage was built in 1991 of precast concrete with cast-in-place concrete slabs. The

lateral resisting system was an exterior perimeter

frame constructed of precast "Trees" (columns with

half-length beams cast as one piece) connected in the field to form a moment frame. The collapse

apparently started at the interior of the building

and the exterior perimeter frame was pulled over

towards the center of the structure (Figure 18),

suggesting that interior beams had lost their

vertical support, causing the floor plates to collapse and pull the exterior walls in with them.

Kaiser Permanente Medical Complex on Cadillac

Avenue The five-level parking garage in West Los

Angeles experienced a similar collapse. All four

facades were pulled inward, suggesting that interior damage precipitated the failure (Figure 19). The

garage was a post-tensioned, cast-in-place concrete structure with shear walls providing the

lateral-force resisting system at the east and west

ends, and moment frames at the north and south

perimeter. The shear walls were cracked

horizontally throughout their height due to the out-of-plane bending (Figure 20).

Trans World Bank Parking Structure The twolevel concrete parking structure adjacent to the Trans World Bank in Sherman Oaks had been

seismically strengthened after suffering serious

damage in the 1971 San Fernando quake. Flared

extensions were cast next to the interior columns

along their north-south axes (Figure 21), creating

short wing walls supporting the precast beams.

The unrehabilitated exterior columns were badly

damaged in the Northridge earthquake (Figure 22).

4. BRIDGES

4. 1 Overview

The vast majority of bridges in the Los Angeles

metropolitan area performed well in the Northridge

earthquake. However, the quake caused the collapses

of six major highway bridges and heavy damage to 157 other high bridges. The estimated cost to replace or repair these bridges is $1.5B. Figure 23

shows the locations of the bridges inspected by the reconnaissance team. This section provides a brief description of field observations of several major

bridge damages or collapses.

4.2 Interstate 5 (the Golden State Freeway) at Gavin Canyon Undercrossing

This bridge was designed in 1964 and constructed

in 1967. It carries Interstate 5 over Gavin Canyon,

about 3 km north of the intersection of

Interstate 5 and State Route 14. An aerial photograph of the damaged bridge, after demolition

had begun, is shown in Figure 24, and a ground level view is shown in Figure 25. The structure consists of two parallel bridges, with about a 66 degree skew

alignment. Each bridge is composed of five spans. Each bridge has four two-pier bents, and there is a

structural hinge near each of the two center bents. The superstructure consists of reinforced concrete box girders, and the central portion (between hinges)

is post-tensioned. The bridge survived the 1971 San

Fernando earthquake with virtually no damage. In 1974, the hinges were retrofitted with cable

restrainers. The seat length at the hinges is 200 mm,

which by current design standards would be

considered inadequate, especially for a structure of

this size and flexibility.

Despite the presence of cable restrainers at the hinges, failure of three of the four end spans of the bridge was apparently initiated by the spans falling off the hinge seats. The fourth end span, at the far western corner of the bridge, partially came off its hinge seat but did not collapse. After support was

lost at the hinges, a triangular portion of three of the end spans collapsed. It is likely that the skewed

alignment of the bridge was an important contributing

factor in this failure. The skewed alignment

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permitted rotation of the superstructure in a counter-clockwise direction, when viewed from

above. The tall, flexible piers of the two center bents would have offered relatively little resistance to rotation in a horizontal plane. Rotation of the superstructure would have resulted in differential displacements at the hinges, and eventually unseating

of the end spans at the hinges.

4.3 State Route 14 (The Antelope Valley

Freeway) Interchange with Interstate 5

(the Golden State Freeway)

Two bridges partially collapsed at this interchange:

the Route 14/5 Separation and Overhead Ramp C,

which is the ramp linking westbound SRI4 to southbound 15; and the North Connector

Overcrossing Ramp M, which is the ramp linking

westbound SR14 to northbound 15. In addition to the two collapsed ramps, there was evidence of

pounding between spans at several hinges, and

permanent differential offsets (both horizontal and

vertical) were observed between the ends of the spans, as shown in Figure 26. The cable

restrainers installed at hinges in this interchange

during the early 1970's may have been responsible

for preventing further collapses during the Northridge earthquake.

This interchange was designed in 1968 and was

under construction in 1971 when portions were damaged by the San Fernando earthquake. At that time, one of the completed ramps in the interchange collapsed, and two ramps which were

under construction were damaged. The portion

which collapsed in 1971 was the South Connector

Overcrossing, connecting southbound 15 with

eastbound SR14. A photograph of the collapsed

ramp, taken in 1971, is shown in Figure 27. This

ramp was later rebuilt, with improved pier

reinforcement, and it suffered no significant

damage in the Northridge earthquake. The

damaged portions of the other two ramps under

construction in 1971 were repaired in place, but

not strengthened; the portions not yet constructed were completed with limited seismic upgrading.

Both of the ramps which partially collapsed in the Northridge earthquake were under construction at the time of the 1971 earthquake.

4.3.1 Route 14/5 Separation and Overhead

^Ramp C

This ramp was designed in 1968 and constructed in 1971. The structure consists of multiple-cell concrete box girders, most of which are posttensioned, supported on single pier bents with

flared tops. The portion of the ramp which

collapsed was between Abutment 1 and the hinge

located near Pier 4. A ground-level view of the collapsed span is shown in Figure 28, and a view

of damaged Pier 3 is shown in Figure 29. The

mode of failure of this ramp has not yet

conclusively been determined, but the following

scenario appears likely. Because Pier 2 was the

shortest and stiffest pier in the structure, it probably attracted a large share of the horizontal

seismic forces. The reinforcing details of the pier

indicate that the pier contained quantities of

lateral reinforcement which would be considered

inadequate by current standards. Thus it is likely

that Pier 2 was damaged and collapsed, pulling the superstructure off the seat at Abutment 1. It was reported that Pier 2 had completely disintegrated

in the collapse. This observation is consistent with

the conjecture that the pier was initially severely

damaged in shear, and was then so badly weakened

that it was crushed in compression. Following the collapse of Pier 2, a failure of the superstructure was

initiated by excessive shear and negative bending

moment at Pier 3. The superstructure sheared off on

either side of Pier 3, leaving the pier standing nearly

intact, as shown in Figure 29. The span between

Piers 3 and 4 then collapsed, pulling the superstructure off the hinge seat near Pier 4.

4.3.2 North Connector Overcrossing Ramp M

This ramp was designed in 1968 and constructed in 1971. Photographs of the collapsed portion of the ramp are shown in Figures 30 and 31. The portion

of the ramp which collapsed was located between

Abutment 1 and Pier 3. This ramp was under

construction in 1971 and was nearly completed when

the San Fernando earthquake occurred, with only the portion from the hinge near Pier 8 to Abutment 11 remaining to be constructed. The portion of the ramp

which collapsed in the Northridge earthquake,

between Abutment 1 and Pier 3, was already constructed when the San Fernando earthquake

occurred. Only minor damage to this ramp was

reported following the 1971 earthquake (State of

California, 1971). This damage consisted of

permanent offsets at the hinges near Piers 4 and 6.

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The collapse in the Northridge earthquake appears to have been initiated by the failure of Pier 2. After

Pier 2 failed, the simply supported span between

Abutment 1 and the hinge near Pier 2 collapsed, and,

as shown in Figure 31, the superstructure failed in bending near Pier 3 due to a large negative moment.

Pier 2 completely disintegrated, as shown in Figure

32. This is the shortest pier in the structure, so it is likely that it attracted a high level of lateral force, was damaged in shear, and finally collapsed in compression. Pier 10 is similar in height to Pier 1, but the quantities of lateral reinforcement in Pier 10 are greater than those in Pier 1. This is because

Pier 2 had already been completed, but Pier 10 had

not yet been constructed, when the 1971 earthquake

occurred. Following the 1971 earthquake the plans

for Pier 10 were revised to provide increased lateral reinforcement.

4.4 State Route 118 (The Simi Valley

Freeway)

SRI 18 is the major east-west transportation route for northern Los Angeles County. The route passes just north of the epicentral region of the quake. A

number of bridges along SRI 18 received minor,

repairable damage, but two bridges were damaged

severely: portions of a bridge collapsed at the

intersection of San Fernando Mission Boulevard and

Gothic Avenue; and nearby there was severe pier

damage and a near collapse of the bridge at Bull Creek Canyon Channel. The performance of these two bridges is described below.

4.4.1 State Route 118 (Simi Valley

Freeway) at San Fernando-Mission

Blvd. and Gothic Ave

Because the Mission-Gothic Undercrossing was

designed in 1972, after the San Fernando

earthquake, the bridge contains seismic details which are improved over the details used in designs

of the 1960's and earlier. Most notably, the spiral

hoops in the bridge piers are spaced closely

together. Nonetheless, the piers of the Mission-Gothic Undercrossing suffered severe damage in this quake, and one of the two parallel spans partially collapsed. Damage to piers at Mission-Gothic is shown in Figures 33 and 34. In addition to pier damage, there was also severe damage to the abutments, including cracked wing

walls, settlement of soil behind the abutments, and

apparent shifting of the abutments.

The likely failure mode of the bridge was that the

piers initially suffered severe shear damage, followed

by crushing of a number of piers in compression.

The structure may also have shifted to the southwest,

as constrained by the geometry of the abutments.

Although unseating at Abutment 5 did occur, this was

probably a secondary effect, as the seat lengths were

generous by current standards. Rather, failure was

probably initiated by damage to the piers. In this case the failure of piers has important theoretical implication, as the piers of this bridge contained

quantities of lateral reinforcement similar to those required by current seismic design standards. Therefore, further research will be required to determine why the pier reinforcement was inadequate

in this case.

4.4.2 State Route 118 (The Simi Valley

Freeway) at Bull Creek Canyon

Channel.

The bridge did not collapse completely, but many of

the piers were severely damaged, as shown in Figures 35 and 36. In addition to pier damage, there was significant damage to the abutments, including

cracked wing walls, settlement of soil behind the abutments, and possible shifting of the abutments.

This bridge was designed just after the 1971 San

Fernando earthquake, so it contains some seismic

detailing which is improved over earlier bridge

designs. However, since it was not until several years after the San Fernando earthquake that seismic

design codes for bridges were significantly changed,

the design of this bridge would be considered

substandard by current design codes.

All ten of the piers in Bent 3 failed in combined

shear and compression near their bases, as shown by

Figures 35 and 36. In Bent 2, Piers on the south side of the span failed near their top end, and the damage

became progressively less for piers to the north in Bent 2. The farthest north piers in Bent 2 showed

relatively little damage - only diagonal shear cracks near the base, with no spalled concrete. This uneven

distribution of damage to piers was probably caused

by the asymmetric plan geometry of the bridge.

The confining reinforcement in the piers is inadequate

by current standards. All pier failures occurred in

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the zone of wide spiral spacing. Another factor contributing to pier failure was that the effective lengths of the piers were much shorter than the distance between the footings and the girder soffit. The lengths of piers varied between roughly 6000 and

9100 mm. In Bent 2, the effective pier lengths were

decreased roughly one-third by compacted backfill

soil above the pier footings. In Bent 3, the pier

lengths were also decreased roughly one-third, but by

a reinforced concrete channel liner wall, which was

cast abutting the piers. The shortened effective pier

length resulted in stiffer piers, which were more

susceptible to shear-dominated failure than

bending-dominated failure.

4.5 Interstate 10

Interstate 10 is a major east-west artery running

between Santa Monica and downtown Los Angeles.

The freeway was constructed in 1966. Major bridge

collapses occurred at two locations: the La Cienega

Blvd. -Venice Blvd. Separation and the Fairfax Ave. -Washington Blvd. Undercrossing.

4.5.1 Interstate 10 La Cienega Blvd. /Venice

Blvd. Separation

Figure 37 is a ground-level view of the portion of the westbound lane that collapsed to the ground at the hinge located between Bents 7 and 6. The pier

in the foreground is the northernmost pier of Bent 7.

It is seen that there was extensive column shortening

due to the failure of the lateral reinforcement to provide adequate confinement of the core concrete. Fractured lateral reinforcement can be seen to the

right of the failed portion of the pier. Figure 38

shows the other two piers of Bent 7 which supported

the westbound lane. Extensive failure and column

shortening is evident. The concrete block wall behind the piers is part of a storage building that was constructed beneath the bridge structure. The

storage building extended from Venice Blvd. to La

Cienega Blvd. The piers of Bent 7 which supported

the eastbound lane suffered relatively minor damage,

and the roadway to the east of Bent 7 remained

largely at its original elevation.

Figure 39 shows the northernmost pier of Bent 6, located just to the west of the hinge. This pier

totally disintegrated, but the storage building

prevented the roadway from collapsing to the ground.

Figure 40 shows the southernmost pier supporting the

eastbound lane at Bent 3. The top of the pier is severely damaged and it is likely that the storage

building prevented the total collapse of the roadway.

In summary, the bridge structure spanning La

Cienega Blvd. and Venice Blvd. was extensively

damaged. The failure is attributed to the small amount of lateral pier reinforcement. As the piers

cracked due to the lateral loading, the lack of

adequate confinement resulted in a reduction of the

vertical load capacity because of core concrete loss and buckling of the longitudinal bars. Had it not been for the storage building located beneath

the bridge, it is likely that more spans would have

collapsed to the ground.

4.5.2 Interstate 10 Fairfax Ave.- Washington Blvd. Under Crossing

The earthquake caused partial collapse of two spans of the eastbound and westbound lanes on either side of Bent 3. The piers in Bent 3 had failed and caused

the girders to sag at this location. However, cable

restrainers at the hinge west of Bent 4 prevented the girders from falling off of the hinge seats (Figure

41). Figure 42 shows the westbound lane of 110 over

Fairfax Ave. It can be seen that the piers in Bent 3 shortened and caused the girder to sag. It can also be

seen that the span to the west of Pier 2 lifted off of

the abutment. Figure 43 is a close up view of one of

the piers in Bent 3. The failure mode is similar to that observed at Venice Blvd. and La Cienega Blvd.

When the piers were subjected to lateral ground

motions, the lateral reinforcement was not able to adequately confine the core concrete. As a result of

lateral load damage, there was a loss of axial load

capacity and the piers shortened under the action of

the vertical loads. As mentioned, the piers at Bent 4

had considerably more longitudinal reinforcement

than the other piers. Examination of the piers in Bent 4 revealed varying degrees of diagonal cracking

and spalling of the concrete cover. At the time of the

site visit, these piers were surrounded with wooden

shoring as a precautionary measure during the demolition of the collapsed spans.

5. LIFELINE SYSTEMS

5. 1 Introduction

Lifelines include water, sewer, gas, fuel, electric power, telecommunications, and transportation

382
systems. These systems are critical to the vitality of

the built environment and the functioning of modern

society. They provide services to the community to maintain its safety, health, financial transactions, and

to stimulate economic activities. Damaged lifelines can impede emergency response following an

earthquake, and can hinder postearthquake recovery. Disruption of regional lifeline systems due to a major

natural disaster such as an earthquake can have a profound effect on the entire nation because of the economic interdependence of lifeline systems and the functions they support.

This section offers a brief overview of the performance of lifeline systems, including

observations made by many investigators, but with

particular emphasis on the sites that the team visited personally.

5.2 Water Supply Systems

In addition to limited sources from the region's local groundwater basins, the main water supply is from

Northern California and the Colorado River. January

17, 1994 was the first time in history that an

earthquake resulted in the breakage of all four

pipelines that feed water to the region's three water

treatment facilities.

Compared with the extensive damage caused by the 1971 San Fernando earthquake to the Jensen

treatment plant in Sylmar (under construction at the

time), the 1994 earthquake caused only minor

damage. The 1994 damage included lateral spreading

of the ground or soil settlement around the facilities, leaks in pipelines, and leaks at construction joints.

While water supply to these facilities was available once the major pipelines were repaired, the system

nevertheless failed to provide water to customers

because of the failure of the water supply distribution network, especially the network serving areas near

the epicenter. Thousands of main line leaks were

reported and the repairs were very time-consuming.

Among the numerous situations that contributed to the disruption of the water supply system was damage to one of the four damaged pipelines, Los Angeles

Aqueduct No. 2 at Terminal Hill. Aqueduct No. 2 is made of 2. 1-m diameter steel pipes. Terminal Hill is located about 20 km north-northeast of the epicenter, southeast of the intersection of 1-5 and SR14. There

is no strong motion record available for the site. However, two stations maintained by the California

Division of Mines and Geology (CDMG) and the U.S. Geological Survey (USGS) about 5 km south of

the site showed peak ground accelerations over 0.9g.

A steel pipe brings water up from the canyon below

(Figure 44). The pipe is supported on concrete

saddles built along the mountain slope. In a few

places, the pipe separated from the saddles creating

50 to 80 mm gaps. In at least one location, the pipe

crashed into the saddle. In two other locations, the pipe sections bulged 80 mm and 150 mm

respectively. However, no rupture or leakage was

noticed. The ruptured section of the 2. 1-m pipe is located near top of the hill. The two sections where

the rupture occurred were connected together using a mechanical coupling system (Figure 45). To restrain the relative movement between the two pipe sections during strong shaking, the sections were connected by

eight pairs of restrainer rods 35 mm in diameter and

2.2 m long. The rods were attached to brackets,

which were welded to the pipes. These welds broke

due to strong shaking and resulted in the separation

of the two pipe sections.

Repair of the pipe sections began immediately after the earthquake. The repair work was completed in the evening of January 19 and the aqueduct started operation at 2:00 a.m. January 20. However, leakage was found shortly thereafter which caused the operation to shut down again. Excavation was

required to repair two other sections. The pipeline was back in operation on January 21.

There were numerous breaks of water lines during

this earthquake. Two water mains were ruptured at a site along Balboa Boulevard adjacent to the rupture

of a 0.6-m gas main. A major fire resulted, which is described later in this section.

5.3 Gas and Liquid Fuels

Natural gas systems consist of transmission,

distribution, and service lines. In the earthquake-affected area, transmission lines are steel pipes with diameters ranging from 0.3 to 0.8 m. Most of the failed lines were of pre-1971

construction. The distribution lines are either steel or

plastic pipes. Compared with the effects of the 1971

San Fernando earthquake, this earthquake resulted in more ruptures in the distribution lines than in the

383
transmission lines. Most of the breaks occurred to old steel pipes. Plastic pipes used in the distribution system seemed to perform well.

As happened in earlier earthquakes such as the 1987

Whittier Narrows earthquake and the 1989 Loma

Prieta earthquake, tens to hundreds of thousands of

gas supply outages occurred. Most outages were the

result of customers shutting off gas valves because of

fear of gas explosions or fires. While immediate

safety is the intent of this action, it can result in a long delay before service is restored because gas company technicians must test each system before

turning the valve back on.

The rupture of a 0.6-m gas main occurred along

Balboa Boulevard between Rinaldi Street to the south

and Lorillard Street to the north, about 0.5 km north

of the intersection of Balboa Boulevard and SRI 18 (Figure 46). Along this stretch, compressional

ground failures occurred between Rinaldi Street and

Halsey Street, and extensional ground failures occurred between Bircher Street and Halsey Street.

Three pipelines ruptured as the result of ground

contraction; a 150-mm gas distribution line, the 1.8-m Rinaldi trunk water line, and a 0.6-m gas main

(Figure 47). The Rinaldi trunk line was of post- 1971

construction and the gas main was of 1930 vintage.

These pipes showed a shortening of 125 to 150 mm.

About one block north, the ground extension caused

rupture of the same 0.6-m gas main and a 1.2-m

water main (Figure 48). The rupture of the gas main

caused a major fire at the site that destroyed five houses (Figure 49). The loss of both water mains at this location and the difficult access to the site due to the fire and flooding in the street made fire fighting

difficult.

An excavated section just north of the fire site revealed some additional underground lines (Figure

50). The 460-mm crude oil line in the middle of the photo performed well. The 150-mm gas distribution

line shown is a replaced section.

In addition to the above damage in the epicentral

area, the strong shaking from the earthquake cracked

welds at several locations along a 250-mm pipeline

transporting crude oil to refineries from the San

Joaquin Valley. As a result there was an oil spill along the Santa Clara River.

5.4 Electric Power

Power was lost to most of the Los Angeles basin area

after the earthquake. Nearly 2 million customers were out of service immediately after the quake.

About half of them had power restored within one

day and over 95 percent had power restored by

midnight Tuesday, January 18. All power was

restored within ten days after the earthquake.

Some transmission towers suffered significant

damage, many as the result of foundation failure. Damage to several high voltage substations near the

epicenter, such as Sylmar, Pardee, and

Rinaldi, led to the widespread power outage in the Los Angeles basin, as well as the isolated outages

throughout seven western states (Figure 51). As

happened during the 1971 San Fernando earthquake,

porcelain elements of equipment of 230 kV and 500

kV classification suffered the most damage (Figure

52). This highlights the urgent need for developing new materials that would be more earthquake

resistant to replace porcelain, which is very brittle. Furthermore, most of the 230 kV circuit switchers,

similar to those in the forefront of Figure 53, at the Sylmar substation, were damaged. However, none were damaged during the 1971 earthquake, a possible

indication of much stronger ground shaking by the Northridge earthquake. Most of the capacitor banks,

similar to those in the background in Figure 53,

performed well in this earthquake, whereas most

collapsed during the 1971 quake. Better performance

of the capacitor banks is the result of higher seismic

requirements in equipment qualification and in

installation practices.

The level and types of damage to electrical systems

experienced in the Northridge earthquake demonstrate

that many valuable lessons have been learned since

the 1971 earthquake. Many improvements have been

made in areas such as requirements for equipment

qualification and installation practices.

5.5 Transportation

All airports in the region survived the earthquake

with no major problems. The airports were shut down immediately after the quake as a precautionary move to allow for the inspection of runways and

taxiways. All airports were re-opened for operation once the inspections had been completed. No

structural damage was observed in airport facilities.

384
However, they did suffer some typical types of

non-structural damage, such as fallen ceiling tiles and

leakage of water pipes.

There was a freight train derailment in Northridge

during the earthquake (Figure 54). The 64-car

freight train belonged to Southern Pacific and was on

its way from Houston to Sacramento. Twenty-five

cars derailed, sixteen of which carried sulfuric acid

or diesel fuel. The derailment resulted in the spill of

30,000 liters of sulfuric acid and 7,500 liters of diesel

fuel. There were no casualties in this incident. About 200 m of railroad tracks were replaced

immediately following the earthquake and rail service was restored at 2:00 a.m., January 19. Removal of

damaged cars and cleanup of debris were completed on January 21.

Strong ground shaking and lateral movement of

subgrade materials resulted in numerous ruptures of

asphalt pavement and concrete sidewalks in the

epicentral area (Figure 55). In some instances, local

traffic was interrupted temporarily waiting for the

repair of these cracks. In most cases, the damage was

minor and local traffic was not interrupted. The

south approach of Balboa Boulevard at SRI 18 caved

in from loss of abutment fill due to a water main

break. This bridge was closed for traffic due to some structural damage to the piers and the loss of

the south abutment fill.

6. POSTEARTHQUAKE FIRES

6.1 Introduction

The Northridge earthquake resulted in fires which

challenged the resources of the fire service due to the number of fires, disruption of the water supply, and

damage to fire protection systems within buildings.

The majority of the estimated 30 to 50 significant

fires were located in the San Fernando Valley and

confined to the building of fire origin either by

separation or by fire department action. Fortunately

there was no loss of life from fire. A principal cause

of the fires involved natural gas leaks. A small number of fires were caused by hazardous chemical

interactions. The only major instances of

building-to-building fire spread occurred in three manufactured housing developments (mobile home

parks). Fire incidents occurred at a greater than normal rate in the clays following the earthquake with

the cause of some of the fires directly attributable to

the restoration of power and gas to buildings shaken

in the initial earthquake and aftershocks. Fire

sprinkler systems sustained damage in some buildings

although the number and extent of damage is not known at this time.

Fire protection in the municipal environment is derived from private and public systems including

building construction, building fire protection

systems, land use, public and private water supplies,

public and private fire departments, and

communication and utility systems. In the aftermath

of a major earthquake the normal interactions between these systems is disrupted. Even though

emergency operational plans exist, the interaction between these systems in reducing the loss from fire

is complex since it involves decisions on the part of

a great many people.

The loss of life and property caused by fire occurs in a different time frame than the structural and property

damage caused directly by the earthquake. While

most of the loss caused by shaking occurs during the time of ground movement, there is basically no fire loss during that time. Fire loss directly attributable

to the earthquake begins immediately following the earthquake and can continue for days after the ground movement has stopped.

This section examines the factors contributing to the cause, spread of and loss from fire in selected buildings affected by the earthquake.

6.2 Fire Events Following the Earthquake

Immediately following the earthquake the Los

Angeles City Fire Department initiated the Earthquake Operational Mode which included placing

emergency equipment on patrol throughout the city

and dispatching fewer pieces of equipment to each

incident in order to accommodate the increased

number of incidents. Earthquake damage was wide

spread but occurred mostly within the City of Los

Angles and most of the fire incidents were within the San Fernando Valley. Immediately following the earthquake electrical power was lost and telephone

service disrupted throughout the city. At 0545 hours

Mayor Riordan declared a state of emergency and by

0645 as many as 50 structure fires had been reported

and over 100 incidents were being handled by the fire department. By 0945 all fires were under control.

385
Although the earthquake was centered within the City

of Los Angeles, there was damage in surrounding

counties and emergency resources throughout the region were utilized. In addition to responding to

fire incidents, the fire department provided

emergency medical, hazardous materials, and urban

search and rescue services. The Los Angles City

Fire Department responds to over 900 fire, medical,

and other emergences on a typical day. This number

increased to over 2200 on the day of the earthquake

and remained at twice the normal level in the following days. The continued high number of

incidents was due in part to fires associated with the restoration of utilities.

Water available for fire fighting was generally

adequate in the San Fernando Valley area during the day following the Earthquake. The exceptions were

in areas near the boundaries of the system

and in areas at higher elevation. In the hours

following the earthquake pressure in the water system

dropped due to disruptions in supply and more than 3000 leaks. By the day after the earthquake, water

tankers had been deployed throughout the San

Fernando Valley to assist in fire fighting operations.

On January 20th fire department pumpers were used

to pump water from areas of adequate pressure within

the system to areas with low pressure.

6.3 Fire Causes

The 30 to 50 fires reported initially following the earthquake occurred in a variety of residential and

commercial occupancies. The majority of buildings

in the San Fernando Valley are four stories or less in height and therefore the fires occurred primarily in these types of buildings. Figures 56 to 59 show

some examples of buildings involved in postearthquake fires. The fire in the rear of the apartment complex shown in Figure 59 was the only

fire observed in completely collapsed portion of a

structure.

Preliminary indications are that a significant number

of fires were associated with natural gas leaks. Natural gas is not in itself a source of ignition but it

is relatively easy to ignite especially in confined

spaces. Natural gas is the predominant fuel used for space and water heating in the Los Angeles area. Although electrical power was lost throughout the area immediately after the earthquake, the most likely source of ignition was a combination of electrical

sources and flames in the gas appliances themselves. Gas leaks occurred both inside and outside of

buildings. The fire resulting from a leak in a gas main under a street destroyed several nearby houses. Figure 60 shows a gas meter in a manufactured

housing development (mobile home park) in which

there were at least six individual ignitions. Although

there was not a leak in the gas service shown, it demonstrates how the movement of the manufactured

homes following the earthquake damaged the gas

service resulting in leaks which were ignited by

unknown sources. There were no postearthquake

fires in the newest section of the development where

an improved gas service design was used. As in past

California earthquakes, water heaters appear to be a source of gas leaks (Mohammadi, et al, 1992).

Inadequately secured water heaters are the gas

appliance most likely to tip over during an

earthquake. The fire which destroyed the multi-family housing unit shown in Figure 61 was

reported to have started as the result of damage to a water heater. An undamaged housing unit similar to the one which was destroyed can be seen in the background. Even though natural gas leaks may have

played a role in a significant number of the 30 to 50

reported fires, this number is very small when

compared to the total number of buildings exposed to significant shaking as a result of the earthquake.

Since residents of this area are aware of the dangers

of gas leaks following an earthquake many fires may

have been averted by individuals shutting off the gas

to buildings or appliances.

The rapid failure of the electrical power distribution system probably resulted in fewer fires than might

have been expected. As electrical and gas service was restored in the days following the earthquake a

significant number of fires were reported. Some of

these fires were a result of earthquake damage to

electrical and gas equipment which went unnoticed or

unattended. Electrical and gas service was not disconnected in all red tagged buildings which were

identified by authorities as unsafe to enter. These

and additional fires may have all occurred

immediately following the earthquake if electrical service had been maintained.

As in past California earthquakes, a small number of

fires appear to have been caused by flammable liquid or chemical spills. Figure 62 shows a science

building at the California State University at Northridge in which the fire was reported to have

386
been the result of a chemical spill. No fires were known to have occurred at service stations. Since the earthquake occurred when most people were asleep,

fire causes such as overturned candles and barbecue

grills and fires associated with industrial processes appear to be nearly nonexistent.

A small number of wildland fires were attributed to earthquake related causes, most likely arcing in overhead power lines. Since the wind was light and

the vegetation was not excessively dry, these fires were easily extinguished.

6.4 Performance of Fire

Protection Systems

Some damage to fire sprinkler systems has been

reported but the full extent is not known. Damage to these systems is frequently the most visible of the fire protection systems since it may result in water leaks. Interviews with persons who have entered buildings

in the earthquake area indicate that many sprinkler

systems remained intact particularly those installed in accordance with latest seismic standards. Typical

damage to fire sprinkler systems included broken

pipes due to differential building movement or the sway generated in long pipe runs without adequate

bracing. Sprinklers installed in the downward or pendent position from piping above ceilings were in some cases sheared off.

In other cases pendent sprinklers installed in drop

ceilings were pulled through the ceiling by the upward movement of the pipes and punched new

holes in the ceilings during the downward movement.

While the punching may not have resulted in leaks,

it damaged the sprinkler deflectors which generate the desired spray pattern.

Damaged deflectors usually result in a significant decrease in sprinkler performance requiring the sprinklers to be replaced.

Sprinkler systems normally have one or more check

valves which prevent water from flowing from the

sprinkler system into the water supply system. In the most common wet pipe system, the pressure in the

sprinkler system piping will be the highest pressure

attained in the water supply system over time. In the days following the earthquake, as the fire department

pumped water from one part of the municipal water

supply system to another, there were significant local

increases in pressure. These higher pressures were

then "trapped" in the sprinkler systems. Although it

is probably not a major problem and can easily be

remedied by bleeding off the pressure, these higher

pressures could lead to premature failure of the system and reduced effectiveness at the time of

activation until the system returns to the design

pressure.

Damage to fire alarm, detection, smoke control, other extinguishing systems, and passive building fire protection systems such as fire and smoke barriers has not been reported at this time. The

disruption of land based communication systems is reported to have affected the ability of systems to dispatch alarms.

7. CONCLUSIONS

7.1 Introduction

The Northridge earthquake caused severe damage to a wide range of structural types because the epicenter was located in a populated urban area. Damaged

structures revealed a number of deficiencies in current construction practices and areas needing

improvements in code provisions. Implementing

lessons learned about structural performance and

postearthquake fires will reduce seismic hazards

throughout the United States.

7.2 General

The Northridge earthquake claimed 58 lives and caused over 1500 serious injuries.

However, fewer than half of the deaths were

attributed directly to structural failures. Because it occurred at 4:31 a.m., on a

holiday, life loss was limited to a small number. Had the earthquake occurred

during business hours, the collapse of

parking structures alone could have caused

a large number of deaths.

Although a number of fires started immediately following the earthquake, calm

winds limited spreading of fires in residential and commercial districts. At

mobile home parks, fires spread from unit to unit. In most instances, fires were caused

by natural gas leaks.

387
Damage to multi-family dwellings

contributed significantly to the over 25,000

dwelling units that became unhabitable.

Providing adequate shelter for displaced persons is a major task after an earthquake.

Special attention should be paid to improving the seismic performance of

existing dwelling stock.

At many locations, peak ground acceleration exceeded 0.4g, the maximum design value

in building codes. However, most buildings

met code expectations for performance. Because many strong motion records and

response measurements are available from

this earthquake, valuable opportunities exist for in-depth studies of building performance

to assess the adequacy of design values for earthquake forces and provisions for seismic

resistance.

Building Performance

In many cases, buildings designed and

constructed in accordance with modern

(mid-1970's or later) seismic requirements

performed well structurally. This clearly

shows the value of incorporating modern

seismic design and construction requirements

into building codes. However, failures of

structures characterized as "Undefined

Structural Systems" (UBC 2333(i)2), such as the parking garage at California State University at Northridge, indicate that the performance of such structural systems

needs to be evaluated carefully to update

code provisions.

Damage to unreinforced masonry (URM)

buildings was widespread. In most cases, those URM buildings rehabilitated with

parapet braces and floor-wall ties, as those

rehabilitated in response to the Los Angeles

division 88 ordinance, escaped total collapse. However, walls sustained severe cracking and, in many cases, pieces fell onto

sidewalks. This posed life-threatening

hazards to pedestrians.

Nonstructural damage caused hospitals,

schools, businesses, and industrial facilities to be inoperative even though structural

damage was minimal or non-existent. An

in-depth review of current code requirements

and standards for nonstructural elements is needed to improve their seismic

performance.

Damage to steel structures was not readily

visible from the exterior of buildings

because most steel members are hidden

behind architectural finishes and fire proofing. Removal of such coverings

revealed brittle failures of welds and

connections in many steel frame structures. Because owners of damaged structures often wish to keep damage reports confidential,

damage information is not forthcoming.

Detailed failure analyses of available data should be performed to understand the underlying causes of these failures.

7.4 Bridge Performance

In general, bridges designed using standards developed after the mid- 1 970' s performed

well. Several bridges near the epicentral

region, which were designed and constructed

in the 1960's and early 1970's, sustained severe damage. Of the seven major bridges

which sustained severe damage, six failed due to inadequate lateral reinforcement of

the bridge piers.

Most older bridges which had been

seismically retrofitted (with cable

restrainers, pier jacketing or foundation

strengthening) performed well in the Northridge earthquake. However, because

this earthquake was relatively moderate in magnitude, it should not necessarily be

concluded that all seismic retrofit methods

for bridges have been adequately proof

tested by this event. Design criteria for cable restrainers may need to be reviewed.

Even though jacketed bridge piers apparently

performed well in this earthquake, it would

be valuable to study the intensity of ground

motions experienced at the specific bridge

sites where jacketed piers have been

employed, so that the performance of

jacketed piers in larger events can be

estimated.

388
It is likely that some bridge piers were damaged because their effective lengths had

been reduced. Several factors reduced the

effective lengths: architectural flares at the tops of piers; backfill soil over a portion of

the pier height; and concrete walls cast integrally with, or directly abutting, piers.

Such constraints need to be minimi zed, or

the effects of the constraints must be

considered carefully in the design of piers.

In bridges with piers of varying heights, the

shortest piers tended to sustain the most

damage. This is apparently because short

piers have high lateral stiffnesses, and

therefore attract a large share of the seismic

loads. A better understanding of the role of

short piers in overall bridge performance, is needed.

Bridge spans which had skewed alignments

or irregular plan configurations often sustained severe damage. Special attention should be paid to the potential for problems

with bridges having skewed alignments or

irregular plans.

Older bridge bearings, of the steel rocker

type, are highly susceptible to damage

during strong ground shaking. Loss of

rocker bearing support can lead to a broad

range of bridge damage, including cracked

girders, loss of roadway elevation

alignment, and complete collapse of the superstructure.

Performance of Lifelines

Damage to older trunk lines and main lines for water distribution caused serious disruptions in water supply in the epicentral

region. About 50,000 Los Angeles

Department of Water and Power customers were without water on the first day after the earthquake. About 10,000 customers were

still without water one week after the earthquake.

Buried pipelines which carry natural gas and

oil fractured at many locations due to ground

motion. Over 1300 breaks and leaks in the gas piping system were reported.

As in the past earthquakes, brittle ceramic

elements, which are often the weak links in

circuit breaker assembles, were damaged.

7.6 Fire

The occurrence of the earthquake before the

start of morning traffic on a federal holiday

allowed the fire department to respond to

fires promptly without delays.

Damage to natural gas pipelines and

appliances resulting in leaks contributed

significantly to postearthquake fires.

In many cases, electric power was restored

to buildings which were identified as unsafe

for entry. The desire to restore utility service as quickly as possible is at odds with

the desire not to cause fires.

In general, fire sprinkler systems designed

and installed in accordance with the latest seismic standards withstood the earthquake

with little damage. However, in some

instances, sprinkler pipes ruptured where

sprinkler systems interacted with suspended

ceilings.

8. References

1. EERC (1994), Jack P. Moehle, editor. Preliminary Report on the Seismological and

Engineering Aspects of the January 17, 1994

Northridge Earthquake, Earthquake

Engineering Research Center Report No.

UCB/EERC-94/01, University of California,

Berkeley, CA.

2. EERI (1994), John F. Hall, editor. Northridge Earthquake January 17, 1994,

Preliminary Reconnaissance Report,

Earthquake Engineering Research Institute,

94-01, Oakland, CA.

3. FEMA (1988). Rapid Visual Screening of

Buildings for Potential Seismic Hazards: A

Handbook, FEMA 154, Federal Emergency

Management Agency, Washington, D. C.

4. Hauksson, E., Jones, L., Mori, J., Heaton,

389
T., and Hutton, K. (1994). Aftershocks of

the Magnitude 6.6 Northridge, California

Earthquake of January 17, 1994, Southern

California Seismographic Network,

California Institute of Technology,

Pasadena, CA.

5. LA Times, February 27, 1994

6. Mohammadi, J., Alyasin, S., Bak, D.N.,

Investigation of Cause and Effects of Fires

Following the Loma Prieta Earthquake,

Illinois Institute of Technology, IIT-CE-92-

01, May, 1992, 181 pages.

7. State of California (1971), "The San

Fernando Earthquake: Field Investigation of

Bridge Damage," State of California,

Business and Transportation Agency,

Department of Public Works, Division of

Highways, Bridge Department, 209 pages.

8. USGS (1994), Porcella, R.L., Etheredge,

E.C., Maley, R.P., and Acosta, A.V.,

authors. Accelerograms Recorded at USGS

National Strong-Motion Network Stations During the M3 = 6.6 Northridge, California

Earthquake of January 17, 1994, U.S.

Geological Survey Open File Report 94-141.

390
Figure / The epicenter of the January 17, 1994 Northridge earthquake was centered about 30

km northwest of Los Angeles in the San Fernando Valley.

391
Figure 2 Peak horizontal and vertical ground accelerations recorded at some of the

instruments in the affected area operated by the USGS's National Strong Motion

Program.

392
0 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15

Time, seconds

Figure 3 Accelerograph record from the grounds of the Sepulveda Veterans Affairs Hospital,

7 km from the epicenter.

393
Figure ty- Serious damage to buildings caused by the magnitude 6.8 Northridge earthquake

was widespread, but, outside of the epicentral area, seriously damaged buildings

tended to occur in pockets.

394
Figure 5 At the Northridge

Meadows apartment complex near the

epicenter the first story in several of the

buildings collapsed, killing sixteen

people. The second and third stories

remained essentially intact. Large

portions of the first story in these

buildings were open space for parking.

This type of collapse occurred in a

significant number of similar apartment

buildings throughout the affected area.

Figure G> Steel

pipe columns

between parking

bays supported the

upper stories of the

Northridge

Meadows apartment

buildings. Few of

the buildings in the

north half of the

complex collapsed,

but all were

damaged.

395
Figure 7 At

Northridge Meadows, the first story

steel pipe columns

generally hinged

near their bases but

retained their

length, controlling

the displacement of

the upper stories.

Figure & Champagne Towers

apartment building in Santa Monica

suffered a classic form of earthquake

damage: shear failure of the coupling

beams between linked shear walls at the

transverse end of the building.

396
Figure 9 The

back of the Champagne Towers

building was badly

damaged. The

railings here are

solid infill, which

effectively stiffen

the lower half of

the columns. The

shortened upper

half of the columns

were forced to accommodate all the displacement,

and they failed in

shear.

Figure / 0 The seven-story northwest wing of St. John's hospital in Santa Monica

experienced severe X-cracking in the piers between the second story windows.

397
Figure / / A portion of the concrete

waffle slab roof and the interior floor

slabs at the Bullock's department store

at the Northridge Fashion Center mall

fell to the basement, leaving many of

the columns standing. No reinforcing

steel can be seen in the remaining

columns that would have connected the

columns to the floor slabs.

Figure f 2 Damage to unreinforced masonry buildings, even those that had been

rehabilitated, was common. Upper stories and corners were particularly

vulnerable to damage. The rows of light-colored diamonds on the face of this

building are the end plates of through-bolts used to connect the wall to the floor

joists, a typical rehabilitation technique. The remains of parapet braces can be

seen on the roof.

398
Figure ' 3 The comer of this rehabilitated unreinforced masonry apartment building in East

Hollywood crumbled on all four stories.

Figure / H- The second story of the concrete frame Kaiser Permanente office building on

Balboa Avenue in Northridge completely collapsed. The first story was

relatively undamaged, suggesting it was far suffer than the story above.

399
Figure / 5 The exposed concrete

frame of the Kaiser Permanente

building, revealing deep beams,

suggests that the building had a strongbeam, weak-column configuration.

Except for the second story columns,

the columns and beams remained intact

while the joints disintegrated.

Figure / £ The stiff infill panels of

the Barrington Building effectively

shortened the concrete columns, forcing

all the displacement into the column

sections bounded by windows.

Diagonal X-cracking decreased in

severity towards the top of the building.
Figure 1 7 The

first-story columns

of the steel-framed

building housing

the American

Savings Bank on

Topanga Canyon

Road in Canoga

Park flexed during

the quake and

ended up slightly

displaced from

their original

positions.

Figure f£> The three-year old four-level parking garage at the Northridge campus of

California State University partially collapsed in the earthquake. The collapse

apparently started at the interior, causing the exterior walls to be pulled in to the

middle of the structure. Precast concrete "trees" were connected to form a

moment-resisting frame at the exterior of the structure.

401
Figure 1 9 The

cast-in-place concrete

parking garage at the

Kaiser Permanente

Hospital complex in

West Los Angeles

collapsed in on itself. The moment frame

that provided lateral

resistance on the south

face of the structure could not resist the

out-of-plane forces

apparently caused by

the collapse of interior

elements.

Figure 2-0 - At the east and west faces

of the Kaiser Permanente Hospital

parking garage, shear walls made up

the lateral force resisting system. Both

walls had horizontal cracks caused by

the out-of-plane bending when the walls were pulled inward. Neither wall had

major diagonal cracking, suggesting that

the lateral load was never fully

delivered to the walls. The stair towers

remained erect. The east wall and stair

tower are shown here.

402
Figure 2? The

interior columns of

the two-level

parking garage at the Trans World

Bank in Sherman

Oaks had been

seismically

strengthened after it suffered damage in

the 1971 San

Fernando quake.

The interior

columns fared well

in the 1994

Northridge quake.

Figure 22 The exterior columns of

the Trans World Bank parking garage,

which had not been strengthened, were

badly damaged.

403
Figure 23 Bridge damage sites studied by the reconnaissance team

404
Figure 2.^ Aerial view of the bridge at Gavin Canyon, looking to the east, showing

appearance of the four end spans shortly after demolition had begun

Figure 2^ Ground level view of the southbound bridge at Gavin Canyon, looking to the

northwest, showing the hinge bearing seat near Bent 3.

405
Figure 27 SRI 4/15 interchange after the 1971 San Fernando earthquake. View looking to the west, showing collapsed South Connector Overcrossing

106
Figure 26 Overview of the collapse of the SRI 4/15 Separation and Overhead, Ramp C,

looking to the south at Abutment 1. Pier 3 can be seen at the right of the

photograph (Courtesy of EERC)

f

Figure 29 Damage at Pier 3 of the SRI 4/15 Separation and Overhead, Ramp C, looking to the west (Courtesy of EERC)

407
Figure 30 Collapsed portion of the SRI 4/15 North Connector Overcrossing, looking east.. Pier 2 was at the left, near the broken hinge on the ground (Courtesy of EERC)

408
Figure ?>\ View of the flexural failure at Pier 3 of the SRI 4/15 North Connector

Overcrossing, looking to the south

409
Figure 33 Damage to piers of the SRI 18 Mission-Gothic Undercrossing, looking to the

northwest at the north face of the westbound bridge, Bents 3 (foreground) and 2.

Figure 3^h Damage to a pier of the SRI 18 Mission-Gothic Undercrossing, looking to the

northwest at the southernmost pier of Bent 3 of the westbound bridge.

410
Figure 36 Damage to piers of Bent 3S of State Route 1 18 at Bull Creek Canyon Channel,

looking to the southeast.

411
Figure 37 Collapsed westbound lane of 110 at Venice Blvd.; pier of Bent 7 in foreground

412
Figure 3 ^ Northernmost pier of Bent 6 supporting westbound lane of 110 (west of Venice

Blvd.)

Figure IJO Piers of Bent 3 supporting eastbound lane of 110 at La Cienega Blvd.

413
Figure V Hinge adjacent to Bent 4 of I 10 at Fairfax Ave.; cable restrainers prevented loss

of support (Courtesy of EERC)
Figure ^3 Failed column in Bent 3 of 110 at Fairfax Ave. (Courtesy of EERC)

415
Figure The steel pipeline at Terminal Hill separated from its supporting saddle at several

places and pipe sections bulged at other locations along the alignment.

8 pairs of restrainer rods 1 -3/8" (35 mm ) diameter v Coupler

1

Bracket 77" (1955 mm) diameter pipe

Gasket ring —

Section A-A

Figure <+S Schematic drawing of connection of the two steel pipe sections which were pulled

apart during the earthquake.

416
Figure Location of the 0.6- m gas main rupture on Balboa Boulevard.

417
JSP'.--'

Figure 4 7 Ground contraction along Balboa Boulevard caused the rupture of a gas

distribution line, the Rinaldi trunk water main, and a gas main.

418
Figure Q-S Ground extension along Balboa Boulevard resulted in the rupture of the same

gas main and water main shown in Figure **7. The rupture of the gas main

here caused a major fire that engulfed five houses.

419
Figure SO Some of the buried pipelines located near the intersection of Balboa Boulevard

and Bircher Street~the potential hazard of co-location of lifeline systems. At

this site are three water lines (two of them main lines); three gas lines; two

sewers; one crude oil line; overhead power, telephone, and cable TV lines; the

street lighting system; and the street itself.

420
Figure SI Damage

to DC equipment at several high voltage

substations, such as

this one at Sylmar,

led to widespread

power outages in the

Los Angeles area as well as isolated

outages throughout seven western states. (Photo courtesy of

Edward Matsuda,

Pacific Gas and

Electric Company.)

Figure 52 Porcelain is an integral part

of high voltage electrical equipment due to

j insulation requirements and is also the

most vulnerable to damage during strong

earthquake shaking. (Photo courtesy of

Edward Matsuda, Pacific Gas and Electric

Company.)

421
Figure 5*3 Most of the 230 kV circuit switchers (in the forefront) at the Sylmar substation

were damaged. Most of the capacitor banks (in the background) performed

well in this earthquake, whereas most collapsed during the 1971 quake.

(Photo courtesy to Edward Matsuda of Pacific Gas and Electric Company.)

Figure ^f" Derailment of a 64-car

freight train in Northridge. The

incident resulted in the spill of 30,000

liters of sulfuric acid and 7,500 liters of

diesel fuel.

422
Figure S5

Transverse cracks

in pavement along

Balboa Boulevard

in Northridge area.

Figure 56 Fire damage in a single story commercial building.

423
Figure 57 Fire damage in a two story commercial building.
Figure 63 Partially collapsed apartment building with fire damage in the rear.

Figure 60 Damaged gas service in a manufactured housing development.

425
Figure & I Multi-family residential building destroyed by fire.

Figure 6 2. University science building damaged by fire.

426
Summary Report on the Northridge, South California, U.S.A.,

Earthquake of January, 1994

by

Yukihiko Sumiyoshi 0

, Tomomitsu Fujii 2)

, Kazuhiko Kawashima3)

, Tatsuo Uwabe4>

, Kouichi Koshiumi 5)

, Eiichi Itoigawa6)

, Yoshiaki Nakamura7)

, Kimihiko Izumi8) and Hideki Sugita9)

ABSTRACT

The Northridge, California, U.S.A.,

Earthquake occurred at 4:31 on January 17,

1994. Destructive damage was developed in the

epicenter region. This paper describes a summary of the damage investigation by the

Japanese Government Study Team for the

Northridge Earthquake.

Key Words .'Northridge Earthquake,

Damage Investigation,

Urban Disasters,

Earthquake Disaster

Countermeasures

1. INTRODUCTION

Destructive damage was developed due to the Northridge earthquake which occurred at north/west of Los Angeles at 4:31 on January

17, 1994. Because the earthquake occurred

close to a major urban area, the urban-type

earthquake damage such as collapse of freeway

bridges and buildings, damage of lifeline

facilities were developed.

The Japanese government sent a study

team consisting of 21 experts of the ministries,

agencies and institutions concerned to contribute to the development of earthquake

disaster prevention measures not only for Japan

but for world-wide countries where the occurrence of earthquakes is a great threat. Two

experts from Railway Technical Research

Institute and Tokyo Gas Company and six experts for fire office of local governments

accompanied to the study team. Mr. Yukihiko

Sumiyoshi, former Director-General of the

Public Works Research Institute, Ministry of

Construction and Tomomitsu Fujii, Director of

Earthquake Disaster Countermeasures Division,

National Land Agency served as the leader and

sub-leader of the study team. The investigation in the epicentral region was made from March 1 to 12, 1994 with a

great assistance and support of many

organizations concerned including U.S. Federal

organizations, county and city governments,

private companies and individuals. The main

objectives of the study team were to investigate

1) damage and restoration of civil infrastructure

and buildings, 2) damage and restoration of

lifeline facilities, 3) fire, and 4) emergency

treatment and support for damaged peoples.

The investigation by the study team was

1) Former Director-General, Public Works Research

Institute, Ministry of Construction

2) Director, Earthquake Disaster Countermeasures

Division, Disaster Prevention Bureau, National Land

Agency

3) Research Coordinator for Underground Development,

Public Works research Institute, Ministry of Construction

4) Chief, Earthquake Disaster Prevention Laboratory,

Structures Division, Port and Harbor Research Institute, Ministry of Transport

5) Assistant Director, Building Guidance Division,

Housing Bureau, Ministry of Construction

6) Senior Researcher, Urban Disaster Prevention Division,

Urban Planning Department, Building Research

Institute, Ministry of Construction

7) Chief, Traffic Management Section, Traffic Operation

Department, Japan Highway Public Corporation

8) Chief, Design Division, 3rd Construction Department,

Metropolitan Expressway Public Corporation

9) Research Engineer, Earthquake Engineering Division,

Earthquake Disaster Prevention Department, Public Works Research Institute, Ministry of Construction

427
successfully completed with great supports by

Dr. R. N. Wright, U.S. Chairman of the Panel on Wind and Seismic Effects, UJNR, Mr. Noel

J. Raufaste, Secretary General of the Panel, Dr. H. S. Lew, Mr. James D. Cooper, Dr. S. C.

Liu, and Mr. James H. Gates (Panel members).

The original full report is being prepared

with the contents as

i)

Preface

1. General

2. Natural and Social Condition

3. Systems of Earthquake Prediction and

Precursors of the Northridge Earthquake

4. Outline of the Event, Crustal Movement

and Strong Motion Observation

5. Soil Liquefaction and Slope Failure

6. Damage and Restoration of Roads and

Bridges

7. Damage and Restoration of Buildings

8. Damage and Restoration of Lifeline

Facilities

9. Break out of Fires and Countermeasures

10. Emergency Response

11. Administrations for Disaster Prevention

12. Concluding Remarks

This paper presents a summary of the

report with an emphasis on the damage of

roads, bridges, buildings and lifeline facilities,

and fires.

2. OUTLINE OF DAMAGE

Based on the information from OEM of

City of Los Angeles, FEMA and OES of State of California, the major damage of the

earthquake was as:

(1) death:

(2) injured:

(3) families affected:

(4) source of fire:

57

9,348

about 18,000

about 100

(5) loss: about 15-30 billion dollars

In addition to the death of 57, there were

several persons who died due to indirect effect of the earthquake such as heart attack. The

families affected means those whose houses

were so badly damaged that they could not stay

there. The source of fire represents the fire which broke out in one hour after the

earthquake in Los Angeles City. Besides those

fires, there were several fires outside Los

Angeles City.

3. GROUND MOTION

Many strong motion records with high

accelerations were obtained by CDMG, USGS

and LADWP.

Fig. 3.1 shows the records at Tarzana

Cedar Nursery where a very high acceleration over 1.8 g was measured. It was measured on a

hill by a SMA-1 strong motion accelerograph

installed in a plastic case as shown in Photo

3.1. Because the SMA-1 is to measure an

acceleration up to 1 g, the acceleration over

scaled from the recording film space as shown

in Photo 3.2. The peak acceleration with 1.8 g was recorded beyond a driving hole at a side of

the film. It is expected that the recording

accuracy for over 1.8 g acceleration is verified

by a shake table test of the SMA-1

accelerograph.

It is interesting in the Tarzana records to note that the duration of high acceleration is about 8 seconds, and that this is quite different

with other near field records which have much

shorter duration. This would suggest the large

amplification by the surface soils. But the shear wave velocity measured at 250 m west of the

Tarzana recording site is 278 m/sec from the surface to 5 m below the surface, and 409

m/sec from 5 m to 30 m from the surface. It is noted that the ground impedance is not so high.

Fig. 3.2 shows the Arleta record. This was recorded at Arleta Fire Station at Nordhoff

Avenue. The vertical peak acceleration was

0.59 g and was lager than the horizontal

components of 0.35 g and 0.29g. It should be

noted that this is not the free field record but was recorded on a concrete floor of the fire

station as shown in Photo 3.3. Fig. 3.3 shows the response accelerations

of the records at Arleta, Newhall, Pacoima and

428
Sylmar. Although the soil boring data are not

available for those sites, it is interesting to know that the response acceleration at longer

natural period such as the value at 1 second is quite large at Newhall and Sylmar. The highest

response acceleration at 0.3 second is close to 0.3 g at Sylmar. It is a very important task to investigate why damage was not developed for most of structures although they were subjected

to such large ground motions. Few number of

inelastic excursion developed in structural members associated with short duration of

ground motion may be a possible reason for

this.

4. DAMAGE OF ROAD FACILITIES

4.1 Outline of Damage

Destructive damage was developed to road facilities. Fig. 4.1 shows the locations of

major road damage. Among various types of

road damage, such as cracks and buckling of

pavements, damage of bridge was significant.

Failure of gas and petroleum pipelines

embedded under ground suffered damage and

this suspended the roads. Overturning of

masonry and concrete block fence occurred.

Because there are enough space in roads and

because the earthquake occurred early in the

morning, this did not cause serious damage.

Seven bridges suffered significant

damage as:

1) I- 10 Fairfax/Washington Undercrossing

2) 1-10 La Cienega/Venice Blvd.

Undercrossing

3) S-118 Mission/Gothic Undercrossing

4) S-118 Bull Creek Channel Bridge

5) I-5/S-14 Interchange, South Connector

6) I-5/S-14 Interchange, North Connector

7) Gavin Canyon Undercrossing

The damage was developed due to insufficient lateral force and inadequate design

details considered in the pre-1971 seismic

design practice.

Among those bridges, two typical

damages are presented in the followings.

4.2 Damage of S-118 Mission/Gothic

Undercrossing

The bridge is of 3 span continuous

prestressed concrete box girder with 152 m

long (west bound) and 4 span continuous

prestressed concrete box girder with 170 m

long (east bound) as shown in Fig. 4.2. It was

designed in 1973 and the construction was

completed in 1976. Although the design was

completed after 1971, the original design

concept depends on the pre-1971 design

practice. Only some design detailing such as

spiral was updated according to the current design method.

The sectional area of concrete was small

as shown in Fig. 4.3. The columns had a hexagonal shape with diameter of 1.8 m. They

had flare at the top. The main reinforcements

and spiral were of 45 deformed bars with

diameter of 39 mm and round bars with

diameter of 15 mm at 8.75 cm interval. Two spans of the east bound bridge

collapsed as shown in Photo 4.1. Although the

west bound bridge suffered significant damage

to reinforced concrete columns, it did not

collapsed.

There were various columns which

suffered different level of damage. By

comparing the damage degree of those

columns, it can be seen that how damage

progressed. When the column was subjected to a seismic lateral force, flexural cracks were

firstly developed as shown in Photo 4.2. It is supposed that the main reinforcements and the

spirals already yielded at this stage.

As the lateral force increases, the shear

failure was initiated from the flexural cracking

as shown in Photo 4.3. Because the sectional area of concrete was small and because the

lateral confinement was not enough, the shear

strength was low. The core concrete was

initiated to be crushed and moved outward

direction. The main reinforcements bucked in outward direction. The spirals were initiated to be ruptured.

As the columns were subjected to the

429
alternative lateral force, the lateral resistance of

the columns was lost, and the columns were

dislodged as shown in Photo 4.4. Finally, the

columns lost the bearing capacity for supporting

the weight of superstructure, and failed as shown in Photos 4.5 and 4.6. The similar damage was developed at Bull Creek Channel Bridge on State Freeway

118 as shown in Photo 4.7, Fairfax/Washington

Undercrossing (I— 10) and La Cienega/Venice

Undercrossing (I- 10) as shown in Photo 4.8.

4.3 Damage of I 5/S-14 Interchange

Various destructive damage was

developed at this interchange. The S-14/I-5

Separation and Overhead which connects

southbound S-14 to southbound 1-5 collapsed

as shown in Photo 4.9. It was of five frames

with 475 m long as shown in Fig. 4.4. Three

spans from the abutment 1 (Al) collapsed. The

collapsed section was of a prestressed concrete box girder, and the both ends were supported

by Al and the gerver hinge near P4. The pier 2

collapsed and the pier 3 stranded as shown in Photo 4.10 with the deck being felt down.

The dimension of piers supporting the

collapsed section may be as shown in Fig. 4.5. They were reinforced by deformed bars with

diameter of 57 mm and the tie reinforcements

with diameter of 15 mm placed at 30 cm

interval.The failure may be initiated by the shear

failure of P2. Because of higher stiffness of P2

associated with shorter pier height, the seismic

lateral force induced in the deck concentrated at P2. Failure of the cable restrainers as shown in Photo 4.11 at the joint near P4 may be

developed as the a result of the collapse of

deck.

Although the failure is attributed to the

insufficient design lateral force and the detail considered in design, it shows the importance

of taking a total structural response into account

in design. Lateral force distribution of a bridge

supported by piers with different height

depends on the stiffness of piers as well as the

stiffness of soils.

The North Connector Overcrossing also

failed as shown in Photo 4.12. The deck

consisting of the second frame from Al

collapsed at P3. This may be caused by the

shear failure of P2. Similar mechanism with the S-14/1-5 Separation and Overhead contributed

to the failure of P2.

Separation of the two adjacent decks

occurred at many hinges. Photos 4.13 and 4.14

show two examples of such large separations.

Cable restrainers ruptured at some hinges as shown in Photo 4.15. Nevertheless the

effectiveness of the cable restrainers was

apparent. More collapse might be developed

unless the cable restrainers.

4.4 Effectiveness of Seismic Strengthening

There are many bridges which were

strengthened against an earthquake. As shown

above, the effectiveness of cable restrainers was

apparent for reducing the damage. No seismic

damage was observed at columns which were

strengthened by steel jacketing. Photo 4.16

shows an example of such columns.

4.5 Prompt Repair and Restoration

It was remarkable to know that every

efforts were made to prompt recovery of the

damage. Demolish of the major bridge damage was initiated from the day following the

earthquake. Explosive material was used at some bridges as shown in Photo 4.17 to reduce

the time required to demolish the damaged

bridges. A special contract among about 5

construction companies was made for reducing

the time required for usual open bids. A

penalty/bonus system was used for reconstructions. Seven days per week and 24

hours per day reconstruction is being made.

Well considered countermeasures

were/are being adopted for mitigating the traffic congestion due to close and damage of roads.

Detours were designated promptly after the

earthquake. High priority was given to high

occupancy vehicles. Time interval of traffic signals was adjusted so that the smooth traffic flow could be made. Many well prepared traffic signs were provided at various locations. Thus

430
serious traffic congestion was avoided.

It should be noted that this could be

made because the U.S. highway system is much

better than that of Japanese system from

various points such as traffic volume, number

of lanes and network systems. For example,

I-5/S-14 Interchange was one of the bottle necks in Los Angeles Area, because they cross

the narrow canyon. But number of lanes at this section including interstate freeway, freeway

and other roads is 22. Although this reduced to 6 lanes soon after the earthquake, it was soon

increased to 13 by detours. Even normal streets and boulevards in residential zone have 4 lanes.

If the similar type of bridge damage occurs in urban areas in Japan, unexpected damage and

effect might be developed. It is therefore

obvious from such evident that further earthquake disaster prevention measures are

inevitably required in Japan for reducing the

critical effects of earthquakes.

5. DAMAGE OF BUILDINGS

5.1 Locations of Building Damage

According to the map showing unsafe

buildings (Red tagged) and limited entry

buildings (Yellow tagged) in Los Angeles city, there are several areas where damage

concentrated because of difference of ground

shaking intensity. Within about 7 to 8 km from

the epicenter, damage of buildings distributed

almost uniformly, and in the outside area from

the epicenter beyond the Santa Monica

Mountain there were some areas where damage

concentrated such as Sherman Oaks, Hollywood

and Adams. It seems to be related with the soil condition. This requires more study.

5.2 Building Damage and Social System

At Sherman Oaks area, many multi-story

apartment and condominium buildings were

significantly affected by the earthquake. Most

of these building were constructed before 1974

at that time the structure code was revised.

According to the zoning code for city planning,

the area is designated as R3 (Multiple Dwelling

Zone). The code requests 1 to 2 parking spaces

for each dwelling unit and 90 to 130 dwelling

units per one hectare. Therefore parking space

is provided at first floor or basement floor as

"pilotis type". This structure type seems to have

increased the damage level of building during

the earthquake as shown in Photos 5.1, 5.2 and

5.3. Because it is required to provide parking

space in dwelling buildings in urban areas, a

city planning control system which matches

with the building planning and structure design

may be required.

In case of complete collapse of precast concrete parking garage at California State University at Northridge as shown in Photo 5.4,

the damage seems to be developed due to an

insufficient connection between elements of the

structure and lack of ability to support the

vertical loads while undergoing to large lateral deformation. It is interesting to note that compared with Japanese buildings, more

freedom seems to be accepted in U.S. in designing buildings. For example, structural engineers can adopt more free use of frame

under the condition that they can verify the

safety of buildings against the seismic loading

by dynamic response analyses. In such case, it

is required for structural engineers to conduct

overall and comprehensive engineering

judgment besides the requirements by the

building code.

5.3 Quick Administrative Response

It can be highly evaluated that California

State and Los Angeles city made a prompt

administrative response for the damage. For

example State Governor of California directed

the SSC to coordinate a study of the specific

policy implications arising from the earthquake,

with particular attention to implications for seismic structural safety and land-use planning.

And the structural codes of chimney and block

fence were revised. It is important to take

necessary measures to safety taking account of

seriousness of social influence based on

building damage in case of large scale earthquake.

431
5.4 Retrofit Program

Most of the retrofitted unreinforced

masonry building (URM) clearly performed

better than the unretrofitted ones as shown in Photo 5.5. Only some retrofitted URM suffered

minor damage as shown in Photo 5.6. On the

other hand, retrofitted program has not yet been

initiated for RC buildings, tilt-up buildings,

steel constructions, and wooden buildings. They

suffered damage as shown in Photos 5.7 and

5.8.

5.5 Post Earthquake Safety Evaluation

of Buildings

The procedures taken for post earthquake

safety evaluation of buildings seem

considerably effective to prevent secondary

disaster by the aftershocks. This activity

exhibited a very good style of social

contribution of the engineers including the

volunteers. Also these results were effectively

provided to other organizations such as FEMA

and OES, and used to support the application

and repair/ reinforcement consultation of

damaged buildings at DAC and ESC. The most

important point evaluated was to collect information for safety evaluation of buildings

and to release the data to appropriate

organizations.

6. DAMAGE OF RAILWAY AND PORT

6.1 Railway

Twenty nine trains of a 64 car freight

train traveling through Northridge district derailed by the earthquake. Twenty tank cars carrying sulfuric acid derailed, and 2,000

gallons of acid spilled from one of these cars. There was no other damage caused by this spilled acid. As there was no damage of derailing of

traveling train due to an earthquake in Japan,

this damage is very impressive and interesting

for Japanese railway engineers.

6.2 Port and Harbor

Some damage occurred at the Port of Los

Angeles. A piled pier as shown in Fig. 7.1 at the American President Lines container terminal was damaged. Several cracks formed in asphalt

pavement in the backland area which was

placed hydraulically from dredged materials.

Sand boils were present in this backland area. The concrete deck of piled pier moved

approximately 4 inches away from the

backland. Photo 7.1 shows crack of the rock

dike. Cracks also formed on the top of the

several prestressed concrete piles.

There was no damage at the Port of Long

Beach adjacent to the Port of Los Angeles.

Over a distance of about 300 feet of the

north pier at the King Harbor, substantial

subsidence occurred to a maximum of about 0.6

m to 0.9 mt, and the concrete bulkhead located on the right side at the north pier moved away

from the backland shown in Photo 7.2. The

maximum lateral offset in the bulkhead is estimated to be about 3 m. There was no damage at the concrete bulkhead on the left side of the north pier. This damage of large

displacement was due to the liquefaction of the

reclaimed land.

7. DAMAGE OF LIFELINE FACILITIES

At the Aqueducts Water Plant of the

Department of Water and Power, City of Los

Angeles, six water pipes suffered damage as shown in Photos 7.1, 7.2 and 7.3. This resulted

the suspension of operation for 24 hours.

However, because there is enough storage of 10

times daily consumption, serious problem for

distribution was not developed. Pipes broke at about 1,200 locations, and it required about 7

days for repair.

Several cracks with 5 cm wide were

developed at the crest of Lower San Fernando

Dam as shown in Photo 7.4. Because many

sand boils as large as 10 m were observed at the floor of the reservoir at upstream side. It seems that the damage was developed due to

soil liquefaction. The dam is empty and is served as a flood control basin.

At Joseph Jensen Filtration Plant,

432
Southern California Water District, one of the

two water intake pipes ruptured as shown in Photo 7.5. It took about 60 hours to repair the

pipe, and the operation was suspended during

the repair.

Electricity in Los Angeles is supplied by

the Los Angeles Water and Power and Southern

California Edison. Electric facilities suffered

damage at conversion stations, power stations,

and towers supporting high-voltage transmitting

lines. Electricity was suspended to all customers

supplied by LADWP (1.4 million homes) and

0.63 million customers supplied by Edison by

the seismic damage. Many equipments have

been replaced with the the ones with high

seismic performance after the San Fernando

Earthquake. They performed quite well. The gas pipes of Southern California Gas

Company suffered damage at 37 main

transmitting lines, 118 main lines and 79

service lines. Among damage of 37 main

transmitting lines, 27 were damaged ai oxy-acetylene girth weld. They were

constructed before 1932.

At Balboa Street of Granada Hill, a

service pipe with 6 inch diameter and a main

pipe with 22 inch diameter were ruptured as shown in Photos 7.6 and 7.7, and was ignited

the fire. Five single family houses along the

street were burned down. A main transmitting

line with 30 inch diameter embedded at the same location did not suffer damage. Large

lateral movement of pavement and surface soils as large as 30 cm was developed in road axis

as shown in Photo 7.8. A rupture of petroleum pipe was

developed at Wolfskill Street, San Fernando.

The oil was ignited, and caused a fire at houses

along the street. Further fires due to breaks of gas pipes were developed at several mobile homes (refer

to Chapter 8). The telecommunication systems suffered

damage due to structural and nonstructural

damage of buildings, rupture of optical fiber

associated with a fire and damage of telephone

exchanger. Emergency electricity back-up

system did not work at some stations. Concentration of telephone traffic caused

suspension of traffic. The traffic increased 60 %

at AT&T on January 17. Wasted water treatment facilities suffered

damage at pipes. Small TV cameras were

effectively used to identify the damage. The

damage information of buildings was

effectively used to identify the areas where

possible damage was developed.

8. STRUCTURAL DAMAGE BY FIRES

8.1 Introduction

Major cities in Japan are quite vulnerable

to urban fires because of a great number of

combustible buildings. It is anticipated that numerous fires would break out in such big

cities, if a large earthquake with magnitude 8

such as the anticipated Great Kanto Earthquake

occurs. Fires would spread uncontrollably over

a wide area. They would grow into large scale

fires called as "Post Earthquake City Fire" in the areas where wooden houses located densely.

Over 140,000 people were killed in the 1923

Great Kanto Earthquake because they lost their way to evacuate by fires. Therefore the investigation was conducted with a great

interest for the structure fires by the Northridge

earthquake.

8.2 Structure Fire Incidents

Many fires broke out after the earthquake

and some of them grew into large scale fire in mobile home parks.

The Los Angeles City Fire Department

(LAFD) was alerted many times including

requests for rescue because of the telephone

congestion. The LAFD made quick response

according to "Earthquake Emergency

Operational Plan". In mobile home parks, there were many

fire breakouts. At Sylmar, LA City, 20 to 60

mobile homes were burnt down in the parks.

It was interesting that except such mobile

home parks fires which broke out at built-up

areas did not spread widely across streets and

433
that most of these fires were isolated. By 9:45 of the day, all fires in LA City were controlled and there were no active major

structure fires in progress.

8.3 Number of Breakouts, Scale of

Fires and Cause

According to LAFD, there were about

100 structure fire incidents within 1 hour after the earthquake in LA City. Although number of

breakouts of each incident has not yet identified

at the present time, the fires during the

earthquake may be recognized as "simultaneous

fire breakouts after an earthquake." About 80%

of these fires were in Division 3 which includes

San Fernando Valley.

One of the significant incidents was the

rupture of a large natural gas main and a water

main at the intersection of Balboa and Rinaldi

as shown in Photo 8.1. Subsequent to the

rupture, the natural gas main ignited, and five

single family homes were burnt down.

Also significant loss of property resulted

from structure fires at three mobile home parks.

According to LAFD, there were 9 fire incidents even in a mobile home park. The structural building codes are not applied to the mobile

homes and mobile home parks. Therefore, the

supports of mobile homes were weak, and

many mobile homes felt down from their supports. Because the natural gas valves were plumped to mobile homes, the natural gas

valves and pipes were ruptured as shown in Photos 8.2 and 8.3. Mobile homes have also no restriction of

fire regulation. Therefore the performance of

fire prevention of mobile home are generally

poor and they are vulnerable to fires as shown

in Photos 8.4, 8.5 and 8.6. Distance between

adjacent mobile homes are also relatively small

as show in Photo 8.7. These made fire spread

inside mobile home parks.

Most of fire breakouts following the

earthquake were developed due to leak of

natural gas. Pilot flame of water heaters ignited

the natural gas. Another cause of the fire breakouts in the evening of the day and the

next day was ignited by electric sparks after electric power supply was restored. Fires in wooden buildings in built-up

area didn't spread as is afraid in Japan because

the distance between buildings is large. The

performance of fire prevention of the wooden

buildings is not the only one reason for having

prevented the fire spread.

9. CONCLUDING REMARKS

The followings are the major findings of

the Japanese Governments Study Team for the Northridge Earthquake.

(1) Many strong motion records with high peak

accelerations were obtained. It is required to investigate how structures behaved during the

earthquake based on the measured accelerations.

It is particularly interesting to study why

damage was not developed in many structures even though they were subjected to such strong

ground shaking.

(2) It was fortunate that the earthquake

occurred very early in the morning. If the

earthquake had occurred in the daytime, much

larger loss of human beings might be

developed. This is particular true for the

damage of transportation and building facilities and for the effects of structure fires.

(3) The bridges which were designed in accordance with the pre-1971 seismic design

practice suffered major damage including

collapse. Small lateral force considered and

small concrete section with insufficient lateral bars caused extensive shear failure. Small seat length at hinge joints also contributed to the

collapse. The bridges which were strengthened

by cable restrainers and steel jacketing

performed quite well.

(4) The pilotis type wooden houses suffered

extensive damage. It should be noted that this structural type was adopted to provide a parking space at the first floor. A precast concrete parking garage collapsed due to

insufficient connection of beams and columns.

434
Most of retrofitted unreinforced masonry

buildings performed well. Reinforced concrete

buildings, tilt-up buildings and steel constructions which have not yet retrofitted suffered damage.

(5) Many pipes suffered damage, and caused

extensive leakage of gas, water and oils. Gas

and petroleum were ignited fires and burned

down houses along roads.

(6) A railway which was travelling at the

earthquake overturned. Because there was no

such damage in Japan, it is impressive and

important to learn why the overturning was

developed.

(7) Many fires simultaneously broke out immediately after the earthquake. But they were

controlled in a few hours. No extensive

spreading of fires extending streets occurred.

The long distance between houses contributed

to avoid the spreading. There was no

automatic/manual system to cut off the gas flow

when the gas leaked immediately after the

earthquake.

(8) Response and countermeasures after the

earthquake were prompt and appropriate. Every

effort was made to minimize and mitigate the

loss and effect of damage.

ACKNOWLEDGEMENTS

Valuable supports were provided from

both Japan and U.S. governments to the

Japanese Government Study Team for the

Northridge Earthquake. Special thanks are

directed to the Ministry of Foreign Affairs,

Consulate-General of Japan in Los Angeles and

San Francisco, National Land Agency and

Ministry of Construction.

Sincere appreciation is also directed to U.S. Federal government organizations, county

and city governments, private companies and

individuals concerned. The valuable support

through the Panel on Wind and Seismic Effects,

UJNR for arranging the visits to various

organizations and sites is highly acknowledged

and appreciated. Sincere appreciation is directed

to Dr. R. N. Wright (U.S. Chairman of the

Panel, and Director, Building and Fire Research

Laboratory, National Institute of Standards and

Technology), Mr. Noel J. Raufaste (Secretary

General of the Panel, NIST), Dr. H. S. Lew

(member of UJNR Panel, NIST), Mr. James D.

Cooper (member of UJNR Panel, Federal

Highway Administration), Dr. S. C. Liu

(member of UJNR Panel, National Science

Foundation), Mr. James H. Gates (member of

UJNR Panel, California Department of

Transportation), and Mr. LeVal Lund.

REFERENCES

1) Japanese Government Study Team for the

Northridge Earthquake : Report on the

Damage by the Northridge Earthquake,

National Land Agency, in preparation (in

Japanese)

2) Kawashima, K., Unjoh, S., Sugita, H.,

Hoshikuma, J., Iwasaki, T., and Yoshida,

Y. : Reconnaissance Report on the

Damage by Northridge, California, U.S.A.,

Earthquake of January 1994, PWRI

Technical Note, No. 3272, Public Works

Research Institute, February 1994 (in

Japanese)

3) CSMIP Strong Motion Records from the

Northridge, California Earthquake of

January 17, 1994, California Earthquake,

Department of Conservation, Division of

Mines and Geology, Office of Strong

Motion Studies, Report OSMS 94-07,

February 1994

4) Processed Strong Motion Records from the Northridge, California Earthquake of

January 17, 1994, California Earthquake;

Release No. 1 and No. 2, Department of

Conservation, Division of Mines and

Geology, Office of Strong Motion Studies,

Report OSMS 94-06B and 94-08,

February 1994

5) Cooper, J. D. and Friedland, I. : 1994

Northridge California Earthquake of

January 17, 1994, FHWA News, 1994

435
6) California Department of Transportation: Post Earthquake Investigation Team

Report on the Northridge Earthquake,

1994

7) Preliminary Report on the Seismological

and Engineering Aspects of the Northridge

Earthquake, EERC, University of

California, Berkeley, Report No.

UCB/EERC-94/01, January 1994

8) Seismological and Engineering Aspects of

the January 17, 1994 Northridge

Earthquake, Earthquake Engineering

Research Center, University of California,

Berkeley, 1994

9) Priestley, M. J., Seible, F. and Uang, C. M. : The Northridge Earthquake of

January 17, 1994, Damaged Analysis of

Selected Freeway Bridges, Report N.

SSRP-94/06, University of California, San

Diego, February 1994

10) Kawashima, K. : Seismic Retrofit of

Existing Road Bridges in the State of

California - Two Years Later after the

Loma Prieta Earthquake, Bridges and

Foundations, 92-5, May 1992 (in

Japanese)

11) EQE: The Northridge, California,

Earthquake of January 17, 1994

12) Fumai, T. E., Gibbs, J. F. and Roth, E. F.: In-situ Measurements of Seismic Velocity

at 19 Locations in the Los Angeles,

California Region, USGS, Open-File

Report 81-399, 1981

13) Earthquake Engineering Research Institute: Northridge Earthquake of January 17,

1994, March 1994

14) Persson, V. H. : Assessment of Damage to State Jurisdictional Dams Caused by the

Northridge Earthquake on January 17,

1994, Seismic Safety Commission

Meeting, Burbank Airport Hilton, March

1994

15) Status Report on Collection System

Damage Assessment to Northridge

Earthquake, Collection System Engineering

Division, Bureau of Engineering,

Department of Public Works, City of Los

Angeles, March 1994

436
(GHU

-500.0

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I SEC

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Fig. 3.2 Acceleration Record at Arleta - Nordhoff Avenue Fire Station (from CDMG)

437
3000

101:

0.05 0.1 0.2 0.5 1 NATURAL PERIOD(sec)

ARLETA

<T3

St 300 - 2>.jj _j

CJ

LUz

0. 05 0. 1 0. 2 0. 5 1 2 NATURAL PERIOD(sec)

NEW HALL

Fig. 3.3 Response Acceleration Spectra with 5% Damping Ratio

438
Joseph Jensen Filtration

Station—Pipe Break

S—126

1-5, S-14 Interchanged Connectors)

>— Pacoima Dame

Uppar San Fernando Reservoir

Los Angeles Reservoir

-Pacoima/Kagel (0.43g, 0.19g)

Lower San Fernando Reservoir

Los Angeles Aqueduct Filtration

Plant — Water Pipe Break

I— 10 Fair Fax/Washington Viaduct

\\ \

I— 10 La Cienaga/Venice Viaduct \

N

Fig. 4.1 Locations of Major Road Damage

439
1 68 '6"

506'

159' 6' 566 ' V4

" 178'

# 6 @12"

(c) (1MB) M (e) SfrS

Photo 4J Initiation of Shear Failure (S-118 Mission/Gothic Undercrossing)

440
RCStS P CMj RC+Hii P CTsfii

Fig. 4.4 Collapse of S-14/I-5 Separation and Overhead

53

26' 6" 26' 6"

li

I—n

n —n

M jl^

f bi jl it

! L, —1_

jj ~ir il

(a) ++-y7't"-A

1480' (P 2»)

(b) W,m (-«05)

12'

6' 6'

|-

Q * 18 #5 ©12" #4 ©12

(c) EM (-&ss;

Fig. 4.5 Typical Section of Piers Supporting the Collapsed Part of S-14/I-5 Separation and Overhead

441
442
Photo 3.1 Recording Site at Tarzana sery
Photo 4.1 Collapse of S-118 Mission/Gothic Undercrossing

444
Photo 4.4 Shear Failure and Extensive Outward

Deformation of Main Reinforcements

(S-118 Mission/Gothic Undercrossing)

Photo 4.5 Dislodgement of Column Axis due

to Shear Failure

(S-118 Mission/Gothic Undercrossing)

Photo 4.6 Complete Failure (S-118 Mission/Gothic Undercrossing)

445
7 Damage of Columns of Bull Creek

Channel Bridge

Photo 4.8 Damage of Column of La Cienega/

Venice Undercrossing

Photo 4.9 Collapse of I5/S14 South Connector

446
Photo 4.10 Damage of P3 (Courtesy of EERC)

Photo 4.11 Failure of Cable restrainers at a Joint near P4

Photo 4.12 Collapse of North Connector Overcrossing

(Courtesy of Patrick K. Hipley, Caltrans)

447
Photo 4.13 Large Separation Developed at Hinge

Photo 4.14 Large Separation and Damage

of Hinge Seat (Courtesy of Patrick

K. Hipley, Caltrans)

448
Photo 4.16 Effectiveness of Strengthening of Columns with

Steel Jacketing

Photo 4.17 Demolish of Damaged I-5/S-14 INterchange by

Explosion

449
Photo 5.1 Damage of Northridge Meadows Apartment

(First floor of the left side building collapsed completely)

Photo 5.2 Damage of Northridge Meadows Apartment (First floor of the right side building was reinforced using wooden columns. Original posts were slender steel pipes)

Photo 5.3 Basement Floor of an Apartment in Hollywood (To provide require parking

spaces required, number of columns and walls at the basement floor is small)

450
451
JL i Photo 5.7 Damage of A Tilt-up Wearhouse in Northridge

452
Photo 6.1 Cracks of Rock Dike

453
Photo 7.1 Rupture of a Main Water Pipe (Aqueducts Plant, DWP,

City of Los Angeles)

Photo 7.2 Damage of a Expansion Joint Water of a Pipe near a Tank on a Hill (Aqueducts Plant, DWP, City of Los Angeles)

454
Photo 7.3 Local Buckling of a Water Pipe Photo 7.4 Cracks on Crest of Lower San

on Downslope of a Hill (Aqueducts Fernando Dam (Aqueducts Plant,

Plant, DWP, City of Los Angeles) DWP, City of Los Angeles)

Photo 7.5 Rupture of a Water Pipes (Joseph Jensen Filtration

Plant, Southern California Water District)

455
Photo 7.6 Rupture of Gas and Water Pipes due to Tension

(Balboa Street, Granada Hill)

Photo 7.7 Rupture of Gas and Water Pipes due to Compression

(Balboa Street, Granada Hill)

456
Photo 7.8 Lateral Movement of Pavement

and Surface Soils as Large as 30

cm (Balboa Street, Granada Hill)

Photo 8.1 Damage of Five Single Family Homes due to Fire at Intersection of Balboa and Rinaldi

457
Photo 8.2 Damage of Mobile Homes due to Weak Supports

Photo 8.3 Rupture of Natural Gas Valves Plumbed to Mobile

Homes Which Fell Down their Supports)

Photo 8.4 Vulnerable Mobile Homes to Fire

(Mobile homes have no restriction of

fire regulation based on fire code)

458
Photo 8.6 A Mobile Home with Plywood Ceiling

Photo 8.5 Structure of External Wall of a Mobile Home (Frame is of wood

and siding is of hard board or aluminum panel)

Photo 8.7 Road spacing Mobile Homes (Though roads in mobile home parks are

relatively wide, distance between mobile homes is small because of terrace)

459

The M8.1 Guam Earthquake of August 8, 1993

by

A. Gerald Brady 1

ABSTRACT

Guam's seismicity is controlled by

its position in the Mariana Islands

at the uplifted eastern edge of the

Philippine plate, under which the

Pacific plate is subducting. This

deep thrust magnitude 8.1 event oc- curred on the junction between the

plates. The source mechanism, geometry, and attenuation factors con- tributed to significant structural

damage to hotels in Tumon Bay, but

not to stiff, one- and two-story

residential and commercial buildings. Future earthquakes on closer,

smaller faults threaten these highfrequency structures. No serious

injury or deaths were attributed to

the earthquake, and therefor media

attention was minimal. The maintenance of three strong-motion instruments had been neglected due to

fiscal constraints, and electronic

and power failures occurred at the

USGS observatory. No records were recovered

.

KEYWORDS: earthquake; ground motion;

Guam; seismicity; structures.

1. INTRODUCTION

Mariana Trench where the two plates

converge. The Mariana Trench con- tains the greatest measured depth of

the world's oceans, approximately

11.3 km (37,000 feet) located to the

southwest of Guam. The trench is

arc-shaped like the island chain,

and its minimum distance from Guam

is 112 km (70 miles) to the southeast, where it is aligned predominantly in the northeast-southwest

direction, at right angles to Guam's

azimuth (Figure 2). Given this geometry, if a deep thrust earthquake

on this fault system were to be

close enough to Guam to cause significant structural damage, it would

seem that the earthquake would have

to be located southeast of Guam, or

at least in the southeast quadrant.

2. SEISMOLOGY

The seismicity of Guam can be briefly outlined by the listing of earthquakes from 1825 to 1936 in Table 1 (from Tracey and others, 1964) and

by the map showing damaging Guam

earthquakes from 1975 to the present

(Figure 2, from the USGS/National

Earthquake Information Center (NEIC)

in Golden, Colorado)

.

Guam is the southernmost island of

the Mariana Islands in the western

Pacific Ocean (Figure 1, from Tracey

and others, 1964). The islands form

a slight arc, convex to the east,

and span from 13° 30' to 20° 30' N,

and 145° to 146° E. The island

chain is 770 km (480 miles) long.

Guam is 46 km (30 miles) long, averages 11 km (7 miles) wide, and is

aligned approximately NE-SW. The

island arc outlines the Mariana

Ridge at the eastern edge of the

Philippine Plate, which is forced to

rise in elevation as the Pacific

Plate dives below it in a northwesterly direction. At the same time,

the descending Pacific Plate, while

dragging the eastern edge of the

Philippine Plate with it, forms the

2.1 The 1993 magnitude 8.1 Guam

earthquake

Updated information on the 1993 epicenter location was available from

NEIC over the first several weeks

following the earthquake. At various times, the epicenter was placed

at all points of the compass around

the island, as different or more re- cent subsets of data and interpretation were considered. As of

8/26/93, 18 days after the earthquake, the magnitude 8.1 earthquake

was located at 12.97° N, 144.74° E

at a depth of 61 km (38 miles).

'USGS, Menlo Park, CA94025

461
This places the epicenter 60 km

(37.5 miles) due south of Agana, the

commercial center of the island. It

is a few kilometers further to Tumon, where most of the tourist hotels are located. The origin time

was 08:34:25.0 on Sunday, August 8,

1993 (GMT), 6:34 pm local time. The

depth estimate of 61 km (38 miles)

is not dependent on the USGS' World

Wide Seismic Station on Guam, so the

failure of the instrumentation

there, both in the electronics and

the power supply, does not affect

the accuracy or precision of the

estimate.

Guam is a remote island in the western Pacific Ocean, and epicenter and

depth calculations cannot be made as confidently as in California. For a 90% confidence level in the epicentral location the error circle has a radius of 660 km (412 miles).

The moment tensor solution for the

source mechanism provides two planes

on which the first motion might have

occurred. With the restrictions

placed on the correct mechanism by

geologic considerations outlined a- bove, the preferred fault plane

strikes at N70E and dips steeply

(77°) to the NNW. The first motion

has the Pacific Plate moving down

dip (to the NNW) and along the

strike (to the WSW) in equal

amounts

.

The aftershocks listed in NEIC's

8/26/93 summary are constrained to

have depths of either 60 or 33 km.

They are plotted in plan in Figure

2. With this constraint, only the

plan view of the portion of the

fault ruptured during the main shock

can be identified using aftershock

epicenters

.

A seismicity map received from NEIC

with plotted epicenters, (10° to 15°

N, 143° to 148° E, M=5.5 and above)

shows that there are many earthquakes on faults near Guam. It is

not likely that magnitude 8 events

will ever be closer than this one. But lesser magnitude events can occur on closer faults, whether such

faults are already known, inferable

from this plot, or not yet found.

These have the potential for much

damage, affecting particularly stiff

structures such as one- or twostoried houses and buildings.

3. STRONG GROUND MOTION

An estimate can be made of the character of the ground motion from observations on the island. Many people interviewed agreed with the

newspaper accounts of the shaking

lasting 60 seconds. The aftershock

distribution indicates that the

mainshock rupture must have been for

the most part in one direction,

namely to the northeast, and could

well have lasted 40 seconds or more. Wave motion arriving from the rupture's start, at the focus, will

have been only slightly stronger

than the motion arriving from the

rupture's end, which is not much

further from Guam. If we maintain

the assumption that the rupture was travelling to the northeast it will

have finished up receding from Guam,

and therefore the later arrivals

will have been further attenuated in

amplitude. We conclude that the

strong shaking lasted 30 seconds,

and then slowly lessened.

A 61-km depth and 60-km epicentral

distance permits geometric threedimensional spreading to occur for a greater distance and longer duration

than would be expected for a shallow

high-magnitude California event. We

conclude that peak acceleration amplitudes will be low (say 0.25 g)

,

and that high frequencies, greater

than 5 Hz, will not be present.

Slight damage to small, solidlybuilt houses and buildings of one

and two stories, and flimsy buildings, too, in some cases, also

points to a lack of high frequencies. On the other hand, the damage

to the tourist hotels of four to

twenty stories indicates prominent

ground motion components having periods between 0.5 and 2 seconds.

The fault plane is approximately

perpendicular to a line to Guam, so few high-amplitude P-waves arrived

there. Furthermore, most shear waves

generated on the fault arrived at

Guam refracted sufficiently to be

travelling vertically, thereby supplying minimal vertical accelerations. There were no indications of

disturbed soil at hill crests, ex462
cept landslides, and no reports of

furniture or equipment moving sideways while airborne. We conclude

that vertical accelerations certainly did not exceed 1 g, and probably

offered a negligible contribution to

damage, considerably less than was provided by the horizontal accelerations . 3.1 Ground motion at damage sites.

The 61-km focal depth ensured that

the attenuation of ground motion

amplitudes was not rapid from the

southern to northern extremities of

the island. All firm foundation

soils or rock in the Agana business

district and the Tumon hotel district probably received shaking of

the same character. Accelerations

are unknown at potentially liquefiable sand layers overlain by fill,

that failed and spread during the

earthquake

.

3.2 Tsunami

A tsunami was generated, but passed

by Guam without fanfare. A parked

pick-up truck on a beach on the east

coast was partially swamped. Much

more common typhoons provide ample

experiences of high ocean waves.

Tide gauges recorded maximum wave

heights of 98 cm at Japan and 19 cm

at Hawaii.

4. US NAVY INSTRUMENTATION

The US Navy over a period of years

ending in 1988 maintained three instruments in Guam as part of their

network of strong-motion accelerographs. No records were recovered

from the instruments, which had re- mained in place.

A network of strong motion accelerograph instrumentation does not generate support, either personnel or financial, by its very nature of patiently waiting for an earthquake

event. A network needs an associated research team of earthquake

scientists or engineers to provide

incentives for good recordings, and

to provide successful appeals for

long-term maintenance funding.

Some preliminary conversations have

been held between representatives of

the Department of Public Works and

the US Geological Survey's Guam Observatory, with respect to the installation and maintenance of a small network of strong-motion instruments. Cooperation with the

USGS in Menlo Park for processing

records and analyzing ground motion

is a possible future goal.

4.1 The 1983 magnitude 6.3

earthquake record.

The U.S. Navy strong-motion earthquake instrumentation on Guam included one Kinemetrics SMA-1 accelerograph installed in 1979 in Building 372, the Transportation Equipment Maintenance Shop. This instrument produced a record from the magnitude 6.3 1983 earthquake (Brady

and Mork, 1993) with an epicenter 30

miles (48 km) northeast of Agana,

the business center of the island,

and 31 miles (50 km) from the instrumented site. It was the first

such record from Guam, and the ar- chitect/engineer contractors working

on building analysis on Guam re- quested response spectra. Through

an arrangement between Stan Takahashi (retired, Navy Civil Engineering Laboratory (NCEL) , Port Hueneme,

California), IOM Towill (at the

time, a digitizing company in Santa

Clara, California), and the USGS,

Menlo Park, California, the record

was digitized and processed, and the

data was sent to NCEL.

5. NOTES ON THE STRUCTURAL DAMAGE

The Royal Palm Resort Hotel consisted of two closely-spaced 12-story,

reinforced concrete, matching towers, with a structural connection.

Several second-story columns near

the southwest corner of one of the

towers failed, shortening by the

height of the window opening. The

tower came to rest leaning against

its twin. The connecting arm suffered major structural damage as it

attempted to reconcile the displacements of the two towers. Both towers have been demolished.

The Grand Hotel was a 4-story reinforced concrete building, L-shaped

in plan. The non-symmetrical plan

and the glass-fronted commercial

first story contributed to the fail463
ure of the first story columns and

resulting horizontal offset of 8

inches (0.2 m) in the building's

final position. The top three stories remained in a relatively undamaged state. The hotel has been

demolished.

The Guam Reef Hotel had a very

heavily constructed canopy over the

entrance to its twin structures.

The canopy was supported on four

circular columns which failed ca- tastrophically , allowing the canopy

to fall to the ground, with almost

no lateral offset. Two taxis were

caught

.

Gas station shelters, built to

shield the pumps and customers from

vertical sun and rain, suffered se- vere structural damage at the top

and bottom of the supporting co- lumns .

6. WIND AND SEISMIC EFFECTS

Damage to one class of structure,

characterized by the canopy and the

gas station shelters in the section

above, and by the umbrella-shaped

bus-stop shelters, is an interesting

consequence of a design philosophy

that is dominated by the typhoons

that threaten Guam every year. The

design wind speed is 155 mph (240

km/hr) which produces a horizontal

force that is not excessively high

on roof cross-sections, and a vertical uplift force that is relatively high. These wind loads are resisted satisfactorily with heavy

roofs, and do not require great ductility in column members. But the

inertial loads at roof level during

strong earthquake excitation are

high, and require great ductility in

support columns. The typhoon design

philosophy does not address this re- quirement, and the structures are damaged during earthquakes.

7. LIQUEFACTION

Parts of the Navy port area, the

commercial port area, and the electric power generation plant site to

the southwest of Agana, are built on filled land. As is usually the case where sufficient precautions against

ground failure, ground spreading and

liquefaction are not taken, these

areas suffered to various extents.

Cracks in the ground surface on the

piers reached down to the water table and beyond, the cranes were or- dered stopped while insurance questions were addressed, and the docksides were no longer straight but

bulged out toward the water.

8. CONCLUSIONS AND LESSONS TO BE

LEARNED

.

No earthquake as large as this, magnitude 8.1, is ever likely to be

closer to Guam, so this earthquake

provided representative shaking for

the big event. The shaking had predominant periods in the range of 0.3

to 3 seconds, coinciding with the

natural resonant periods of the se- riously damaged high-rise structures. Causes of the damage, once the ground motion is given, are typical: architectural layout of the

structures, lack of ductility details, an understanding of the

structural dynamics, material

quality and construction techniques.

Smaller earthquakes have occurred on closer faults, and will do so again,

with a greater content of high frequency ground motion. These will

affect the stiffer low-rise residential structures.

The island absorbed the tourism industry very rapidly, and may have

occasionally outstripped its ability

to provide quality assurance during

all stages of a hotel project.

Filled land, and other bad ground

(for example, soft, amplifiable, and

liquef iable ) , must be identified,

rectified, or must not be built

upon.

Strong ground motion records in the

regions of strong shaking and structural damage are imperative for ac- curate assessment of a damaged building's performance. The effort re- quired to design, fund, and install

a network, no matter how small,

should be followed with successful

efforts for maintenance for the ex- pected instrumentation life.

464
120° 130° 140° 150°

Figure 1. Plate tectonics in the western north Pacific Ocean

465
9. ACKNOWLEDGEMENTS 3. Tracey, J.I., S.O. Schlanger,

J.T. Stark, D.B. Doan, and

H.G. May (1964). "General

geology of Guam." Geological

Survey Professional Paper 403-

A, 105 p.

TABLE 1.

GUAM EARTHQUAKES, 1825-1936.

The author was on the Earthquake

Engineering Research Institute re- connaissance team that visited the

island for a week shortly after the

earthquake. A complete report is in

preparation for Earthquake Spectra.

This paper includes some of the

views of other members of the team.

Paul Hattori, of the USGS Guam Observatory, was a great help with

transportation and local knowledge.

Personnel from Dames and Moore, with

an office on Guam, and from H.J.

Degenkolb Associates, contributed to

the author's understanding of the

earthquake's effects.

10. REFERENCES

1. Brady, A.G. and Mork, P.N.

(1993). "Processing of the US

Navy strong-motion record from

the February 13, 1983 magnitude mb 6.3 Guam earthquake."

USGS Open File Report in

press

.

2. EERI (1993). "The Guam earthquake of August 8, 1993."

EERI Newsletter, v 27, n 10,

12p.

143° 144° 145°

I,.„l,.,.l,,,.l„,,l I I I .I... .I Inn l lllll Nl l l llll l ll l l l llllll l ll l li

Date Max

MMI

April, 1825 7-8

May, 1834 7-8

Jan 25, 1849 8-9

Jul 1, 1862 6

Dec 7, 1863 5-6

Jun 24, 1866 5-6

May 13, 1870 5-6

May 16, 1892 7-8

Sep 22, 1902 8-9

Dec 24, 1902 5-6

Feb 10, 1903 6

Dec 10, 1909 7-8

Oct 26, 1912 5-6

May 10, 1917 5-6

Nov 24, 1917 5-6

Jun 12, 1932 5-6

Oct 30, 1936 7-8

Nov 12, 1936 5-6

Dec 14, 1936 6

^A'-.

13°-!

1975 M7.1 # 1983 M6.3

GUAM

Tumon ^V,

Agana

J^197&

M5.2

1993 M8.1

o ^

$

146°

0 50 KM

1

' >—iA**

l|llll|llll|l l ll

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|

I MI

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|

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|

l l|llll|llll

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| M II

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Figure 2. Earthquakes damaging Guam since 1975, and 1993 aftershocks

466
Infrastructure Damage by 1993 Hokkaido-Nansei-Oki Earthquake

by

Takashi Iijima", Yutaka Iida 2)

, Kazuhiko Kawashima31

, Hisanori Otsuka4

', Osamu Matsuo5

', Shigenobu Tanaka6

', Akira Nakamura71

, Masanori Nakano8

', and Ryosuke Tsunaki9'

ABSTRACT

Seismic damage caused by the

Hokkaido-nansei-oki, Japan, Earthquake of

July 12, 1993 is described. Description is given

to the damage of road, road bridges, river

hydraulic facilities, dams and sewage facilities,

and effect of tsunami, slope failure and soil liquefaction.

Key Words: Earthquake Damage,

Hokkaido-nansei-oki Earthquake,

Road, Bridge, Lifeline Facilities,

Dam, Tsunami, Slope Failure,

Soil Liquefaction

1. INTRODUCTION

The Hokkaido-nansei-oki earthquake

occurred at 22:17 on July 12, 1993 at sea bottom of the south west Hokkaido Island. The

magnitude was 7.8 on Richter scale. Extensive

damage was developed by tsunami, fire, slope

failure and soil liquefaction at the south-west

Hokkaido. In particular, damage was destructive

at Okushiri Island where tsunami attacked soon

after the earthquake.

This paper summarizes the damage

caused by the Hokkaido-nansei-oki earthquake

(Ref. 1). Emphasis is given to the damage of

roads, road bridges, river hydraulic facilities,

dams, sewage facilities, and effect of tsunami,

slope failure and soil liquefaction.

2. THE EARTHQUAKE AND

EARTHQUAKE GROUND MOTIONS

Magnitude of the earthquake was 7.8

according to the Japan Meteorological Agency.

The epicenter was about 80 km from the west

coast of Hokkaido Island, and the focal deptl was about 34 km. The aftershock area extendec

220 km in north/south and 70 km in east/west.

Fig. 2.1 shows the distribution of JMA

Seismic Intensity. It was V at Otaru, Suttsu,

Esashi and Fukaura, and IV at Kutchan,

Tomakomai, Muroran, Hakodate, Mutsu and

Aomori.

A number of strong motion records were

obtained as shown in Fig. 2.2. The peak

horizontal ground acceleration was 496 gal at Kuromatsunai and 386 gal at a free field near

the Shichiho Bridge. It should be noted that the

Kuromatsunai record is not a free field ground

motion. It was recorded on a concrete floor of a

first story JR station. Fig. 2.3 shows the strong motion

accelerations measured at Shichiho Bridge site. The duration is quite long. The bracketed

duration (Ref. 2) with the acceleration larger

than 50 % of the peak acceleration is about 10

seconds. Response accelerations of the records

for 0.05 damping ratio are presented in Fig. 2.4. A high response acceleration of 1,652 gal was

developed at a natural period of 0.15

1) Director-General, Public Works Research Institute, Ministry of Construction

2) Director, Technical Department, Trans-Tokyo Bay

Highway Corporation (Former Director, Earthquake

Disaster Prevention Department, PWRI)

3) Research Coordinator for Underground Development,

PWRI

4) Head, Earthquake Engineering Division, PWRI

5) Head, Soil Dynamics Division, PWRI

6) Head, Coastal Engineering Division, PWRI

7) Head, Fill-type Dam Division, PWRI

8) Head, Foundation Engineering Division, PWRI

9) Head, Slope Failure Division, PWRI

467
second. The ground where the records were

obtained is classified as the Group I (stiff, Ref.

3). Fig. 2.5 shows the attenuation of peak

ground accelerations. Attenuation by an

empirical equation by Eq. (2-1) (Ref. 4) is also

presented in Fig. 2.5 for comparison.

- 987AXl002im X(A+3Q)-l2VS

(stiff) 232.5 X io°-313M x Gd+30)-1 - 213 (moderate)

403.8 X 100M5M X (A+30)~12i3

(soft)

(2-1)

where, Umax : peak ground acceleration (gal), M : earthquake magnitude, and A: epicentral

distance (km). It is seen in Fig. 2.5 that the

acceleration at Kuromatsunai is extremely

higher than the empirical evaluation, and that the attenuation of the measured records in terms

of distance was lager than Eq. (2-1).

3. OUTLINE OF DAMAGE

Table 3.1 shows the human loss and

damage developed in Hokkaido Island. Although some more damage was developed in the Main Island along the Japan Sea, most of

damage occurred in Hokkaido Island. The total damage was about 100 billion yen. The earthquake caused 201 deaths (200

in Hokkaido Island), 33 missings and 236

seriously or slightly injured. Among the 201

deaths, 171 were killed at Okushiri Island by

Tsunami. This corresponds to 3.6 % of the total residents (4,719) in the Island. The number of houses which were

partially damaged or completely destroyed was

3,555 in Hokkaido. Among 567 houses which

were completely destroyed, 410 were located in Okushiri Island. A fire broke out soon after the

earthquake, and it burned down 192 houses.

The damage to road facilities occurred at 88 locations on 7 National Roads and 291

locations on municipal roads. A serious failure

of embankments occurred on National Road 5

at Chirai, Oshamanbe Town. Among rivers with

the jurisdiction of Hokkaido Development

Bureau, shore protections in Okushiribetsu

River and Shiribetsu River were damaged at 43

locations, and other damage was developed at 143 locations in the rivers with the jurisdiction

of Hokkaido government. The damage to civil engineering facilities was about 32 billion yen, which is 32% of the total loss. Water and gas supply was out of service

at 16,000 and 15,000 houses, respectively. The

facilities were recovered by July 25. The

electric power supply was suspended to about

33,000 houses, and the recovery was completed

by July 16. The railway service was restored on

July 18, and the ferry boat and the air flight

service were recovered on July 17.

4. DAMAGE TO ROADS

Damage of roads occurred at the south-west Hokkaido including the Okushiri

Island. The damage included land slides,

surface submergence by tsunami, failure of

embankments, collapse of road shoulders,

settlement/uplift of road surfaces. Fig. 4.1 and

Table 4.1 show the location of major road

damage.

One of the major damage was developed

at a road embankment at Chirai in Oshamanbe-cho on National Road 5 as shown

in Photo 4.1. A 9.5m-wide road embankment

collapsed to a depth of 7.5m along 100m. The

volume of collapsed soil amounted to 5,000m

3

. Fig. 4.2 shows the cross section of the embankment before and after the earthquake.

The embankment was filled on the soft ground

consisting of peat layer. A temporary road for the emergency restoration was constructed on

harder ground near the collapsed embankment

and was put into operation after 12 days

following the earthquake (July 24).

Permanent undulating deformation as shown in Photo 4.2 was developed at a road

section between Oshamanbe and Yakumo on

National Road 5 along Uchiura Bay.

Nakanosawa Elementary School located along

the Road 5 suffered shear failure at the top of

468
piles supporting a new wooden frame building.

Extensive sand boils, cracks, vertical discontinuities and lateral flow in the yard

occurred. Uplift of manholes installed in the

pedestrian way along the Road 5 also occurred.

In this area, iron sand had been produced, and

the holes were re-filled with loose sand. This

caused extensive soil liquefaction.

On a bypass of National Road 5 at Oshamanbe, an embankment and retaining

walls, which were under construction along a

sea shoreline, moved laterally toward the sea by

a distance of several tens centimeter. Manholes

installed in the embankment for surface water

drainage and sewage also uplifted.

On the National Road 228 at Arikawa,

Kamiiso-cho located at west of Hakodate City,

cracks were developed at edges of a concrete pavement slab with length of 10m and width of

7m both in the longitudinal and transverse

directions. The slab moved by about 10 cm in longitudinal direction. The pavement crashed

and uplifted as shown in Photo 4.3. On the new

road under construction along the damaged

existing Road 228, boiling holes and circular

subsidence holes were developed and diagonal

cracks to the longitudinal axis were developed.

The curb moved and slope of the embankment

collapsed with width of about 1.5 m. Cracks were developed at manholes and concrete pipes

installed in the pedestrian way.

A PC arch-type snow-shelter collapsed

as shown in Photo 4.4 on a regional road in Okushiri Island. The structure was just

completed but was not yet in service. The

shelter collapsed due to the overturning of the

foundation toward outside. Since two arch

members were connected only at the arch crown and the other sides were supported by

the foundations with rubber pads, the shelter was statically determinant.

5. DAMAGE TO ROAD BRIDGES

Twenty four bridges were damaged by

the earthquake. Among them, 7 bridges were of

the jurisdiction by the Hokkaido Development

Bureau, 12 bridges by Hokkaido Government

and 5 bridges by municipal governments.

Table 5.1 shows the major bridges

damaged. Damage occurred at reinforced concrete piers, wings and parapets of

abutments, and mortar and anchor bolts of

bearing supports. Major causes of the damage

were as follows (Ref. 5):

(1) Damage to reinforced concrete piers with

termination of main reinforcement without

enough anchoring length

Significant damage was developed at the reinforced concrete piers of Motoe Bridge on a municipal road. The bridge was of 7-span

simply supported plate girder with a deck

length of 165.7m. The deck was supported by

the T-type reinforced concrete piers with

circular section of 1.8 m. The bridge was

constructed in early 1970s. Medium to heavy

damage was developed at 3 piers among 6

piers.

The most significant damage was

developed at Pier 3 as shown in Photo 5.1. Crack and spalling-off of cover concrete, and

outward buckling of longitudinal reinforcement were developed at the section where the part of

longitudinal reinforcements were terminated.

The requirements for anchoring length of the

reinforcements described in the pre- 1980

Design Specifications (Ref. 6) was not

sufficient. This is a typical damage developed

in the bridges designed in accordance with the pre-1980 Design Specifications (Ref. 5). In accordance with the Guide

Specifications for Post Earthquake

Countermeasures (Ref.7), this damage

corresponds to the "Substantial Damage" which

requires to close the bridge for traffic. The similar type of damage was

developed at two more bridges as shown in Photos 5.2 and 5.3. Fig. 5.1 shows the evaluation of seismic

vulnerability for 14 bridges which suffered/did not suffered this type of damage at reinforced concrete piers during this earthquake and the

Kushiro-oki Earthquake of January 1993 (Ref.

469
8). A method proposed to evaluate seismic

vulnerability of such type of reinforced concrete

piers in terms of the Failure Mode Factor S and

Safety Factor F? (Ref. 9) was applied after the earthquakes to show the accuracy of the

proposed method. It is obvious that the

proposed method provides a reasonable

prediction for the damage.

(2) Damage caused by soil liquefaction

Oshamanbe Bridge at Oshamanbe on the

National Road 37 was damaged by the effect of

soil liquefaction. It is of 5 -span simply

supported post-tension T-shaped girder bridge

with deck length of 150.1 m. The foundations were of a caisson type. The bridge was

constructed in 1960, and the pedestrian bridge was added at the upper stream side in 1976.

The bridge piers inclined in the upper

stream direction as shown in Photo 5.4 resulting

in the residual lateral displacement of the deck

of 42 cm. The inclination was 1-2 degree in angle. This caused some damage to the deck

slab and the expansion joints.

Fig. 5.2 shows the soil boring data at the

site. It is considered that the sandy layer at depth of 3.5 - 7 m caused liquefaction resulting

loss of bearing capacity. The Seismic Design

Specifications of Road Bridges (Ref. 10) in which the first requirement for soil liquefaction was included was issued in 1971. Because this bridge was designed before the Seismic Design

Specifications was issued, the consideration

against the soil liquefaction was not adequate.

The bridge piers have an antisymmetric

shape due to the added pedestrian deck, and

this caused tilting of the foundation only in one

direction.

6. DAMAGE TO RIVER DIKES

Flood control earth-fill river dikes

suffered strict failure. Damage occurred at 224

sites with a total length of 29 km. The repair cost including other river works amounted to about 3.9 billion yen.

The damage of river dikes consisted of

cracking, settlement, lateral spreading and

slumping. In many cases, sand boils were

observed at the ground near the toe of dikes.

Such dikes are filled on alluvial deposits

without exception. It is considered that soil liquefaction triggered the damage.

The most extensive damage occurred at the dikes of Shiribeshi-Toshibetsu River and

Shiribeshi River. Typical damage at the rivers are as follows:

(1) Kabutono Dike at Shiribeshi-Toshibetsu

River

Kabutono Dike suffered the most serious

damage at its right dike at the mouth.

Longitudinal and transverse cracking,

settlement, slumping and lateral spreading were

developed for about 2.2 km. Sand boils were

also observed at the sites. Fig. 6.1 shows the cross section. It is 6 m high and 6 m wide at its

crest, with side slope of 2 H : 1 V. The cracks

developed at the crest extended to as deep as 2.3 m associated with slumping of 2 m and

width of 1 m. Occurrence of fissures and sand

boils in ground near and away from the dike

indicates that the liquefaction took place mostly

in material beneath the dike. As shown in Fig.

6.1, the top soil is underlain by fluvial sand

deposit of 3 m thick. This layer has a relatively

low SPT-N value of 10 or less.

(2) Hatsuta Dike at Shiribetsu River

The right-hand dike of the River suffered

damage as shown in Photo 6.1 for 0.89 km

long upstream from the 4.8 km-post. The

damage included longitudinal cracks at the crest associated with slumping of as deep as 1.5 m,

cracks at both slopes, and lateral spreading. Fig.

6.2 show the cross section of the dike as well

as soil profile. There is an extremely loose

fluvial sand deposit from 7 m below the ground

surface. Many fissures and sand boils were

observed between the dike and river. It is obvious that the damage was developed due to

soil liquefaction.

Fig. 6.3 shows how damage rate of river dikes decreases as the epicenrral distance

increases. The damage ratio is defined as a ratio

470
of the length of river dikes damaged and the

total length. The damage ratio for other

previous earthquakes such as 1964 Niigata

Earthquake and 1978 Miyagi-ken-oki

Earthquake (Ref. 11) is presented in Fig. 6.3 for comparison. It is seen that the damage ratio for

the Hokkaido-nansei-oki earthquake shows the

similar attenuation with the previous

earthquakes. Such result may be valuable to evaluate the damage degree for a certain

combination of an earthquake magnitude and

epicentral distance.

Fig. 6.4 shows the damage ratio vs. surface geology relation. The surface geology is represented in terms of the micro-landform

classification. It is obvious from Fig. 6.4 that the damage ratio is considerably high at the

former river bed. It is important to note that the

river dikes on the river beds tend to cause

substantial damage due to an earthquake.

7. DAMAGE TO DAMS

Immediately after the earthquake, an emergency inspection was made for 52 dams: 6

dams are owned by the Ministry of

Construction (M.O.C.), 11 dams were

subsidized by M.O.C., and 35 other dams are

for water utilization. The inspection revealed

that among dams with height of 15 meters or more, the Niwa-Ikumine Dam suffered

moderate damage. Minor damage was

discovered at the Pirika Dam and the

Makomanai Dam. But they were not so severe

as to affect the safety of the dams (Ref. 12).

(1) Pirika dam

Pirika Dam is a combined dam consisting

of a concrete gravity dam and a rockfill dam.

The dam was located 89 km from the epicenter. In the structure between block 0 of the concrete dam body and the connecting corridor

leading to the control office, a floor panel

concrete of the connecting corridor was pushed

up and spalled in the length of about 1.5 meters

due to compression in axial direction of the

dam. Slight concrete spalling was also

discovered in other cross joints of the

connecting corridor. This connecting corridor

rides on the rock zone of the core wall at the

left bank connected perpendicularly to the concrete dam. It was therefore subjected to severe compression caused by differences of the

earthquake response between the dam body and

the rock zone. Because the connecting corridor

and the concrete dam body are completely

separated and there was no deformation in the

inspection gallery for block 0, the deformation

did not effect the dam body itself. After the earthquake, water leaked from

the intersection of the inspection observation

gallery and the blocked section of the

temporary diversion channel inside the dam on

the block 28. This section was sealed after leakage was observed during the previous test

filling. Because since then, no leakage was

observed at the sealing, it is considered that a small gap opened in the sealing due to the

earthquake causing leakage. The leakage was

small (between 80 and 100 milli-liters per

minute) after 10 days, and it was not turbid. Therefore the leakage is not considered to be a

serious problem.

(2) Niwa-Ikumine Dam

Niwa-Ikumine Dam is a central core type

earth fill dam. It is an old dam completed in 1927. Fig. 7.1 shows a plane of the dam.

The earthquake caused bulging and

cracking of the dam body at two elevations,

and undulating deformation of the upstream

slope. The bulging appeared in the middle and

close to the toe of the upstream slope. The

maximum level difference of the bulge at the toe of the slope was about 1 meter over the 40

meter length. The cracking consisted of five

longitudinal cracks at the crest. The longest

crack extended for a total length of

approximately 120 meters on the crest of the

dam, with the depth of 2 meters. It settled on

the upstream side, and its maximum level difference was about 1 meter. Fig. 7.2 shows

the cross sections at 10 m interval measured

from the point 80 m apart from the right bank

471
spillway. This point is almost the center of the

dam. It is believed that the earthquake caused

the liquefaction of the sandy stratum and the

settlement of the upstream side of the dam

body. Sand boil was seen on the surface of the

sediment on the bottom of the reservoir near

the toe of the upstream slope of the dam body.

The bottom of the dam body and river-bed

sedimentary stratum consist of loose sand with

a small N value.

(3) Makomanai Dam

Makomanai Dam is a central core type

rockfill dam with a height of 34 meters. It was

completed in 1987 and located 66 kilometers

from the epicenter.

The dam body damage included

deformation of the concrete frame on the

upstream slope, and the riprap had shifted. In the ground, the facing concrete blocks on the

right bank had undergone undulating

deformation, sprayed concrete on the left bank

slipped, and the small-scale failure had

occurred on the slope of mountain. None of

these damages presented any particular

problems related to the safety of the dam body.

Immediately after the earthquake,

increase in leakage water was observed at six dams, and the amount of leakage water

decreased at 2 dams. The turbidity of the

leakage water increased at one dam. However,

these changes in leakage water were not considered to indicate serious problems,

because the leakage water returned to its pre-earthquake state within a period ranging

from a few days to a few weeks.

Furthermore two small irrigation ponds

with dike of less than 15 meters high located

only 65 and 70 kilometers from the epicenter in Setana-cho in Hokkaido were destroyed. And

along the Japan Sea Coast near the epicenter, some dikes of this kinds were damaged by

settlement or cracking.

8. DAMAGE TO SEWAGE FACILITIES

Damage of sewage facilities was

developed at 8 cities and towns at the west

south of Hokkaido. About 2/3 of the total damage concentrated at Oshamanbe Town, in which about 10 % of the sewage pipes suffered

damage. Hume pipes and polyvinyl chloride

pipes are generally used for sewer. Damage included break and offset of

pipes, leakage, damage of side wall of

manholes and uplift of manholes (Ref. 13).

Only minor damage such as nonstructural

damage and break of window was developed at sewage treatment plants. Damage of small

diameter pipes was surveyed by small video cameras which either crawled or are pulled into the pipes. Break of pipes and joints, dislodging

of pipes, pull-out of pipes at joints and leakage

of water could be clearly identified. Photo 8.1 shows the damage detected by such a small

video camera. Table 8.1 shows the type of pipe and

joint failures developed at Oshamanbe Town.

Permanent settlement of pipes and leakage of

water were the most frequently developed

damage. The damage of polyvinyl chloride

pipes was less than the damage of hume pipes.

Most of uplift of manholes as shown in Photo 8.2 was developed due to soil liquefaction. The uplift of manholes was as high as 50-80 cm. Fig. 8.1 (a) shows the uplift

of 10 manholes on a same sewer. For analyzing

the uplift, the Liquefaction Resistant Factor Fl

(Ref. 14) was evaluated for each manhole, and

the safety of uplift was evaluated as (Ref. 15)

in which Fs: safety factor for uplift, Wm\ weight

of manhole, F: friction between the soils and

manhole, Us: uplift force due to static water

pressure, and Ud: uplift force due to pore water

pressure. The pore water pressure was evaluated

from the Liquefaction Resistance Factor Fl

(Ref. 14).

Fig. 8.1 (b) and (c) shows the Liquefaction Resistance Factor Fl and Safety Factor Fs. The safety factor Fs is minimum at

472
manhole No. 6 where the largest uplift was

developed. Thus it is obvious that some

correlation is seen in Fig. 8.1. More data are required to verify the accuracy for predicting

the uplift.

9. DAMAGE DUE TO TSUNAMI

Okushiri Island suffered the most

destructive damage due to tsunami. In

particular, the damage concentrated at Aonae

district (southern part) and Inaho district (northern part) in the Island. Photo 9.1 shows a village backyard at Inaho fishery harbor. Even houses located on a

hill side of a road were completely destroyed.

The tsunami inundation height was as high as

8.9 m at this site. Aonae district suffered much more

destructive damage due to the tsunami. Photo

9.2 shows the north-east of Aonae fishery

harbor. Even though this site was sheltered by

Aonae Cape from the tsunami source, it suffered such destructive damage. It is considered that tsunami came over the site by

refraction and diffraction. Some witnesses who

saw that the highest tsunami attacked the Aonae

district from the north-east to south-east

confirm this. The tsunami inundation height was 7.4 m at this site. The tsunami inundation height measured

by PWRI members is presented in Fig. 9.1

(Ref. 16). It was high at Shimamaki village and

the coast from Setana to Taisei town. The

tsunami height was as high as 3 m. At Okushiri

Island, the survey of tsunami height was made

only at east coast, because population is more

dense at this side. The highest tsunami

inundation height was 19.8 m at Kuki cape. The tsunami inundation height decreased from

south to center, because the tsunami attacked

from the west. It should be however noted that the maximum tsunami height of 15.9 m was

observed at Hamatsumae district even though

this district is sheltered by Aonae Cape. It must

be also pointed out that the maximum

inundation height was 31.7 m at Monai. This is

the highest tsunami inundation height ever experienced in this century in Japan.

10. DAMAGE DUE TO SLOPE FAILURE

Slope failures concentrated in Okushiri

Island and in the western region of the Oshima

Peninsula. This section shows two slope failures

at Okushiri Island as shown in Fig. 10.1.

(1) Failures in the Okushiri District Most of death in the Okushiri District in Okushiri Island was developed by landslides. A

slope collapsed as shown in Photo 10.1 at the

terrace behind Okushiri Harbor. The failure was

200 m wide and 120 m high. Downslope of

about 100,000 m

3 of debris destroyed the

Hotel Yoyo-so.

The geological structure of the collapsed

slope consisted of surface soil, sediments, tuff breccia, tuff sandstone and tuff of the Neogene

lacustrine epoch and Pliocene Hotokesawa layer

from the surface. The average inclination of the

slope after failure was 40 degree. The debris

moved a maximum 70 to 80 meters toward the

sea. A considerable amount of earth and rock mass which was observed in the middle and

bottom parts of the slope, retained its original

shape, and is estimated to have fallen along the shear plane. A flat plane, which nearly

corresponds to the terrace plane before the earthquake, remains intact and is wooded at the

uppermost part of the remaining earth mass. Rock masses of maximum diameter of 5 to 7

meters are included in the earth mass. Such slope failures have led to speculation that the earth mass did not move at once, instead, a certain range of the lower slope

section probably collapsed first, then the

unstable upper part slipped down.

(2) Failures in the Monai District Numerous slope failures and falling rocks were observed along roads in the Monai

District as shown in Fig. 10.1. A considerable

number of rocks fell, especially near the

Hoyaishi District, from steep cliffs and slopes

of tuff breccia, greatly damaging retaining walls

473
and roads. Earth movement similar to landslides, and fluctuation of colluvial deposits

of considerable range and thickness, which

were believed to include weathered rock layers, were also recognized.

(3) Influence of knick lines on failure risk

Investigation on slope failures caused by

earthquakes confirmed that slopes with knick

lines easily collapse by earthquakes.

Topographic analysis was conducted for the

slopes near the Horonai River, where most

slope failures were observed. The influence on

failure risk by projecting forms on slopes

formed by knick lines was investigated.

The central region of Okushiri Island was

subjected to analysis. The number of slope

failures within this range was 44. They have the

slope height ranging from 30 to 120m.

Although the slope height is one of the most

controlling factors for damage vulnerability, a

precise evaluation for the effect of slope height was not made in this analysis. This requires

further analysis.

Fig. 10.2 shows the results of the

analysis. Here it is seen that the steeper the

slope, the higher the risk if the degree of

knicking is the same. It is also seen that the

vulnerability slope failure increases as the

degrees of knicking increases if the slope

inclination is the same.

11. CONCLUDING REMARKS

The preceding pages presented a summary of the damage caused by the

Hokkaido-nansei-oki Earthquake of July 1993.

The feature of the earthquake damage may be

summarized as:

(1) The tsunami caused destructive damage and

loss of many human lives. The highest tsunami

inundation height ever experienced in Japan in

this century was recorded. It is important to take the refraction and diffraction into account

for tsunami countermeasures.

(2) Bridges suffered at the reinforced concrete

piers with termination of main reinforcement

without enough anchoring length. The piers

which suffered damage can be predicted with

sufficient accuracy by a vulnerability evaluation

method proposed at PWRI. Seismic

strengthening of such type of piers is required.

(3) Most of soil structures such as road

embankments and river dikes suffered damage

due to soil liquefaction under and around the embankments. It is required to develop

appropriate and cheap measures to decrease the damage.

(4) Aside from old dams and dikes of irrigation

pond constructed with experience-based

construction technology, those dams constructed on rock foundations with modern design and

execution technology survived this powerful

earthquake without any adverse effects to maintain their safety, and that they are still in sound condition.

(5) Extensive slope failure was developed at Okushiri Island. It was confirmed that the steeper the slope, the higher the risk if the degree of knicking is uniform. It was also confirmed that slopes tend to collapse more

easily at higher degrees of knicking if slope

inclination is uniform.

(6) Sewage damage concentrated at pipes and

manholes. Small TV cameras were effectively

used to identify the damage. Uplift of manholes was as high as 50-80 cm, and they were

developed due to soil liquefaction.

ACKNOWLEDGEMENTS

For conducting the post earthquake

damage investigation, invaluable supports were

provided by Hokkaido Developing Bureau,

Hokkaido Government, and other local governments. The authors express their sincere

appreciation for the warm supports of the

organizations and personels concerned. A post

earthquake damage survey was jointly made on

July 20-23, 1993 with the U.S. Damage Study

Team headed by Dr. Riley Chung, Building and

474
Fire Research Laboratory, National Institute of

Standards and Technology as one of the

technical cooperation through the Panel on Wind and Seismic Effects, UJNR. Various

discussions with the U.S. Damage Study Team

members were extremely valuable for the Japan-side. A great appreciation is directed to Dr. Richard N. Wright, U.S. Chairman of the

Panel and Mr. Noel J. Raufaste, Secretary of

the Panel for their cooperation.

REFERENCES

1) Public Works Research Institute: Report on

the Hokkaido-nansei-oki Earthquake,

PWRI report, in preparation (in Japanese)

2) Kawashima, K. and Aizawa, K. : Bracketed and Normalized Duration of

Earthquake Ground Acceleration,

Earthquake Engineering and Structural

Dynamics, Vol. 18, 1989

3) Japan Road Association : Part V Seismic

Design, Design Specifications of Road

Bridges, 1990

4) Kawashima, K. and Aizawa, K.: Attenuation of Peak Ground Acceleration,

Velocity and Displacement Based on

Multiple Regression Analysis of Japanese

Strong Motion Records, Earthquake

Engineering and Structural Dynamics, Vol.

14, 1986

5) Kawashima, K., Unjoh, S. Hoshikuma, J. and Nakajima, T. : Damage of Road

Bridges by Hokkaido-nansei-oki

Earthquake of July, 1993,Bridges and

Foundation, Vol. 94-3, March 1993 (in

Japanese)

6) Japan Road Association : Part IV

Substructures, Design Specifications of

Road Bridges, 1980 (in Japanese)

7) Japan Road Association : Guide

Specifications for Earthquake Disaster

Prevention Measures of Road Facilities (Post Earthquake), 1988 (in Japanese)

8) Kawashima, K., Hoshikuma, J. and Unjoh,

S. : Damage Analysis of Reinforced

Concrete Bridge Piers by the Kushiro-oki

Earthquake of January 1993 and the

Hokkaido-nansei-oki Earthquake of July

1993, "nd U.S.-Japan Workshop on

Seismic Strengthening of Bridges,

Berkeley, California, U.S.A., January 1994

9) Kawashima, K., Unjoh, S. and Iida, H. : Seismic Inspection and Seismic

Strengthening of Reinforced Concrete

Bridge Piers at Mid-height where Main

Reinforcements are terminated with In Adequate Anchoring Length, Report of

PWRI, Vol. 189, Public Works Research

Institute, September 1993 (in Japanese)

10) Japan Road Association : Seismic Design

Specifications of Road Bridges, 1971(in

Japanese)

11) For example, Hashimoto, H. et al: Chapter

8 Damage to River Facilities, Report on

the Disaster by the Nihon-kai-chubu

Earthquake, Report of PWRI, No. 165,

Public Works Research Institute, 1985 (in

Japanese)

12) Nakamura, A., Yasuda, N. and Iwashita, T.

: Damage and Analysis of Dams during

1993 Kushiro-oki Earthquake and 1993

Hokkaido-nansei-oki Earthquake, 26th

Joint Meeting, UJNR Panel on Wind and

Seismic Effects, Gaithersburg, MD,

U.S.A., May 1994

13) Ninomiya, Y., Otsuka, H. and Azuma, T. : Damage of Sewage Facilities due to 1993

Hokkaido-nansei-oki Earthquake, 29th

Japan National Conference on Soil

Mechanics and Foundation Engineering,

Morioka, June 24027, 1994 (in Japanese)

14) Japan Road Association : Design

Specifications of Common Utility Ducts,

March 1986 (in Japanese)

15) Matsuo, O., Otsuka, H., Ninomiya, Y. and

Koseki, J. : Uplift of Sewage Manholes

during the 1993 Kushiro-oki Earthquake,

26th Joint Meeting, UJNR Panel on Wind

and Seismic Effects, Gaithersburg, MD,

U.S.A., May 1994

16) Tanaka, S., Sato, S. and Noguchi, K. : Propagation of Hokkaido-Nansei-Oki

Earthquake Tsunami around Cape Aonae,

475
26th Joint Meeting, UJNR Panel on Wind

and Seismic Effects, Gaithersburg, MD,

U.S.A., May 1994

For example, Yasue, T., Iwasaki, T.,

Kawashima, K. and Nakano, M. : Seismic

Response of Slopes, Civil Engineering

Journal, Vol. 23, No. 4, 1981 (in Japanese)

476
Table 3.1 Damage and Loss in Hokkaido Island

Category Hokkaido (Unit)

Death

Missing

Serious Injured

Slight Injured

200 (Person)

33 (Person)

39 (Person)

197 (Person)

Number Residental HousesDamaged

Number Non-residental HousesDamaged

3,913 (Houses)

736 (Houses)

Loss

Agricultural Products 11,909,918 (Thousand Yen)

Public Works 31,422,824 (Thousand Yen)

Fishery 13,229,326 (Thousand Yen)

Forestry 21,629,977 (Thousand Yen)

Sanitation 989,701 (Thousand Yen)

Commerce and Industry 9,763,154 (Thousand Yen)

Education 2,523,941 (Thousand Yen)

Social Education Facilities 468,461 (Thousand Yen)

Welfare Facilities 472,530 (Thousand Yen)

Others 721,881 (Thousand Yen)

Total 100,702,194 (Thousand Yen)

477
Table 4.1 Major Damage to Road Facilities

No. Damage Type Site

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1

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17 Road Surfaces Mena in Imakane Town on National Road 230

18 IVUllllUl 111 KJay alllallUC 1UW11 Ull l>iallUllal IVUaU £DVJ 19 Yamazaki in Yakumo Town on National Road 277

20 Nodaoi in Yakumo Town on National Road 5

71 rVXalgaWd AVai oa\x/a in111 IMori IVlUll Tow/n 1UW11 on Nation a 1 R oaH ^ Ull l^iallUllal IVUaU J

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23 Uplift of Kamiiso in Kamiiso Town on National Road 228

IxUdU OUXlaCca

24 Land Slides No.2 Shiraito Tunnel on National Road 229

25 Section Facing Okushiri Port in Okushiri Island

26 Hntamata in Taisei Town nn National Road 929

27 Surface Setana Town on National Road 229

28 Submergence Miyano in Taisei Town on National Road 229

29 by Tsunami Esashi Town on National Road 227

30 Damage to Senjouzaka in Okushiri Island

Snow Shelters

(Note) Refer to Fig.4.1 for location.

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Fig. 2.1 JMA Seismic Intensity

145

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Fig. 2.2 Peak Ground Accelerations (Two Horizontal Components)

481
482
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483
484
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Safety Factor F/

Fig. 5.1 Evaluation of Seismic Vulnerability of Reinforced Concrete Piers without

Enough Anchoring Length at Termination Section of Main Reinforcements

485
Fig. 5.2 Soil Boring Data at Oshamanbe Bridge

N- value

0 103)304050 N- value

n in m m tn Sv-:- N- value

0 103)304050

Fig. 6.1 Cross Section of Kabutono Dike (Shiribeshi-Toshibetsu River, Left-Side Dike)

486
N- value

Fig. 6.2 Cross Section of Hatsuta Dike (Shiribetsu River, Right-Side Dike)

40-

30

20-

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Niigata Eq.

JL

Hokkaido - nansei - oki Eq.

JL

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Miyagi - ken - oki Eq.

fro, n > -^n i 20 30 40 60 60 70 80 90 100 110 120 130 HO ISO IG0 170

Epicentral distance(km)

Fig. 63 Epicentral Distance vs. Damage Rate of River Dikes

487
Fig. 6.4 Micro Land Form vs. Damage Rate of River Dikes

488
489
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Boring No.

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Boring No.

(c) Safety Factor Fs by Eq. (8.1)

Fig. 8.1 Dependence of Uplift of Manhole on Liquefaction

Resistance Factor Fl and Safety Factor Fs by Eq. (8.1)

490

Fig. 10.1 Locations of Slope Failures

492
V

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6.000-

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^ Total length of knick line

£ km

3 T/

0. 44 km

3.09km

17V

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£>>t/ /4 97km

^0 96 km

10.52 km

Types of convex

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Type C - I

30 -45*

Type B - 3

30>43'

Type C - 2 Type D

Angle^of

knick point 150 - 180* 150 - 180* 135 - 165* 150 - 180* 135 - 165" 120 - 150

Slope

gradient 15 - 30* 30 - 45* 45 ~ 60*

Number of places where slope failure occurred per 1km of knick line

Fig. 10.2 Dependence of Slope Failure Occurrence Rate on Type of Convex Cross-Sectional Shape

493
Photo 4.1 Collapse of An Embankment (National Road 5)

Photo 4.2 Permanent Undulating Deformation of Road Surface (National Road 5)

Photo 4.3 Crash and Uplift of Pavement (National Road 228)

494
Photo 4.4 Collapse of A Snow Shelter (Okushiri Island)

Photo 5.2 Damage of Pier P3 at Shin-shiriuchi Bridge

495
496
Photo 8.1 Pipe Damage Detected by A Small Video
Photo 9.2 Tsunami Damage at Aonae Fishery Harbour

Photo 10.1 Slope Failure at Okushiri District

498
12 July 1993 Hokkaido-Nansei-Oki Earthquake

by

Riley M. Chung 1

ABSTRACT

An earthquake measuring 7.8 on the Richter scale struck northern Japan at 9:17 a.m., EDT, Monday,

July 12, 1993. The quake, with epicenter about

80km west of Hokkaido Island, was the largest

Japanese earthquake in the past 15 years. Total life loss has reached 23 1 , most of them from the tsunamis

and fires. Total dollar loss was estimated at $1B.

Immediately after the earthquake and under the auspices of the U.S. -Japan Panel on Wind and

Seismic Effects, the U.S. dispatched a team to jointly

investigate the impact of this earthquake to the built environment. The earthquake provided an

opportunity to study the performance of wood-framed

buildings, the causes of and response to fires, the

effect of tsunami, and the performance of lifelines and other infrastructure systems.

Damage to the built environment was primarily

caused by tsunamis, fires, landslides, and ground

failures, i.e., liquefaction, sand boils, and lateral spread. Okushiri, a small island off the coast of

Hokkaido and 48km south of the epicenter, was

devastated by the fires, tsunamis, and landslides. Buildings and houses performed well, with minor

damages noticed in a some instances. Most bridges

suffered no damage. Those with damage experienced

failure at the interface of the bridge deck and their abutments. Lifeline systems, especially those in Hokkaido, performed well. Those that had been

damaged were restored quickly after the quake.

Tsunami warnings were issued promptly by Japanese

authority. However, they were still not quick enough

to reduce the casualties in coastal villages of

Okushiri Island and a few others along the west coast of Hokkaido. Experience from the 1983 NihonkaiChubu tsunami helped the villagers' prompt

evacuation.

INTRODUCTION

The powerful earthquake measuring 7.8 on the Richter scale struck northern Japan at 9:17 a.m., EDT, Monday, July 12 (or 10:17 p.m., local time).

The quake, with epicenter about 80 km west of

Hokkaido Island, was the largest Japanese earthquake

in the past 15 years (Figure 1). The earthquake

provided U.S. and Japanese researchers and

practitioners an opportunity to study the performance

of the built environment against earthquake forces. Immediately after the quake, under the auspices of

the U.S. -Japan Panel on Wind and Seismic Effects,

United States, dispatched a team to visit the disaster

sites. The team stayed in the field for about one

week, teaming up with a Japanese team led by the Public Works Research Institute, to conduct

postearthquake investigations.

Okushiri, a small island southwest of Hokkaido and

48 km south of the epicenter, was devastated by the quake, resulting from tsunamis, fires, and landslides (Figure 2). Less damage to buildings, roads,

railroads, tunnels, bridges, and other lifelines were

observed on Hokkaido. As of September 1993, the

final count of life loss was at 231, most of them

tsunamis, and the total dollar loss was placed at around $1B.

The teams were to collect perishable data and observe

and assess the performance of various

structures against seismic loading. Key observations

and conclusions from each of the studied areas are given in this paper.

1 . National Institute of Standards and

Technology

499
EARTHQUAKE MECHANISM AND

STRONG GROUND MOTION

The subduction plate boundary along the Japan Sea

coast of northern Honshu and Hokkaido has only

been recognized recently. It is estimated that subduction may have begun about one million years ago at a rate of 1 cm/yr, which would have resulted

in only 10 km of subduction to date. The earthquake

struck on July 12 at 22:17 p.m. local time at latitude 42°47' North, longitude 139°12' East. The

aftershocks define a zone about 150 km long (northsouth) and 30 km wide (east-west) having a predominant depth range of one to 20km (Figure 3).

The earthquake was not preceded by any anomalous

geophysical measurements. The currently available data appear to be mostly consistent with a rupture

model in which the northern part of the earthquake

involved westward subduction and the southern (and

predominant) part of the earthquake involved

eastward subduction. However, the interpretation is preliminary, and a conclusive interpretation will evolve only after a considerable period of data

analysis and interpretation.

Strong ground motions were recorded on a variety of

instrument types by numerous agencies. There were

no strong motion recordings from Okushiri Island or

the immediately adjacent southwest coast of Hokkaido

between Taisei and Motsutanosaki. At a cemetery on

Okushiri, 80 percent of the tombstones were

overturned, suggesting peak accelerations in the ranges of 0.4 to 0.5g. On the immediately adjacent

mainland of Hokkaido, there was no major structural damage to buildings, and few gravestones were

overturned, suggesting that the ground motions were

not unusually high.

The duration of strong ground motion was

approximately one minute at most locations within

about 100 km from the rupture surface, which is compatible with the source duration of about one

minute inferred from teleseismic seismograms, and is expected from an earthquake of this magnitude.

Comparing the attenuation relationships for subduction earthquakes for soil sites currently used in the United States, the peak accelerations recorded

during this earthquake are generally larger for distances ranging from 70 to 150 km and attenuate more rapidly than the attenuation relations (Figure 4). This rapid attenuation is interpreted as being due to higher anelastic absorption as the result of the thermal effects of volcanism.

TSUNAMIS

The earthquake produced one of the largest tsunamis

in Japan's history. Within 2 to 5 minutes, extremely

large waves engulfed the Okushiri coastline and the

central west coast of Hokkaido (Figure 5). At

Okushiri, the tsunami was refracted by the shoaling

bathymetry at both ends of the island. Extensive

damage occurred at both southern and northern tips

of Okushiri, particularly in the south at the town of Aonae. Tsunami vertical runup measurements varied

between 15 to 30 m over a 20-km portion of the southern part of Okushiri Island, with several 10-m

values on the northern portion of the island (Figure

6). Along the west coast of Hokkaido, no survey

values exceeded 10m, but damage was extensive at several coastal towns. Given the sudden onset of the tsunami and its high energy, it is amazing that the life losses were comparatively low.

CONFLAGRATION

The town of Aonae, a fishing village of about 1 ,600

population, experienced the only known fire ignitions as a result of the earthquake. The fires eventually

consumed 190 houses and buildings over an 11 hour

period. Buildings (or houses) are typically one- and

two-story post and beam (heavy timber) construction,

although some steel and concrete (commercial)

structures were also present. Exterior coverings are generally non-combustible cement board cladding over wood, with corrugated metal roofing. Typically

each home has a 490 L kerosene tank for heating

which sat on a steel rack on cement blocks, and were

not anchored down; and also a pair of propane tanks

for cooking, typically 20 kg. The propane tanks are normally not secured.

Two fires broke out after the earthquake. The first

fire began in a structure above the area directly affected by the tsunami. Fire department officials thought the fire began from an overturned kerosene

500
tank; the ignition source is unclear since power was

knocked out by the quake. Villagers told of the quake overturning all their furniture, so numerous

ignition sources such as cooking and heating

appliances are possible.

The fire spread was relatively slow because of low

wind velocity. The fire fighters had been battling the

fire for about 2 hours when a second fire broke out

to their rear near the shore in an area which had been

devastated by the tsunami. Finding themselves

between two fires, the firefighters took up a new

position to the south of the second fire and continued

to fight the fire. After 5-1/2 hours, they ran out of

water in the last accessible cistern and decided to

draft sea water but access to the sea was blocked by

debris. Eventually the fire fighters had to knock

down houses on the fire side of a road to widen the

fire break and as a result, saved the remainder of the town.

The principal mechanisms of building-to-building fire spread involved ignition of exterior wood trim, radiant ignition of combustible contents through

windows, spread along scrap wood between

buildings, and the burning of kerosene and propane

tanks.

LANDSLIDES

Landslides were triggered throughout Okushiri Island and throughout much of the southern peninsula of

Hokkaido. The slides consisted mainly of rock falls and rock slides on steep (>50°) natural slopes and

engineered cut slopes. Volumes ranged from several cubic meters to the 800,000 m

3 rockslide that buried

the Yo Yo Hotel at Okushiri Port (Figure 7). As of

July 24, 1993, there were 10 confirmed casualties and 10 still missing at the hotel site. The scarp of

this major rock slide is a planar to slightly concave outward surface that is highly weathered. This

suggests that the scarp is a pre-earthquake fracture

that provided a structural discontinuity or surface of

weakness along which failure took place.

Numerous rock falls and slides took place along the coast of the island where extremely steep slopes of

volcanic breccia, basalt, and tuffaceous sandstone

shed rock debris that was shaken apart along highly

weathered fractures.

The occurrence of rock falls and rock slides triggered

by the earthquake on the southern Hokkaido

Peninsula was similar to that on Okushiri Island. These types of failures appeared to be most numerous on the western coast of the peninsula where the

steepest slopes exist. About 15 km north of Setana on Highway 229, one of the sections of a talus shed,

made from concrete ducts, collapsed when debris

from a rock fall struck the concrete duct work

(Figure 8). Total volume of the rock fall was several hundred cubic meters and originated from a near

vertical cliff about 60 m above the highway.

The distribution and types of landslides triggered by

the earthquake on Okushiri Island and the southern

Hokkaido Peninsula were characterized by the absence of small rockfalls and rock slides from both

natural and cut slopes. This absence was, in part, due to the wide fracture spacing within the andesitic volcanic breccia that is predominant along the coasts of these geographical locations. The frequency

content of strong shaking during the earthquake may

also be another factor influencing the size of failures and fragment size. Preliminary strong motion

recordings indicate that the accelerations were somewhat low for a magnitude 7.8 earthquake and

that frequencies of 5 Hz and greater are lacking. The

long duration of about one minute and relatively low

frequencies of shaking would thus favor displacement

of larger blocks of rock and individual boulders of one meter diameter and larger.

GEOTECHNICAL EFFECTS

The earthquake caused widespread liquefaction in the southwestern part of Hokkaido and at one location near Aonae on Okushiri Island. The areas affected included alluvial river valleys on the west coast of

Hokkaido, alluvial plains along the east and south

coasts of Hokkaido, and filled ground at several

ports. On Okushiri Island, liquefaction developed in an alluvial plain near the Aonae airport. The

observed effects included sand boils, fissure, lateral spreads, ground oscillation, loss of bearing strength,

ground settlements, and the buoyant rise of manholes

and buried tanks. These effects caused damage to

501
constructed works, but in several notable instances

well-built structures survived the occurrence of

liquefaction without significant damage.

Most of the ports on southwestern Hokkaido and on

Okushiri Island suffered some damage, typically

seaward tilt of quay walls accompanied by fissures,

ground settlements, and sand boils in filled areas behind the walls. None of the damages was

catastrophic, nor were there any major disruptions to port operations as a consequence of liquefaction. The

most severe damage occurred at Pier 1 of Hakodate

Port (Figure 9). Fortunately, Pier 1 is an older part

of the port which was used primarily for the docking

of unused ships. Its severe damage therefore did not

affect the port operation. At Pier 1, large ground

fissures and settlements disrupted an asphalt-paved

apron behind the quay wall, and the wall displaced or

tilted seaward by as much as 3 m. The fissures were

as wide as 1 m and generally paralleled the wall. Ground settlements in the filled area were as great as 0.5 m.

Several storage tanks were damaged during the quake. At Pier 16 of Hakodate Port, a cement

storage tank, 20 m high and 8 m diameter, tipped due

to loss of bearing strength as a result of liquefaction

of surrounding soils (Figure 10). The tank tipped

about 2.5° southward during the quake. The most

spectacular failure of a tank was the buoyant rise of

a buried oil storage tank at an industrial facility

southwest of the Highway 229 bridge over the Assabu river, about 6km north of Esashi. The tank was 7.2 m long with a diameter of 1.8 m, and was

about one-third full of oil at the time of the earthquake (Figure 1 1). It was shallowly buried with

a cover slab and few tenths of a meter of soil above

the tank.

Liquefaction and consequent ground failure also contributed to the outages of many lifeline systems,

which were reported after the quake. Japanese utility

personnel responded quickly, and most of the damage

was repaired and service was resumed by the time the study team arrived at the sites one week after the quake.

The study team also observed many buildings that were constructed in areas where minor to rather

severe liquefaction effects occurred. Most of the buildings performed well, some exceptionally well; but a few suffered rather severe damage. The most

spectacular damage was at the Oshamanbe

Nakonosawa Elementary School, located west of

Highway 5 in Oshamanbe. The school consists of

two main structures, a wood-frame classroom

building with glue-laminated beams and a steel-frame gymnasium (Figure 12). They were surrounded by

grassy open areas and playing fields. The area beneath and surrounding the buildings had been

filled. Liquefaction occurred in sediments beneath

the site generating widespread sand boils, ground

fissures, and ground displacements. Despite these ground disturbances, other than fractured piles and

minor cracks in foundation walls, the buildings were undamaged. An overwhelming majority of houses

and small buildings performed well in areas of minor

liquefaction-induced ground disturbance examined by

the study team. One reason for the excellent performance of these structures may have been the strength of their foundations, which consist of a gridwork of well reinforced and interconnected wah

footings.

Several bridges suffered minor damage as a consequence of liquefaction-induced ground

displacements. The most common disruption of

bridge sites was settlement of the approach fill due to compaction of the embankment materials. These

settlements produced vertical separations as great as a few tenths of a meter between the approach

pavements and the bridge decks. The most common

bridge damage associated with liquefaction was

generated by the lateral displacement of abutments

toward river channels (Figure 13). These

displacements were most likely caused by lateral spread of floodplain sediments toward river channels,

but may also have been caused by inward rotation of

abutment walls due to compaction-induced increases

in lateral forces. The abutment displacements

crowded walls into bridge stringers and compressed

railings and other linear features spanning the bridges.

The flow of traffic was disrupted at more than 350

locations in the highway system. Many of those

disruptions were caused by the settlement of bridgeapproach fills, or by landslides or tsunami debris.

502
Another major source of such disruption was

liquefaction. In several areas where highways

crossed terrain underlain by liquefiable soils, the highway fills settled differentially, deforming the roadway into long vertical waves with troughs as deep as 0.6 m. And the roadway took on the appearance of a low roller coaster.

Kilometers of flood-control dikes along rivers in southwestern Hokkaido were damaged due to liquefaction of subsurface sediments (Figure 14). The dikes usually split longitudinally, approximately

down the centerline of the crest, and subsided.

Liquefaction of sediments beneath the dike allowed

the core to penetrate downward into the liquefied

layer. The downward penetration apparently

produced a type of bearing capacity failure that thrust the flanks of the dikes outward and upward,

generating fissures with dike-facing scarps along the outer margins of the flanks. Simultaneously, the dike

split into two approximately symmetrical segments,

each penetrating into the liquefied soil, but each

being carried outward by the displaced soil. STRUCTURAL ASPECTS

In general, shaking damage to buildings on Hokkaido

and even on Okushiri Island was light to non-existent,

while tsunami damage on Okushiri Island varied from

total loss for wood buildings severely inundated, to moderate for concrete or steel buildings well anchored to their foundations. For concrete or steel buildings, tsunami inundation resulted in partial or

total loss of contents of cladding on the first story, stripping the building to its bare structure, while

leaving the building above the inundation line almost

undamaged.

A lighthouse at Aonae on Okushiri Island, a one-story

reinforced concrete building with a reinforced concrete tower approximately 8 m high, was severely

damaged by the earthquake (Figure 15). The tower

failed at its base, rotating en masse onto the one-story

building and coming to rest at approximately a 45°

angle. Inspection indicated an apparent bond failure of plain "J" anchorage bars, approximately 24 mm in diameter. Perhaps the building most sustaining the most damage was a three story school in the northern

part of Aonae. Significant diagonal shear cracking

occurred in columns, spandrels, and walls.

LIFELINE EFFECTS

Highways performed well with most damage to pavements resulting from general ground settlement,

and settlement at approaches to bridge abutments.

Few bridges exist on Okushiri Island. In Hokkaido,

bridge performance was generally very good, with

few bridges sustaining more than minor damage.

The most heavily damaged bridge observed was a six span two lane 1970 vintage steel plate girder highway

bridge over the Assabu river, at Assabu-Cho (Figure

16). Lateral spreading was observed on the south

bank of the bridge, which did not appear to affect the

superstructure. Significant cracking was observed

just above the water line at the 1.8 m diameter

reinforced concrete bridge piers founded in the river; significant spalling and broken hoop reinforcement were noted.

Just north of Aonae a recently constructed snow shed

collapsed as the result of rotational failure of

perimeter footings (Figure 17). The most heavily

damaged tunnel in Hokkaido was the Shiroito tunnel north of Setana on Highway 229, as described earlier in the landslide section.

Electric power does not appear to have been

significantly impacted. However, significant number

of distribution poles in the tsunami inundation areas, particularly on Okushiri Island, were swept away by

the wave runups. Water supply to Aonae was not

affected and structural performance of the Aonae

Sewage Treatment Plant did well during the earthquake. However, the sewage treatment plant was shut down following the quake as the result of

contamination by seawater of pumps, motors,

electrical equipment etc, located in a pump room

below grade.

The Okushiri-Cho remote control switches system

performed well during the earthquake (Figure 18).

It is located in a two story reinforced concrete

building and built on firm soils. Atop of the building

is a 40 m free-standing steel truss microwave tower. No damage was reported. Telecom equipment in the building was well-braced in both directions, to floor and ceiling. Backup diesel generators functioned well.

503
SOCIETAL ISSUES

The tsunami warning system worked well in this earthquake. Five minutes after the earthquake, the Japan Meteorological Agency (JMA) issued a warning that a major tsunami had been generated.

JMA transmitted the warning via seven television and

radio stations to local jurisdictions in the affected

areas. In addition, Nippon TV issued a notice 1.5 minutes after the earthquake about a possible

tsunami.

The importance of prior public awareness in areas

potentially vulnerable to a tsunami was clearly

demonstrated by this event. The first tsunami waves

arrived on the east side of Okushiri Island and were

observed during the quake. On the other part of the

island, the first waves arrived approximately 2 to 5 minutes after the shaking. A second and larger waves arrived after the first waves. People reported

that they began to run immediately for higher ground.

In Aonae, people vividly remember the 1983

Nihonkai-Chubu tsunami, and knew that they had to run very quickly to higher ground when they felt the strong shaking. Many of the fatalities were children

who had been sleeping and mothers who tried to wake them up and carry them to safety. Many of the men were out fishing. In Taisei, tsunami warning

sirens were not activated because of power failure,

again demonstrating the importance of backup

emergency power supply.

Adequate shelters were provided immediately after the quake for both displaced residents and for rescue workers and responders. In Okushiri Island,

approximately 90 percent of displacees initially took

advantage of temporary shelter in two schools and

community meeting houses. The ratio dropped to approximately 2 percent by the end of the week. In

Taisei, damage was concentrated in three coastal

districts where approximately 60 percent of displacees

took shelter.

RESTORATION AND

RECONSTRUCTION

The frequent occurrence of earthquakes requires that the Japanese government and its people to be well prepared to cope with their damaging effects.

Emergency headquarters were established at all levels of government within an hour after the quake to coordinate damage assessment and emergency

response including search and rescue efforts. Such

early efforts were continued through the fall of 1993

to restore the services of all lifeline systems. In October 1993, the restoration program had

progressed to the phase of construction and repair of

permanent facilities. On Okushiri Island, multi-story

buildings were being erected of pre-fabricated

components on higher ground for resettlement.

CONCLUSIONS

The study team members were impressed by what

they had learned during the post-event investigations

of this earthquake. In all respect, the event demonstrated the well-prepared nature and resilience of the Japanese community, from preparedness,

emergency response, sheltering, repair, to restoration and reconstruction. All of these elements were

carefully planned and orderly executed. It also demonstrated their insight in engineering design and

construction practices, particularly of their in-depth

understanding of the effects of liquefaction to superstructures.

While it is understood that the earthquake struck a region which is far from heavily populated with only

low rise buildings and less complex lifeline systems,

and the potential impact of an earthquake of such

magnitude to a populated region in Japan is remain to be seen. The general attitude of Japanese toward the impact of earthquakes nevertheless still can set an

excellent example for us to follow in the United

States.

Can major fires and tsunami damages happen here in the States? Actually, major fires have already

happened; but we have been fortunate enough in recent earthquakes that none of them had spread out

to become conflagrations. For instance, in 1989

Loma Prieta earthquake in California, 22 ignitions

occurred in San Francisco alone. Water supply in the densely built Marina District was completely lost, and

a large fire developed could not be extinguished until a fireboat supplied water from the Eay through a

special large diameter hose system developed by the

fire department for just such a purpose. The 1993

504
Cape Mendocino earthquake in Northern California

resulted in a fire that destroyed most of the business

district of the small town of Scotia. The most recent 1994 Northridge earthquake in Los Angeles area

further put us on notice that major fires or

conflagrations are indeed possible. They can

devastate a major metropolitan area like Los Angeles

if less favorable conditions, such dry weather, strong

winds, and inadequate preparation in fire fighting,

prevail during future earthquakes. It is important that these factors be carefully examined in all jurisdictions

to develop comprehensive plans of attack to avoid

conflagrations in future earthquakes.

Fortunately, we have not had to face the devastation

of a tsunami in any of our recent earthquake events. Nevertheless, a subduction similar to that in Japan

runs through Seattle and into Puget Sound. If an

earthquake of magnitude 7.8 struck there, major

tsunami damage could be expected. On Okushiri

Island, the residents ran for high ground as soon as they felt the shaking, expecting the arrival of a tsunami. In contrast, when a tsunami warning was

issued for the northwest coast following the 1964

Alaska earthquake, people drove to the shore to watch the wave come in. Will we now have a

different attitude toward such event thirty years later?

The Hokkaido-Nansei-Oki earthquake provided a number of successful stories in the performance of

buildings and bridges. Joint effort should be

established to bring together experts from both

countries to examine such performance in detail so

that better and improved engineering design and

construction practices in both countries can be

achieved.

ACKNOWLEDGMENT

The author would like to thank all U.S. team

members and members of the Japanese team for the opportunity to work together in their field investigations immediately after the earthquake. U.S. team members are: Les Youd, Richard Barnes, Eddie

Bernard, Richard Bukowski, Frank Gonzales, Edwin

L. Harp, Jane Preuss, Charles Scawthorn, Dennis

Sigrist, Paul Somerville, David Tyree, Peter Yanev,

and Robert Lo.

Particularly thanks will go to Messrs. Iida,

Kawashima, Unjoh, and Nakajima, and colleagues at the Hokkaido Development Bureau, for their technical and logistical support to make the investigation effort a success one. The author also would like to express our appreciation to a number of

organizations who supported this reconnaissance

effort. These organizations are the National Institute of Standards and Technology, Earthquake

Engineering Research Institute (EERI), National Center for Earthquake Engineering Research

(NCEER), National Science Foundation who funded

the postearthquake investigation activities of EERI

and NCEER, National Oceanic and Atmospheric

Administration (NOAA), U.S. Gelogical Survey and

the U.S. Forest Product Laboratory.

505
Figure I Rupture zones of offshore earthquakes in Japan larger than magnitude 7.4 since

1923, with plate boundaries shown on dashed lines. Source: Modified from

Japan Meteorological Agency

506
Figure 2 Map of southwestern Hokkaido and Okushiri Island showing principal localities

referred to in this report

507
R.C.E.P. Hokkaido Univ.

Figure 3 Mainshock and aftershock hypocenters, July 1 7 through 1 8, 1 993. Depth cross

sections O-O', S-N, and B-B' are shown on the right with vertical exaggeration of

a factor of 2. Source: Research Center for Earthquake Prediction, Hokkaido

University

508
Hokkaido-Nansei-Oki, 12 July 93, M=7.8

Closest Distance (km)

Figure ^ Very preliminary uncorrected peak horizontal acceleration plotted against closest

distance to the rupture surface, compared with attenuation relations for

subduction earthquakes used in the United States. Source: Hokkaido

Development Agency, Japan Meteorological Agency, Japan Rail, Ministry of

Construction's Port and Harbor Research Institute

509
510
Hokkaido Nansei Tsunami

Figure 6 Distribution of vertical tsunami runup for Okushiri Island. The legend indicates

data source, and runup values are projected onto offshore scales. A projection

from a data point onto the coastline will identify the geographical position of the

measurement

511
Figure 7 Rock slide near the ferry terminal in Okushiri Port that buried the Yo Yo So

(Hotel). The slide is comprised of Pliocene tuff, tuffaceous sandstone, and

conglomerate and is approximately 800,000m3 in volume.

512
Figure 8 Concrete talus shed collapsed by rock fall along highway 229 north of Setana.

513
Figure 7 Pier 1 at the Port of Hakodate was the most severely damaged port facility we

visited. (A) View southward of quay wall that tilted or shifted seaward by as much

as 3m. (The warehouse in background is further illustrated in Figure 19.) (B)

View northward of paved area that was disrupted by settlements as great as 0.5m

and fissures with separations as great as 1m. Source: T.L Youd

514
Figure \0 Cement storage tank near Pier 16 at the Port of Hakodate. The tank tipped about

2.5 degrees southward due to transient loss of bearing strength. Source: T.L

Youd

'

515
Figure 1 1 Buried storage tank that floated out of the ground at an industrial facility about

6km north of Esashi. The tank was about one third full of oil at the time of the

earthquake. Source: T.L Youd

516
Figure i2, (A) Southward view across campus of Oshamanbe Nakonosawa Elementary

School showing classroom building (center) and gymnasium (right). The playing

fields in foreground were disrupted by ground fissures and ground settlement.

(B) Disrupted pavement in roadway near gymnasium; overlapped pavement

indicates compressional impacts due to ground oscillation. Source: T.L Youd

517
Figure J 2> View of rotated bridge bearing and a girder that penetrated into abutment wall at west end of Kamiiso-shin bridge. This damage was caused by displacement or

tilting of the abutment toward the river. Source: T.L Youd

518
Figure H-f- Views of dike along the north bank of the Assabu River, illustrating the

deformations diagrammed in Figure 6-39. (A) View westward showing

longitudinal fissure along the center line of dike crest. (B) View along

north flank of dike showing fissure with southward (toward dike) facing

scarp, indicating flank had risen relative to dike.

519
Figure IS A Aonae lighthouse--the structure failed because of the pull-out of

inadequate ties (dowels) between the foundation and the superstructure.

The dowels were 24mm (1 inch) in diameter.

Figure 15 S Aonae Lighthouse-detail of reinforcing anchorage failure. 520
Figure I £A A six span, two-lane, 156m long, 1970 vintage steel plate girder bridge at Assabu on Highway 227. Severe shear damage to all visible columns

which are about 2m (6 ft) in diameter. The bridge is located about 80km

(50 miles) from the epicenter. No significant damage was observed to the

superstructure.

521
K ii

Figure lG 6> View of Assabu bridge piers-note cracking near waterline.

Figure >6C Detail of a damaged column of the bridge over the Assabu River. Note

the buckling of the vertical reinforcing steel and the wide spacing of shear

(horizontal) reinforcement. The estimated damage is 25 to 50% of the

value of the bridge.

522
Figure 17 A A collapsed snow shed (tunnel) just north of Aonae. The just completed

structures is on fill in rice paddies. About 1/3 of the three-hinged arch

structure has collapsed. The rest is severely damaged. The

approximately 2m (6 ft) wide quarter circle panels are precast concrete

and are bolted to the footings and to each other at the top.

Figure \J 8> Exposed footing and stem wall of snow shed (west side - undamaged).

523
Figure 17 C Failed east side of snow shed-lateral spreading of the fill and/or

liquefaction caused spreading, rotation, and failure of the footings.

Figure 17 I> Note that, although the arch has collapsed and dropped the precast

planks onto the pavement, the planks area relatively undamaged, testifying to the general quality of construction.

524
Figure f 8 Okushiri-Cho Remote Control Switch-it is a concrete shear wall building.

Atop the building is an approximately 40m (1 30 ft) free-standing steel truss

microwave tower.

525

MANUSCRIPTS

AUTHORED FOR PANEL

MEETING BUT NOT

PRESENTED ORALLY

Technological Development for Automatic Construction

Methods of Caissons

by

Masanori Nakano*, Jiro Fukui** and Shigeru Takagi***

Abstract

In recent years, the construction Industry has been facing such problems as skilled labor shortages and

an aging work force, while at the same time, foundation structures are becoming wider and deeper.

Therefore, it lias now become necessary to Improve the efficiency of field work for smooth and expeditious construction of foundation structures.

Willi this In mind, the Ministry of Construction Is undertaking a comprehensive development project to automate the construction oT caissons. This development project deal with open caissons'", including

the precasting of caisson bodies and development of technologies for automatic excavation, earth removal and press-In settlement, and pneumatic caissons'3

', which entail the development of technologies for automatic excavation and earth removal Inside a working chamber under high pressure

of air.

Key words: foundation, open caissons, pneumatic caissons, precasting, automation, excavation,

earth removal, press-In settlement

1. Introduction

In recent years, demands have arisen In the construction Industry to Improve construction

efficiency and working environments to counteract problems such as skilled labor shortages and an aging work force. With this In mind, the Public Works Research Institute (PWRI) Initialed a five-year project In 1990 to Improve

the efficiency of field works and promote the technological development for automating

several construction methods presently In use on

site. This paper will Introduce the technological

development for automating the construction of open caissons and pneumatic caissons.

2. Technological development for automating the construction of open caissons

2.1. Present state of the construction method

for open caissons

The open caisson method has been widely used for constructing foundations, shafts of tunnel

structures, etc-. However, alternative methods,

such as the pile foundation method, are gradually replacing It. Figure I shows the transition of methods for foundation structures. Caisson foundations accounted for 2G % of the

total In 19GG; however they decreased to G % In I97G. and 4 % In 1985. Open caisson foundations

accounted for 4.3 X of all caisson foundations In 197G (12.3 out of a total of 2,837 foundations

investigated): however, this proportion

decreased to 2.4 % In 1985 (90 out of a total of

.3,097 foundations Investigated). One plausible reason for this may be the failure of the open

caisson method to stay competitive with the pile

foundation method. New techniques, such as I. ho enlargement of diameter, were developed for the pile

foundations, while Integral technologies, such as methods for caisson production and excavation, were not actively developed for the open caisson method. Thus, the open caisson method could not go beyond the confines of convention.

The procedure for constructing open caissons

using the conventional method is as follows: a caisson body is constructed using cast-ln-place

reinforced concrete; the earth inside the caisson

Is removed with a clamshell bucket or a hammer

grab bucket: then the caisson is settled using a press-In device (oil pressure Jacks). However,

its construction efficiency In terms of the number of workers and the term of construction

Is not satisfactory, since several construction

works must be done In situ, such as assembly of reinforcing bars and molds, concrete placement,

and curing. In addition, problems such as reduced precision or failure of caisson settlement occur when the caisson is constructed

deep underground, since the ground under the cutting edge cannot be precisely excavated

during excavation and settlement. As a technique

to solve these problems, the PC well caisson

method'3 ' has already been put to practical use. With this method, the caisson body is constructed

by assembling ring-shaped precast blocks at the point of settlement, to which tension Is applied

by PC steel bars.

*)Chlef, Foundation Engineering Division. Structure & lirldge Department, PWRI, MOC

»»)SenIor Researcher, ditto. •»»)Researcher, ditto.

525
However, this method Is applicable only to caisson foundations with small diameters of 4.0 m or less due to the limitations Imposed by

conveyance of precast; blocks. In addition, the construction Is Imprecise, since the settlement,

using a press-in Jack heavily relies on the experience and observations of workers.

2.2. Objectives of the techno logica 1 development

and out lines of construction methods

To Improve the open caisson method, the PWRI is proceeding with the development of construction

technologies 1 ' 711

. The purpose of this development;

Is to sharply reduce both labor and construction

terms In field works, expand the applicable

domain of open caissons and ensure high

construction precision by Introducing precast

caisson bodies and automating the excavation and

press-in works.

This development comprises the following three

systems. The major advantages of each system

are summarized In Table 1. Any combination of

these systems can be used depending on the conditions at each work site (see Figure 2). (1) Development of a precast body system

This system Is designed to assemble precasted

materials in situ, the scantling of which is smaller than that of the PC well caisson method.

This system enables the construction of a caisson

body of high quality with less labor during a shorter period of time than a caisson body

constructed using cast-ln-p lace concrete.

(2) Development of an automatic excavation and

earth removal system

With this system, the ground under the cutting

edge of caisson body can be accurately

excavated, allowing for deep, highly efficient underground construction. Thus, the range of applications for open caisson construction can

be expanded.

(3) Development of an automatic settlement control system

This system automatically controls the press-In

settlement. and the position of caissons,

processing data measured by various sensors.

With this system, press-In settlement can be conducted with high precision.

2.3. Precast caisson body system

2.3.1. Structure of a caisson body

The structure of a caisson body using precast

blocks Is similar to that of shield segments; a caisson body is constructed by connecting

precast blocks using horizontal and vertical

Joints. The horizontal joints are connected by

applying tension to PC steel bars, as Is the case

with the PC well caisson method, while the vertical joints are connected by wedge

connecting joints. As shown In figure 3, external forces act on the vertical joints. Wedge connecting joints are

resistant against, bending force by wedge metals,

out-plane shearing force by square steel pipe

keys (see Figure 4), and in-plane shearing force by the confining force of upper and lower blocks.

With regard to tiie structure of a wedge metal,

blocks arc connected to each other by inserting an H-shaped wedge metal Into a ('-shaped metal planted In a body In advance; resistance against

bonding is provided by the tensile force of the H-shaped metal (see Plate 1). When the automatic excavation and earth removal system and the automatic settlement control system are used for constructing a caisson, guide

rails and various sensors, such as an earth pressure gauge, a hydraulic pressure gauge and

an inclinometer, are installed on the caisson

body.

2.3.2. Design method of a caisson body

The precast caisson body system differs from the conventional method in that a caisson body is constructed by connecting precast blocks In situ using joints. The performance of the .joint

sections, such as stiffness and strength, affects greatly the deformation and the sectional force

of flie entire caisspn body due to external forces such as the bending moment and horizontal force.

In accordance with the Design Specification"1 for Caissons presently In effect, a caisson must be a monolithic body constructed with cast-in-place

concrete. Therefore, In order for the precast

caisson body system to meet the specification, the stiffness and strength of joint sections must be made equal to those of monolithic materials.

In order to confirm the performance of the various types of joints mentioned above, the Public Works Research Institute conducted a loading tests using models with thickness

identical to that of the side wall of actual-size

caissons. As shown in Figure 5, the test results proved

that the vertical joint possessed stiffness and strength roughly equal to those of a monolithic concrete body that uses no Joints, at the level of allowable stress In the present design method. A loading test was also conducted for horizontal

joints In order to understand the features of deformation and the strength of Joint sections

(see Plate 2). This test elucidated the reduction

of stiffness that occurred locally to horizontal

joint sections and the progress of bending and

shear failure due to the yield of PC steel bars.

In addition to these loading tests, the Influence

of joint sections on deformation and sectional

force was examined by conducting an analysis based on the finite element method. Figure fi shows the model for analysis and an example of

530
the results. In this model, the stiffness against

the tensile force In the joint sections of members

was neglected. However, the test results proved

that no region of tensile stress appeared on the caisson body and that the Influence of the joint

sections on deformation and sectional force was

almost negligible.

2.3.3. Production of precast blocks of a caisson

bodjr Precise work Is required for the production of precast blocks, since any deficiency is likely to aTfoct the performance of joint sections: an aperture or projection on a Joint surface may

cause cracks when the blocks are prestressed by

applying tension to the PC steel bars. In the precast caisson body system, blocks are produced

by the match-cast method in order to eliminate irregularity on joint surfaces. This method Is to

utilize the surface of a block made by casting concrete as the mold of an adjacent block. Specif leal 1}-, the two block surfaces become

concavo-convex, the Irregularity on the surface of the first block being offset by the second

block. Thus, there exists no aperture on the joint surface. This production method Is also used for the Joint surfaces of precast block

girders of bridge superstructures, and Its reliability is high. The problem, however, is that breaks caused by joints exist on the Joint

surfaces. Therefore, the production precision and the assembly method must be kept under close

watch. The construction procedures of a caisson body

are as follows: PC steel bars are set up prior to

the Installation of blocks; an adhesive agent Is applied on the horizontal Joints; blocks are

Installed: tension Is applied to PC steel bars; then, vertical Joints are Installed.

2.4. Automatic excavation and earth removal

sys ten

This system comprises a back hoe-type automatic

underwater excavator, which travels along guide

rails, and a portal crane-type automatic earth

lifter coupled to the excavator that removes

earth. The automatic underwater excavator

excavates the earth Inside a caisson Including

the earth under the cutting edge, traveling

along the guide rails installed on the lower end

inside the caisson body (see Plate 3). Several

excavation patterns to meet various soil conditions and Inclination of the caisson body

arc programmed In the excavator; the excavator

automatically conducts excavation, selecting the most appropriate excavation pattern (see Figure

7). The portal crane-type automatic earth lifter detects Its own location so that it does not touch the arm of the excavator. It travels along

the circumference of the caisson In a link motion

with the circular travel of the excavator, and removes the earth with Its clamshell bucket (see Plate 4). With this system, the above operation Is commenced automatically upon designation of an excavation depth, which Is repeated until the caisson reaches that depth.

2.5. Automatic settlement control system

The automatic settlement control system is an improved version of the conventional press-in open caisson method. This Is to control the settlement and the position of a caisson by

controlling the press-In force based on the data

measured by various sensors Installed on the caisson body. This system consists of measurement and control systems.

By the measurement system, data collection arid processing, display of calculated data and

figures, and issuance of warnings are all done

automatically. The measurement factors are the settlement. Inclination, eccentricity, the reaction force of the cutting edge, skin friction, and earth and hydraulic pressures. Using the data obtained by the measurement

system, the control system calculates the most appropriate press-In force and press-In moment

for controlling the settlement and the position

of the caisson. The press-in force of each Jack

Is adjusted to the most appropriate one In real

time.

2.G. Trial construction test

In order to confirm the applicability of this proposed method, tests using an actual-size

caisson model have been conducted in 1993. in these tests, to confirm their performance and

applicability, the automatic excavation and earth removal system and the automatic settlement

control system In the test are Introduced at the stage where a caisson body is constructed using

precast blocks. For the purpose of examining the applicability of the systems to different types of

soil, the ground conditions are modeled after

sandy, clayey and mudstone layers. Major

confirmation Items In the test are shown In Table

2. The test results concerning the precision of caisson construction, automatic excavation, and

automatic press-In settlement arc discussed

below:

(1) Construction of a caisson body

The construction precision and applicability of a caisson body using wedge connection joints and concrete connection joints were examined. The

precision of the construction of a caisson body

proved that the errors are Invariably less than G mm by the test.

531
(2) Excavation precision

Figure II shows the cross sections of both sandy

and mtidstone layers to Illustrate the excavation

precision; excavation patterns Instructed by the control system and the outcomes of the actual excavation are compared. Although the curves of

the actual excavation differ conspicuously from

the targetted excavation curves around the point

where the excavator turns upward under the cutting edge, the excavation errors were within a

lew centimeters across the board, almost

negligible for practical use.

(3) Settlement precision

Figures 8 and 9 show, respectively, the

inclination angle and the eccentricity in terms of

the precision of the position control. When

going through the sandy and mudstone layers,

which is equivalent to the initial stage of actual caisson settlement, the caisson showed slight and

transient inclination; however It immediately

assumed the proper position automatically. When

each lot had settled, the caisson was built with

its eccentricity and Inclination angle within the standard control values (allowable limit).

3. Technological development for automating the construction of pneumatic caissons

3.1. Present situation of the pneumatic caisson

method

The pneumatic caisson method Is designed to settle a body to a designated depth by Iterating

the procedures of caisson building and

settlement; when the base ground Is excavated,

compressed air Is sent: to the airtight working

chamber Installed underneath a caisson in order

to prevent the groundwater from flowing Into the work yard. The construction reliability of this method is high, since, for example, the bearing

capacity of the bottom of a caisson can be

confirmed directly. Therefore, this method Is widely used for shafts, and foundations of bridges and other underground structures. However, with the present method, construction

work depends heavily on manual labor; skilled workers are sent down to the working chamber to excavate the ground using electric excavators

and exchange a earth bucket to carry out the excavated earth. The working environment worsens as the construction depth Increases. Since the air pressure Inside the working chamber

Increases, the workers are susceptible to "caisson disease," and the allowable terms to work in the working chamber Is decreased for the health control. Therefor, to protect the safety and health of workers, the maximum construction

depth must be limited.

In order to solve these problems, private sector corporations have been conducting research on

labor saving and automation of the pneumatic

caisson method since 1980. Presently, the focus

of the research Is the remote operation of an excavator inside a working chamber. The

technology for the remote operation of the excavator from an operation panel installed either Inside a pressure-resistant: capsule or on the ground is shortly put to practical use; some automatically-controlled excavators are already

in use (see Plate 5). Nevertheless, the earth removal works have not yet been automated, and workers are still sent to the high-pressure

caisson box for exchanging buckets. In addition, a worker (lock attendant) must be assigned for opening and closing the air lock in the upper

part of the material shaft; this worker is also vulnerable to accidents such as falls, since his work post Is elevated, as the depth of caisson

construction Increases. The excavator Is also used for loading the excavated earth in the buckets, during which the excavation work is suspended. Tor improving the work efficiency, the earth loading and removal must be kept separate from the excavation work. Problems still unresolved concerning the construction of pneumatic caissons are as follows:

(1) Earth removal work inside the working

chamber and the lock switching operation should

be automated.

(2) To maintain excavation efficiency during the earth removal work, the earth removal system

must be separated from the excavation system.

3.2. Goal of the development and the outline of the construction method

PWR1 Is developing an automatic conveyance

system for excavated earth, which does not require either workers stationed Inside a caisson

box for earth removal or a lock attendant2

'. The

goal of this development Is to completely automate the works inside a pneumatic caisson, by

combining this system with the automatic excavator being developed In the private sector. Figure 10 shows a schematic diagram of the automatic conveyance system for excavated earth devised and examined by the Institute. The pre-conditions for this system are reasonable

production costs and the feasibility of switching

to the conventional method without suspending

the construction work when problems sucli as breakdown of the machines occur.

With this system, the excavated earth is automatically removed using the following

procedure:

(1) The residual earth excavated by the automatic excavator In the working chamber is collected on

the bottom of caisson center;

(2) The accumulated earth is lifted out of the

532
working chamber by a telescopic clamshell

bucket;

(3) The earth receiving- board Is closed, and the earth Is led to the shoot;

(4) The slide gate Is opened, and the earth is fed

into the bucket; and

(5) Using the automatic-control skater, the bucket is rolled up, the shaft is lifted, and the earth Is dumped into the earth hopper.

Here, operations such as the collection of excavated earth, the switching of the earth

receiving board, the slide gate, and the lock, the

raising and lowering of the bucket, and the dumping of earth Into the earth hopper, are all conducted automatical ly. When system trouble occurs, the construction

continues, being switched to the conventional

method; the bucket is put into the working

chamber which Is being loaded with earth by the excavator.

3.3. Element test

A test using an actual-size model of the earth

removal system was conducted In order to confirm

In detail the performance and durability of each element of this system. In the test, emphasis was placed on the earth flow In the shoot, the durability of the slide gate, and the precision

and durability of various sensors for the automatic control (see Plate 6).

4. Conclusion

As mentioned earlier, PWRI has been developing

technology for automating the construction of open and pneumatic caissons. The results of

tests using actual-size models, all of which had

been completed by last year, are now being used

to prepare design and construction manuals for putting the methods to practical use. Road construction in Japan Is expected to become

Increasingly difficult In the future, since It will be required In mountainous areas and congested

urban areas. Therefore, construction terms and

on-site works must be reduced. Moreover, the

number of deep underground construction works

Is expected to increase. The technological

development for automating construction works

Introduced in this paper will be helpful for saving labor and meeting such social needs.

Notions:

(1) Open caisson method

Method to cause a caisson to settle Into a desired bearing layer by using a clamshell or grab bucket to excavate and remove earth within the caisson interior.

(2) Pneumatic caisson method

Method to Install a working room at a bottom of a caisson to which compressed air is supplied for discharging water and to cause the caisson to settle onto a desired bearing layer by using

human labor or machines to excavate and remove

earth within the caisson Interior.

(3) PC well caisson method

PC well caisson method is same to open caisson

methods for excavation and settlement. The

difference of this method Is using precast, concrete blocks for the main body, and using PC

steel bars to conect the each blocks. The height, of black Is 2. 0-2. 5m and its diameter is l.G-4.0m.

References

1) Ministry of Construction, et al.; Research on

the Technological Development for Automating

the Construction of Bridge Foundations (Open

Caissons), 1992 Joint Research Report, 1993.3 (In Japanese)

2) Ministry of Construction, et al.; Research on

the Technological Development for Automating

the Construction of Bridge Foundations

(Pneumatic Caissons), 1992 Joint Research Report,

1993.3 (In Japanese)

3) Nakano, et al.; Technological Development for Automating the Construction of Open Caissons,

Civil Engineering Journal, 1994.1 (In Japanese)

4) Japan Road Association; Specifications for Highway Bridges, Part IV. Substructures, 1994.2

(In Japanese)

533
TabJe - 1. Advantages of the development

Developments items Safety

improvement

Quality

improvement

Reduction of

the number

of workers

Reduction of construction

terms

Scale-down

of work

yard

Precasting of bodies O O

Development of an automatic excavation

and earth removal system

Development of an automaticcontrol press-in settlement

system

Table - 2. Verifications items in the trial construction test

Purposes of the test Check items in the test

Ceneral

Understanding the overall applicability and work efficiency of the constructionApplicability and work efficiency

Construction control data

Body assembly works

Understanding the applicability and work efficiency con- cerning the assembly of pre- casted materials

Efficiency in the transportation of precasted materials

Change in the stress of the joint sections and the precision

of the body assembly

Volume of consumable materials used, such as mortar and adhesive agents

Safety and Feasibility

Understanding the practicability of the connection between

the precasted materials and cast-in-place sections

Construction efficiency and installation precision of PC

steel bars

Volume of consumable materials used

Excavation and earth

uplift works

Understanding the performance

of machines(the excavator, the earth lifter and the press-in

equipment)

Applicability of disassembly and assembly

Performance of the automatic control during operation

Operational feasibility, safety, and maintenance feasibility

Understanding the excavation efficiency

Excavation efficiency in each layer

Understanding the linkage be- tween the excavator and the earth lifter

Linked operation of the excavator and the earth lifter

Operational feasibility, safety, and maintenance feasibility

Understanding the linkage be- tween the earth lifter and the press- in equipment

Linked operation of the machines and the systems

Safety and maintenance feasibility

Settlement control

works

Understanding the performance

of the press- in position control

Performance of the press-in settlement and the precision of settlement

Operational feasibility, safety, and maintenance feasibility of the press-in equipment and measurement instruments

Behavior of the caisson and the evaluation of the settlement resistance

Understanding the affect on

the surroundings

Deformation of the surrounding ground

534
100

80

£ 60

Others

0.0

40

20

0

Caisson foundations

26.1

0.7 -6.0-

(4.3)

48.8

44.

5

0.8

- 3.9 -

(2.4)

54.0

41.3

i I" ' Jl 1966 1976 1985

* Figures in parentheses indicate the proportion of open caissons Figure - 1. Transition of foundation usage

Automatic earth lifter Command for automatic earth uplift

Data on the state of earth uplift

pommanc^for_au^tomatic excavation

Data on the state of excavation

Ooeration room for excavator and earth lifter

Automatic

excavation . control system

Automatic

earth uplift control system

is .2 j «

5 - 3=

Central control room

—^ Measurement

svstem

Control system

Figure - 2. Overview of the systems

535
Bending moment

Out-plane shearing force

In-plane shearing force

Horizontal joint

Figure - 3. The precast body using

horizontal & vertical joints

Figure - 4. Structure of wedge connection joint

Load(tf)

Load(tf) 250.0 T j 1 j r

Displacement (mm) Displacement (mm)

(a) Bending test results < b > Out-plane shear test results

Figure - 5. Results of the loading tests on a vertical joint

5M>
Model for the finite element ! Distribution of the axial analysis and the distribution stress intensity of the body of the displacement

Stress intensity

Figure - 6. Sample results of the finite element analysis

Excavation patterns Side view of excavation patterns Plan view of excavation patterns

Guide rails Center of the caisson

1/ Conjectured ground

in case of sudden

settlement

I 1.

Soft soi

Medium-hard soi

Cutting edge

(Continuous excavation

around the caisson

in one direction)

(Discontinuous excavation

in one direction)

Super-hard soil

(Continuous excavation in a designated area

in one direction)

Figure - 7. Automatic control excavation patterns

537
(Standard control value) 30 / 9000

(Standard control value) 30 / 9000

Q -10

Horizontal displacement (mm)

(Standard control value) (Standard control value) -100 0 100

0

-10

-15

(T direction (solid line) I80c direction isolid line) direclinn idolted line) Z70c direction I dolled lino

. U Commencement Df settlement (GL - 1.2 m)

..„ ...

-

- —

11.47

Figure - 8. Caisson inclination Figure - 9. Eccentricity on the top

of the caisson

Tower crane

1 Control room

Excavator inside the _ \J working chamber

Figure - 10. Schematic diagram of the automatic earth removal system

538
Plate - 1. Wedge metal Plate - 2. Loading tests on

the horizontal joints

Plate - 5. The automatic excavator Plate - 6. Overview of the element

for Pneumatic caisson test equipment

539

Strong-Motion Records on the Hokkaido-Nansei-Oki Earthquake in 1993

by

Masashi Satoh", Yuji Ono2)

, Toshio Yamauchi 3)

, and Hiroaki Nishi 4>

ABSTRACT

On July 12, 1993, an earthquake occurred

whose epicenter was off the southwest coast

of Hokkaido. Large-scale tidal waves

(tsunami) struck immediately following the

earthquake and fires broke out, killing many

people on Okushiri Island. Road embankments, bridges, river embankments, and harbor facilities particularly those on the Oshima Peninsula were seriously damaged.

The road division of the Hokkaido Development Bureau has installed strong-motion

seismographs at 28 bridges in Hokkaido to observe strong motion. In the July 1993

earthquake, strong-motion records were

obtained from 12 of these seismographs.

Various spectrum analyses and polarization

analyses were conducted on typical strongmotion records of these 12, and the following facts were found: no predominant fre- quency was found, and there were many

strong-motion records with short-periodic

components.

KEYWORDS: The 1993 Hokkaido Nanseioki Earthquake, strong-motion records, polarization analysis

1. Introduction

A large earthquake occurred in the southwest region of Hokkaido at 22:17, June 12,

1993. In this earthquake, strong seismic

motions of the V degree on the seismic in- tensity scale of the JMA (Japan Meteorological Agency) were recorded in Otaru City

and Suttsu and Esashi towns, while tremors were felt in a wide range of areas from

throughout Hokkaido to the Tohoku district. The epicenter was approximately 100 km off

the southwest coast of Hokkaido, at 42° 7' N. lat. and 139° 2' E. long., at a depth of 34

km. The quake measured 7.8 on the Richter

scale of magnitude. This earthquake oc- curred in a region where few earthquakes

had occurred before. It was the first earthquake along the shore of the Sea of Japan

since the Central Sea of Japan Earthquakes

(Richter magnitude 7.7) of 1983, and the

scale of the Nansei-oki Earthquake was the same as of the 1993 Off Kushiro Earthquake.

The Hokkaido Nansei-oki Earthquake was

the nation's 5th largest earthquake since the

Great Kanto Earthquake of 1923 as shown

in Table 1. After the July 1993 earthquake,

many aftershocks occurred, including over 150 tremors, and strong seismic motions of

V degree on the seismic intensity scale were

recorded again on Okushiri Island on August

8.

As of 1994, the Structures Section, the Civil

Engineering Research Institute of the Hokkaido Development Bureau, had installed

strong-motion seismographs at 28 bridges on

national highways from 1966 when a strongmotion seismograph was installed at Chiyoda

Bridge in the Tokachi district of Hokkaido.

In the 1993 Hokkaido Nansei-oki Earthquake,

1) Head, Structures Section, Civil Engineering Research Institute

2) Research Engineer, Structures Section,

Civil Engineering Research Institute

3) Deputy Head, Structures Section, Civil

Engineering Research Institute

4) Research Engineer, Structures Section,

Civil Engineering Research Institute

541
strong-motion records were obtained from

1 2 of these seismographs, and strong-motion

records of over 100 gal were obtained at 5

strong-motion seismic stations. The largest

acceleration was 249 gal, recorded at the

Shichiho Bridge 155 km away from the

epicenter. This paper reports on response

acceleration spectra, running spectra and

polarization analysis performed on these

major records.

2. Outline and Damage of the Earthquake

In the 1993 Hokkaido Nansei-oki Earthquake, which was not as big as the Off

Kushiro Earthquake of January 1993 when

compared on the JMA seismic intensity scale

of 1-7, strong seismic motions of V degree on the seismic intensity scale were recorded

at 3 JMA observation stations. When

measured on the Richter scale, however, it was equivalent to the Great Kanto Earthquake of 1923, the Tokachi Offshore Earthquake of 1968, and the Kushiro Offshore

Earthquake of January 1993. The depth of

its epicenter was 34 km, which was almost

the same as depths of previous earthquakes,

while that of the Off Kushiro Earthquake

was very deep.

A large-scale tsunami warning was given in the area along the Sea of Japan of Hokkaido

and of the Tohoku district immediately after

the occurrence of the earthquake. In fact, a tsunami struck the coastal area of the Sea of

Japan, and - especially on Okushiri Island - the tsunami and subsequent fires caused

great harm to human life.

It also damaged public infrastructure including 711 roads, 14 bridges and 124 railway

facilities. In road traffic networks/ 16 sections of 5 public roads were closed, and it took 3 months to reconstruct some roads

that had suffered collapses and landslides.

The damage to traffic networks was so seri- ous that some towns and villages near the

epicenter remained isolated temporarily from

the outside world. This speaks of the

enormity of the earthquake. The major

damage to bridges included the subsidence

of embankments at the bases of bridges, and

the breakage of bearing support mortar, of

bridge anti-collapse devices, and of T-shaped

columnar bridge piers (which also had been

damaged in the Off Kushiro Earthquake).

Table 2 shows the damage conditions.

3. Observation Stations and Strongmotion Records

Figure 1 shows strong-motion observation

networks of the road division of the Hokkaido Development Bureau. Figure 2 indi- cates that, in this earthquake, records were

obtained from 12 out of 28 observation stations. Circles in the figure show that the

seismic wave records at the observation sta- tions were digitized. Table 3 shows the names of bridges with digitized records, the

ground classifications of the stations under

the specification of road bridges, the types of

strong-motion seismographs installed, and

distances from the epicenter. Records are recorded on stylus paper for SMAC-B2 or on scratch film for SMAC-Q and D except at the Hakucho Bridge. However, since the

frequency characteristics of records differ

depending on machine types, records are

corrected, after digitization, in relation to

main axis, circular arc, and frequency characteristics.

Table 4 shows the maximum acceleration of

the following 3 components with corrected

records at each bridge: the bridge axis direction (Longitude: LG), the right angle direction of the main axis (Transverse: TR) and

the up-down direction (UD). The maximum

acceleration measured in this earthquake was

542
249 gal at the Shichiho Bridge (ground

classification Class-I by the specification of

road bridges), which was 155 km away from

the epicenter. Compared with the Off

Kushiro Earthquake of 1993, no acceleration

of over 300 gal was recorded, probably be- cause bridges were far away from the epi- center.

Strong-motion records at the Kamitorisaki

Bridge and the Shichiho Bridge, which are comparatively near the epicenter, are shown

in Fig. 3-(a),(b). As indicated by these

strong-motion records, the P-wave recorded

at both the Kamitorisaki Bridge and the

Shichiho Bridge (both Class-I ground,

roughly equidistant from the epicenter) was

equal and was 30 seconds. However, comparing the S-waves, that of the former was

50 seconds, of the latter 25, respectively; in other words, the length of principal shock

differed greatly. No common term was

found for original waveform characteristics.

4. Running Spectra and Acceleration Response Spectra

Running spectra were obtained to conduct

waveform analysis, including dynamic characteristics of the ground. The analysis was made by obtaining the power spectrum for 9

seconds waveform in time axis, by moving

step width by 5 seconds. A Lag window of

0.25 Hz was used to smoothen spectrum.

Figure 4-(a) shows the results for the Kamitorisaki Bridge of Class-I and the Hakucho

Bridge of Class-Ill, by the ground classification. The dotted lines in the figure show

characteristic value (TG) of the ground cal- culated from the specification of road

bridges. Concerning frequency characteristics at the Kamitorisaki Bridge, the fre- quency predominated at 1.5 Hz near second

50 of the principal shock, but it shifted to 3 Hz, which is almost the same as TG of the

ground, near second 70. Figure 4-(b) shows

running spectra at the Hakucho Bridge

which stands on Class-Ill ground. The fre- quency characteristics at the Hakucho

Bridge had no typical predominant period,

but it was found that the bigger acceleration

amplitude of principal shock, the more the

low-frequency component tended to be

amplified. This is thought to have been due

to a non-linear tendency of the ground which

became remarkably apparent along with ac- celeration of amplitude.

Figure 5-(a),(b) shows acceleration response

spectra (damping constant of 5%) at a right

angle to the bridge axis at the Kamitorisaki

Bridge and the Shichiho Bridge, while Fig.

6-(a),(b) shows that at the Hakucho Bridge

and the Sapporo IC viaduct. The response

spectrum tended to decrease gradually with

the lengthening of natural period.

5. Polarization Analysis

For back azimuth estimates, a time-domain

non-linear polarization filter analysis originally proposed by Flinn has been used. In

this study, "Rectilinear Motion Director Fil- ter Analysis" was adopted, which is widely

used to emphasize the phase difference in a direction of seismic waves. The characteristics of this analytical method are expressed

by the dual parameters of the rectilinearity of

the elliptic, particle orbital motion in a three

dimensional space and in the direction of

polarization.

A data set was built from the analysis of

three-component recordings from the seis- mic station at the Kamitorisaki Bridge. This

technique, however, was used for calculating an accelerogram in a given direction.

Figure 7 shows the results of the analysis. In the figure, sigma indicates power in the ma543
jor axis direction of time width at a given

time, theta indicates azimuth from due north

in the direction of the major axis, and phi

indicates angles between major axis and vertical axis. Figure 8 shows component waves

calculated from polarization analysis. They

are, from top to bottom, the up-down direction component wave, the east-west direction component wave, the south-north direction component wave, the horizontal

component wave in the maximum major axis

direction at the peak power, the horizontal

component wave at a right angle to the

maximum major axis at the peak power, the

P-wave, and the SV-wave. Consequently,

horizontal seismic motion dominated at the

Kamitorisaki Bridge. As indicated in Table

5, the tendency of horizontal seismic motions was also obtained from the other re- cords.

6. Conclusion

The characteristics of the 1993 Hokkaido

Nansei-oki Earthquake are:

1) On the Richter scale, it was the 5th largest

earthquake in Japan since 1923.

2) On the seismic intensity scale of the Japan

Meteorological Agency, the earthquake was

smaller than the 1993 Off Kushiro Earthquake, but it caused serious harm to human

life.

3) Even though ground classifications and

the distances from the epicenter were the

same, the lengths of principal shock differed.

4) According to various spectrum analyses, no predominant frequency was recognized.

5) Except at the Sapporo IC viaduct, the

seismic wave had waveform characteristics

containing high-frequency component waves.

6) According to the polarization analysis of

seismic motion, horizontal motion dominated.

References

1) "Control System of Strong-motion Records"; Civil Engineering Institute Monthly

Reports No 100

2) Flinn, E.A.: Signal Analysis using Rectilinearity and Direction of Particle Motion

3) "Outline of Strong-motion Records on the

Off Kushiro Earthquake of 1993" 25th joint

UJNR

544
table-1 Details of Recent Major Earthquake Date Name Mj Focal Depth

September 01, 1923 Great Kanto Earthquake 7.9 . December 21, 1946 iNankai Earthquake 8.0 20km

March 04, 1952 Off Tokachi Earthquake 8.2 0 August 12, 1961 Off Kushiro Earthquake 7.2 80

June 16, 1964 Niigata Earthquake 7.5 40 May 16, 1968 Off Tokachi Earthquake 7.9 0 June 17, 1973 Off Nemuro Peninsula Earthquake 7.4 40

June 12, 1978 Off Miyagi Peninsula Earthquake 7.4 40 March 21, 1982 Off Urakawa Earthquake 7.1 40

May 26, 1983 Off Akita Prefecture Earthquake 7.7 14 January 15, 1993 Off Kushiro Earthquake 7.8 107

July 12. 1993 SW Off Hokkaido Earthquake 7.8 34

table-2 Outline of the 1993 Hokkaido nansei-oki earthquake Date July 12, 1993 Time 22:17 Magnitude on the JMA scale 7.8 Hypocenter Long.l390 12'E, Lat.42°47'E Depth of focus 34km

Killed 202persons

Wounded 305persons

Missing 29persons

Damage to buildings (half to full) 805 houses Damage to buildings (less than half) 2,193 houses Damage to public utilities 445 houses Damage to roads 711 Damage to bridges 14 Damage to railroads 124 Damage to harbor facilities 11 Damage to fishingports 58 Damage to ships 1,748

Damage to river facilities 326

Damage to agriculture facilities 2,953

Damage to commerce and industry facilities 2,264

Damage to life lines 38

landslips 14 As of September 15, 1993

545
tablc-3 Ground and Other details of Seismic Stations Bridge Name Ground Classfication Type Distance

Isoya 3rd class SMAC-D 96

Kamitorisaki 1st class SMAC-Q 140 Hakucho 3rd class AJE-306 152

Shichiho 1 st class SMAC-B2 155

Ishikari-Kako 3rd class SMAC-B2 182 Sapporo IC 3rd class SMAC-Q 184

Shin-Ishikari 3rd class SMAC-B2 193

tablc-4 Maximum Accelerations

Bridge Name Latitude Longitude

Acceleration (gal) LG TR UD

Isoya 42.87° 140.37° 118 148 53

Kamitorisaki 42.10° 140.57° 181 231 77 Hakucho 42.35° 140.95° 151 170 81 Shichiho 41.59° 140.31° 235 249 97

Ishikari-Kako 43.22° 141.35° 31 36 12 Sapporo IC 43.07° 141.43° 41 57 27

Shin-Ishikari 43,14° 141.52° 107 98 28 LG : Longitude Direction TR : Transversal Direction UD : Up and Down

table-5 Direction of Major Axis & Angle between this Axis and Epicenter

Bridge Name

Incident Angle

Kamitorisaki 84° 50° 106° 95° Hakucho 82° -23° 48° 31°

Shichiho 90° -53° 22° 13°

Ishikari-Kako 90°

Shin-Ishikari 82° 46° 148° 91°

546
kako

20 Hiroo

21. Kamitorisaki

8.lshikari-kako 22. Sorachi

9.Shari 23.Shimamatsuzawa

24. Sapporo IC 25. Hakucho

26. Sapporo

27 Sh i n hama a t su ma

28. Tokachi-kako

29. Tokachi

139 140 141 142 143 144 147 148

Figure -1 Location of seismic stations of the Stractures Section,

547
48

Figure - 2 Epicenter and location of seismic

stations investigated in this study

548
300

LG 0

-300

300

TR 0

-300

300

UD 0

-300

(gal) Kamitorisaki GL 1993.07.12 22117

Max = I 8 I .Oga

Max= 230. 9ga

Max = -76.7 ga

0

-i 1 1 r ~

i r 50

Time (sec)

Figure-3(a) Storong motion record at the

kamitorisaki Bridge. (Grand)

100

(gal) Shichiho Bridge

300r GL 1993.07.12 22: 17

Max =-235.3gal

Max=-248.5gal

Max= -96.8gal

i r t r -i r 0 50

Time ( sec)

Figure-3(b) Storong motion record at the

shichiho Bridge. (Grand)

100

549
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553

Development of Integrated Numerical Research System

for Prevention and Estimation of Coastal Disaster

by

Tomotsuka Takayama 1 ', Chiaki Goto2)

, Hidenori Shibaki 2)

, and Toshio Aono2)

ABSTRACT

This paper presented an outline of

the design and applications of the

Integrated Numerical Research System for

Prevention and Estimation of Coastal

disasTers (INSPECT)system. This system

was developed to efficiently perform

studies for preventing disasters on Japanese coasts due to wave, tsunami, and

storm surge.

INTRODUCTION

Japan is located in an area with a severe natural environment. Typhoons and

earthquakes hit frequently, and strong,

northwest seasonal winds blow in the

winter. For this reason, the coastal area

shave often been heavily damaged by

natural sea disasters. Typical of such

disasters were the Ise-Wan Typhoon storm

surge disaster in 1959 and the Chilean

Tsunami disaster in 1960. By experiment of these two disasters,

research on the prediction andprevention

of disasters has attracted considerable

attention, and the construction of

facilities to prevent disasters has been

active. In 1994, the Hokkaido-Nansei

Earthquake Tsunami hit Okushiri Island

and caused severe damage. At that time,

the island had disaster prevention

facilities that were designed using

external forces based on the NihonkaiChubu Earthquake Tsunami, which had

occurred 10 years before. Nevertheless, a tsunami exceeding this level hit the

island. This disaster provided the

precious lesson that it is almost

impossible to build facilities that can fully prevent damage from natural

disasters, and that the development of

disaster prevention measures based on previous disaster conditions is limited. At the same time, we strongly

believed in the necessity to develop

systematic analytical techniques for

preventing coastal disasters that can respond quickly to unexpected events and

that can be applied to the wide range of

Japanese coastal conditions.

To analyze coastal disasters, we must first study the phenomena causing

the disasters. The use of numerical

simulation experiments to scientifically

reproduce these phenomena and predict the

irreoccurrence in the future is believed

to be well suited for the study of

disaster phenomena. The development and

wide-spread use of computers, in addition

to the dramatic expansion of their

capabilities, have made simulation

practical. However, the analysis of

phenomena using numerical simulation

requires advanced, special techniques. To solve this problem, we developed the

Integrated Numerical Research System for

1) Director of Hydraulic Engineering

Division, Port and Harbour Research

Institute, Ministry of Transport,

Yokosuka City, Japan

2) Former Member of Ocean Energy

Utilization Laboratory, Hydraulic

Engineering Division, Port and Harbour

Research Institute, Ministry of

Transport

555
Prevention and Estimation of Coastal

disasTers (INSPECT) system, which

effectively combines past studies and

numerical analysis for disaster

prevention and systematizes the analysis

process.

The INSPECT system is designed to be

used for the Japanese coasts and is

intended to be effective for processing

follow-up calculations (hindcasting) of

previous disasters, as well as for

predicting future disasters. Compared to conventional analysis methods that

require high-level knowledge, this system

semi-automatically performs a series of

processing steps, from the preparation of

data to numerical simulation and the

evaluation of results.

DESIGN OF THE INSPECT SYSTEM

1. Equipment configuration

The INSPECT system is an expert

system for using numerical analysis to prevent disasters. This system has been

developed to perform numerical

simulations and a series of analyses for

wave, tsunami, and storm surge extreme

sea conditions through simple operations.

Therefore, the system was designed to quickly output the same level of results

as that of specialists, without requiring

difficult analysis or high-level

knowledge. The analysis is basically

performed in an interactive mode, and

consideration was given to the ability to perform processing using only simple data

input.

The system is composed of

engineering workstations (EWS), personal

computers (PC), and peripheral equipment,

as shown in Fig. 1. The EWS performs

numerical simulations of wave, storm

surge, and tsunami. The PC prepares the

input data necessary for calculation,

analyzes the observed or estimated values,

and outputs the calculation results on a screen or paper. The EWS and the PC are connected on a network that transmits

input and output information between them.

Peripheral equipment for the system

include externalhard disks for the EWS

and the PC, and a laser printer, optical

disk unit, and digitizer connected to the

PC. The (1GB) hard disk for the EWS

temporarily stores estimation results,

while the (300HB) hard disk for the PC

stores the observed value database. The

laser printer provides paper output, the

optical disk unit stores large-volumedata,

such as computed values, weather charts,

and the water depth charts database, and

the digitizer converts analog information

such as weather charts and water depth

charts into digital data. The system also

has video editing and playback equipment

and a special -purpose EWS to perform

dynamic picture processing of computation

results.

2. System configuration and analysis

The wave climate statistical

analysis system has two subsystems: wave statistical analysis, which can statistically process observed wave data

at all places in Japan, and probable wave

analysis, which determines the design wave (50-year probable wave) to be used

for designing harbors.

The wave estimation system has four

subsystems: wave hindcasting on Japanese

coasts from a plane output type spectrum

method model, wave hindcasting in inner

bays (short fetch sea areas) using a plane output type parameter method model, wave prediction during a typhoon by a single point output type spectrum method

model, and wave prediction on Japanese

coasts by a physical factor multiple

regression model. An output system for

estimation results is also provided.

The tsunami calculation system has a

556
near-field tsunami calculation subsystem,

which is used for tsunamis originating in

the nearby sea area of Japan, a far-field

tsunami calculation subsystem, which is

used for tsunamis propagated across the ocean from a remote source, and an output

system for the results.

The storm surge calculation system

has an inner bay sea wind calculation

subsystem, which uses a hybrid wind

hindcasting model that combines an empirical formula with the MASCON model,

a storm surge calculation subsystem,

which uses a multi-layer model, and the

results output system.

The support system is used for the

preparation of input data for wave, tsunami, and storm surge calculations.

The weather chart processing system

digitizes the analog atmospheric pressure

information plotted on a weather chart

and generates a digitized database. This

atmospheric pressure information is used

for estimating wind, which is an external

force for developing waves and storm

surge. The water depth chart processing

system digitizes the analog information

on commercially available sea bottom

topographical charts, marine charts,

ground topographical charts, sea bottom

sounding charts, and similar sources into

a database. This digital, water depth and

elevation information is used as basic

data for preparing the calculation grid

and the grid point water depth data that are used for calculating wave, tsunami,

and storm surge characteristics.

In shallow water deformation

calculations, wave deformation is

calculated for shallow sea waves of

various directions and periods. The

resulting refraction and shallow water

coefficient table (collectively called

the coast coefficient) for a wave of a given direction and period at a given

point is entered into a database. By

using the coast coefficient, the time

series information for offshore waves

(which is calculated by wave hindcasting)

is converted into shallow water

information. The dynamic picture

processing system processes the

computations for wave, tsunami, and storm

surge into dynamic pictures. These

pictures make dynamic evaluation of the

phenomenon possible.

Six databases are available for

supporting these subsystems: the observed

value database for observed waves, tsunami mark height, storm surge

deviations, etc. , the weather chart

database, the water depth database, the

coast coefficient database, the input

value database for numerical simulation

such as typhoons, fault parameters,

calculation grid, grid point water depth,

and the estimated value database for wave, tsunami, and storm surge. The use of

these databases enables processing such

as the preparation of input data for

numerical simulation to be done rapidly

and provides comparisons of observed and

estimated values.

ANALYSIS THEORY OF THE SYSTEM

1. Wave analysis

(1) Wind estimation

The sea wind, which provides the

external force for developing waves, is

estimated using atmospheric pressure

information. The first step in the sea wind estimation is to estimate the wind

in a free atmosphereusing a hybrid model

that combines a gradient wind model with

a typhoon model. The next step is to convert the results into a sea wind using

a boundary layer model n .

(2) Spectrum method wave prediction

One method for handling irregular

557
waves is to assume that the wave is

composed of component waves, each with a different height, period, and direction,

and to represent the waves by their wave energy spectrum. Wave estimation by the

spectrum method is based on the energy

equilibrium equation and follows the time

series change of direction spectrum,

which represents frequency and wave

direction. The MR I method of Isozaki and

Uji 2) is used in this system. This model

has been used in many applications, such

as for wave predictions performed by the

Meteorological Agencyand for calculations

of the design wave for harbor facilities.

(3) Parameter method wave estimation

The parameter method 3) traces the

change in non-dimensional wave energy and

is used for wave hindcasting in short

fetch sea areas such as inside of a bay.

The fundamental wave development equation

is formulated by the relationship between

non-dimensional fetch and non-dimensional

energy and by the 3/2-power relationship

between non-dimensional wave height and - non-dimensional-wave period.

(3) Wave deformation calculation

In shallow sea areas, effects become

important that otherwise need not be

considered in the offshore wave

calculation, such as shallow water

effects, refraction, diffraction,

reflection, and wave breaking. In this

system, evaluation is performed by wave deformation calculations using the energy

equilibrium equation only for the

propagation term.

(4) Wave prediction

In addition to design wave analysis

using a plane output type wave

hindcasting model, the prediction of

coastal waves is also important for

preventing disasters. Offshore wave prediction is performed using a physical

factor multiple regression model 4)

. In this model, prediction is performed using a linear algebraicequation for the

development, propagation, and decay of

waves, while the wind-driven wave energy

and swell energy are used as descriptive

variables. Two models are available:

model 1, which uses only the estimated

wave energy, and model 2, which uses both

the observed and estimated wave energies.

When high wave prediction is

performed only where a typhoon is going

to attack, a calculation method is used

that traces the wave spectrum along the wave direction lines. The spectrum

method is used for the wave development

and propagation equation, and the output

is for a single prediction point.

Therefore, this method is named the

single point output type spectrum method 5)

.

2. Tsunami analysis

(1) Numerical calculation of near-field

tsunami

When the initial wave is formed, it

propagates as a surface wave by the water

pressure differential between that point

and the surroundings. The tsunami is

expressed by the long-wave theory

equation since a long-wave approximation

is possible. Numerical simulation

quantitatively traces the water level and

flow velocity of the tsunami, which is

deformed in the process of propagation by

the effects of shallow water,

concentration, and resonance that are

caused by the changing water depth and

coastal topography. In this system, one

of two fundamental equations is used

(linear or nonlinear long wave),

depending on the minimum water depth of

the subject area.

558
(2) Numerical calculation of far-field

tsunami

Spherical coordinates must be used

for assessing phenomena propagating over distances greater than 10.000km, such as with the Chilean Tsunami. The scale is so large that the wave number dispersion

property cannot be neglected. Therefore,

linear dispersed wave theory considering

the Coriolis force is used as a governing

equation. In far-field tsunami

calculations, as in near-field tsunami

calculations, the initial waveform is

calculated from fault parameters.

3. Storm surge analysis

(1) Inner bay sea wind prediction

The sea wind in an inner bay

surrounded by land is different from that

of an offshore sea wind because it is

subject to the effects of land topography.

This topographical effect greatly

influences the accuracy ofinner bay wave

and storm surge estimations. In this

system, a hybrid wind model is used that

combines an empirical formula with the

MASCON model.

(2) Multi-layer storm surge model

Among the storm surge phenomena, the

rise in waterlevel due to drift is caused

by concentration of the drift current

driven by the strong typhoon wind in the

coastal area. The drift current has a vertically distributed flow velocity. The

flow velocity is at its maximum at the sea surface and decreases with water

depth. Therefore, if the drift current is

approximated by a single-layer model, the

vertically distributed flow velocity is

incorrectly replaced by an average

velocity that is uniform throughout the

water depth. The flow rate can then be

approximated, although the upper-layer

flow velocity is underestimated, and the

lower-layer flow velocity is

overestimated. In the storm surge

calculation, therefore, a multi-layer

model is used to enhance the accuracy of

the drift current estimate.

APPLICATION TO THE ANALYSIS

1. Wave climate analysis

(1) Analysis of design wave

The major reason for the development

of design wave analysis is to identify

the causes of wave disasters and to determine the probable magnitude of waves.

In calculating design offshore waves, the

extraction of weather turbulence,

preparation of weather chart data, wave estimation in weather turbulence,

calculation of the coast coefficient,

verification of the estimated value,

statistical analysis of extreme values,

assumption of model weather turbulence, wave estimationin model weather

turbulence, and study of the design

offshore wave are performed in that order.

The related analysis systems are the wave condition statistical analysis system and

the wave estimation system. Subsystems

for weather chart processing and shallow

water wave calculations are used as support systems.

To extract weather turbulence, weath

er conditions that caused high waves at t he point of interest over a long period

of time are determined. During processing,

a list of high waves by year, the maximum

significant wave distribution during the

period, and other parameters are generated by the wave statistical

analysis subsystem. Such information is

used as the basis for extracting the

weather conditions corresponding to the

559
yearly maximum and extreme waves. Wave prediction is performed after

the weather data is prepared.

Calculations are made on the time series

changes of the significant wave height,

significant wave period, and wave direction at the point of interest when

the high wave occurs. In the wave hindcasting system, when the area and

period of time are specified, the EWS

automatically performs gridpoint

interpolation of the atmospheric pressure,

wind estimation, and wave estimation in

that order. The gridpoint pressure is

calculated using the spline plane

interpolation method converted into

latitude-longitude coordinates. As described in the theory of wave analysis,

wind estimation is performed by using the

gradient wind/typhoon hybrid model and

the atmosphere boundary layer model. In wave estimation, either the plane output

spectrum method or the parameter method

isselected depending on the sea area. For wave hindcasting using the plane

output spectrum method, the latitudelongitude coordinates are used. For main

harbors on Japanese coasts, the wide area

where wind-driven waves develop is

approximated by a 1/2 degre grid. As it

approaches the harbor, the area is

approximated by a 1/8 degree grid and

then by 1/24 degree grids to accurately

represent the shielding effect of the

land. The hindcastings are summarized as alist of maximum significant wave height

and significant wave period by direction

at all grid points for each turbulence.

This list is stored as the wave estimated

value database. The design wave at the

point concerned is calculated by using

this database and the probable wave analysis system.

The probable wave analysis system

performs the various processing tasks

that are required for calculating the

probable offshore wave. Such calculations

include the extraction of estimated data

at the point of interest, correlation

analyses of the estimated and observed

values, and statistical analysesof

extreme values. The time series change of

the maximum significant wave and the

extreme wave by direction for each year are obtained from the estimated wave

value database over a long period of time, as shown in Fig. 2. This is used as basic

data for calculating the design wave at

the point concerned. Because of the

calculation errors in wave estimation,

the correlation between the estimated and

observed values is analyzed to improve

the empirical formula and produce

estimations closer to the observed value.

To perform this correlation automatically,

observed values for the same weather

turbulence as the one being estimated are searched and extracted from the observed wave database.

However, when the observed value is

in shallow water, a table of coast

coefficients for the observation point

should be prepared by wave direction and wave period using the wave deformation

calculation subsystem. The estimated

value must be converted into a shallow

water value before comparison to the

observed value. To perform the wave deformation calculation, the calculation

grid and the grid point water depth are needed as input data. This data is

automatically prepared for the point

concerned by using the coastline and

water depth information that is stored in

the water depth database.

Statistical analysis of the extreme

wave value is performed using the

estimated value corrected by the observed

value. Fig. 3 shows the distribution of wave height of a 50-year probable wave calculated for each direction.

(2) Coastal wave prediction 41

560
The object variables for wave prediction using thephysical factor

multiple regression model are the

estimated wave energy at the time to be

predicted, the estimated and observed

wave energy at the present time (when the

prediction is performed). Therefore,

estimates of the wind at the present and

at the time tobe predicted are required.

The wind at the present can be calculated

from the wind estimation model by

converting the flash weather chart into

digital information using the weather

chart processing subsystem. The predicted

wind can be calculated by digitizing the

predicted weather chart. Figure 4 shows

an example of a 12-hour wave prediction

at Hitachi Naka Harborin February, 1983,

obtained using model 2. Wave prediction is shown in two components, wind wave and swell. Shown

from the top to the bottom in the figure

are the predicted wave direction of the

composite wave for the wind wave and

swell, the time series change between the

predicted and observed significant wave

height, the time series change between

the predicted and observed significant

wave period, the component wave direction

for the wind wave and swell, and the time

series change in equivalent significant

wave height and equivalent significant

wave period for the wind wave and swell.

The prediction results are satisfactory

for both low- and high-wave periods.

During wave development, the direction of

the composite wave coincides with the

direction of the wind wave, while during

wave decay, the composite wave direction

agrees with the swell direction.

Regardng the wind wave and swell

components of significant wave height,

the wind wave component was predominant

on February 17 when the maximum height

wave developed, while the swell component

was predominant from the 20th to the 21st,

when the wave was decaying. During wave

development, the period of the wind wave componentis almost equal to that of the

composite wave.

Thus, the wave prediction model can predict not only the significant wave

height, but also the significant wave

period and its direction. It also can separate the wind wave component from the

swell component. Moreover, the predicted

offshore wave can beconverted into

shallow water wave information by

providing the coast coefficient database

at multiple points (e.g., at the front of

a breakwater) in the shallow water area.

The single point output type wave prediction modelis intended for wave prediction when a typhoon hits. Fig. 5

5) shows the wave predictions for Typhoon No. 8213 off Kochi, as output on the PC

screen. The time series change of the

typhoon path, the typhoon scale, and the

observed wave state are displayed on the

screen. The estimated and observed values

of significant wave height, period, and

direction are also displayed during the

progress of calculation. Prediction of

the maximum wave for the typhoon is

performed on the PC by simple input of

the typhoon information, including the

predicted path and scale, using a mouse

or keyboard. The time required for

prediction of the maximum wave is about

one minute for a five-day wave prediction.

3. Tsunami and storm surge analysis

(1) Analysis of tsunami and storm surge

Numerical calculations of tsunami

and storm surge are often used to investigate the effect of disaster

prevention structures. First,

calculations for reproducing previous

great tsunami and storm surge events are performed to determine the tsunami wave source and the storm surge weather

561
turbulence. Predictions are then

calculated for the planned location.

Tsunami waves and storm surge are numerically computed by sequentially

combining zones with different grid

intervals from the offshore sea area, where the wave or weather turbulence

source is located, to the coastal or harbor area concerned. This is to keep

the discrete and cancellation errors

small. Normally, a grid interval of 1/20

to 1/30 or finer of the tsunami or storm

surge wave length is used.

A feature of the system is that the

calculation program and input data can be

prepared automatically simply by editing

control data for the specified mode.

Control data is edited interactively. One

piece of control data connects the

calculation grid and grid point water

depth data for tsunami/storm surge

calculations with input data such as the

calculation range, grid interval, time in

terval, fault parameter and typhoon

constant, and completes the calculation

preparations. The calculation grid and

the grid point water depth can be

prepared using any calculation rangeand

grid interval for any coastal area in

Japan, whether off-shore or on-shore, by

using the water depthchart processing

system and the water depth chart database.

When detailed topographical approximation

of a harbor and coast isquired, the plan

view of the planned structure and the

harbor bottom sounding chart are entered

into a database, and the calculation is

performed in the same manner. In the

tsunami/storm surge calculation system,

the Japanese coast is separated into

distinctive sea areas, and each sea area

is independently systematized. The

database for the harbor is added to the

corresponding sea area system if

necessary.

Input data is provided for previous

tsunamis from the fault parameter

database by selecting the tsunami number.

Likewise, input data for a previous storm

surge is provided from the typhoon

database by selecting the typhoon number.

The prediction results are output on

the screen or paper. Information about

the maximum tsunami /storm surge water

level and other parameters for each grid

point is stored in the tsunami /storm

surge estimated value database. The

estimated value is correlated with the

tsunami mark height data or the storm

surge deviation data to verify the

calculation accuracy.

(2) Sanriku coast tsunami analysis 60

The Sanriku coast is divided into 33 zones for calculation. Each calculation

zone has six grid interval stages from A to F. The deep sea area (zone A) has a grid interval of 5,400m and ranges from

off of the Nemuro Peninsula in Hokkaido

to the coastal area in Miyagi Prefecture.

This covers the sea area where near-field

tsunamis develop that hit Sanriku. The

coastal area is composed of zones with

grid intervals from 1,800m (two zones of

Bl and B2) to 50m (four zonesof Fl to F4).

Four harbors are included (Ohfunato Bay,

Kamaishi Bay, Kuji Bay, and Hachinohe Bay).

Regarding previous Sanriku tsunamis,

the Meiji Sanriku Tsunami. Showa Sanriku

Tsunami, Tokachi-Oki Earthquake Tsunami,

and Chilean Earthquake Tsunami can be

specified from the fault parameter

database. For theChilean Earthquake

Tsunami, the calculated results of

propagation from the Pacific Ocean onto

the Sanriku coast is extracted from the

tsunami estimation databaseand is used.

The topographies of the harbors at the

time of the previous tsunamis (e.g., the

Meiji, Showa Sanriku and Chilean

Earthquake tsunamis) were prepared from

old topographical charts and various

tsunami records at each location and were

562
put into the database.

Figure 6 is a propagation chart for

the Chilean Earthquake Tsunami obtained

by numerical calculation. The numerals

represent the elapsed time from the start

of the tsunami. The Pacific Ocean

propagation calculation for the Chilean

Earthquake Tsunami covers the area from

60 degre South to 60 degree North and

from 120 degree East to 70 degree West

(i.e., the whole Pacific Ocean surrounded

by Far East Asia, Australia, North

America, and South America), as shown in

the figure. The calculation grid interval

is 10 minute in both latitude and

longitude (about 18.5km distance at the

equator), and the total number of

calculation points is about 730, 000. The upper side in Fig. 7 shows the

output of the maximum water level

distribution in Ohfunato Bay on the

Sanriku coast for the Chilean Tsunami.

This result wasobtained by connecting the ocean propagation and near-field tsunami

calculations, the latter using a finer

grid in the sea area near the Sanriku

coast. The figure indicates the contour

lines for the maximum water leveland the

tsunami water level at points where the

tsunamimarks are apparent. The numerals

in the figure are based on T. P. (m), and

the numerals in the parentheses indicate

tsunami height marks. The shaded area

denotes the flooded land area. The water

level for the Chilean Earthquake Tsunami

in Ohfunato Bay was about 1.8m at the

mouth of the bay and about 4. 5m at the

innermost point of the bay, gradually

increasing from the bay mouth to its

inner most point. This indicates resonance with the natural vibration of

the bay with the mouth being a node and

the inner most point being a loop.

Flooding from the tsunami reached the

urban area, suggesting that the damage

caused by the tsunami was heavy.

The lower side in Fig. 7 shows the

results of numerical calculations for the

imaginary maximum water level

distribution from the Chilean Earthquake

Tsunami, assuming a bay mouth breakwater

is installed and using the same fault

parameters as those in the verification

calculation. The calculated water level

from the tsunami is about 1.2m at the

mouth of the bay and 1.9m at the inner

most point of the bay. This indicates

that the water level is halved by the

addition of a bay mouth breakwater.

Flooding is limited to narrow area, where

a tidal embankment is not installed and

the ground elevation is low.

(3) Ise-Wan storm surge analysis

In the storm surge calculation

system, wind estimated by the typhoon

model is added to that from the tsunami

calculation. Storm surge is very

different from a tsunami in that the

external storm surge forces are from

atmospheric pressure and the typhoon wind,

which changes constantly, while the

external force in atsunami is the initial

water level determined from the fault

parameters. In general, the water level

rises remarkably when a typhoon

approaches. However, even when a typhoon

is distant, the water level may rise in a bay if the free waves developed in the

typhoon zone propagate through the ocean

and enter the bay. It is also assumed

that the drift current developed in the ocean is blown to the coast, affecting

the rise in water level not only at the

mouth of a bay, but also in the entire

bay. Therefore, the region used for storm

surge calculations should be a wide ocean

region including the continental shelf

and the sea area where the typhoon passes.

Along the coast, grid intervals of

about 100m or narrower are necessary to evaluate the effect of delicate

topography such as harbor structures on

563
storm surge. In an inner bay with shallow

water, the effect of astronomical tide

cannot be neglected, particularly when

the amount of land flooding due to a storm surge is calculated, so a highly

accurate calculation of astronomical tide

is necessary. However, using current tech

nology, it is difficult to simultaneously

perform atide calculation for the ocean

and an accurate tide calculation for an inner bay, because the information on boundary conditions is insufficient. For

this reason, the storm surge calculation

is performed by separating the area into ocean and inner bay regions. Fig. 8 shows the distribution of maximum storm

surge deviations in the ocean region

during the Ise-Wan Typhoon. The ocean

region is subdivided into six zones, from

the ocean to Ise Bay to perform the

calculation efficiently. In each zone, the topography is approximated by grid

intervals that decrease gradually at a rate of 1/2 or 1/3 as the bay is

approached.

Fig. 9 shows the flow velocity

distribution that was estimated by using

a two- layer model for Ise and Mikawa bays

during the Ise-Wan Typhoon. Water depth

in the open sea is several thousand

meters, and the sea area in front of the

mouth of Ise Bay has a depth of several

hundred meters or more with a steep sea bottom slope. Under such topographical

conditions, the drift current generated

by the storm surge develops mainly in the

upper layer. If a single layer model is

used for the open sea, the transfer of sea water due to the drift current is

approximated as being uniform throughout

the layer. This results in part of the

water layer being reflected back to the

open sea at the steepsea bottom slope, so that it does not enter the bay. For this

reason, multi-layer modeling is used.

CONCLUSIONS

This paper presented an outline of

the design and typical on-site

applications of the Integrated Numerical

Research System for Prevention and

Estimationof Coastal disasTers (INSPECT)

system. This system was developed to efficiently perform studies for

preventing disasters on Japanese coasts

due to wave, tsunami, and storm surge. The main conclusions follow.

(1) The system was developed to systematize the numerical analysis of

wave, tsunami, and storm surge and to efficiently perform a series of

processing steps, from the summarization

of observed information and the

preparation of input data for numerical

simulation, through numerical calculation,

to the analysis and evaluation of results,

by a simple operation. The design premise

is to permit an engineer to quickly

output the results by a simple operation

without difficult analyses or high-level

special knowledge.

(2) The equipment used in this system

consists of engineering workstations,

personal computers, and theirperipheral

devices, which provide a general -purpose

configuration. Processing is performed in

teractively, and the results can be

output on screen or paper.

(3) The system is composed of five

analysis systems (wave statistical

analysis, wave estimation, tsunami

calculation, storm surge calculation, and

calculation support), and six support

databases (observed value, weather chart,

water depth chart, coast coefficient,

numerical simulation input value, and

estimated value). Each analysis system

has multiple subsystems. Analyses are performed by using combinations of the

subsystems and databases.

564
(4) Numerical simulation is the basis of

this model andis performed by advanced

numerical calculation techniques. The

models used include a spectrum method

wave estimation model, a physical factor

multiple regression wave prediction model,

a near- and far-f ieldtsunami calculation

model, and a storm surge multi-layer

model.

(5) Design wave analyses for Japanese

coasts, wave predictions for the Hitachi

Naka Harbor, tsunami calculations for the

Sanriku coast, and storm surge

calculations for the Ise Bay were presented as typical applications of this

system. The analysis procedure, the inter

relations between the analysis systems,

and output examples were described for

these applications.

REFERENCES

1) Goto, C. and H. Shibaki : A hindcast

of maritime surface wind including

effects of land topography, Report of

PHRI, Vol.32. No. 3, 1993. 2) Isozaki, I. and T. Uji : Numerical

model of marine surface winds and its

application to the prediction of ocean

wind waves. Papers in Met. and Geo.

,

Vol.25. No. 3, 1974. 3) Goto, C. , K. Suetsugu and T. Nagai : Wave hindcast model for short fetch

sea, Report of PHRI, Vol.29, No. 3, 1990. 4) Goto, C. , H. Shibaki and T. Aono : Multiple regression wave forecast

model described in physical parameters.

Report of PHRI, Vol.31, No. 5, 1993. 5) Goto, C. and T. Aono : A spectral wave prediction system for a single point.

Report of PHRI, Vol.31, No. 2, 1992.

6) Goto, C. and K. Sato : Development of

tsunami numerical simulation system

for Sanriku Coast in Japan, Report of

PHRI, Vol.32, No. 2, 1993.

Laser Shot Printer o U—V-7*)>2-

(Ethernet) oBiST*- messes

Digitizer

Optical Desk

(600 MB)

cm- v-^-jio

(32 Bit) (a<I;S*7'nt7+f--)

0/\- Kf < X?

(300 MB)

Personal Computer

1Pq

(> *>J- 64 MB)

(16 M (lops)

(1 GB)

Engineering Workstation

Video oii!Iff££Sg

Ti tr

Fig. 1 INSPECT system equipment configuration

565
Wave Height

o : ffi ± :& si

Z 18

(m)

ii jii i i i i:ii i i|iniini: m i n i m i m ili um

i

! o J L

12

i

8

6

A

2

Hi!iii iii i| i i i i! i ;! iii |i i ii ii ii i n|!iiiii ii : ii |ii i ii:iiii i

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i iiiii ii i ii iii i i ii i i i iiiiii ii iii i ij i Diiiiii i jiiii i nii i iJ i miiiim ^

1951 1952 1953 1954 1955 1956 1957 1958 1959 1960 1961 1962 1963 1964 1965 1966 1967 1968 1969 19705

(m)

1

1

. o_

* •

•

•

•

•

V

•

\ a

• •

1 o

© • 0 n ( o o

' o r-o-4 n

i Lu

<; 0

c i J 0

I

> o 1

1 1: 1i 0

:iii:l!liii 1 minimi 'iiiiiinii illlMiHii,!:!:!!!"!! •

Year

Fig. 2 Time series distribution of yearly maximum and extreme waves for the

statistical period

wsw

ENE WNW

ESE WSW

ENE

ESE

Fig. 3 Directional distribution of the probable wave and wave period corresponding

to probable wave height

CENT. PRESSURE AND TYPOON CONST.

DP (mb) 82 RO (kml

I 60 "1 1

1 1 L <00

140" -350

120' .. .300

100' .250

80" 200

60 '

| s . 1 50

40' 100

20' 1 50

0 li V o y

.......... |. ...... ....| 0 9. '23 9.24 9.25 CENTRAL PRESSURE

TYPOON CONSTANT

//////////////// I I/////////// WAVE HINOCAST BY SPECTRUM METHOD

18

16

14

m) ,T(») B2

t i i ri rrrnT;TTTTTTTTTTT |Ti rrrj-. » r . i " " i '~~

9.23 9.24 9.25' CAL . WAVE HEIGHT

CAL. WAVE PER I 00 OBS. WAVE HEIGHT

-OBS. WAVE PERIOD

Fig. 5 Screen output tor high wave prediction during a typhoon using the single

point output type wave hindcasting model

566
C/3 CDC -H ° § N COC3

C/3 E—C

S3

03

CO SZ

CO -r-> bO U

co co — x= o o

rHo. co

COQJ 8

cobO

cocx o 3

•i-l Vh — CX O "O

CO

C

O- CO

bo

fcL. o —1 4->CO CO 3= c o

~o t/"j c c

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+-» +J CO CO rH ,—1 e 3 c cO o c c r—1

CO

+-> 03 CJ 3 E-i -0 C

rH o

f-i • 1-1

-t-i CO +J

LO 3 CJ H cr • rH rH -a X3 cy

+-> cu

t—

H

Ih CO D. >

cu

a ca

CD CO CO (h rH cu _Q

d> CO

+-> c

ca xz c M •rH +J no e s_ CO e c —i E 3 >> • i—t o CO X >, rD

CO CO CO s: CO o CO • rH

U-.

-H-4-H I * C

0)>- 1—i

f-H -ai -ac

CO

CO

a>oecoo

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CG o T3

CO CD

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CO • rH c C/3 ge o >> o • H JS C • 1—1

+J

a o

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<d rH CD C/3 t-H CO XZ C/3 Q. ca +-> CD

CD rH

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s* 1—1

C/3 CD 3 CD rH 5* 3 C/3 Vh a O O •rH J= CD -O -r-> I > t-H r—l OJ CO CO 3

33 EE

be

567
S 10 li W 25 M J5 40 « 50 55

Fig. 8 Open sea calculation zones for storm surge estimation and distribution of

maximum storm surge deviations during Ise-Wan Typhoon

Fig. 9 Flow velocity distribution in Ise Bay and Mikawa Bay estimated by

2- layer model

568
Structural Testing of Confined Masonry for the

Third World Housing

by

Hatsukazu Mizuno Tetsuro Goto2

', Masanori Iiba 3

', Hiroto Kato4)and Yutaka Yamazaki

ABSTRACT

This paper presents structural testing of

confined masonry structure, which is selected as an appropriate candidate of apartment house

building in urban areas of the Third World

countries. It also includes a background

description of housing construction conditions of

the Third World. In static loading tests, about

half-scale specimen of confined masonry wall was used. Main experimental parameters are cross section size of reinforced concrete columns, axial reinforcement and shear

reinforcement of reinforced concrete columns,

and lateral reinforcements in brick wall

horizontal joints. Confinement effects of

reinforced concrete frame on brick wall, lateral reinforcement effects in horizontal joints and

axial stress are examined.

KEY WORDS : the Third World, Housing,

Economic Aspects of Construction, Confined

Masonry Structure, Structural Testing

1. INTRODUCTION

Urban areas of the third world countries

are subjected to excessive population

concentration due to some social and economic

conditions, and lack of healthy housing.

Development of economic and earthquakeresistant apartment house is required to supply massive housing in urban areas. Japanese Government has been and is carrying some technical cooperation programs,

especially a few research center projects with some of the Third World countries. A few

earthquake reconnaissance and technical

consulting in repair and retrofit of damaged

buildings were also done in a few earthquakes.

Through those experiences, We found something

in common as well as some differences in

structural engineering and construction

technology between them and us. Our

understanding on housing construction

technology motivated us to develop structural design method of appropriate masonry-based

construction method in urban areas of the Third

World countries.

2. PROBLEM DESCRIPTION

We surveyed earthquake damage,

conventional housing structures, prices of

industrialized basic construction in materials, e.g. steel bar, cement and masonry units in the third world countries, as a preliminary study. Our

understandings are as follows1 *;

(1) Earthquake Damage

Damage patterns of masonry structure types

are summarized as follows;

Adobe houses frequently suffered from

severe damage, especially total collapse due to heavy mud roof, and a lot of human lives were

lost, such as the Iran Manjil Earthquake of June

21, 1990. Unreinforced masonry also had

severe damage. Damage patterns of these two

types are overturning of walls in the out-ofplane direction and shear diagonal failure of

walls.Almost the same damage to brick masonry

buildings is found in the 1891 Mino-Owari

earthquake in Japan. Japan transplanted brick

masonry buildings in the second half of the

1) Program Director of International Codes and

Standards, Building Research Institute, 2)Senior

Research Engineer, Housing Construction

Division, BRI, 3) Senior Research Engineer,

BRI, 4) Research Engineer, Full-Scale

Structural Testing Division, BRI, 5) Director of

Research Planning Department, BRI

569
nineteenth century from Europe as a token of

modernization or industrialization. In the 1923

Kwanto earthquake, similar damage is reported.

Floors of these two types of masonry

structure are usually wood floor slab, concrete

joist slabs etc. In-plane stiffness of these slabs,

and box effects of floors and walls are smaller

than those of reinforced concrete slab. Damage to confined masonry is found in the 1985 Chilean earthquake2)

. Confined

masonry is composed of confining reinforced

frame and confined walls. After masonry walls

are setup, reinforced concrete frame is constructed in confined masonry structure. Typical damage patterns are 1) shear diagonal

failure of walls, 2) shear and bending failure of

heads and feet of reinforced concrete columns

and 3) separation between brick walls and

columns. The damage to confined masonry,

however, is not severe, compared with adobe

houses and unreinforced masonry.

Damage to masonry-infilled wall structure

is reported in the 1985 Mexican earthquake etc. After concrete frame is constructed, masonry

units like bricks are infilled as walls in the

structure. We have no data of damage to reinforced

masonry, in which hollow bricks or concrete blocks are used as permanent forms for reinforced concrete casting.

(2) Construction and Economic Aspects

1. Masonry structures prevail in many

earthquake-prone countries of the third world.

2. From building cost and dissemination

viewpoints, however, masonry-based apartment

buildings are most appropriate to supply

housing to the urban people, because masonry

units like bricks are comparatively cheaper than

industrialized materials. Steel bars and cement

have a tendency to be international trade goods

as well as crude oil, and the prices of these are expensive for people of lower GNP per capita.

3. We selected confined masonry structure, which is composed of confined masonry walls

and confining reinforced concrete frame

(sometimes steel frame), as an appropriate

apartment building on the basis of abovementioned building cost.

4. Confined masonry structure has been

studied and is already used in some countries.

But, more studies, especially experimental and

analytical ones are needed to improve structural design of the structure, because structural design data are not much accumulated.

3. RESEARCH OBJECTIVES

Building Research Institute has begun a research project of housing construction

technology transfer including this study since

1989 fiscal year. The objectives of the study are

to analyze damage patterns of masonry

structures, to examine improvement methods for minimizing earthquake damage, and to prepare

guideline of structural design with special

emphasis to common aspects of confined

masonry. Modification of the guideline to each

country is supposed to be carried out in other

research projects, because level of structural performance should be decided by each country,

and local conditions, especially social economic

conditions much differ.

4. STRUCTURAL TESTING

Static cyclic loading tests are intended for clarifying basic structural performance of

confined masonry structures. Figure 1 shows an example of Chilean

confined masonry apartment house. That seems

a little stiff, compared with other countries ones. We considered a wall in Fig. 1, as a reference model of the tests. Materials of

specimens are domestic (Japanese). Scale ratio of length in specimens is about half. Specimen

brick walls composed of full-scale bricks and

cement mortal joints, are considered to be

properly scaled on the basis of "spatially

integrated effects". We carried out the following two series of

static cyclic loading tests of confined masonry

wall units to clarify basic structural performance

of confined masonry, and to find most effective reinforcement arrangements of confined masonry

structures on the above-mentioned basis; No.l

series of the tests were executed to clarify

570
confinement effects of reinforced concrete frame

(column) on brick wall, which is one of the key

factors of confined masonry structure. No.

2

series of the tests were done to clarify to lateral reinforcement effects in brick wall horizontal

joints, which is one of possible measures of

improving structural performance of confined

masonry, and also to clarify effects of axial

stress on structural performance of walls due to overburden of the upper story of apartment

house. A set up of the tests is shown in Fig. 2.

(1) Confinement Effects of Reinforced

Concrete Columns

Figure 3 presents specimens of No. 1 series of the tests. Table 1 summarizes nine specimens

of the series, and experimental parameters and

conditions. Main experimental parameters are cross size of reinforced concrete column, axial reinforcement and shear reinforcement of

reinforced concrete column. Figure 4 depicts

range of axial reinforcement and shear

reinforcement of columns, which is determined

making reference to Peruvian, Chilean and

Mexican design codes.

Figure 5 shows envelope curves of relation

between shear stress and deflection angle of the

specimens. Shear stress of the figure is defined

as stress derived from dividing shear force by

total cross section area of brick wall and

columns.

Figure 6 shows rapture sketches at final loading stage of specimen CR-A, the strongest one, and specimen CR-I. The specimen CR-A

behaved stably and sudden rapture did not occur

during the test, though reinforced concrete columns were cracked due to bending moments,

and four corners of brick wall were cracked due

to shear. On the other hand, in the specimen

CR-I, shear cracks occurred at the right and

lower corner of the brick wall at about 0.0014

radian of deflection angle propagated into the

upper part of left reinforced concrete column,

and then loading capacity deteriorated. All

specimens except for CR-A were subjected to shear diagonal cracks at deflection angles less than 0.0025 radian, and loading capacity

deteriorated. Shear diagonal cracks penetrated

into reinforced concrete columns at final loading

stages of the specimens.

Figure 7 presents relationship between

maximum shear strength values of the specimens, and axial reinforcement and shear

reinforcement of reinforced concrete columns.

Maximum shear strength tends to increase with

axial reinforcement of reinforced concrete columns. But maximum shear strength is not found to be related with shear reinforcement of

reinforced concrete columns. Table 2

summarizes tested values and calculated values

of maximum strength of the nine specimens. In the specimen CR-B, a loading capacity value at the first large shear diagonal crack is defined as maximum strength, though the CR-B recovered

somewhat loading capacity due to compressive

struts of brick wall in cyclic loading process,

after loading capacity deterioration by the first large diagonal crack.

In Table 2, proposed evaluation equation

Ql of maximum strength in confined masonry

structure is derived from adding crack strength

evaluation of brick wall to maximum strength

evaluation equation of reinforced concrete columns. The derivation procedure is as follows;

Qcr = wQcr + £CQcr> where

WQCT ; Shear Crack Strength of Brick Wall

(Principal Stress Evaluation Formula),

cQa ; Shear Crack Strength of Column

(Ohno-Arakawa Evaluation Formula)

Explicit expression of WQCT is; wQcr = 0.5(op)a5tL/K

Ql = WQCT + ECQSU , where

cQsu : Max. Shear Strength of Column

(Ohno-Arakawa Mean Formula).

Explicit expression of ,,0^ is;

CQSU = {0.0679 Pla23(180+aB)/(M/Qd+0.12) +

2.7(Pw a^)05 + 0.1 aj bj

where

pt

: Ratio of Tensile Axial Reinforecement

Area and Cross Section area of Column

j : w

7d/8. where d: effective height of cross

section

571
aB

: Compressive Strength of Column

Concrete

M/Qd: Shear Span Ratio of Confined Masonry

pw : Ratio of Shear Reinforecement in Reinforced Concrete Column

: Yielding Strength of Shear

Reinforecement Bar in Reinforced

Concrete Column

a0: Axial Stress of Confined Masonry Wall

op: Prism Compressive Strength of Brick

Wall

t : Thickness of Brick Wall

L : Length of Brick Wall

K : Shape Factor (1.0)

The equations were modified from shear

crack evaluation equation and maximum strength

evaluation equation of reinforced concrete members used in Japan.

Figure 8 shows relationship between ratio of tested value exQu and calculated value Ql,

exQu/Ql and axial reinforcement of reinforced concrete columns.

Another evaluation equation Q2 of

maximum strength is used for reinforced concrete block structure in Japan6)

, as shown in Table 2.Figure 9 compares tested values of

maximum strength exQu with proposed equation

Ql and another equation Q2. Proposed

maximum strength evaluation equation Ql

agreed with the tested values. Evaluation

equation Q2 overestimates the tested values.

(2) Effects of Lateral Reinforcement in Brick

Wall Joints and Axial Stress

Table 3 summarizes nine specimens of

No.2 series of the tests. Figure 10 shows the specimens and experimental parameters, namely

lateral reinforcement arrangement in horizontal

joints of brick walls, and axial stress due to overburden of the upper story of apartment

house. Figure 11 demonstrates envelope curves

of relationship between loading force and

horizontal deflection.

Specimens I, P and Q without lateral reinforcement in horizontal joints of brick wall was damaged in separation between brick wall

and reinforced concrete columns at stages of

small deflection. And then shear diagonal cracks

occurred and lost bearing capacity of vertical loads. Figures 12 demonstrates loading force versus deflection relationship of specimen M,

compared with that of specimen I without lateral reinforcement. From the figure, lateral reinforcement improves behavior, especially

ductility.

Specimens M, N and O with column-tocolumn lateral reinforcement showed ductile

behavior up to large deflection, namely up to deflection angle 0.0020 radian.

Specimens J, K and L with partial

reinforcement showed shear failure at stages of

deflection angle 0.0020 radian, after axial reinforcement yielding of reinforced columns.

Partial reinforcement means to improve ductility,

compared with specimens I without lateral reinforcement. Improvement by partial

reinforcement is less than that of specimens M,

N and O with column-to-column lateral reinforcement.

Figure 13 demonstrates effects of axial

stress due to overburden by comparing specimen

Q of axial stress 14kgf/cm2 with specimen I of

axial stress 4kgf/cm2

. High axial stress makes

maximum strength large, but sudden

deterioration of loading capacity is remarkable.

Figure 14 shows sketches of final stage

rapture of specimens Q and M.

Maximum Shear strength is improved 1.2 -

1.3 times by adding lateral reinforcement.

Specimen Q with high axial stress, 14 kgf/cm2 maximum shear strength is about 1.5 times as large as specimen I with low axial stress. From construction cost viewpoint, partial

reinforcement at four corners of brick wall is recommended, because specimen K with the smallest lateral reinforcement showed good

behavior, compared with the other lateral reinforcement.

5. CONCLUDING REMARKS

Basic structural performance of confined

masonry structure is revealed as follows;

1. Axial reinforcement of reinforced concrete columns increases maximum strength of confined

masonry structure from semi-quantitative

572
viewpoint.

2. Lateral reinforcement in horizontal joints of

brick wall and shear reinforcement of reinforced concrete columns both improves ductility of

confined masonry structure. The former effects

is more remarkable than the latter.

ACKNOWLEDGMENT

No.2 series of the tests was carried out as cooperative research project with "Structure Subcommittee" (Convener: Ex-Director of

Structural Department, BRI, Now, Team leader

of CENAPRED, Mexico) under the auspice of

"Committee on International Asesmic Network

of Buildings and Housing" (Chairman: Professor

T. Okada, Tokyo University). The fund of the

cooperative research was supported by Japan

International Cooperation Agency. The authors

would like to express their thanks to whom

concerned.

REFERENCES

1. Mizuno, H. et al., "Some basic considerations on appropriate housing construction technology

in the Third World countries," Kenchiku

Kenkyu shiryo, No. 68, Building Research

Institute, Mar. 1990 (in Japanese)

2. Cruz, E. et al., "Lecciones del sismo del 3 de

Marzo de 1985," Instituto Chileno del Cement y

del Hormigon, 1988 (in Spanish)

3. Ishibasi, K. et al., "Aseismic structural testing

of confined masonry in Mexico (Part 1) & (part

2)," Proceedings of A.I.J. Convention, ppl969-

1972, 1992, (in Jpanese)

4. Katsumata, H. et al, "Nonlinear analysis of

confined masonry walls (Part 1) & (Part 2),"

Proceedings of A.I.J. Convention, pp325-328,

1993, (in Japanese)

5. Kato, H., Goto T., Mizuno H. and Iiba, M.

"Cyclic loading tests of confined masonry wall

elements for structural design development

apartment houses in the Third World,"

Proceedings of the tenth World Conference on

Earthquake Engineering, Balkema, vol.6,

pp3539-3544, 1992

6. Architectural Institute of Japan, "Ultimate

strength and deformation capacity of buildings in seismic design," 1990, (in Japanese)

573
L

i ioo soo 2500

gso

~o 8

cs L

1 1 trrva u j j—c

i—/—L-±l oo

r-&

1

1 ]\A 8OO 8

7200

3500

\ -

\

l 200 1700 |

\

/ '/' l innn

\ MOOl 1600 "l30ol 1500 1 1700

2700 4500

7200 Fig.l Apartment House damaged in 1985

Chilean Earthquake (Plan & Elevation Views)

CR-A-D

Hydraulic Jack

for Axial Load

Hydraulic Jack

for Horizontal Lonrj

Load Cell

Drift angle

R= d/U

Fig. 2 Setup of Static Loading Test

=fcrlck fall Brick :210XI00X£0»b _ Layer :|9 d llortal Joint: 10bi

i j_a i n u r I X

1 200 | 435 I 435

100

.(un I t : as) Fig. 3 Specimens & Reinforcement

Arrangement (Confinement Effects)

574
Table 1 WALL SPECIMENS OF STATIC TESTING ( No . 1 Series)

(CONFINEMENT EFFECTS OF R.C. FRAME)

Speci- men Size Axial

Reinf.

of RC Column

Shear

Reinf. of RC Column

Strength of Brick

Prism

Stren- gth of Concre

te

Yield

Stren- gth of Steel

CR-A Wall : 870x100

nun Col. : 200x100

mm

6-D13 D6-@50411.1

kgf/

cm2

207.2

kgf/ cm2

D13 : 3544.2

D8 : 4145.4

D6 : 3778.0

kgf /cm2

CR-B D6-@220

CR-C

4-D8 D6-@50

CR-D D6-@220

CR-E Wall : 870x100

mm

Col. : 100x100

mm

4-D8 D6-@50

320.9

kgf/

cm2

258.0

kgf/ cm2

D8 : 4208.4

D6 : 4694.4

D4 : 3369.5

D3 : 3805.6

kgf /cm2

CR-F D4-g80

CR-G 4-D6 D4-@40

CR-H 4-D4 D3-Q45

CR-I 4-D8 D4-@40

5.

4.

^ 2.

1.

0.

V A Designed Building in Chili A A Designed Building in Peru

!

CR-D \

i

1

l

CR-A \ • •

CR-F \CR-I CR-E

-

\CR-G

i i CR-C \

QR-1I i

? • i

I , i

0.5 1.0 Pw (%)

1.5 2.0

Fig. 4 Experimental Parameters

(Axial Reinf. of RC Columns

versus Shear Reinf. of Columns)

.0 -1.5 -1.0 -0.5 0 0.5 1.0 1.5 2.0 2.5 Drift Angle (%)

20.

15.

E 10.

5.

-5.

-15.

-20.

CR-I CR-H

CR-G

CR-F

CR-E

-2.5 -2.0 -1.5 -1.0 -0.5 0 0.5 1.0 1.5 2.0 2.5 Drift Angle (%) Fig. 5 Envelopes of Shear Stress - Deflection Angle Relation

575
20.

CNE

16.

12.

«j

8. a

Shi 4.

CR-A

CR-C

I

'

I

CK-fl

i

_ ...! CR-D

= •

CR-i.f

CR-G

CR 1,1.1

I

'

I

' I

1 I ' I

! i i CR-F

j CR-C j !•;!!! CR-I) * 5 C7?-f

CK-/<

cr-d

aCR-l

i CR-G

i ' i i

2.0

1.8

1.6

1.4 CJ>1.2

HM.O

0.3

0.6

0.4

'

1

'

1 i • i i 1

'

1 li j CR-F

j s : CR-B

'

• •- lilt

CR-D

j

! 1

! CR-A . •

0

...CR-C... — ~^

c*-c j «-£

-

i I.I.I. I

2.0 1.5 1.0 0.5 0 0.5 1.0 1.5 2.0 2.5 3.0 3.5 4.0 /V (%) Pg (%)

20. 40. 60. 80. 100. 120. Pg f, (kg/lcm*)

140.

Fig. 8

Fig. 7 Max. Strength versus Experimental

Parameters (Axial Reinf. of RC Columns,

Shear Reinf. of RC Columns)

exQu /Qi Versus Axial Reinf,

of RC Columns

56

30.

25.

20.

15.

i

! 10.

5,

0.

—1—1— ' 1

"

1

'

1

1 +20%/

j "CRTA /

\ CR-B •' /

CR-F A A _ m

...;..2/.~... cr-c

. CR- E ^i^CR--D \ CR-l-^Yy-

.\ACR-G

CR-H A. i . i . i . i . i .

35.

30.

25.

§ 20. a 15.

H

5. 10. 15. 20. 25. 30. Ql(CaL) (ton/) (a)Ql t<D&&

10,

5.

0.

, j . j

1 | 1 | 1 | y , ,y

\ +207* / \/

-20%

~Cr-a~

CR-FyS CR-'C\

'• CR-B -

. '/ a • •• CR-D /-< "

«

/Ay cr-g

.•' \

\ \ \r-i

A.\ \ CR-H A. i . i . i . i . i . i .

5. 10. 15. 20. 25. 30. 35. Q2 (Cal.) (ton/) (b)Q2 t Oft^t

Fig. 9 Comparison of Tested & Calculated Max. Strength

Table 2 COMPARISON of tested and calculated maximum strengths

(tonf

)

Specimen Tested

Max. Streng.

exQu

Cal. Shear Crack

Streng

.

Qcr

Calculated

Maximum Shear

Strength

Ratio of Tested

and Cal. Max. Strengths

Ql Q2 exQu /Ql exQu /Q2

CR-A 22.0 12.43 17.43 32.07 1.26 0.69 CR-D 20.3 12.43 14.59 32.07 1.39 0.63 CR-C 16.3 12.43 17.22 21.68 0.95 0.75 CR-D 14.7 12.43 14.43 21.68 1.03 0.68 CR-E 12.8 10.89 11.89 18.89 1.08 0.68 CR-F 14.9 10.89 10.22 18.89 1.46 0.79 CR-G 10.6 10.89 10.75 16. 66 0.98 0.64 CR-H 6.0 10.89 10.29 12.74 0. 58 0.47

CR-I 11 . 8 10.89 10.75 18.89 1.10 0. 62

LEGEND

0cr - uQcr + E c Qcr , where uQcr : Shear Crack Strength of Wall

(Principal Stress Formula), c Qcr : Shear Crack Strength of Column

( Ohno-Arakawa Formula)

Q 1 " wOcr + ^cQiu< where C QC<J : Max. Shear Strength of Column (OhnoArakawa Mean Formula),

020 ' V*p< h/D+o.7

+ ° 04 ' ^+ ° 108 V/P)^V°>0 2o o> t'j

576
Table 3 WALL SPECIMENS OF STATIC TESTING (No. 2 Series)

(EFFECTS OF LATERAL REINFORECEMENT & AXIAL FORCE)

SPEC

IMEN COMMON EXPERIMENTAL

PARAMETERS

BRICK

TYPE

( Groov

es

)

LATERAL REINFORCEMENT

IN HORIZONTAL JOINT

OF BRICK WALL

REMARKS

No. I •a/D : 0.62 •Column Size

10cm x 10cm •Ax.Reinf .of

NoGroove

(0)

No Reinf. The same specimen

as No. P

& No . Q J Column : 4-D8 (P, = 1.98%)

• ShearReinf

.

of Column

D4-Q40 (P„

- 0.64%)

•Brick with

& without

Groove

(1)

D4 : Single Through

Upper Five Layers

Five

Joints ti H/4

K D4 : Single

Partially Reinf. at Four Corners of Wall

I Up. ol LiOW . O J . )

Anchor: 1

.

5 times

of brick

leng th

L Groove Size : 100 x 210 x 60mm,

No. J - N : 1 Groove,

No. Q : 2 Grooves

•Concrete Strength

210kgf/cmJ

•Axial

S tres s ( o )

D4 : Single Through

Each Two Joints

The same

reinf. as

No. J

M D4 : Single Through

Each Joint

N D8 : Single Through

Each Joint

The richest

reinf

.

0

* 4kgf/cm! Groove

(2) D4 : Double Through

Each Joint

The 2nd richest

reinf

.

P The same parameters as those of No.

I - 0 except

NoGroove

(0)

No Reinf. AXIAL STRESS (Normal stress due to Axial

Force) : 7kgf/cm2

Axial Stress of

No. I : 4kgf /cm2

Q for AXIAL

STRESS No Reinf. AXIAL STRESS (Normal Stress due to Axial

Force) : 14kgf/cmJ

3 15 100 8 7 0 100 3 1 5 3 I 5 100 J 7 0 100 3 15

1" "( Brick v-IF/

1" "I Bnck

9

>J:1-D8 wall bar

RC :100x100mm aJ~W:10d>DWI))

j 20mm groov c3

0:2-D-1

joint ihicknraa :10mm

unit '. mm

Fig. 10 Specimens & Reinforcement Arrangement (Effects

of Lateral Reinf. of Wall & Axial Stress)

577
Fig. 11 Envelopes of Loading Force Versus Hor. Deflection

578
20

I

10

a

-10

Unreinfoi

Q (High

;

I (Norm

ced wall

jxial force)

al axial for cc)

fjk

N

- " /

/Ax- (/

H,

- 0 J "(i j fiitl f j

»' ii

1 '/ 7

i ' —

I

\ r

f

-30 -20 -10 0 10 20 30

Horizontal Deflection,8(mm)

Fig. 13 Loading Force Versus Hor. Deflection of Specimen Q

5n

Z5E» I 1 ! I Vi 1 St

TTT

1

Speicimen M

Classified (Shear Type :I.P,Q.) , Flexure and Shear Type :KJJL)

Failure Mode: (Flexure Type :M,N,0)

Fig. 14 Sketches of Final Stage Rapture of Specimens Q & M

579

Guideline for Safety Evaluation of

Structural Response-Control Building (Tentative)

by

Kitagawa Yoshikazu u

, Midorikawa Mitsumasa2)

, Iiba Masanori3

', Fujitani Hideo3) and Kashima Toshihide 4'

ABSTRACT

At present, structural response-control

technology for practical use in buildings is developed mainly by private companies aiming

to improve the living comfort against strong

winds or moderate earthquakes. A general

social consensus of this type of new building is not yet to be reached. Therefore, it is desirable

that safety evaluation procedures of structural response-control buildings are established to not interfere with but encourage free development of the state-of-the-art. The

Building Research Institute promoted the

project of the government/private sector join

research & development titled "Research &

Development to Achieve Active Structural

Response-control of Building Structures" from

1990 to 1993. The part of the results was

reported as the "Guideline and Commentary for

Safety Evaluation of Structural Responsecontrol Buildings (Tentative)". This guideline

mainly consists of three chapters; "General",

"Safety Evaluation of Structures", "Safety

Evaluation of Response-control Devices".

Presented is the Tentative Guideline in this report.

Key words: Safety evaluation, Structural

response-control, Building, Earthquake, Wind

INTRODUCTION

Structural response-control technology

for practical use in buildings is presently

developed mainly by private companies aiming

to improve the living comfort against strong

winds or moderate earthquakes. Structural

engineers, however, have not yet reached a consensus on the design philosophy and safety

of structural response-control buildings, and a general social consensus of this type of new

building is not yet to be developed. Under this situation, it is desirable that safety evaluation

procedures of structural response-control

buildings are established to not restrict but

promote the development of this new

technology. The Structural Response-control

Research Committee was established for this purpose in the Building Center of Japan in 1990 as part of the project, partly sponsored by

the Ministry of Construction, of the

government/private sector join research &

development titled "Research & Development

to Achieve Active Structural Response-control

of Building Structures". This committee was

in charge of developing the safety evaluation

guideline for structural response -control

buildings, and was chaired by Professor

ENOUE Yutaka of the University of Osaka.

The work of the committee was reported as the

"Guideline and Commentary for Safety

Evaluation of Structural Response-control

Buildings (Tentative)" in 1993 1) on which this report is based. This guideline mainly consists

of three chapters; "General", "Safety

Evaluation of Structures", "Safety Evaluation

of Response-control Devices". In this report,

we present the Tentative Guideline.

CONTENTS

1. General

1.1 Scope

1.2 Definitions

1.3 Classification of structural responsecontrol buildings in safety evaluation

1.4 Safety evaluation of structural response-control buildings

1) Director, International Institute of

Seismology and Earthquake Engineering,

Building Research Institute, Ministry of

Construction, 1 Tatehara, Tsukuba, Ibaraki,

305 Japan.

2) Head, Structure Division, Structural

Engineering Dept., BRI, MOC.

3) Senior Research Engineer, Structural

Engineering Dept., BRI, MOC.

4) Research Engineer, USEE, BRI, MOC.

581
2. Safety Evaluation of Structures

2.1 Outline of buildings and ground

conditions

2.2 Design principles and safety evaluation contents 2.3 Setting of design conditions

2.4 Design of structures 2.5 Detail design at response-control device

location

2.6 Earthquake and wind response analysis

of structural response-control buildings

2.7 Effects to structures in unusual

condition of active control devices

3. Safety Evaluation of Response-control

Devices

3.1 Composition and specifications of

response-control devices

3.2 Control design of active control devices

3.3 Operation program of active control

devices

3.4 Performance validation and durability

of response-control devices

3.5 Maintenance inspection of responsecontrol devices

3.6 Environmental safety

4. Others

4.1 Construction

GUIDELINE

1. GENERAL

1.1 Scope

This guideline is applied to the safety

evaluation of a structural response-control

building which has an active control device.

The items that are not shown in this guideline

shall be followed by the Building Standard

Law of Japan and the related specifications and

recommendations.

1.2 Definitions

The definitions of the following apply

to this guideline.

Structural response-control building: A

building that has an active or hybrid control

device receiving the supply of external energy

to carry out the response control of the building

for earthquake ground motions or wind

loading.

Response-control device: A device carrying out

the response control of a building for

earthquake ground motions or wind loading.

Active control device: A response-control

device receiving the supply of external energy

to carry out the response control of a building

for earthquake ground motions or wind

loading, which is composed of the sensing

system, the control system, and the drive

system.

Passive control device: A response-control

device that needs no external energy to carry out the response control of a building for

earthquake ground motions or wind loading.

Hybrid control device: A response-control

device composed of an active and passive

control devices that are arranged in a series or

in parallel to carry out the response control of a building for earthquake ground motions or wind loading.

Level 0 earthquake ground motion: An

earthquake ground motion that has a smaller

level than level 1 earthquake ground motion.

Level 1 earthquake ground motion: The largest

earthquake ground motion expected to occur at a building site once or more during the life time

of a building.

Level 2 earthquake ground motion: The largest

earthquake ground motion believed to have

occurred, and that considered to possibly occur

in the future, at a building site. Level 0 wind loading: The wind loading that has a smaller level than level 1 wind loading.

Level 1 wind loading: The largest wind loading

at the wind velocity expected to occur at a building site once or more during the life time

of a building.

Level 2 wind loading: The largest wind loading

at the wind velocity believed to have occurred,

and that considered to possibly occur in the

future, at a building site. Relaxation of design conditions: Setting the

relaxed design conditions, based on the

judgment of a designer, such as design loads

and structural specifications that correspond to the response reduction or the structural characteristics of a structural response-control

building.

Operation limit of active control devices: A

level of upper limit of earthquake ground

motions or wind loading at which an active

control device operates effectively.

Criteria for earthquake and wind resistance:

The criteria that structural response-control

buildings should satisfy in order that structural response-control buildings shall possess the

seismic and wind safety equivalent to or more

than that of ordinary buildings for earthquake

ground motions and wind loading.

Target performance: A target value of the

response-control effect that is set by the

mi
designer or the client regardless of the seismic

and wind safety of a structural response-control

building.

Safety margin: The degree of safety allowance

that a structural response-control building holds

for the design conditions.

Block diagram: A summary figure that shows

the connection and composition of various

kinds of elements of an active control device.

Backup system: A system that substitutes the

function and that secures the safety when a response-control device is in unusual

conditions.

Quantity of state: The information that expresses the motion of structure and responsecontrol device, and the control state of

response-control system.

Sensing system : A system sensing the motion

of structure and response-control device, and

the condition of transmitted signals.

Control system: A system generating the

control signal to drive the drive system, based on the quantity of state sensed, in order that the

response-control system may show the

response-control effect. Drive system: A system that receives the

control signal and drives the response-control

device.

Signal processing: Applying operations on the

signal indicating the quantity of state sensed.

Control theory: A theory used to make the

control signal that is sent to the drive system

based on the quantity of state. In unusual condition of active control devices:

In unusual condition that is impossible to prevent and is unpredictable in the control

design, maintenance, and restoration design of an active control device.

At power failure: In condition that the all electricity supply to an active control device is cut off from some unpredictable causes.

In excitation: Shifting from the normal control

state to the state exciting for a building from

some unpredictable causes. Performance monitoring: Monitoring the sound

performance of an active control device.

Environmental safety: Safety in relation

between the response-control device and its surroundings.

1.3 Classification of Structural Responsecontrol Buildings

Structural response-control buildings

are classified based on the relaxation of design

conditions of structure and the operation limit

of active control device as shown in Table 1.1. (Refer to Fig. 1.1)

1.4 Safety Evaluation of Structural Responsecontrol Buildings

Safety of a structural response-control

building is evaluated for the structure and the response-control device. Necessary evaluation

items are determined according to the

classification of structural response-control

buildings in Table 1.1. (Refer to Table 1.2)

2. SAFETY EVALUATION OF

STRUCTURES

2.1 Outline of Buildings and Ground

Conditions

The use, scale, shape, structure,

dynamic characteristics of a building, and the

site/subsoil conditions and circumstances of the

construction site shall be made clear.

2.2 Design Principles and Safety Evaluation

Contents

In the safety evaluation of a structural response-control building, the design principles

and evaluation contents shall be confirmed

according to the following. Necessary

evaluation items are determined according to the classification of structural response-control

buildings in Section 1.3.

1) Design principles such as the goal of

structural response-control building, the criteria for earthquake and wind resistance of structural response-control building, the operation limit of active control device, and the relaxation of

design conditions.

2) The structural design and the response

analysis results for earthquake ground motions

and wind loading that satisfy the criteria for

earthquake and wind resistance.

3) Safety margin.

2.3 Setting of Design Conditions

The design conditions such as external

disturbances on a structure, design loads acting on a structure, and others shall be determined

taking into account the site, subsoil, and

structural conditions of structural responsecontrol building.

2.4 Design of Structures

2.4.1 Design for Permanent Loads

Structural members shall be

investigated for dead and live loads, and snow

load where applicable, so as not to exceed the

583
permanent allowable stress as specified in the

Enforcement Order of the Building Standard

Law, in addition to an appropriate check for

serviceability.

2.4.2 Design for Design Seismic and Wind

Loads

Structural members shall be

appropriately investigated for the design

seismic and wind loads considering the

dynamic characteristics of structural responsecontrol building and the site conditions.

2.4.3 Ultimate Limit State of Structures

The ultimate limit state of structure shall be investigated corresponding to the structural characteristics.

2.5 Detail Design at Response-control Device

Location

The details at the location of response

-

control device shall be designed appropriately

considering the aspects of architectural

planning and design, building equipment, and

structural details.

2.6 Earthquake and Wind Response Analysis

of Structural Response-control Buildings

2.6.1 Response for Earthquake Ground

Motions

(1) Criteria for Earthquake Resistance

In general, a structural response -control

building must be shown to respond to two

levels of earthquake ground motions as follows.

1) Under level 1 earthquake ground motion, the structure remains in elastic range, and the

response-control device maintains the function.

2) Under level 2 earthquake ground motion, the structure remains within general yield of stories

(or within permissible ductility factor), and the

response-control device maintains the function.

(2) Input Earthquake Ground Motions

Earthquake ground motions used as

input in the response analysis of structural response-control building shall be determined

appropriately reflecting the locality of the

construction site. The intensity of earthquake

ground motions must be the two levels as follows.

1) Level 1 earthquake ground motion: The

largest earthquake ground motion expected to occur at the building site once or more during

the life time of the building.

2) Level 2 earthquake ground motion: The

largest earthquake ground motion believed to have occurred, and that considered to possibly occur in the future, at the building site. (3) Analytical Method and Mathematical Model

An analytical method shall be adequate

for the calculation of the significant features of

its dynamic response. A mathematical model

of the physical response-control building shall be used which adequately represents the

dynamic characteristics of structure and

response-control device within the expected

response.

(4) Evaluation of Seismic Safety

Criteria for earthquake resistance in two

levels must be shown to be satisfied by

earthquake response analysis.

2.6.2 Response for Wind Loading

(1) Criteria for Wind Resistance

In general, a structural response-control

building must be shown to respond to two

levels of wind loading as follows.

1) Under level 1 wind loading, a building

suffers no loss of serviceability and function

caused by the structural deformation without

damage, and the response-control device

maintains the function.

2) Under level 2 wind loading, the structure remains in elastic range, and the responsecontrol device maintains the function.

(2) Wind Loading

Wind loading used as input in the

response analysis of structural response-control

building shall be determined reflecting the wind

environment of the building and the

construction site. The intensity of wind

loading must be two levels in terms of

reference wind speed based on mean wind

velocity as follows.

1) Level 1 wind loading: The largest wind

loading at the wind velocity expected to occur

at the building site once or more during the life time of the building.

2) Level 2 wind loading: The largest wind

loading at the wind velocity believed to have

occurred, and that considered to possibly occur

in the future, at the building site. (3) Analytical Method and Mathematical Model

An analytical method shall be adequate

for the calculation of the significant features of

its dynamic response. A mathematical model

of the physical response-control building shall be used which adequately represents the

dynamic characteristics of structure and

584
response-control device within the expected

response.

(4) Evaluation of Wind Safety

Criteria for wind resistance in two

levels must be shown to be satisfied by wind

response analysis.

2.7 Effect to Structures in Unusual Condition

of Active Control Devices

2.7.1 Principles

This section is applied when there is inconsistency in the safety evaluation of

response-control devices in Chapter 3, or when

there is no reliable guarantee of normal

operation of response-control devices.

2.7.2 Unusual Condition

There are two conditions at power

failure and in excitation for a building as unusual condition considered.

2.7.3 Safety of Structure in Unusual Condition

Under the unusual condition

considered, a structure must be shown to be

satisfied by either of the following.

1) Safety of structure taking the unusual

condition into consideration.

2) Safety of structure in the state of certain stop

of active control device.

3. SAFETY EVALUATION OF RESPONSECONTROL DEVICES

3.1 Composition and Specifications of

Response-control Devices

The composition and specifications of

response-control device shall be determined

adequately so as to meet the target

performance.

3.2 Control Design of Active Control Devices

The control system of active control

device shall be designed adequately

corresponding to the target performance.

3.3 Operation Program of Active Control

Devices

An active control device shall be

adequately operated, within the safety

guaranteed range, based on the operation

program according to the control design of

active control device.

3.4 Performance Validation and Durability of

Response-control Devices

3.4.1 Performance Validation Test

The sound performance of responsecontrol device concerning with control shall be

validated by the following.

1) The fabrication and product inspection

procedures

2) The test program validating the operation

and performance of response-control device,

and the performance test and evaluation of

response-control device according to the test program.

3.4.2 Durability

The contents and time of maintenance

inspection shall be determined based on the life time and durability of each component of

response-control device.

3.5 Maintenance Inspection of Responsecontrol Devices

3.5.1 Type of Inspection

The maintenance program of responsecontrol device and its vicinities shall be

determined adequately. In general, the

inspection of this purpose shall consist of three

types of inspections; ordinary, periodical, and

temporary inspections.

3.5.2 Inspection Contents

Inspection points, inspection items,

inspection procedures, criteria, and

countermeasures when not meeting the criteria, shall be determined in each of ordinary,

periodical, and temporary inspections.

3.5.3 Performance Monitoring of Responsecontrol Devices

The performance monitoring system

keeping check the sound operation of responsecontrol device shall be prepared.

3.5.4 Countermeasures of Malfunction

Diagnosis procedures, criteria of

malfunction, and countermeasures shall be

established in advance against malfunction of

response-control device, and the

countermeasures shall be taken immediately

when malfunction of the device was

discovered.

3.5.5 Maintenance Organization

The maintenance organization shall be

established which comprises members such as

the owner, administrator, designer,

constructor, and maintenance administrator of a building.

585
3.6 Environmental Safety

3.6.1 Active Factors

Countermeasures reducing the effects of the existence and operation of responsecontrol device on existing human beings and

things around it shall be adequately taken

investigating the factors to be considered and

the degree of effects.

3.6.2 Passive Factors

Countermeasures keeping a responsecontrol device in normal condition shall be

adequately taken investigating the factors and

influence from the condition that is required to keep the device normal by its surroundings,

and that possibly affects the device externally.

4. Others

4.1 Construction

A structural response-control building

shall be constructed appropriately considering

the specifications of building and the safety of

construction work.

CONCLUDING REMARKS

The Tentative Guideline for Safety

Evaluation of Structural Response-control

Buildings is presented that may be needed

further to be polished because the development

of structural response-control buildings is still in the initial state. The commentary following

the guideline will be introduced in another

report. In general the dynamic characteristics of

a physical building include the uncertainties

caused by the scatter and secular change of

structural strength, materials, and construction

quality. Therefore the uncertainties of various

factors is needed to be investigated when

constructing a structural response-control

system. Further studies on the reliability,

safety margin, and response-control effect index are needed because of insufficient data of

uncertain factors.

Reference

1) Report of the Project on Research &

Development to Achieve Active Structural

Response-control of Building Structures,

Ministry of Construction, Government of

Japan, March, 1993 (in Japanese).

Table 1.1 Classification of Structural Response-control Buildings

Relaxation of

Design Conditions

Operation Limit of Active Devices

Level 0 Earthquake

Motions or Wind

Loading

L-0

Level 1 Earthquake

Motions or Wind

Loading

L-l

Level 2 Earthquake

Motions or Wind

Loading

L-2

Not Relaxed

R-l

A B C

Relaxed

R-2

D E F

Notes:

1) Relaxation of design conditions means that some design conditions of a structural response-control

building are relaxed considering the effects of response-control device in comparison with an ordinary

building.

2) Operation limit of active control devices means a level of upper limit of earthquake ground motions

or wind loading at which an active control device operates effectively.

586
Start

Passive Control

Yes

Out of Application of Guidline

Fig. 1.1 Flow of Classification of Structural Response-control Buildings

587
Table 1.2 Safety Evaluation Items for Structural Response-control Buildings

Class of Response

-

control Buildings Evaluation TtPtrK RprfiorVc

A B c D E F

(1) Outline of Building to be Evaluated (To understand outline of a building to be evaluated)

1 . Outline of Ground

0 0 0 0 0 0 a. Site

0 0 0 2 2 2 b. Conditions around site (topography, environment)

(J u u 2 2 2 c. Subsoil conditions

0 Outline rvF Rinlrlinrr Z. VJULIITIC OI JjUllULflg

0 !) 0 0 0 0 a. Use

A A

u

A

u

A

u u u b. Scale and shape

0 0 0 2 2 2 c. Structure (structural type, foundation)

2 2 2 2 2 2 3. Outline of Response-control Device

a. Type of system Active, Hybrid

b. Size and shape of device

c. Setting location

4. Design Principles

2 2 2 2 2 2 a. Goal of response-control Living comfort, Function,

Earthquake and wind resistance

2 2 2 2 2 2 b. Criteria for earthquake and wind

resistance

1

1 11 1 1

11 1 c. Target performance of earthquake and

2 2 2 3 3 3 o. uperanon umn oi active control oevice

0 0

->3 3-> 3 e. KeidAdiion oi uesign conuinons Keiaxeci or noi, ixeiaxeu contents

(2) Safety Evaluation or Structure (To confirm that a structure is designed appropriately considering

5. Design Conditions of Structure earthquake and wind response of

0 0 0 0 0 0 a. Materials response-control building)

0 0 0 3 3 3 b. Design loads (permanent load, seismic

load, wind load, others)

{) 0 0 3 3 3 c. Story drift limitation

U u u 3 3 3 d. Structural specifications

0 0 0 0 0 0 a. Design for permanent load

0 0 0 0 0 0 b. Design for temporary load

0 1 1 0 3 3 c. Ultimate limit state

7. Detail Design at Response-control Supporting structure, equipment,

2 2 2 2 2 2 Device Location others

588
Table 1 .2 Safety Evaluation Items for Structural Response-control Buildings (Continued)

Class of Responsecontrol Buildings Th x 7Q n i q ti Itpinc -CVdlUaliUIl ILCIIJb A B C D E F

8. Response for Earthquake Ground

Motions

0 2 3 2 3 3 a. Input earthquake ground motions Locality, Level 1 , Level 2

0 2 3 2 3 3 b. Analytical method Time history response analysis

0 2 3 2 3 3 c. Mathematical model Structure, Response-control device

0 1 2 2 3 3 d. Evaluation of seismic safety Criteria for earthquake resistance,

oatety margin

9. Response for Wind Loading

0 2 3 2 3 3 a. Wind loading Level 1, level 2

0 2 3 2 3 3 b. Analytical method Time history response analysis,

rrequency response analysis

0 2 3 2 3 3 c. Analytical model Structure, Response-control device

0 1 2 2 3 3 d. Evaluation of wind safety Criteria for wind resistance, Safety

margin

10. Effect to Structure in Unusual

Condition of Active Control Device

1 2 3 1 3 3 a. Necessity of consideration on unusual Reliability on normal operation,

condition Simultaneous effects of earthquake

and wind

1 1 1 1 3 3 b. Safety of structure at power failure Out of operation, Impulsive force

1 2 3 1 3 3 c. Safety of structure in excitation Excitmg force, Simultaneous

effects of earthquake and wind,

Operation limit

1 1 1 1 1 1 d. Emergency interruption mechanism T"» 1

* 1 ' 1' A. J? Reliability of emergency

interruption mechanism, Effect of

sudden stop on structure

(3)Safety Evaluation of Response-control (To confirm that a responseDevice control device is in normal

operation)

2 2 3 2 3 3 11. Composition and Specifications of

Response-control Device

a. System Composition Block diagram, Backup system

b. Components and specifications Size, Weight, Input and output,

(Sensing system, Control system, Accuracy, Capacity, Computing

Drive system) performance, Environmental

conditions

2 2 3 2 3 3 12. Control Design of Active Control

Device

a. Sensing of quantity of state b. Signal processing

c. Control theory

d. Driving program

589
Table 1.2 Safety Evaluation Items for Structural Response-control Buildings (Continued)

Class of Responsecontrol Building Evaluation items Remarks

A R r D F F

i

i Z

i

! -2

.j 13 Oneration Program of Active Control

Device

a. Period, conditions and method of Program of start and ston operation

14. Performance Validation and Durability

of Response-control Device

9z 9z

1Z j

'X ,J a Fabrication and nroduct insnection Oualitv V^/ LA CIA 1 L V and CCA A Vl standards O ICAA1 VJ CIA VlO ofUl device vi-W V procedures material

9Z

0i-, j1Z J h feneration test and nerformance test Oneration of device Performance

of response-control building

1J '1L j . 9J J cWe JiS Durabilitv LAA LALyAAAl T Environmental JJ Jk. 1 T LI XJ'liillVlllUl conditions VV/1 l\JllllL/llUi| durable VJU1 UL/lV vear and dnrahilirv of sensor

(~*PT V_^JL KJI

, drivintr vll IV lllli device nower UUWtl

source, and others

1

! 1

J z

"J J .3 15 Maintenance Inspection of Resnonsecontrol VAA1 i LA V./A Device LSW1W

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a. Performance validation test data

Notes:

1) "0" denotes that evaluation equivalent to ordinary building is necessary.

"3" denotes that detailed evaluation as response-control building is necessary.

"2" denotes that evaluation as response-control building is necessary.

"1" denotes that evaluation as response-control building is not necessary except for special cases.

2) Earthquake and wind response analysis is for a structural response-control building including the

structure and response-control device. Degree of investigation for each item is different according to a balance between the potential performance of response-control device and the lateral load-carrying

capacity of structure.

590
Tsunami Disaster Caused by Hokkaido-Nansei-Oki Earthquake in 1993

by

Shunji Niwa, Takekazu Akagiri, Masakatsu Horino, Masaharu Tsuzawa

ABSTRACT

The earthquake was accompanied by big

tsunamis. Okushiri island, which is in the southwest of Hokkaido, was severely damaged by the

tsunamis with a loss of 201 persons. The maximum heights of the traces of the tsunami were

measured by using a parallax bar on aerial photographs. The naximum height of the tsunamis is 21m in the west coast of the island.

Key words: Okushiri island, maximum height of

the tsumamis, parallax bar, aerial photograph.

1 .Introduction

At 22:17 on July 12 in 1993, a big earthquake hit Okushiri Island which is located in

the southwest of Hokkaido, off the west coast of

the Oshima Peninsula , of which magnitude was

7.8. In Okushiri Island, very strong shock hit the

island, and subsequently, about 300 sites of failures were caused mainly along the coastline and

in the mountainous area. In Oshima Peninsula,

many collapses, liquefaction of soil, subsequent

disruptions of roads were caused in many sites. The earthquake was accompanied by big

tsunamis. (Tsunami is very strong and high wave

which is caused by a strong earthquake in the

sea.) It hit Okushiri Island and other areas and

caused intensive damages which included loss of many people and various kinds of damages in

the coastal areas in Japan, Korea, Russia and

other areas. In the case of Okushiri Island, tsunamis hit the island only several minutes after the

earthquake. And subsequently 201 persons were lost and many houses and buildings were

destroyed.

Geographical Survey Institute(GSI) took

aerial photographs to catch actual conditions of

damages in the southwestern coastal zone of

Hokkaido main land and Okushiri Island on July

1 3 and 14, and worked in the field from 15 to 20

of July and from 3 to 5 of August. Based on the

interpretation of these aerial photographs with

fieldcheck, the authors clarified actual conditions of tsunamis using traces of tsunamis,

mainly on the areal distribution of drowned area and the height of the tsunami. (Fig.2)

In this report, we will report the general

conceptions of the damage and heights of tsunamis.

2. Outline of the Hokkaido Nansei-oki Earthquake

According to the Meteorological Agency, the

hypocenter of the earthquake is 34km deep, 42°

47 ' N, 1 39° 12 ' E. It is located at about 70km off

to the northwest of Okushiri Island. According

to the distribution of hypocenter of the aftershocks, it is assumed that the fault plane is tilted westside down. Okushiri Island subsided about

20 cm to 80 cm by the earthquake. We clarified

Geographical Survey Institute, Ministry of

Construction, Kitasato 1 , Tsukuba-shijbarakiken, 305 Japan

591
the quantitative value of subsidence using

change of landscapes, which are subsidence of

rocks along the coastline, change of shore protection, the change of the vertical distribution of

small life in tidal zones in comparison with before and after the earthquake based on

photointerpretation, photo-grametric way and

fieldcheck.

The source area of the tsunami is distributed

in the area about 1 50 km from north to south in

the area of south-west off of Hokkaido. Based

on the initial motion, this earthquake caused the

submergence in the east, and the upheaval in the

west. Several models of the fault with two

faults planes were proposed and mostly that the

fault planes are westside down.

3. Outline of Okushiri Island

Okushiri Island is an island which is 22km

from north to south, 1 3km from west to east. In

the island, the highest peak is Mt.Kamui with

585m above the sea level and various terraces can be seen from the coast to the top of the

mountainous area. Terraces develop in the southern part, east coast and north edge of the island,

but few in the west coast. The central area of the

island is remarkably dissected. The terraces,

which had been formed about one hundred and

twenty thousand years ago, have heights of 70m

to 80m in the south, 1 20m to 1 30m in the north.

The island repeatedly raised up by tectonic upheaval by earthquakes and the island has been

being formed. But this time, the island subsided.

Villages are distributed only along the east

coast and both north and south ends of the island, but very few along the west coast and in

the mountainous area.

4.Damage Caused by Tsunami

As of 1 1 of August, there were loss of

life:200 persons, missing:34. Among them, 169

loss, 32 missing at Okushiri Island. Most of

them were by tsunami. As for damage of houses:

complete destruction 558, partial destruction

247, drown, inundation and slight destruction

3,348 (as of 27 of July). Most of houses were

destroyed or washed away by tsunami at Aonae,

Hatsumatsumae in the south of the island and

Inaho in the north end.

Fire was caused after the hit of tsunami and

subsequently a few hundreds of houses burned

down at Aonae. The fire started in a few sites,

but it spread to the broad area because of a few

promoting factors: l)anxiety about following

tsunami waves, 2)accumulation of destroyed

houses, damaged cars and logs and wastes from

the sea by tsunami waves were seen on town

roads. These reasons interrupted fire service

team to enter and work in the area. Subsequently fire spread to the broad area. Tsunami

waves also caused damage of port facilities,

destruction of ships, washed ashore ships and

many other things were burned and washed

ashore or washed away to the open sea.

5. Drowned Area by Tsunami Waves and the

Distribution of Maximum Flooded Level (Flood

Mark)

It is clearly and easily found to distinguish

the drowned area and the maximum flooded

level of tsunami waves based on color change

of grass-covered area, washing away and destruction of houses, various kinds of wastes,

driftwood and others. Vegetation color in the

drowned area was changed from green to brown

and the boundary was very clear. Salt in sea water which overflowed let grass dry and changed

color. And the distribution of driftwood and

592
other kinds of wastes were seen in the same

zone. This boundary was seen clearly in the

southwest and north end of the island, but it was

not so clear in the northwest and east coasts, be- cause the heights of tsunami waves were not so high and very few area is covered by vegetation

in the low area along the coast line, and sometimes, such area consists of rock.

6.Measurement of Tsunami Waves using Aerial

Photographs

We used a parallax bar on the aerial photographs to measure heights of tsunami. And we

checked the value at several sites in

the field. The method using a parallax bar is to

get relative height between two points and a parallax bar is a tool to measure parallax of the same point on two leaves of aerial photographs.

Using this tool, it is possible to measure parallax

between two points. Based on the parallax difference, it is possible to calculate relative

heights between two points. The heights of tsu- nami waves must be calculated from the sea

level, for the reason, it is possible to measure maximum heights of tsunami waves easily

based on the relative heights of maximum flood

mark on aerial photographs. Thus, we could get

the distribution map of tsunami heights which

went upstream in the land area(Fig.2).

The maximum value is 21m in the west

coast, but 3-7m in the east coast. Tsunami damage was intensive at villages at low sites in the

areas of north and south ends.

In aerial photographs, traces of tsunami waves were recorded very clearly on the aerial photographs taken on July 13 and 14. The

record includes drowned area and heights

which tsunami reached upstream in the land

area. And we measured the heights of tsunami

waves using aerial photographs as a part of ur- gent survey in a very short period and officially

announced on August 2 as a part of "The urgent

report of the Hokkaido nansei oki earthquake".

The method to measure relative heights is never special one, rather very fundamental

method of photogrammetry, but we have not

seen an example to have measured the area and

heights of tsunami waves, then, here we will report the result.

7.Interpretation of Aerial Photographs

Information of aerial photographs consist

of l)location, 2)tone, 3)pattern, 4)texture,

5)color, 6)form. This time, we could use color

aerial photographs. Then, it was possible to use color information enough for interpretation.

Naturally, various information of actual conditions of damage was very important, that is, destruction of houses, erosion, accumulation of

wastes change of landscapes along the coastline. Especially, color change of grass-cover

land. Based on the preliminary interpretation

and measured map, the survey team made sure whether or not the map was correct. In most

cases, the result was very good with very few exceptions.

Aerial photographs used are as follows: 1

)

the whole area of Okushiri Island., taken by

GSI, flight height is 2,600m, color, 2) the

coastal zone, taken by Kokusai-kogyo Co. flight

height is 1 ,275m. color.

8. Method

Procedure is shown in the followings: ^Interpretation of damaged area using aerial photographs stereoscopically. The boundary of

damaged area is marked on the photographs. At

this stage, every site, one stereo pair of photo593
graphs was prepared. 2)using each pair of aerial photographs on which the interpretation was

marked, and the relative heights (area and maximum heights of tsunami waves) were calculated based on parallax differences which were measured using a parallax bar. The values were

not the same and values above 40 sites were measured using aerial photographs and were made sure in the field. The result of interpretation and measurement was almost the same as

the values which were measured in the field. The

result is shown in the Table 1

.

9.Procedure

1 preparation of two leaves of aerial photographs on which drowned area is shown. Plotting principal point of each photo.

2)Transferring principal point to each other

and to measure the air base length between principal point and transferred principal point on

each photo.

3) Set the stereo pair of photos. Two principal

points and two transferred principal points must

be set in straight to be able to get stereoscopically using a mirror stereo scope.

4)On the pair of photos, using a parallax bar, measure value of the same point A (coast line) by

adjusting, measureing mark of the parallax bar. Adjusting point B and measure value. If the val- ues of A and B are the same, A and B are the same heights. If A and B are not the same, the

ground height of A and B are different. This dif- ference is parallax difference. Based on this

value, the relative height is calculated.

5)The relative heights (maximum height of

tsunami) is calculated as follows:

dh=H x dp

b+dp

dh : relative height

H rflight height

b :base length

dprparallax difference

6)Repeat measurement of parallax differ- ences at least several times. And average the

result except maximum and minimum to de- crease error. This time, because the parallax

difference is relatively small to base length, the

parallax difference in the denominator of the formula can be neglected.

This method can be available to get uniform

quality of the result in a short period. If it is impossible to visit the sites, the result using

aerial photographs is almost the same as the result from fieldcheck.

It is possible to distinguish distance by eyes

when intersection angle is above 20"

(Kimoto: 1971). This shows that 1.3m is the

minimum resolution when the intersection angle

above 20" is 0.09mm in the case of 25cm in

clear vision of the photos which were taken at

1 ,275m in flight height. For the reason, the measured values were counted fractions of 5 and as

a unit and cut away the rest. Objectives to research using photos must be

shown in the photos. It is pretty difficult to interpret and measure the sites which are in shadow, in the deep valley bottoms, in a deep

transparent sea. To get better results, surveyors must use the

center of the photos as possible as they can. And also to get better results of measurement,

aerial photographs should be taken along the

594
coast line. It means that decision of flight courses is pretty important.

10. Conclusion

As of the maximum height of tsunami waves, some researchers reported that the maximum

value was over 30m, but based on our research,

the maximum value was 21m. Usually it is pretty

difficult to measure in the shadow, in valley

deep floor, but the difference was too large.

It is possible to clarify the distribution of

tsunami waves using aerial photographs with

parallax bar. This method is very fundamental

method in photogrammetry. Such a fundamental

method is available to catch actual conditions of

tsunami waves.

1 1 .Acknowledgment

The authors are full of gratitude to Mr.Haruo Fujimaki, Mr.Minoru Hoshino,

Mr.Yohta Kumaki and other members of the

team for the urgent research on the Hokkaido

Nansei-oki Earthquake.

1 2. Reference

Ujihisa KIMOTO(1971):Pracrica/ photogrammetry. Sankaido, Tokyo, p267, in Japanese. The Emergent Research Group for the

Hokkaido Nansei-oki Earthquake

Hazard( 1 993):The emergent survey report of the

geomorphological hazard induced by the

Hokkaido Nansei-oki Earthquake of 1993.

Kokudochiriin jihou, GSI, 20-27, in Japanese.

595
Table 1. Calculation of the height that tsunami reached

1 0

1 1

1 2

1 3

1 4

1 5

1 6

1 7

1 8

1 9

2 0

2 1

2 2

2 3

2 4

2 5

2 6

b

8 3

8 2

8 2

8 3

7 8

8 0

8 2

8 7

7 9

8 7

8 6

7 7

9 8

8 2

9 8

9 9

8 6

84

8 4

8 4

8 4

8 9

35.28

3 2.33

32.99

31.19

34. 92

33.3

34.39

29.56

28. 75

29.35

31.46

33.31

31.82

31.05

21.55

14. 40

21. 22

20.89

25.91

16.70

3. 99

13.30

16.57

15.90

3. 63

8. 12

35.57

32.63

33.27

31.47

35.02

33.46

34. 59

29.78

28. 98

29.74

31.89

33.65

32.31

31.41

21.77

14. 61

21.63

21.14

26.27

18.01

5.36

14.30

17.88

16.92

4. 42

8.83

d P

0. 29

0. 3

0. 28

0. 28

0. 10

0. 1

6

0. 2

0. 22

0. 23

0.3 9 0.4 3

0.3 4 0.49

0.3 6 0.2 2

0. 2

1

0. 41

0. 25

0.3 6

1.3 1

1.3 7 1.00

0. 84

1. 02

0. 79

0. 7

1

d h

6. 9

7. 2

6. 8

6. 7

2. 5

3. 9

5. 0

5. 8

8. 9

9. 9

8. 7 11.0

. 1

4. 4

5. 1

8. 3

5. 0

8. 3 19.9

20. 8 15.2

10. 5

14. 8 11.4

10. 2

Airphotos

> GSI

Kokusai Kogyo

Co. Ltd. **

No. :Locality numbers. b :The length of baselines, (mm) p s :Parallax at

the sea level, (mm) p t :Parallax at the highest level that tsunami

reached. (mm) d p :Parallax difference, (mm) d h :The highest level

that tsunami reached, (m) A i r pho t o s : A i r pho t o s used for measurement.

* Geographical Survey Institute, July 14 1993, 1:13,000

** July 13 1993, 1:8, 000

596
597
g. 2 Distribution of the height of the tsunami

598
5^9

The Dam Safety Program

by

William S. Bivins* and Harold W. Andress*

ABSTRACT

The United States does not have a legislated National

Dam Safety Program. Devastating dam failures in the 1970's raised public concern but surprisingly did not produce legislation to mitigate the loss of lives and property. In 1980, the Federal Emergency

Management Agency (FEMA) initiated a program to pull together concerned professionals. It was an

attempt to substitute a concensus-building process for the missing legislation (with the attendant funding).

The process produced The Partnership, a term

defined as the intertwining of individual interests in the dam safety community to enhance the safety of

dams. The Partnership in turn produced the leadership, technical assistance, and public awareness

to mitigate the risks from the nation's dams.

KEYWORDS: concensus-building, dams, dam

safety, guidelines, inventory, training.

1. INTRODUCTION

Currently there is no National Dam Safety Program.

There is a need for one and the need will continue.

In November, 1982, the National Research Council

undertook a study for the Federal Emergency

Management Agency (FEMA) and published a report

titled, Safety of Nonfederal Dams - A Review of the Federal Role. One of the conclusions of that study was "[bjecause of inadequate state supervision and

the magnitude of the consequences of widespread

dam failures, a National perspective on dam safety is needed. The problems will not be solved entirely

without a Nationwide perspective or the impetus that a broad-based group can offer."

However, no legislation to create such a National

perspective has been forthcoming. Each federal agency had autonomous dam safety programs. Some

states also had programs but they varied widely.

In 1980, the Federal Emergency Management Agency

(FEMA) began a process to substitute consensus

building for legislative mandate. This paper

describes the evolution of The Partnership — an

intertwining of the interests of the dam safety

communities of the federal government, of states, and

of the private sector. While legislation is still needed, the FEMA initiative has resulted in the

partial mitigation of risks from the nations dams.

2. BACKGROUND

The Federal dam safety program had its beginning in disasters of the 1970's. It was after the failure of

Buffalo Creek, Teton, and Kelly Barnes dams that Federal efforts in dam safety began in earnest.

2.1 Buffalo Creek

Saturday morning, February, 1972, a mine tailings

dam failed at Buffalo Creek, West Virginia, killing

125 people. The Buffalo Creek Dam was neither designed nor built to accepted engineering standards. Nor was it operated, maintained, inspected or

regulated in accordance with sound practices of dam

safety. This disaster created public concern about the

safety of the Nation's dams which in turn stimulated

Congress to pass the National Dam Inspection Act

(Public Law 92-367). The Act authorized an

inventory and an inspection of the Nation's dams.

The inventory began, but the inspection had to await

another disaster.

2.2 Teton

The Teton Dam, near Newdale, Idaho, failed June 5, 1976, killing 14 people. This induced the federal government to take a hard look at existing dam safety programs, to assess what they accomplished and to define what they needed to be. The assessment led

to the development of the Federal Guidelinesfor Dam

Safety.

2.3 Kelly Barnes

It was a rainy night in Toccoa Falls, Georgia,

November 6, 1977, when the Kelly Barnes Dam

failed without warning; killing 39 people, mostly

* Federal Emergency Management Agency

Washington, DC 20472

601
Bible college students. This disaster provided the impetus to fund the inspection portion of P.L. 92-

367. The U.S. Army Corps of Engineers began a program to inspect 8818 high hazard potential nonfederal dams. The inspection revealed that 2925

dams were unsafe.

In October 1979, President Carter directed the agencies to implement the Federal Guidelines for

Dam Safety and report to the newly created Federal

Emergency Management Agency (FEMA). The

Federal program for dam safety was born.

3. THE PROGRAM

The goal of the dam safety program is to identify and

mitigate the risk of dams to protect the lives and

property of the public. This is accomplished by

establishing and enforcing acceptable design,

construction, operation and maintenance standards for dams. The goal is further elaborated upon in the FEMA dam safety doctrine:

"Dams must be designed, built, operated and

maintained safe. Ensuring the safety of dams is not

a passive activity. The responsibility for protecting

lives and property never ends. New approaches and

policies must be developed, implemented and

evaluated. New players must be continually recruited and drawn into the group of involved participants.

New programs must be promoted; weak programs

must be revitalized. Leadership, awareness,

dedication and action must be advocated and

repeatedly fostered. Opportunities must not be

overlooked.

The dam safety program objectives are to improve

federal dam safety through implementation of the Federal Guidelinesfor Dam Safety and to foster nonfederal dam safety by assisting States, local governments and the private sector in developing and

implementing effective programs. The objectives are expressed in the following vision statement:

"FEMA will accomplish its operational objectives for national dam safety through the creation of quality

policies, procedures and products to enhance personal

safety, protect property and benefit the dam safety

community. By becoming pacesetters as resource stewards and leadership providers, through

commitment to high quality, reliable service and

effective management, FEMA will achieve results beyond the dam safety community and public's

expectations.

4. FEMA'S ROLE

FEMA's role in National dam safety is unique.

FEMA neither owns dams nor has regulatory

responsibility for dams; but, FEMA coordinates

hazard mitigation dam safety activities. FEMA's role includes activities in the federal and non-federal

sectors.

4.1 Federal:

The federal role embodies all the activities undertaken

by federal agencies to safety build, operate and

maintain dams. FEMA coordinates the activities of

the federal agencies to implement the Federal

Guidelines for Dam Safety through the Interagency

Committee on Dam Safety (ICODS). FEMA is the chair for ICODS and its administrator.

4.2 Non-Federal:

Non-federal program activities involve the practices

undertaken by State and local governments and the private sector to design, build, operate and maintain

safely the dams they own and/or regulate. FEMA

coordinates non-federal program activities primarily

with the Association of State Dam Safety Officials (ASDSO). FEMA's role in this sector is to foster

state efforts to establish effective dam safety

programs patterned after the Model State Dam Safety

Program and to provide technical assistance and

public awareness support.

4.3 Partnership

National dam safety goals and objectives are pursued

in partnership with Federal, non-Federal and private

sector participants. This consortium embodies the power of partnership:

Partnership is essential to National dam safety;

partners enhance capability, minimize duplication and

maximize efficient use of resources. Success is directly related to success of the Federal and nonFederal partnership. Federal agencies working

together form one element of the partnership while

State and local governments and the private sector working together form the other. The Federal and

non-Federal elements working together form the

'ultimate' partnership. Together these partners make

the National Dam Safety Program truly National.

602
4.3.1 Federal Partners

4.3.1.1 Interagency Committee on Dam Safety

ICODS exemplifies the Federal partnership. ICODS

is a potent force. It is the only Federal forum where

Federal agencies initiate cooperative efforts and

offerings of talent and resources to meet National

dam safety needs. FEMA chairs ICODS which consists of

representatives from; the Departments ofAgriculture,

Defense, Energy, Interior, Labor and State (International Boundary and Water Commission), the Federal Energy Regulatory Commission, Nuclear

Regulatory Commission and the Tennessee Valley

Authority.

ICODS was established in 1980 and one of its function is to assist FEMA in fulfilling its National

dam safety responsibilities. It continues to excel in

its role of providing an arena where Federal dam

safety issues are identified, discussed and resolved,

and where progress and contributions are made in cooperative efforts to resolve non-Federal dam safety

issues.

4.3.2 Non-Federal Partners:

The year 1993 marked the completion of a successful decade of partnership between the Federal and nonFederal dam safety communities; between ICODS,

ASDSO and the United States Committee on Large

Dams (USCOLD).

4.3.2.1 Association of State Dam Safety Officials

ASDSO brings together the partnership of the States and the private sector. It provides the non-Federal

voice to address National dam safety issues. ASDSO, organized in 1984, continues to carry out

far reaching projects for which it has become known

and which have made it one of the leading

organizations dealing with dam safety issues,

nationally and internationally. ASDSO produces

products which benefit the entire dam safety

community while maintaining its primary focus of

supporting State dam safety programs and

maintaining intergovernmental relations.

ASDSO continues to prosper as the word of its good

work continues to spread. Membership from other

countries has steadily increased. Countries

represented include; Australia, Brazil, Canada, India,

Iran, Italy, Japan, South Africa, South Korea, and

Turkey. Total ASDSO membership should reach

1000 in 1994, up from 379 in 1988.

4.3.2.2 United States Committee on Large Dams

USCOLD is the United States member of the

International Commission on Large Dams (ICOLD),

an international organization composed of nearly 80

countries. USCOLD is a nation-wide professional

organization dedicated to advancing the technology of

dam design, construction, operation and maintenance,

and to promoting awareness of the role of darns in the beneficial development of the nation's water

resources. USCOLD members come from many professional

disciplines, including Civil, Mechanical, and

Electrical Engineering; Environmental Sciences,

Social Sciences; Finance; Economics and Law. They

are contractors; consulting engineers; representatives

of local, State and Federal government agencies;

academicians; managers and owners of water

resource projects.

FEMA is represented on USCOLD's Committee on

Public Relations. The role of the committee is to communicate about dams to the public and is closely

aligned with FEMA public awareness activities in dam safety. USCOLD adds the third leg to the triad

- the partnership of Federal sector, non-Federal

public sector, and private sector - which broadens the base of support for national dam safety.

4.3.2.3 "Ultimate" Partnership

Federal and non-Federal partners working together

form the "ultimate" partnership by consolidating their efforts to meet the goals and objectives of National

dam safety. FEMA coordinates these efforts.

5. FEMA ACTIVITIES

FEMA conducts its dam safety activities primarily in three areas:

5.1 Leadership

FEMA provides strong National leadership by

chairing ICODS and coordinating Federal actions in non-Federal dam safety.

An important leadership project in dam safety today

is maintaining a current National Inventory of Dams

(NID). In 1986 FEMA funded ASDSO to determine

the feasibility of updating the NID. The update was

603
feasible and the states agreed to participate.

However, the states did not have adequate funding

for update. In November 1986, Congress passed the Water Resources Act of 1986 which authorized U.S.

Army Corps of Engineers to update the inventory.

Funds were appropriated in fiscal year 1988. On

March 16, 1989 FEMA and the U.S. Army Corps of

Engineers signed an agreement whereby FEMA

assumed responsibility for a program using the appropriated funds. The update program was

completed in 1992. The inventory, available in CD

ROM form, is updated annually.

As previously mentioned, FEMA chairs ICODS

which meets quarterly to further dam safety goals.

Workshops and meetings are sponsored to review,

discuss, and debate current issues relative to national dam safety. An example is the 1990 Probable

Maximum Precipitation/Probable Maximum Flood

(PMP/PMF) workshop held to discuss the formulation of PMP/PMF guidelines. PMP and PMF

estimates have taken on greater importance in liability

and dam construction and modification costs. Since no uniform standards existed for estimating the PMP

and PMF, differences of opinion concerning these estimates had caused considerable confusion.

Uniform standards allay this confusion and allow

practitioners to develop estimates that are both widely

acceptable and reproducible.

5.2 Technical Assistance

Technical assistance is delivered through the exchange of dam safety knowledge, skills, experiences and resources between the Federal and

non-Federal sectors.

One of the most successful technical assistance programs is the Training Aids for Dam Safety

(TADS). TADS was developed as an ICODS

project, with non-Federal participation and adept lead agency management by the Bureau of Reclamation.

TADS consists of 21 training modules, text and

video, grouped in three components: 1) The

Inspection component, designed to teach regulators

how to conduct a dam safety inspection; 2) The

Awareness component which emphasized dam safety

mitigation; and 3) The Analysis component, which

provides evaluation techniques for dam safety

program managers, owners/operators and experienced

engineers. FEMA distributed TADS materials to the

States free of charge. They in turn can train their own engineers and regulators in dam safety.

Despite the efforts to minimize the risk, dams

occasionally fail. When this happens it is very

important that dam owners and operators have a plan

to deal with the emergency. For this reason a State requirement for an emergency action plan for each

high hazard dam is one of the eight essential elements

of an effective dam safety program. The plan is absolutely essential for dams which have a high

hazard potential for failure and highly recommended

for dams having a significant hazard potential. The

Federal Energy Regulatory Commission (FERC)

developed a training program to teach hydroelectric

licensees how to develop and exercise emergency

action plans. FEMA is cooperating with FERC to adapt this course for all dam owners/operators.

5.3 Public Awareness

This activity includes proactive and aggressive public awareness enterprises to bring the dam safety

message to the general public, emergency

management officials, Federal, State and local

officials, and the private sector dam safety

community.

Most people know very little about dams, even if they live or work near one. Therefore, knowledge

about how dams operate and what to do in the event of an emergency is important for everyone. It is through public awareness that we raise the level of

consciousness of dam safety in both the Federal and

non-Federal sectors.

Public awareness takes many forms. It can be as straightforward as a State Dam Safety Public

Awareness Workshops or the biennial National Dam

Safety Program Progress Report to the President. FEMA publishes many materials emphasizing dam

safety; most written with the assistance of other

agencies. These publications are designed for diverse

audiences such as dam owners/operators, State and

local officials, and the general public.

FEMA also disseminates other types of information

to the public and Federal agencies. For example,

several years ago dam owners in the Great Lakes

region became aware of a new crustacean which

clogged water intake valves, bred prolifically, and was extremely difficult to eradicate. This crustacean,

the Zebra mussel, found its way to the United States from the Caspian Sea via shipping in the St. Lawrence Seaway. The mussels are now spreading

throughout the waterways of the Midwest, East and

Southern United States. FEMA, through ICODS, has

604
created a clearinghouse for information on the Zebra

mussel so dam owners can keep abreast of new

developments on how to eradicate these creatures or impair their ability to attach themselves to critical structures and associated equipment.

All of FEMA's public awareness programs have the proactive goal of disseminating information to the widest possible audience in order to save lives,

mitigate property damage and conserve resources.

6. CONCLUSIONS

The concensus-building process, tentatively embarked

upon in 1980, worked. A de facto National Dam

Safety Program does exist. It has produced

phenomenal results.

The National Inventory of Dams, the Training Aids

for Dam Safety, the development of numerous

guidelines and the enhancement of public awareness,

all evolved from the development of The Partnership.

Legislation, with adequate funding would have been

desirable. The results produced in 14 years would

have been available sooner. Loss of lives and

property damage may have been avoided. However,

funding was not available.

The Partnership of literally thousands of dedicated

dam safety professionals from federal and state governments and the private sector proved an

acceptable substitute.

6. REFERENCES

6.1 Safety of Nonfederal Dams - A Review of the Federal Role, National Research Council for FEMA, 1982.

6.2 Federal Guidelines for Dam Safety, Federal

Coordinating Council for Science, Engineering

and Technology, (FEMA-93) 1979.

6.3 Model State Dam Safety Program, Association

of State Dam Safety Officials for FEMA, 1987.

6.4 Proceedings of Probable Maximum

Precipitation and Probable Maximum Flood

Workshop, FEMA, 1990.

6.5 Emergency Action Planning Guidelines for

Dams, Emergency Action Planning

Subcommittee, ICODS, (FEMA 64) 1985.

6.6 Federal Guidelines for Earthquake Analyses

and Design ofDams, Design Earthquake Task

Group, ICODS (FEMA-65) 1985.

6.7 Federal Guidelines for Selecting and

Accommodating Inflow Design Floods for

Dams, Inflow Design Floods Working Group,

ICODS, (FEMA-94) 1986.

605

Multi-Hazard Mitigation for Non-Engineered Structures

by

Michael Mahoney*

ABSTRACT

The Federal Emergency Management Agency

(FEMA) is the Federal agency responsible for

coordinating all of the Federal government's

activities relating to disasters. One of this agency's primary goals is to reduce, or mitigate,

the nation's losses that result from natural

hazards, such as earthquake, high winds,

flooding and fire. In order to accomplish this goal, FEMA is committed to encouraging

building practices that address these different

hazards and minimize the damage that can

result. To accomplish this, FEMA has worked

with the design and engineering communities

and the nation's three model code organizations

to encourage the use of loss reduction design

standards. While this has generally been

effective in their own way, there are two reasons why they could be more successful.

One reason is that each approach only dealt with

its own hazard. The development of design

criteria has historically only examined the impact on the hazard being addressed, and did not examine what the impact was on other hazards.

There is a growing awareness that mitigation

must look at all hazards; that coordination is needed to avoid conflicts where action taken to reduce the threat of damage by one hazard may

increase the threat of damage from another.

A second reason is the building codes

themselves and how they are used. Model

building codes usually work well with an

engineered building, where the engineer can

translate the performance requirements into

construction requirements, or plans. The

problem is with marginally- or non-engineered

buildings, where that interpretation is usually left up to the builder or contractor. The success of

that building in resisting the forces generated by

the different natural hazards then becomes

dependent upon their level of expertise and

attention to detail. As shown by several recent

disasters, such as the Northridge earthquake and

Hurricane Andrew, this is where the process

often breaks down.

To address the need for technical guidance for

non-engineered buildings, and to do so using a multi-hazard approach, FEMA has recently

undertaken two projects that will be of

significant benefit. However, the coordination

of efforts between the different hazards may

require an independent, non-governmental body

that is capable of coordinating the present

standards writing organizations as well as the

various outside interests, such as the architects,

engineers, materials interests, contractors and

other similar groups.

1. INTRODUCTION

The Federal Emergency Management Agency

(FEMA) is the Federal agency responsible for

coordinating all of the Federal government's

activities relating to disasters. One of this agency's primary goals is to reduce, or mitigate,

the nation's losses that result from natural

hazards, such as earthquake, high winds,

flooding and fire. In order to accomplish this goal, FEMA is committed to encouraging

building practices that address these different

hazards and minimize the damage that can

result.

To help encourage building practices that can

reduce the threat presented by different natural

hazards, FEMA has worked with the design and

engineering communities and the nation's three

model code organizations to encourage the use

of loss reduction design standards. This effort has taken several forms.

* Federal Emergency Management Agency

Washington, DC 20472

607
To address the seismic hazard, the National

Earthquake Hazards Reduction Program

(NEHRP), which is made up of four Federal

agencies (FEMA, National Institute of Standards

and Technology, National Science Foundation,

and the U.S. Geological Survey) worked with an

outside organization, the Building Seismic safety

Council (BSSC) to develop a resource document

for use by all regulatory organizations. This

document, the NEHRP Recommended Provisions

for the Development of Seismic Regulations for

New Buildings, was developed using a consensus

procedure, and has now been either adopted or

utilized by all three of the model code

organizations.

To address the flood hazard, FEMA, which is responsible for administering the National Flood

Insurance Program (NFIP), started with design

criteria based on the legislation that authorized

the program. We then funded the development

and publication of technical guidance materials

based on the design criteria, and then worked

directly with the model code organizations to have this material incorporated into the nation's

model building codes.

With regard to the wind and fire hazards, there are currently no specific government mitigation

program that addresses these hazards. This is primarily because these hazards have generally

been addressed by non-governmental concerns, such as the American Society of Civil Engineers

and their ASCE 7 standard for wind and the

National Fire Protection Association for fire. Most government agencies accept and use the current model code requirements for these

hazards.

While all of these approaches have generally

been effective in their own way, there are several reasons why they could be more

successful.

2. MULTI-HAZARD APPROACH

One of the reasons why the mitigation of

different natural hazards has not been completely

effective is that, until now, each approach only

dealt with its own hazard. The development of

design criteria, from initial research up to actual

code language, has historically only examined

the impact on the hazard being addressed, and

did not examine what the impact was on other

hazards. There is a growing awareness that mitigation must look at all hazards; that coordination is needed to avoid conflicts where

action taken to reduce the threat of damage by

one hazard may increase the threat of damage

from another.

One example of a potential conflict between

hazards is whether seismic loads acting on a

structure that has been built on an raised

foundation to elevate the building above flood

levels would result in increased damage.

Seismic loads could cause an elevated foundation

to act as a moment arm, and greatly increase the

shaking of the building. However, if such a

structure and its elevated foundation were

properly constructed and reinforced to resist coastal wind forces, the additional steps needed

to provide adequate seismic resistance may well

only consist of a strengthened pile to beam

connection or the addition of knee bracing, both

of which are already common coastal

construction techniques.

A second example is the issue of whether

protecting a structure against flooding by

elevating it above flood levels increases its potential of wind damage. In this case, elevating a structure does reduce the shielding

effects of vegetation and other buildings.

However, the ANSI Standard and the building

codes all treat wind loadings as the same up to a height of 30 feet, so a building properly

constructed to withstand wind forces should

generally do so whether it is on the ground or

elevated above flood levels.

The model codes address this coordination issue

every year during their annual code change

process, when staff members review every

change to avoid such conflicts. However, this is not the most effective way to address this problem, since each code may ultimately solve

the problem in a different manner. It would be more effective to address this problem during

the development of these individual standards.

608
FEMA has also attempted to deal with this issue

in two ways. With regard to the NEHRP

Recommended Provisions, this was avoided by

using a consensus process, which incorporates

such a review. Since the flood design standards

of the NFIP are not consensus-backed, FEMA

recently had a "Code Compatibility Report"

prepared. This review also looked at other

natural hazards as they relate to flood, and

resulted in a series of recommendations that now

need to be resolved.

3. CODES AND NON-ENGINEERED

BUILDINGS

A second reason why the mitigation of natural

hazards on the built environment has not been

completely effective is the building codes

themselves and how they are used. Model

building codes usually need to be general enough

to be relevant to all types of construction. It is for this reason that they are normally

performance-based in nature, meaning that they

specify how a building should perform as opposed to how they should be built. This

generally works well with an engineered

building, where the engineer can translate the

performance requirements into construction

requirements, or plans. The problem is with

marginally- or non-engineered buildings, where

that interpretation is usually left up to the builder or contractor. The success of that building in

resisting the forces generated by the different

natural hazards then becomes dependent upon

their level of expertise and attention to detail. As shown by several recent disasters, such as the Northridge earthquake and Hurricane

Andrew, this is where the process often breaks

down.

It is for this reason that several reports prepared

after Hurricanes Hugo and Andrew

recommended that the local building code

examine the possibility of including prescriptive

design details within their code where possible.

For example, this was particularly relevant after Hurricane Andrew with regard to the design of

residential gable roof systems, where it was recommended that the South Florida Building

Code require the use of blocking and/or bracing

so that the roof trusses would be capable of

withstanding horizontal wind loads independent

of roof sheathing.

It is this need for prescriptive language that various standards groups are trying to address

with what have been referred to as "deemed to comply" standards. One example of such a standard is the "Standard for Hurricane Resistant

Residential Construction" (SSTD 10-93),

developed by the Southern Building Code

Congress International.

The biggest percentage of non-engineered

buildings are residential structures, and it is this class of building that has been responsible for a disproportionate amount of the losses FEMA has

had to handle in the most recent natural

disasters. It is for this reason that FEMA has

begun to target residential dwellings and the

home building industry. This problem was

never more apparent than in Hurricane Andrew,

where virtually every home in the 300 square

mile impacted area suffered some degree of

damage. However, even in events where

conditions have been well below the design

criteria, this class of building has often still suffered significant amounts of damage,

especially when construction techniques have

been shown to be less than adequate. What is particularly important from FEMA's point of

view is that, given the sheer numbers of this type of building, even minor damage represents a significant loss potential that needs to be

addressed by all levels of government if we are going to even begin to control the cost of

disasters.

However, FEMA has also observed that this does not have to be the case. In every natural

disaster we have investigated, there have been

examples of residential structures that have

successfully resisted design conditions with little significant damage. The residents of these

buildings suffered little loss to their homes or

possessions, and they were able to quickly

reoccupy their dwellings. The difference was

that in these cases, the builder not only properly

followed the local code, but also took the steps

needed to ensure that the critical connections

609
worked to allow the home to withstand these

loads.

While it is important that home construction

remain affordable, it is equally important that it follow the current codes and be capable of

surviving a moderate hurricane or earthquake.

It seems to us that the home building industry

would rather have homes that are still standing

shown on the evening news as opposed to what

we all saw after Hurricane Andrew.

4. GUIDANCE FOR NON-ENGINEERED

BUILDINGS

To address the need for technical guidance for

non-engineered buildings, and to do so using a multi-hazard approach, FEMA has recently

undertaken two projects that will be of

significant benefit.

The first project addresses residential structures

that are exposed to both flood and earthquake

hazards. As described above, to mitigate against

flood hazards, FEMA has developed a series of

flood resistant design criteria, one of which

requires that all new construction as well as all existing structures that have been substantially

damaged or improved, be elevated above

anticipated flood levels. In a seismic area, such an elevating foundation can increase seismic

loads on the entire structure.

To resolve the problem of how to elevate a flood-prone building in a manner that still complies with current seismic building codes,

FEMA has contracted for the development of a guidance document that will present basic

designs for a series of sample building

foundations that can be used to elevate a building above flood levels while being capable

of resisting seismic and wind loads. This effort

is being performed under FEMA's National

Earthquake Technical Assistance Contract

(NETAC).

The sample foundation designs provided in this publication will be in compliance with the

NEHRP Recommended Provisions as well as the

seismic requirements of all three of the model

building codes. The designs will be for areas of

both moderate and high seismicity and will be

applicable for different flood conditions. The

designs used will be representative of

foundations used throughout the country. The

publication will introduce the problem of

designing for both flood and earthquake

resistance and present the foundation designs and

supporting background information will be in a manner understandable by the non-engineer,

since the target audience is the homeowner and

their contractor.

The second project is the development of a

series of publications designed to present

construction guidance in a manner that can be

used by home builders and other non-engineers

to mitigate the effects of specific natural hazards

for one and two family dwellings. The

documents will also include prescriptive building

plans based on the state-of-the-art in design that could assist home builders and others in building

a non-engineered residential structure that complies with the model codes and is resistant to damage from natural hazards. While each

volume will address a specific hazard, related

material on other hazards will also be provided.

To date, FEMA has contracted for the

development of two documents. The first volume will be the "Home Builder's Guide to Seismic Resistant Construction" and will address

earthquake hazards. This volume will supersede

the one FEMA manual that already targets the

home builder; the "Home Builder's Guide to Earthquake Design" (FEMA-232). That manual was originally published by the Department of

Housing and Urban Development (HUD) in

1980 based on data from the San Fernando

earthquake and is out of date.

The second volume will be the "Home Builder's

Guide to Wind Resistant Construction" and will

address high wind hazards. Since high wind

areas are normally associated with coastal areas,

this second document will also address coastal

flood hazards.

The development process of both documents will

include a peer review is to ensure that the final

610
publication contains material that is technically correct and complete. Possible reviewers will

include home builders, contractors, engineers,

architects, building code officials, materials

interest groups , and knowledgeable homeowners

.

Such an outside review will serve as a "reality

check" for the final product.

Both documents will make it very clear that they

are not a substitute for using the local building

code or working with the local code enforcement

office, but instead are meant to be a resource to provide several techniques by which the home

builder can meet the local building code in an manner that addresses the risk presented by that particular hazard.

FEMA is very excited about these projects, and

believes that they will be of real benefit to both

the home builder community and the emergency

management community.

5. MULTI-HAZARD COORDINATION

The coordination of efforts between the different

hazards will require more than a few guidance

documents that take a coordinated look at several

different hazards. What is needed is an

independent, non-governmental body that is capable of coordinating the present standards

writing organizations as well as the various

outside interests, such as the architects,

engineers, materials interests, contractors and

other similar groups.

Such a body must recognize and utilize the

existence of existing consensus bodies that now

address specific hazards. Examples of these

groups include BSSC for seismic, NFPA for

fire, and the Wind Engineering Research

Council for wind. Such a body would not

replace these established groups, but would

instead act as a coordinating body. A possible

model for this body would be ASCE, which now

brings together these various groups in the

periodic update of its loads standard, ASCE 7.

In addition to coordinating the efforts of existing consensus bodies, such a body may need to be

capable of developing its own material if

necessary. If this body is to be capable of

developing or revising these standards, then it must itself be a recognized consensus body.

The role of the Federal government in such an

undertaking may be to initially fund a study to develop a plan of action. One goal of such a plan would be to identify a coordinating body

that could oversee all of these diverse efforts. Once such a body has been identified, the

Federal government could then fund the

operation of this body to examine and coordinate

the improvement of existing building standards,

technical resource materials and model code

language that address wind, flood and seismic

hazards.

6. CONCLUSION

If this country is going to reduce the cost of

natural disasters, we will need to improve the

way we build our structures, particularly those

that do not presently have the benefit of

engineering design. Just as there are different

hazards that must be addressed when designing

a building anywhere in the country, there are

different mitigation techniques that can be

employed to reduce losses. The key to reducing

losses is knowing how, and where, to properly

build.

If building practices as they now exist are going

to be revised to take into account the different

hazards that may be present, this nation will

need to coordinate the activities of the existing

groups that address these hazards. This will

allow us to resolve any incompatibilities and fill in the gaps that exist in the current standards and

codes.

611

Use of Structural Response Data from

Small Earthquakes and Aftershocks

by

Erdal §afak*

ABSTRACT

With some simplifying but reasonable as- sumptions, the seismic response of a structure to different sized earthquakes can analytically be correlated with the physical

parameters of the earthquakes. This cor- relation allows prediction of the structural

response to future earthquakes by scaling

the recorded responses from past earthquakes. In this paper, three methods are introduced to predict structural response

to large earthquakes by using recorded re- sponses from smaller earthquakes and af- tershocks. The accuracy of the methods

are tested by applying them to the data

from a highway overpass.

KEYWORDS: Earthquakes, vibrations,

structural dynamics, seismic response,

seismic scaling.

1. INTRODUCTION

Studying the correlation of ground motions from co-located large and small

earthquakes has been a popular research

subject in seismology. Methods to simulate large earthquakes by summing smaller

earthquakes have been developed (e.g.,

Hartzell, 1978; Mueller, 1985; Wennerberg, 1990). The limitations of such simulations and the constraints of the summation are discussed in Joyner and Boore

(1986). Similar simulation techniques can also be developed for structural re- sponse. The problem is somewhat simpler

for structures, because the structure acts

as a narrow-band filter, thereby eliminating the need for matching the entire fre- quency band in the simulation.

This paper introduces three methods to

predict structural response to large earthquakes from the recordings of smaller

earthquakes. Although the basic philosophy is the same, the methods are less elaborate than those suggested by the seismologists, so that they can be adopted for

practical applications. The first method

is the simplest, and uses only the magnitudes of the earthquakes to scale the

recorded peak amplitudes from the small

earthquakes, whereas the second method

uses both magnitudes and the corner fre- quency of the small earthquakes for scaling. The third method treats small earthquakes as impulsive point sources, and

simulates the entire time history of the

response for the large earthquake.

The methods are applied to data from

a highway overpass in Rio Dell, California (Shakal et al., 1992). The data consist of two sets of recordings from two

co-located earthquakes. The first set is from the Cape Mendocino earthquake of

magnitude 5.1 occurring on 21 November

1986, and the second set is from the Petrolia earthquake of magnitude 6.4 occurring

on 25 April 1992. The epicentral distances

* Research Structural Engineer, U.S. Geological Survey, 922 National Center, Reston, VA 22029, USA.

613
and the depths are 26 km and 18 km, re- spectively, for the smaller earthquake, and

24 km and 15 km for the larger earthquake. The highway overpass is a twospan, cast-in-place concrete bridge, and

is instrumented with 20 acceleration sensors. A schematic of the bridge and the

instrumentation, taken from Shakal et al. (1992), is given in Fig. 1.

2. SCALING BY MAGNITUDES

We can write the FAS (Fourier amplitude

spectrum) of the response recorded by a

sensor in an instrumented structure as a product of a series of filters:

Ay(f) = C • Ax (f) Hp (f) H( (f) HAf)

(1)

where Ay (f) is the FAS of the recorded

response component y{t), and Ax (f) is the FAS of the accelerations in the source

region, known as the source spectrum.

Hp (f),Hi(f), and Hs (f) are the filters

representing the effects of the propagation

path, local geology, and the structure on

the recorded motion, respectively, and C

is a constant. The response y{t) can be

made to represent displacements, velocities, or accelerations by selecting the fil- ter H3 (f) accordingly. For a specified re- sponse component, and linear wave propagation and structural response, the filters

Hp (f), Hi(f), H3 (f), and the constant C

can all be assumed to be identical for all co-located (i.e., same source zone) earthquakes. Therefore, we can write the fol- lowing equation for the ratio of the FAS of

structural responses from two co-located

earthquakes:

AylU) =

Axl {f)

(2)

Ay3 {f) AM)

^

where the second subscripts / and s denote

the larger and the smaller earthquakes, re- spectively. Equation 2 shows that the ratio of the FAS of the responses from two

earthquakes is equal to the ratio of their source spectra. A simple model for the source spectra for accelerations is given

by the following equations (Brune, 1970):

Ax (f) = Mo • 1 ;\ 2 (3)

with

/c = 4.9xl06 /Jo(^)

1/3

(4)

where Mq is the seismic moment (in dynecm), fc is the corner frequency (in Hz), /?o is the wave velocity in the source region

(in km/s), and Act is the average stress drop (in bars) during the rupture. It can

be shown from Eq. 2 that for / > fc : Ax (f) » (2tt/c ) 2 Mo (5)

From random vibration theory (e.g., Lin,

1976), the mean-square value a

2

y of the

response y(t) is equal to the area under

the power spectral density, which can be

expressed as the square of FAS divided by

the effective duration; that is:

1 /'°° °» =

TJo

A2yif)df (6)

where T denotes the effective duration.

The effective duration can be approximated as the inverse of the corner fre- quency, i.e., T = l//c . Using this in

Eq. 6, and also Eq. 5 in Eq. 2, we can

derive the following equation for the ratio of the large to small earthquake RMS

(root-mean-square) response:

614
oy^Moj_ (fci_\ 2 - 5 (?)

cr ys M0a V/cs /

where the second subscripts / and s again

denote the larger and the smaller earthquakes, respectively. If we assume that

the average stress drops, Act, in both

earthquakes are approximately equal, we

can write from Eq. 4 that

Using this in Eq. 7 gives:

Instead of seismic moments, Eq. 9 can be expressed in terms of moment magnitudes, M, by using the definition of Hanks

and Kanamori (1979):

M= |log10 M0 -10.7 (10)

which gives

M^L - 10 i.5(M,-M5

) , u

s

Using Eq. 11 in Eq. 9 results in the fol- lowing equation for the ratio of RMS re- sponses:

ZlL _ 1 q(M,-m3 )/4 / 12)

where M\ and M3 are the moment magnitudes of the large and small earthquakes,

respectively. Random vibration theory

shows that the peak response can be obtained by multiplying the RMS response

with the peak factor, which is calculated

from the moments of the power spectral density function of the response (Lin,

1976). For typical structures (e.g., with

dominant first mode and first mode fre- quencies varying from 0.8 to 2.5 Hz) and

earthquakes (e.g., with effective durations

varying from two to ten seconds) the peak

factors do not vary significantly (§afak,

1988). Considering all the other uncertainties involved, we can assume for practical purposes that the ratio given by

Eq. 12 also represents the ratio of peak

responses.

To test the accuracy of the scaling relationship, Eq. 12 is applied to the large and

small earthquake data from the US101/

Painter Street overpass. It is assumed

that the difference of moment magnitudes

is equal to the difference of local magnitudes (i.e., Mi - Ma = 6.4 - 5.1 -

1.3). Figure 2 shows the comparison of

recorded and predicted large earthquake

peak accelerations and displacements for

the 20 sensors. The match is good for

accelerations but not for displacements.

The reason for this is that the scaling

factor is valid for frequencies above the corner frequency of the smaller earthquake. The displacement response is influenced by the frequencies from zero to

the resonant frequency of the structure,

whereas the acceleration response is influenced only by the frequencies near the

resonant frequency. Therefore, this model

is appropriate to estimate peak acceleration response, but not the displacement

response, and applicable only to structures whose dominant frequency is above

the corner frequency of the smaller earthquake.

3. RECURSIVE FILTERING

The simple method given above cannot be

used to scale the whole frequency band.

To account for the entire frequency band,

we use Eq. 3, rather than Eq. 5, for Ax (f).

Using Eq. 3, Eq. 2 becomes:

615
and

Ayl(f) M0/ l + if/fcsf

Ay3 {f) M0s l + (f/fcl ) 2 R(f)

(13)

where R(f) is the frequency- domain filter

relating the FAS of responses from large

and small earthquakes. The shape of R{f)

is similar to that of a low-pass filter and

can be converted into a discrete-time fil- ter, R{q), of the following form (§afak,

1994):

R(q) =

Po+Piq

1 — aq~

-l

(14)

where q~ l is the backward time-shift operator, defined as q~J y(t) — y(t—jA) with

A denoting the sampling interval. The

discrete-time filter given by Eq. 14 leads

to a recursive equation to calculate the

response time histories of the large earthquake:

yi(t) = ayi(t - A) + (l0 ys {t) + (3 iys (t - A)

(15)

The parameters {3q, fli, and a are calculated from the following equations:

f3Q =RM -a(RM -R1^) (16)

(17)

a =2 - cos(2tt/J9 A) - 03 - cos(2tt/p A)][1 - cos(/p A)]

(is;

fP = V2-

R

1/3

M

RM R

Rm Moi

Mto s (20)

M

The derivations of Eqs. 14-20 can be found

in §afak (1994). As the above equations

show, the filter relating the large and

small earthquakes depends on the ratio

of seismic moments, the same as in the

previous method, but not on the ratio of

the corner frequencies. The filter requires

the specific values of the corner frequencies because of Eq. 19. Since data from

the small earthquake are available, we can

easily calculate fcs . Again by assuming

an identical average stress drop for large

and small earthquakes, we can calculate

fd from Eq. 8, and also use moment magnitudes instead of seismic moments by utilizing Eq. 11. Thus, by knowing the moment magnitude and the corner frequency

of the small earthquake, and specifying

the moment magnitude of the large earthquake, we can determine the filter to predict the response for the large earthquake.

To give an example, the acceleration and

displacement time histories for each sen- sor in the USlOl/Painter Street overpass

are simulated for the magnitude 6.4 earthquake from the records of the magnitude

5.1 earthquake by using Eq. 15. The

corner frequency of the small earthquake

used in the calculations is fC3 = 0.5 Hz.

Figure 3 gives the comparison of the peak

values of the simulated and recorded ac- celerations and displacements. The comparison is good, particularly for the peak

displacements.

A key element of this and the previous

method is the assumption that the large

and small earthquakes have identical aver- age stress drop. When this assumption is removed, one needs to specify not only the

magnitude, but also the corner frequency

of the large earthquake in order to make

616
a prediction of the resulting structural re- sponse.

4. ORTHONORMAL EXPANSION

A more accurate simulation of large earthquake motions can be done by using the

small earthquake motions as Green's functions. The rupture process for the large

earthquakes can be assumed as the sum of

smaller ruptures, each of which generates a scaled version of the smaller earthquake.

With this assumption, we can write a record from the large earthquake as the

sum of scaled and time-shifted finite number of records, each of which is identical

to the record from the small earthquake.

That is:

p-i

yi(t) = Y,rk y3 (t-Tk ) (21)

fc=0

where r*. is the constant scaling factor and

Tfc is the delay time (where tq = 0) for the

kth. small earthquake in the sum, and p is the number of small earthquakes required

to match the large earthquake. Since r^.,

Tjt, and p are initially unknown, we can

use a more generalized form of Eq. 21 and

assume that:

m

yi (t) = Y,a,iy3 {t-i&) (22)

i=o

where m is a large number such that

raA >> rp _!. If Eq. 21 is valid, we

would then find in Eq. 22 that al — 0

for all i ^ Tfc/A, and a z

- = for all

i = rjt /A. The sequence a t is called the

impulse-response or the source function.

By taking the Fourier transform of Eq. 22,

we find that

Fy,(f) = Fa (f) • Fyt {f) (23)

where F denotes the complex Fourier

transform. If both the large and small

earthquake records are available, we first determine Fa (f) from Eq. 23 by taking

the ratio of the Fourier transforms of

the records, and then determine a; by

taking the inverse Fourier transform of

Fa (f). For discrete-time data, the maximum number of a x coefficients would be

determined by the length of the records.

The Fourier transform method of calculating d{ coefficients involves taking the ratio of the Fourier transforms of two noisy

records, and therefore is not very robust.

When signal to noise ratio is low, this approach can lead to erroneous estimation

of a 2 coefficients. An alternative would

be to use the time domain MA (MovingAverage) time series methods. However,

these methods are also not appropriate

because the number of a; terms required

to have a good match is usually too large

(e.g., i > 100). For such high orders, the

MA methods do not work because of the

accumulation of numerical round-off er- rors. A more robust method of obtaining the

cii's can be developed by generating a set of m orthonormal time series <f>i(t) from

ys (t) by using the following transformation (§afak, 1994):

m — 1 &W= 52 *> iiy,(t (24)

where bij is the mxm transformation matrix. The transformation matrix is determined from the condition that the <^(t)'s

form an orthonormal set; that is:

n s + m — 1 £ 4> l {t)<t> 3 {t) = 8lJ (25)

<=i

617
where i,j = 0, • •

• , (m— 1); t = 1, • • •

, +

m — 1); ns is the number of points in ys ; and 6{j is the Kronecker delta. The coefficients bij can be calculated from Eq. 25

by utilizing the singular value decomposition technique, and the orthonormal series

(f>i(t) are calculated from Eq. 24. Once

(j)i(t) are determined, the series yi(t) can

easily be expanded into its orthonormal

components by using the equation

m — 1

where the coefficients cj are calculated

from

rin+m — 1 c;= E M*)y.(*) (27)

The sequence Cj<f>j(t) in Eq. 26 represents

the contribution to the large earthquake

from the jth orthonormal mode. It can be shown by using Eqs. 22 and 26 that

the coefficients at

- are related to the coefficients bij and cj by the following equation

(§afak, 1994):

m— 1

To give an example, we consider the

displacements recorded by the three or- thogonal sensors (channels 12, 13, and

14) at the free field site near the east

abutment of the overpass during the two

earthquakes. For each orthogonal direction, m = 100 orthonormal displacement

time series from the displacements of the

small earthquake are generated by utilizing the singular value decomposition technique. The displacements of the large

earthquake and the impulse-response coefficients (source functions, a,-) are calculated from Eqs. 26, 27, and 28. Contributions from the three most significant

orthonormal modes to the free field displacements are given in Fig. 4 for the longitudinal, transverse, and vertical directions. The source functions for the same

directions are given in Fig. 5.

For linear wave propagation and structural response, the source functions for a

specified response direction should be similar for all sensors in that direction, irrespective of their location. This suggests

that if the source function for a reference

location is known, it can then be used

to predict the response in another location where we only have the small earthquake record. To test this, we use the displacements from the small earthquake and

the source functions calculated from the

free-field site, to calculate the displacement response to the large earthquake for

three sensors, namely: (1) the longitudinal sensor at the deck at the east abutment (channel 11), (2) the transverse sen- sor at the girder of the center bent (channel 7), and (3) the vertical sensor at the

deck at the center of the west span (channel 6). The comparison of the calculated

and the recorded displacements are shown

in Fig. 6. The comparison gives a good

match.

A practical application of this method

would be to predict the response of a

structure to the mainshock by recording

its response (using portable instruments)

during aftershocks. The same aftershocks

should also be recorded at the nearby permanent ground stations which recorded

the mainshock. To estimate the structure's response to the main shock, we first calculate the source function by using the

mainshock and the aftershock data from

618
the ground station, then convolve it with

the aftershock recordings from the structure.

4. CONCLUSIONS

Recordings of structural response from

small earthquakes and aftershocks can be

used to predict the response to larger

earthquakes of the future. Simple scaling laws for the structural response can be developed by using the analytical models of the seismic source spectrum, and

assuming constant stress drop and linear

behavior. Two such methods introduced

here give satisfactory results when applied

to recorded data. The records from a

large earthquake can be expanded into

orthonormal series that are generated by

transforming the records from a smaller

earthquake. This expansion permits the

prediction of the response to the large

earthquake (or the mainshock) by using

recordings from the small earthquake (or

the aftershocks).

5. REFERENCES

1. Brune, J.N. (1970). Tectonic stress and the spectra of seismic shear

waves from earthquakes, J. Geophys.

Res., 75, 4997-5009.

2. Joyner, W.J. and D.M. Boore (1986).

On simulating large earthquakes by

Green's-function addition of smaller

earthquakes, Earthquake Source Mechanics, Geophysical Monograph 37

(Das, et al., Eds.), American Geophysical Union, Washington, D.C.,

269-274.

3. Hanks, T.C. and H. Kanamori (1979).

A moment magnitude scale, J. Geophys. Res., 84, 2348-2350.

4. Hartzel, S.N. (1978). Earthquake af- tershocks as Green's functions, Geophys. Res. Letters, 5, 1-4.

5. Lin, Y.K. (1976). Probabilistic Theory of Structural Dynamics, Robert

E. Krieger Pub. Co., Huntington,

New York.

6. Mueller, C.S. (1985). Source pulse

enhancement by deconvolution of an

empirical Green's function, Geophys.

Res. Letters, 12, 33-36.

7. §afak, E. (1988). Analytical approach to calculation of response spectra from seismological models of ground

motion, Earthq. Eng. & Struc. Dyn.,

16, 121-134.

8. §afak, E. (1994). Scaling of seismic

structural response, Journal of Structural Engineering, ASCE, submitted

for publication.

9. Shakal, A. et al. (1992). CSMIP

strong-motion records from the Petrolia, California earthquakes of April

25-26, 1992, Report No. OSMS 92-

05, California Department of Conservation, Division of Mines and Geology, Sacramento, California.

10. Wennerberg, L. (1990). Stochastic

summation of empirical Green's functions, Bull. Seism. Soc. Am., 80,

1418-1432.

619
620
RECORDED VS. PREDICTED (BY CONSTANT SCALING) PEAK RESPONSE

104

S 103 J

5 102

ST

s 101 u

J

ffl 8 io°

10"1

. . Recorded

0 ^ Predicted

Accelerations

Displacements

0 2 4 6 8 10 12 14 16 18 20

SENSOR NUMBER

Figure 2. Comparison of recorded and predicted (by using constant scaling) peak accelerations and displacements for all channels.

RECORDED VS. PREDICTED (BY RECURSIVE FILTERING) PEAK RESPONSE

10" i 1 1 1 1 1 1 1 1 > 1 0 2 4 6 8 10 12 14 16 18 20

SENSOR NUMBER

Figure 3. Comparison of recorded and predicted (by using recursive filtering) peak accelerations and displacements for all channels.

621
THREE SIGNIFICANT MODES OF FREE-FIELD DISPLACEMENTS

12

LONGITUDINAL MODES

TRANSVERSE MODES

VERTICAL MODES

Mode 3

Mode 49

Mode 83

Mode 3

Mode 8

Mode 84

Mode 4

Mode 52

Mode 81

lO 15 20

TIME (S)

25 30 35

Figure 4* Contributions from the three most significant orthonormal

modes to the free field displacements in the longitudinal, transverse, and

vertical directions.

622
SOURCE FUNCTIONS FOR FREE FIELD DISPLACEMENTS

.on i , , 1 1 ,—_ . , , 1 0 123456789 10 TIME (S)

Figure 5. Source functions for the free-field displacements in the longitudinal, transverse, and vertical directions.

RECORDED VS. PREDICTED (FROM FREE-FIELD SCALING) DISPLACEMENTS

20 1

. .

i

1

" 150 5 10 15 20 25 TIME (S)

Figure 6. Comparison of recorded and predicted displacements at three

sensors in the structure.

623

Three Recent Tsunamis—Warning System Response

by

Michael E. Blackford*

ABSTRACT

Three recent earthquakes, located in the Pacific

basin, that generated tsunamis are selected to represent the response of the United States tsunami warning systems. The earthquakes

selected are the northern California earthquake

of April 25, 1992, the Nicaraguan earthquake of

September 2, 1992, and the Guam earthquake of

August 8, 1993. The Nicaraguan earthquake is the only earthquake of the three that resulted in significant destruction and loss of life. All three

earthquakes, however, did generate tsunamis that were recorded in Hawaii, remote from the sources. Two of the earthquakes occurred

during regular office hours of the warning

systems and one was reported to a warning

center by phone within six minutes of initial shock. Timing of message dissemination from

the warning centers was controlled mainly by

earthquake magnitude determination. This delay can become quite long for source areas at some

distance from the warning centers, rendering

them ineffective as regional warning centers. Additional warning centers are needed in many

areas of the Pacific, especially for the coastal areas of Central America and the islands of the

western Pacific.

KEY WORDS: body wave magnitude; directed

tsunami wave; surface wave magnitude; tsunami

warning response

1. INTRODUCTION

Three earthquakes that occurred within the last few years in the Pacific basin are selected as examples of the response of the United States tsunami warning centers to situations that have

the potential of generating destructive tsunamis.

Only one of the three, the earthquake in Nicaragua, actually resulted in loss of life and

significant property damage as a result of a tsunami. Other recent earthquakes such as

Indonesian earthquake of December, 1992 and

the Japan Sea earthquake of July, 1993, although

causing considerable loss of life and property

damage, occurred in marginal basins of the

Pacific or Indian Oceans and did not generate

tsunamis that traversed a significant portion of

the main Pacific basin. This paper shall take the

reader through a time history of the actions

taken by the centers from the onset of the

earthquake through the termination of their

activities related to the event.

There are two warning centers in the United

States. The Pacific Tsunami Warning Center,

located in Hawaii, acts as a regional warning

center for the Hawaiian Islands and as the

operational center for the Tsunami Warning

Center for the Pacific. The Alaska Tsunami

Warning Center, located in Palmer, Alaska, acts as a regional warning center for the northeastern

Pacific from the western end of the Aleutian

Islands to the border between the United States and Mexico.

One of the earthquakes selected occurred within

the Alaska Tsunami Warning Center's region.

The other two earthquakes occurred in areas without regional warning centers.

2. EARTHQUAKE OF APRIL 25,1992 IN

NORTHERN CALIFORNIA

At 8:06 A.M. Hawaii Standard Time (HST)

(18:06:04. 2Z) an earthquake (MS 7.1) occurred

off the coast near Cape Mendocino in northern

California. About one minute later the seismic waves reached Arnold Ranch seismic station near San Francisco and about eighty seconds

*NWS/Pacific Tsunami Warning Center, Ewa

Beach, Hawaii 96706

625
after that the waves reached the seismic station

at Palomar Observatory in southern California.

The signals from these stations, which are relayed to the Pacific Tsunami Warning Center

(PTWC), through the National Earthquake

Information Center (NEIC) in real time, were

noticed by the geophysicists. They began to take readings of the P-wave arrival times at these stations and at other United States mainland and Alaska seismic stations that are relayed to PTWC. At about 8:13 A.M. a final reading was taken from the local instrument at PTWC.

The P-wave arrival times were entered into the

computer and an initial location for the

earthquake was determined. The estimated

arrival times for data to determine the magnitude

of the earthquake were 8:19 A.M. for the S- wave data used to determine the body wave

magnitude, mb, and 8:22 A.M. for the Rayleigh wave data used to determine the surface wave

magnitude, Ms. The initial solution had a

relatively large r.m.s. time error indicating a potential reading error. The arrival time at the

Newport, Washington seismic station had indeed

been misread and, when it was corrected, the

solution time residual fell to an acceptable level. The final location of the epicenter was 40. 4N

and 124. 1W. This compares favorably with the

NEIC solution of 40. 4N and 124. 3W, however

the NEIC solution is much closer to the

coastline.

At 8:16 A.M. the Alaska Tsunami Warning

Center (ATWC), regional warning center for

this area, issued an information message for the

earthquake, giving it a magnitude of 7.0 based on the P-wave amplitudes. ATWC did not issue a warning, however it did indicate that some

local tsunami wave action may occur.

The body wave magnitude was determined to be

6.8 shortly after the S-wave was recorded at PTWC. A few minutes later, at 8:26, a surface wave magnitude of 7.0 was determined. Twenty

minutes had elapsed since the earthquake

occurred. PTWC issued an information message

indicating that no Pacific-wide tsunami was

expected and reported its surface wave

magnitude.

Indeed local tsunami action did occur with a maximum peak to trough amplitude of 110

centimeters being recorded at Crescent City,

California. Amplitudes of twenty centimeters or

less were reported at other sites in northern

California and southern Oregon. The tsunami was recorded at two sites in Hawaii, with fifteen

centimeters reported for Kahului, Maui and ten centimeters for Hilo. There was no reported

loss of life and no property damage specifically

attributed to the tsunami was reported.

3. EARTHQUAKE OF SEPTEMBER 2, 1992

OFF THE COAST OF NICARAGUA

At 2:16 P.M. HST, September 1 (00:16:01.6Z,

September 2) an earthquake was initiated about

eighty kilometers off the Pacific coast of

Nicaragua. Five minutes later P-waves from the

earthquake reached the seismic stations at French Village, Missouri and Blacksburg,

Virginia. A little over a minute later the seismic waves had passed the stations at Golden,

Colorado and Palomar, California. This activity was noted by the PTWC geophysicists and they

went into action determining arrival times and

locating the earthquake. Initial efforts indicated

the earthquake was centered in Honduras, well

inland from the Pacific coast. A body wave

magnitude of 7.5 and a surface wave magnitude

of 6.8 were determined from seismic waves that arrived at PTWC at twenty minutes and thirtyfour minutes after the origin time of the

earthquake, respectively. Because of a relatively

large time residual error for the solution, the

location was redetermined and found to be about

250 kilometers south of the original location,

placing the earthquake well into the Pacific.

Since procedures are based on the surface wave

magnitude, however, no further action was taken

other than to determine expected tsunami travel times from the new epicenter.

The closest water level station was Baltra in the

Galapagos Islands. Travel time to this station was about two hours and twenty minutes. The

time to the first transmission after this arrival

time was about two hours and forty minutes.

626
While waiting for this data, reports came from

NEIC and from other sources that a significant

tsunami had struck the coast of Nicaragua.

Concern over whether or not this tsunami would

be destructive Pacific-wide were heightened

when data from Baltra indicated a peak-to-trough

height of 111 centimeters. This concern was

lessened somewhat when data arrived from La

Libertad, Ecuador a short while later showing

only an eighteen centimeter tsunami wave

height. The concern was lessened further when

data from the Soccoro station was only twentyeight centimeters peak-to-trough and other, later arriving data indicated smaller amplitudes.

Valparaiso, for example, had a maximum

amplitude of only ten centimeters. Easter

Island, located along the path of the directed

tsunami wave that passed by Baltra, had a peakto-trough amplitude of eighty-three centimeters.

The great circle path of this directed wave did not pass near any large population centers so the

threat of a destructive Pacific-wide tsunami was

essentially eliminated. With regard to Nicaragua, by the time the location and

magnitude of the earthquake had been

determined, the destructive wave was already on

shore. No action by PTWC could have affected

the outcome of this event in Nicaragua.

4. EARTHQUAKE OF AUGUST 8, 1993

NEAR GUAM

At 10:40 P.M., August 7 (08:40Z, August 8)

the primary watchstander at PTWC received a telephone call from the observer at the U.S.

Geological Survey Seismic Observatory in

Guam. He reported the occurrence of a severe earthquake that had caused the loss electrical power at the observatory. He did manage to obtain a P-wave time before power was lost. He

passed this on to the watchstander who

immediately went over to the PTWC office at 10:42 P.M. to begin processing the earthquake.

A check of the NEIC automatic processor

showed that the event had not completed enough

of its transit across the national seismic network

to generate a preliminary solution. The

alarm sounded at 10:44 P.M., nearly ten minutes after the origin time of the earthquake.

Over the next six or seven minutes data was

read from the seismograms and an initial location at 15. ON and 145. 3E was determined.

Additional data from seismic stations in Asia and

the western Aleutians moved the epicenter closer

to the north end of the island of Guam. The

final location published by NEIC gave the

location as 13. ON and 144. 8E, about thirty

kilome 1 *rs "-"v1

.' sc itheasl of th< ;outhern tip oi Guam. The arrival times for the seismic waves

that could be used to determine the body wave

and surface wave magnitudes were 10:51 P.M.

and 11:00 P.M. respectively.

At 10:54 P.M. HST a body wave magnitude of

7.8 was determined from the S-waves recorded on the horizontal long period seismometers. Six

minutes later, at 11:00 P.M. HST, a surface wave magnitude of 8.0 was determined from the

vertical long period seismometer. Preparations were made to issue a regional warning/watch

message because of the size of the earthquake.

By this time, however, had there been a destructive local tsunami, it would have already

struck. The expected arrival time for Agana,

Guam was 08:41Z, nineteen minutes before the

magnitude determination.

At 11:08 P.M. HST a regional warning/watch

message was issued. Areas within three hours

tsunami travel time were placed in warning

status and areas within six hours tsunami travel time beyond the warning area were placed in a watch status.

Efforts now turned to determining whether or

not a Pacific-wide destructive tsunami had

indeed been generated. This earthquake was the

largest earthquake ever to have occurred in the

Marianas Islands historically. Therefore there was no previous history of tsunamis in this area

for earthquakes of this size. A calculation of

tsunami travel times indicated that arrival times

would 08:41Z at Guam, 09:47Z at Yap, and

09:56Z at Chuuk. Data was not initially

available via satellite telemetry from Guam,

however, data at Yap and Chuuk indicated

maximum peak-to-trough amplitudes of less than

fifteen centimeters. These stations are located

along a nodal direction with respect to the

627
azimuth of a directed tsunami wave so additional

data from stations more in line with the azimuth

of a directed tsunami were sought. The tsunami

arrived at Wake Island at 11:30Z and had a maximum amplitude of fifteen centimeters.

While awaiting the Wake data, data from the

Agana tide gauge was obtained from a secondary

source. The maximum amplitude at Agana was

eighteen centimeters peak-to-trough. This

information was sufficient to cancel the

warning/watch at 01:45 A.M. HST (11:45Z),

two hours and thirty-seven minutes after the

warning/watch was initiated.

Following the cancellation, reports of tsunami

amplitudes of up to ninety-eight centimeters were reported for sites in central Japan and in the Bonin Islands. Since there was typhoon

activity in the northwestern Pacific, it is not

clear whether these wave heights are totally

attributed to the tsunami or are a combination of

the tsunami and storm surge associated with the

typhoons. The tsunami was recorded on several

tide gauges in the Hawaiian Islands with a maximum amplitude of nineteen centimeters

being observed on the gauge at Port Allen on

Kauai Island.

5. CONCLUSIONS

The response time of the United States warning

centers is barely adequate for regional events and is inadequate for events in more remote

areas of the Pacific where regional warning

centers do not exist. The centers can effectively

track the progress of tsunamis across the Pacific

and can provide satisfactory warning of tsunami

hazard to areas beyond a few hours tsunami

travel time from the source. Areas within an hour tsunami travel time require a regional

system for warning. Ongoing education of

coastal inhabitants of areas subject to tsunami

hazard on procedures to take at the time of large

local earthquakes is an important adjunct to a regional warning system.

628
Hurricane Storm Surge Analysis for the Coast of Delaware

by

David J. Mark 1 , Norman W. Scheffner2 and Keith D. Watson3

Abstract

Coastal structures are designed to provide

protection against various forces of nature

acting in the coastal zone. Structures such as seawalls are expected to provide onshore

protection from the effects of storm surge. However, the degree of protection is usually

limited by the economics of constructing for a

specified storm surge elevation. This elevation

is determined from a frequency-of-occurrence

analysis of historical events. When adequate storm surge data are not available, statistical methods are used in generating reliable stagefrequency relationships. This paper describes

the development of a database of storm surge

elevations and the application of the Empirical

Simulation Technique used in developing

stage-frequency relationships for the coast of

Delaware.

Keywords: Delaware; empirical simulation

technique; hurricane; numerical model; stagefrequency; storm surge.

1. INTRODUCTION

The U.S. Army Engineer District, Philadelphia

is presently developing a storm damage

reduction and shoreline protection program for the open coast of Delaware (Figure 1). Accurate stage-frequency relationships and

storm surge time-histories are required for estimating long-term shoreline erosion. Because of the infrequent occurrence of

hurricanes at a given site, and therefore lack of historical data, a standard ranking method

could not be used in a stage-frequency

analysis; thus, statistical procedures were

applied in this analysis.

Traditionally, a joint probability

method (JPM) is employed in synthesizing storm events subsequently used in hurricane

stage-frequency analyses. However, the JPM

usually assumes that all hurricane parameters,

such as central pressure deficit and maximum

wind speed, are independent, ignoring the interdependence of storm parameters. Consequently, unrealistic hurricanes are created by

arbitrarily combining parameter values. For

example, one parameter can be assigned a value typical of a weak storm, whereas a second is assigned a value representative of an

intense storm. Thus, a level of uncertainty is incorporated into the stage- frequency

computations.

An alternative approach to the JPM is the empirical simulation technique (EST) which

preserves the interdependence of hurricane

parameters. The EST is a statistical resampling

procedure which uses historical data to develop

joint probability relationships among the various measured storm parameters. The

resampling scheme generates large populations

of data which are statically similar to a much

smaller database of historical events. Using

this expanded data set, the EST generates a database of peak storm surge elevations by

simulating multiple year periods (e.g., 200-yr

periods) of storm activity a multiple number of

times (e.g., 100 simulations). Stage-frequency

relationships are then generated using the database of peak storm surge elevations.

Research Hydraulic Engineer, Senior

Research Hydraulic Engineer, US Army

Engineer Waterways Experiment Station, 3909

Halls Ferry Road, Vicksburg, MS 39180-6199;

3 Hydraulic Engineer, US Army Engineer District, Philadelphia, Wanamaker Bldg. , 100 Perm

Square East, Philadelphia, PA 19107-3390.

629
This technique does not rely on assumed

parametric relationships but uses the joint

probability relationships inherent in a database

of storm parameters. In this approach, probabilities are site specific, do not depend on

fixed parametric relationships, and do not assume parameter independence. Thus, the EST is "distribution free" and nonparametric.

The only assumption made is that future events

will be statistically similar in magnitude and

frequency to past events.

This paper describes the process used in developing the hurricane-induced stagefrequency analysis for the coast of Delaware.

Their development consists of three

interrelated tasks, each employing a numerical

model. In the first task, historical hurricanes

impacting the study area were analyzed to determine storm statistics and correlations. From these data, a reduced set of hurricanes,

representative of all storms impacting the area, were chosen and subsequently simulated with

a tropical wind field model to generate wind

and atmospheric pressure fields.

In the second task, storm surge events developed with the wind model output were

simulated using a long-wave, finite elementbased hydrodynamic model to obtain peak

storm surge elevations. With the hurricane

parameters serving as input to the wind field model together with the corresponding storm

surge elevations predicted by the storm surge model, the EST is employed in developing

frequency-of-occurrence relationships in the

third task. A description of the EST and the methodology used in performing these tasks is presented in the following sections.

2. Description of Empirical Simulation

Technique

The EST is a statistical resampling technique

which uses historical data to develop joint

probability relationships among the various

measured storm parameters. For this study, the EST was developed to generate numerous

multi-year intervals of possible future hurricane events for the coast of Delaware.

The ensemble of modeled or simulated events

are consistent with the statistics and correlations of past storm activity at a site. Furthermore, the EST permits random

deviations in storm behavior (when compared

to historic events) that could occur in the

future. For example, simulated hurricanes are permitted to make landfall at locations other than those made by the historical storms. These random deviations can also result in more intense storms than the historical events themselves, allowing for the possibility of a

future hurricane being the storm of record.

The simulation approach requires specifying a

set of parameters which describe the dynamics

of some physical system, such as hurricanes.

These parameters must be descriptive of both

the process being modeled and the effects of

that process. In the case of tropical storms,

pertinent input vectors include: 1) the central pressure deficit; 2) the radius to maximum

winds; 3) maximum wind speeds; 4) minimum

distance from the eye of the storm to the

location of interest; 5) forward speed of the eye; and 6) tidal phase during the event. These

values are defined for each specific location corresponding to each particular historical or

hypothetical event of the total set of storm

events used in the study.

The second class of vectors involve some

selected response resulting from the input- vector parameterized storm. For this study,

the maximum total water surface elevation,

reflecting the combined tide plus storm surge,

is the response vector of interest. Although

response vectors are related to input vectors, the interrelationship is highly nonlinear and

involves correlation relationships which can

not be directly defined, i.e., a non-parametric

relationship. For example, in addition to the storm input parameters, storm surge is a

function of local bathymetry, shoreline slope

and exposure, ocean currents, temperature, etc. as well as their spatial and temporal gradients.

It is assumed, however, that these combined

effects are reflected by the response vector.

From the entire set of storms, a subset of

storms is selected which is representative of all storms that have impacted a particular

location. This subset is referred to as the

630
training set. This set is used as input for appropriate numerical models, such as the hydrodynamic model, for computing the desired response vector(s). The training set usually includes historical events but may

include historical storms with slight

perturbations. For example, a historical storm

with a slightly altered path.

The training set of storms can be augmented

with additional storms contained in the

historical data set. Storm events augmenting

the training set are referred to as the

"statistical set" of storms. Whereas numerical

models are used for generating response

vectors for events residing in the training set, response vectors for the statistical set of storms

are interpolated using the training set response

vectors. Thus, stage-frequency relationships can be generated using the entire historical data set without need of simulating all storms

in that data set.

With the augmented storm data set (i.e., training and statistical storm sets), the EST

produces N simulations of a T-year sequence

of events (hurricanes), each with their associated input vectors and response vectors. Because there are N-repetitions of a T-year

sequence of events, an error analysis of the

results can be performed with respect to median, worst, least, standard deviations, etc.

The simulation of realistic events is accounted

for in the nearest-neighbor interpolation,

bootstrap, resampling technique developed by

Borgman (Borgman, et al 1992). The basic

technique can be described in two dimensions

as follows. Let Xj, X2 , X3 , Xn be n

independent, identically distributed random

vectors (storm events), each having two

responses [Xj — {x

; (l), X;(2)}; i = l,n].

Each event X; has a probability p; as 1/n,

therefore, a cumulative probability

relationship can be developed in which each storm event is assigned a segment of the total probability of 0.0 to 1.0. If each event has an

equal probability, then each event is assigned

a segment Sj such that s- - Xj.

A random number is selected to identify a

storm event from the total storm population or augmented storm set. (The procedure is equivalent to drawing and replacing random

samples from the total storm event population.)

Using this storm event, the model performs a random walk from the event Xj with Xj and x2 response vectors to its nearest neighbor

vectors. The walk is based on independent

uniform random numbers and has the effect of

simulating responses which are not identical to the historical events but are similar to those

events which have historically occurred. Thus,

the EST is not simply a resampling of

historical events technique, but rather an

approach intended to simulate the vector

distribution contained in the augmented set of

storms.

Because the simulated events correspond to a

specific location, the criteria that the total number of storm events selected in a T-year

simulation must be statistically representative

of the number of historical events impacting

that location must be satisfied. For this study,

33 hurricane events were identified which

impacted the coast of Delaware during the 104

year period extending from 1886 through 1986

(Jarvinen, et al 1988). Given the mean

frequency of storm events for a particular

region, a Poisson distribution is used to determine the average number of expected

events in a given year. For example, the Poisson distribution can be written in the following form:

for s=0, 1,2,3,... The probability Pr(s;X) defines

the probability of having s events per year where X is a measure of the historically-based

number of events per year. For this study, X is computed as 0.32 (33/104).

A 10,000 element array is initialized to the above Poisson distribution. The number

corresponding to s=0storms per year is 0.7261,

thus if a random number selection is less than or equal to 0.7261 on an interval of 0.0 to 1.0, then no hurricanes would occur during that year of simulation. If the random number is

631
between 0.7261 and 0.7261 +P[N=1] =0.7261 + 0.2324 =0.9585, one event is selected. Two

events for 0.9585 +0.0372 =0.9957, etc. When

one or more storms are indicated for a given

year, they are randomly selected using the nearest neighbor interpolation technique

described above.

An empirical estimate of Fx

(x^) , denoted by

Fx

(X(

r) ), is given by Gumbel (1954) (see also Borgman and Scheffner, 1991 or Scheffner and

Borgman 1992).

Output of the EST program is N repetitions of

T-years of simulated storm event responses. It

is from these responses that frequency-of- occurrence relationships are computed. The

computational procedure followed is based on

generating a probability distribution function

(pdf) that corresponds to each of the T-year

sequence of simulated data.

Estimates of frequency-of-occurrence begin

with the calculation of a pdf for the response

vector of interest (e.g., peak storm surge

elevation). Let Xp X2 , X3 , Xn be n

independent, identically distributed, random

response variables with a cumulative pdf

Fx

(x) = Pr[X z x] (2)

where Pr[ ] represents the probability that the random variable X is less than or equal to some

value x and Fx

(x) is the cumulative probability

density function ranging from 0.0 to 1.0. The

problem is to estimate the value of Fxwithout

introducing some parametric relationship for probability. The following procedure is adopted because it makes use of the probability laws defined by the data and does

not incorporate any prior assumptions

concerning the probability relationship.

Assume that we have a set of n observations of

data. The n values of x are first ranked in order of increasing size such that

XW * *(2) * *(3) * - • " * X(n) (3)

where the parentheses surrounding the subscript indicates that the data have been

rank-ordered. The value x^ is the smallest

in the series and x,. represents the largest. Let r denote the rank of the value x^ such

that rank 1 is the smallest and rank r =n is the largest.

for {x

(r)

, r = 1, 2, 3,...., n}. This form of

estimate allows for future values of x to be

less than the smallest observation x^ with

probability of l/(n+l), and to be larger than

the largest value x

(n) also with probability

l/(n+l).

This form of the cumulative distribution function allows for values of x to be greater

than the maximum or less than the minimum

observed values in the historical data base. In implementing the EST, tail functions (Borgman

and Scheffner 1991) are used to define the pdf

for events larger than the largest or smaller

than the smallest observed event so that there

is no discontinuity in the pdf.

The cumulative pdf as defined by

Equation 4 is used to develop stage-frequency

relation-ships in the following manner.

Consider that the cumulative probability for an

n-year return period storm can be written as:

F(n) = 1 - - (5) n

where F(n) is the cumulative probability of

occurrence for an event with a return period

of n-years. Frequency-of occurrence relation- ships are obtained by linearly interpolating a

stage from Equation 5.

3. Description of Wind and Atmospheric

Pressure Field Model

The Planetary Boundary Layer (PBL) wind

field numerical model (Cardone et al. 1979) was selected for simulating hurricanegenerated wind and atmospheric pressure

fields. The model employs the verticallyaveraged primitive equations of motion for

632
predicting wind velocities experienced within

a hurricane. The model includes

parameterization of the momentum, heat and

moisture fluxes together with the surface drag

and roughness formulations.

The model requires a series of "snapshots" for input consisting of a set of meteorological

parameters defining the storm at various stages

in its development or at particular times during

its life. These parameters include: latitude and longitude of the storm's eye; track

direction and forward speed measured at the eye; radius to maximum winds; central and

peripheral atmospheric pressures; and an

estimate of the geostrophic wind speed and

direction.

Some meteorological storm parameters were

obtained from the HURricane DATabase

(HURDAT) developed by the National

Oceanic and Atmospheric Administration

(NOAA) National Hurricane Center (NHC)

(Jarvinen, Neumann and Davis 1988). This

database summarizes all hurricanes and

tropical storms that occurred in the North

Atlantic Ocean over the 104-yr period from

1886 through 1989. Information contained in

this database includes latitude and longitude of

the storm, central pressure, and maximum

wind speed.

Radius to maximum winds is approximated

using a function that incorporates the maximum wind speed and the atmospheric

pressure anomaly (Jelesnianski and Taylor,

1973). Track directions and forward speeds

required by the PBL model are approximated

hourly, using a cubic spline interpolation

technique, from the storm's 6-hr latitudinal and longitudinal positions provided in the HURDAT database. Peripheral atmospheric

pressures were assumed equal to 1013 mb.

Geostrophic wind speeds were specified as 6

knots, typical conditions within the Gulf of

Mexico (Thompson 1993), and have the same

direction as the storm track.

Whereas some models employ a fixed grid

system to simulate a hurricane (i.e., stationary

grid with a moving storm), the PBL model

simulates the hurricane as a stationary storm

with a moving grid. A hurricane's translational or forward motion is incorporated into model calculations by adding the forward and

rotational velocity vector components. Wind

velocities and atmospheric pressures are computed at each node in the translating

uniform grid.

The model uses a nested gridding technique,

composed of five layers or subgrids, for computing the wind fields. Each subgrid measures 21 by 21 nodes in the x- and ydirections, respectively, and the centers of all subgrids are defined at the eye of the hurri- cane. Although the number of nodes

composing each subgrid is the same, the area of coverage and spatial resolution differs for each grid. For this study, the subgrid with the

finest resolution had an incremental distance

of 5 km between nodes. Incremental distances

for the remaining subgrids were 10, 20, 40,

and 80 km.

For each snapshot, the equations of motion are

first solved using the grid covering the greatest

area, which is the grid having an incremental

distance of 80 km between nodes. Computed

wind velocities are then used as boundary

conditions on the second largest grid, and the equations of motion are solved again. This same procedure is followed for the remaining

grids where wind fields are computed using

sequentially smaller grids together with wind

velocities computed with the next larger grid

serving as boundary conditions.

After wind and pressure fields have been

computed for each 6-hr snapshot, hourly wind

and atmospheric pressure fields are

interpolated using a non-linear blending

algorithm which produces a smooth transition from one snapshot to the next. Hourly wind

and pressure fields are then interpolated from

the PBL grid onto the hydrodynamic grid and

subsequently stored for use by the

hydrodynamic model.

4. Description of Storm Surge Model

The ADvanced CIRCulation (ADCIRC)

numerical model (Westerink et al. 1992) was

633
chosen for simulating the long-wave

hydrodynamic processes in the study area. This program employs a two-dimensional,

depth-integrated finite element solution of the Generalized Wave-Continuity Equation

(GWCE). The fundamental components of the GWCE equation are the depth-integrated

continuity and Navier-Stokes equations for conservation of mass and momentum. The

time-differentiated form of the conservation

of mass equation is combined with spacedifferentiated form of the conservation of

momentum equation to develop the GWCE

equation.

The GWCE-based solution scheme eliminates

several problems associated with finite element

programs which solve the primitive forms of

the Navier-Stokes equations, including

spurious modes of oscillation and artificial damping of the tidal signal. Forcing functions

contained in this code include time-varying water surface elevations, wind shear stresses, atmospheric pressure gradients and the Coriolis

effect.

The storm surge model was adapted from the

East, Gulf and Caribbean Sea hydrodynamic

model developed in the USAE Dredging

Research Program (DRP). The numerical grid,

specified in spherical coordinates, was devel- oped by increasing the resolution of the DRP

East Coast grid in the study and surrounding

areas. The grid consists of 11,829 nodes and

21,917 elements. Figure 2 presents the numerical grid used in this study and Figure 3 provides a "zoomed-in" picture of the grid

along the study area.

Using a grid which incorporates the entire Eastern seaboard provides several benefits.

First, extending the seaward boundary beyond

the continental shelf enables a linear tidal signal to be specified at the mid-Atlantic open

boundary; overtides, generated in shallow

water, are implicitly included in the model

computations.

Second, the non-linear effects described above

must also be addressed for the lateral openwater boundaries, which extend from shore to the ocean's abyssal plain. Typically, lateral

boundaries are placed sufficiently far from the study area so that errors in the time-series do

not corrupt model-generated results within the area of interest. With the grid extending from

Nova Scotia, Canada to Venezuela, this problem is minimized.

Third, the majority of hurricanes affecting the coast of Delaware approach the mainland from

the southeast, then veer away to the northeast. As the hurricane enters the model area, the

tidal signal along the lateral open-water

boundary can be significantly altered by a

storm's atmospheric pressure anomaly. Thus,

the flow field can be corrupted in much the same manner as described above. As before,

this problem can be alleviated by modeling the

entire eastern seaboard.

Bathymetry was supplied from the National Center for Atmospheric Research's ETOP05

data base. These data are stored at a resolution of 5 min latitude by 5 min longitude and

depths have the units of meters. Bathymetry

in estuarine and nearshore areas were obtained

from NOAA nautical charts.

Time-series of astronomical tidal elevations

specified at the open water boundary were

computed from tidal amplitudes and phases

obtained from Schwiderski and Szeto (1981).

Modeled constituents include M2 , S

2 , N2 , Oj,

Pj, K2 , Kj and Qj. Constituent data are provided at one degree increments in latitude and longitude. A bilinear interpolation

algorithm was used for estimating tidal amplitudes and phases at the 61 grid nodes

composing the open water boundary.

5. Tidal Calibration of Hydrodynamic Model

The hydrodynamic model was calibrated by

adjusting local bottom friction coefficients so

that model-generated water surface level timeseries compare favorably to those

reconstructed from tidal constituents.

Locations at which comparisons were made

include: Cape May, NJ; Reedy Point, DE;

Lewes, DE; Indian River Bay Inlet, DE; and,

Ocean City, MD. Calibration simulations were

conducted over a 30-day period beginning at

634
0:00 Eastern Standard Time on 1 September

1985. A 30-sec time-step was used in these simulations. Comparison of computed and

constituent-generated water surface levels for the Lewes gage is presented in Figure 4.

Of the five stations, the greatest difference

between computed and constituent-generated water elevation extrema was experienced at Reedy Point where the model overpredicted

tidal amplitudes by 11 percent. Differences in extrema for the remaining stations were less than 4 percent. The discrepancy at Reedy

Point is attributed to the station's close proximity to the grid boundary; placed

downstream of the head-of-tides, the boundary acts as a dam preventing tides propagating upstream/downstream of the

station and, consequently, increases tidal amplitudes at the station.

Phase differences between the computed and

constituent-generated tidal oscillations ranged

from 0.18 to 0.54 hours. Computed tidal oscillations at the Cape May, Lewes, and

Reedy Point stations led the constituent- generated oscillations, whereas at Indian River

Inlet and Ocean City, the computed tidal oscillations occurred later than the constituent- generated phases. The difference in phases at the Indian River Inlet and Ocean City stations

is attributed to the lack of grid resolution defining these areas.

6. Storm Surge Validation of Hydrodynamic

Model

Model validation was achieved by performing

a storm surge simulation of Hurricane Gloria

which impacted the study area in September

1985. The hindcast began on 1 September

1985 at 0000 Greenwich mean time (GMT) and

ended on 1 October at 0000 GMT. A 30-sec

time-step was used in the simulation and tidal forcing was specified at the open water

boundary.

From the beginning of the hindcast simulation

to 16 September at 1200 GMT, no wind or atmospheric pressure gradients were specified

in the model. This 16-day period provides

sufficient simulation time to develop an

accurate tidal current field and to dampen any

start-up errors. Thereafter, from 16 September at 1200 through the end of the hindcast at 0000 GMT on 1 October, wind and

atmospheric pressure fields were supplied to the model at hourly increments. Furthermore,

nodal wind and pressure values were linearly

interpolated at time-steps falling between

whole hours.

Figure 5 presents a comparison of modelgenerated and measured water surface

elevations recorded at Lewes, DE for the period of 25 September at 0000 GMT to 0000

GMT on 30 September. The model predicted

a peak water surface elevation of approximately 6.21 ft mean sea level (MSL), whereas

the Lewes gaging station measured a peak

elevation of 5.99 ft MSL. Thus, the model

overpredicted the peak water level by 0.2 ft.

As shown in the figure, the hydrodynamic

model predicted the peak storm surge elevation

approximately 1 hr later than its measured

occurrence. This lag is attributed to the assumption that the hurricane's forward speed was constant during the 6-hr period between

"snapshots" or entries contained in the NHC

database.

Whereas the model accurately predicted the peak storm surge elevation, model-generated

water surface elevations underpredicted the measured data by as much as 1 ft during the 36-hr period prior to the storm's passage. This

discrepancy is believed to be due to differences between PBL-generated wind

directions and those experienced during the storm; it is believed that the westerly

component of the computed winds were

greater than those experienced during the storm, forcing greater water volumes into the upper reaches of Delaware Bay while lowering water levels along the lower bay's southwest

shoreline (which is where the Lewes gaging

station is located).

7. Selection of Hurricanes

Because stage-frequency relationships are

635
based on a statistical analysis of storm surge

elevations resulting from historical storm

events, a thorough analysis was performed to define those storms which impacted the study

area. As a first step in selecting hurricanes,

track positions of all hurricanes in the NHC

database were processed to determine those storms which affected the Mid-Atlantic states. Thus, any hurricane that did not track through

the "box" extending from latitude 36 OO'OO" N.

longitude 77 OO'OO" W. to latitude 41 OO'OO" N.

longitude 68 OO'OO" W. was eliminated from

consideration.

Of the 875 hurricanes in the NHC database,

only 66 hurricanes passed through this box.

After reviewing these 66 hurricanes, an

additional 36 hurricanes were eliminated from

the storm ensemble. Reasons for eliminating

these hurricanes include: some storms only

"clipped" the edges of the box and did not come into close proximity with the study area; a few hurricanes were downgraded to severe storm status before entering the box and were

thus ignored in the storm surge analysis; and, some storms tracked along the western edge of

the box and did not impact the coast of

Delaware.

With this set of 30 hurricanes, a set of 15 storms (referred to as the "training set") was

selected and simulated using the storm surge model. Hurricanes composing the training set were selected such that the set is representative

of the entire set of 30 storms which impacted

the study area.

Storm surge elevations generated using these storms can be considered as approximations of

the historical events. Although the frequencies

associated with their maximum surge may be

considered accurate, the value of peak surge may not correspond to historically observed

surge elevations. Differences in peak surge are

attributed to estimates of the hurricane

parameters which are only approximate; all data necessary to simulate each event are not known and have not been calibrated. Therefore, the hydrographs should not be

considered as hindcasts of the historical events.

8. Application of Storm Surge Model

All 15 storms in the training set together with

an additional 5 hurricanes were simulated with

the storm surge model. These extra hurricanes were added to the training set to ensure the accuracy of the response vector population for input to the EST model. Each storm surge

simulation was performed independently of

tidal action, eliminating the task of extracting surge levels from a time-series of combined

tide- and surge-induced water surface levels. A 30-sec time-step was used in each

simulation. Time-series of water surface

elevations were recorded, at 15-min intervals,

at 40 stations in the study area. Modelgenerated and measured peak storm surge

levels for Lewes are summarized in Table 1.

Differences between computed and measured

surge levels ranged from 0.2 ft to 0.9 ft. As

with the model validation exercise, these

differences are attributed, in part, to discrepancies between the PBL-generated wind

directions and those produced by the storms. Winds generated with the PBL model were

computed independently of adjacent weather

systems; without the adjacent systems, the PBL

model is unable to precisely hindcast wind

directions necessary for reproducing the measured surge levels. Because the Lewes gage

is partially sheltered by Cape Henlopen, small

discrepancies in computed wind directions can have an appreciable impact on the hydrodynamic model's ability to predict surge levels.

9. Application of EST Model

Input to the EST model consists of hurricane

parameters, or input vectors, describing each

storm in the training set together with their associated peak total water surface elevations

(or response vectors). This data set is referred

to as the training set of storms. Furthermore,

the training set was augmented with those storms that occurred in the study area but were

omitted from the training set. This second set of storms is referred to as the statistical set. The combined storm set consisting of the

training and statistical sets is referred to as the augmented storm set. No response vectors are

636
specified with the statistical set of storms be- cause the EST model computes these vectors

internally. Tables 2 and 3 present the training

and the statistical set input vectors for the Lewes gaging station.

Hurricane parameters used as input vectors in the EST model are: maximum wind speed;

radius to maximum winds; atmospheric

pressure anomaly; translational or forward

speed; track direction; and the minimum

distance between the study area and the eye of

the hurricane. Parameter values are selected at the time a hurricane makes its closest approach

to the study area.

The NHC database was processed to determine

the necessary parameter values. Data

contained in this database, however, are provided at 6-hr increments. Therefore, for each storm in the augmented storm set, a cubic

spline interpolation or curve-fitting procedure was followed to compute hurricane positions at hourly intervals. Using the interpolated

hurricane positions, the minimum distance

between the hurricane's track and the study

area was determined. The required hurricane

parameters were then interpolated from NHC's

6-hr incremental data.

In developing response vectors, peak storm

surge elevations were extracted from the storm

surge time-histories created by simulating the storms in the training set. Each surge

elevation was then combined with four tidal elevations. The basis for this procedure is that a storm surge event, and therefore its contribution to the peak total water surface

elevation, is independent of the tidal cycle. In

other words, peak storm surge levels can occur

at any time during the tidal cycle. Thus, surge

levels must be combined with a range of tidal elevations in order to accurately represent the temporal phasing of surge and tide.

Four tidal elevations were specified for each storm event in order to represent the tidal water level component of the total water

surface elevations. One elevation represented

high tide elevations whereas a second depicted

low tide levels. The two additional elevations equaled zero; thus, the combination of four

elevations represent one tidal cycle. Further- more, the EST model randomly chooses the

tide elevation, based on the four elevations, in the simulation procedure. For this study, it was assumed that the maximum and minimum

tidal elevations were equal to the summation of

the M2 S

2 , N2 , Op

Pj, K2 , Kj and Q{ constituent amplitudes, which represents

the average spring tide elevation.

With four tidal elevations being specified with

each set of input/response vectors, the total number of storms in the training set were

effectively increased from 20 to 80 events. Four identical sets of input vectors or

hurricanes were added to the statistical storm

set, increasing the number of storms in this set from 25 to 100 storms. Thus, the augmented

storm set contained 180 storms. (Adding the

additional input vectors to the augmented

storm set avoids biasing the training set's input

vectors in the stage-frequency computations.)

With the EST model, 100 simulations each

having a 200-yr duration were performed for each of the 40 stations. Because the computation is repeated 100 times, 100

individual pdfs and 100 stage-frequency

relationships are computed. This family of

curves is averaged and the standard deviation

computed, resulting in the generation of a

stage-frequency relationship containing a measure of variability of data spread about the mean value. Figure 6 shows the plot of the computed mean value as well as all 100

frequency relationships for the Cape May, NJ

station.

This figure demonstrates a reasonable degree

of variability in the 100 simulations which

should be expected from a stochastic process

involving a large number of events. Because

elevation data are computed continuously, the

full population of data are subsequently

reduced to a plot of mean value bracketed by

plus-and-minus one standard deviation. This

presentation of data provides a mean

frequency estimate for design as well as an

error, or confidence, band which can be

incorporated into a risk analysis.

The EST- generated stage- frequency

637
relationships were compared to those computed

using the JPM for the Ocean City and

Rehoboth Beach stations (Po, et al 1976).

These comparisons are shown in Figures 7 and

8 for the Ocean City and Rehoboth Beach

station, respectively. As shown, the ESTgenerated frequency relationships agree closely

with the relationships developed with the JPM.

Generally, stages produced with the JPM are within 0.25 ft and also one standard deviation

of those stages generated with the EST.

Furthermore, the EST required 20 model

simulations as opposed to the 468 simulations

needed for the JPM.

10. Summary

This paper describes an application of a

stochastic modeling procedure for generating

frequency-of-occurrence relationships for hurricane-induced surges for the coast of

Delaware. Three models were employed in

this analysis, including: a wind and

atmospheric pressure field model; a hydrodynamic long wave model; and, an empirical

simulation model. This paper provides an overview of each model together with

description of the procedures taken in generating the frequency-of-occurrence

relationships.

The approach uses an expanded data set of

historical storms to develop joint probability

relationships among the various descriptive storm parameters and responses such that multiple repetitions of multiple years of storm

activity are generated. Each simulation

reflects the joint probability inherent in the

original database of historic events and does

not rely on assumed parametric relationships.

Therefore, computed probabilities are site specific, do not depend on fixed parametric

relationships, and do not assume parameter

independence.

The accuracy of the statistical procedures

described herein is demonstrated for two

stations located along the open coast. Because

the approach defines not only the mean

frequency-of-occurrence but also a measure of

the error, the described procedures show great

promise for defining risk-based frequency

relationships for any coastal location.

11. Acknowledgments

The technical contributions of Drs. Leon E. Borgman and Joannes J. Westerink to the modeling study described herein are

acknowledged. This study was funded by the U.S. Army Engineer District, Philadelphia.

Permission was granted by the Chief of

Engineers to publish this paper.

12. References

1. Borgman, L. , Miller, M., Butler, L., and

Reinhard, R. 1992. "Empirical Simulation

of Future Hurricane Storm Histories as a Tool in Engineering and Economic

Analysis," Fifth International Conference on Civil Engineering in the Oceans, ASCE,

College Station, TX, 2-5 November 1992.

2. Borgman, L.E. and Scheffner N.W., 1991,

"The Simulation of Time Sequences of

Wave Height, Period, and Direction," U.S.

Army Engineer Waterways Experiment

Station, Vicksburg, MS, Technical Report

DRP-91-2.

3. Gumbel, E.J., 1954, "Statistical Theory of

Extreme Value and Some Practical Application, " National Bureau of Standards

Applied Math. Series 33, U.S. Gov. Publ.,

Washington, DC.

4. Jarvinen, B. R., Neumann, C. J., and

Davis, M. A. 1988. "A Tropical Cyclone

Data Tape for the North Atlantic Basin,

1886-1983: Contents, Limitations, and

Uses," NOAA Technical Memorandum

NWS NHC 22, National Hurricane Center,

Miami, FL.

638
5. Jelesnianski, C. P. and Taylor, A. D. 1973. "A Preliminary View of Storm Surges

Before and After Storm Modifications,"

NOAA Technical Memorandum ERL

WMPO-3, Weather Modification Program

Office, Boulder, CO.

6. Po, F.P., Tracey, R.J., Myers, V.A., and

Foat, N.S., August 1976, "Storm Tide

Frequency Analysis for the Open Coast of

Virginia, Maryland, and Delaware,"

NOAA Technical Memorandum NWS

HYDRO-32.

7. Scheffner, N. W. and Borgman, L. E. 1992. "Stochastic Time-Series Representation of Wave Data," Journal of

Waterway, Port, Coastal and Ocean

Engineering, ASCE, Vol. 118, No. 4, pp. 337-351.

8. Schwiderski, E. W. and Szeto, L. T. 1981. "The NSWC Global Ocean Tide Data Tape,

Its Features and Application, RandomPoint Tide Program," Technical Report

NSWC 81-254, Naval Surface Weapons

Center, Dahlgren, VA.

9. Thompson, E. F. 1993. Personal

communication. U.S. Army Waterways

Experiment Station, Vicksburg, MS.

10. Westerink, J. J., Luettich, A. M., Baptista, A. M., Scheffner, N. W., and Farrar, P. 1992. "Tide and Storm Surge Predictions

using Finite Element Model," Journal of

Hydraulic Engineering, ASCE, Vol. 118,

No. 10. pp. 1373-1390.
640
Figure 3 Numerical grid in vicinity of study area.

CALIBRATION: SEPTEMBER 85

Lewes

381.0 108.0 132.0 1S6.0 180.0 501.0 TIME (HRS)

S28.0 S52.0 576.0

Figure 4 Comparison of water levels at Lewes Station

641
HURRICANE GLORIA: SEPTEMBER 85

LEHES, OE

642
>3

0.0 25.0

REH0B0TH BEACH, DE

EST: MEAN

EST: MEAN +/- STD DEV JPM

50.0 75.0 100.0 125.0 RETURN PERIOD (YRS)

150.0 —I 175.0 200.0

Figure 7 Comparison of EST- and JPM-generated frequency relationships for Ocean City

>

CD

0.0 25.0

OCEAN CITY, MD

EST: HEAH

EST: MEAN +/- STD DEV

JPM

SO.O 75.0 100. 0 125.0 RETURN PERIOD ( YRS)

150.0 175.0 200.0

Figure 8 Comparison of EST- and JPM-generated frequency relationships for Rehoboth

Beach

643
Table l. Comparison of Predicted and Measured Peak Storm

Surge Elevations at Lewes, DE.

Name

Hurricane

Number Date

Peak Surge

Predicted

(ft)

Elevation

Measured

(ft)

Unnamed 327 Aug 193 3 2.1 na

unnamed 332 Sep 1933 2 . 5 1.7

unnamed 370 Sep 193 6 2 . 2 na

unnamed 386 Sep 1938 1.7 na

unnamed 436 Sep 1944 4.7 4.5

unnamed 440 Oct 1944 1.0 na

unnamed 476 Aug 1949 0.5 na

Carol 535 Aug 1954 0.5 na

Connie 545 Aug 1955 3.0 2.3

Daisey 575 Aug 1958 0.3 1.2

Donna 597 Aug 1960 1.1 na

Doria 657 Sep 1967 2.5 3.2

Belle 748 Aug 1976 0.8 na

Gloria 835 Sep 1985 4.5 4.3

Charley 842 Aug 1986 0.1 na

na - not available

644
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646
APPENDIX:

TASK COMMITTEE

REPORTS

Report of Task Committee on

(A) Strong-Motion Data and Applications

Date : May 17, 1994

Place: National Institute of Standards and Technology

Gaithersburg, MD

Attendees: U.S. Side - G. Brady (Chairman)

H. Meyers

J. Hill

(USGS)

(NOAA)

(DOE)

Japan Side S. lai (Acting Chairman)

K. Ohtani

T. lijima

(PHRI)

(NIED)

(PWRI)

1. OBJECTIVE AND SCOPE OF WORK

The objectives of the Task Committee are (1) to coordinate and promote sharing of

strong motion earthquake data among researchers and practicing engineers, and (2)

to develop techniques and exchange information for evaluating the destructive effects

of earthquake motion.

The scope of work includes:

(1) instrumentation,

(2) recording, processing and analyzing strong-motion data,

(3) engineering characterization of ground motion,

(4) design applications, and

(5) seismic zonation.

The activities of the Task Committee include:

(1) regular exchange of data and publications,

(2) creating procedures for the dissemination of significant strong-motion

digital data with regard for the rights and expectations of (a) owner, (b)

the users of data and (c) the earthquake engineering community,

(3) planning and conducting T/C workshops and meetings, and

(4) coordinating relevant research activities.

649
2. ACCOMPLISHMENTS

(1 ) Exchange of Data

From U.S. Side:

• Catalogs of Earthquakes in the U.S.

• Three sets of slides for the M6.7 1 994 Northridge earthquake: epicentral area,

far-field, and a comparison with the M6.6 1971 San Fernando earthquake,

produced by the World Data Center A and NGDC/NOAA, provided by H.

Meyers.

A set of 2 CD's of digital slides on Geologic Hazards, (volcanoes, tsunamis, and

earthquakes), produced by NGDC/NOAA from slides from various sources,

provided by H. Meyers.

From Japan Side:

° Catalogs of Earthquakes in Japan

• Floppy disk containing accelerograms from the 1993 Off-Kushiro earthquake,

recorded by the Building Research Institute, M.O.C., and provided by Y.

Kitagawa and T. Kashima.

(2) International Workshop

Conducted a successful International Workshop on Strong Motion Data, December

1993, Menlo Park, California. The workshop dealt with the current state of strong

motion recording in the U.S., Japan, and the rest of the world, and compared

attenuation relations for peak ground accelerations and response spectra. It was

hosted by PHRI and USGS, and sponsored by the Science and Technology Agency,

the International Science and Technology Exchange Center, and PHRI, Japan. The

published Proceedings contain 36 contributions from US, Japanese and other

researchers (from New Zealand, France, and Chile).

(3) Recordings of the Northridge Earthquake

The constant promotion of high-quality strong motion recording networks throughout

the world has assisted in the production of useful sets of data from the Northridge

Earthquake (on the order of 200 from California's state program (CDMG/SMIP), 100

from USGS, and 60 from the University of Southern California).

650
3. FUTURE PLANS

An international workshop on strong-motion data concentrating on local site effects

is planned for early 1996 in Japan. Momentum from the December 1993 workshop

and the large amount of Northridge data will be well-used. Topic detail includes

regional geology, topography, local site geology, amplification, non linearity, and

liquefaction. Invited attendance of 30-50 is planned, including participants from

countries other than the U.S. and Japan. Financial support from international and

U.S. sources, and the Japanese Government, will be sought.

651

Report of Task Committee on

(B) Testing and Evaluation Procedures for Building Systems

Date : May 18, 1994

Place : National Institute of Standards and Technology

Gaithersburg, MD

Attendees U.S. Side - H

S

Lew (Chairman)

Mahin

(NIST)

(UCB)

Japan Side - K

H

H

Ohtani (Chairman)

Okada

Shiohara

(NIED)

(BRI)

(BRI)

1. OBJECTIVE AND SCOPE OF WORK

The objective of the Task Committee is to develop rational test procedures and to

collect performance data of dynamic response of structures through laboratory testing

of prototype structures and field testing of structures in situ.

The scope of work includes:

(1) Plans and conducts workshops and joint meetings to identify research topics

and develops joint research programs.

(2) Coordinates research projects carried out by various laboratories in the U.S. and

Japan. Facilitates publication of research results and implementation of

findings in codes and standards.

(3) Facilitates exchange of research personnel, technical information and available

testing facilities.

(4) Develops uniform testing procedures including test loading history for case of

comparison of results of tests carried out by various researchers, and for

establishment of data base.

(5) Develops guidelines for interpretation of test results in consideration for design

of structures.

(6) Develops methodology for evaluation and interpretation of physical test results.

653
2, ACCOMPLISHMENTS

(1) Conducted the fourth U.S. -Japan Joint Technical Coordinating Committee

(TCC) meeting on Precast Seismic Structural Systems Program (PRESSS), May

1994, Tsukuba, Japan.

(2) Published seismic design and construction guidelines and design manuals for

precast reinforced concrete buildings by Building Center of Japan, July 1993.

(3) Initiated the Japan research program on Composite and Hybrid Structures

(CHS), 1993. The Japan TCC meetings were held during the past year.

(4) The Panel endorsed the implementation of the joint research program on CHS

during the 25th Joint Panel Meeting in May 1993.

(5) Conducted the first U.S. -Japan Joint TCC meeting on CHS in November 1 993,

Tsukuba, Japan.

3. FUTURE PLANS

(1) The third phase of the U.S. PRESSS has been planned to be initiated during

1994-1995. The fourth U.S. meeting will be held in October 1994.

(2) The second U.S. -Japan joint TCC meeting on CHS will be held in July 1 994 in Chicago, Illinois during the occasion of the 5th U.S. National Conference on

Earthquake Engineering.

(3) The joint research program on CHS will be formally initiated in the U.S. in 1994. The first U.S. TCC meeting will be held in the fall of 1994.

(4) In support of the International Decade for Natural Disaster Reduction (IDNDR)

Program, techniques will be explored to disseminate finding of the Joint

Research Projects carried out under the auspices of T/C "B" to countries with

high seismic risks.

(5) To maximize the usefulness of experimental data, it is desirable to develop

standards for storage, documentation and retrieval of test data. Researchers

and members of the user community from the U.S. and Japan should be

encouraged to convene for the purpose of establishing such standards. Efforts

should be made to utilize available resources and expertise in Japan and the

U.S.

654
Report of Task Committee on

(C) Design Evaluation and Improvement of Structures

Date : May 18, 1994

Place : National Institute of Standards and Technology

Gaithersburg, MD

Attendees : U.S. Side - K. Chong (Chairman) (NSF)

C. Anderson (USBR)

R. Fuller (HUD)

J. Hill (DOE)

L. Hultengren (DOS/FBO)

J. Scalzi (NSF)

A. Taylor (NIST)

J. Wilcoski (COE)

Japan Side - H. Okada (Acting Chairman) (BRI)

M. Kanda (PWRI)

S. Kobayashi (PWRC)

M. Koga (CRS)

K. Nishikawa (PWRI)

H. Shiohara (BRI)

K. Yokoyama (PWRI)

1. OBJECTIVE AND SCOPE OF WORK

The objective of the Task Committee is to develop disaster mitigation policies and

programs which will improve the performance of new structures and the capacity of

existing structures in resisting wind and seismic forces. In order to establish adequate

evaluation of performance, each country will coordinate development of condition

assessment, screening, and structural analysis and design methodologies. Enhance

performance can be obtained by the intelligent renewal, repair and retrofit of existing

buildings and other structures. Structures will be analyzed and instrumented, and

then evaluated after disasters. Also, effective utilization of new advanced materials

and structural systems, associated with appropriate design methods, will provide

excellent structural performance in the life cycle.

655
The scope of work includes:

(1) Exchanging information, planning, and hosting workshops on new design

technologies, evaluation technologies, and repair and retrofit techniques.

(2) Developing a uniform system for screening and analyzing wind and seismic

resistance capacity of structures in each country.

(3) Developing sensor technology, instrumentation, and "expert" systems to

provide reliable condition assessment of existing structures.

(4) Evaluating masonry, precast concrete and composite/hybrid building structures;

and providing documentation for post-disaster evaluation of performance.

(5) Studying new materials and methods for repairing and retrofitting existing

structures and for new construction.

(6) Coordinating research projects in the U.S. and Japan to minimize duplication

and maximize benefits.

2. ACCOMPLISHMENTS

(1 ) T/C "C" and "G" co-sponsored a workshop on Structural Control and Intelligent

Material Systems in May, 1 993 in Japan. Proceedings of this workshop will be

available soon.

(2) A member of Task Committee C participated in the EERI Reconnaissance Team

(Dr. A.G. Brady) which investigated the Guam Earthquake of August 8, 1993.

(3) After the January 1994 Northridge Earthquake, several U.S. agencies

conducted cooperative damage assessment studies to determine the necessity

to improve building codes and to develop repair and strengthening requirements

for existing buildings. (HUD, USGS, NIST, FEMA, NSF)

(4) BRI has sent to the U.S. side the proceedings of the second U.S. -Japan Study

Group Workshop on New Design Methodology for Building, which was

published by BRI. Although the workshop consisted of the U.S. and Japan Ad

Hoc Committees, the activities will lead to an official U.S. /Japan joint study

group that is recommended in the item (3) of "Future Plans."

3. FUTURE PLANS

The following planning was performed on Design, Evaluation and Improvement of

Structures.

656
Develop a uniform system for screening and analyzing wind and seismic

resistant capacity, and improve sensor/inspection technology, instrumentation

and expert systems to provide condition assessment of existing structures.

Compile a database of advanced materials/systems that have potential for

improving structural performance of new construction, as well as for use in rehabilitating and strengthening existing buildings.

Investigate new design methodologies based on required structural

performance, which also will address advanced materials and systems. In order

to effectively take this new approach, it is recommended that each country

should make an effort to establish a U.S. /Japan joint study group of members

of appropriate institutions in each country.

Develop design guidelines for applying advanced structural and materials

systems to be used for composite and hybrid structures under the joint

U.S. /Japan Cooperative Research Program.

Participate in activities of the International Decade for Natural Disaster

Reduction (IDNDR), including the "World Conference on Natural Disaster

Reduction," held in Yokohama, Japan during May 23-27, 1994.

Continue to encourage participation by private industry, consulting engineers,

universities, State and local government agencies, and other Federal agencies

involved in instrumentation, evaluation, condition assessment, and retrofit and

strengthening of existing buildings for seismic and wind resistance.

The second workshop on "Structural Control and Intelligent Material Systems"

is scheduled in the U.S. during 1995, in coordination with T/C "G".

OTHER ACTIVITIES

MOC delegates visited NSF in September and December 1993. An agreement

was reached to promote research programs on "Building Safety and Related

Infrastructure Improvements" under the framework of the "New Economic

Partnership between the U.S. and Japan".

A workshop on Composites for Construction will be held at NSF - July 8, 9, 10,

1994, to plan a 5-year research program on the basic research topics required

for the use of composites in general construction.

A U.S. -Japan Workshop on "Development and Future Directions in Structural

Testing" co-chaired by T. Okada and S. Mahin, held in Honolulu, Hawaii in June

1993.

657
(4) Continued U.S. research sponsored by NSF has developed reports on seismic

rehabilitation and retrofit strategies, including innovative techniques for

strengthening existing buildings and bridges.

(5) NSF completed panel review of the 3rd year "Repair and Rehabilitation for

Earthquake Resistance" proposals and will announce awards in the near future.

(6) A new and improved Seismic Safety Guide is to be published by the

Department of Energy in 1994, Donald Eagling, Editor. The "Guide" provides

information for engineers and building managers for developing practical and

cost-effective seismic mitigation programs for application at sites with several

buildings.

(7) Benefits versus costs for retrofitting existing buildings has been studied at the

Lawrence Berkeley Laboratory in California. A paper to be completed soon

shows benefit/cost ratios saving from 25 to 1 to 100 to 1.

(8) "Second International Conference on Intelligent Materials" co-funded by NSF

and other U.S. agencies, will be held June 1994 in Williamsburg, VA.

(9) Continue development of NSF's Civil Infrastructure Systems Research Program,

including strong emphasis on evaluation and improvement of structures.

658
Report of Task Committee on

(D) Earthquake Engineering for Dams

Date : May 18, 1994

Place : National Institute of Standards and Technology

Gaithersburg, MD

Attendees : U.S. Side - W. Roper (Co-Chairman)

R. Hall

(COE)

(WES)

(COE)

(USBR)

(WES)

(CERL)

L. Guthrie

F. McLean

A. Franklin

J. Wilcoski

Japan Side K. Nishikawa (Acting Chairman) (PWRI)

1. OBJECTIVE AND SCOPE OF WORK

The objective of the Task Committee is to develop technical insights into better

understanding of the response of dams to seismic effects. The T/C will plan,

promote, and develop research initiatives to assist in assuring seismic safety and

economical protective countermeasure against earthquake loading of these structures.

The scope of work includes:

(1) Methods of analysis for seismic design of dams. Compare U.S. and Japan

design methods. Perform investigations using dynamic analysis methods as

tools (modeling, calculation codes).

(2) Dynamic characteristics of dam construction materials. Identify strength and

deformation characteristics during an earthquake (concrete, soil and rock).

(3) Analysis of observed behavior of dams during an earthquake. Investigate the

mechanism of damages to earthquake loading. Analyze the observed behavior

during an earthquake to the earthquake-resistant design.

2. ACCOMPLISHMENTS

Mr. Nario Yasuda, senior research engineer of Fill-Type Dam Division, PWRI, visited

Research and Development Directorate, Headquarters U.S. Army Corps of Engineers

on January 28, 1994, to discuss future possible activities of T/C (D), feasibility of

cooperative research, exchange of technical information and so on.

Dr. Dean Norman, Senior Research Structure Engineer of the Structures Laboratory,

659
WES, visited the Earthquake Engineering Division, PWRI during February and March

1 994. He conducted cooperative studies on retrofit techniques for bridges and other

concrete structures, and concrete constitutive models for dynamic finite element

applications. During his stay he attended the International Workshop on Construction

and Management Technology for Dams, and Enhancement of the Dam Lake Basin

Environment for the Next Generation in Tsukuba.

3. FUTURE PLANS

(1 ) Exchange results of research/investigation and technical information about dam

earthquake engineering is encouraged, and feasibility of joint research and

investigation is in continuous consideration.

(2) Exchange visits to the institutes concerned, of scientists and engineers

between U.S. and Japan, to enhance effective communication.

(3) In accordance with the progress of activities (1 ) and (2), T/C (D) will coordinate

a Workshop on Earthquake Engineering for Dams and the development of a joint

publication on the current status of research in this area.

660
Report of Task Committee on

(F) Disaster Prevention Methods for Lifeline Systems

Date : May 18, 1994

Place: National Institute of Standards and Technology

Gaithersburg, Maryland

Attendees : U.S. Side - R. Chung (Chairman)

S. Mahin

(NIST)

(UCB)

Japan Side - K. Kawashima (Acting Chairman)

C. Minowa

H. Sugita

(PWRI)

(NRIESDP)

(PWRI)

1. OBJECTIVE AND SCOPE OF WORK

The objectives of the Task Committee are to improve: 1) the performance of lifeline

systems during earthquakes and 2) engineering and other seismic countermeasures

including the capability in damage estimation techniques and inspection procedures.

The scope of work includes:

(1) Planning and conducting workshops on lifeline systems

(2) Facilitating exchange of technical information

(3) Promoting development of design guidelines and standards for lifeline systems

2. ACCOMPLISHMENTS

(1) Published the proceedings of the 5th Joint Workshop on Earthquake Disaster

Prevention for Lifeline Systems, which was held in October 1992 in Tsukuba,

Japan. The proceedings were distributed to members of the UJNR panel and

others in the lifeline earthquake engineering community.

(2) Organized a joint post-earthquake reconnaissance effort for the July 1993

Hokkaido-Nansei-Oki earthquake. The U.S. team members collaborated closely

with the Japanese team members in their field investigations.

Results of their joint studies, especially those related to the performance of

Lifeline systems, have been presented in a number of meetings, workshops,

and conferences.

661
Assisted in the arrangement of the Japanese Government Study Team's visits

to Los Angeles area in January and March, 1994 after the January 17, 1994

Northridge earthquake. The Japanese team has visited a number of sites where

critical lifeline systems, such as aqueducts, substations, to receive briefings and

assess their performance.

FUTURE PLANS

Encourage current efforts in both countries to establish seismic design

guidelines and standards for lifeline systems. Existing UJNR channels should

be fully utilized to facilitate the exchange of relevant information concerning

standards development. Possible collaboration of joint development of

guidelines and standards for lifeline systems should be pursued.

Encourage collaborative researchers and development in the following areas:

planning for urban disaster prevention

systems approach to lifeline performance

soil-structure interaction

performance of underground pipelines and cables

fire protection/prevention

geographic information systems (GIS)

cost-effective ground improvement techniques

development and operation (including restarting) of safety shut-down

systems

Hold the 6th Joint Workshop on Earthquake Disaster Prevention for Lifeline

Systems, December 1994, Los Angeles, CA. The Workshop focus will be on

soil-structure interactions, performance of underground pipelines and cables,

and cost-effective ground improvement techniques.

662
Report of Task Committee on

(G) Structural Control and Intelligent Material Systems

Date : May 17, 1994

Place : National Institute of Standards and Technology

Gaithersburg, MD

Attendees : U.S. Side - K. Chong (Acting Chairman) (NSF)

C. Anderson (USBR)

R. Fuller (HUD)

J. Hill (DOE)

L. Hultengren (DOS/FBO)

J. Scalzi (NSF)

A. Taylor (NIST)

J. Wilcoski (COE)

Japan Side - H. Shiohara (Acting Chairman) (BRI)

M. Kanda (PWRI)

S. Kobayashi (PWRC)

M. Koga (CRS)

K. Nishikawa (PWRI)

H. Okada (BRI)

K. Yokoyama (PWRI)

1. OBJECTIVE AND SCOPE OF WORK

The objectives of the Task Committee are to: (1) develop research plans in control

of equipment and structures and in high performance structural and material systems;

(2) implement control techniques for motion reduction or modification; (3) implement

use of advanced materials in actual design and construction of buildings and other

infrastructure systems under seismic or wind environments; (4) promote U.S. -Japan

cooperation in structural control and intelligent material systems research; (5) bring

together governmental, academic, and industrial participants in joint pursuit of these

efforts; and (6) contribute to IDNDR by organizing joint research and other technical

activities in structural control and intelligent material systems research on the basis

of international cooperation.

The UJNR T/C "G" works closely with other organization to provide the leadership in

this emerging research by facilitating the exchange of technical data and information

through UJNR mechanisms. The scope of work includes:

663
(1 ) Providing technical assistance, consultation and coordination to UJNR affiliated

research organizations in the initiation, development, and execution of their

programs in research areas.

(2) Promoting joint government-university-industry collaborative efforts to facilitate

technology transfer and practical implementation.

(3) Sponsoring and conducting inter-disciplinary workshops and meetings to

identify key areas of research and opportunities for cooperation, and to

exchange new knowledge and experience in practice.

(4) Developing promotional and demonstrative activities to stimulate public

awareness and interest in this field of research.

(5) Providing information useful for the establishment of performance standards,

design specification, guidelines and code recommendations for application in new construction as well as retrofit/rehabilitation of existing structures.

(6) Initiating research in intelligent material systems, sensors, actuators, optimal

control system design, and encourage laboratory and field experiments of

prototype and full-scale structures.

2. ACCOMPLISHMENTS

(1) Conducted the "3rd U.S Japan Workshop on Earthquake Protective Systems

of Bridges" at Berkeley, California during January 24 and 25, 1994 under the

joint auspices of T/C "G" and "J". A study tour to visit base-isolated bridges

in the San Francisco area was conducted by participants prior to the Workshop.

Proceedings of the Workshop was published by NCEER in March 1994. The

workshop was attended by 45 U.S., 32 Japanese researchers, and experts

from Taiwan and New Zealand.

(2) To pursue the technical objectives of the T/C "G" and T/C "C", a U.S. domestic

workshop was held in Williamsburg, Virginia, on November 17 and 18, 1993

to further define the objectives and research value and needs in intelligent

materials/systems. Published the workshop proceedings, "Rebuilding and

Enhancing the Nation's Infrastructure: A Role for Intelligent Material Systems

and Structures;" it was distributed by the Workshop organizer.

(3) Joint Research Program for Hybrid Control of Bridges is being performed

through T/C "G" and "J" by PWRI, NSF, FHWA, NIST, NCEER, and UC Irvine.

Under this program, experimental studies have been conducted at PWRI and

NCEER (SUNY at Buffalo). The experimental studies at PWRI tested and

confirmed the effectiveness of the sliding base-isolation systems developed at

NCEER.

664
(4) International Workshop on Structural Control was held in Honolulu, Hawaii on

August 4-5, 1 993, with major participation by Japan and U.S. Panel Members.

Proceedings was published by the University of Southern California in December 1993.

(5) NIST developed standard test procedures for seismic isolation systems were

presented at the 26th UJNR Joint Meeting.

(6) At BRI, research programs on structural control of buildings have been

developed and implemented. Collaborative research programs involving

possible organizations and private companies were developed. The final report

was published by BRI in 1993.

3. FUTURE PLANS

(1) Planning efforts for U.S. -Japan cooperative research projects on seismic

response control involving PWRI, BRI, NSF and industries from both sides are continuing. Research cooperation between the U.S. and Japan will be actively

pursued using forums such as the UJNR and other inter-governmental

mechanisms.

(2) Joint research projects on active and hybrid control of structures, such as bridges and buildings, will be continued and developed by PWRI, BRI, NCEER,

CUREe, and others.

(3) In cooperation with T/C "J", the 4th US-Japan Workshop on Earthquake

Protective Systems for Bridges is being scheduled in 2 years in Japan. Topics

are expected to include: (1 ) innovative protective systems; (2) design methods

for seismic isolated/menshin bridges; (3) full-scale verification of performance;

and (4) comparative design examples.

(4) Under the Joint Research Program for Hybrid control of Bridges through T/C

"G", the experiment to confirm the effectiveness of variable damper systems

developed at PWRI will be performed at NCEER.

(5) A Workshop on "Structural Control and Intelligent Material Systems" is planned

for 1995 in cooperation with T/C "C" of the Panel on Wind and Seismic

Effects.

4. OTHER ACTIVITIES

(1) The NSF Research on Structural Control will hold a coordination meeting in Pasadena, on August 6-7, 1994.

(2) The First World Conference on Structural Control will be held in Pasadena,

California on August 3-5, 1994.

665
"Second International Conference on Intelligent Materials" co-funded by NSF

and other U.S. agencies, will be held in June 1994 in Williamsburg, VA.

The U.S. Panel on Structural Control Research worked closely with one of its counterpart "Japanese Panel on Structural Response Control" in exchanging

technical information and developing joint research plans.

666
Report of Task Committee on (H) Soil Behavior and Stability During Earthquakes

Date : May 17, 1994

Place : National Institute of Standards and Technology

Gaithersburg, Maryland

Attendees: U.S. Side - A. Franklin (Chairman)

F. McLean

(WES)

(USBR)

Japan Side K. Kawashima (Acting Chairman)

S. lai T. lijima

T. Kagawa

C. Minowa

H. Sugita

(PWRI)

(PHRI)

(PWRI)

(WSU)

(NIED)

(PWRI)

1. OBJECTIVE AND SCOPE OF WORK

Government agencies responsible for public works must assure seismic safety and

provide economical protection against earthquake hazards.

The objective of the Task Committee is to assist in meeting these needs by enhancing

the available technology for predicting the dynamic behavior of soils, analyzing

dynamic soil-structure interaction, and modifying the earthquake behavior of

foundations and earth structures to assure their safe performance during earthquakes.

In accordance with the objective, the scope of work includes:

(1) Present technical papers at annual joint panel meetings on technological

developments and on state-of-the-art and practice related to soil behavior and

stability during earthquakes.

(2) Exchange information and technical data relating to field performance, research,

and methods of practice.

(3) Plan and conduct programs of cooperative research and/or workshops in coordination with the proposed or ongoing programs.

(4) Make other needed efforts including exchange of researchers between U.S. and

Japanese research institutions, publication of research results and

recommended practice.

667
2. ACCOMPLISHMENTS

(1 ) The Japan-side provided to the U.S. -side the "Handbook on Remedial Measures

for Reclaimed Land Soil Liquefaction," published in Japanese by the Ministry

of Transport. The Waterways Experiment Station (WES) has translated the

document into English and has provided a copy of the manuscript to the Japan

side for review prior to publication.

(2) The National Research Institute for Earth Science and Disaster Prevention

(NIED), the Building Research Institute (BRI), and Wayne State University

(WSU), have conducted a cooperative research program on "Research on

Earthquake Pile Resistance in Liquefied Soils" (1992-1993) under the auspices

of UJNR Panel on Wind and Seismic Effects and have reported results at the

26th Joint meeting. Dr. C. Minowa and Dr. N. Ogawa, NIED, visited WSU, and

Professor T. Kagawa visited NIED in connection with this cooperative research.

3. FUTURE PLANS

(1) T/C (H) will hold its 4th U.S. -Japan Workshop on Remedial Treatment of

Potentially Liquefiable Soils at the Public Works Research Institute (PWRI) in Tsukuba, 4-8 July 1994.

(2) The Port and Harbour Research Institute (PHRI) and WES will publish the

English version of the "Handbook on Remedial Measures for Reclaimed Land

Soil Liquefaction" and deliver it to the panel members and participating

organizations in 1995.

(3) NIED and WSU will perform a cooperative research program "Physical and

Numerical Simulation of Structural Damages Due to Liquefaction and

Development of CountermeasureTechniques" (1 994-1 998) underthe auspices

of UJNR Panel on Wind and Seismic Effects.

(4) T/C (H) tentatively plans to hold a workshop on Earthquake Simulation in Centrifuges during 1996.

668
Report of Task Committee on

(I) Storm Surges and Tsunami

Date: May 18, 1994

3 lace: National Institute of Standards and Technology

Gaithersburg, MD

Attendees: U.S. Side - H. Meyers (Chairman)

M. Blackford

W. Roper

J. Wilcoski

(NOAA)

(NOAA)

(COE)

(COE)

Japan Side H. Sato (Acting Chairman)

S. lai (PWRI)

(PHRI)

1. OBJECTIVE AND SCOPE OF WORK

The objective of the Task Committee is to mitigate damages from storm surges and

tsunamis through shared technologies, research, information sharing, and cooperative

work. Storm surges are caused by hurricanes and typhoons. Tsunamis are caused

by underwater earthquakes, volcanic activities and landslides. Both hazards may

cause serious disasters along coastal region.

The scope of work includes:

(1) Exchange results of research on storm surge and tsunami occurrence,

generation, propagation, and coastal effects. This includes observations on

historical, current, and theoretical tsunamis. Of particular interest is the effort

by U.S. and Japan to acquire deep ocean tsunami measurements.

(2) Exchange results and status of countermeasures against storm surge and

tsunami activities including analysis of the problem, planning, warning, and

engineering approaches.

(3) Exchange information on development of technologies such as computer

programs to predict travel times, land-fall locations, run-up heights, and wave

characteristics, improved instrumentation, and use of satellite communications

for detection and warning.

(4) Facilitate dissemination of scientific and technical information through

exchanges of literature, technical reports at joint meetings, special workshops,

joint projects, and direct interaction among participants.

669
2. ACCOMPLISHMENTS

(1) A joint field investigation was performed during 17-24 July 1993 on the

damage caused by the 12 July 1993 M7.8 Hokkaido Nansei-oki

Earthquake/Tsunami.

(2) Findings from the investigations on the Tsunami caused by the 12 July 1993

Hokkaido Nansei-oki Earthquake were reported in the 26th Joint Panel Meeting

by U.S. and the Japan-side members.

(3) Conducted the Third Tsunami Workshop in Osaka, Japan on 28 August 1 993.

(4) In the continuing exchange of publications the U.S. -side provided the Japanside with a copy of Tsunamis Affecting the U.S. West Coast 1806-1992 and

a set of slides of tsunami damage resulting from the recent tsunamis in

Nicaragua, Indonesia, and Japan.

3. FUTURE PLANS

(1 ) Planning is proceeding to conduct a joint T/C "I and K" Workshop on Wind and

Earthquake Engineering for Offshore and Coastal Areas during January 1995

in the United States.

(2) Continue exchanging data and information on tsunami measurements including

deep ocean measurements.

(3) 1996 will be the centennial anniversary of the great Sanriku earthquake and

tsunami in Japan and the 50th anniversary of the April 1, 1946 Aleutian

earthquake and tsunami. These events led to the establishment of warning

system and an increased awareness of the damages wrought from these

events. The Task Committee agreed to plan a Fourth Workshop to be held in Hawaii in 1996 in observance of these natural disasters.

(4) Publish the proceedings of the 3rd Tsunami Workshop.

4. OTHER ACTIVITIES

(1) The storm surge research community is less well organized and lacks the

specialty societies, journals, and international organizations to define and

stimulate work in the field. The Task Committee's "I" mission and interests

can help fill this gap.

(2) The tsunami problem is less severe in the U.S. than in Japan which leads to a smaller community of researchers. There are few U.S. researchers focusing on

tsunamis at any specific location; thus highlighting the importance of tsunamis

at T/C "I" meetings and workshop are important to transfer information.

67d
Report of Task Committee on

(J) Wind and Earthquake Engineering for Transportation Systems

Date : May 19, 1994

Place : National Institute of Standards and Technology

Gaithersburg, MD

Attendees: U.S. Side - J. Cooper (Chairman)

A. Franklin

A. Taylor

J. Scalzi

P. Yen

(FHWA)

(COE)

(NIST)

(NSF)

(FHWA)

Japan side K. Yokoyama (Chairman) (PWRI)

K. Nishikawa (PWRI)

K. Kawashima (PWRI)

M. Kanda (PWRI)

1. OBJECTIVE AND SCOPE OF WORK

The objectives of the Task Committee are to: (1 ) plan, promote, and foster research

on the behavior of highway bridges when subjected to wind and seismic forces and

(2) disseminate research results and provide specifications and guidelines based on

the Task Committee's findings. Surface transportation systems play a vital role in the

movement of goods and people. Highway bridges are especially influenced by the

forces of wind and earthquakes because of their open exposure to those forces.

The scope of work includes:

(1 ) Focus research on highway bridges without limitation on their size and function.

(2) Investigate existing and new bridge designs and the behavior of whole bridge

systems and/or single components of a bridge.

2. ACCOMPLISHMENTS

(1 ) Held the Tenth U.S. -Japan Bridge Workshop in Lake Tahoe, Nevada, May 10-

11, 1994. Fifty-eight participants representing government, academia,

industry, and private practice presented 39 papers in 8 technical sessions.

Technical site visits followed to gain knowledge about testing facilities at the

University of Nevada, Reno; and to study ongoing bridge construction, and

seismic retrofitting at bridge sites in the Seattle, Washington and the San

671
Francisco Bay Area.

(2) Held the Second U.S. -Japan Workshop on Seismic Retrofit of Bridges at

Berkeley, California, January 20-21, 1994. Technical site visits after the

workshop included bridge construction, retrofit, and base isolated bridges in the

San Francisco Bay Area. The site visits were held in conjunction with the third

U.S. -Japan workshop on Earthquake Protective Systems for Bridges. (NOTE:

Special trips were made to Los Angeles to document bridge damage from the

January 17, 1994, Northridge Earthquake.)

(3) Conducted, in cooperation with Task Committee (G), the third U.S. -Japan

workshop on Earthquake Protective Systems for Bridges at Berkeley, California,

January 24-25, 1994. Technical site visits to bridge construction, retrofit and

base isolated bridges in the San Francisco Bay Area preceded the workshop.

3. FUTURE PLANS

(1 ) Conduct the 1 1th U.S-Japan Bridge Engineering Workshop in Tsukuba, Japan,

in conjunction with the 27th Joint Panel meeting in May 1995.

(2) Initiate plans to conduct the Third U.S. -Japan Workshop on Seismic Retrofit of

Bridges in 1996 in Japan. The workshop will focus on practical seismic retrofit

measures.

(3) In cooperation with Task Committee (G), initiate plans to conduct the Fourth

U.S. -Japan Workshop on Earthquake Protective Systems for Bridges in Japan.

The workshop will focus on research and application including: innovative

protective systems; design methods; full-scale verification of performance; and

comparative design examples.

(4) Continue to investigate and exchange technical information on improved

seismic retrofit and strengthening procedures for highway bridges based on

experimental, analytical, and field studies. This exchange should include

information on maintenance of existing bridges.

(5) Continue experimental research on the seismic performance of bridge piers and

columns, and encourage research on base isolation and hybrid control of

bridges in cooperation with Task Committee (G).

(6) Develop a coordinated research study on seismic, aeroelastic, and aerodynamic

response of cable-supported bridges with emphasis on behavior of composite

materials, cable inspection, vibration control, and corrosion protection.

(7) Continue the coordinated research study to compare the seismic design criteria

for bridges in Japan and the U.S. and discuss the method and analysis

procedures for bridge column design. Continue information exchange on the

672
application of the limit state design method.

Encourage a coordinated research study on seismic response and control,

system identification techniques, nondestructive evaluation of bridge structures,

use and performance of structural materials including new materials, and

performance of jointless bridges.

673

Report of Task Committee on (K) Wind and Earthquake Engineering for Offshore

and Coastal Facilities

Date : May 18, 1994

Place

:

National Institute of Standards and Technology

Gaithersburg, Maryland

Attendees: U.S. Side - W. Roper (Acting Chairman)

J. Wilcoski

(COE)

(COE)

Japan Side S. lai (Acting Chairman)

M. Kanda

H. Sato

(PHRI)

(PWRI)

(PWRI)

1. OBJECTIVE AND SCOPE OF WORK

Criteria for the design of offshore and coastal facilities may differ greatly from their

onshore counterparts. These differences can arise due to: unique design of mass

distribution; fluid/structure or wind/structure interaction; placement of foundation

elements in or on soft fully saturated soils that can be subject to large hydrodynamic

pressures; and lack of specific environmental data or the engineering experience that

has been developed for most onshore sites.

The objective of the Task Committee is to develop technical insights necessary to

mitigate damage to offshore and coastal facilities due to extreme wind and seismic

effects. The Task Committee will plan, promote and develop research initiatives to

meet this objective, and will disseminate the results of their research for incorporation

into future specification or design guidelines.

The scope of work includes:

(1) Sponsoring and conducting workshops and meetings to identify key area

research, opportunities for cooperation, and the exchange of knowledge.

(2) Predicting strong ground motions for offshore and coastal sites including

assessing the effects of basin geometry, linear, and nonlinear local geological

effects using actual seafloor response measurements.

(3) Determining the dynamic response and the interaction of

structure/foundation/soil systems to seabed motions and/or extreme wind

forces.

675
(4) Assessing the dynamic response and behavior of various operational facilities

mounted on offshore and coastal structures.

(5) Developing assessment methodologies for seismicity and other characteristics

of potential seismic sources (e.g. faults) for offshore and coastal sites in regards to how these conditions relate to structural design criteria.

(6) Promoting the implementation of new research results into current design and

construction process.

(7) Developing research efforts to include laboratory and field programs to obtain

data on the response of offshore and coastal facilities to extreme wind and

seismic forces.

(8) Creating performance standards, design specifications, guidelines , and code

recommendations for application to new construction as well as remedial action

for existing facilities.

2. ACCOMPLISHMENTS

(1 ) Conducted the 1 st U.S. -Japan Workshop on Wind and Earthquake Engineering

for Offshore and Coastal Facilities at the PHRI, Yokosuka, Japan during May 1

and 13, 1993.

3. FUTURE PLANS

(1) Hold the 2nd U.S. -Japan Workshop on Wind and Earthquake Engineering for

Offshore and Coastal Areas in the U.S. in January 1995. This will be a joint

workshop with Task Committee (I) Storm Surges and Tsunami.

(2) Develop workshop agenda and finalize list of speakers from the U.S. and Japan

Side.

(3) Publish the Proceedings of the first Workshop; it will be distributed by NIST.

(4) Coordinate, where possible, on-going research on wind and earthquake

engineering for offshore and coastal facilities of interest to the members of the

Task Committee.

<T US GOVERNMENT PRINTING OFFICE: 1994— 386-627 /22782

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