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OF

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NBS SPECIAL PUBLICATION 560

U.S. DEPARTMENT OF COMMERCE/National Bureau of Standards

National Bureau of Standards

library, E-01 Admin. Bfdg.

oct i m

131079

WIND AND SEISMIC EFFECTS

Proceedings of the

Tenth Joint UJNR

Panel Conference
NATIONAL BUREAU OF STANDARDS

The National Bureau of Standards' was established by an act of Congress on March 3, 1901. The Bureau's overall goal is to strengthen and advance the Nation's science and technology

and facilitate their effective application for public benefit. To thi:. end, the Bureau conducts

research and provides: (1) a basis for the Nation's physical measurement system, (2) scientific and technological services for industry and government, (3) a technical basis for equity in trade, and (4) technical services to promote public safety. The Bureau's technical work is per- formed by the National Measurement Laboratory, the National Engineering Laboratory, and

the Institute for Computer Sciences and Technology.

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Headquarters and Laboratories at Gaithersburg, MD, unless otherwise noted; mailing address Washington, DC 20234. 'Some divisions within the center are located at Boulder, CO 80303.
WIND AND SEISMIC EFFECTS

national Bureau of Staotett

library. E-01 A<Jmln. BW|.

Proceedings of the Tenth Joint OCT 2 9 1980

Panel Conference of the U.S.-Japan , r , r> r „ Cooperative Program in

Natural Resources

May 23-26, 1978 HO. ShO

National Bureau of Standards fQ gfi

Gaithersburg, Maryland

H.S. Lew, Editor

Center for Building Technology

National Engineering Laboratory

National Bureau of Standards

Washington, D.C. 20234

U.S. DEPARTMENT OF COMMERCE, Philip M. Klutznik, Secretary

Luther H. Hodges, Deputy Secretary

Jordan J. Baruch, Assistant Secretary for Productivity, Technology and Innovation

NATIONAL BUREAU OF STANDARDS, Ernest Ambler, Director

Issued October 1980
Library of Congress Catalog Card Number: 79-600134

National Bureau of Standards Special Publication 560

Nat. Bur. Stand. (U.S.), Spec. Publ. 560, 644 pages (Oct. 1980) CODEN: XNBSAV

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1980

For sale by t lie Superintendent of Doniments, U.S. (iovcnvmenl Printing Office Washington, D C. 20402 - Price $11
PREFACE

The Tenth Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects was

held at the National Bureau of Standards in Gaithersburg , Maryland on May 23-26, 1978.

This panel is a part of the U.S. -Japan Cooperative Program in Natural Resources (UJNR)

,

The panel meets annually, alternating between Washington and Tokyo to exchange technical information on the latest research and development activities within governmental

agencies of both countries in the area of wind and seismic effects.

These proceedings include the membership list, the formal resolution, the technical

papers presented at the joint meeting, and the reports of the task committees.

It should be noted that throughout the proceedings certain commercial equipment,

instruments or materials are identified in order to specify adequately experimental

procedure. In no case does such identification imply recommendation or endorsemement

by the National Bureau of Standards, nor does it imply that the material or equipment

identified is necessarily the best for the purpose.

H. S. Lew, Secretary

U.S. Panel on Wind and

Seismic Effects

iii
ABSTRACT

This volume includes thirty eight technical papers presented at the Tenth Joint Meeting

of the U.S. -Japan Panel on Wind and Seismic Effects. It also includes the formal resolutions

and the reports of the Panel's eight task committees. The subjects covered in the Joint

Meeting include: (1) natural wind characterization and extreme wind records, (2) characterization of earthquake ground motions and strong-motion earthquake data, (3) engineering

seismology, (4) response of hydraulic and earth structures to seismic forces, (5) structural responses to wind loading, (6) recent developments in seismic design criteria,

(7) design and analysis of special structures, (8) damage evaluation, repair and retrofit,

(9) earthquake hazard mitigation, and (10) storm surge and tsunami.

Key Words: Accelerograph ; codes; design criteria; disaster; earthquakes; ground motion;

seismology; standards; structural engineering; storm surge; structural response;

wind loads; winds; tsunami.

iv
SI CONVERSION UNITS

Length

In view of the present accepted practice for wind and seismic technology, common units

Df measurements were used throughout this publication. In recognition of the position of the

Jnited States as a signatory to the General Conference on Weights and Measures, which gave

Dfficial status to the International System of Units (SI) in 1960, the table below is presented to facilitate conversion to SI Units. Readers interested in making further use of the

coherent system of SI units are referred to: NBS SP 330, 1972 Edition, The International

System of Units; and ASTM E380-76, Standard for Metric Practice.

Force

Pressure

Stress

Energy

Torque

or ending

oment

Weight

.Unit Weight

Velocity

Acceleration

Table of Conversion Factors to SI Units

Customary Units

inch (in)

foot (ft)

pound (lbf)

kilogram (kgf)

pound per square

inch (psi)

kip per square

inch (ksi)

inch-pound (in-lbf)

foot-pound (ft-lbf)

pound-inch

pound-foot (lbf-ft)

pound (lb)

pound per cubic foot

(pcf )

foot per second

(ft/sec)

foot per second

per second (ft/sec 2

)

International

(SI) UNIT

meter (m)a

meter (m)

newton (N)

newt on (N)

newton/meter

newton/meter

joule (J)

joule (J)

newton-meter (N-m)

newton-meter (N-m)

kilogram (kg)

kilogram per cubic

meter (kg/m3

)

meter per second

(m/s)

meter per second

2 \ per second (m/s )

Conversion

Approximate

1 in = 0.0254 m*

1 ft = 0.3048 m*

1 lbf = 4.48 N

1 kgf = 9.807

1 psi = 6895 N/m2

1 ksi = 5985 x 106 N/m2

1 in-lbf = 0.1130 J

1 ft-lbf = 1.3558 J

1 lbf-in = 0.1130 N-m

1 lbf-ft = 1.3558 N-m

1 lb = 0.4536 kg

1 pcf = 16.018 kg/m3

1 fps = 0.3048 m/s

1 ft/sec 2 = 0.3048 m/s 2

a Meter may be subdivided. A centimeter (cm) is 1/100 m and a millimeter (mm) is 1/1000 m.

* Exactly

v

CONTENTS

Page

PREFACE iii

ABSTRACT iv

SI Conversion Units v Program of Joint Meeting . . x List of Members xvi

Formal Resolutions xx

TECHNICAL PAPERS

Natural Wind Characterization and Extreme Wind Records

. 1. On the Distribution of Extreme Winds Expected in Japan 1-1

2 . Boundary Layer Winds 2-1

3. On the Damage to Buildings In Oki-erabu from the

September 1977 Typhoon 3—1

4. Recent Developments in Atmospheric Remote Sensing and Their

Implication for Wind Engineering 4-1

5. Extreme Wind Data Base Development at the National Climatic

Center 5-1

Characterization of Earthquake Ground Motions and Strong-Motion

Earthquake Data

6. A Strong-Motion Record Information Retrieval System 6-1

7 . Integration of Strong-Motion Accelerograms 7-1

8. Some Recent Developments in National and International

Seismic Data Exchanges 8-1

9. Report on the International Workshop on Strong-Motion Earthquake

Instrument Arrays

May 2 - 5, 1978, at Hawaii 9-1

Engineering Seismology

10. Expectancy of Maximum Earthquake Motions in Japan 10-1

11. Determination of Design Earthquake for the Dynamic

Analysis of the Fort Peck Dam 11-1

12. The Izu-Ohshima Kinkai Earthquake of January 1978 12-1

Response of Hydraulic and Earth Structures to Seismic Forces

13. Empirical and Analytical Methods of Estimating Soil

Liquefaction Risk '. 13-1

14. A Practical Procedure for Assessing Earthquake-Induced

Liquefaction of Sandy Deposits 14-1

Structural Response to Wind Loading

15. Wind-Resistant Design of Cable-Stayed Bridges in Japan 15-1

16. Aerodynamic Stability of Proposed Ohio River

Cable-Stayed Bridge 16-1

17. Some Aerodynamic Considerations in the Design of the

Ruck-A-Chucky Bridge 17-1

18. A Summary of Wind Tunnel Test Results for the Luling,

Louisiana, Cable-Stayed Bridge 18-1

19. A Technique for Measuring Fluctuating Wind Loads on a Tall-Building Model Irrespective of Model Motion 19-1

20. A Study of Building Damage Caused by Wind 20-1

i
Page

Recent Developments in Seismic Design Criteria

21. Recent Earthquake-Resistant Design Methods for

Different Types of Bridge Foundations in Japan .................... 21-1

22. Recent Developments in Seismic Design Codes ......................... 22-1

23. Logical Analysis of Seismic Design Provisions ....................... 23-1

24. Highlights of California School and Hospital

Building Regulations 24-1

25. Summary of Research Projects in the LargeStructures Testing Laboratory from 1967-1977 25-1

Design and Analysis of Special Structures

26. Dynamic Behavior of Rectangular Water Tanks

Assembled with Panels ............................................. 26-1

27. Inelastic Behavior of Non-Bearing Walls in an 11-Story

Steel-Reinforced Concrete Frame

Aseismatic Safety of External Surface Finishes and Coatings ......... 27-1

Damage Evaluation, Repair, and Retrofit

28. Seismic Evaluation of Existing Multistory Residential

Buildings 28-1

29. Damage to Engineering Structures During the

Near Izu-Ohshima Earthquake of January 1978 ....................... 29-1

30. Repair and Retrofit of Buildings .................................... 30-1

31. The Distribution of Property Losses Caused by

Historical Earthquakes ............................................ 31-1

Earthquake Hazard Mitigation, Including Socioeconomic Aspects

32. The Disaster-Resistance of Cities and Their Lifelines ............... 32-1

33. Rescue and Rehabilitation After the Izu-Ohshima Kinkai

Earthquake of 1978 .. .......................................... 33-1

34. National Science Foundation Activity in Earthquake

Hazard Mitigation ................................................. 34-1

35. Rapid Seismic Analysis Procedures for Buildings ..................... 35-1

36. US-Southeast Asia Symposium on Engineering for

Natural Hazards Protection ........................................ 36-1

Storm Surge and Tsunami

37. Storm Surge 37-1

38. Operation of the Tsunami Warning System in the

Pacific 38-1

Members of Task Committee 39-1

Task Committee Reports

A. Task Committee on Strong-Motion Instrumental Arrays and

Data .............................................................. A-l

B. Task Committee on Large-Scale Testing Programs ...................... B-l

C. Task Committee on Repairs and Retrofit of Existing

Structures C-l

D. Task Committee on Structural Performance Evaluation ................. D-l

viii
Page

E. Task Committee on Land Use Programs for Controlling

Natural Hazards Effects E-l

F. Task Committee on Lifeline Systems F-l

G. Task Committee on High Speed Wind Data ................................. G-l

H. Task Committee on Soil Behavior and Stability During Earthquakes ....... H-l

IX
TENTH JOINT MEETING

OF

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS

May 23 - 26, 1978

at the

National Bureau of Standards

TUESDAY - May 23 Lecture Room B

OPENING SESSION

10:00 A.M. Call to order by Dr. H. S. Lew, Secretary, U.S. Panel

Remarks by Dr. John W. Lyons, Director, National Engineering

Laboratory, National Bureau of Standards

Remarks by Mr. Kiichiro Nagara, Counselor, Embassy of Japan

Remarks by Dr. Edward 0. Pfrang, Chairman, U.S. Panel

Remarks by Mr. Kazuto Nakazawa, Chairman, Japan Panel

10:30 A.M. Introduction of U.S. Panel Members by U.S. Chairman and Japan Panel

Members by Japanese Chairman

10:45 A.M. Election of Conference Chairman

10:50 A.M. Adoption of Agenda

11:00 A.M. Group Photograph

11:30 A.M. Lunch - Dining Room C

SESSION I NATURAL WIND CHARACTERIZATION AND EXTREME WIND RECORDS

Chairman: Dr. Edward 0. Pfrang

1:00 P.M. On the Distribution of Extreme Winds Expected in Japan -

T. Okubo and N. Narita

1:25 P.M. Boundary Layer Winds - D. A. Haugen

1:50 P.M. On the Damage to Buildings in Okierabu Island Caused by Typhoon

Number 7709 of September 9, 1977 - T. Murota, Y. Ishiyama,

S. Fujiwhara, M. Okuda and K. Tsukada

2:10 P.M. Recent Developments in Atmospheric Remote Sensing and Their

Implications for Wind Engineering - W. H. Hooke

2:30 P.M. Extreme Wind Data Base Developed at the National Climatic Center -

M. Changery

2:50 P.M. Discussion

3:10 P.M. Break

x
SESSION II CHARACTERIZATION ON EARTHQUAKE GROUND MOTIONS AND

STRONG-MOTION EARTHQUAKE DATA

Chairman: Mr. Kazuto Nakazawa

3:50 P.M. Strong-Motion Record Information Retrieval System - A. M. Converse and R. B. Matthiesen

4:15 P.M. Integration of Strong-Motion Accelerograms -

S. Iai, E. Kurata, H. Tsuchida and S. Hayashi

4:40 P.M. Some Recent Developments in National and International Seismic

Data Exchange Activities - J. Lander

5:00 P.M. Development of Dense Strong-Motion Earthquake Instrumentation

Arrays - T. Okubo and H. Tsuchida

5:20 P.M. Discussion

5:30 P.M. Adjourn

WEDNESDAY - May 24 Lecture Room B, Administration Building

SESSION III ENGINEERING SEISMOLOGY

Chairman: Dr. Edward 0. Pfrang

9:00 A.M. Study on Expectancy of Maximum Earthquake Motions in Japan - M. Watabe and Y. Kitagawa

9:25 A.M. Determination of Design Earthquake for the Dynamic Analysis of

Fort Peck Dam - W. F. Marcuson, III and E. L. Krinitzsky

9:50 A.M. Analysis of the Izu-Oshima Kinkai Earthquake of January 14, 1978 - M. Tajima, H. Sato, M. Otsuka, K. Sudo and K. Ishibashi

10:10 A.M. Discussion

10:25 A.M. Break

SESSION IV RESPONSE OF HYDRAULIC AND EARTH STRUCTURES TO SEISMIC FORCES

Chairman: Mr. Kazuto Nakazawa

10:40 A.M. Empirical and Analytical Methods of Estimating Soil Liquefaction

Risk - R. K. McGuire, F. Tatsuoka, T. Iwasaki and K. Tokida

11:05 A.M. A Practical Procedure for Assessing Earthquake-Induced Liquefaction

of Sandy Deposits - M. Ohashi, T. Iwasaki, F. Tatsuoka and K. Tokida

11:30 A.M. Discussion

11:40 A.M. A Presentation on Site Investigation of Dam Failures During the

Izu-Oshima Kinkai Earthquake of January 14, 1978 - W. Marcuson, III

12:00 noon Lunch - Dining Room C

xi
SESSION V STRUCTURAL RESPONSE TO WIND LOADING

Chairman: Dr. Edward 0. Pfrang

1:00 P.M. Wind Resistant Design of Cable-Stayed Bridges in Japan - T. Okubo,

N. Narita and M. Katsuragi

1:25 P.M. Aerodynamic Stability of Proposed Ohio River Cable-Stayed Bridge -

L. R. Cayes

1:50 P.M. Some Aerodynamic Considerations in the Design of the Ruck-A-Chucky

Bridge - R. H. Scanlon

2:10 P.M. A Summary of Wind Tunnel Test Results for the Luling, Louisiana,

Cable-Stayed Bridge - H. R. Bosch

2:30 P.M. A Technique for Measuring Fluctuating Wind Loads on a Tall Building

Model Irrespective of Model Motion - T. Reinhold

2:50 P.M. Discussion

3:15 P.M. Break

3:30 P.M. Task Committee Meetings

a. Task Committee on Strong-Motion Instrumentation Arrays and Data

b. Task Committee on Large-Scale Testing Programs

c. Task Committee on Repairs and Retrofit of Existing Structures

d. Task Committee on Structural Performance Evaluation

5:30 P.M. Adjourn

THURSDAY - May 25 Lecture Room B, Administration Building

SESSION VI RECENT DEVELOPMENTS IN SEISMIC DESIGN CRITERIA

Chairman: Mr. Kazuto Nakazawa

9:00 A.M. Recent Earthquake Resistant Design Methods for Different Types of

Foundations in Japan - Y. Shioi, T. Furya, M. Okahara and

Y. Mitsuie

9:25 A.M. Recent Developments in Seismic Design Code - R. L. Sharpe

9:50 A.M. Logical Analysis of Seismic Design Provisions - J. Harris

10:10 A.M. Highlights of California School and Hospital Building Regulations -

J. F. Meehan

10:30 A.M. Break

10:45 A.M. Summary of Recent Projects Carried Out in the Large-Size Structures

Testing Laboratory During the Last Eleven Years - M. Hirosawa,

Y. Ishiyama and T. Goto

11:05 A.M. Discussion

xii
11:55 A.M. Inelastic Behavior of Non-Bearing Wall in a 11-Story SteelReinforced Concrete Frame and Aseismic Safety of External Surface

Finishes and Coatings - M. Watabe, T. Kubota, T. Fukuta, A. Baba

and H. Ito

12:15 P.M. Discussion

12:25 P.M. Lunch - Dining Room C

SESSION VIII WIND AND SEISMIC DAMAGE EVALUATION, REPAIR AND RETROFIT OF STRUCTURES

Chairman: Dr. Edward 0. Pfrang

1:30 P.M. Seismic Evaluation of Existing Multistory Residential Buildings -

G. R. Fuller

1:55 P.M. Damage Features of Engineering Structures Due to the Near IzuOhshima Earthquake of January 14, 1978 - K. Nakazawa, T. Iwasaki,

K. Kawashima, M. Watabe, H. Yamanouchi and Y. Yamazaki

2:20 P.M. Repair and Retrofit of Buildings in Seismic Areas - R. D. Hanson

and J. K. Wight

2:40 P.M. An Evaluation Study on the Distribution of Property Losses Caused

by Earthquakes - E. Kuribayashi and T. Tazaki

3:00 P.M. Discussion

3:15 P.M. Break

3:30 P.M. Task Committee Meetings

e. Task Committee on Land Use Programs for Controlling Natural

Hazards Effects

f. Task Committee on Disaster Prevention Methods for Lifeline

Systems

g. Task Committee on High-Speed Wind Data

h. Task Committee on Soil Behavior and Stability During

Earthquakes

5:30 P.M. Adjourn

IFRIDAY - May 26 Lecture Room B, Administration Building

SESSION IX EARTHQUAKE HAZARD MITIGATION INCLUDING SOCIOECONOMIC ASPECTS

Chairman: Dr. Edward 0. Pfrang

9:00 A.M. Present Status and Problems on Countermeasures to Earthquake

Disasters with Emphasis on Disaster-Resistivities of Cities and

Their Lifelines - K. Nakazawa and E. Kuribayashi

9:20 A.M. The Investigation on the Rescue and Rehabilitation of the IzuOhshima Kinai Earthquake of 1978 - E. Kuribayashi, T. Tazaki and

T. Hadate

9:40 A.M. The National Science Foundation Activities on Earthquake Hazard

Mitigation - C. Thiel and J. Scalzi

xiii
10:00 A.M. Rapid Seismic Analysis Procedures for Buildings - T. K. Lew,

S. K. Takahashi and C. V. Chelapati

10:20 A.M. U.S. - Southeast Asia Symposium on Engineering for Natural Hazards

Protection - A. H-S. Ang

10:40 A.M. Discussion

11:00 A.M. Break

SESSION X STORM SURGE AND TSUNAMI

Chairman: Mr. Kazuto Nakazawa

11:15 A.M. Storm Surge - C. S. Barrientos

11:35 A.M. Operation of the Tsunami Warning System in the Pacific - M.Spaeth

11:55 A.M. Discussion

12:05 P.M. A Presentation on the Teton Dam Failure - J. Dodd

12:30 P.M. Lunch - Dining Room C

SESSION XI TASK COMMITTEE REPORT

Chairman: Dr. Edward 0. Pfrang

1:30 P.M.

1:45 P.M.

2:00 P.M.

2:15 P.M.

2:30 P.M.

2:45 P.M.

3:00 P.M.

3:15 P.M.

3:30 P.M. Break

Task Committee Report

a. Task Committee on

b. Task Committee on

c. Task Committee on

d. Task Committee on

e

.

Task Committee

Hazard Effects

on

f. Task Committee

Systems

on

g. Task Committee on

h. Task Committee

Earthquakes

on

xiv
SESSION XII CLOSING SESSION

Chairman: Dr. Edward 0. Pfrang

3:40 P.M. Adoption of Formal Resolution

Closing Remarks - Mr. Kazuto Nakazawa and Dr. Edward 0. Pfrang

4:10 P.M. Adjourn

XV
U.S. PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

MAY 1978

Dr. Edward 0. Pfrang

Chairman

Chief, Structures and Materials Division

Center for Building Technology, NEL

National Bureau of Standards

Washington, DC 20234

301-921-2196

Dr. S. T. Algermissen

Office of Earthquake Studies

Denver Federal Center

Branch of Earthquake Tectonics

USGS

Stop 978, Box 25046

Denver, CO 80225

303-234-4014

Dr. Celso Barrientos

Research Meteorologist

NOAA

8060 13th Street

Silver Spring, MD 20910

Mr. Billy Bohannan

Assistant Director

Wood Engineering Research

Forest Products Laboratory, USDA

P.O. Box 5130

Madison, ,WI 53705

608-257-2211

Dr. Roger D. Borcherdt

Office of Earthquake Studies

Branch of Earthquake Hazards, USGS

345 Middlefield Road

Menlo Park, CA 94025

415-323-8111

Dr. Charles G. Culver

Disaster Research Coordinator

Structures and Materials Division, NEL

National Bureau of Standards

Washington, DC 20234

301-921-2196

Mr. Jerry Dodd

Bureau of Reclamation

P.O. Box 25007

Denver Federal Center

Denver, CO 80225

303-234-3089

Dr. Michael P. Gaus

Division of Problem Focused Research

Applications

National Science Foundation

1800 G Street, NW Washington, DC 20550

202-632-5787

Mr. John J. Healy

Chief, Research and Development Office

Department of the Army

DAEN-RDM

Washington, DC 20314

202-693-7287

Dr. William B. Joyner

Office of Earthquake Studies

Branch of Earthquake Hazards, USGS

345 Middlefield Road

Menlo Park, CA 94025

415-323-8111

Mr. John W. Kaufman

Aerospace Environment Division

ES42-Space Science Laboratory

National Aeronautics and Space Administration

Marshall Space Flight Center, AL 35812

205-453-3104

Mr. James Lander

National Oceanic and Atmospheric

Administration

Environmental Data Service

National Geophysical and Solar-Terrestrial

Data Center

Boulder, CO 80302

Dr. H. S. Lew

Structures Materials Division

Center for Building Technology

National Bureau of Standards

Washington, DC 20234

301-921-2647

Dr. E. V. Leyendecker

Structures and Materials Division

Center for Building Technology

National Bureau of Standards

Washington, DC 20234

301-921-3471

xvi
Dr. Richard D. McConnell

Office of Construction

Veterans Administration

811 Vermont Avenue, NW Washington, DC 20234

202-389-3103

Dr. William F. Marcuson, III Research Civil Engineer, WES-SH

Department of the Army

Waterways Experiment Station

Corps of Engineers

P.O. Box 631

Vicksburg, MI 39180

Dr. Richard D. Marshall

Office of the Associate Director for Programs

National Bureau of Standards

Washington, DC 20234

301-921-3132

Dr. R. B. Matthiesen

Chief, Branch of Seismic Engineering

Office of Earthquake Studies, USGS

345 Middlefield Road

Menlo Park, CA 94025

415-333-8111 (FTS 467-2881)

Mr. John F. Meehan

Department of General Services

Office of Architectural Construction

Sacramento, CA 95805

Mr. Howard L. Metcalf

Deputy Director for Construction Standards

and Design

Office of the Assistant Secretary of Defense

(MR&L) Room 3E763

The Pentagon

Washington, DC 20301

202-695-2713

Dr. Thomas D. Potter

Deputy Director

Environmental Data Service, NOAA

3300 Whitehaven Street, NW

Washington, DC 20235

202-634-7319

Dr. John B. Scalzi

Program Manager

Earthquake Hazards Mitigation

National Science Foundation

1800 G Street, NW Washington, DC 20550

Mr. Charles Scheffey

Director, Office of Research

Federal Highway Administration

Department of Transportation

Washington, DC 20590

202-426-3943

Dr. Warren A. Shaw

Head, Civil Engineering Department

Civil Engineering Laboratory

Naval Construction Battalion Center

Port Hueneme, CA 93043

805-982-5407

Alternates

Dr. A. Gerald Brady

Physical Scientist

Office of Earthquake Studies, USGS

345 Middlefield Road

Menlo Park, CA 94025

415-323-8111

Dr. James D. Cooper

Structures and Applied Mechanics Division

Federal Highway Administration

Office of Research HRS-11

Washington, DC 20590

202-557-4315

Mr. G. Robert Fuller

Architectural and Engineering Division

Department of Housing and Urban Development

Washington, DC 20411

Dr. Jerry Harbour

Site Safety Research Branch

Office of Site Safety Research Branch

Nuclear Regulatory Commission

Washington, DC 20555

301-427-4370

Dr. Charles T. Thiel

Deputy Director

Division of Advanced Environmental Research

and Technology

National Science Foundation

Washington, DC 20550

202-632-5734

Mr. Drew A. Tiedemann

Bureau of Reclamation

Engineering and Research Center

Denver Federal Center

Denver, CO 80225

303-234-3029

Mr. Robin McGuire

U.S. Department of the Interior

U.S. Geological Survey

Box 25046

Denver Federal Center

Denver, CO 80225

303-234-2874

xvii
JAPAN PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

MAY 1978

Mr. Kazuto Nakazawa, Chairman

Director

Public Works Research Institute

Ministry of Construction

2-28-3 2 j Honkomagome, Bunkuo-ku, Tokyo, 113

Dr. Keikichi Naito

Head, Meteorological Satelite Division

Meteorological Research Institute

Japan Meteorological Agency

4-35-8, Koenji Kita, Suginami-ku, Tokyo, 166

Mr. Shigemi Fujiwhara

Head, Typhoon Research Division

Meteorological Research Institute

Meteorological Agency

4-35-8, Koenji Kita, Suginami-ku, Tokyo, 166

Dr. Sadaiku Hattori

Chief, Seismology Section

International Institute of Seismology and

Earthquake Engineering (I.I.S.E.E.)

Building Resejarch Institute

Ministry of Construction

3-28-8, Hyakunin-cho , Shinjuku-ku, Tokyo, 160

Dr. Satoshi Hayashi

Director

Port and Harbour Research Institute

Ministry of Transport

3- 1-1, Nagase, Yokosuka-shi

,

Kanagawa-ken , 239

Mr. Masaya Hirosawa

Chief Researcher

Building Research Institute

Ministry of Construction

5-26-25 Daitakubo, Urawashi, Saitamaken, 336

Mr. Toshio Iwasaki

Chief, Ground Vibration Section

Earthquake Disaster Prevention Division,

Chiba Branch

Public Works Reserch Institute

Ministry of Construction

4- 12-52, Anagawa, Chiba-shi, Chiba-ken, 280

Mr. Eiichi Kuribayashi

Chief, Earthquake Engineering Section

Earthquake Disaster Prevention Division,

Chiba Branch

Public Works Research Institute

Ministry of Construction

4-12-52, Anagawa, Chiba-shi, Chiba-ken, 280

Mr. Tatsuro Murota

Chief, Structure Section

Structure Division

Building Research Institute

Ministry of Construction

3-28-8, Hyakunin-cho, Shinjuku-ku, Tokyo, 160

xvi

Dr. Kiyoshi Makano

Director

Building Research Institute

Ministry of Construction

3- 28-8, Hyakunin-cho, Shinjuku-ku, Tokyo, 160

Mr. Nobuyuki Narita

Chief, Structure Section

Structure and Bridge Division, Chiba Branch

Public Works Research Institute

Ministry of Construction

4-12-52, Anagawa, Chiba-shi, Chiba-ken, 280

Dr r Masamitsu Ohashi

Head, Earthquake Disaster Prevention Division

Chiba Branch

Public Works Research Institute

Ministry of Construction

4-12-52, Anagawa, Chiba-shi, Chiba-ken, 280

Mr. Keiichi Ohtani

Chief, Earthquake Engineering Laboratory

National Research Center for Disaster

Prevention

Science and Technology Agency

3-1, Tennodai, Sakura-mura, Niihari-gun,

Ibaragi-ken, 300-32

Mr. Michio Ohtsuka

Acting Director, International Institute of

Seismology and Earthquake Engineering

(I.I.S.E.E.)

Building Research Institute

Ministry of Construction

3-28-8, Hyakunin-cho, Shinjuku-ku, Tokyo, 160

Dr. Hiroshi Sato

Chief, Research Division

Crustal Dynamics Department

Geographical Survey Institute

Ministry of Construction

3- 24-13, Higashiyama, Meguro-ku, Tokyo, 153

Mr. Kenkichi Sawada

Chief, Soil Dynamics Section

Construction Method and Equipment Division,

Chiba Branch

Public Works Research Institute

Ministry of Construction

4- 12-52, Aangawa, Chiba-shi, Chiba-ken, 280
Mr. Yukitake Shioi

Head, Foundation Engineering Section

Structure and Bridge Division, Chiba Branch

Public Works Research Institute

Ministry of Construction

4-12-52, Anagawa, Chiba-shi, Chiba-ken, 280

Mr. Akira Suwa

Head, Seismological Laboratory

Meteorological Research Institute

Meteorological Agency

4-35-8, Koenji Kita, Suginami-ku, Tokyo, 166

Mr. Hiroshi Takahashi

Head, Second Research Division

National Research Center for Disaster

Prevention

Science and Technology Agency

3-1, Tennadai, Sakura-mura, Niihari-gun,

Ibaragi-ken, 300-32

Mr. Hajime Tsuchida

Chief, Earthquake Resistant Structures

Laboratory

Structure Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi , Kanagawa-ken, 239

Dr. Makoto Watabe

Head, Structure Division

Building Research Institute

Ministry of Construction

3-28-8, Hyakunin-cho, Shinjuku-ku, Tokyo, 160

Dr. Tadayoshi Okubo, Secretary-General

Head, Planning and Research Administration

Division

Public Works Research Institute

Ministry of Construction

2-28-32, Honkomagome, Bunkyo-ku, Tokyo, 113

Secretariat

Mr. Keisuke Takagi

Chief, Planning Section

Planning and Research Administration Division

Public Works Research Institute

Ministry of Construction

2-28-32, Honkomagome, Bunkyo-ku, Tokyo, 113

xix
RESOLUTIONS OF THE TENTH JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS

U. J.N.R.

May 23 - 26, 1978

In response to the urgent needs expressed by the U.S. Congress

in its "Disaster Mitigation Act of 1977" and by Japanese

Diet's intention of adopting a similar act within this year,

the U.S. -Japan Panel on Wind and Seismic Effects proposes

(adopts) the following resolutions for future activities:

1. The Tenth Joint Meeting was an extremely valuable

exchange of technical information which was beneficial

to both countries. In view of the importance of

cooperative programs on the subject of wind and seismic

effects, the continuation of Joint Panel Meetings is considered essential.

2. The exchange of technical information, especially

revised codes and specification relevant to wind and

seismic effects, and the promotion of research programs

including exchange of personnel and available equipment

should be strengthened.

3. The Panel on Wind and Seismic Effects recognizes the

importance of the U.S. -Japan Cooperative Program on Large Scale Testing and it urges early implementation

of the program through this Panel.

4. Considerable advancement has been made by various

Task Committees. Cooperative research programs should

be promoted through the Task Committees.

5. Chairpersons of both sides should encourage broadening the

activities to include the topic on tsunami and storm

surge.

6. Chairmen of both sides should explore, by correspondence, the feasibility of establishing an additional

task committee on wind and earthquake engineering for

nuclear facilities construction.

7. Date and location of the Eleventh Joint Meeting of the

Panel on Wind and Seismic Effects will be Fall 1979 in

Japan. Specific dates and itinerary will be determined

by the Japanese Panel with concurrence by the U.S.

Panel. Subsequent meetings will be held in May of each

year.

XX
1

ON THE DISTRIBUTION OF EXTREME WINDS EXPECTED

IN JAPAN

Todayoshi Okubo

Nobuyuki Narita

Public Works Research Institute

Ministry of Construction

On the Distribution of

Extreme Winds Expected in Japan

This paper deals with the basic wind speed — the determination of

the expected value of extreme wind, which forms the basis of the

design wind speed. Wind speed predictions based on historical data

are generally made by using the extreme value theory [1], Distribution function of yearly maximum ten-minute average wind speed can

be fitted by double exponential distribution:

F(V) = f(y) - exp(-exp(-y))

y = a(V - b)

where F(V), a, and b are the distribution function and constants.

R. Saito [2] and E. Kikuchibara [S] have shown that historical wind

data at the meteorological observatories satisfy the distribution

mentioned above. E. Kikuchibara [4] has also shown that Gringorten's

method, in which Hazen's distribution is saved (F(V) = (2i - 1 ) / 2N

,

where N and i denote the number of samples and the order of samples

in N respectively), is the best for the estimation.

On the other hand, the multiple regression method has also been

used for estimating the expected wind speed at an arbitrary point,

in which the topographical factors (altitude, undulation, ratio of

land or sea area to surroundings, and so on) are used. R. Satio

[5] applied the method to determine expected wind speed every two kilometers between Tokyo and Osaka. The result formed the basis of

design wind speeds for railway structures on the New Tokaido Line.

Wind Speed Data

Yearly maximum wind speeds recorded by the meteorological

observatories of the Meteorological Agency are the foundation of

basic wind speeds. Wind speeds are commonly reported as tenminute average speed and instantaneous speed. The instantaneous

wind speeds are valid for about two seconds depending on the

sensitivity of the measuring device.

The wind records were collected from the 163 meteorological

observatories. The records were from 1929 to 1975. As the observation method (measuring device, probe height, location, data process regulation, and so on) has been modified during the period,

the wind data have been homogonized to give yearly maximum ten- minute speed at a height of ten meters.

The data were checked if they satisfied the Hazen's formula on

the double exponential distribution diagram, and the result was as follows

(number of data)

109 (67%)

34 (21%)

20 (12%)

Effects of Statistical Length and Period on Expected Speeds

The wind data at 55 observatories ranging from 1929 to 1975 were classified into four groups according to the statistical length of

years as follows:

(fitness)

better

good

not so good

1-1
(I) 47 years (1929 - 1975)

(II)

(III)

(IV)

33 years

21 years

12 years

(1943 - 1975)

(1955 - 1975)

(1964 - 1975)

The mean and the standard deviation of the ratio of expected

wind speeds referring to the specific return period are tabulated

in Table 1

.

The wind data were also divided into four groups, in which

every group had the same statistical period of twelve years:

(I)

(II)

(III)

(VI)

12 years

12 years

12 years

12 years

(1964 - 1975)

(1954 - 1965)

(1944 - 1955)

(1934 - 1945)

Results are given in Table 2. The expected wind speed estimated by the data of the oldest period gives the largest value.

It is desirable for carrying out the multiple regression analysis of topographical factors to select a set of data with the same statistical length and period of years. It is also desirable to select wind data with observation periods as long as possible. Wind

data at 101 observatories were selected, which had the statistical

period of 33 years (1943 - 1975). Eighty-one data among them were

used for deriving the regression equation; the rest (twenty data) were used for checking the equation.

Estimation of Expected Values by Multiple Regression

Definition of topographical factors (X^):

(1) altitude of anemometer (Xj)

(2) ratio of the sea to the whole area in a specific circle

surrounding the point (X2, X^ and X^)

(3) ratio of land to the whole area in a specific circle

surrounding the point (X^, —,Xg)

(4) openess of the surrounding area (X^,—— ,X^2^

(5) undulation of the ground of surrounding area

(6) slope of the ground (Xjy, —»^20^

(7) distance from the point to obstacles (Xj^)

(8) covergence of surrounding 31" £3 ^ ^22 ^25

(x13 ,--,x

16 )

Relations between expected wind speed and topographical factors were analyzed by multiple regression, in which regression equations

were assumed to be linear. Table 3 shows how the multiple correlation coefficients increase as the number of topographical factors

increase. The expected wind speed is highly correlated with the

openess factor X^. the radius being 40 kilometers.

Table 4 shows the compatibility of the regression equations.

The standard deviation by the regression equations (81 points) is

nearly equal to that of the checking points (20 points).

1-2
Estimation of Expected Values by Modified Method

Estimation errors, i.e., the differences between expected wind

speeds from Gringorten's method and those from the multiple regression method, have a regional distribution, and the averaged estimation errors were added to the equations in the modified method.

This suffices for the meteorological effects that cannot be

expressed by the five topographical factors and constants.

Although the standard deviation can be decreased to 1.1 m/s in the

case of 81 points (Table 5), this seems lacking in generality^ as seen in Table 6.

As the second step, regression equations were derived by the

use of 131 stations ranging from 1943 to 1975. The results are given in Tables 7 and 8, in which the standard deviation decreases

to 1.1 m/s. The average estimation errors are given in Figures 1 to

3, and the modified regression equations can be written as follows:

V

2Q

= 30.1 + 0.0107X^+0.879X2+0.0126X^1 .08Xg-0.00669X13 +AV20

V50

= 34.0+0.00985X12 +0.983X2 +0.0137X1 -1.24Xg-0.00790X13 +AV50

V 100

= 37 - 0+0 - 00919xi2 +1 " 07x

2 +0 ' 0145Xl" 1 - 37x8" 0 - 00883X13 +A^100

where

X

12 R = 40 kilometers openess

x

2 R = 3 kiometers sea

X

l alt itude X8

: R = 40 kilometers land X

13 R = 5 kilometers undulat ion

Estimation of Expected Wind Speeds at Points Except Observatories

Ninety-seven specific points except meteorological observatories

were selected to draw up the contour of expected wind speeds.

Figure 4 shows the final distribution map. Similar maps have

also been drawn up for the return periods of 20 and 50 years.

The topographical distribution of instantaneous wind speeds

has been obtained by using a similar procedure, and Figure 5 shows

the map of 100-year expected instantaneous wind speed.

Concluding Remarks

The design wind loads in the present Design Specifications of

Bridges in Japan are based on the ten-minute wind speed of 40m/s.

However, as seen in Figure k, the expected wind speed exceeds this

specified value in some southern coastal regions on the Pacific

Ocean side and in some western coastal regions on the Japan Sea

side. For the best bridge designs, the basic wind values

representing the region should be used instead.

The authors wish to acknowledge their debt to Dr. A. Masatsuka

and Mr. T. Kimura, who prepared the meteorological data.

References

1. Gumbel, E. J., Statistics of Extremes .

2. Saito, R. , "Probabilistic Distribution of Strong Wind—Study

on High Wind in Japan III," Research Review , 9, No. 7, 1955

(in Japanese)

.

1-3
Meteorological Agency, Estimation of Return Period of Strong

Wind at the Construction Site of the Proposed Honshu Shikoku

Bridges , 1965 (in Japanese).

Association for Electrification of Railways, Wind-Strong Wind

and Railway Tracks , 1961 (in Japanese).
TABLE 1: VARIATION OF EXPECTED WIND SPEED WITH

STATISTICAL LENGTH OF YEARS

return period (yrs) 20 50 1 00

ratio ©/® ®/® ®/® ©/© ®/® ®/® ©/® ®/® ®/®

mean 0.97 0.94 0.87 0.97 0.94 0.87 0.97 0.94 0.86

standard deviation 0.057 0.081 0.100 0. 063 0.089 0.111 0.067 0.094 0. 1 16

where

(I) expected wind speed derived from the data, 19291975

( II ) expected wind speed derived from the data, 1943 1975

(III) expected wind speed derived from the data, 1955-1975

(IV) expected wind speed derived from the data, 1964 1975

TABLE 2: VARIATION OF EXPECTED WIND SPEED WITH

STATISTICAL PERIOD OF YEARS

return period (yrs) 20 50 100

ratio ®/® ®/® ®/® ®/® ®/® ®/® ®/® ®/® ®/®

mean 1. 16 1.15 1. 21 1. 16 1. 16 1. 22 1. 17 1. 16 1.23

standard deviation 0. 139 0. 180 0. 194 0. 159 0.202 0.216 0. 172 0.217 0.232

where

(I) expected wind speed derived from the data, 1964 1975

( II ) expected wind speed derived from the data, 1954 1965

(ID) expected wind speed derived from the data, 1944-1955

(IV) expected wind speed derived from the data, 1934-1945

TABLE 3: MULTIPLE CORRELATION COEFFICIENTS

return period 20 50 100

rank topographical factor multiple correl- ation coefficient topographical factor multiple correl- topographical factor

multipl e correlation coefficient

1 X,i (R=40km) 0.629 X,i (R=40km) 0. 610 X,, ( R — 40km) 0. 599

2 X, 0.661 X, 0.637 X, (R = 3km) 0. 627

3 X3 (R= 5km) 0.721 X3 (R= 5km) 0. 699 X, 0.679

4 X!0 (R =20km) 0.732 X!0 (R=20km) 0.713 X„ (R = 15km) 0.692

5 Xs (R = 10km) 0. 740 Xs (R = 10km) 0. 721 X 5 (R=10km) 0. 709

1-5
TABLE 4: COMPARISON OF ESTIMATION ERRORS

return period 20 50 100

type (I) ( n

)

( 1 ) in) ( I ) (m

mean (m/s) 0.0 0.5 0. 0 0.0 0.0 0.2

standard deviation Im/s) 5. 3 5. 0 6. 1 6. 0 6.8 6. 7

where

(I) data of 81 stations used for determining the multiple regression equations (D) data of 20 stations used for checking the multiple regression equations

TABLE 5: COMPARISON OF ESTIMATION ERROR

(81 STATIONS)

return period (yrs) 20 50 100

type (I) (II) (I) ( n

)

(I) (II) mean ( m/s) 0. 0 - 0. 1 0.0 - 0. 1 0. 0 0. 1

st.mH.Trd deviation (m/s) 5. 3 1. 1 6. 1 1.

1

6. 8 1. 1

where

( I ) multiple regression method

( [I ) modified method

TABLE 6: COMPARISON OF ESTIMATION ERRORS

(20 CHECK STATIONS)

return period (yrs) 20 50 100

type (1) (D) (I ) cm (I) ( n

)

mean (m/s) 0.5 0. 6 0.0 - 0. 1 0.2 0. 5

standard He* iation (m/s)

5. 0 4. 5 6.0 5. 2 6. 7 5. 3

where

( I ) multiple regression method

(II) modified method

1-6
TABLE 7: MULTIPLE CORRELATION COEFFICIENTS

return period 2 0 50 1 on

rank topographical factor multiple corel- atiun coefficient topographical factor multiple Corel- topographical factor multipk* correl1 Xi, (R=40km) 0.599 X,s (R = 40km) 0.577 X„ (R = 40km) 0.563

2 X, (R= 3km) 0.643 X, (R = 3km) 0.625 X2 (R= 3km) 0.614

3 X, li 7(1(1 X, 0.674 X, 0.659

4 X8 (R=40km) (I 7 1 5 X, (R=40km) 0.690 X8 (R=40km) II 6 75

5 X,, (R = 5km) 0.734 X,, (R = 5km) 0.712 X, 3 (R = 5km) 0 (,99

TABLE 8: COMPARISON OF ESTIMATION ERRORS

return period 20 50 100

type (I) (H) (I) (H) (I) (I)

mean (m/s) 0.0 . - 0. 1 0. 0 - 0. 1 0. 0 - 0.1

standard deviation (m/s) 5. 2 1. 3 6. 0 1. 2 6. 7 1. 1

where

(I) multiple regression method

( II ) modified method

1-7
1-8
1-9

FIGURE 4: DISTRIBUTION OF EXTREME WIND SPEED

(RETURN PERIOD OF 100 YEARS, 1 0-M IN. AVERAGE)

1-11
FIGURE 5: DISTRIBUTION OF EXTREME WIND SPEED

(RETURN PERIOD 100 YEARS, INSTANTANEOUS)

1-12
BOUNDARY LAYER WINDS

Duane A. Haugen

NOAA/ERL/Wave Propagation Laboratory

Boulder, Colorado 80302

Boundary Layer Winds

Introduction

The atmospheric boundary layer has received a great amount of

attention by research scientists over the past several decades.

Within the last ten years, particularly, the behavior of boundary

layer winds over flat, horizontally homogeneous, uncluttered land

surfaces has become well known. A large number of empirical laws

have been established that describe the height variation of mean

wind speed, gustiness ratios, and turbulence scale sizes for these

idealized flow conditions. Unfortunately, much of this knowledge

can be used only for general guidance in wind engineering problems.

Rarely, if ever, does the engineer find the problem located in

terrain approximating a flat plane. Furthermore, strong wind con- ditions are generally of limited interest to boundary layer scientists, but of primary interest to designers and builders of houses,

office buildings, and bridges. Indeed, certain types of high winds,

viz., tornadoes, thunderstorm wind surges and topographicallygenerated winds such as chinooks, are considered intermittent,

discrete events by the boundary layer scientist and rarely gain his

or her attention.

The purpose of this paper is threefold: (a) to outline briefly

those results of boundary layer research that seem applicable to wind engineering problems; (b) to emphasize the need for obtaining

local wind data sets in generating extreme wind statistics; (c) to draw attention to a few references which might be useful.

Boundary Layer Wind Characteristics

The atmospheric layer of most interest to us is the so-called

surface layer. This layer has received the most widespread attention from research meteorologists who find it well-suited as a "laboratory" for testing and developing theories of turbulent flow.

As a result, extensive reference material exists on the behavior of

surface layer winds. A recent summary article by Panofsky (1973),

for example, provides an excellent set of references for further

study.

We will restrict our attention to cases in which average wind

speeds in the layer exceed roughly 12 m/sec, excluding thunderstorms, tornadoes, and hurricanes. Under these conditions, a fairly

simple characterization of the surface layer is possible. The depth

of the layer will range between roughly 50 m and 200 m. In general,

the stronger the winds, the deeper the layer. The variation of

average wind speed with height is given by the so-called log-law,

U 9 tn(Z /Z ) u* - Mz'/z°2

)

> Zl ' 22 >y Z ° (1)

where U^ and U

2 are the average wind speeds at heights Zj and Z2

, respectively, and Z

Q

is the surface roughness parameters. The

overbar denotes a time average. The averaging period should be at least 15 minutes, but should not exceed 60 minutes for the law to be most dependable (Wyngaard et al., 1974).

The validity of the log-law is well-established. However, it

is often difficult to apply the law because of variations in the

surface roughness parameter. Order-of-magnitude estimates of Z

Q

for three broad categories of surface are given in Table 1. The

reader is cautioned that Z for any given site might well fall

2-1
outside the ranges shown. Furthermore, if the site is not horizontally homogeneous, the appropriate value of Z

Q will be a function

of wind direction.

TABLE 1. Z -values for three general types of surface

Surface Z (m)

Open meadow 0 01 to 0 10

Forests 0, 10 to 1. 0

Cities 1 0 to 10 0

Another means of specifying wind speed variation with height

is the so-called power law.

U

2 = U

1 (Z2 /Z1

) P

; Z

2

> Z

l (2)

for which typical values of p are given in Table 2 (Davenport,

1968).

TABLE 2. p-values for three general types of surface

Surface P

Open meadow 0. 14 to 0 2

Forests 0. 25 to 0 33

Cities 0 33 to 0 5

As with the log-law, application of the power law depends on knowledge of the local site. There is no way that this can be avoided.

It is therefore common practice for the design engineer to choose a wind profile law based largely on personal preference and to specify

surface characteristics according to experience and the safety

factor judged necessary.

In general, the average wind direction does not vary with

height in the surface layer under strong wind conditions. Therefore, it is appropriate to assume no directional change when making

first-order estimates of wind loading. The single most important

exception to this rule of thumb will be those cases where the structure of interest is in the wake of other structures in which case both wind speed and direction profiles will be strongly dependent on local conditions.

We now briefly consider wind variability or gustiness. By

definition, the gust factor is the ratio of the peak wind speed (of

a given duration) to the mean wind speed (for a given averaging

period). The gust factor may be considered independent of height

under strong wind conditions, but its value is highly dependent on

2-2
the gust duration and the length of the averaging period (Lettau

and Haugen, 1960). The most probable gust factors one should use range from 1.4 to 1.7, whereas maximum gust factors reported range

from 1.7 to 2.2 (Sissenwine et al., 1973).

Extreme Wind Statistics

The preceding discussion has not considered discrete phenomena

such as severe storms and topographically caused winds. Incorporating extreme wind statistics in engineering design problems is

well-established practice. Furthermore, the statistical techniques

used for estimating extreme values are also well-tested and widely

accepted (Gringorten, 1963; Gumbel, 1958). There is a serious

weakness in the application of extreme wind statistics, however,

that is frequently overlooked by the practicing engineer. All

of the techniques depend on the existence of long years of wind

records; the reliability and the statistics increase as the

length of record increases. However, many design and/or construction efforts take place in geographical regions where wind records

are non-existent and where local wind effects introduce extremes

not expected from the closest observing stations. One well-documented example of such anomalous extreme winds is the downslope

winds in Boulder, Colorado (Julian and Julian, 1969; Lilly and

Zipser, 1972). These winds occur frequently every winter; they

often exceed speeds of 95 mph ; they cause property damage of

roughly $1 million each year (Bergen and Murphy, 1978). Yet it

is only in recent years that local building practices have been

substantially modified to minimize wind damage to homes and businesses. To my knowledge there are no climatological extreme wind

charts that would have accurately predicted such severe wind

effects. Instead, local experience was necessary to permit laypersons

and engineers alike to appreciate the seriousness of the situation.

Another example of local wind anomalies is the highlypublicized problem with the John Hancock Insurance building in

Boston, Massachusetts. Reference material in the open literature

is not yet available because of pending lawsuits concerning blame

for the wind damage that has occurred. But this skyscraper was constructed in the middle of a large urban area that seemed relatively free of high wind problems. Whatever the resolution of this

engineering nightmare, its very occurrence stands as a strong case

for obtaining local wind data as a standard operating procedure for

all large construction efforts. If traditional wind specification

formulas prove to be adequate, all well and good. v If not, considerable savings in time and effort will inevitably result.

Wind Threshold Sensor

In closing, it seems appropriate to reference a new wind speed

threshold detection device that could be used by the construction

industry. Quite frequently designers need only extreme wind statistics to learn of possible wind loading problems either before or after the construction has been completed. A wind threshold sensor

designed by Bedard and Fujita (1978) seems ideally suited for this

purpose. Its data output consists of time of occurrence and duration of wind speeds above any desired thresold. The device is inexpensive, easy to operate, and reliable. Details of the sensor may

be obtained in the referenced article or by direct correspondence

with either of the authors. Mr. Bedard is at the Wave Propagation

Laboratory, NOAA, Boulder, Colorado; Prof. Fujita is at the University of Chicago, Chicago, Illinois.

2-3
References

1. Bedard, Jr., A. J., and Fujita, T. T. , 1978: An Omnidirectional, Tilt Insensitive, Wind Speed Thresold Detector, Proc

.

Fourth Symposium on Meteor. Obs. and Inst. , 10-14 April 1978,

Am. Meteor. Soc, Boston, Massachusetts.

2. Bergen, W. R. , and Murphy, A. H. , 1978: Potential Economic

and Social Value of Short-range Forecasts of Boulder Wind

Storms, Bull. Am. Meteor. Soc , Vol. 59, pp. 29-44.

3. Davenport, A. G. , 1968: The Dependence of Wind Load on Meteorological Parameters, Wind Effects of Buildings and

Structures , Univ. of Toronto Press, pp. 19-82.

4. Gringorten, I. I., 1963: A Simplified Method of Estimating

Extreme Values from Data Samples, J. Appl . Meteor

.

, Vol. 2, pp. 82-89.

5. Gumbel, E. J., 1958: Statistics of Extremes , New York,

Columbia Univ. Press, 375 pp.

6. Julian, L. T. , and Julian, P. R. , 1969: Boulder's Winds,

Weatherwise , Vol. 22, pp. 108-112, 126.

7. Lettau, H. H. , and Haugen, D. A., 1960: Wind, Handbook of

Geophysics , New York, The Macmillan Co., Chapter 5, Section 1.

8. Lilly, D. K. , and Zipser, E. J., 1972: The Front Range

Wind Storm of 11 January 1972 - a Meterological Narrative,

Weatherwise , Vol. 25, pp. 56-63.

9. Panofsky, H. A., 1973: Tower Micrometeorology , Workshop on Micrometeorology , Am. Meteor. Soc, Boston, Massachusetts,

pp. 151-176.

10. Sissenwine, N. , Tattelman, P., Grantham, D. G. , and

Gringorten, I. I., 1973: Extreme Wind Speeds, Gustiness, and

Variations with Height for ML-STD 210B, AFCRL-TR-73-0560 , Air

Force Camb. Res. Lab., Hanscom Field, Massachusetts, 72 pp.

11. Wyngaard, J. C, Arya, S. P. S. , and Cote, 0. R. , 1974: Some

Aspects of the Structure of Convective Planetary Boundary

Layers, J. Atmos . Sci

.

, Vol. 31, pp. 747-754.

2-4
3

ON THE DAMAGE TO BUILDINGS IN OKI-ERABU FROM THE SEPTEMBER 1977 TYPHOON

Tatsuo Murota

Yuji Ishiyama

Building Research Institute

Ministry of Construction

and

Shigemi Fujiwhara

Minoru Okuda

Kiyoshi Tsukada

Meteorological Research Institute

Ministry of Transportation

On the Damage to Buildings in Oki-erabu

from the September 197 7 Typhoon

Typhoon No. 7709 was first detected on September 2, 1977, near the

Caroline Islandsj moving westward slowly. On September 69 east of

Philippine Islands, it turned a NNW direction. At 0900 JST on the

8th, moving to the south-east of Taiwan, it developed into an intensive typhoon with central pressure of 905mb and maximum average wind speed of 55m/s. At 2000 JST on the 9th, it passed about

100 km east of Okinawa toward Oki-erabu Island. Since that time, the

typhoon had increased it speed gradually. A little before 2300

JST, it passed Oki-erabu toward East China Sea (see Figure 1).

Turning to the west, it moved through the East China Sea and

landed near the mouth of the Yangtse River on the morning of the

llthjand shortly declined into a temperate depression.

Mukade (1977) of the Okinawa Meteorological Observatory

calculated the detailed typhoon path near Oki-erabu based on radar

reports of two stations at Naha and Nase located nearly the same distance from Oki-erabu.

According to him, the typhoon had a circular eye wall 35 km in

diameter and the two paths calculated from the two radar reports

closely agreed with each other. As a consequence he decided that

the typhoon landed near Furusato in the central part of Okierabu

and moved in a NNW direction at a speed of 35 km/h as shown by the

solid line in Figure 2.

According to weather observations at a few points on the

island

* At Oki-erabu Weather Station, eye phenomenon was not clearly observed but winds calmed down temporarily.

The east edge of the eye wall probably passed near

the station.

° At the top of Mt . Oyama on the western part of the

island, wind direction changed counter-clockwise

during the typhoon and a clear eye phenomenon was observed for 25 minutes from 22:30 to 22:55.

* At Wadomari, 6.5 km south-west of Oki-erabu Weather

Station, an eye phenomenon was observed.

Summing up, it is inferred that the diameter of clear sky area in the typhoon eye was about 20 km and its center moved on a course more west than that determined by radar observations and

also east of Mt . Oyama. A dotted line in Figure 2 shows the

inferred path of the typhoon.

The 15 km difference in eye diameter can be explained as follows: calculating radar beam height at Oki-erabu based on the

data that height angle of Nase radar was 1 degree during the

typhoon observation and distance from Nase to Oki-erabu is about

150 km, the result comes to nearly 4000m. On the other hand,

according to rainfall, observations at Oki-erabu Weather Station,

rather weak rain continued for about 30 minutes while the typhoon

eye was passing on the island (see Figure 6). Such a weak rainfall

can often be caused by a nimbus lower than 4000m in height. From

these it can be assumed that the eye wall diameter would be 35 km.

Inside the wall there would be stratocumulus regions about 20 km

in diameter surrounding a clear sky area.
Table 1 shows the hourly changes of surface pressure, wind

speed, wind direction, and rainfall at Oki-erabu Weather Station

during the typhoon. Some of those are illustrated in Figures 3 to 6. The extreme values of meteorological elements observed at the same station during the typhoon passage are listed in Table 2.

Surface pressure continued to fall slowly until 21:20; after

that time it fell rapidly.. The minimum value of 907. 3mb was recorded at 22:50. This is the lowest value that has been observed

in Japan.

From 19 to 21 o'clock there were severe winds of around 20m/s

in average and more than 30m/s at peak. Then, wind speed increased

rapidly; at 22:00 peak gusts exceeded 50m/s. The maximum peak gust

of 60.4m/s (wind direction ENE) was recorded at 22:16. At that

time a severe gust blew down the wind tower of the station and wind

observations were forced to stop. The maximum average wind speed

recorded was 39. Ws (ESE) at 22:10.

A rather heavy rain of 20mm/h occurred at 21:10 at the beginning of the rapid pressure fall, but soon stopped. Another heavy

rain with an intensity of about 80mm/h occurred at about 22:00 and

continued until 22:30. From 22:30 to 23:00, a weak rain continued.

During this period Oki-erabu Weather Station must have been inside

of the typhoon eye. At 23:00 a heavy rain with an intensity of

lOOmm/h rose; it slowed after 20 minutes. During these two intense

rainfalls Oki-erabu Weather Station was probably in the eye-wall

belt. Based on this, the thickness of eye-wall belt can be

estimated at 18 km and 12 km in the forward and backward part of

the eye-wall, respectively.

Statistics of Damage

Table 3 shows the statistics of damage in Kagoshima Prefecture

announced by prefecture authorities at 8 a.m., September 19. According to this table, no one was killed but as many as 135

persons were injured. The number of buildings damaged amounted to 6290; the corresponding loss forms the major part of the total

property loss. Wind damage rather than water damage was the major

factor.

Damage to personnel and buildings in every administrative district is shown in Table 4. Note that most of the damage occurred on Oki-erabu, which was on the track of this typhoon. The neighboring islands, Yoron and Tokunoshima, which are situated 25 km east and 40 km west of the typhoon path, respectively, were remarkably

free from damage. The high wind zone of this typhoon was confined

to the eye-wall belt; these two islands were outside this zone.

In Tables 3 and 4 and also in the later sections, the

following are used:

* Major damage: floor areas damaged in a building

amount to more than 70% of total floor areas, or repairs amount to more than 50% of the building's

current price.

* Moderate damage: 20 to 70% of total floor area damaged, or repairs amount to 20 to 50% of current

price.

3-2
Minor damage: less than 20% of total floor area

damaged, or repairs amount to less than 20% of

current price.

Nmaj

: number of buildings with major damage.

Njjjq^: number of buildings with moderate damage.

^min : numDer of buildings with minor damage.

N

t

: total number of households in a district

1^ • : rate of the number of households in a district

= Nmaj/N

t

R^jQd : rate of the number of dwellings with major or moderate damage to the total number of household in a district = (Nmaj + Nmod )/N

t Rmin : rate °f the number of dwellings with damage to the total number of households in a district

= (Nmaj + Nmod + Nmin>/N

t Distribution of Damage to Dwellings in Oki-erabu Island

Oki-erabu Island is a flat coral-island having the topography shown

in Figure 7. About 40 villages consisting of a few decades or hundreds of households are loosely distributed there, mainly along the

coast

.

The statistics of damage to dwellings in those localities are shown in Tables 5 and 6. Damage rates R^: and in the Tables

are illustrated in Figures 7 and 8. Note that the damage rate

differs remarkably from place to place; R^j ' ^or exampl e » i- s 44.8%

in a western locality and in another locality in the central part

of the island it is only 2.0%. This is caused by topography and

ground roughness, as follows:

" The higher rate of damage in the eastern part of the

island appears to result from its flat farm land with

little ground roughness as shown in Photo 1. It was exposed to severe wind throughout the typhoon.

* Some localities southeast of Mt . Koshiyama in the

central part of the island suffered severest damage.

Those localities are near the top of gentle slopes

facing southeast and therefore were exposed to severe

winds from ENE-SE-S during the typhoon.

* In a few localities 2 or 3 km west of the area described above, damage was very light. These localities are in a shallow valley open to southeast between

Mt . Oyama and Mt . Koshiyama. They were not exposed to strong winds.

* Many localities along the coasts in China-cho suffered

heavy damage. These localities are situated on flat

gentle slopes of Mt . Oyama. They were exposed to high

winds from the sea or parallel to the coast. It also

might be that they were exposed to local high winds

around conical hills, as observed by Geiger (1950).

3-3
Description of Damage to Buildings

Damage to roofs, structural frames, and exterior wall openings of

wooden buildings were most frequently observed in our field survey

of tbe island.

Damage to Roofs

Most of the dwellings in Oki-erabu were of wooden frames with

corrogated galvanized iron sheet roofing. Usually those sheets are nailed to laths set on bed boards. In some cases, laths are nailed

to rafters directly.

The typical feature of damage to the former style of roofs is shown in Photo 2. The latter is shown in Photo 3. In the latter

case, secondary damage to ceiling and furnishings was also observed

As shown in Photo 4 the flying sheet metal can seriously

threaten nearby people and buildings.

Tile roofs were less common than sheet metal roofs. When tile

roofs had bed boards under them, damage was localized to ridges or eaves; the damaged tile area was generally smaller than damaged

sheet metal areas. But on tile roofs without bed boards damage was spread widely. Photos 5 and 6 show examples of damage to tile

roofs.

The roof structures of some sheet metal roofs were completely

demolished. This damage occurred because roof member connections

such as column-beam, beam-post, post-purlin, and purlin-rafter were

in most cases short-tennon connections and therefore could not resist to tensile forces generated by the suction pressure on the

roof. Photos 7 to 9 show some examples of damage to roof

structures.

Total Collapse

The traditional wood framing is a kind of rigid-connection framing

consisting of columns and rails. Rigid connections are made by

driving wedges into clearances to tennon holes (Photo 10). The

efficiency of such structures is said to be uncertain when main

members have rotted or the wedging has not been maintained. In recent times, the cross-sectional size of columns and rails has

become smaller. It is difficult to expect such a system to have

enough strength to resist severe winds. Although diagonal bracings

have been installed on many such houses, many of them were inefficient, as shown in Photo 11.

Damage to Exterior Wall Openings

On many houses, strong sliding shutters were used to cover windows.

These were definitely effective in protecting glazed openings from

wind-generated missiles. Grills were also effective, as shown in

Photo 13.

Aluminum sashes have been coming into vogue in Oki-erabu and

many of- the openings framed by those sashes were observed to have

been broken. The absence of shutters in turn lead to broken

windows, which then lead to roof loss (Photo 14).

3-4
Reinforced Concrete Buildings

Dwellings, hotels, schools, and administrative buildings were of

reinforced concrete; they were less than 4 stories. Damage here

was confirmed to openings shown in Photo 15. This induced secondary

damage to interior furnishings. Most of this damage was caused by

wind-generated missiles. Shutters or grills could have prevented

this loss.

Steel Buildings

Some factories and roofs of school gymansiums were of steel.

Roofings and sidings on these buildings were of slate or galvanized

iron sheet. Damage is shown in Photos 16 to 20. It was caused by

wind-generated missiles or insufficient strength of finishing material connections to back-up members. The insufficient strength of

connections was sometimes caused by the corrosion of steel materials. Note that the damage here was very diffusive. The aim of

steel-building design should thus be to prevent initial local

failures in some parts of the finishing and to prevent those

failures from spreading widely.

Concluding Remarks

The damage seen on Oki-erabu came from winds that are expected more

than once per 20 years in that area of the Pacific. Nor were the

winds of this typhoon extraordinarily strong. Nevertheless, the

wooden-frame construction could not cope sufficiently with today's

lighter-weight roofs. The significance of traditional preventives,

such as shutters, was forgotten and modernization went on without

checking its appropriateness. Since these two complaints about

structures can be repeated for many communities, it is important

that the lessons of Oki-erabu be remembered.

Acknowledgement

The authors wish to express their thanks to Fukuoka District

Meteorological Observatory, Oki-erabu Weather Station, Construction

Division of Kagoshima Prefecture, and local administration offices

in Oki-erabu for their cooperation in supplying information.

References

1. Geiger, R. , The Climate Near the Ground , Harvard Univ. Press,

1950.

2. Kagoshima Local Meteorological Observatory, Abnormal Weather

Report About Oki-erabu Tyhpoon , 1977.

3. Okinawa Meteorological Observatory, Okinawa Bulletin , No. 64,

1977.

3-5
TABLE 1: HOURLY CHANGES OF METEOROLOGICAL ELEMENTS AT OKI-ERABU

WEATHER STATION

Date Time Pressure

(mb)

Wind

dir.

Wind

speed

(m/s)

Precipitation

(mm)

Weather

Sept.

8

03 1003 .

1

NNE 3.9 0. 5

Uo N E

3 d A c: U . 3

uy 1UU3 .

3

NNE A Q 4 . y

A C U . 3 vi>

12 N L

13 l ah i a 1UUX .

U

ENE 3 . J y>

JLo NNt 3.X

2X 1 AAA C XUUU .

3

ENE "7 "3 1 A X . u V

*) A E"7 £ £b A. U

y A "3 yyo . a E b . J J3 . U 9

Ub ESE b . 2 X . 3

uy y y4 . x SSE b . b

"7 A

12 ESE b . b 0.0

Id A A O *3 yyy . i N E

1 "3 "7 12 . / 0.0. V

16 986 .

7

N E 12 .

6

o c 8 . 3

at, W

17 983 .

7

ESE 13 .

6

3 .

0

18 981 .

9

N E 19.6 11. 0

19 980 .

1

N E lb . 4

C A 3 . 0

20 976 .

3

N E 21.4 4 . 0

21 969 . 0 ENE •3 A A 24 . U

A A 4 . U

22 A A C C y4b . b ENE 3 c: c £3 A

*3 "J 2 J yio . 4 TO A

24 964.7 53.0 #

10 01 974.8 0.5

03 984.3 0.0 e

05 988. 5 0.0 e

3-6
TABLE 2: WEATHER RECORDS OF TYPHOON NO. 7709 AT OKI-ERABU

WEATHER STATION

lateorological elements Value Time (JST)

Min. sea-level pressure 907 . 3mb 2250

Max . wind ESE 39.4m/s 2210

Max. peak gust ENE 60.4m/s 2216

Total precipitation 177 mm 1820 (Sept. 8)-

0510 (Sept. 10) Max. daily precip. 168.0mm Sept.

9

Max. hourly precip. 61 . 0mm 2220-2320

(Sept. 9)
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3-9
TABLE 5: STATISTICS OF DAMAGE TO DWELLINGS IN WADOMARI-CHO

Locality

No. of

holds

Population

Nma3 N

mod N . min Rmaj Rmod

R . min

X • nCtUUlllal X 540 1666 110 101 150 20.4 39 .

1

66 .

9

2. Wa 116 414 17 30 40 14.7 40.5 75.0

3. Tedechina 253 810 31 64 113 12.3 37.5 82.2

4 . Kamitedechina 61 219 13 12 21 21.3 41.0 75.4

5 . Kibiru 156 504 57 60 59 36.5 75.0 112.8

vj • rvu i 1 ± y cuu j. 354 1461 124 117 \J ~J • o Qft 3D D QJ 7. Nishibaru 87 348 39 20 17 44.8 67.8 87.4

8. Degi 101 327 32 15 23 31.7 46.5 69.3

9 . Inobe 31 107 9 7 8 29.0 51.6 77.4

10. Azefu 105 377 22 16 75 21.0 36.2 107.6

XX c LlCUJL X 69 289 4 19 21 5.8 33 .

3

63 .

8

12. Tamashiro 180 589 25 115 45 13.9 77.8 102.8

13. Oshiro 69 232 22 20 11 31.9 60.9 76.8

14. Minagawa 50 132 9 16 28 18.0 50.0 106.0

15. Furusato 65 188 14 7 33 21.5 32.3 83.1

xOt ucni.jj.io H P q 1 \J i n fid A 17. Kamiuchijiro 45 118 6 9 31 12.9 33.3 102.2

18. Goran 35 124 3 6 17 8.6 25.7 74.3

19. Taniyama 26 89 4 0 22 15.4 15.4 100.0

20. Nishi 25 94 3 11 7 12.0 56.0 84.0

21. Nagamine 54 175 9 8 17 16.7 31.5 63.0

22. Sena 64 200 10 10 14 15.6 31.3 53.1

Total 2531 8615 572 665 879 22.6 48.9 86. 3

3-10
TABLE 6: STATISTICS OF DAMAGE TO DWELLINGS IN CHINA-CHO

Locality

No. of

housenoIds

Populatxon

Nma 3 Nmod N . min R . ma j mod R . min

( *)

1. China 358 1075 68 78 40 19.0 40.8 52.0

Z • I ciJS.(jITtU X*iX ccJJ 7Z Qy 7

1

z X

7 q n ->y . d 1 H . 0

7 ^"\+*£M^/T^ j . utsuge 7C, o / / X Xz1 7 7 n n z U . U . y

c; "7 1 J / . 1

*i . lUKULOKl C.Q 5(11 QO

7 Q o z o . o 4yA Ct O. ^ 5. Sumxyoshi. lo1 P.7/ RQQ AO 7 1

/ X

7 c n ZD./ /I "7 1 OD.U

6, Masana 124 486 25 48 38 20.2 58.9 89.5

7 » Tainins J JJ XUX44 X Xz 17 7 X Jz 7Q 7 T C yb . / o . onxuiosnxxo 7QZ J

c 1IdA Qo 17 7 X / • z 77 /l / Z , *i XUU . U

7. KdJ.fi -L SflllTO COdo 1 7 P. X / O 7

1

zX D JO7<^ 7. Z **D . D

qiCSX n. U iu. 1 O juinjo CVci n -1a 77 Z D 11* 7 "3 Z J Z7 J7 7 C 7Qzy Q. y CQ 7 ->y . /

iXUnc 7 j . z

11. Gushiken 43 143 4 i 16 9.3 11.6 48.8

12 • Ak.axni.ne K 1 1 J J X

7 1 Z X ^ . u ji.U ci r» 13 o Saozu C "7 J /

1 7 "5 X / J

c

_> z xy Q Q y o ZO*± XD1 C ZZ7 7 Qy

1 c o 13. J 44d . y 15. Kamihirakawa ZXU C C Q 4z z4 ^0 . U OA A jy . u DU . j

16. Shimohirakawa 63 184 14 6 n 22.2 31.7 49.2

x / . i a j a jy7Q Xz1 7 CD XD1 c o

1 c X D JO. 7 Q CJ JJ.O C "3 Q 07 7 yz^ . j 18. Ashikyora 148 448 29 64 21 19.6 62.8 77.0

X_7. JS.UX (JilllJS. X ZD XX J7 Z7 HD • J_ OH • x 20. Serikaku 278 807 107 89 20 38.5 70.5 77.7

21. Kogome 231 766 86 64 61 37.2 64.9 91.3

22. Oyama 77 77'

Total 2778 8335 725 691 576 26.1 51.0 71.7

3-11
FIGURE 1: TRACK OF TYPHOON NO. 7709

3-12
3-13
3-14
3-15


1

3-18

3-20


PHOTO 7: GABLE ROOF , HALF OF WHICH WAS BLOWN AWAY

3-23
3-24
PHOTO 13: GRILLS ON EXTERIOR WALL OPENINGS
PHOTO 16: DAMAGE TO SHEET METAL SIDING OF A STEEL FRAME FACTORY

PHOTO 18: DAMAGE TO SLATE ROOFING OF A STEEL FRAME FACTORY

3-26
PHOTO 19: DAMAGE TO SHEET METAL ROOFING OF A GYMNASIUM

3-27

4

RECENT DEVELOPMENTS IN ATMOSPHERIC REMOTE

SENSING AND THEIR IMPLICATION FOR WIND

ENGINEERING

William H. Hooke

Atmospheric Studies Program Area

Wave Propagation Laboratory

Boulder, Colorado

NOTE: FIGURES NOT AVAILABLE FOR THIS PAPER

Recent Developments in Atmospheric Remote Sensing and

Their Implications for Wind Engineering

This paper reviews recent progress in developing a most promising

source of meteorological observations — remote-sensing systems

that measure the effect of the atmosphere on optical, radio, and

acoustic waves. We begin by highlighting several major needs

within the wind-engineering community, and then discuss the

relative advantages and disadvantages of remote sensing and

in-situ approaches to meeting these needs. We find that in general

remote-sensing methods offer enormous advantages in area and volume

coverage, spatial and temporal resolution, representativeness, technological feasibility, and cost. We then summarize the substantial

impact of atmospheric remote-sensing on wind engineering, by identifying and isolating several specific areas of present application,

examining each in detail. These areas of application include the

development of devices for the study of high winds associated with

severe storms, the detection of low-level wind shear at airports,

and the transport and diffusion of pollutants. Remote sensors are giving us our first real look at the circulation of air within

thunderstorms and hurricanes, even on scales as small as the tornado

vortex. Arrays of pressure sensors, lidar, and even modified airtraffic-control radars can provide much information on low-level

wind shear in the vicinity of airports. And lidars are now able to track pollution plumes, determine the velocity of the air within the

plumes, examine even their chemical composition. Then we look at possible future applications of remote-sensors to wind-engineering

problems. These applications include the quantitative study of

plume diffusion over irregular terrain and urban areas, the use of

remote sensors in wind power-park siting, and the use of lidar to study the influence of high-rise structures on urban airflow. An appendix contains a review of existing remote-sensing capabilities

and developments.

Remote-Sensing and In-Situ Approaches to Meeting Wind Engineering

Data Requirements

There are many ways we could classify the data requirements of the

wind-engineering community. However, for present purposes we shall

satisfy overselves by considering wind-engineering data requirements

in four categories:

° structural design

° transportation

0 pollutant transport and diffusion

° wind power

Structural design is a major concern. In the United States

alone, wind damage to buildings and structures amounts to a billion

dollars per year; much of this could be prevented were more

detailed wind climatologies available to foster improved design

practice. Particularly vulnerable are structures whose design

extends preexisting technology — high-rise skyscrapers with glass

clading, long cable-stayed bridges, domed and open sports stadia,

and high towers. Also economically important because of their large

numbers. are smaller buildings, such as residential structures, particularly those in high-risk areas (for example, tornado belts and

the southern Atlantic and Gulf coastal zones). In addition to problems posed by the need to design structurally sound buildings

that won't collapse or be destroyed by high winds, there is also

the need to design functionally adequate structures, in which the

elevators will continue to operate in high winds, which won't

oscillate or sway excessively, and which do not create unacceptably

4-1
high street-level winds or snow-drifting in their immediate

vicinity. On the other side of the coin, the risks inherent

with strong winds lead to over-design of structures such as nuclear power plants, which in some areas must be constructed to withstand a tornado. Because tornado maximum winds and gusts

remain ill-defined, reactor design must continue to be correspondingly conservative. This over-design of such expensive structures

represents a substantial incremental cost.

Proper structural design requires accurate, reliable data, for

long periods of time, on boundary-layer winds. Near-surface winds

are relatively well documented, but there is a pressing need for

wind and turbulence statistics at heights of 100-1000 m. There is

particular interest in data from extreme-wind conditions associated

with hurricanes, tornadoes, severe convective storms, downslope

winds, and low-level jets. Finally, measurement of these quantities in urban high-rise areas is particularly desirable.

Surface and airborne transportation are quite vulnerable to high winds. For example, the foothills of the Rocky Mountains

frequently experience severe downslope winds able to overturn trucks

and campers, making it necessary to close the roads to such traffic.

Far more serious and widespread, however, is the threat posed to aircraft takeoffs and landings by low-level wind shear. Curiously

enough, this hazard can be greater for the most modern jetliners

than it is for the smaller, ligther aircraft of general aviation.

This is because the latter can respond quickly to changes in

air-speed, while the larger aircraft require more time. In some respects, a jetliner landing may more properly be referred to as nearly a controlled crash, with the pilot anticipating future con- ditions along the glide path at every point in his descent, changing

control settings some tens of seconds before those changes are to be

in effect. This procedure works satisfactorily when the meteorological conditions prevailing over the airplane glide path are homogeneous and stationary. It is hazardous, however, in the

highly-variable and inhomogeneous thunderstorm environment, where

gust fronts and thunderstorms downbursts of quite limited extent

are difficult to detect. Here the need is for operational systems

able to monitor such dangerous phenomena and provide timely warnings

to the pilots.

Shipping and boating are also vulnerable to winds, both

directly and indirectly, through wind effects on sea state. The

trend towards larger and larger ships, while in some respects minimizing such risks through the incorporation of improved design, in

other respects enhances the risk, through increases in the size

of cargo spills, for instance. Here the need is for accurate

information on sea state — its present state and forecasts for

the future, often for periods as long as several days.

Pollutant transport and diffusion pose a whole new set of

engineering needs and data requirements. However, here the need

is greatest for information pertaining to low-wind conditions, when

pollutants do not disperse rapidly but instead build up to dangerous

concentrations, and when the effects on the airflow of terrian and

buildings, are most pronounced. There is a need for more measurements of winds and turbulence at heights above 100 m, and for more observations of the diffusion of pollutants on a regional scale,

particularly in urbanized areas.

Finally, the nation's energy crisis has made wind power a subject of considerable concern. Engineering interest in efficiency

must be reconciled with a need for ruggedness, or for an ability to

4-2
forecast weather events and to protect a threatened wind generating

system (through feathering of propeller blades, for instance). To provide an optimal reconciliation requires more information pertinent to wind siting, including extensive wind and turbulence climatologies for a large number of promising sites. In these cases also,

data to substantial heights are needed, in order to accommodate

design of the large systems contemplated for operational use.

Underlying all these needs are certain common threads. First,

we desire data over extensive regions. To understand the ultimate

fate of pollutants in the urban environment or to monitor weather

conditions along an entire jetliner glide slope into an airport

requires this extensive coverage. Note that we require the information not merely over certain substantial areas, but also over large

volumes. For example, since wind speeds increase with height in

the boundary layer, the effects of winds on buildings are greatest

near the tops of the buildings. Similarly, it is hard to predict

pollutant transport and diffusion without knowing the boundary-layer

depth (^1000 m). The critical portion of the aircraft glide slope

is between heights of 100 m and 1000 m. Second, within this volume

of coverage we are interested in both representativeness and

resolution. That is to say, on the one hand we are interested in

knowing that our estimates of a certain parameter, e.g., wind speed, or maximum turbulent gust intensity, are representatives or characteristic of the area as a whole, while on the other we need fine

spatial and temporal resolution, in order to carry out such tasks

as monitoring tornadoes and estimating tornadic winds, or to define

the effects of individual buildings and terrain features on pollutant transport and deposition. Third, for engineering purposes we generally require (or at least prefer) site-specific information.

It does little good, for example, to estimate the prospects for

wind power from the Pali lookout on Oahu by using data from Erie,

Colorado. Fourth, and finally, we must often take the measurements

without interfering with the medium. A most obvious example concerns airport operations. To place instrumented tall towers in

the glide path is no solution.

With these preliminaries, we are now ready to contrast the two basic approaches to measuring wind — in-situ instrumentation and

remote sensors. As its name implies, the former method involves

sampling of the medium in the immediate vicinity of the instrument,

using some direct effect of the variable in question in order to measure it (for example, the rate at which an airflow turns the

cups of an anemometer immersed in it, or the cooling effect of airflow past a hot wire). On the other hand, remote-sensing techniques

use the interaction of radio, optical, and acoustic waves with the

medium through which they propagate. This interaction depends upon

the properties of the medium in a well-defined way. Thus, for

example, sound waves propagate through air at a speed that is

determined by the local temperature and the local wind velocity.

Likewise the Doppler shift introduced into a radio wave by scattering from atmospheric turbulence is proportional to the velocity

of the turbulent eddies relative to the radio source. Absorption

of an optical wave propagating through a mixture of gases depends

upon the composition of the mixture and its density as well as the

frequency or color of the incident light.

Passive remote sensors use observed properties of existing

radiation of natural origin to draw inferences concerning the

medium. For instance, microwave radiometry uses the spectrum of

microwave emissions from the clear sky to determine temperature

and humidity. Although they are attractive for military purposes,

where detection of the sensor itself is a concern, passive sensors

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are generally much less desirable than active sensors , which supply

their own wave radiation. This added variable gives active sensors

the edge in flexibility, improved spatial resolution, and scanning

capability.

In general (though not always), because they are typically

more complex, remote-sensing systems cost more to make the first

measurement than do their in-situ counterpart. However, they offer

the major intrinsic advantage that they do not perturb the medium

being measured (provided, in the case of the active remote sensors,

that we neglect the slight heating of the medium necessarily from

any absorption of the probing wave). By contrast, in-situ instruments such as cup anemometers or thermistors necessarily interfere

with the flows of interest, and the platforms on which they must be

mounted (e.g., towers or aircraft) introduce still more significant

obstacles to the flow. Moveover, where remote-sensing equipment

really excels is in the cost of incremental measurements, because

most such equipment offers a scanning capability which permits

rapid sequential interrogation of many measurement cells within

large volumes. Thus the cost of remote sensors should be compared more appropriately to the cost of in-situ sensor arrays in

aggregate.

Let us now look at the particular advantage of improved spatial

and temporal resolution and areal coverage which remote-sensing

systems offer, using dual-Doppler radar data on boundary layer winds

by way of illustration. Figure 1 shows dual-Doppler radar data on the wind field at a height of 300 m above St. Louis on a summer afternoon during the 1975 METROMEX experiment. The data were taken

using chaff dispersed from a light airplane; Doppler shift of the

radar echoes yields the wind. The mean wind field has been removed

from the data in order to reveal the small-scale structure. Lines

of convergence and divergence are evident from the figure, and

appear to be associated with an elongated heat island extending

from Granite City across the St. Louis central business district

(Kropfi and Kohn, 1977, 1978). Although the data are of considerable meteorological interest, we show them here simply to indicate

the ability of dual-Doppler radar to provide data with appreciably

better than 1-km resolution over an area greater than 100 km^. An individual anemometer with telemetry and recording system would

cost perhaps $1 K if installed on the surface, much less than the

$500 K cost of the dual-Dopler radar system. But the array of some

300 anemometers required to reproduce the pattern revealed here

would cost $300 K, a figure beginning to approach that of the

radars. And if we were to take full advantage of the radar resolu3 tion (the equivalent of 10 anemometers) the cost of the corresponding anemometer array would be double that of the radars. But the

comparison does not end here. The anemometer system described

above could give only surface data, whereas the radar data were

taken at a height of 300 m. In fact, the radar system provides

complete altitude information, up to the top of the boundary layer,

as shown in Figure 2, which shows the vertical circulations asso- ciated with the horizontal airflow pattern of Figure 1. To provide

this information would require four times as many anemometers. And

this ignores the major cost — that of the towers on which the

anemometry would have to be mounted. Figuring a cost of $10 /tower

for 300 1000-m towers (an extremely conservative figure) we find a total system cost for the anemometry of $300 x 10 , some 600 times

the cost of the radar system (a system that doesn't interfere with

the airflow or blight the landscape, and which is transportable as well). Note moreover that the radars quite naturally provide data

throughout the depth of the mixed layer, because chaff will dispere

throughout this volume no matter how deep it happens to be (up to
several km in extreme cases). By contrast, the technology does

not exist for constructing towers meeting this more stringent

height requirement.

Remote sensors also offer much improved temporal resolution.

We might imagine, for example, that rather than put up 300 tall

towers, we could deploy 300 radiosondes, which would each provide

a single wind profile for the 70-min period shown in Figures 1 and

2. But the cost of radiosonde expendables is some $ 1 00/prof ile , so that the increment cost of one data set represented by Figures 1 and 2 would be $30 K. The incremental cost of the dual-Doppler

radar data is orders of magnitude less than this value. Moreover,

the data shown represent averages of seventeen such data sets, and

to reproduce this by radiosondes would have cost some $500 K. Besides, to acquire a single radiosonde profile through the atmosphere can take up to hour. Both these factors preclude more

frequent measurements that would give us a better look at smallscale and short-lived changes in atmospheric conditions on pollutant transport and diffusion, wind gustiness, etc. By

contrast, in the dual-Doppler example given above, the entire

velocity field can be mapped out every few minutes. The result

is a tremendous improvement in the picture available of the time

history of such events.

Finally, the deployment of in-situ instruments in necessarily

sparse arrays gives rise to questions concerning the extent to which the reported observations accurately reflect conditions over

the whole of the area being monitored by each instrument. Instrument siting remains something of a black art, whose precise influence on experimental results is poorly understood. Remote sensors alleviate these difficulties in two ways. First, sensors such as the dual-Doppler radars provide many data points over a large area;

in some cases this permits evaluation of the representativeness of

various sites. Second, remote-sensing data necessarily represent

a line, area, or volume average. In the case of the dual-Doppler

radars, for example, each wind vector shown in figures is the

spatial average over a second volume some 300 m on a side. Thus

the remote sensor acts as a low-pass spatial filter on the data,

smoothing out smaller-scale fluctuations, some of which reflect

siting problems.

All this might seem to belabour an obvious point. But we are trying to emphasize the fact that remote sensors are not merely

making a 10%-20% improvement in our observational capabilities.

Rather they make a quantitative difference amounting to orders of

magnitude, and they so greatly extend our capability that they

imply qualitative changes in our insights into atmospheric behavior

and our ways of approaching engineering design. The result is that

remote sensors are the instruments of choice, where available, for

field experiments, because they provide a look at small-scale,

short-lived phenomena, over wide areas, with high reliability.

Remote Sensors Applied to Storm Research

Dual-Doppler radars have also been used to study the circulation in severe convective storms. Such work is easier in the sense that

nature supplies the tracer producing the radar echoes; no aircraft

dispersal of chaff is required. Figure 3 shows an early result of

such a study — a vertical plane cut through the convective storm.

The data reveal a number of the storm circulation features, including the major updrafts and downdrafts, the surface cold-air outflow

and gust front preceding the rainshaft itself, and the airflow

producing the thunderstorm anvil. A current model of thunderstorm

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circulations is included for comparison. Although the present data

show only the gross features of the thunderstorm, it should be possible to improve the radar performance to the point where we can examine cloud microstructure and short-lived aspects of cloud

dynamics. Radar studies of this type have been a major part of the

National Hail Research Experiment (NHRE), which has focused on the

dynamics and behavior of hail-producing severe convective storms. Multiple-Doppler radars have also been used at NOAA's National

Severe Storms Laboratory (NSSL) to observe and study tornadic

storms. In addition to carrying out basic research, NSSL has

developed single-radar Doppler approaches to the detection of

tornadic meso-cyclones within severe storms, regions that are associated to very high probability with tornado development and

maintenance (Lemon et_ al_. , 1977 ; Brown et_ a_l . , 1978). Figure 4 shows a typical tornado vortex signature (TVS). Although at these

ranges of the resolution of the radar is less than that necessary

for isolation of the tornado itself, TVS is proving to be a reliable indicator of tornadic activity. NOAA is now considering

making such Doppler information operationally available in stormprone areas.

The Wave Propagation Laboratory (WPL) of NOAA has also

developed a frequency-modulated continuous-wave (FM/CW) infrared

(IR) Doppler lidar which has been used to measure vortex velocities

in dust devils and waterspouts (e.g., Schwiesow and Cupp, 1976;

Schwiesow et_ al_. , 1977), using truck and light-aircraft platforms.

On a stronger aircraft, the instrumentation would be suitable for

making profiles of tornadic-vortex velocity. Such data would be

of considerable value to those responsible for designing structures

(e.g., nuclear reactors, schools) to withstand torando hits.

Figure 5 shows a schematic of the approach used to estimate waterspout velocities; Figure 6 shows some of the resulting data.

On a much larger scale, NOAA' s highly sophisticated P-3 storm

research aircraft uses three radars (nose, belly, and tail) to map

hurricane precipitation fields. Attempts are underway to Dopplerize

these radars in order to provide data on hurricane wind fields.

When completed, these systems should open new horizons in hurricane

research.

Remote Sensors Applied to Low-Level Wind-Shear Detection

As mentioned in the introduction, strong shears in low-level

winds constitute a major hazard to commercial aircraft takeoffs and

landings. Roughly one aircraft disaster per year results from this cause, with an attendant loss on the order of one hundred lives, tens of millions of dollars in property damage, and millions more

at stake in the ensuing lawsuits. The first efforts tc develop

remote-sensing systems for monitoring this hazard focused on lowlevel wind shear of wind areal extent, associated with frontal

passages. For such wind shear it suffices to provide a single

vertical profile of the wind speed in the airport vicinity. The

Wave Propagation Laboratory developed a prototype system for windshear detection at Dulles International Airport which relied

primarily on acoustic-sounding Doppler techniques for wind-speed

determination. Since the acoustic sounder could not be used in

heavy rain, because of noise on the sounder antenna, the Dulles

system incorporated a radar unit for making VAD scans and Doppler

wind determinations. The resulting hybrid system provided an allweather capability (Hardesty et a_l. , 1977).

4-6
As the work proceeded, however, it became apparent that much

of the dangerous low-level wind shear was very localized in nature.

For example, a careful analysis of the meteorological causes of the

Eastern airlines crash at Kennedy International Airport in June 1975

showed that it resulted from a thunderstorm downburst, perhaps no more than a few km in horizontal dimension (Fujita and Byers , 1977).

During such severe-weather events, wind data from single height profiles can no longer be taken as representative of the aircraft glide

slope as a whole. What is needed instead is some sort of area coverage or scanning capability. The Wave Propagation Laboratory

has developed several approaches to this monitoring needed. The

first involves the use of cheap pressure switches to monitor the

sharp surface-pressure rise associated with the passage of thunderstorm gust fronts and downbursts. By deploying such instruments in

dense arrays, linked by phone lines to a central station, it is

possible to monitor the progress of such atmospheric structures as they approach, pass over, and leave the airport area. Figure 7 illustrates this concept. Bedard et_ al_. ( 1977) and Bedard and

Hooke (1978) have described operation of a prototype system used at Dulles International Airport for this purpose. It is worth noting

that the pressure switches used are remote-sensors in two senses.

First, unlike thermometers or anemometers, pressure sensors measure

the integrated mass of the air column above the sensor; as a result,

they are less subject to local influences. Second, unlike thermometers or anemometers, pressure sensors can detect gust fronts even when mounted inside buildings. Both features combine with the

sensor's mechanical simplicity to make them an ideal choice for

gust-front monitoring. A second approach to the problem involves

the use of scanning remote-sensors. Chadwick et al. (1976a and b)

describe one such technique, based on an extension of FM-CW radar

sounding. And recently (Strauch, private communication) WPL has

found that the new-generation FAA air-traffic-control radars are sufficiently sensitive to provide clear-air echoes and wind-velocity

estimates. Lidar methods have also been considered.

Lidar Studies of Pollutants

This is a rapidly growing discipline, and a difficult one to review

completely. Hall (1974) gives a clear exposition. The most commonly

used lidar technique is the scatter of pulses from aerosols; work

in this area dates back to the early 1960's. Figure 8 shows a time

history of plume development monitored by one such pulsed lidar

system, shown schematically in Figure 9. As indicated in Section 3,

there now exist FM-CW lidar systems for estimating velocity fields

within regions of high aerosol density; a number of coherent pulsed

lidar systems are also under development for this purpose. One plan

includes the possibility of global wind monitoring by satellite in

this way. Finally, a variety of techniques, including Raman and

differential absorption, are being used to determine quantitatively

concentrations of gaseous pollutants such as CO, N0

x , SO2, and O3.

In the future such devices should offer quantitative diagnostics

for air-pollution photochemical studies.

Future Applications of Remote-Sensors to Wind Engineering Problems

Remote-sensor application to wind engineering is in its infancy.

Many remote-sensing systems have hardly advanced beyond the

prototype stage. Still ahead lie the majority of their research

applications, to say nothing of long-term operational uses. Moreover, behind these first-generation systems lie even more advanced

technology, promising greater spatial and temporal resolution,

greater ranges and areal coverage, and providing measurements of

4-7
quantities other than wind — e.g., humidity and temperature. With

these will come more powerful, definitive wind-engineering studies.

For example, it should be possible to develop mono-dispersed

aerosol generators that could provide artificial plumes for study

by lidar systems — study that could then be quantitative rather

than merely qualitative, revealing plume concentrations as well as plume positions. Such systems could then be used to study plume

diffusion over irregular terrain, in mountainous regions slated

for mineral or energy development, and over urban areas. They

would thus provide more reliable environmental impact assessment,

both through direct measurement and through enhanced data sets fostering more accurate models.

With their scanning capability and areal coverage, remote sensors are ideally suited for studies of wind power-park siting.

The sensors could monitor winds and turbulence in large volumes of

a prospective wind power-park. From the data it would be possible

to select those specific sites offering the best tradeoff between

maximum averaged wind speeds and minimum gustiness.

In another important application, lidar could be used to study

the interaction of airflow with high-rise buildings and other manmade structures. Currently, these structures are designed using

only data from scale models in wind tunnels. There has never been

a major attempt to verify model predictions by comparing the model

results with measurements in the field. We do know, however, that

the existing approach produces buildings with many design flaws.

The failures of glass cladding in Boston's John Hancock building

and in the Sears structure in Chicago constitute important although

by no means solitary examples. It is now possible in principle,

however, to make definitive comparisons between wind-tunnel models

and flows observed in the field using high-resolution lidar techniques. In this way it should be possible to develop a new understanding of the relation of wind tunnel flows to those actually

obtained, to develop new scalings of wind-tunnel findings that

might be more appropriate, and generally to raise user confidence

in the modeling procedure.

Far more interesting than these individual efforts, however,

in terms of its long-range potential for weather observations and

forecasting in general, is the rise of new system concepts of

improved local weather prediction. These concepts incorporate

recent advances in remote-sensing, mesoscale numerical modeling,

data processing, and information dissemination. MAA is currently

examining one such proposal, entitled PROFS (Prototype Regional

Observing and Forecasting Service), which calls for exploratory

development of such a service in the Denver area and then expansion to a nationwide system, focusing on urban areas. Such systems

would attempt to improve local weather services to meet increasingly

urgent needs of aviation, agribusiness, the construction industry,

and the general public. New system concepts have incorporated a great concern with wind engineering problems, including hazardouswind forecasts and monitoring, and pollution dispersal. Should

such a concept be implemented, it would mean a tremendous increase

over the long term in the quality and quantity of routine National

Weather Service data available to the wind engineering community.

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39. Russell, P. B., and Uthe, E. E., "Sodar Network Measurements

of Regional Mixing Depth and Stability Patterns for an Air

Quality Model," Proc. Joint Conference on Applications of Air

Pollution Meteorology , American Meteorological Society, Boston,

Mass. (1977).

40. Schwiesow, R. L. , and Cupp, R. E., "Remote Doppler Velocity

Measurements of Atmospheric Dust Devil Vortices," Appl. Optics , 15, 1-2 (1976).

41. Schwiesow, R. L. , Cupp, R. E. , Post, M. J., Abbey, R. F., Jr.,

and Sinclair, P. C. "Velocity Structures of Waterspouts and

Dust Devils as Revealed by Doppler Lidar Measurements," Pre - prints Tenth Conf. on Severe Local Storms , American Meteorological Society (1977).

42. Strauch, R. G. , Derr, V. E., and Cupp, R. E., "Atmospheric

Temperature Measurement Using Raman Backseat ter , " Appl. Optics , 10, 2665-2669 (1971)L

43. Strauch, R. G. , Derr, V. E ., and Cupp, R. E. "Atmospheric

Water Vapour Measurement by Raman Lidar," Remote Sens. Env. , 2, 101-108 (1972).

44. Wesely, M. L. , "A Comparison of Two Optical Methods for

Measuring Line Averages of Thermal Exchanges above Warm

Water Surfaces," J. Appl. Meteorol. , 15, 1177-1188 (1976).

45. Westwater, E. R. , "Improved Determination of Vertical

Temperature and Humidity Profiles of the Atmosphere by a Combination of Radiometers and Active Ground-Based Sensors"

(Unpublished Manuscript, 1978).

4-11
46. Wyngaard, J. C, Kaimal, J. C, Ochs, G. R. , Hill, R. J. , and Sorenson, D. C. , "An Optical Heat Flux Experiment,"

Proc. Fourth Symposium on Meteorological Observations and

Instrumentation American Meteorological Society , Boston,

Mass. (1978).

Appendix: Recent Developments in Remote Sensing

It is appropriate to include a partial listing of recent advances

in atmopsheric remote sensing, since the method of presentation used

above gives a somewhat fragmented view of the arsenal of techniques

now available. In this rapidly evolving field, two of the latest

major reviews available (Derr, 1972; Little, 1972) are somewhat out

of date. However, our treatment is not by any means exhaustive; we have simply attempted to supply the reader with a sampling of the

material available and a few references to direct his first steps.

Acoustic Echo Soundings

Acoustic echo sounding is analogous to radar except that it uses acoustic pulses to interrogate the atmosphere, detecting acoustic

energy scattered by atmospheric turbulence. The subject dates back

to papers of McAllister ejt _al. ( 1969) and Little ( 1969), although

even these have antecedents in both western and Russian work.

Brown and Hall (1978) have recently reviewed the field thoroughly.

From the start, acoustic echo sounding has been used to advantage

in visualizing conditions in the atmospheric boundary layer —

studying the development of the convective boundary layer, monitoring inversion height, etc. In recent years, however, interest has

turned to the use of acoustic echo sounding as a quantitative tool.

One piece of information yielded readily by the frequency shift of

the echo returns is the atmospheric wind velocity. The technique

has been used for years; a recent system is described by Kaimal

and Haugen ( 1977). Hardesty et al_. ( 1977) describe a quasioperational system installed at Dulles International Airport to monitor low-level winds and wind shear. More recently, researchers

have begun to investigate the potential of acoustic echo sounding

for other quantitative information, including height profiles of

the turbulence structure function Cj,^ and C v

, surface heat flux,

surface stress, and Monin-Obukhov length, and rudimentary methods

of temperature profiling involving heavy use of similarity theory

(Brown, 1978) or phase-coherent techniques (Brown, private communication). Many of these quantities are of fundamental importance

to wind engineering design.

Although in many cases acoustic methods for measuring these

parameters do not seem as promising as radar or lidar techniques,

they are worth pursuing because of the lower cost of acoustic

methods. This lower cost is making it practicable to install

acoustic echo sounders in large arrays, to determine mesoscale

structure in PBL height, for example (Russell and Uthe, 1977).

Various pattern recognition schemes are currently under development for use in automated processing of acoustic-sounder facsimile

records (Clark et al., 1977; Hall, 1978).

Radio Techniques

The use of pulse Doppler radars in PBL studies has been described

at length in the body of the paper. For the most part, the work

cited has involved the use of chaff; however some installations,

e.g., NOAA's National Severe Storms Laboratory, have demonstrated

clear-air pulse dual-Doppler radar capability out to ranges of 40

or 50 km (Doviak and Jobson, 1977). As a rule, the work has

4-12
involved the use of two radars to provide two components of the

wind directly; the third velocity component has typically been

inferred, by assuming incompressibility and using the equation of

continuity. However, the use of multiple radars is becoming

increasingly frequent, to provide increased area coverage, to provide redundancy and therefore verification of the dual-Doppler

approach, and to examine the influence of the lower boundary condition on data recovery. Other research directions are also being

pursued in response to the needs of the operational community.

Lemon et al. (1977) report use of single Doppler radars for detection of tornado vortex signatures. Strauch (private communication)

is currently exploring a variety of single-radar techniques for

estimating both horizontal wind components. It has been pointed

out recently, by Gal-Chen and by Leise (work in progress) that from

the dual-Doppler radar fields and knowledge of the surface boundary

conditions it is possible in principle to deduce mesoscale temperature fields from the Doppler radar data. Tests are currently underway to determine whether the approach is practicable. If it is, or if it is true that mesoscale meteorological development is much

less sensitive to thermal effects than the initial velocity fields

j(Anthes, private communication), then radars offer tremendous

potential for mesoscale models.

FM-CW radar has been in use for the past decade as a powerful

tool for studying the clear-air boundary layer with high resolution

(e.g., Gossard et al. , 1970). In recent years, however, the utility

and flexibility of the technique have been greatly increased by the

development of a Doppler capability for determining wind velocities,

using a VAD scan (Chadwick e^ al_. , 1976a, b).

Another radar advance has been the use of VHF radars for probing the clear troposphere and lower stratosphere (e.g., Green et al., 1975); Balsley e_t_ al_. , 1977 , and the references therein).

While the technique has not been applied to the problem of lowertropospheric profiling in any detail, it does appear that it can

be (Balsley, private communication) and we can anticipate that it will be in the future.

Optical Techniques

Foremost among the optical techniques is lidar, or laser radar.

The most common type is the pulsed lidar, which senses echo returns

from aerosol scattering; work in this area dates back to the early

1960's and has been reviewed by Collis (1969) and Hall (1974).

['Recent papers of interest include a paper comparing FM-CW radar and

Uidar returns (Noonkester e_t_ al_. , 1974) which shows, as one might

|iexpect, that the PBL presents a different characteristic appearance

Jto the two sensors, and a paper describing lidar observations of

the convective boundary layer (Kunkel et_ al_. , 1977).

As in the case with both the acoustic sounders and the radars,

however, lidars offer tremendous potential for quantitative measurement as well. Huffaker (1970, 1974, and the references therein)

describes the principles of Doppler lidar velocity determination.

Benedetti-Michelangeli et_ al_. ( 1972) describe preliminary tropospheric wind measurements using lidar. Schwiesow and Cupp (1976)

and Schwiesow et al. (1977) have used truck-mounted and airborne

infrared CW Doppler lidar to measure atmospheric velocities in dust

devils and waterspouts, paving the way for analogous studies to tornado vortices. And in a recent work Huffaker e_t al_. ( 1976) put

forth their concept for satellite-based global wind measurement

using pulsed coherent lidar systems.

4-13
Lidar also offers some promise for temperature profiling.

Fiocco et_ al_. (1971), using molecular thermal broadening of the

lidar return, and Strauch et_ al_. (1971), using Raman-scatter lidar

estimate of N2 scale heights, provided early estimates of atmospheric temperatures. In more recent work, Cooney (1977) has used

rotational Raman scattering to estimate temperature profiles.

Optical sensing seems to hold the greatest long-term potential

for sensing gaseous pollutants, where optical radiation offers in

principle a tremendous chemical specificity. Unfortunately however,

such spectroscopic effects are often very weak, and, despite its

glowing future, optical sensing has only a brief and rudimentary

history. The work of Strauch et al . (1972) illustrates the

approaches, which include Raman spectroscopy and differential

absorption. Some recent work, including that of Swedish and

Japanese groups, may be found in the Invited Papers and Conference

Abstracts of the Eighth International Laser Radar Conference, held

at Drexel University in June 1977 and sponsored by the American

Meteorological Society and the Optical Society of America.

Lidar is currently being used to probe cloud microphysical

processes. For example, Piatt (1977) has used lidar to study phase

changes within an altostratus cloud. Derr (private communication)

has found recently that cloud glaciation can occur at a much earlier

stage in cloud development than previously thought.

Other optical techniques for remote probing include line-ofsight observations of light-beam scintillations which have been

used to measure wind (Lawrence et al., 1972), surface convergence

(Kjelaas and Ochs, 1974), and surface heat flux (e.g, Weseley, 1976;

Wyngaard ei^ aj_, 1978).

Hybrid Systems

As remote-sensing matures, attention is turning away from the

individual technologies themselves, and emphasis is shifting to the development of hybrid systems tailored to particular meteorological tasks. Here we discuss a few of these; however, from

recent work such as that of Noonkester et al . (1974) comparing

FM-CW radar and lidar returns from the boundary layer, and from

recent joint radar and lidar studies of clouds (Derr, private

communication), we can expect that much more such research is in

the offing.

Microwave radiometry has long been used to retrieve temperature and humidity profiles in the atmosphere from measurements of

atmospheric microwave brightness at different frequencies and

zenith angles. Recently, Westwater (1978) has shown that should

radiometric data be supplemented with data on inversion height

then the radiometric inversion techniques can yield much greater

resolution of the humidity and temperature variations in the vicinity of the inversion layer itself. Figure Al shows an example

(from a paper study) of the advantages of such a system. Note that

since estimates of inversion strength are unnecessary, these supplementary data could readily be supplied by acoustic sounders or FM-CW radar.

Another hybrid technique of considerable novelty is the so- called RASS (Radio-Acoustic Sounding System) approach, which

provides temperature profiles by using radiowave pulses to follow

an acoustic pulse as it propagates through the atmosphere. In this

way it is possible to measure the speed of sound and hence the

temperature (Frankel and Peterson, 1976; Frankel et al. , 1977).

4-14
RASS suffers from the major disadvantage that it is vulnerable to high winds, which advect the acoustic pulse out of the region scanned

by the radar. The result is that even with clever engineering a rather formidable array of antennas is required to overcome this

difficulty (Priestley, private communication).

4-15
FIGURE 1: EDDY WIND FIELD (MEAN WIND FIELD REMOVED) AT 0.3 KM ABOVE

GROUND LEVEL ON 28 JULY 1975. WINDS ARE OBTAINED FROM

AN AVERAGE OF 17 VOLUME SCANS TAKEN DURING A 70 MINUTE

PERIOD ENDING AT 1557 CDT. THE VERTICAL CUTS OF FIGURE 2 LIE AT THE LOCATIONS Y = 2,4,6,8 KM. THE GATEWAY

ARCH IS INDICATED BY THE DOT JUST TO THE WEST OF THE MISSISSIPPI RIVER AT THE LOCATION X = 22.6 KM, Y = 8.5 KM KROPFL AND KOHN , 1978).

FIGURE 2: THE VERTICAL CIRCULATION PATTERNS ASSOCIATED WITH THE CIRCULATION OF FIGURE 1 (KROPFI AND KOHN, 1978).

FIGURE 3: WIND FIELD OBSERVED BY DUAL-DOPPLER RADAR IN A THUNDERSTORM, COMPARED WITH A MODEL OF THUNDERSTORM CIRCULATION

(VERTICAL PLANE CUT).

FIGURE 4: TORNADIC VORTEX SIGNATURE (STIPPLED AREA) WITHIN A

SINGLE-DOPPLER VELOCITY (METERS PER SECOND) FIELD. VELOCITIES ARE RELATIVE TO TVS MOTION. SHORT DASH MARK

INDICATES NO DATA, AND DOT INDICATES TORNADO LOCATION.

DARK RECTANGLES SHOW RELATIVE RANGE GATE SIZE. RADAR

AZIMUTHS AND RANGES ARE INCLUDED (FROM LEMON ET AL.

,

1977; BROWN ET AL. , 1978).

FIGURE 5: A SCHEMATIC REPRESENTATION OF AIRBORNE LIDAR STUDIES OF

WATERSPOUTS, INDICATING THE APPROACH USED TO OBTAIN THE

DATA OF FIGURE 7.

FIGURE 6: WATERSPOUT VELOCITY PROFILE DEDUCED FROM LIDAR MEASUREMENTS.

FIGURE 7: SCHEMATIC REPRESENTATION OF A SYSTEM FOR DETECTING

THUNDERSTORM GUST FRONTS.

FIGURE 8: VERTICAL SLICES OF A COLSTRIP, MONTANA POWER-PLANT PLUME

AS MONITORED BY PULSED LIDAR.

FIGURE 9: SCHEMATIC OF THE WPL PULSED-LIDAR SYSTEM FOR AEROSOL

STUDIES.

FIGURE Al : TEMPERATURE AND HUMIDITY PROFILES RECOVERED BY RADIOMETRY WITH AND WITHOUT KNOWLEDGE OF INVERSION HEIGHT,

COMPARED WITH RADIOSONDE DATA.

NOTE: FIGURES NOT AVAILABLE FOR THIS PAPER

4-16
EXTREME WIND DATA BASE AT THE NATIONAL

CLIMATIC CENTER

Michael Changery

National Climatic Center, NOAA

Department of Commerce

Extreme Wind Data Base Development

at the National Climatic Center

ssign wind speeds included in currently accepted building codes and

andards are presented in probabilistic terms. A daily extreme

Lnd speed is extracted for a location for as many years as are

/ailable. This set of observations comprises a population and a requency distribution with the lower bound at zero. From this set

if daily values, the series of annual extremes form an extreme value

Lstribution. Statisticians have investigated such distribution

id have found that the limiting distribution can take only three

:>rms regardless of the distribution of the original data. These

arms are called Type I, II, and III. Type I is unbound at both

nds and Types II and III are bounded by zero, below and above,

2spectively. Since wind has no strict upper bound, the probability

f a known extreme being exceeded exists, although such probability

ay be very small. A specified mean recurrence interval corresponds

d an expected design wind speed for this location and the associted stresses may not exceed the stress level considered in strucjral design. Currently, the American National Standard A58.1

Specifies that the 50-year mean recurrence interval (frequently

ailed the "return period") value is the design wind speed for all

tructures excepting those with an unusually high or negligible

azard to life. This standard also assumes that the Type II ?rechet) distribution best models the behavior of wind extremes.

The national maps (in Standard A58.1) of wind speeds for

arious recurrence intervals were obtained by utilizing the Type II Lstribution and the annual fastest mile speed for approximately

50 locations for an average of 15 to 20 years. In areas subject

3 the extreme winds of tropical storms, a mixed distribution was

sed since tropical storm winds have a different distribution than

Dn-tropical ones.

For a number of reasons, the current design wind standards

ave been a source of considerable disagreement among researchers

nvestigating design wind and structural wind loading. The first

3 concern that the fastest mile may not be the most useful extreme

d be used in modeling of extreme winds. No readily available

Iternative set of data exists. Many investigators also believe

tiat other additional statistical techniques can be employed to

pre accurately define the extreme wind environment. No documented

2t of annual fastest mile speeds used in deriving the standard

alues is available. In addition, the period of record used in the

tandard covers generally 15 to 20 years — a period many believe

do short to adequately model the extremes at a location.

The National Climatic Center (NCC) , repository for the majority

f meteorological data gathered in the United States, is the largest

rganization of its kind in the world. In addition to its standard

chival function, the NCC is responsible for extracting and prossing c limatological data in a variety of forms for users such as

e general public, industry, academia, and other government

rganizations.

In response to requests from other agencies, the Center is

eveloping additional sets of wind extremes for a greater number of

tations and years than currently available. We are also developtig documentation on current and historical wind data for the locaions in the Center's archives — including availability of multiple

agister and gust recorder charts (the source of wind extremes),

nd instrument height, location, dates of height changes and pertient exposure problems.

5-1
This paper will present the status of extreme wind data bases

the Center is developing for the Department of Energy (DoE),

National Bureau of Standards (NBS) and Nuclear Regulatory Commission (NRC). Initially however, a brief discussion of the National

Weather Service (NWS) instrumentation and data types is necessary.

NWS Instrumentation

Although various research instruments are available for sampling

the wind speed fluctuations at intervals of seconds or fractions

of seconds, the standard anemometer of the Weather Service (and its

predecessors — the Weather Bureau and Army Signal Corps) has been

the vertical axis rotating cup anemometer. Many modifications were

applied to this instrument in an attempt to more accurately record

wind speed. The number, shape, and strength of the cups and length

of the supporting arms have all been changed. Basically however,

these instruments measure wind passage determined by motion applied

to a freely rotating mechanism.

Beginning in 1870, the 4-cup (hemispherical) Robinson anemometer was introduced by the Army Signal Corps. This was a contacttype instrument that accumulated wind passage in miles and fractions

of miles on a dial. Frequently, this instrument was used with a mechanism which allowed a current to flow at each contact which

actuated a light or buzzer. The observer reported wind speed by

counting the number of contacts per unit time. A more sophisticated

mechanism actuated a pen arm at each contact which recorded each

mile of wind passage on a clock-driven drum chart called a single

or multiple register.

The 4-cup instrument was used exclusively until 1927 when

replaced by a 3-cup anemometer with significantly better accuracy.

The 4-cup was known to highly overestimate wind speeds — particularly the extreme winds. For example, a 100 mph observed speed

required a minus 24 mph correction to indicate the true speed.

Only the observed speed is entered in the official records and all

data before 1928 must be corrected by a table of 4-cup corrections

before use. Many investigations early in the century demonstrated

the superiority of the 3-cup (hemispherical) anenometer (designated

"S" type) and beginning January 1, 1928, the Weather Bureau began

using this instrument for official wind measurements. The 4-cup

was retained as back-up. Observations made with the 3-cup were

also entered uncorrected in the official records on the erroneous assumption that the anemometer ran so close to the true speed that

corrections were not necessary. It was later determined that the

originally published corrections were in error and hence the published values must also be corrected to true speed before use. Readings from any 4-cup back-up instrument used were corrected to a true speed for entry in the official records. For a few years

beginning in 1932 the 4-cup instrument was again the official

station instrument, but at most locations the 3-cup was eventually

re-adopted for official observations. Most importantly all wind

data entered in official records since January 1, 1932 are true speeds with the necessary corrections applied.

Additional investigations at this time demonstrated that cups

with "beaded" edges — produced by rolling the edges outward to form a ring — ran considerably smoother in a turbulent air-stream

than cups with straight edge. Beaded edges were used on 3-cup

(semi-conical) small airways contact anemometers (designated "SA")

introducted at smaller airports in the 1930' s. While beaded cups were found to be advantageous, they were not used on the standard

3-cup ("S") anemometer at major airport stations. However, at

5-2
selected locations, a special beaded 4-cup experimental instrument

was used in the 1930's and 40' s. All instruments mentioned were contact anemometers, although aviation requirements in the 1930's

led to the development of contacts for each l/60th of a mile of

wind passage. Thus the number of contacts per minute was converted

directly to a mph equivalent.

A second type of instrument employing a different method of

recording was introduced during the 1940's at most major airports.

This was the direct-reading type (similar to those in operation

currently) in which the wind-induced rotation governs the magnitude

of electrical current generated by a magneto. The current is

linearly related to the true wind speed, which is determined by

dial displacement. The direct-reading instrument is frequently

connected to a device, called a gust recorder, which provides a continuous record of wind speed. Direct-reading instruments have

generally replaced contact-type instruments at all locations except

approximately 130 major airport locations which also employ the

older-type instrument in recording data on multiple registers.

Types of Wind Measurements

As mentioned in the preceding section, contact-type anemometers

accumulated wind passage in miles and fractions in the earlier

years, and in l/60th of miles in the 1930's and 1940' s. It is

clear that measurement of short-term gusts of wind speed had to await development of the direct-reading anemometer. Early stations

measured the wind at 2 or 4 observation times per day. This was done by converting the miles and fractions of wind passage in

5-minutes to a mph equivalent. After the turn of the century, stations began accumulating hourly totals of wind passage on a 24-hour

basis. Aviation requirements by the 1920 's demand hourly or more

frequent measurements determined again by 5-minute measurements

made each hour.

At stations equipped with register charts marked by successive

passages of miles of wind on a clock-driven drum chart, two measures

of wind extremes were made:

(1) Beginning in 1872, any station so equipped determined on a daily basis the greatest number of miles of wind to pass by

the instrument in any 5-minute period. This was called the

5-minute maximum speed and was tabulated on a daily basis

until ceased by the Weather Bureau of 1957.

'(2) Beginning in 1887 and continuing through 1904 and then beginning again in 1912 and continuing through the present, a second type of daily extreme was extracted. Called the

"extreme" or fastest mile, this measurement was obtained by

determining the shortest time interval between successive

marks and converting this to miles per hour. Thus, marks

1-minute apart represents a 60 mph speed; 2-minutes apart —

a 30 mph speed; and 30-seconds apart — a 120 mph speed. A special graduated scale permitted direct conversion of a "mark spacing" to a mph equivalent.

With introduction of the direct-reading instruments in the

1940' s as a standard equipment, the hourly observed winds were obtained by an "eyeball" estimate of the wind for a 1-minute

period. Assuming the speed met certain criteria, wind gusts

(highest instantaneous speed) observed were included on hourly

reports. However, only stations with continuous recording instruments carried a daily peak gust since this gust could occur between

5-3
hourly observations and not be seen by the observer. Military sites

(Air Force and Navy) were generally equipped by such recorders from

the mid-1940's. Natinal Weather Service locations, as a rule, did not have recorders until the late 1960's.

The choice of suitable measurements to be used in extreme value

wind analysis is limited to three: 5-minute maximum, fastest mile,

and peak gust. As mentioned above a national data base of annual

peak gusts must be restricted to the last 10 years excepting military sites - too short to be reliable statistically. It has been

assumed that a relatively short burst of extreme winds produces the

most dangerous wind load on a structure. An extreme as determined

by fastest mile measurement which may be limited to a time frame of

1-minute or less, may be a better parameter to use. In addition,

the other available parameter (5-minute maximum) has not been

extracted since 1957. For airport locations, at which the contact

anemometers have been located since the 1940' s, the period of record

would be limited to those years prior to 1957 and after the transfer

of instrumentation to the airport - again too short a period to be

statistically useful. The remainder of this paper discusses the

types of extreme wind data base development in which the Center is

currently involved.

The NBS Project

The National Bureau of Standards has long been active in investigations of wind load on structures. At their request the NCC is providing additional extreme wind data on a national basis for use

in developmental work. The primary purpose of the Center's role

in this project is two-fold: (1) To update the set of fastest mile

speeds by including the last 10-12 years of data available. As mentioned earlier, the data on which the current Standard is based

was restricted to the 15-20 year period prior to the mid 1960's;

and (2) To provide a set of 5-minute maximum speeds coincident with

the fastest mile.

The original set of fastest-mile data was restricted to airport locations. Neither the period used nor the actual values have

been available for use by other reseachers. Therefore, the Center

is determining the first full year of fastest mile data at a given

airport location and listing all annual values through 1977 or the

last year available. It is apparent, unfortunately, that the

contact-type anemometer is slowly being phased out as a source of

wind data. An additional station is lost to the data base every

few years. For each occurrence of the fastest mile, we are also

extracting the coincident 5-minute maximum speed. This may or may

not be the true annual highest 5-minute maximum. In checking a number of stations for another project, we determined that the

annual extreme fastest mile and 5-minute maximum were normally

coincident. However, occasionally a fastest mile extreme associated

with a short period thunderstorm outflow may have a 5-minute maximum

lower than a second episode of sustained strong winds in the same year. We are not providing the true annual value since this would

require close analysis of daily charts for approximately 130 locations for the period 1958-1977 — a very expensive and time consuming

proposition. This analysis would be required because 5-minute

values were not extraced from charts after 1957.

Two problems have been encountered in this NBS project. We

have been consolidating the fastest mile values entered on daily

manuscript forms and listed in various publications. Obtaining the

coincident 5-minute values after 1957 requires inspection of the

register charts for the day of the fastest mile. We have come

5-4
cross a number of occurrences where the published fastest mile

falue is seriously in error - often by a factor of two. The probem is apparently caused by misusing the scale provided for the

xtraction of fastest mile data. To conserve space, the scale has

; een constructed so that four speeds (multiples of one another)

an be read in essentially the same location on the scale. It is

ery possible for an unskilled observer to misread the scale's

rinted speed and report a speed twice the actual. For example, a

'0 mph speed requires a mark interval of two minutes and misreading

he scale can result in a 60 mph reading. Charts for a given day are

.nalyzed on station the following day, and the observer could check

he extracted value against actual surface observations made at the

ame time. Apparently, this is not always done. The erroneous

eading is then entered on manuscript and climatological forms used

|t the NCC as the sources for published data. No quality control

rocedures exist for the registered charts and data is not recogized as erroneous until a special data requirement, such as this

roject, is funded. In cases such as these, we are not providing

[he published value, but are checking the next highest speed and

roviding this value if it appears reasonable.

The second problem encountered is not as serious. It has been

tation practice to log the scale value, together with time of ccurrence on the chart for a given day and then apply the necessary

nemometer correction before entering the observation on official

orms used for published data. Beginning in the late 1960's, we

ave found many instances where the speed logged on the chart is

he same as entered on the official forms. Either the observer is

orrecting the scale value before logging it on the chart or he is

ntering uncorrected speeds on the official forms. The differences

.re minor since most corrections for the 3-cup extracted values are

ess than 5 mph. It is not possible to re-analyze the charts for

he purpose of obtaining a definitive answer because of the

ifficulty in distinguishing (as an example) a 44 mph spacing from

47 mph spacing. We prefer to use, wherever possible, the

n-station observer's estimate. Unfortunately, at stations where

his has occurred, the problem is not consistent. A few intermedite years may have data entered in the "correct" manner. We have

hecked with a number of stations and all Weather Service Regional

,'ffices — all of whom stated that uncorrected values should be

ogged on the register chart, although there are no official guideines which so dictate. Proceeding under this assumption, we have

.pplied corrections to the appropriate published uncorrected speeds

,nd used these unless the corrected value was less than another

: peed (already corrected) which occurred during the year. We have

rurnished the available fastest mile and coincident 5-minute speeds

to NBS together with the anemometer heights and dates of changes,

ncluded are also a number of stations considered as "airports"

iue to their location, such as Cape Hatteras, NC, Block Island, RI,

orth Head and Tatoosh Islands, WA. Data for these begin early in

he century. The entire listing will be published for distribution.

One of the problems in utilizing contact anemometer data is

hat the station location changed from an urban city office to a jural airport office 25-30 years ago. The marked change in surface

oughness has precluded merging the data for both locations into a ingle set. For a few stations, a considerable period of overlap

xisted with registers existing at both city office and airport

ocations. For these stations, we are providing not only the

available annual extremes at both locations, but also the daily

xtreme at one location to match the date of the annual extreme at he second if the annuals at both sites did not occur on the same

ates. This will enable NBS to investigate possible techniques

5-5
for providing a homogeneous set of data for the station's entire

period of record.

The last product to be provided will be limited to the period

of record for 5 airport stations. Together with the annual fastest

mile and associated 5-minute maximum, we will provide a true annual

5-minute maximum. We will also provide the associated 10-minute

and true annual 10-minute maximum. To provide the true annual 5-

and 10-minute maximum requires detailed analysis of daily charts

after 1957 for the 5-minute speed and the period of record for the

10-minute speed. This portion of the project will be restricted to only a few stations to test the validity of providing associated

values in lieu of the true annual, and to provide an additional

set of measured extremes.

The NRC Project

The Nuclear Regulatory Commission requires detailed wind loading

analyses for proposed nuclear power sites — particularly those along

and off the immediate coastline. To provide the largest possible

extreme wind data base, the Center is developing a data base for

all locations within 10 miles of the Atlantic and Gulf of Mexico

coastlines consisting of all fastest mile values for the station's

complete period of record. This will be for both city office and

airport locations.

In addition to the first-order stations currently in use, we found many additional locations with extracted data — including

city office locations which made observations until 20-30 years

ago. The project requires the quality control of available fastest

mile data and extraction of data prior to 1887 and between 1905 and

1911 - two periods during which the fastest mile was not extracted

on station. We are also extracting data by examining daily traces

for those stations with register charts which have never been

analyzed. As an example, nearly 70 years of data exist for Delaware

Breakwater, DE, which have never been extracted. In addition, 15 to 20 years of data exist for a larger number of Coast Guard sites

along the Florida coastline, including the Florida Keys. However,

the quality of the charts are, in many cases, less than adequate,

with frequent equipment malfunctions and long periods of missing

data.

The requirement to analyze charts for annual extremes never extracted has reinforced the conclusion that an accurate estimate

of extreme winds becomes very tenuous for the higher wind speeds —

particularly those in association with tropical storms. For example, to determine an 84 mph fastest mile speed requires measuring the spacing between two consecutive marks out of the 6 or 7 which may be marked in a 5-minute space occupying 1/4-inch of

chart paper. The overrun problem of the old 4-cup anemometer

accentuated the problem at high speeds by registering additional

marks in a short spacing — although the overrun was corrected out. We have found that in most instances, the older register charts are of higher quality than current charts. The oldest charts were originally marked in pencil and these have become very faint with

time. However, generally from 1900 through the 1940's, the charts were finely marked by ink and adjacent marks are distinguishable.

This instrument quality control presumably was a reflection of the

stations' dependence on the chart as the only source of wind data.

Currently, the direct-reading instrument is the primary source of

wind data, and the quality of register data has suffered. As a result, during severe wind episodes adjacent marks tend to run together and become nearly illegible. We have found a few

5-6
historical cases where, because of anticipated very high winds, the

station had doubled the speed of the drum, making analysis of the

actual extreme significantly easier.

In addition to providing all historical annual fastest mile

extremes, the Center is also extracting the coincident 1-, 2- (if

available), 5-, 10-, 15-, and 30-minute and 1-, 2-, and 3-hour

speeds together with associated directions. This will provide

additional sets of wind extremes as well as data from which interrelationships for various time intervals could be developed. The

NRC project is scheduled for completion in late 1978. The final

data set will be consolidated and available to all requesters.

The DoE Project

The final project in which the Center is engaged is not directly related to development of actual data sets, but could be

very useful to the research community. The Department of Energy

requested that the Center publish a concise listing of all stations

in its archives with any type of wind data available. Anemometer

information was to be included. The listing was designed for those

interested in investigating wind power as an alternative source of

energy. The publication (to be available within a few weeks) lists

all locations with basic wind observations, their location, elevation, approximate number of observations per day, and period of

record. Also included were the period with digitized data, multiple

register charts, and gust recorders charts. Most important,

instrumentation information is provided, such as anenometer height,

dates of height changes, exposure location, and exposure problems.

The instrument history is of importance in any effort to standardize wind extreme observations to a given height. We have found numerous examples of erroneous anemometer information in published

sources. In particular, most sources do not discriminate between

contact and direct-reading anemometer and assume similar heights

for both if only the direct-reading height is known. This is

frequently not the case.

Future Projects

We anticipate two additional developmental projects involving

extreme wind data. The Center may analyze additional station-years

of data for NBS to determine the accuracy of published values. If numerous examples of inaccurate data exist, we may need to check

and correct all the airport extremes.

A second project for the NRC may involve the extraction of

annual fastest-mile speeds for the entire historical period of

record for all locations east of the Rocky Mountains. This would

cover all areas of th U.S. subject to extreme winds excepting the

Northwest coastline. If funded, this project would take at least

two years to complete.

5-7

6

A STRONG-MOTION RECORD INFORMATION RETRIEVAL

SYSTEM

A. M. Converse and R. B. Matthiesen

Seismic Engineering Branch

U.S. Geological Survey
i

r
A Strong-Motion Record Information Retrieval System

To make information about strong-motion records readily

available, the information has been entered into a computerized

data-base management system. The system provides ready access to information about strong-motion records, stations, events, and

level of data processing and analysis of each record. The system

should prove to be of considerable value in the aftermath of a major

event as a means of providing immediate information to persons in

research, design, operations, and regulation.

The basic organization of the data is illustrated in Figure 1. Of primary importance is the information about the records themselves and the level of processing and analysis that has been

performed. Information about the event generating the motion and the

stations at which it was recorded are also of major importance.

These three elements constitute the major data sets included in the

data base. Supplementary data sets include general information

about the data base and how to access data, additional information

about the stations, and the recorders and transducers. The organizations that own the instruments have additional information about the

stations and sites, and archive the original records. The specific

information that is included in each data set is indicated in

Table 1.

At present, the information included in the data is restricted

to that for which the USGS has a primary responsibility. The data

sets will be expanded to include information about all strong-motion

programs in the U.S. In addition, the computer program will be

available to other organizations that operate strong-motion programs

in other parts of the world so that information about strong-motion

data can be exchanged efficiently.

6-1
EVENTS RECORDS STATIONS

GENERAL

INFORMATION

DATA

ANALYSIS

SUBSTATIONS

SUPPLEMENTAL

INFORMATION

FIGURE 1: STRONG-MOTION DATA ORGANIZATION

TABLE 1: DATA SETS

EVENTS event Identification

(date & suffix)

time (local or GKI) event name location: latitude

longl tude magni tude maximum intensity references

RECORDS

event identification station number substation identification transducer level or location

recorder identi f ication(s)

s-trigger interval, sec. epicentral distance, km. intensity at the site

total length of record peak acceleration duration >0.1g who has the original record references

DATA ANALYSES who digitized the data

who has the digitized data length of record digitized

highest stage of analysis

stage 2 frequency band data processing type references

STATIONS

station number station name

(or street address)

station location

(or city)

state (or country)

nearby stations references

SUBSTATIONS substation Identification

date of installation

date of removal location: latitude longitude

geology: class code

near surface shear wave vel. structure: class code

size short description

remarks

ARRAYS array name

list of stations and their substations In the array

remarks

NEARBY STATIONS

station number

list of nearby stations remarks

RECORDERS

recorder Identification

owner agency

date of Instal latlon

date of removal remarks

TRANSDUCERS

transducer serial number date of installation

date of removal location: code

short description direction of acceleration

orientation

trace location on record

remarks

TRANSDUCER CALIBRATIONS

calibration date sensitivity

period

damping

remarks

RECORDER TYPES

recorder type code descriptions

AGENCI ES owner-agency code

remarks about the owner-agency

name, address, and contact

6-2
7

INTEGRATION OF STRONG-MOTION ACCELEROGRAMS

Susurau Iai

Eiichi Kurata

Hajime Tsuchida

and

Satoshi Hayashi

Port and Harbour Research Institute

Ministry of Transport

Integration of Strong-Motion Accelerograms

Ground displacements have a great influence on structures such as submerged tunnels and pipelines, and also on large-scale structures.

Because of the need for ground-displacement records of strong earthquakes, several methods have been proposed for integrating strongmotion accelerograms. Although the methods proposed so far are basically quite similar, the displacements obtained by them are often very different.

An accelerogram always produces some error. Moreover, integration overemphasizes the low-frequency component. Therefore, direct

integration of a digitized accelerogram generally produces an abnormal displacement because of low-frequency errors. Therefore,

one of the important parts in an integration is the design of an appropriate high-pass filter. The report deals with this problem

based on an examination of records accuracy and proposes two new

methods for integration of accelerograms.

There are many sources of error in a integrated displacement

if the error is defined as difference between a true displacement

and an integrated displacement. Some of the errors generated

during recording and processing before digitization come from the

seismometer itself; some of these errors can be corrected. Although

i such errors should be examined carefully, the report deals only

with the errors generated during the digitization.

Accelerograms examined in the report are those recorded by the

ERS accelerographs and those recorded by the SMAC-Bj accelerographs

.

The accelerograms were digitized for each kind of record. However,

the discussion in this report may be applied to many types of

accelerograms and digitizers.

Accelerograms and Digitizers

The ERS accelerograph is composed of three transducers of the

moving-coil type and an electromagnetic oscillograph. The frequency

characteristics are such that the magnification is maintained at a 90% level in the frequency range 0.18 - 50 Hz. The sensitivity of

each accelerograph is adjusted depending on the prevailing seismic

activity at the site and other conditions. The sensitivity of the

records dealt with later in the report is 10 gal/mm. The recording

speed is 4 cm/sec.

'The digitizer for making a record from an ERS accelerograph is

such that with a cursor an operator traces waveforms on an oscillogram fixed on the table. The sampling intervals along X-axis are

0.1 mm, and the digitization unit has the same values. Length of

the record to be digitized without shifting of the oscillogram is

70 cm at the maximum.

The SMAC-B2 accelerograph is a mechanical one; the natural

frequency of the pendulum is 7.14 Hz and its damping is critical.

The sensitivity is 12,5 gal/mm and the recording speed is 1 cm/sec.

Its digitizer is similar to that used with the ERS accelerograph.

The difference is that the digitizer for the SMAC-Bj accelerograph

has a cursor that moves automatically in an incremental manner in

0.1 mm intervals along the X-axis. Digitization units in the Y-axis

are 0.008 mm. The maximum length for the digitization is 45 cm. A detailed description of these accelerographs and digitizers is

available in a separate report.

7

1 7-1
Accuracy of Digitized Accelerograms

To investigate the accuracy of digitized accelerograms, Fourier

spectra of an accelerogram and its errors are studied. The accuracy

is specified in terms of ratios between them at individual frequencies. Hereafter, Fourier spectra of an acceleration record will

be called SIGNAL (S); Fourier spectra of errors will be called NOISE

(N); and the ratio will be called S/N.

There are two kinds of NOISE generated during digitization; one

is generated in the process of tracing an acceleration record due

to the difference between a recorded line and an actually traced or digitized line. The causes of this difference are parallax, fluctuation of a hand in motion, and errors in the digitizer mechanism and

electronics. The other is due to the difference between true zero

axis in the record and the base-line located before or after the

of tracing. The former will be called digitization errors and the

latter will be called base-line location errors.

Digitization Errors

For the examination of digitization errors, accelerograms by the

ERS accelerographs were digitized several times by the digitizer

described previously. A part of the results of the digitization

are shown in Figures 1 and 2. The digitized lengths and the

operators are listed in Table 1.

An average of the digitized records is calculated and its

Fourier spectra is approximately looked upon as SIGNAL. The difference between the averaged record and each digitized record is

calculated and its Fourier spectra is calculated. Then, an average of absolute value of the Fourier spectra is calculated and

approximately looked upon as NOISE. Figures 3 and 4 show SIGNAL

and NOISE (the figures at the bottom) and their ratio S/N (the

figure at the top) in the frequency range up to 1 Hz, where the

accuracy is often questionable for integration of acceleration

records. S/N in Figure 3 is mostly lower than 1 through the

frequency range shown. Increase of SIGNAL with decrease of frequency may be due to the transverse play of the recording paper

because in this examination no correction on the transverse play

has been applied. By the way, predominant frequencies of SIGNALS

are about 10 Hz in both records and S/Ns around this frequency are about 10 in both cases, as shown in Figures 5 and 6.

These results show average characteristics of NOISE and S/N.

That is:

° S/N and SIGNAL are similar in the frequency domain. The

greater SIGNAL is, the greater S/N is.

° In the frequency range up to 1 Hz, NOISE is almost

constant

.

These facts imply that digitization errors" are approximately

white noise. Comparison between Figures 3 and 4 shows the following: NOISE level varies slightly with a digitized accelerogram.

However, if we take into consideration the digitizer and the digitization process, it can be expected that this variation will be

limited to some specified amount.

These results lead to the following conclusion about digitization errors. Digitization errors are white noise whose level varies

with a record to be digitized. The variation is, however, limited.

7-2
The average lev|l of NOISE in terms of power spectral density is

about 0.05 (gal /Hz) in the case of the accelerogram obtained by an ERS-accelerograph whose sensitivity is 10 gal/mm.

With an accelerogram from the SMAC-B^ accelerograph , a similar

examination was carried out. The conclusion was similar to that

described above. NOISE level was found to be 0.05 (gal /Hz) too.

Base-line Location Errors

If significant part of an accelerogram is longer than the available

digitizer table length, it is divided into sections for the digitization. In this case, a base-line is located for each section.

Therefore, base-line location errors are piece-wise constant or stepwise errors. Fourier spectra of the errors of this kind are approximately proportional to 1/f, where f is a frequency.

I The amount of the step error is approximately equal to the

standard deviation of digitization error. In the cases of accelero

grams examined previously, the standard deviations were 2.7 gal and

0.8 gal, respectively. This result implies that the amount of the

step error is 0-3 gal on an accelerogram from the ERS accelerograph

whose sensitivity is 10 gal/mm. In the case of an accelerogram

from the SMAC-B2 accelerograph, the same amount was estimated.

Errors in a Digitized Accelerogram

Here, an equation will be constructed to represent the approximate

errors — digitization errors and baseline location errors — for

the estimation of S/N in a later section.

Digitization error is approximated by white noise whose level

is 0.05 (gal 2 /Hz), as described previously. An accelerogram to be

integrated is corrected with a digitized fixed trace before the

integration to eliminate errors due to the transverse play of the

I recording paper. Since the corrected accelerogram contains

digitization errors in the fixed trace, the digitization errors in

the corrected accelerogram are approximated by white noise with the

level of 0.1 (gal 2 /Hz).

On the other hand, power spectra of base-line location errors

for each digitization section are expressed as - [a* sin( 2irfT) / ir f ] ;

where, T (sec) is length of the section and a (gal) is the amount

of the step errors. Power spectra of the baseline location errors over total record length may by approximated simply by

2T ir f

2

The following equation expresses these errors approximately:

E (f) = [0.1 +

2 ± (f)

2

• ijx lO" 2 P

2 (1)

where, a = 0 ^ 3 (gal), and

p (gal/mm) is the sensitivity of the ERS accelerograph.

. The sensitivity of the SMAC-B2 accelerograph is 12.5 (gal/min)

however, the digitization error of an accelerogram from the SMAC-B2

accelerograph is almost equal to that from the ERS accelerograph

7-3
with sensitivity of 10 gal/min. In the case of a record by the

SMAC-B2 accelerograph, the error may be obtained by substituting

p = 10 into the above equation.

A Relation Among S/N, Frequency Band, and Records

While NOISE, as described previously, becomes higher at low frequencies, SIGNAL (ideally true Fourier spectra of an accelerogram)

becomes, for most cases, lower at low frequencies. Therefore, S/N

of a digitized record mostly becomes lower at low frequencies. On

the other hand, the level of SIGNAL varies with records. Therefore,

a degree of the decline of S/N at low frequencies varies with

records. In the usual situation where the number of records available is limited, these facts imply the following relation among S/N,

frequency band, and records (Figure 7).

° If a frequency band is specified and fixed, S/N maintained

in the frequency band varies with records. For each record,

the wider the frequency bandwidth is, the lower the S/N

ratio is. For a fixed frequency band, the greater the

number of records is, the lower is the overall S/N level.

0 If a S/N is specified, a frequency band maintaining the

S/N varies with records. For each band, the higher the S/N

is and the greater the number of records is, the narrower

the frequency bandwidth is,

° If an S/N and a frequency band are set, corresponding

records are selected. The higher the S/N is and the wider

the frequency band is, the smaller the number of corresponding records is.

These three relations are summarized as follows:

0 If levels are specified for two items among S/N, frequency

band, and records, the level for the rest will be automatically determined. The higher the level of one item is, the

smaller the level of the other is, if the level of the rest

is fixed. Here, the heights of levels of items denote S/N

level, frequency bandwitdth, and number of records respectively.

Design of Filters

Appropriate S/N Frequency Band, and Records

For the Annual Report on Strong-Motion Earthquake Records in

Japanese Ports published by the Port and Harbour Research Institute,

S/N, frequency band, and records are set as follows:

(1) Records whose maximum horizontal component is greater than

50 gal are integrated. For the vertical component, no limitation is applied.

(2) Two kinds of filters are applied; one maintains a fixed frequency range and the other is expected to maintain constant

S/N.

Condition (1) assures some frequency bandwidth and some S/N level and at the same time assures the number of records.

Condition (2) meets the desire to use the integrated displacements

for a wide range of research.

7-4
It should be considered in fixing frequency band that it may

usually be required to retain at least 0.1 Hz - 0.3 Hz. However,

if width of frequency band is extended excessively, S/N implicity

required (say 1.0) cannot be maintained for most records. In con- nection with fixing S/N, it will be pointed out that there is no clear consensus of required or recommended S/N level among data

users. However, if S/N level is fixed to be 10, for example,

corresponding frequency band is limited just around the predominant

frequency for most of the records, and does not meet the implicitly

required (say higher than 0.5 Hz) frequency range. These considerations lead to the conclusion that, in the case of fixing frequency

band, appropriate (or expendient) lower boundary is 0.1 - 0.3 Hz,

and in the case of fixing S/N, the appropriate S/N level is 2 - 5. Because the upper boundary of the frequency band does not have a great influence on integrated displacements, the report does not treat it. For this matter, a separate report is available.

Sharpness of Roll-Off

To design a high-pass filter, sharpness of roll-off of the filter

should be discussed as well as the cut-off frequency.

Three out of the many filters proposed so far are shown in

Figures 8(a) through 8(c) as examples of sharpness of roll-off.

In the figures, combinations of filters and double integrals

' (l/(2irf ) 2

) are shown for comparing the influence of the sharpness

on integrated displacements. The filter in Figure 8(a) has a stronger effect on artificial predominant components around the

cut-off frequency. On the other hand, the filter in Figure 8(c) is too mild to filter some limited frequency band. The sharpness of

roll-off of the filter in Figure 8(b) was considered appropriate.

Proposed Filters

A Fixed Filter (which has a fixed cut-off frequency) and

Variable Filter (which has a cut-off frequency varying with records)

are both proposed in this section. These filters are both being

applied to the routine standard processing of the strong-motion

accelerograms in the observation network of the Port and Harbour

Research Institute.

Fixed Filter

This filter is designed to make it easy to compare the integrated

displacements with records observed by the seismographs deployed by

the Japan Meteorological Agency (the natural period of horizontal

pendulum is 6 sec. and the damping factor is 0.552); see Figure 9.

1 1 Hi (f) = • ____ (2)

i - (ii) 2 - 2h (ji) i -Z1 + ( t~

)2

where, fQ = 1/6 (Hz), h = 0.552, and fj - 0.1 (Hz).

Cut-off frequency (3 dB down) of this filter is 0.154 (Hz).

Errors for integrated displacement by this filter, estimated

by Equation 1, are O.Olp (cm) to 0.06p (cm) in terns of standard

deviation, where p (gal/mm) is the sensitivity of the ERS accelerograph. In the case of an accelerogram from the SMAC-B2 accelerograph, estimated errors are 0.1 (cm) to 0.6 (cm).

7-5
Variable Filter

Variable Filter is defined, with the parameter fc , by the following

equation; see Figure 10:

2 2 H2 (f) = [1 - exp { - (|_) }] (3)

c The cut-off frequency (3 dB down) is 1.36fc (Hz).

The parameter fc varies so as to make a equal to standard quantity E related to the required S/N, where a is a quantity to express

the low-frequency component level and defined by

| x (f) | 2

• [1 - exp {- (fT) 2

}] 4 • [1 - H

2 (f)] 2 df (4)

where, S is length of digitized accelerogram,

T is a minimum length of a section for digitization, and

x(f) is Fourier transform of an acceleration record.

fc is fundamentally determined so as to eliminate some constant amount of the low-frequency components of an accelerogram

(Figure 11). Components lower than about 1/T are eliminated for

the decision procedure of fc, because, in this frequency range,

fluctuation of baseline location errors are considered to be

remarkable. The decision procedure is introduced so as to assure

some stability against possible fluctuation of NOISE.

The selected values for E are listed below:

E = 0.05 (gal) for a record by the ERS accelerograph whose

sensitivity is p(gal/mm),

E = 0.5 (gal) for a record from the SMAC-B2 accelerograph.

This value E is expected to retain S/N level 2 - 5 in the

frequency range higher than the cut-off frequency under the following condition: that is, NOISE is expressed by Equation (1), fc,

is about 0.1 - 0.5 Hz, and power spectrum of a digitized record

does not extremely vary in the frequency range up to the cut-off

frequency

.

Specified in terms of standard deviation, expected errors from

Equation 1 in the integrated displacement are approximately:

.1 ,

5 x 10"4

• p • f

c

2

• /l + 0.01 • ( 3T

°) • a2 f c

2 (cm)

where, a = 0 - 3 (gal), T (sec) is length of a section for digitization, and p (gal/mm) is the sensitivity of the ERS accelerograph.

In the case of a record from the SMAC-B2 accelerograph,

expected errors are expressed by the same equation with substitution

of p = 10.

Comparison of the Filters

The filters proposed so far including the two filters proposed previously in this report are shown in Figure 12. The combination of

the filters and double integrals are shown in Figure 13. In the

figures, examples of Variable Filters are shown with parameters

7-6
fc = 0.07 Hz and fc = Hz . Although Kurihara and Sakurai investigated three cut-off frequencies , in the figures, the filter with

the lowest one is shown. Figure 14 shows a sample calculation of

displacements. In this case, the parameter fc of the variable

filter is 0.824 Hz.

Errors in the displacements calculated with filters proposed

by the several researchers were estimated by Equation 1. Here

records from the SMAC-B2 accelerograph and digitized by the

digitizer described previously are considered. The errors are expressed in terms of standard deviation. The results are as follows

Fixed Filter

Variable Filter (fc = 0.824 Hz)

Kurihara's Filter [3]

Kurata's Filter [4]

Tanaka's Filter [5]

Trifunac's Filter [6]

0.1 - 0.6 cm

0.01 cm 0.04 - 1.2 cm

0.1 - 0.3 cm

0.3 - 1.2 cm

0.4 - 2.3 cm

It is seen that the error in the calculated displacement with

the variable filter is remarkably small in comparison with others.

It will be noted, however, that the filter with the smallest error

is not necessarily the best filter because the smaller error is a result of narrower frequency range.

Conclusions

The following are concluded about the integration of strong-motion

acceleration records:

(1) Errors generated during the digitization process exist in a wide frequency range. They generally become higher as the

frequency decreases. The errors depend on each record, but it

is considered to be limited to some amount.

(2) S/N, frequency band, and records have the following relation.

If levels are specified for two items among them, the level

for the other will be automatically determined. The higher

the level of one item is, the lower the level of the other

is, if the level of the rest is fixed. Here, the heights of

levels denote S/N level, frequency bandwidth, and number of

records

.

(3) For the Annual Report on Strong-Motion Earthquake Records

in Japanese Ports published by the Port and Harbour Research

Institute, records whose maximum horizontal component is

greater than 50 gals are integrated. For these records,

two kinds of filters are applied; one maintains a fixed

frequency range and the other is expected to maintain constant S/N.

In the case of fixing frequency band of the fixed filter,

it is considered that the appropriate lower boundary is

0.1 - 0.3 Hz, and in the case of fixing S/N, appropriate S/N

level is 2 - 5.

Sharpness of roll-off of the filter is desired to be neither

extremely sharp nor extremely mild.

16) Two filters have been proposed; one is a Fixed Filter for

fixing frequency band and the other is Variable Filter for

maintaining S/N.

7-7
References

1. Iai, S. , Kurata, E. , and Tsuchida, H. , Digitization and

Correction of Strong-Motion Accelerograms , Technical Note of

the Port and Harbour Research Institute, No. 286, March, 1978,

56 p. (in Japanese).

2. Kurata, E. , Iai, S. , and Tsuchida, H. , Annual Report on StongMotion Earthquake Records in Japanese Ports (1976 and 19/7) , Technical Note of the Port and Harbour Research Institute,

No. 287, March 1978 (in English).

3. Kurihara, C, and Sakurai, A., Integration of Earthquake

Acceleration, Journal of Technical Laboratory Central Research , Institute of Electric Power Industry, No. 68077, June 1969

(in Japanese).

4. Trifunac, M. D. , Low Frequency Digitization Errors and a New

Method for Zero Baseline Correction of Strong-Motion Accelerograms , California Institute of Technology, Report No. EERL

70-07, September 1970.

5. Tanaka, T. , Morishita, T. , Oosawa, A., and Yoshizawa, S., On

the Correction of Records Obtained by Strong-Motion A^ccelerographs , Proc. of Symposium of Architectural Institute of

Japan, October 1972, pp. 411-412 (in Japanese).

6. Tsuchida, H. , and Kurata, E. , Observation of Earthquake

Response of Ground with Horizontal and Vertical Seismometer

Arrays , Proc. the Forth Japan Earthquake Engineering Symposium,

November 1975, pp. 137 - 144 (in English).

7-8
i j v

/ v A

A,1

A,'

j

1 i i i

l I i 1 1 i 1 i 3. 00 0. 50 1 • 00

TIME f SEC )

1. 50

FIGURE 1: A PORTION OF REPEATED DIGITIZATION (M-125 N-S)

as

I. 00

TIME (SEC!

1. 50

FIGURE 2: A PORTION OF REPEATED DIGITIZATION (M-133 N-S)

TABLE 1: SPECIFICATION OF REPEATED DIGITIZATION

Record

Number

M-125 N-S

M-133 N-S

Length of Number of Repeated Digitization Sampling

Digitized (Wrators

Portion Total Operators

(sec) ABC (sec)

Intervals

12.5

10.0

2 2 2

2 0 2

0.0025

0.0025

7

7-9
a

*

I'M hUU /[ HhflH

. C 1 0. Z 0. 3 0. < 0. 5 ".5 0. 7 0. fREOUENC -"2 I

AVE RAGE /ERROR

0. 0 0. 1 0.2 0. 3 0. * 0. S 0. 6 0. 7 0. 9 0. 9 1.0 FREQUENCY IhZI

0.0,3.1 0.2 0.3 0. * 0.5 0.6 0.7 0. 8 0.9 1.0

FREQUENCY I hi )

0. 0 0. 1 0. 2 0. 3 0. * 0. 5 0. 6 0. 7 0. 8 0. 9 10

FREQUENCY I hZ I

FIGURE 3: ACCURACY OF DIGITIZATION

IN LOW FREQUENCIES (M-125 N-S)

FIGURE 4: ACCURACY OF DIGITIZATION

IN LOW FREQUENCIES (M-133 N-S)

average /error

0. 10. 20. 30. 40. 50. 60. 70. 80. 90. 100. FREQUENCY I HZ I

AVERAGE /ERROR

0. 10. ZO. 30. *0. 50. 60. 70. 80. 90. 100.

FREQUENCY IMZI

FIGURE 5: ACCURACY OF DIGITIZATION

(M-125 N-S)

FIGURE 6: ACCURACY OF DIGITIZATION

(M-133 N-S)

7-10
7-11
I HIGH f li TR Hi D0U81F 1NIFC.R3L I Br TRIFuNSC

0. I 0. 2 a 3

FHEOUENCT IH2I

0. < 0. 5

FIGURE 8(a): DOUBLE INTEGRATED FILTER BY TRIFUNAC4

l/(*lf)' | MICH-P05S F ILTFH I >f double intfofw. i FJT KURflTfi

0. I 0. 2 0. 3

FREQUENCE I hZ 1

0. * 0. 5

FIGURE 8(b): DOUBLE INTEGRATED FILTER BY KURATA6

I HIO-PASS F| L TER I I DOUBLE INTEGRR. I BT KuRIMqRP

0- 0 0. I C. 2 0. 3 0. < 0. S FREQUENCE i hZ 1

FIGURE 8(c): DOUBLE INTEGRATED FILTER BY KURIHARA3

7-12
Seismometer

o of J.M.A. ~ (t=6sec,h=0.55)

0. 0

til&ifT f ^IC.h-PRSS FIlTFR 1 ¥( DOUBLE INTEGRAL)

FIlTER NO.

0. 1 0. 2 0. 3

FREQUENCY f hZ

)

FIGURE 9: FIXED FILTER (DOUBLE INTEGRATED)

l/(2TTf)

hlCh-PRSS FIlTER]

¥( DOUBLE 1NTEGRRL ) FIlTER NO. 2

0. 2 0. 3

FREQUENCY ( hZ

)

0. 4

FIGURE 10: VARIABLE FILTER (DOUBLE INTEGRATED)

7-13
FIGURE 11: SIMPLIFIED ILLUSTRATION OF DECISION PROCEDURE OF FC

7-14
TRIFUNAC

TAMAKA

VARIABLE FILTER (PROPOSED) <Fc-0.07Kt)

I FIXED FILTER (PROPOSED)

i KURATA

0. 0 0. 1 0. 2 0. 3 0. 4 0. 5 0. 6 0. 7 0. 8 0. 9 1

FREQUENCY (He)

FIGURE 12: COMPARISON OF FILTERS PROPOSED SO FAR

0. 0 0. 1 0. 2 0. 3 0. 4 0

FREQUENCY (It)

FIGURE 13: COMPARISON OF FILTERS PROPOSED SO FAR

(DOUBLE INTEGRATED)

7-15
5-1043 DOWN ONPHGMq-S

n. *9l faccl ration

25. 30.(»e) 4t

AAA f\ f\ A a r\ r\j IN

30.<««)

•0.

C. 5. 10. 0 IS . ca OHplKwt by OMIIMM't Hilt*

15. 20.

0.

0. IS

o. v

C 3, » OltstKCBMt by nWTk'I FUTCB

15. ZO. 25. 30.

30. <«•*>

30. <•«)

30. <•«)

?0>

FIGURE 14: AN EXAMPLE CALCULATION OF DISPLACEMENTS BY THE FILTERS PROPOSED SO FAR

7-16
SOME RECENT DEVELOPMENTS IN NATIONAL AND

INTERNATIONAL SEISMIC DATA EXCHANGES

James F. Lander

National Geophysical and Solar-Terrestrial Data Center

Environmental Data Service, NOAA

Department ,of Commerce

Some Recent Developments in National and International

Seismic Data Exchanges

lmost since its beginning, seismology has been an international

cience. The fact that stronger earthquakes are recordable at great

istances was first observed in 1889 and gave rise to the need to entralize certain data-handling activities internationally. The

se of this data has given invaluable insight into the distribution

f earthquakes, the properties of the earth's interior, the moveents of continents, and the cause of earthquake.

Activities in data observations and exchange are expanding and

hanging under the pressure of new applied research programs, such

s the U.S. Earthquake Hazards Mitigation program, and new opporunities are arising in the use of digital technology. This paper

riefly reviews some of the trends in seismic data exchange, reports

n the status of a number of data-related activities particularly

s this involves World Data Center-A and NOAA data activities, and

jutlines some of the problems to be faced by data centers in the

mmediate future.

rends

s an overview, three phases in international data exchanges are onsidered. The first is characterized by the exchange of observed

eismic phase data (station bulletins) and graphical or handalculations of epicenters. Beginning in 1918 the International

eismological Summary (ISS) began to collect and publish seismic

eports for numerous stations and to calculate the source of the

arthquake. This continued through 1963. It essentially ceased

Ihen ISS was unable to cope with the large quantities of data that

egan to become available after World War II. It could not adapt

So the use of the digital computer then coming on the scene. The

jask was taken over by the International Seismological Centre,

hich from the beginning used digitial computers, including comuter composition of manuscripts. The Bureau Central International

e Seismologies was active in the early 1950s and 1960s in compilng data on many earthquakes that ISS was unable to include. The

unction was no longer needed as the ISC was able to provide full

overage. In 1976 the BCIS was terminated and replaced by the

egional European-Mediterranean Seismological Center.

The preliminary Determination of Epicenters program also

ncreased its activity in the 1950s to provide a fast data service,

nd was able to make the switch to computer-processing in the early

960s to meet needs for improved quality and quantity of data.

In 1957 the International Geophysical Year saw the installaion of stations in Antarctica and other remote places. It also

jaw the establishment of the World Data Center system, but the data

'xchanged were primarily station bulletins. Seismograms were xchanged only by writing to many observatories to borrow original

ecords or copies. The records so borrowed were of unknown calibraion and in different formats.

The second phase was marked by the installation of the Worldide Standard Seismograph Network beginning in 1961. For the first

ime a global network of data in standard form with excellent timing

nd known response characteristics was available at a reasonable

ost. To date, the master file contains about 5 million records.

staggering total of about 70 million copies have been distributed

round the world. The benefits to science of this network are

ieyond calculation.

8-1
In 1974 the "Guide to Data Exchange through the World Data

Centers" was changed to provide for automatic seismogram exchange

for large or high-interest earthquakes. This brought in many more non-standard stations including those from the USSR network. To date, 22 events are included in the exchange and up to 300 stations

are included in the program.

Lastly, we are moving into a phase characterized by digitial

recording and processing of data. Initially, phase-arrival times were converted to digitial values and large quantities of data

could be handled. The resulting digital hypocenters could be

plotted automatically on maps.

In 1970, building on earlier experience including LASA, a Long-Period High-Gain network (HGLP) was established with instruments in environmentally controlled containers. An eleven-station

network was installed worldwide. The digital recording allowed more sophisticated wave analysis than simple arrival time and

amplitude. Soon, however, the higher gains that were possible by

recording in bore holes at depths of 100 meters justified a network of Seismic Research Observatories (SRO). Some of the HGLP

stations are being modified to have electronic characteristics

similar to the SROs. A global network of up to 20 digitally

recording gravimeters is being installed to provide information in

the ultra low-frequency part of the spectrum. Also it was decided

to modify 15 of the WWSSN stations for digital recording. Altogether this constituted a global digitally recording network of

about 50 stations. Data from the network will be put on two network data tapes.

Digital recording is also being increasingly used for strong

motion recording — most of the important earlier analog records

have been manually digitized and a new worldwide program for digital

arrays is being planned. Ocean bottom seismometer recordings are also increasingly being used.

The IASPEI at its Durhan meeting recognized the importance of

digital seismology by creating a Commission on Digital Seismology

to promote its use and a Working Group on Digital Data Exchange in

its Commission on Practice.

WDC-A is compiling an inventory of digital recording seismic

stations worldwide and USGS is compiling a manual on the digital

stations in the global Digital Seismic Network.

It is clear that the impact of this digital data will be very

important for seismology and that the data centers must be prepared

to modify their traditional services to provide the new services

that will be needed.

This phase is also characterized by the growth of heavily

instrumented, often temporary local and regional networks. These

present unusual and largely unsolved problems for data archiving.

Current Activities

The historical development outlined above provides a perspective on current activities as is discussed over the next few paragraphs.

Analog Data

Advances in technology have made it possible to use essentially

standard microfiche cameras to reproduce copies of seismograms on

8-2
108-mm film, 24 images to the chip. They have essentially the same resolution as cameras specially designed for seismogram filming 18

years ago. There will be considerable savings to the users. In July

1978, NGSDC will begin filming data from the WWSSN network in this

format.

Historical Microfilming Project

WDC-A is beginning a project to microfilm seismograms from around

the world with the support and technical guidance from the USGS.

Through the project we hope to save a collection of records from

selected stations worldwide for every day and from many stations

for selected events. This year will be used for planning and

acquiring equipment. Filming will begin next year with some North

American data; and foreign participants will be contacted. The program may last five or more years and involve a photography team

working with local experts and station directors. A Working Group

in IASPEI will give advice to this project.

Digital Data

Of the Global Digital Seismic Network mentioned earlier the IDA data

are already being distributed by NGSDC. The SRO digital data are available for use at the data center and I expect that within 3 to

9 months all the data from this station will be available through

NGSDC and the World Data Center as network day tapes (two tapes per

day) and stacked plots and punched cards. Additional data products

may be offered as they become standard requests.

On May 2-5, 1977, there was an International Workshop on Strong-Motion Instrument Arrays in Honolulu sponsored by the (U.S.)

National Science Foundation and UNESCO. It was decided to recommed

as highest priority the instrumenting of six sites worldwide with

ian extensive array of strong-motion accelerometers in the hope of

recording near-source activity in several kinds of geologic settings.

Some instruments will be in bore holes, some centrally and digitally

recorded. In all, some 28 sites were recommended for instrumentation. This program promises to be as important to the near field

studies as the WWSSN program was to global seismology. We expect

to have this data in the World Data Center system.

The need to develop seismic histories has long been felt.

Having such information on computer readable files is even more

advantageous in that huge files can be searched for events of particular characteristics such as locations, size, depth, etc. In the

lmid-1960s the Coast and Geodetic Survey started to compile such a computer file. The file now has data from 11 sources including the

PDE and ISS and information on nearly 150,000 earthquakes. This

file is actively being expanded and will have data from 23 sources

I iby this summer (including the UMA file). We are collaborating with

jthe USGS in this (see Table 1).

Internationally, an even more ambitious project is being proposed. An IASPEI Working Group on Global Digital Data Bases met in

jParis in March with UNESCO support to design a definitive Global

Seismic Data Base. It was recommended that an activity be attached

to the ISC to develop a comprehensive file representing the best

possible values. This file would include a substantial amount of

pr e-instrumental data, some of it several thousand years old. Funding for this project is being sought.

One of the oldest seismic data bases is intensity data. This

data form is particularly useful for engineers. It, too, can benefit

8-3
from digitization and analysis. We have developed a computer file

of about 115,000 felt reports mostly from the U.S. Earthquake

periodical (1928-1975). This file has information on the date,

location and magnitude of the earthquake, and location and intensity

of the community reporting the effect. It allows the listing of

felt reports in locations adjacent to sites of interest, construction of maximum reported intensity maps, and analysis of regional

variations of intensity and its attenuation with distances. This

file is being checked and will be available this fall.

The Modified Mercalli Scale of 1931 consists of 12 steps. Each

step is defined by a number of descriptives or elements. Mr. R. J. Brazze, an NGSDC scientist until recently, completed an analysis

of how each element behaved with respect to attenuation and found a number of apparently misplaced elements in the scale. His analysis

"Re-evaluation of the Modified Mercalli Scale for Earthquakes Using

Distance as a Determinant" shows where items not formally in the

scale would fit. His study has opened up the way for a quantitative

re-evaluation of the MM scale, including using weighting factors,

and for computer evaluation of past and future intensity reports.

The World Data Center-A for tsunamis was relocated from

Honolulu to Boulder in 1974. It had not been particularly active

as part of the International Tsunami Information Center. In the

intervening period, a catalog of tsunamis for Alaska and Hawaii were updated and reissued. All of the mareograms showing tsunamis are being microfilmed. We have copied the most recent ten years of

U.S. records and will complete the microfilming project in a year

for nearly 100 years of records. We are making a collection of

photographs of tsunami effects. The present Guide for data

exchange in tsunamis lists only a few stations as participating in

the exchange. A new Guide is in preparation. WDC-A is in a position to digitize selected tsunami records and to prepare special

packages of seismic data including focal mechanisms of interest to tsunami records. The international exchange of data for tsunamis

is far less well developed than the seismic data and probably

should be improved.

The World Data Center-A has been supportive of various international projects. For the current International Geodynamics

Project we are publishing data and information publications such as the Geodynamics International Series and data publications such as the "Summary of Earthquake Focal Mechanism for the Western Pacific —

Indonesian Region" and a World Heat Flow Map. For the U.S. Geodynamics program we are just completing a Directory of U.S. Data

Bases that are relevant to the Geodynamics Project. We expect to provide similar or other needed services for the upcoming Crustal

Dynamics Program.

Next year there will be a project involving observations of

seismic activity on the ocean floor called Project ROSE. Approximately 60 ocean bottom seismometers will be used in an array on

the River Fracture zone near the tip of Baja California. The data

will be digital and eventually released through the data center

system

.

Finally, I should mention our guest worker program. Because

of the growing size of the data bases and the need to use sophisticated search and retrieval systems it is often useful for scientists

to visit the data center and spend more days, weeks, or years in

residence. NGSDC supports such visits through a variety of mechanisms and to varying degrees. Scientists in the Solar-Terrestrial

Physics part of our data center have availed themselves of this

8-4
opportunity to a considerable degree — particularly Japanese

scientists. Dr. Yukutake, a Geomagnetician, spent 6 months with

the center last year. Inquiry is invited.

Problems and Opportunities

From the activities described above it is clear that major changes in

data services will be made possible because of the new digital environment. Data centers must continue to supply analog and interpreted

data to less sophisticated users, be active in proselytizing the new

techniques, and must provide advanced services for advanced users.

The latter includes expertise in use of computers, remote terminals,

interactive graphics, advanced plotters and digitizers, and data

base management systems, as well as advances in science.

A second problem area is the expansion of the user base to areas outside classical seismology. The earthquake prediction,

energy exploration, geodynamics, and natural hazards programs are examples of new emphasis activities that bring a wide range of user types and related but different data sets such as tilts,

gravity, and heat flow mechanisms. There is a growing need for data

synthesis for different uses and users.

A third set of problems come from a change in scale of many

needs and projects. We find our data bases that are so good in

solving global problems are inadequate on the local scale. At present, there are many projects to observe small areas intensely

in a variety of ways for a limited period of time. We must discourage the use of global data for local problems. What part of these

data may be of use to other and future users? This question has

not been examined in detail.

The opportunities are great for providing data worldwide as a resource for science and its applications. The lesson from the

past is clear! Keep up or drop out.

8-5
TABLE 1: SEISMIC DATA FILES IN HDF - MAY 1978

1. Preliminary Determination of Epicenters 1937-1977

2. United States Earthquakes 1928-1975

3. Gutenberg and Richter "Seismicity of the Earth" 1892-1952

4. Southern California - Cal Tech 1932-1975

5. Northern California - Berkeley 1910-1975

6. Earthquake History of the U.S. 1638-197Q

7. Central California Regional Network 1969-1273

8. Coastal and Offshore Earthquakes of the Pacific Northwest 1853-1973

9. Seismicity of the European Area - Karnik 19Q1-1955

10. International Seismological Summary 1218-1161

11. Bulletin Mensuel BCIS 1950-1961

12. OCSEAP 1973-1277

8-6
9

REPORT ON THE INTERNATIONAL WORKSHOP ON STRONGMOTION EARTHQUAKE INSTRUMENT ARRAYS

MAY 2-5, 1978, AT HAWAII

Tadayoshi Okubo

Ministry of Construction

and

Hajime Tsuchida

Ministry of Transport

Report on the International Workshop

on Strong-Motion Earthquake Instrument Arrays

May 2-5, 1978, at Hawaii

l July 1977, the International Association for Earthquake Engisering (IAEE) appointed a Steering Committee for planning an Interitional Workshop on Strong-Motion Earthquake Instrument Arrays,

le first meeting of the Steering Committee was held October 17 -

3, 1977, at Caltech and resolved the following:

* Primary Workshop Goal

The primary goal of the Workshop was to develop a workable

plan for the possible future deployment of dense strongmotion earthquake instrument arrays with primary emphasis

on ground motion studies.

* Workshop Content

The major application areas that should be considered are

0 Source Mechanism Studies

° Wave Propagation Studies (Source to Site)

° Local Effects

and the major subject areas of the Workshop are

* Favorable Array Locations

0 Tectonic environments

° Site conditions

° Short-term and long-term rate of data recovery

° Operating costs over direct instrument related costs

° Local technical and operating assistance

* Array Design

° Number and distribution of arrays

° Array configurations

0 Number of dimensions

° Extent

° Geometry

° Density

Instrumentation

0 Specifications

0 Capital cost

Array Construction; technical and economic

considerations

0 Site studies

° Site preparation

0 Equipment installation

Array Operation; technical and economic consideration

0 Instrument maintenance

0 Data retrieval, processing, archiving and distribution

9

9-1
* Implementation

° Organization Plan

° Adequacy of existing organizations

° Modes of international cooperation

° Unique contributions of different countries

Because the major subject areas had to intersect with each of

the application areas and vice versa, the content of the Workshop

was thought of in terms of a type of matrix as indicated in

Table 1.

Resolution

After five days, the Workshop published the following resolution:

Resolution of International Workshop on Strong-Motion

Earthquake Instrument Arrays

May 2-5, 1978

Honolulu, Hawaii

The protection of life and property from the devastating

effects of earthquakes is an urgent worldwide problem. An understanding of the nature of strong earthquake motions is of crucial

importance in solving this problem. At the present time, however,

there is a scarcity of engineering data acquired near the center of

destructive earthquakes and existing instrument arrays are inadequate to provide the necessary data. Yet there is a high probability of occurrence of destructive earthquakes in different parts of

the world in the next decade. The participants in this international workshop unanimously recommend the earthquake-threatened

countries and other concerned countries and organizations make a concerted effort to establish a comprehensive worldwide system of

specialized strong-motion earthquake instrument arrays capable of

resolving the nature of the earthquake source mechanism, wave propagation, and local site effects. As a first step, the following

specific recommendations should be implemented.

* The International Association for Earthquake Engineering in

collaboration with the International Association for Seismology and Physics of the Earth's Interior form an International Strong-Motion Arrays Council.

* The earthquake-threatened countries individually and

collectively initiate the immediate installation of minimal

arrays of 10 strong-motion instruments at least at the 28

worldwide sites identified by this Workshop.

* High priority be given to the design and installation of more elaborate source mechanism, wave propagation, and

local effects arrays particularly at the 6 critical sites

identified.

* A mobile strong-motion instrument array capable of making

source mechanism, wave propagation, and local effects

measurements be established and maintained for deployment

to sites immediately following the occurrence of a major

earthquake for measurements of aftershocks.

9-2
9-3

10

EXPECTANCY OF MAXIMUM EARTHQUAKE MOTIONS

IN JAPAN

Makoto Watabe

Yoshikazu Kitagawa

Building Research Institute

Ministry of Construction

Expectancy of Maximum

Earthquake Motions in Japan

In the Seismological Bulletin of the Japan Meteorological Agency[2],

all the relevant information on earthquakes are reported monthly.

The location of the J.M.A. stations and the characteristics of

strong motion seismometers used by J.M.A. are shown in Figures 1 and 2. The phase time of P- and S- waves, the maximum displacement

amplitude for EW, SN, and UD components, the predominant period of

seismic wave, the initial motion for each component, the epicentral

distance, and the P-S time are described in List 4 of the J.M.A.

Bulletin. These data are transferred to a single card and stored

on magnetic tape under the station name.

The data used in this paper are mainly the maximum amplitude

and the predominant period for the period of 1967 to 1973. The

number of data for each year is shown in Figure 3. In addition,

the magnitude reported in the J.M.A. Bulletin (List 3) is determined by Tsuboi's formula, so that we used only the earthquake data

with depth D<60m because the earthquake damage to structures is

caused by shallow earthquakes with depth D<60 Km. Further, Tsuboi's

| formula is limited to earthquakes with depth D< 60 Km.

! Analytical Method

! The ground motion in frequency domain (G(jw)) will be generally

expressed in the following equation:

G(jw) = FS( jw) *EA( jw) *MT( jw) *EC( jw) *ST( jw) (1)

' where FS(jw) means the spectrum of focal mechanism, EA(jw), MT(jw),

i and EC(jw) represent the transmission function around the hypocenter,

the function in the mantle-rock, and the function of the wave propa-

| gation in soil layers. Now equation (1) can be reexpressed as the

following equation:

G(jw) = FS( jw) *TF( jw) *ST( jw) (2)

where TF(jw) means the overall transmission function for the generation of waves at the hypocenter and the propagation of earthquake

i waves. Then all functions in equation (3) can be combined as

f o 1 lows

,

G(jw) = FS( jw) «TF( jw) *ST( jw) (3)

= FS ( jw) «TF( jw) *ST( jw) (4)

I The equations (3) and (4) stand for the theoretical seismic records

based on the focal mechanism and for the ground motion.

On the other hand, the long period (10.0>T>1.0 sec.) corresponding to the ground characteristics in a fairly large area is

J amplified by depth varying from a few hundred meters to 1 to 2 kilometers. The short period (T^l.O sec.) corresponding to the

ground characteristics beneath a structure is affected by depth of

soil less than a few hundred me ter s . [ 3 , 4 , 5

]

In this paper, the study is carried out from the viewpoint of

equation (4) in the range of long period (6.0^T22.0 sec).

The magnitudes (M) reported in the J.M.A. Bulletin are deteri mined by following Tsuboi's formula.

10-1
M = -i- log (ANS + Aew) + 1-73 log a - 0.83 (5)

where ANg , A

EW , and A are the maximum displacement (p) in NS and EW

components, and the epicentral distance (km), respectively. When

the values of A^g and A

£W in equation (5) for a single earthquake

are observed at various stations of J.M.A., the mean value of

magnitude is reported as the magnitude of the earthquake in Bull,

of J.M.A.

From equation (5), the root mean square value (R.M.S.) of the

maximum displacement amplitudes in NS and EW components is

expressed in the following equation,

NS A

2 = 10* EW

(6)

where g = M - 1.73 log A + 0.83

As mentioned above, since the magnitudes in the Bulletin are

the average ones determined by equation (5) from many observation

stations, then the value of 10^ can be regarded as a standard value

of R.M.S. of maximum displacement amplitudes.

Under the assumption that the value of 10° is always regarded

as the standard one in the range of long period (6.0^T^2.0 sec)

reflecting the deep ground characteristics in fairly large area,

the following values are calculated at every J.M.A. station,

F = _ z m

i=1 JJ

A

NS (T

i

} + AEW (T I> / 10t (7)

f =± I m

i=l

'A

NS

(T

I

} + A

EW (Ti

} " 106 (8)

st log | f

where T and m denote the period of the phase giving maximum displacement amplitude and the total number of the earthquake data,

respectively. Consequently, the values of F and f (or F st

) denote

the mean and the variation of the observed values to standard ones

at each station of J.M.A.

On the other hand, since the period T being reported has an interval of 0.1 sec in the Bulletin, it is most advisable to vary T. However, the maximum displacement amplitude in NS and EW components

are neither always in the same phase nor in the same period. In this study, the concept of the period window AT is introduced, such

as AT = T

i + 0.2 sec (in case of J=3), AT = T- + 0.3 sec (in case

of J=4), and AT = T^ + 0.4 sec (in case of J=4), where T^ dnotes

the medium value of the period window and represents the predominant

period of seismic wave.

In addition, the values of F, f (or Fst) with period T^ are smoothed by the weighted mean method.

10-2
Regional Distribution of Ground Characteristics

Figures 4 and 5 show the curve of F and F ^ values calculated by

equations (7) and (8) and the number of the earthquake data for

the case of J=3 , 4, and 5 at Tokyo. Judging from these figures,

the number of earthquake data for the case of J=3 and 5 are more

than 15 and under 50 in the range of period T ^ 4.0 sec, respectively,

and the values of F are little larger than 1.0 and 2.0 - 8.0 in

the range of period T<3.6 and period T>3.6, respectively. There

is good agreement among the curves of F values vs. period for the case of J=3, 4, and 5 and the characteristics of F and F

s(

_ values

correspond well. Consequently the degree of the "quakability"* is

high in the range of fairly long period.

On the other hand, the characteristics of F values look to be more suitable than F

gt

values as the magnification factor of ground

characteristics. As a result, the characteristics of F values are mainly dealt with here.

Judging from the characteristics of F value at all stations

of J.M.A. , the pattern of the characteristics mentioned above is

divided into two main classes: (1) there is no great difference

among J=3, 4, and 5; (2) there is a remarkable difference among

J=3, 4, and 5. In the first case, it means that the earthquake

data for all cases are almost identical or enough so not to affect

the whole characteristics in number. In the second case, quite new data are introduced by making AT larger; these data have a great effect on the characteristics of F values.

Therefore, the criterion for the evaluation in this paper is

as follows: the result of J=3 in the first case, and the result of

a small J number in the second. Under these criteria, the regional

distribution maps of F values corresponding to the period of 1.0,

2.0, 3.0, 4.0, 5.0 sec. are contoured in figures 6(a)-(e). As seen

in these figures, the areas where the magnification factor of the

ground characteristics is considered to be high are generally as follows: the coast on the Japan sea in the Tohoku district from

the Hokkaido district; the southern area of the Kwanto district

from the area of the Kinki district; the southwestern area of the

Shikoku district from the Kyushu district. However, these are

general tendencies only.

Expectancy Based on Seismic Activity and Soil Characteristics

It is fairly difficult to know the detailed underground structure throughout Japan. However, to examine the ground structure

down to a fair depth several experiments in the vicinity of Tokyo

were carried out by the research groups of Dr. Y. Ohta[3-6] and

Dr. E. Shims[7]. The former group analyzed the part regarded as S-wave on the seismographs at Tokyo under the Niigata earthquake

(Model 1 and 2) and made their experiment by means of a SH-wave

generator and in a deep borehole (Model 5). The latter group made

seismic exploration tests in the Tokyo district (Model 3 and 4).

These results are shown in Figure 7. As seen in Figure 7, the depth

of the base rock with the shear wave velocity V

g = 2 - 3 km/sec is

assumed to be 2 - 3 km. Figures 8(a)-(e) show the magnification

factor (U

Q /U^

n

) of the ground structure mentioned above by means of

Haskell's method, assuming the incident wave travels from the base

"Quakability" represents the degree of quake to the standard

value

.

10-3
rock. As seen in Figure 8, there is a tendency for the magnification factors in the range of long periods to resemble each other,

but to be remarkably different in the range of short periods less

than 1.0 sec.

The underground structure, representative of a fairly large area in the Tokyo district, can be geometrically determined.

Figures 9 and 10 show the average underground structure in the Tokyo

district and the magnification factor, respectively. Under the

assumption that the average underground structure shown in Figure

9 is the underground structure at the Tokyo station of J.M.A., the

standard ground characteristics shown in Figure 11 are inferred by

dividing the magnification factor shown in Figure 10 with the

characteristics of F value in Figure 4.

Consequently, the expectancy of the maximum earthquake motion

on ground with V

g = 0.6 - 0.8 km/sec in the range of the long period

( 6£T£2 sec.) is represented by the products of three factors: the

expectancy of maximum earthquake motion based on seismic activity

on the base rock with V

g = 2 - 3 km/sec; the magnification factor

of the standard ground characteristics; and the regional "quakability" of the soil layers. However, it should be remembered that the

final regional distribution of the expectancy of earthquake motion

will be different from that based on the seismic activity.. In addition, the effects of ground characteristics just beneath the

construction site and the concept of soil profile should also be

introduced in future research.

Acknowledgments

This study was partially supported by the New Aseismic Design Method

project monitored by Ministry of Construction, Japan[8] . The authors

wish to express their thanks to Prof. M. Ozaki, Chiba University,

and Dr. S. Hattori, Applied Seismology Division of I.I.S.E.E., for

valuable discussion and advice in the course of this research; to

Mr. Y. Serizawa and Miss K. Naito, ex-Member of B.R.I. , for drawing

figures and making fair copies; to Dr. S. Kawamura, Res. Eng. of

Taisei Corporation, for his help with a large amount of earthquake

data; to Miss M. Ota, Member of I.I.S.E.E., for typing.

Re f erences

1. Ozaki, M. , Kitagawa, Y. , and Hattori, S., "Study on Regional

Distribution of Maximum Earthquake Motions in Japan," Ninth

Joint Meeting, U.S.A. Japan Panel on Wind and Seismic Effects , UJNR, Tokyo, May 24-27, 1977.

2 . The Seismological Bulletin of the Japan Meteorological Agency , 1967-1973.

3 . Earthquake Excitation with Soil Classification Comparison of

Ground Characteristics up to 10 second Period of Tokyo and

Choshi by Low-Sensitivity Displacement Seismometer (in Japanese),

Report for Establishment of New Aseismic Design Method, Ministry

of construction, Japanese Government, 49-l-2-(2) Ken. Hou. 1,

1975.

4 . Earthquake Excitation with Soil Classification — On Seismic

Wave Spectra and Subsurface Structure , Report for Establishment of New Aseismic Design Method (in Japanese), Ministry of

Construction, Japanese Government, 50-l-2-(2), Ken Hou. 1, 1976.

10-4
5. Earthquake Excitation with Soil Classification — A Point by

Point Evaluation of Amplification Characteristics in Japan on Five Sec. Seismic Motions in Relation to Deep Soil Deposits

(in Japanese), Report for Establishment of New Aseismic Design

Method, Ministry of Construction, Japanese Government, 5 1—1—2—

(2), Ren. Hou. 1, 1976.

6. Ohta, Y. , Goto, N. , Shiono, K, Takashashi, T. , Yamamizu; F.,

"Shear Wave Velocities in Deep Soil Deposits — Measurement

in Borehole to the Depth of 3500 Meters at the Iwatsuki" (in

Japanese), Proc . of Seismological Inst, of Japan , No. 1, 1977.

7 . Sh ima , E . et_ al^ ; Experimental Study on Generation and Propaga - tion (in Japanese), Report of Tokyo Prevention Disaster Council,

1975.

8 . Earthquake Excitation with Soil Classification — On the Con- siderations of Earthquake Danger in and Around Japan (in

Japanese) , Report for Establishment of New Aseismic Design

Method, Ministry of Construction, Japanese Government,

5 l-l-l-C 2) , Ken. Hou. 2 , 1977.

10

10-5
/ha FIGURE 1: LOCATION OF J.M.A. IN JAPAN. FIGURE 4: CHARACTERISTICS OF F AND F VALUES AT TOKYO STATION.

st

&10.0

1.0

10.0

H:

V.

Horizontal Ca

(To =6

Vertical Comf

(To=5 ,sec.ho*Q5o)\ \

sec.ho*Q55)

.

0.1 1.0 1Q0

T(sec)

FIGURE 2: CHARACTERISTICS OF SEISMOMETER.

a

E

. O

2.0 -j1.0 -r

For Period of 1967-1973

,67*68

,69, 7o, 7i '72 '73

1—- year

FIGURE 3: EARTHQUAKE DATA IN PERIOD OF 1967 - 1973.

< 1000 -3

E

500-

1.0 2.0 3.0 40 5.0 60

1 PERIOD T (sec)

FIGURE 5: NUMBERS OF EARTHQUAKE DATA IN CASE OF J=3, 4, AND 5.

10-6
FIGURE 6: REGIONAL DISTRIBUTION OF F VALUES

IN PERIODS OF 3.0, 4.0, and 5.0

SECONDS

.

10-7
FREQUENCY(HZ)

FIGURE 7: PROFILES OF UNDERGROUND STRUCTURES (AFTER DR. OHTA,

DR. SHIMA, ET AL.

)

Model 1 to,et Model 2 J O**0

- - Model 3 l oHw

Model 4l sr,m0 Model 5 atto OHO 1 0

20

1.0 2.0

I H. '

' I

1

Density (g/cm

3

) Vs (km/sec)

Mean Value of S-V\tave Velocity

(knrVsec) Mean Value of Geological Density

(g/cm)

FIGURE 8: MAGNIFICATION FACTORS OF UNDERGROUND CHARACTERISTICS.

FIGURE 9: PROFILE OF AVERAGE

UNDERGROUND STRUCTURE .

40-r

15 2.0

FIGURE 10: MAGNIFICATION FACTOR OF AVERAGE

UNDERGROUND STRUCTURE.

IS 2.0 30 40 5.0 6.0

'—=— T (sec) FIGURE 11: STANDARD CHARACTERISTICS OF

UNDERGROUND CHARACTERISTICS.

10-8
11

DETERMINATION OF DESIGN EARTHQUAKE FOR THE DYNAMIC ANALYSIS OF THE FORT PECK DAM

W. F. Mar cuson III and E. L. Krinitzsky

Soils and Pavements Laboratory

Waterways Experiment Station

Vicksburg, Mississippi

Determination of Design Earthquake for the

Dynamic Analysis of Fort Deck Dam

Because of the slide in the Lower San Fernando Dam during the San

Fernando Earthquake of 9 February 1971, the U.S. Army Corps of

Engineers began earthquake studies of its hydraulic-fill structures

located in seismically active areas. Fort Peck Dam is a large

hydraulic-fill dam in northeast central Montana, on the Missouri

River, built during the period from 1933-1940. This paper describes

the investigations performed to determine the so-called "design"

earthquake used in the dynamic analysis of Fort Deck Dam.

Geological Setting

Fort Peck Dam is on the edge of the Williston basin. The sedimentary section is about 7,500 ft (2,500 m) thick at Fort Peck and

about 13,000 ft (4,300 m) thick in the Brockton-Froid area to the

southeast. (See Figure 1 for locations.) Strata dip a few tens of

feet per mile to the southeast.

The sediments at and near the surface consist of a thin,

irregular veneer of glacial deposits that are underlain by thick

Cretaceous shales. The shale encountered over much of this area is

the Bearpaw Formation, a dark gray clay shale, approximately 1100 ft thick, with bentonite seams. The Cretaceous deposits are underlain

by a succession of sandstones, shales, limestones, and dolomites. In some areas, evaporite deposits are found. The section extends to

the Cambrian and, in turn, the Cambrian lies disconformab ly on crystalline basement rock.

Structural features, in addition to that of the Williston

basin, are the Bowdoin dome, the Opheim syncline, the Blood Creek

syncline, and the Redwater anticline (Figure 1). All of these

features, including the Williston basin, date to Paleozoic time.

They are essentially gentle features and have not undergone intense

tectonism at any time in their history. They have probably been

inactive since the Cretaceous.

Normal faults trending from northeast to southwest and downdropped to the southeast are shown in Figure 1 as the Hinsdale,

Weldon, and Brockton-Froid faults.

/ Geological Studies

Since earthquakes are caused by movements along faults, the geologic

and seismologic studies centered on the earthquake-generating potentiality of each fault. The study was complicated by uncertainty as

to whether Tiger Butte is an active tectonic feature, and the

interpretation of linears seen on airborne imagery of the area. Preliminary examinations were made of (1) Earth Resources Technology

Satellite (ERTS) imagery, (2) air photography by the Soils Conservation Service (scale 1:20,000), (3) U.S. Geological Survey (USGS)

fault studies in the Fort Peck area, (4) Side-Looking Airborne Radar

(SLAR) imagery, and (5) aerial reconnaissance and field examination

of faults and lineations in the Fort Peck area. Also during this

time, inquiries were made to oil companies known to be working in

the Fort Peck area to learn if they might provide information on

faults from their seismic profiles. Such information was obtained

with the stipulation that the sources and certain aspects of the

information to kept confidential.

11-1
Seismological Studies

Concurrent with the geological work, a seismological study of the

Fort Peck area was conducted .[ 1

]

The historic earthquake events, both in the Fort Peck environs

and in the larger region in which Fort Peck is located, were reviewed. Figure 2 shows the distribution of earthquakes in this

region for those events with Modified Mercalli (MM) intensities

greater than V. The Canadian Department of Energy and Mines was contacted to obtain what was then unpublished data on the earthquake events in Canada. Contemporary newspaper accounts of the

1909 earthquake that was felt throughout the Fort Peck area were restudied. On the basis of the restudy, the maximum intensity of

the 1909 earthquake was downgraded from VIII to VI and its epicenter

relocated. A maximum magnitude earthquake of 7.5 was assigned to the zone of active seismicity in western Montana and the ground

motions were attenuated to the Fort Peck Dam. Maximum earthquakes

for the faults in the Fort Peck area were assigned and their motions

attenuated to the dam.

Ground motions in eastern Montana have been monitored by an array of seismometers set up to measure long-distance events such

as nuclear explosions in other countries. The Montana array was

part of the Large Aperture Seismic Array (LASA) program conducted

under the Vela project by the U.S. Air Force. Teledyne Corporation, the contractor for the LASA work, made a limited study of

unexamined small events in Montana on the chance that they might

identify activity along faults.

In collaboration with Waterways Experiment Station consultants

earthquake records were selected and rescaled to provide time histories of ground motion for use in dynamic analyses of the dam.

Seismicity in the Fort Peak Area

Historic seismicity for events greater than intensity V on the MM

scale in Montana and adjacent areas is shown in Figure 2 for the

period 1852 to 1967. For the eastern half of Montana the period is

extended to 1975 since no subsequent events have occurred. It is

noteworthy that all of the eastern half of Montana is seismically

quiescent. Further, there are no groups of earthquakes observed

that might be interpreted to trend into the eastern Montana area. Western Montana, on the other hand, is seismically active along the

Rocky Mountain front.

The major event in the Rocky Mountain area of Montana was the

Hebgen Lake earthquake of 17 August 1969 which occurred at the upper

end of Hebgen Reservoir. It had a magnitude of 7.1 on the Richter

scale and an intensity of X on the MM scale.

The seismicity east of the Rocky Mountain front is illustrated

in Figure B5 of Reference 1 by all observed events for eastern

Montana and the adjacent portion of Canada. The total historic

record consists of 14 events of which only two, those of 1909 and

1943, were as great as intensity VI. All others were intensity IV to as little as a possible intensity II.

Of all the earthquakes, only the events of 1956 have an epicenter that plots near a known fault; it is adjacent to the Hinsdale

fault. The earthquakes of 1943 and 1946 may be erroneously located

and may be relatable to the Brockton-Froid fault, but that is, at best, only a distant possibility.

11-2
A more significant possibility is the lineup of earthquakes

along a projection to the northeast of the Hinsdale fault. The

1956, 1972, 1968, and 1909 earthquakes appear to trace along an extension of the limited mapped extent of the Hinsdale fault.

The 1909 earthquake appears along the above-mentioned trend

because of reinterpretation that was made during this study.

Reevaluation of the 1909 Earthquake

The 1909 earthquake, the strongest in this region, was rated at intensity VIII though very little damage was reported. The earthquake was felt over a large area, approximately 500,000 square

miles (1,300,000 square km); its epicenter was located at 50° N and 105° W. Contemporary newspaper accounts were reexamined and on

the basis of this information the earthquake was downgraded from

intensity VIII to VI and the epicenter moved to the new positio[l].

As a result of the reexamination, the epicenter of the 1909 earthquake lined up with epicenters of the three other earthquakes. The

'four earthquakes indicated a possible trend of seismic activity

that extended from the Hinsdale fault to the northeast and into

Canada for approximately 200 miles (320 km). Alternatively, these

four earthquakes might be unrelated to a single fault or associated

with minor cross faults.

Fault-Plane Solution for the 1972 Earthquake

The 26 July 1972 earthquake, known as the Bengough , Saskatchewan,

earthquake, was a minor event with an MM intensity of IV. However,

the data were suitable for a fault-plane solution. The interpretation indicated a strike-slip motion on a nearly vertical plane

striking approximately N 30° E. These results were later

published[ 2] .

The fault strike roughly matches that of the Hinsdale fault

'and can be used as an evidence for interpreting an extension of the

Hinsdale fault or of discontinuous faults that follow this trend.

Hicroearthquakes

the restudy of the LASA data for 1966 to 1968 located 10 epicenters

jf tnicroearthquake s

. * These microear thquakes had very low orders of

magnitude, roughly less than 2 on the Richter scale, and probably

are not recordable at all in terms of MM intensity. Locations are

shown in Figure 3.

A large group of events not shown in Figure 3 were recorded

southeast of the Brockton-Froid fault. These events were found to

lave been blasts associated with strip mining for coal. A check

was also made against oil extraction and water flooding- in oil

fields; no tremors were associated with these activities. Thus the

epicenters in Figure 3 have been closely scrutinized and are believed

:o be those of very small but valid earthquake events.

I

The small number of these events suggests that the Fort Peck

irea is one of very low seismic activity. In central California

:here would be hundreds to thousands of such events for a comparable

irea and period of time. Also the events would generally line up

fo reflect the patterns of traces of active faults. In the Fort

j?eck area there is nothing comparable. Major fault trends are

Letter from Teledyne-Geotech to OD, 28 June 1974.

11-3
hardly reflected at all by these events. There is one event beyond

the southwest end of Hinsdale that might be related to the Hinsdale.

Two events occur parallel to the Weldon fault but each is about

15 miles (24 km) away from the fault. Similarly, two events are about 20 to 15 miles (16 to 24 km) from the Brockton-Froid fault.

Thus, none of the microearthquakes are closely associated with

any of the faults. There were no microearthquakes relatable to the

Tiger Butte feature.

Relation of Earthquakes to Faults

The historical record of earthquakes in the Fort Peck area[l]

suggests that:

° The area is one of low seismicity.

° A low level of seismicity appears to be possibly related to the Hinsdale fault and to a trend to the northeast for a total length of about 200 miles (320 km).

° The fault-plane solution for the 1972 earthquake may reflect

the trend of the Hinsdale fault but it does not necessarily

imply that there is a continuity to this trend.

° No earthquakes have been associated with the Weldon or the

Brockton-Froid faults or for the area between them. The

1943 and 1946 earthquakes might have been mislocated and

might be relatable to the Brockton-Froid fault, but the

possibility is tenuous at best.

0 The microearthquake data (Figure 3) do not confirm activity

on any of the faults though one event lies along the trend

of the Hinsdale fault.

In general, the implication of the earthquake data is that

there may be some relationship between seismic activity and the

trend of the Hinsdale fault for a distance of 200 miles (320 km).

However, there is no certainty of this relationship and the trend

may be there without there being any continuity in the faulting.

The data suggest that the major structural features exhibit very

little earthquake activity.

The area close to the Fort Peck Dam and its reservior shows no evidence of seismic activity.

Geological and Seismological Interpretation of Faults

Following are geological observations on the main structural

features of the Fort Peck area that are relatable to earthquake

activity. The observations are made with the benefit of the seismological background developed above.

For the faults under study, it must, first, be established

that they are deep-seated and have roots in the crystalline basement rocks. Such information is essential to qualify a fault for

its capacity to generate earthquakes. In clay shales of the Fort

Peck area, faults are generally know to be "rootless," or formed

as large slumps during sedimentary deposition. The faults are no longer active and do not act as a source of earthquakes. Second,

one must establish whether faults that reach into the basement

rocks are active or inactive. To be active, the fault must show

11-4
evidence of movement in the recent past,* thus suggesting that it

could move in the near future. The geologic evidence suggestive of

movements may be of several kinds, and there is no single set of

criteria.

Following are discussions of the major faults in the Fort Peck

area.

Weldon Fault

Oil company sources confirm that seismic profiles show the Weldon

fault has roots in the crystalline basement rocks. The fault is

interpreted to have moved more than once during the lower Paleozoic.

The fault extends some 40 miles (64 km) farther to the southwest

and 9 miles (14.4 km) farther to the northeast than current geologic and tectonic maps indicate. However, seismic profiles did.

not indicate whether the fault was continuous up to the BrocktonFroid fault. There is a lack of deep subsurface information in

eastern Montana because the presence of Devonian evaporite

deposits prevents discrimination of the basement in seismic

interpretation. There is evidence that solution and collapse in

the evaporites are responsible for some faulting in this area.

Such faults could not generate damaging earthquakes.

Microseismic data, discussed earlier, suggest that the Weldon

fault is either inactive or has a remote possiblility for an extreme

low order of activity. Historic seismic records show no earthquakes

at this fault.

A study of air and satellite imagery and two reconnaissance

flights failed to turn up any field evidence of activity along this

fault. It was concluded that this fault is inactive.

Brockton-Froid Fault

The record of historic earthquakes and microearthquakes shows a low

order of seismicity in the general area, but the evidence is inconclusive with regard to the relation of seismicity to the fault.

None of the events were located on the fault, and it is unlikely

that any should have been interpreted as falling on the fault.

Even if one or two had been so interpreted the level of activity

would still be very low.

Subsurface displacements along this fault were mapped from

boring data and oil well records by the USGS in Denver. [3] Also

surface maps of the fault area were prepared . [4 , 5 ] Indications

are that Pleistocene deposits were involved in the faulting and

were dropped into the fault zone in many places. This implies

that the Brockton-Froid fault was active.

Several linears, or lineations, formed by the Brockton-Froid

fault were seen on the airphotos and during the overflights. One

linear is a topographic high composed of gravel. This linear was

mapped and interpreted to be gravels derived from deposits older

| than the adjacent till and cut by active faulting. [4,5] This

linear runs into sand dunes to the northeast. A boundary of the

! sand dunes correspond to the linear and to a depression between

Froid and Big Muddy Creek.

Reference 1 provides a full review of the evidence and judgments

used in classifying fault activity.

11-5
During overflight, other linears were observed in the BrocktonFroid area. Figure 4 is a photograph taken under conditions of lowsun angle of traces that extend undisturbed over rolling hills and across several drainage lines. A possible interpretation for these

features was that they are the result of faults that have moved

very recently.

These linears and the postulated fault traces were examined in

the field.

When the linears first seen in the overflights were followed

on the ground, they were found to have none of the features associated with fault movements. There were no vertical displacements

and no horizontal translations in spurs. Instead, there were only

gentle indentations that were traceable over the topography. Oldtimers among the Indians living here said that cattle drives took

place in this area early in the century and that as many as 80,000

head of cattle were driven yearly along a route that paralleled the

Missouri River and was far enough away from it to avoid the bluffs

and deep ravines that occur near the river valley. These trails

went through valley bottoms and over adjacent spurs and ridges.

The trails are covered with colluvium now, but they are still traceable. It was decided by all those persons who made the examinations

in the field that the linears such as those in Figure 4 were old

cattle trails.

During the field examinations, no evidence could be found for

the faults that had been mapped.

The mapped faults are real in the deep subsurface where

Jurassic marker beds are shown to be displaced several hundred feet

in oil well borings. Surface manifestation of the faults is another matter. A linear ridge of brown chert gravels was found on the surface. These gravels were described as being different in

age from lighter colored gravels, called till gravels, that occur irregularly but frequently on topographic highs throughout this area. The brown gravels are called outwash deposits of late

Wisconsin age derived from the older Flaxville Formation. The

Flaxville Formation outcrops about 50 miles (80 km) distant and

contains brown quartzite. The brown color of these gravels was used

as a time-stratigraphic designation. Though faults that form the

boundaries for these gravels could not be found, the faults might

be present. Test trenches would be needed to determine if they were or were not present where they are mapped.

There is reason to question the use of brown color or gravels

to interpret age and origin of the gravels. Other hill crests were examined and more brown gravel was found. The brown color could be

a secondary effect, resulting from soil-forming processes. In this case, the brown color and soil-forming processes may have contributed to an error in interpretation.

The historic earthquakes are a problem because their epicenters

cannot be located with any degree of exactness. Old-timers among

the Indians living in the Brockton-Froid area were asked if they

knew of ground breakages associated with earthquakes that they could

remember from their own experience or oral traditions. They knew

of no ground breakage, though some remembered the 1935 and 1943

earthquake events. They could recall no other events from their

oral history.

It was concluded that this area is not active on the basis of an absence of Holocene scarps along the mapped faults. However,

11-6
the seismic record and the uncertainties in the geological interpretations may allow a very low order of seismicity to be assigned to this area.

Hinsdale Fault

Seismic profiles run by oil companies show that the Hinsdale fault

cuts the basement rocks. The fault is not as well understood as the Weldon. It is interpreted to be about 32 miles (51 km) long,

extending somewhat farther to the northeast and southwest than

previously mapped.

There was no surface evidence of fault activity that could be

recognized either along the portion of the Hinsdale fault that had

been mapped or along the extended portion to the northeast that is suggested by the lineup of four historic earthquakes.

This fault was interpreted to be inactive by geological criteria. However, the alignment of epicenters observed suggests the

possibility of a low order of seismicity .[ 1

]

Tiger Butte Fault

Tiger Butte is a complicated structural feature of small size, a hill about 1 mile (0.62 km) across with steeply dipping beds. It occurs, anomalously, in an area where there is no other structural

disturbance. A careful study was made of this feature.* Though

there are some uncertainties in proving the conclusion reported, it

is believed that the structure resulted from ice thrusting. No evidence was found for deep-seated faults and the structure is considered to be an unlikely source of earthquake.

It has been noted that there is no evidence of seismicity at Tiger Butte.

No evidence of active fault scarps was found. The structure

was classified as probably nontectonic in origin. Because there

are uncertainties in justifying such a view, in such a small

structure, no earthquakes are likely to occur that are greater than

Richter magnitude 5.5 to 6.0.

Linears

Linears seen in SLAR imagery were reconnoitered by air and on the

ground. One notable linear followed the trend of Bear Creek and

projected to the Fort Peck Dam. Its location is shown in Figure 3. This feature was carefully examined on the ground but no ground

breakage and no displacement of strata was found anywhere. This

linear is nonfault in origin and is inactive as a structural

j feature

.

|j Selection of Design Earthquakes

Attenuation from the Rocky Mountains

j! Figure 2 shows the belt of earthquakes in the Rocky Mountains of

[Western Montana with a center at Helena. A maximum credible

Unpublished report: Tiger Butte Geologic Report , by

Kenneth Schulte (1975T!

11-7
earthquake of body-wave magnitude M^* = 7.5 at a distance of 250

miles (400 km) from Fort Peck was assigned to this area. The bedrock ground motions attenuated to Fort Peck Dam provided a maximum

acceleration of 0.001 g, a value too small for consideration.

Maximum Credible Earthquakes in the Fort Peck Dam

The geological evidence generally is that the major faults in the

Fort Peck area are not active though there are some questions in

the Brockton-Froid area that are unresolved. The seismological

evidence suggests that there is low seismic order activity along

the projected Hinsdale trend and there is questionable low order

activity along the We ldon-Brockton-Froid trend. Both fault trends

are about the same distance from Fort Peck Dam. Tiger Butte is believed to be inactive from both a geological and seismological

standpoint. However, since there are some uncertainties in the

geological interpretation, and the structure is small, a small

possible earthquake hazard was assigned to it.

Floating earthquakes are those that are assumed possible anywhere in an area. All active faults may not be found and the

floating earthquake provides for possible oversight. It can be

assigned a magnitude based on the area's seismic history. The

Fort Peck area has such a low order of seismic activity that the

investigation turned up nothing in the vicinity of the dam that

would even remotely be considered a geological hazard. [1]

Thus, chiefly on the basis of the seismological evidence,

maximum credible earthquakes (the greatest that can reasonably be

expected to occur) in the Fort Peck area were assigned as follows:

Distance

from Maximum

Dam Credible

Fault Zone miles km Earthquake

Hinsdale and

Weldon-Brockton-Froid 30 48 M

fe = 5.8

Tiger Butte 10 48 Mb

= 5.5

Selected Design Earthquakes

Time histories for the design earthquakes were produced by rescalin

an appropriate existing earthquake record. On the basis of comparability for the sites involved, the Helena record of 1935 was chosen

.

* The conventional equations that are used to calculate body- and

surface-wave magnitude M are of the form

M = B + C log x + log A/T

where

B and C = constants

x = horizontal distance from epicenter to seismograph

station

A = displacement amplitude of the body of surface wave

T = period of wave

11-8
The most conservative assumption was for a magnitude 5.5 event

at Tiger Butte, as no activity has been proven on this structure.

Maximum credible earthquakes, peak accelerations, and records taken

for rescaling were as follows:

Peak

Accelerat ions

at Fort

Peck Dam, Earthquake

Fault Zone g's Record

Hinsdale and

Weldon-Brockton-Froid

Tiger Butte

0.08

0.2

Helena

(1935)

Helena

(1935)

Tiger Butte

Tiger Butte is the most critical event. For this reason, it is the

one that was used in the dynamic analysis. Its duration is 10 sec. Figures 5 and 6 show the time histories and reponse spectra,

respectively, for the Helena record from which this event was

derived

.

Acknowledgments

Grateful appreciation is expressed to those colleagues at the

Office, Chief of Engineers, the Missouri River Division, the Omaha

District, and the Waterways Experiment Station who assisted in this

work. In addition, the support of the Omaha District, who funded

this project, is gratefully acknowledged.

References

1. Marcuson, W. F. , III, and Krinitzsky, E. L. , "Dynamic Analysis

of Fort Peck Dam," Technical Report S-76-1, March 1976, U.S. Army Engineer Waterways Experiment Station, CE , Vicksburg, Miss.

2. Horner, R. B., Stevens, A. E. , and Hasegawa, H. S. , "The Bengough , Saskatchewan, Earthquake of July 26, 1972, " Canadian

Journal of Earth Science , Vol. 10, 1973, pp. 1805-1821

.

3. Colton, R. B., and Bateman, A. F., Jr., "Geologic and Structure Contour Map of the Fort Peck Indian Reservation and

Vicinity, Montana," Miscellaneous Geologic Investigations

Map 1-225, 1956, U.S. Geological Survey, Washington, D.C.

4. Colton, R. B., "Geologic Map of the Brockton Quadrangle,

Roosevelt and Richland Counties, Montana," Miscellaneous

Geological Investigations Map 1-362, U.S. Geological Survey,

Washington, D.C.

5. Colton, R. B., "Geologic Map of the Poplar Quadrangle,

Roosevelt, Richland, and McCone Counties, Montana," Miscellaneous Geologic Investigations Map 1-367, 1963, U.S. Geological

Survey, Washington, D.C.

11-9
ioi" tor' to«' ios' to*' 101*

FIGURE 1. GENERAL STRUCTURAL GEOLOGY IN THE FORT PECK AREA

FIGURE 2. EARTHQUAKES OF INTENSITY MM V OR GREATER FROM 1852 to 1975

11-10
FIGURE 3. EPICENTERS OF MICROEARTHQUAKES RECORDED BY THE LASA ARRAY,

1966-1968 (FROM TELEDYNE-GEOTECH)

11-11
0 6 r

-LS L

i 1 1 1 1 1 I 1 1 1 1 1 ooo 500 boo is oo a oo sen noo boo woo ex moo ss on TIME iSECOtlOSl

FIGURE 5. THE HELENA EARTHQUAKE RECORD

i. j O

5 »'

no I . i_»

Hf LEU* M0HT4I1 EQK OCT 31, I13S

S 90 1 COMPONENT

RELATIVE RESPONSE SPECTRIN

perioO'Sicohos

curves for 0 s. d. mo 3 perceut 01

HELO!«,»C«TM»EC«

OCT 31. H3S SVICOWOHE3IT

RELATIVE RESPWSE SPECTRIN

pbhooisecohosi curves foro s, ii mo.s peacfjit ojmpmocwTi

»ELFJlA.iaiTIIUEC«. Oa 31 U35

RELATIVE RESPOUSE SPECTRUM

PER100ISECOA0S

FIGURE 6. RESPONSE SPECTRA FOR THE HELENA RECORD

11-12
12

THE IZU-OHSHIMA KINKAI EARTHQUAKE OF JANUARY 1978

Minoru Tajima(*)

Hiroshi Sato(*)

Michio Otsuka(**)

Ken Sudo(**)

Katsuhiko Ishibashi(**)

Geographical Survey Instituted)

Building Research Institute (**)

Ministry of Construction

The Izu-Ohshima Kinkai Earthquake of January 1978

[The Izu-Oshima kinkai earthquake caused severe damage to the Izu beninsula.

The main shock occurred at 03^ 24m

January 14, 1978, under the

iicean bottom between the Izu-Oshima island and the Izu peninsula,

phe seismic parameters determined by the Japan Meteorological

kgency are as follows:

Magnitude: 7.0

Location: 139°15* E 34°46' N Depth: 3 km

Study of the first P wave recorded by the WWSSN network has

|:hown that in many stations initial kicks are equivocal, which is

Jme of the characteristic features of this earthquake. This makes

he determination of the fault plane by use of the P wave push-pull

pattern difficult. Thus, the fault generation of the main shocks

:'egan with a slow process; its fault plane was of purely the strike

llilip type.

Alternatively, use was made of the distribution of polarization angle of S wave read-off from the WWSSN LP seismograms

Figure 1). The fault plane solution thus obtained is as follows:

dip direction dip angle

a-plane 0.3° 82.8°

b-plane -91.5° 76.0°

As was suggested by the appearance of initial P wave motions

ecorded at many stations, the above solution shows that two nodal

lanes are of the pure strike-slip type directed almost to E-W

nd N-S directions. This solution is in good harmony with the push

ull distribution obtained from the reading of seismograms from

apanese stations (Figure 2). The slip direction is that of right

ateral fault, which is confirmed by visual dislocation amounting

o 60 cm at Inatori on the Izu peninsula.

hock Locations

he characteristic feature through the entire process is the high

ctivity of foreshocks around the epicentral area of the main shock

hnd large aftershock swarms which were activated at the middle part

f the Izu peninsula unexpectedly distant from epicenter.

The precise determination of epicenters were made from 35 data

lirovided from stations belonging to Japan Meteorological Agency,

Earthquake Research Institute, University of Tokyo, Nagoya University, and the National Research Center for Disaster Prevention,

ipicentral distances for the above stations are less than 150 km.

|he crustal structure assumed for this study is shown in Table 2. jigure 3 and Table 1 show its result. Onset times of P phases are

jxclusively analyzed and the overall error involved in locating

picenters are estimated to be 2-3 Km. As is shown in Table 1,

|he location of epicenter of the main shocks is only slightly diferent from that reported by JMA (No. 4 in Table 1).

he most remarkable fact shown in Table 1 and Figure 3 is that the

aximum aftershock (No. 8 in Table 1) with magnitude of 5.8 has

12-1
occurred at the central part of Izu peninsula, a location which is

unexpectedly distant for the size of the main shock. Moreover, the

direction of epicenter of the maximum aftershock seen from location

of the main shock indicates a contradictory fact: that is, when

speculation is made that rupture has proceeded from the epicenter

of the main shock toward the location of the maximum aftershock,

the direction of the fault is discordant with this solution. This

problem is further discussed in detail later.

Fault Mechanisms

The push-pull distribution is obtained by seismogram reading of the

stations used in locating epicenters plus data recorded at the telemetry network stations belonging to Kyoto, Tohoku, and Nagoya

Universities (Figure 4).

What is to be noted in comparing mechanism diagrams of foreand aftershock is that there is a strong contrast between these

two. While push-pull distribution for foreshock obtained by superposing two large events is very similar to that of the main shock,

that for aftershocks is almost in harmony with that of the IzuHanto Oki Earthquake (M = 6.9) which occurred in 1974.

Rupture Process

Figure 5 shows an amplitude-distribution diagram of the long-period

Love waves drawn from comparison of seismograms equalized at 90° of

the epicentral distance. This pattern is quite concordant with the

pattern suggested by the polarization angle of S waves. The seismic

moment of this earthquake is estimated to be 9.4x10^ dyne cm from

study of the focal mechanisms and amplitude distribution of surface

wave s

.

As was described above, the fault plane solution obtained by use of the polarization angle of long period S waves is contradictory with the speculation that rupture of the main shock has proceeded toward the maximum aftershock.

Figure 6 shows the distribution of aftershocks that occurred

during half a month after the main shock. Several characteristic

features can be pointed out from this figure. Linear trend of

distribution of the aftershocks, which stretches toward the west

starting from the epicenter of the main shock, is bent sharply near Inatori, which is located at the east coast of Izu peninsula,

by the angle of about 23° toward north. This connects the turning

point with the location of the epicenter of the maximum aftershock,

again with a linear trend.

The distribution of aftershocks shown in Figure 6 suggests

the possibility of introducing the hypothesis that the Izu-Oshima

kinkai earthquake actually is made up of two events: one having

its epicenter as usually assigned and fault azimuth of E-W, and the

other having an epicenter somewhere near Inatori and fault proceeding toward Mochikoshi, which is situated 23° northwest from Inatori.

Introduction of the hypothesis of successive occurrence of the two events avoids the necessity of curved growth of the fault. It is not known yet, however, whether the latter earthquake is technically independent of the former and was just triggered by it.

12-2
It is evident that the latter event is closely related to the

>heaval of the ground which has been observed locally in Izu minsula and also the occurrence of Kawazu Earthquake

1 = 5.4 X=138*57'E, 6=34*47N) in August 1976. However, the posbility of the correlation between the occurrence of the former

'en with these anomalous geophysical activities can not be ruled

it.

As for the latter event, the origin time and location is not

liquely determined. Judging from the fact that the aftershock

Jo. 5 in Table 1), which took place immediately after the occur-

-.nce of the main shock, is located at the middle part of the Izu

tninsula, and that the damage at the east coast of the Izu peninila is regarded as being caused not long after the main shock, it

; reasonably inferred that these two events have occurred within

short interval of time.

It is to be noted here that the above inferrence brings about new contradictory problem that entails some further explanation.

iat is, if the assumption is adopted that successive events were

volved, it should be reflected in the determination of fault

ane through the records of S waves and surface waves. In other

rds, the fault-plane solution that is obtained through analysis

these waves should be that of the average of the two events,

ile, as is described before, the solution is quite concordant

th only the former event, it does not fit the latter. This is a zzling problem.

Two possible reasonings are given:

(1) The main shock is related only to the outbreak of the

former event, and the latter fault formed gradually with

earthquake swarms activated by the first event. This

hypothesis reconciles the contradiction that exists

between the distribution of aftershocks and that of the

amplitudes, but is unsuccessful in explaining the result

of the length measurement survey and leveling.

(2) Rise time was very slow and the amount of dislocation was

small for the latter event. According to this reasoning,

the latter event should have emitted an extremely long

period seismic wave. However, it was too small in amplitude to disturb the pattern of amplitude distribution

partly because of the frequency window of the observing

system and partly because of the smallness of the size

of rupture.

As is seen in Figure 8, the expected pattern of ground defortion derived from that fault model shown in Figure 7 is in good

rmony with that found from the length measurements and leveling

ipveys (Figures 9 and 12).

Although authors prefer reasoning (2) for the possible mechasm of the two events, it should be admitted that it is not

mpletely free from weak points. One relates to the question of

ether a slow deformation of the ground can give rise to the severe

mage that was observed on the east coast of the Izu Peninsula,

e answer to this question might be that the damage of this earthake was caused mainly by landslides, landslips, and breakage of

€ ground rather than the seismic wave itself. Does breakage of

e ground tend to be accompanied by the long-period deformation of

e ground?

12-3
In addition to the study of the vibration characteristic, the

dynamic response of the ground is also of essential importance

not only for earthquake engineering but for seismology.

Land Deformation

Precise leveling and length-measurement surveys were carried out after the earthquake by GSI and other institutes. The results of

the surveys have indicated an apparent vertical deformation near

Inatori, which suffered the severest damage from the earthquake

(Figure 9). At Inatori, visual right lateral dislocation amounting to 60 cm was observed on the ground surface (Figure 10), which

is in harmony with the rupture mechanism inferred from study of the

seismic waves. A series of fault traces elongate in the direction

of NW starting from Inatori. Minor right lateral fault traces have

been reported at Neginota, west of Inatori, on the extention of the

linear trend of aftershocks.

Since 1975, abnormal upheaval at the north-eastern part of Izu peninsula has drawn the attention of geophysicists . Inatori is

located just at the edge of the upheaval, while no significant

movement was reported at the center of the activity of the ground

deformation.

The horizontal movement of the ground derived from the length

measurement survey is in harmony with that expected from the

theoretical model (Figures 11 and 12).

According to the result of the length measurement made across

Suruga bay on the other hand, no significant change in length was recognized. This implies that the Izu-Oshima Kinkai earthquake

gave little direct influence on the strain field of Suruga bay,

where a gigantic earthquake is anticipated.

Precursory Phenomena

Abnormal geophysical phenomena such as ground upheaval and high

microseismic activity have drawn the attention of geophysicists since

1975 to the middle part of Izu peninsula. Among numerous data of

geophysical observation made in connection with the earthquake prediction project, some precursory phenomena can be pointed out.

Professor Mogi of ERI has the opinion that stress accumulation

was accelerated by the occurrence of Izu Hanto-Oki earthquake which

took place at the southern tip of the Izu peninsula in 1974.

Apart from the detailed explanations, it seems generally

plausible to say that the anomalous ground upheaval and high microseismicity resulted from this stress-accumulation process.

It was found out later that the b values of the foreshocks were

a significantly small as 0.5-0.6 (Figure 13). JMA issued an earthquake information bulletin at ll*1 on the 14t '1 of January warning

about the possibility that some bigger events might follow. And

the main shock occurred at 12*1 24m

on that day.

The Borehole volumetric strain meters have been installed at Irozaki and Ajiro and continuous observations have been carried out by JMA. Some unusual change commensed one month before the occurrence of the main shock, and it was amplified several days before

(Figure 14).

12-4
Significant variation in radon concentration in the ground

ater before the main shock is also reported at the middle part of

zu peninsula. However, no significant change in the velocity of

rave propagation was observed between Izu-Oshima and Izu peninsula

efore the occurrence of the main shock.

'j . Date Origin ti mc Lonj> . (K) Lnt . (N) Depth M Remarks

1 Jan. 14 09 h 36 m 14 . I s 139° 18'. 2 34° 44'. 7 5 . 8 km 4 . 6 Fore-shock

t 09 45 34 . 0 139 16 . 3 34 45.3 1.1 4 . 9

?

09 47 37.5 139 16 . 8 34 44. 6 7.6 4 . 9

3 12 2 4 38. 3 139 14 . 3 3 4 44. 8 3.4 7 . 0 Main shock

& 13 42 00. 3 138 53.2 34 51.1 1.9 5 . 1 Aftershock

5 Jan

.

15 03 46 21. 8 133 52. 4 34 49.4 3.7 4. 9

t 07 31 47.1 133 52. 4 34 50 . 6 7.1 5. 0 L 07 36 11.0 138 51.9 3 4 48.3 10.5 5 . 4

) Jan

.

16 14 23 50. 8 133 53.5 34 50. 3 15.0 4 . 7

TABLE 1: CHRONOLOGY OF SHOCKS

H (km) Vp(km/sec)

5 5.5

15 6.0

10 6.6

7.7

BLE 2: CRUSTAL STRUCTURE USED IN RELOCATION OF HYPOCENTER

12-5
05

FIGURE 2: DISTRIBUTION OF PUSH-PULL MOTIONS OF P-WAVES IN STATIONS

IN JAPAN.

12-6
Y

7

8

139'

20 km 3A°A5'N

1

1 1 1

1

FIGURE 3: EPICENTER OF REMARKABLE SHOCKS. REDETERMINED

BY AUTHORS.

H hp I

i—

\

*

'r " "

12

[GURE 4: FAULT PLANE SOLUTIONS FROM P-WAVE FIRST MOTIONS FOR FORE- SHOCKS (NOS. 1 AND 2 IN TABLE 1) (A, COMPOSITE) AND FOR

THE LARGEST AFTERSHOCK (NO. 7 IN TABLE 1) (B) OF THE IZU- OSHIMA KINKAI EARTHQUAKE. LOWER FOCAL HEMISPHERES ARE

PROJECTED ON EQUAL AREA NETS. SOLID CIRCLES INDICATE

COMPRESSION AND OPEN CIRCLES, DILATATION.

12-7
N

2 cm

S

i

FIGURE 5: RADIATION PATTERN OF LOVE WAVES. SOLID CURVE DENOTES

THAT CORRESPONDING TO 1.0 x 10 26 DYNE-CM.

FIGURE 6: AFTERSHOCK DISTRIBUTION DURING THE PERIOD JANUARY 18 - JANUARY 31. LARGE SYMBOL DENOTES EPICENTER OF MAIN SHOCK

DETERMINED BY TSUMURA ET AL. (EARTHQUAKE RESEARCH

INSTITUTE, TOKYO UNIVERSITY).

12-8
j •»•« <-{lm) u(k») D (.) '(»«c) v i(k»/.«c> T *P«

i lac 20 10 1.5 1-J 2.5 Bl-l.ter.l

2nd 12 7,1 1-1 2.0 , Unl-l.ter.l

GURE 7: THE 1ST EVENT INITIATES AT THE LOCATION DENOTED BY CLOSED CIRCLE BI-LATERALLY , THAT IS, TO THE EAST, WITH

RUPTURE VELOCITY OF 2 KM/ SEC , DISLOCATION 1M, FAULT LENGTH

7KM AND TO THE WEST, WITH 2 . 5KM/SEC , 1.5M AND 13KM. THE

2ND EVENT INITIATES AT THE WESTERN END OF THE 1ST EVENT

WITH 2KM/SEC, 1M, and 17KM. DIP ANGLE AND DIP DIRECTION

OF THIS SHOCK ARE 83° AND N23°E, RESPECTIVELY.

IGURE 8: VERTICAL DISPLACEMENT FIELD (IN CM) AND HORIZONTAL DISPLACEMENTS OF TRIANGULATION POINTS ASSOCIATED WITH THE FAULT MODEL OF THE IZU-OSHIMA KINKAI EARHTQUAKE . AN ASSUMED FAULT TRACE IS INDICATED BY A SOLID LINE.

12-9
1978.2-1976^-^77.1

FIGURE 9: VERTICAL MOVEMENT IN THE EAST COAST AREA OF THE IZU PENINSULA BY THE LEVELING SURVEY.

12-10
IGURE 10: FAULT TRACES DISCOVERED IN THE EAST COAST AREA OF THE

IZU PENINSULA.

Jfcft cm

I97»r H" . ~~ 12

'IGURE 11: DISTANCE CHANGES OF THE SIDE LENGTHS.

12-11
FIGURE 12: HORIZONTAL MOVEMENT OF TR I ANGULATION POINTS IN THE

IZU PENINSULA.

FIGURE 13: FREQUENCY OF FORESHOCKS PRIOR TO THE EARTHQUAKE AND THE

B VALUE OF THE GUTENBERG-RICHTER FORMULA.

12-12
IURE 14: STRAIN CHANGES BY THE BOREHOLE VOLUMETRIC STRAINMETER

AT AJIRO AND IROZAKI OBSERVED BEFORE THE EARTHQUAKE

(AFTER JMA).

12-13
I\i
13

EMPIRICAL AND ANALYTICAL METHODS OF ESTIMATING

SOIL LIQUEFACTION RISK

R. K. McGuire

U.S. Geological Survey

F. Tatsuoka

University of Tokyo

T. Iwasaki

K. Tokida

Ministry of Construction

Empirical and Analytical Methods of Estimating

Soil Liquefaction Risk

The importance of assessing seismically induced liquefaction hazard

has been recognized in recent years, primarily as a result of the

spectacular liquefaction failures that occurred during the Alaska

and Niigata earthquakes of 1964. An additional stimulus for determining liquefaction hazard is the increasing number of structures

built or proposed on liquefaction-susceptible deposits. These

circumstances have stimulated studies on liquefaction, and many

findings have already been reported in this field [Seed, 1976;

Yoshimi, 1977; Ohashi et al_. , 1978].

In assessing soil liquefaction hazard, the principal methods

used in the past have been deterministic in nature, involving

comparison of soil strength estimated from geotechnical surveys at the site with estimated stresses predicted during a design earthquake. More recently, several probabilistic methods have been

proposed. Youd and Perkins [1977] suggested using empirical observations of liquefaction as a function of earthquake size and

distance, and geologic classifications of soils, to indicate the

liquefaction hazard. Several investigators [Yegian and Whitman,

1977; Ferritto and Forrest, 1977; Crouse e_t a_l. , 1977] have proposed

probabilistic methods that account for uncertainties in the dynamic

soil strengths or in the seismically induced shear stresses. However, to date no direct comparison has been made between an empirical method for assessing liquefaction hazard and an analytical

method that compares estimated soil strengths and induced shear

stresses, for a site where there is available both good geotechnical

data and abundant seismicity data.

Such a comparison is the purpose of this study. The site

studied is near Tokyo Bay at a location where abundant soil samples

have been recovered and analyzed in the laboratory. The historical

seismicity in the area is known in some detail for a period of over

1000 years, and observations of liquefaction as a function of

earthquake size and distance have been made for a large number of

Japanese earthquakes. Thus there is a good opportunity for a direct, detailed comparison of methods of probabilistically assessing seismic liquefaction hazard. This comparison will allow the

advantages and disadvantages of the methods to be understood for

future applications.

Site Conditions and Seismicity

The site of interest in this study is located near Tokyo Bay (see

Figure 1). This location was formerly in Tokyo Bay and was reclaimed about 10 years ago by a pumping method using sandy soils

from surrounding seabottoms. At this site, sandy reclaimed deposits

with a depth of about 6 m overlie deposits of alluvial origins with

a depth of about 4-11 m. Upper parts of the alluvial deposits are sandy and lower parts are clayed. The soil deposits that have high

susceptibility of liquefaction are the reclaimed sandy deposits and

the alluvial sandy deposits.

To define seismicity in the area around Tokyo Bay, the catalog

of earthquakes maintained by the Public Works Research Institute,

Tokyo, was examined. Figure 2 shows the locations of earthquakes

within a 200-km radius of the site of interest (longitude 130°40'E

and latitude 35°34'N) for the years 416 to 1977 with magnitudes of

6.5 or greater. The catalog was judged to be complete for these

magnitudes from the year to 1977; to expand the data base, a radius

of 500 km was considered, and the numbers of events in chosen

13-1
magnitude ranges were determined. The slope of a straight line that

describes the log number versus magnitude relation is -0.9, which

is consistent with the data reported by Ibukiyama e_t al [1972].

For seismic risk analysis made during this study, a single

seismic source, circular in plan and centered at the site of

interest with radius 200 km, was used. This is a reasonable

representation of the historical ' seismicity for the purposes of a preliminary assessment of seismic hazard and will not affect the

comparison among the methods examined. Based on the historical

seismicity over the last 100 years, an activity rate of 11.2

events per year with magnitude 5.0 or greater was used for this

circular source. This is consistent with the data reported by

Ibukiyama et al [1972], A maximum-possible magnitude of 8.6 was used, based on the entire historical record.

Emprical Method

The simplest method of evaluating the probability of liquefaction

at a site is first to relate past observations of liquefaction to the magnitudes and distances of the earthquakes causing the liquefaction. It then is possible to calculate the probability that an earthquake will occur with sufficient magnitude within a critical

distance to induce liquefaction at the site. Such an analysis is

extremely simple and does not account for soil strength at the site

which might be estimated from subsurface borings or samples.

Because relationships between the magnitude of an earthquake and

the distance to the farthest liquefied site are made without regard

to soil strength (i.e., they include observations of liquefaction

at the very weakest sites), the analysis will be conservative for

sites with stronger soils. This empirical method is essentially

that proposed by Youd and Perkins (1977) for calculating "ground

failure opportunity."

A relationship between earthquake magnitude and distance to the farthest liquefied site was available for Japanese earthquakes

[Kuribayashi and Tatsuoka, 1975]. The data are shown in Figure 3,

along with a lower-bound line and a mean line. The mean line was taken to be the best estimate of the radius within which liquefaction will occur for future shocks; this line was manipulated into

the form of a typical attenuation equation:

LI = -4.5 + .87 m - 0.43 In A (1)

where LI is a liquefaction indicator, M is earthquake magnitude,

and A is epicentral distance in kilometers. A standard computer

program [McGuire, 1976] was used to calculate the probability per

year that liquefaction would occur (that is that LI _> 0) at the

site of interest, using the assumptions on seismicity described

above

.

Surface Acceleration Method

The second method used to evaluate liquefaction hazard was based on

the probability distribution of maximum surface acceleration at the

site, which was obtained from a seismic hazard analysis with the

assumption described above. From this distribution of surface

acceleration, distributions of shear stresses in the soil layers were calculated, and these were convolved with the distributions of

soil strengths estimated from soil properties to calculate the probability of liquefaction.

13-2
To esimate the maximum earthquake-induced stresses in soil

layers from the maximum horizontal acceleration at the surface,

the method proposed by Seed and Idress [1971] was used. In this

method, the ratio of the horizontal shear stress t to the vertical

effective stress a v

'in layer i is estimated by

<r/°v 'h = * (rd) iYZ i/(gav

') (2)

where a is the surface acceleration, g is the acceleration of gravity, r^ is a depth-varying function that accounts for elasticity

of the soil, y is the the unit weight of the soil, and is the

depth of layer i. The values used for (r^)- were determined by

analytical methods (McGuire et_ a_l, 1978) anct were consistent with

the values proposed by Seed and Idriss (1971). If the water table

is not at the ground surface, equation (2) must be rewritten to reflect the different unit weights of soil in the saturated and

unsaturated conditions.

To estimate dynamic soil strength, many laboratory tests were

performed on soil samples taken from the site. Details of the

sampling methods and tests are explained in McGuire et al (1978).

Dynamic triaxial tests were made to determine the dynamic shear

strength defined as the stress ratio at 20 loading cycles

which causes 3 percent axial strain in single amplitude. Estimation of the _in_ situ dynamic shear strength was then made by applying a number of constants to RjoQ' t0 correct: f° r laboratory,

loading, and soil disturbance effects.

Further investigations involved methods of estimating the

dynamic shear strength when less extensive geotechnical data are available, as was the case for most of the soil layers at the site.

Consideration of previous work in this field, plus effects observed

in comparing estimated strengths with laboratory-determined

strengths, led to the following formula for the estimated dynamic

shear strength (shear stress divided by effective confining

pressure)

,

Rv

= 0.0047 D * + 0.0029 S (3)

Ei r where S is the percentage content of fine soils and D * is relative

density estimated by

D

r * = 21 VN/av

' + 0.7

(Gibbs and Holtz, 1957). In equation (4), N is the blow count

number for the standard penetration test. Estimation of shear

strength by equation (3) assumes a linear relationship between

strength and estimated density, as do other proposed relationship,

but accounts for the effect of fine soil content. This effect is

shown in figure (4) which compares laboratory shear strength with

estimated relative density. Uncertainty in the soil strength ratio

was characterized by a normal distribution with a standard derivation of 0.057 about the value estimated by equation [3].

To calculate the probability of liquefaction, the density

function F R

for shear stress ratio = (t/o v

')^ in layer i was computed from equation (2) using a risk analysis to obtain the

distribution of ground acceleration (this was considered the only

random variable on the right side of equation (2)). The probability

of no liquefaction in layer i during 1 event can then be expressed

as

13-3
P[no liquefaction in layer i during 1 event]

= $ P[no liq. in layer i|r]f R .(r)dr (5)

where the probability in the integrand is calculated from the

normal distribution with a mean determined from eq. (3) as described above.

It is commonly assumed in seismic risk analyses that earthquakes conform to a Poisson process in time and that successive

events are independent in size and location. Using these assumptions, the probability of no liquefaction in layer i during time

t is

P[no liq. in layer i during time t] = 1 - exp(vt P[no liq., layer i, during 1 event])

= vt P[no liq., layer i, during 1 event] (6)

where v is the rate of earthquake activity and where the approximation is valid for probabilities less than 0.1.

Often several layers in a soil profile are susceptible to liquefaction, and this is the case at the study site. Consequently,

the term "site liquefaction" must be defined in terms of the

stresses and strains which occur in the various layers. The most

conservative approach is to consider liquefaction in one or more

layers to imply site liquefaction. To calculate probabilities con- sistent with this definition, it is assumed that errors in estimated

strengths of layers in the soil profile are uncorrelated ; this

assumption is justified by examination of errors in shear strengths

as a function of depth (McGuire et al . , 1978). The probability of

no site liquefaction is then

P[no site liquefaction | a] = n^Pfno liq. in layer i|a] (7)

The probability of no site liquefaction in time t can be calculated

by an equation analogous to equation (6), substituting the product

indicated in equation (7). A similar procedure is used if site

liquefaction is defined to be the occurrence of liquefaction in two

or more layers, three or more layers, etc.

Results

Using the function described above for the empirical relation

between magnitude and farthest distance to liquefaction, and using

the seismicity assumptions outlined above, the "Empirical Method"

indicates a probability per year of liquefaction of 0.059. This

implies that liquefaction will occur, on the average, once every

17 years. Such a liquefaction risk seems high but is consistent

with the seismicity assumptions for the area. The risk analysis

for acceleration, for instance, indicates that 20 gals or larger

(an acceleration generally large enough to cause liquefaction) will

occur, on the average, once every 17 years, at the site of interest.

For the "Surface Acceleration Method," six layers were used,

each 2 m thick with a mean shear strength determined from soil parameters. Several analyses were done to calculate the probability of

liquefaction, using several definitions of "site liquefaction" in

terms of "layer liquefaction." (This is defined as the occurrence

of a shear stress in a layer greater than the strength.) These

results are presented in Figure 5. It is evident that the most

conservative criterion for estimating liquefaction risk is to define liquefaction as the occurrence of shear stresses greater

13-4
than shear strengths in one or more layers. This may be too conservative a criterion for estimating the effects of liquefaction

damage on structures; many structures may be undamaged if only one

thin subsurface layer reaches a liquefiable state. If the criterion

for liquefaction is that one specific low-strength layer has shear

stresses equal to shear strengths, the risk of this phenomenon

(0.08 per year) is about equal to the risk that two or more layers

liquefy.

Further detailed comparison between the two methods (discussed

I in McGuire e_t al_. , 1978) shows that the empirical method accounts for

the liquefaction hazard produced by low amplitude, long duration

shaking typically observed long distances from large earthquakes,

while the surface acceleration method does not. On the other hand,

the surface acceleration method indicates more liquefaction hazard

| for small earthquakes. When integration is done over all magnitudes,

j the two methods indicate about the same annual risk, although there

are major differences in the methodologies.

It is thought that the estimates of liquefaction probability by

both methods are conservative. This results, in the case of the

empirical method, because soil strength is not accounted for, so that

the analysis is appropriate for sites with only the weakest liquefaction resistance. For both methods the seismicity assumptions

(appear extremely conservative, especially in light of acceleration

and liquefaction experience in the Tokyo area in the past 50 years

(recall that the representation of seismicity for the risk analysis

used an activity rate and upper-bound magnitude based on the entire

historical record; the last 50 years of seismicity indicate a lower

activity rate and a smaller upper-bound). Hence it is not thought

that the surface acceleration method itself is an inherently conservative method. It is concluded, however, that the surface

acceleration method is a viable one for calculating liquefaction

hazard. Further research will be done to account for motion

duration by this method, both in the seismic loadings and in the

soil resistance.

Acknowledgment s

This research was conducted at the Public Research Institute,

Ministry of Construction, Japan, while the first author was staying

at the Institute during the tenure of a Japanese Government Research

Award for Foreign Specialists. He expresses his appreciation to

the Science and Technology Agency for providing him with the opportunity to study in Japan. He also expresses his appreciation to the staff members at the Public Works Research Institute for their

efforts on his behalf as a visiting researcher in the Earthquake

Disaster Prevention Division in the Chiba Branch.

References

j 1. Crouse, C. B., Guzman, R. , and Espana, C., ( 1977). "Probabilistics Evaluation of Liquefaction with the Application to a Site Near a Subduction Zone," Proc . 6th World Conference on Earthquake Engineering , New Delhi, India, Jan., Vol. 6, pp. 34-48.

2. Ferritto, J. M. , and Forrest, J. B. (1977). "Siting Structures

in Seismic Liquefaction Areas," Proc. 9th International Con- ference on Soil Mechanics and Foundation Engineering , Tokyo,

July, Vol. 2, pp. 225-229.

13-5
3. Gibbs, H. J. and Holtz, W. G. (1957). "Research on Determining the Density of Sand by Spoon Penetration Test," Proc . 4th

International Conference on Soil Mechanics and Foundation

Engineering , London, Vol. 1, pp. 35-39.

4. Ibukiyama, S., Juribayashi, E. , and Iwaski, T. ( 1972). Earthquake Resistibility of Submerged Tunnels," Proc. 4th Joint

Meeting, U.S.-Japan Panel on Wind and Seismic Effects , Washington, D.C., May.

5. Kuribayashi, E., and Tatsuoka, F. (1975). "Brief Review of

Liquefaction during Earthquakes in Japan," Soils and Founda- tions , Vol. 15, No. 4, Dec.

6. McGuire, R. K. (1976). "Fortran Computer Program for Seismic

Risk Analysis," U. S. Geol. Survey Open-File Report 76-67,

90 pp.

7. McGuire, R. K. , Tatsuoka, F., Iwasaki, T. , and Tokida, K. (1978). "Probabilistics Procedures for Assessing Soil

Liquefaction Potential," Journal of Research , Vol. 19, Public

Works Research Inst., Ministry of Construction, Tokyo, March.

8. Ohashi, M. , Iwasaki, T. , Tatsuoka, F. (1978). "A Simplified

Procedure for Assessing Seismic Liquefaction of Silty Sand

Deposits," Central American Conference on Earthquake Engineer - ing , San Salvador, El Salvador, Central America.

9. Seed, H . B. (1976). "Evaluation of Soil Liquefaction Effects

on Level Ground during Earthquakes," State-o f-the-Ar t , Preprint of ASCE Annual Convention and Exposition on Liquefaction

Problems in Geotechnical Engineering , Philadelphia.

10. Seed, H. B., and Idriss, I. M. (1971). "A Simplified Procedure

for Evaluating Soil Liquefaction Potential," Journal of the

Soil Mechanics and Foundations Division, ASCE , Vol. 97, No. SMA, Sept., pp. 249-274.

11. Yegian, M. K. , and Whitman, R. V. (1977). "Soil Liquefaction

Analysis Based on Field Observations," Proc. 6th World Confer - ence on Earthquake Engineering , New Delhi, India, Jan., Vol. 7, pp. 25-30.

12. Yoshimi, Y. , Richard, F. E. , Jr., Prakash, S., Barkan, D. D.

,

and Ilyichev, V. A. (1977). "State-of-the-Art Report of Soil

Dynamics and Its Application to Foundation Engineering," Proc.

9th International Conference of Soil Mechanics and Foundation

Engineering , Tokyo, June, Vol. 2, pp. 605-650.

13. Youd, T. L. , and Perkins, D. M. (1977). "Mapping of Liquefaction Potential Using Probability Concepts," Preprint of ASCE

Annual Convention and Exposition, San Francisco, Oct.

13-6
THE SITE WHERE

LIQUEFACTION POTENTIAL WAS

EVALUATED

FIGURE 1: LOCATION OF THE SITE WHERE LIQUEFACTION POTENTIAL

WAS EVALUATED.

137"

6 5< MAGNIIUD£<7 0=O

7 C< MAGNITUDE <7 5=C

7.5< MAGNITUDE <8 0=0

80S MAGNITUDE<9 0=^

138° 139° + + 13

200 kn

?IGURE 2: OBSERVED SEISMICITY WITHIN 200 KM OF SITE DURING

YEARS 416 TO 1977.

13-7
FIGURE 3: RELATIONSHIP BETWEEN FARTHEST DISTANCE TO LIQUEFIED SITES AND MAGNITUDE (AFTER KURIBAYASHI AND TATSUOKA, 1975).

FIGURE 4: COMPARISON OF R^q AND DR

* USING FIGURE 5: PROBABILITY OF SITE LIQUEFACTION

USING VARIOUS DEFINITIONS, FOR SURFACE

DATA OBTAINED IN THIS INVESTIGATION. ACCELERATION METHOD.

13-8
14

A PRACTICAL PROCEDURE FOR ASSESSING EARTHQUAKEINDUCED LIQUEFACTION OF SANDY DEPOSITS

Masamitsu Ohashi

Toshio Iwasaki

Fumio Tatsuoka

Ken-ichi Tokida

Public Works Research Institute

Ministry of Construction
E

1

j

I]
A Practical Procedure for Assessing Earthquake-Induced

Liquefaction of Sandy Deposits

Many structures were damaged due to severe soil liquefaction and

associated ground failures during such major earthquakes as the

Niigata earthquake of 1964 and the Alaska earthquake of 1964.

During these earthquakes, it also became evident that liquefaction

occurred primarily in sandy alluvial deposits and reclaimed lands

(Kuribayashi and Tatsuoka, [1975], Kuribayashi, Iwasaki and

Tatsuoka, [1977], Youd and Hoose [1977]). Recently, the number of

structures constructed or to be constructed in liquefactionsusceptible deposits has been increasing daily. These circumstances have stimulated studies on liquefaction and, at present,

many findings have been reported in this field (Seed [1976], Yoshimi

et a_l [ 1977]). Furthermore, evaluation of liquefaction potential of

soil and, if necessary, countermeasures against liquefaction are gradually becoming common in ordinary aseismic design of structures.

In actual practical procedures for evaluation of liquefaction

potential of sandy soil deposits, blow counts or N-values by standard penetration tests have been often used. Since a standard penetration test is easy to conduct in situ , it is a common engineering

practice in Japan. At present, for most civil engineering designs

this test is generally conducted. Therefore, procedures to estimate

liquefaction potential with use of N-values are convenient to practical engineers. Figure 1 shows several proposals to evaluate the

liquefaction potential of sandy deposits utilizing N-values summarized by (Tsuchida (1975)). Those methods can be called "the critical

-value method." In some of those methods, however, several factors

uch as ground water level and grain size are not taken into account.

This paper presents a simplified design procedure for evaluating

.iquefaction potential summarizing the recent research in this field.

iy this method proposed with the information about (1) ground water

|level, (2) unit density of soil yt, (3) the standard penetration

jfl-values, and (4) mean grain size D^q, the soil liquefaction potential

ban be easily evaluated.

utline of the Method Employed

igure 2 shows schematically the dynamic behavior of a horizontal

urface ground during earthquake motions. An element in the surface

oil deposit is subjected to cyclic shear stress. In this paper,

ynamic loads to soil elements caused by earthquake motions are

efined as

L = (t/o ' ) (1) v max

n which t means the maximum amplitude of shear stress in single

amplitude and a

1 is the effective overburden pressure at the static

:ondition (see Figure 2). Corresponding to L, the resistances of

oil elements to dynamic loads will be defined as R. The method to valuate R will be described later. Then, the ability to resist

he occurrence of liquefaction of the soil elements can be expressed

FL

= R/L (2)

H

is the liquefaction resistibility factor. From the special disribution of the value of F^ in the ground concerned, the liquefacion potential can be evaluated. In this method, the effects of

seepage of excessive pore pressure caused by cyclic earthquakenduced shear stress are not accounted for. This is because when

14-1
these effects are included in the analyses, the procedures will

become too complicated. A simplified procedure that accounts for

the effects of the seepage has to be developed.

Evaluation of Resistance

Figure 3 shows a typical result of undrained cyclic triaxial tests

on sand specimens. These were obtained by special sand samplings.

The methods of sand sampling employed or referred to in this study

are listed in Table 1. These methods can be considered highly

advanced ones in Japan. Triaxial specimens were made by cutting

the frozen samples which were obtained by the sand samplings.

Undrained cyclic triaxial tests were performed on the specimens

isotropical ly consolidated with the initial effective overburden

pressure o v

'- In Figure 3, the number of cyclic loadings N

c where

the double-amplitude axial strain DA becomes 1, 2, 6 or 10 percent

or the effective confining pressure becomes zero are plotted against

the axial deviator stress in single-amplitude divided by two times the initial effective confining pressure o

c

'. The detailed

procedures of sampling, transporation, sample preparation, triaxial

tests or so were described in the previous papers (Ohashi, Iwasaki

and Tatsuoka (1977) and Tatsuoka, Iwasaki, Tokida, Yasuda, Hirose,

Imai and Kon-no (1978)). In this study, the dynamic shear strength

in dynamic triaxial tests is defined as

which is the stress ratio ( a dp/2

1 ) at the number of loading cycles

N

c =20 where the amplitude of axial strain in double-amplitude (DA)

becomes 5 to 6 percent. The values of Rj used in this study were

read from figures such as Figure 3. In general, to obtain a value

of Rj , three or six specimens obtained from one liner were tested.

Of course, the definition of strength as a function of the number

of loading cycles and amplitude of axial strain should depend on

the purpose of the study, for this research, the definition of

Equation (3) was considered adequate. Effects of changes in and

in amplitude of axial strain will be considered in future studies.

It was found that the difference of R^ between DA=5 percent and for

6 percent is quite small (3 percent of most) and that R^ for DA=5

percent and Rj for DA=6 percent can be considered to be utilized

for the same analyses. And in all of the tests referred, the

Skempton's B-values were larger than 0.96. Note that except for

the values of isotropical confining pressure a

Q

' and frequencies

of cyclic loading, the dynamic triaxial tests referred in this

study were performed by the almost same method using the almost

similar apparatus. The other specifications employed are listed

in Table 1.

In situ dynamic strength R can be evaluated from R after several corrections as

R = C

1 'C

2 *C3 'C 4

«C

5 «RS

, (4)

Those correction factors to are to take into account the

random variation in the time history of L, the in situ anisotropic

stress condition, the sample disturbance, the densif ication of

samples during sampling and handling, and the effects of the twodimensional cyclic loading, respectively. These are described in

detail in the previous papers (Ohashi, Iwasaki and Tatsuoka (1977),

McGuire, Tatsuoka, Iwasaki and Tokida (1978)). It can be considered

14-2
that when the in situ coefficient of earth pressure at rest K

Q

is

0.5,

R = Rl (5)

Note that the value of R or Rl depends on the value of DA in

equation (3). For this research, the definition of Equation (4) was considered adequate.

In the simplified procedure for liquefaction potential evaluation it is desirable that the value of R can be estimated easily.

Actually, it is quite expensive to estimate R using Equations (3),

(4) and (5) from the methods described above. Therefore, it has

been expected and tried to evaluate R from the N-values, the unit

density ft, the ground water level and the grain size distribution

of disturbed samples.

In most previous studies, in situ dynamic shear strengths

of sands for liquefaction potential analyses were evaluated using

relative densities which were in turn estimated from measured re- values and effective overburden pressure a

' (Seed and Idriss

(1967)). This method is denoted as the A-method in Figure 4. In the A-method, two different equations are utilized, namely (i) the

relationship among N, a ', relative density Dr and other soil and

ground parameters, and (ii) the relationship among undrained cyclic

strength, a ' , Dr and other soil parameters. The most important

factor which is firstly evaluated in the A-method is relative

density Dr defined as

Dr = emax ~ e x 100 (%) (6)

e - e • max mm

in which e is the estimated in situ void ratio, and e„^„ and e •„

. . . . . max . min

are the maximum and minimum void ratios of the soil, respectively.

In order to estimate in situ dynamic shear strength from values

of relative density which have in turn been estimated from measured

N-values, it is necessary that the following three points be confirmed .

(1) Relative density can be estimated from measured N-values

within the limit of errors allowable for engineering

purposes

.

(2) The standard methods of measuring the minimum and the

maximum densities of sands have been established.

(3) Dynamic shear strength can be estimated from estimated

relative densities within the limit of errors allowable

for engineering purposes.

To confirm item (1), a number of studies have been performed

already. Among them, Meyerhof [1975] proposed the following

empirical equation on the basis of laboratory tests using clean

sands performed by Gibbs and Holtz (1957):

Dr* = 21 ^/(ov

' + 0.7) (7)

in which Dr* is the estimated relative density (as distinguished

from measured relative density Dr by Equation (6)) (see Figure 5).

Recently Japanese Society of Soil Mechanics and Foundation Engineering conducted sand samplings at Niigata using a reformed Bishoptype sand sampler (JSSMFE [1976], Bishop [1948], Hanzawa and Matsuda

[1977] ). The liner had an inside diameter of 5.3 cm and a length of

65 cm. The liner was driven into the sand deposit with the position

14-3

)
of the piston fixed. For sampling, the liner was pulled up into

the chamber filled with air, causing negative pore pressure and

minimizing disturbance to the sample. The methods for estimation

of the maximum and minimum void ratios are shown in Table 1. Void

ratio determined by this procedure will be denoted as ep hereafter.

Figure 6 shows the comparison between measured relative density Dr and estimated relative density Dr* by Equation (7). As Castro

(1975) did, the standard penetration resistances corresponding to undisturbed samples which are referred in this paper were obtained

either in adjacent borings at the same boring and the blowcount

(N-value) was not considered representative of the undisturbed

sample if the soil description or grain size, or both, were different. The sand deposit at Niigata consists of relatively clean

medium grain size sand. While a scatter is observed in these

figures, the correlation between Dr and Dr* is rather good. And

it is evident that Equation (7) becomes quite misleading when the

grain size becomes finer. Figure 7 shows the relationship Dr-Dr*

and D^q from the data of Gibbs and Holtz (1957). Note that Equation

(7) was established only on the basis of the data of air-dry and

moist coarse sand. It can be seen from Figure 7 that the relative

density is underestimated by Equation (7) for fine sand.

Figure 8 shows a summarized relationship between D

r - D * and

D^q for several sites. The methods of sampling and measurements of

e, e and e • „ adopted for these sites are listed in Table 1. At

; max mm r

. . site J, a frozen column method was utilized by Yoshimi, Hatanaka

and Oh-oka [1977] and Hatanaka [1977]. In th is method a large

scaled frozen column of sand, around 5 m in length and around 4 cm

in diameter, was made in the ground and this was pulled out with

a force of 5 tons or more to the ground. Small pieces of frozen

specimen were cut from the large frozen column and their void ratios were measured while being kept frozen. It was confirmed by them from

other basic experiments that void ratios determined by the method

described as above are almost identical to in situ values of void

ratio. It is seen in Figure 8 that there is a general trend showing

D

r - D * decreases with the increase in D^q. Nevertheless, a scatter

shown in Figure 8 is too large for Equation (7) to be used in precisely evaluating relative densities of various sands with a large

range of D^q. Especially for silty sands which include a large

amount of fine soils, it is obvious that relative densities are usually underestimated by Equation (7).

As to item (2), the standard method of measuring the values

of e„ and e • have not been established in Japan so far. Several max min . r methods have been proposed by different researchers, and it is wellknown that the values for silty sands depend on the method employed

significantly. Therefore, it is not possible to determine uniquely

the relative density of silty sands in Japan even when samples are given. It can be seen in Table 1 that the values of e and e • 6 . max . min

for an identical sand (Toyoura Sand) are different among different

methods. This may be one of the reasons which cause a scatter in

the data shown in Figure 8.

Finally, as to item (3), it is necessary that the relationship

between dynamic shear strength and relative density D

r be established

for undisturbed specimens. For reconstituted specimens, Lee and

Fitton (1969) and Seed and Idriss (1971) show that there is a variation in dynamic shear strength for an identical relative density

with the variation in D^q. In this study, available data of undisturbed specimens were utilized to examine whether there is a correlation between dynamic shear strength and D^q as follows.

Figure 9 is the summary of the realtionship between Rj and D

r for

14-4
undisturbed specimens. It is obvious that there is no correlation

among the data. To examine whether the variation in D^q is a main

cause for a large scatter in the data in Figure 9 or not, a parameter DRj was defined as

DR1 = R

l ~ °- 0042 D

r (8)

Equation (8) is based on the following equation:

R = 0.0042 D

r (9)

This was established on the basis of the test results on reconstituted clean sands by various investigators (Ohashi, Iwasaki and

Tatsuoka (1977)). These samples were made by almost the identical

method

.

Figure 10 shows the relationship between and DR^ for the

data shown in Figure 9. It is seen from Figure 20 that there is

jnot a good correlation between DR^ and D^q- From Figures 9 and 10,

flit is obvious that even if relative density could be estimated precisely, it is still difficult to estimate undrained cyclic triaxial

strength from estimated relative density on the basis of the data

shown in these figures.

Figure 11 shows the relationship between fine contents and the

ratio of R^ of undisturbed specimens from site A to R-^ of reconstituted specimens which were made from completely disturbed soils

obtained from the undisturbed specimens. The reconstituted specimens were made by raining de-aired soil into a mold filled with

de-aired water and had equal relative density with that of undisturbed specimens. It is seen from Figure 11 that the differences

of R^ for equal density between undisturbed specimens and reconstituted specimens are considerably large for this case. This

nas been also reported by Ishihara and Tanaka (1974), Seed, Mori

and Chan (1975) and Mulilis, Mori, Seed and Chan (1977). This

neans that there are some other unknown factors causing a scatter

in R, and in DR-^ . Ladd (1974, 1976) and Mulilis, Chan and Seed

(1975) reported that one of these reasons is the fabric of sand.

They showed that dynamic shear strength of reconstituted sands are

greatly affected by the mode of sample preparation. The variation

in dynamic shear strength due to the variation in fabric for the

same density may also be possible in in situ dynamic shear strengths

las shown in Figure 11. Therefore, it is evident that relative

Jensity is not a unique parameter which determines the dynamic shear

[strength of sand.

In summary, it is evident that with present knowledge it Is extremely difficult to estimate undrained cyclic strength R^ of

I various sands by the A-method using standard penetration resistances

j v

' and other factors within the iimit of error allowable for engi-

'.eering purposes.

I The B-method shown in Figure 4 is a more direct method than the

.-method as described below. First, it is logical that in situ

iynamic shear strength is in general related to several factors

Juch as N-values, overburden effective pressure, a ' , lateral earth

ressure, °h' =K

o

av

l » 8 rac*ing properties and strain or stress

iistories. As for parameter, K

Q

, the value of 0.5 may be assumed

jor all deposits examined in this study. To account for the effects

'f oy

' on N-values, Equation (7) was adopted in correlating Rj with

v

', N and gradings of sands. However, it should be noted that in

hese procedures, the value of Dr* does not necessarily mean relaive density but represents an in situ parameter of soil conditions.

14-5
Figure 12 shows the summary of the correlation where the relationship between R2 and D * is presented. Also shown is the line representing the equation

Rj = 0.0042D

r * (10)

This can be derived by substituting = D * into Equation (9). The

data for a ' = 0.5 to 2.5 kg/cn^ from Castro (1975) are included in

Figure 12 in which R2 for = 20, R^Q were converted from Rj and

N

c = 10, R]iq- Obviously, it is seen from Figure 12 that there is no correlation between R^ and D *. Castro (1975) has reported that

the liquefaction of laboratory samples extracted from zones of sand

having a high penetration resistance is little better than that of

samples extracted from zones of low penetration resistance. He suggested that this is due to a loosening of the dense sand during

the sampling process. The data of such dense sands by Castro (1975)

are shown by three points D * of which are larger than 100. However, it can be assumed that the effects of such loosening may

be relatively small for the sand deposits referred in this study.

This is because all the specimens referred in this study were extracted from loose or medium sandy deposits which have D

r* less

than 100. Furthermore, it should be noted that D

r * or N-values

can be largely affected by grain size. This means that a large D

r* or large N-value may be caused by that the zones are gravelly and

may not be caused by high density. Therefore, it can be anticipated that there can be a relatively high correlation among R^,

D * and parameters which represent grading properties of sands.

To find this correlation, a parameter was defined as

DR

: * = Rj = 0.0042 D

r

* (11)

in which R^ is measured dynamic strength by Equation (3) and D

r*

is measured value by Equation (7). Note that Equation (11) is

analogous to Equation (8). Figure 13 shows the relationship

between DR^* and fine material content FC for fine sands D^q of

which are smaller than 0.3 mm (Ohashi, Iwasaki and Tatsuoka (1977)).

It can be seen from Figure 13 that there is a good correlation

between DR^* and FC. The average line can be represented by

D p * = 0.0035 FC (12)

in which FC is fine material content in percentage. From Equations

(11) and (12),

R

1 = 0.0042 D

r

* + 0.0035 FC (13)

For fine sands D^q of which are smaller than 0.3 mm, approximate

values of Rj can be estimated from D

r * and FC using Equation (13).

For a wider range of D^g, FC is not a good parameter for representing grading properties of sands. The mean diameter D^q can be

a more general parameter than FC. Figure 14 shows the summary of

the relationship between DR^* and D^q from the data available at present. It can be seen from Figure 14 that for a wide range of

D^q, there is a high correlation between DR^* an D^q. The average

line drawn in Figure 14 appears to be a reasonable representation

of the relationship between DR^* and D^q. This line was determined

to fit the data as well as possible, but not to be too com- plicated compared with their scatter. Especially for D^q larger

than 0.6 mm, the constant value of DR^* was considered appropriate.

This average line can be represented by

14-6
DRj* = -0.225 log 1Q (D 5Q /0.35) for 0.04<D 5Q

<0.6 mm

nd

DRj* = -0.05 for 0.6 <D5Q

<1.5 mm (14)

'rom Equations (11) and (14)

Den

R, = 0.0042 D * -0.225 log, 0

/ b(j

s for 0.04<D 50 <0.6 mm

nd 0.35 (15)

R

1 = 0.0042 D

r * -0.05 for 0.6 <D

50 <1.5 mm

n which D

r * = 21 ^/(v* + 0.7).

or D^q ranging from 0.04 to 1.5 mm, Equation (15) can be used to stimate approximate dynamic shear strength R^ using D

r * and D^g.

t can be noted that for the same value of D

r *, Rj increases with

he decrease in D^n in Equation (15). This means that if Equation

10) is used to estimate R, from D *, R^ can be underestimated for

iner sands. Therefore, it can be pointed out from the facts shown

n the above that if liquefaction potential is estimated directly

:'rom N-values without taking into account grading properties of a land, liquefaction potential can be overestimated for finer sands.

Equation (15), which can be considered to be one of the best ones mich fit the data available at present, has an advantage over the

-method as follows. In Equation (15), factors affecting undrained

yclic strength such as fabric of sands, static and dynamic stresstrain time histories can be considered to have been taken into ccount in a simple manner. This is because these factors also

iffect standard penetration resistances in the similar manner to indrained cyclic strength (Seed (1976)). This makes the B-method

lonsiderably simpler than the A-method.

valuation of L

he dynamic load L defined by Equation (1) can be estimated from

he following equation;

v This is based on the simplified procedure for liquefaction

'Otential evaluation by Seed and Idriss (1971). In Equation (16),

l g = <* g /g in which <* s means the design maximum acceleration on the

;round surface in gals and g is the acceleration of gravity (980

;als), av

is the total overburden pressure, a

1 is the effective

verburden pressure and means the reduction factor for dynamic

hear stress accounting tor the elastic performance of the ground,

'he most important part in evaluating L is to estimate K

g in

quation (16). In general, K

g can be considered to be affected

?y the magnitude, the frequency characteristics of the input motion

itt base rock, the response characteristics of the surface soil

eposit. Furthermore, the duration of earthquake motion affects

he liquefaction potential. Nevertheless, it is rather difficult

o evaluate or to establish "the design earthquake motion" for each

f numerous construction sites of bridges. On the other hand, in

he Specifications for Earthquake Resistant Design of Highway

ridges , Japan Road Association 1972, it is specified that the

lagnitude of earthquake motion is represented by the seismic coeffi

ient K^. The aseismic design of highway bridges are performed

y the static seismic coefficient method. Therefore, it can be

onsidered reasonable to introduce such a parameter as into

he evaluation of earthquake— induced liquefaction potential for the

seismic design of highway bridges. Although additional research

14-7
is necessary, the value of K g which is used in Equation (16) can

be proposed tentatively as

Ks = |Kh

= fvi.v2 .v3 .K0 (17)

in which the parmeters , v^*V2*v>3 and Kq are the ones specified

in the Section 4:2 "Design Seismic Coefficient" in the Spec if ications for Earthquake Resistant Design of Highway Bridges . is

the design seismic coefficient for bridges, vj is the zone factor,

V2 is the ground condition factor, V3 is the importance factor and K

Q

is the standard design seismic coefficient for bridges (K

q =0.2).

The value of K2 is assumed to be 75 percent of which is the

seismic coefficient of a bridge, in viewing the results of underground strong-motion observation.

An alternative method to estimate K

g besides Equation (17) can

be proposed as

Ks =!i= 18.4 x 10°- 302Mx

A "°' 80

°/g (18)

g

in which M is the Richter's magnitude of the design earthquake and

A is the corresponding epicentral distance of km. Equation (18) can be utilized when the values of M and A can be specified for the

construction site concerned. Equation (18) was derived from the

statistical analysis of the strong motion records measured in Japan

(Ohashi, Iwasaki, Wakabayashi and Tokida (1977)).

It is to be noted that the value of a /

a

v

1 affects L considerably as seen in Equation (16). Figure 15 shows the effects of the

ground water level on the value of av /av

'. It can be seen in

Figure 15 that the value of av/o

v

' decreases with lowering the

ground water level and in turn the value of L decreases. This means

that to lower the ground water level is quite an effective way to reduce liquefaction potential. It is also noted that the value

of yt affects considerably ov /av

'. It was found that the value of

t^j is considerably affected both by the frequency characteristics

of input motions and by the rigidity of subsoils. Figures 16 and

17 show the typical results of earthquake response analyses using

several earthquake records for an alluvial deposit, and soft alluvial deposit, respectively. These analyses were performed by the

computer program "SHAKE" by Seed, et_ al_. (1971). It can be seen in those figures that the value of r. is, in general, smaller for

the input earthquake motion with smaller predominant period T

g and

for the subsoils with large natural frequency Tg. It can be also

expected that T

g is correlated with the magnitude of earthquake M,

the epicentral distance A, or so. As a simple correlation among

r d> ^e ^ or M ' ^ or so ^ s not estat> lished yet, it was con- sidered reasonable to estimate r^ for the simplified method from

the average relationship as

Xd = 1-0.015Z (Z:depth in meters) (19)

In some cases, there exists extremely soft deposit of the thickness

of 2-4 m whose unit density is around 1.2-1.3 t/m above the liquefaction susceptible sand layer in river beds. In those cases, the

value of av

/a

v

' at the top of the sand layer becomes extremely high.

As it can be expected that the def ormability of those extremely soft

clay layers are quite large and therefore the value of r^ can be

considered very low. Although additional research necessary, it can

be considered reasonable to estimate L assuming that those extremely

soft clay layers do not contribute to both L and R when r.^ is

estimated from Equation (19).

14-8
Some Case Studies

The liquefaction potentials of several sites in Niigata City were

analysed by the method proposed in this paper. In the procedures,

it was assumed that K

g = 0.17 and = 1-0.015Z. This is because

the maximum acceleration of the recorded earthquake motion at the

ground surface at Niigata was around 170 gals and it was found from

response analyses of sand deposits that = 1-0.015Z is reasonable

for the sandy deposits in Niigata City. Figures 18 and 19 show the

variation of F^ with depth for the sites where no liquefaction phenomena were observed. It can be seen in these figures that the

values of Ft are larger than 1.0. Figures 20 and 21 show the variation of F^ with depth for the sites where severe liquefaction pheno- mean were observed. It can be seen in those figures that F^ is, in

general, less than 1.0 for the liquefied zones. The ranges of

liquefied zones were estimated from the damage to the pile of a bridge (in the case of Figure 20) or from the comparison of the

grading of the spouted sand with those of the soil deposits (in

the case of Figure 21). Figure 22 summarizes the results of the

analyses of the sites in Niigata city in the F^ <f Z relationship.

The ranges of liquefied zones were estimated from several observations as described above. It can be seen in Figure 22 that the

most data points where the values of F^ less than 1.0 are those

of liquefied zones and that the most dat'a points where the values

of F^ were are larger than 1.0 are those of non-liquefied zones.

This tendency is especially noted for a depth shallower than 10

m. From those facts, it can be concluded that, while some modifications are necessary, the simplified method proposed herein can

be used for evaluating approximate liquefaction potential.

A Method for Evaluating F^

It is obvious that the damage of foundations due to soil liquefaction is considerably affected by the severity of liquefaction phenomena at each zone and the depth and the width of liquefie d zones.

For this purpose, the damage to foundation structures was compared

with the distribution of the estimated values of F^ in the case of

the Niigata Earthquake of 1964. It was found that the severe damage

to foundation structures was found only when the value of F^ is

considerably less than 1.0, say 0.6, and the zone where F^ is considerably less than 1.0 is not thin, say thicker than 5 m. From

those observations, the following method can be proposed tentatively

for the aseismic calculation of lateral behavior of pile foundations

of bridges:

(20)

;in which q^ and are the lateral bearing capacity and the lateral

aspring constant, respectively, which are estimated when neglecting

| liquefaction phenomena. q^ ^ and are the lateral bearing capacity and the lateral spring constant accounting for liquefaction

phenomena. C is a correction factor accounting for liquefaction

phenomena, and is assumed to be given as follows:

C and C

= 1.0 for F L

2. 1 .0,

= 2/3 for 1.0 ^ F L

> 0.8,

= 1/3 for 0.8 Z?L

> 0.6,

= 0.0 for FL ± 0.6

(21)

14-9
Acknowledgments

The principal part of this research project was concluded at the

Public Works Research Institute, the Ministry of Construction,

while the third author was a staff member at the Ground Vibration

Section of the Institute. Extensive in situ soil surveys and

dynamic shear tests have been conducted by the Kanto Regional Construction Bureau of Ministry of Construction, Honshu-Shikoku Bridge

Authority and Yokohama City. The authors express their cordial appreciation to the staff members concerned.

References

1. Bishop, A. W., (1948). "A New Sampling Tool for Use in Cohesionless Sand below Ground Water Level," Geotechnique , Vol. 11, pp. 125-131.

2. BRI, "Report on Damage to Buildings during Niigata Earthquake,"

The Building Research Institute, Ministry of Construction,

No. 42, 1965.

3. Castro, G. (1975). "Liquefaction and Cyclic Mobility of Saturated Sands," Journal of the Geotechnical Engineering Division , ASCE, Vol. 101, No. GT6, pp. 551-569.

4. Gibbs, H. J. and Holtz, W. G. (1957). "Research on Determining

the Density of Sand by Spoon Penetration Test," Proc. 4th

International Conference on Soil Mechanics and Foundation

Engineering , London, Vol. 1, pp. 35-39.

5. Hanzawa, H. , and Matsuda, E. (1977). "Density of Alluvial

Sand Deposits Obtained from Sand Sampling," Proc. of Special - ity Session No. 2,, 9th ICSMFE , Tokyo, pp. 7-14.

6. Hatanaka, M. (1977). "Fundamental Studies on Undisturbed

Sampling of Saturated Sands by Freezing", A Thesis for the

Degree of Doctor of Eng., Tokyo, Institute of Technology.

7. Ishihara, K. , and Tanaka, Y. (1974). "Liquefaction of Undisturbed Sand includng Fine Content," Proc. 9th Annual Meeting

of JSSMFE , pp. 379-382 (in Japanese).

8. Ishihara, K. (1977). "Simple Method of Analysis for Liquefaction of Sand Deposits during Earthquakes," Soils and Founda- tions , Vol. 17, No. 3, pp. 1-18.

9. Ishihara, K. , and Silve, M. L. (1977). "Lager Diameter Sand

Sampling to Provide Specimens for Liquefaction Testing, " Proc . Speciality Session, No. 2, 9th ICSMFE , Tokyo, pp. 1-8.

10. Ishizawa, M. , Nakagawa, S, and Kurohara, I. (1977). "Liquefaction Test of Undisturbed Samples Containing Fine Content,"

Proc. the 12th Annual Meeting of JSSMFE , pp. 397-400 (in

Japanese)

.

11. Japanese Society of Soil Mechanics and Foundation Engineering

(1976). "Report on Earthquake Damages of Subground Streets

and Structures," March (in Japanese).

12. Kolbuszewski, J. J. (1948) "An Experimental Study of the Maxi- mum and Minimum Porosities of Sand," Proc. 2nd ICSMFE , Vol. lib, pp. 158-165.

14-10
Kuribayashi, E. , and Tatsuoka, F. (1975). "Brief Review of

Liquefaction during Earthquakes in Japan," Soils and Founda- tions , Vol. 15, No. 4, Dec.

Kuribayashi, E., Iwasaki, T. , and Tatsuoka, F. (1977). "A History of Soil Liquefaction in Japan," Proc . 6th World Con- ference on Earthquake Engineering , New Delhi, India, Jan.,

Vol. 7, pp. 31-36.

Ladd, R. S. (1974). "Specimen Preparation and Liquefaction of

Sands," Journal of GT Div., ASCE , Vol. 100, No. GT10, pp. 1180-

1184.

Ladd, R. S. (1976). "Specimen Preparation and Cyclic Stability

of Sands," Pre-print of ASCE Annual Convention and Exposition

on "Liquefaction Problems in Geotechnical Engineering,"

pp. 199-226.

Lee, K. L. , and Fitton, J. A. (1968). "Factors Affecting Cyclic

Loading Strength of Soil," Vibration Effects of Earthquakes on Soils and Foundations , ASTM, STP450, pp. 71-95.

Meyerhof, G. G. (1957). "Discussion of Session 1," Proc. 4th

International Conference on Soil Mechanics and Foundation

Engineering , London, Vol. 3,

Marcuson III, W. F., Cooper, S. S., and Bieganousky, W. A. (1977) . "Laboratory Sampling Study Conducted on Fine Sands,"

Proc. of Specialty Session, No. 2, 9th ICSMFE , Tokyo, pp. 12-22.

McGuire, R. K. , Tatsuoka, F. , Iwasaki, T. , and Tokida, K. (1978) . "Probabilistic Procedures for Assessing Soil Liquefaction Potential," Journal of Research , Public Works Research

Institute, Vol. 19, March, 1978.

Mulilis, J. P., Chan, C. K. , and Seed, H. B. (1975). "The

Effects of Method of Sample Preparation on the Cyclic StressStrain Behavior of Sands," EERC-Report No. EERC 75-18, College

of Engineering, University of California, Berkeley.

Mulilis, J. P., Mori, K. , Seed, H. B., and Chan, C. K. (1977).

"Resistance to Liquefaction due to Sustained Pressure," Journal

of GT Div . , ASCE, Vol. 13, No. GT7, July, pp. 793-797.

Ogura, K., Imai, T. , and Suzuki, K. (1978). "Twist-type Sand

Sample," Proc. 13th Annual Meeting of JSSMFE (in Japanese).

Ohashi, M. , Iwasaki, T. , and Tatsuoka, F. ( 1977). "Studies on

Soil Liquefaction Related to Earthquake Resistant Design on Structures," 9th Joint Meeting, U.S. -Japan Panel on Wind and

Seismic Effects , U.J.N.R. May.

Ohashi, M. , Iwasaki, T. , Wakabayashi, S., and Tokida, K. (1977).

"Statistical Analysis of Strong-Motion Acceleration Records,"

9th Joint Meeting, U.S. -Japan Panel on Wind and Seismic Effects , U.J.N.R. May.

PWRI, "Report on Niigata Earthquake," Report of the Public

Work Research Institute, Ministry of Construction, Vol. 125,

June, 1965.

14-11
27. Saito, A. (1977). "Characteristics of Penetration Resistance

of a Reclaimed Sandy Deposit and their Changes through Vibratory Compaction," Soils and Foundations , Proc. of JSSMFE,

Vol. 17, No. 4, pp.

28. Seed, H. B. , and Idriss, I.M. (1967). "Analysis of Soil Liquefaction: Niigata Earthquake," Journal of SME Div., ASCE,

Vol. 93, No. SM3, pp. 83-108.

29. Seed, H. B. , Mori, K. , and Chan, C. K. ( 1975). "Influence of

Seismic History on Liquefaction of Sands," Journal of GT Div. , ASCE , Vol. 102, No. GT4 , pp. 257-270.

30. Seed, H. B. (1976), "Evaluation of Soil Liquefaction Effects

on Level Ground during Earthquakes," State-o f-the-Art Report,

Preprint of ASCE Annual Convention and Exposition on Liquefaction Problems in Geotechnical Engineering, Philadelphia.

31. Seed, H. B. , and Idriss, I. M. (1971). "A Simplified Procedure

for Evaluating Soil Liquefaction Potential," Journal of the

Soil Mechanics and Foundations Division, ASCE , Vol. 97, No. SM9, Sept., pp. 249-274.

32. Tanimoto, K. , and Iwasaki, T. (1975). "Method of Measurement

of Minimum Density of Sand," Proc. 10th Symposium on Soil

Mechanics and Foundation Engineering, JSSMFE, pp. 11-14.

33. Tsuchida, H. (1975). "Present State and Problems in Earthquake

Resistant Design of Grounds against Liquefaction," Tsuchi-tokiso, Proc. JSSMFE , Vol. 13, No. 6, pp. 5-10.

34. Yoshimi, Y. , and Tohno, I. (1972). "Statistical Significance

of the Relative Density," ASTM, Special Technical Publication,

No. 523.

35. Yoshimi, Y. , Hatanaka, M. , and Oh-oka, H. (1977). "A Simple

Method for Undisturbed Sand Sampling by Freezing," Proc . of

Specialty Session 2, ICSMFE , Tokyo, pp. 23-28.

36. Yoshimi, Y. , Richard, F. E., Jr., Prakash, S., Barkan, D. D. , and Ilyichev, V. A. (1977). "State-of-the-Art Report of Soil

Dynamics and Its Application to Foundation Engineering," Proc . 9th International Conference on Soil Mechanics and Foundation

Engineering , Tokyo, June, Vol. 2, pp. 605-650.

37. Yond, T. L. , and Hoose, S. N. (1977). "Liquefaction Susceptibility and Geologic Setting," Proc. 6th World Conference on

Ear t hguaki' Eni'.i neer i , New Delhi, India, Jan., Vol. 6, pp. 37-42.

14-12
i

5

I

i

i

i J

iii

I :5

!l

II

II

1

1 s

li

is

I

1

f

1

1

f

1

I

I

23

1

I •£

lis

111

.li

!a

ss

a s

x"

s

x"

1 1X

i

1 n

:

,

\^

i o °

2

I

. s s T 5

: i

: 1 Ji

R i>l

• d ii

'1

5 o li.

II!

IHi. i

2

•5

1

fa

3

1 Ui

i

2 s2

If: ill

i

2 ti

i

>! III! iii

Vibrating

no

Id

ualng

air-dry

sand

on

.haklng

table

with

1

kg/c»2

being

applied

on

the

top

(0.61) if!

ill

i

° s s l?i

iH

i

!!

I!

Il e

,:

estimated

average

void

ratio

in

the

liner

fl J m;

c

o J 0

I

!

I itii

n

ill

ill

S i 1 Hi

i

2

1

fi

i!

s

2 2

t

I

i.i

in,

ill!

§

I

1

1

fi i ] i ! 1 ! ! I I !

s

!

3 * § Hi

|s

ail)

in,

frill

I! .

iiiii

IH

a!

li

ail Site

F

(In

Tokyo,

along

Tokyo

Bay)

Alluvial

is

il!,

III! Site

H

(Kavaglshcho,

Nligata

City)

Alluvial

and

S

I

ai Site

J

(In

Yokohama,

along

Tokyo

Boy)

Reclaimed

i 2

14-13
JSifBI 83BJJT1S jo Bssropiqj,

o S b b=b

-+

-V

o

z

<

CO

coU

OSH

Ow

po

z

<X

o-X

H

os os <

w aH fa O

O Z

w CO OS g H O CO M

H O i—I H

Z 2 H <

fa Z

W 5h O Q

H

fa

-~- o

— K1 _ r- r-- r--

cr> ai ci

^ CSJ " 01

o

|UJ) Mid 3Q

Buil

on, a

UJ

cr Poi a Cj. c O co

a lO CD O o CL o

LO a cr> cn cn

ode "o

ion

cX> "OO

o

Q_ o =3 a

cj

I/) a> Co

CL o

c o c a E o

nim< cn

u~> Ins u

a>

CJi —

>

1

zu

1 0) O- OJ o x: o O O

o a,

E o v> III

e

gjg CD

<U (XI _

<CM

H

Z

w

H

O

faZo1—1 HU

<

faw

X

CDOSo

faCOWO U1 X r- < Oi >

z <

J Q < 1-1 U X

H CJ H X 1-1 CO

OS H

CJ - O UJ Q

W

OSX

14-14
1% .

Site G

Oc'= Ov = 1.0 kg /cm2

Silty Fine Sond

Average D50 =0 I6 mm 12 5 10 20 50 100 200

FIGURE 3. TYPICAL TEST RESULT OF DYNAMIC TRIAXIAL TEST ON

UNDISTURBED SANDY SPECIMENS FOR SITE G

A -Met hod

N- value 07

JCo*_

qra d inq * fobnc "

static and dynamic

stress strain time history*

,.e

* ; in general not taken into account

B- Method

N - volue <v

Ko

grading

-emr .100%

Trioiiol Strength

R« - J

Triaxial Strength - Ri -

In Situ Strength

FIGURE 4: COMPARISON BETWEEN A-METHOD AND B-METHOD

Air-Dry and Moist Coarse Sand

0 Air - Dry . 0"V= 0 kg/cm 2

• Moist. 0",'= Okg /cm 2

x Air-Dry & Air -Dry » Moist

.

D Air — Dry

Moist .

0V = 0.67 kg/cm2

0"»'= I 73 kg /cm2

Ov' = 1.73 kg/cm2 0»'=2 93 kg/cm2

Cv = 2 93 kg/cm2

J

/* of (Dr-D*

1

0" of (Dr-D*

l = -2.3

1= 9.3 FIGURE 6: D

r AND D * RELATION BY THE

DATA OF JSSMFE (1976)

(after Gibbs and Holtz ( 1957))

Figure 5: D AND D * RELATION OF AIR-DRY

r r AND MOIST COARSE SAND

14-15
After Gibbs and Holtz (1957)

[ Saturated

I Fine Sand

.-Saturated / Coarse Sond I

' ^

"—Air-Dry Sond

Alt- Dry M-CT Moisi Coarse

Fin Sond Sond

[ Note, ^i; mean value, a standard deviation 1

01 02 05

0 50 (mm)

Underwater Sampling

j w * i \

3 V • • " \

Site

° C (Ishihara(l976, 1977)1 + 0 (Ishizaoa et oe(l977l) H (JSSMFEII976)) A I (Soilo(l977)l

• J (Yoshimi.Hotartaka ond Oh-oko(l977)

and Hatanaka(l977)l

It 9

C

Normally Consolidated

Saturated Reclaimed

ond Alluvial Sandy Deposits

001 002 005 0 1 0 2 0 50 imm i

FIGURE 7: D

R -DR

* AND D

5Q RELATION AFTER FIGURE 8: DR-DR

* AND D 5Q RELATION

GIBBS AND HORTZ (1957)

oo

o

DA Ot

Site (7.) .(kg/cm2

) A 6 a, =02-1 15 B 5

0", = 045-1 75

C 5 15 ! Ishchara ( 1976.1977)1 0 5 05 ( Ishizawa et 08(1977)

& 6 CV = 08- 1 2

-0 3

Underwater Sampling

Site

. A

N c = 20

DA -• 5 or 6 %

•

0 •

. * *

, •»* I *

*

LEGEND

A A

A B

» C

« DG

• • • • -x

"a * *

A A

! A

ft A A A

A A A

A

A

I ' 001 002 005 0.2 0.5

^50

I mm)

FIGURE 9: R (NC=20) AND DR RELATION FIGURE 10: DRj=R )l -0.0042 D

R AND D50 RELATION

14-16
o

E

'IGURE 11:

Sllty Fine Sands

Site A

o o o

fa (DA) = 6% Nc=20

10 20 30 40

Fine Content less than 74 jj (%)

50

8 »X * *A...

LEGEND

COMPARISON OF OF UNDISTURBED

SPECIMENS WITH RECON SITUTED

SPECIMENS FROM SITE A

DA

Site (%) (kg/cm2)

r • A 6 Cv =0.2-1.15 A B 5

0"

v =0 45-1 75 A C 5 1 5 ( Ishihara ( 1976,1977)1 D 5 0-5(Ishizawa eta«(l977))

0 E 6 OV =037-0 77

0 F 6 07=0 5. 10

G 6 07=0 8-12

I + 5 0~v'=0 5~2 5(Castro(l975)f

R, = ^0 715 R

,0

FIGURE 12: R AND D

R

* RELATION

if

40 60

F C I % )

Avero qe

001 002

^fo «A7.m \,

SiteA

. B

C

0

1 E

i F

i G

07 l

1 kg/cm 2

) W = 0 2-1 15

0~»'= 045-1 75

I 5 (Ishihwa(l976.l977ll 05(Ishizowaetaf(l977)

0~»' = 0.37-0.77 07-0 5. 10

: 08-!2

Nc = 20

D50 (mm)

'IGURE 13: DR *=R

t -0.0042 D

R

* AND

F.C. RELATION

FIGURE 14: DRj* AND D

R

* RELATION

1 14-17
2 5

Ql 1 ' 1 1 1 0 0 2 0.4 0.6 0 8 10 hw/Z

FIGURE 15: EFFECTS OF WATER TO o la 1

14-18
C/3 HHOWo>J

(J<H

faO

LOH<

PiOHO

<!

fa2O

t-i HQw

OSO

to MZ <n d9a wHi—I Wo

PL) faQJMO

<cT> H I oHU<

fa53ol-l HUQ

faw

ECDOH

fa

(in)Z qidaQ

14-19
2 e o E o E

o- o

_C in

nz<zw

Ed

MZ

Wda

d

w

a;po nwM

EnWCHI 3oawHH

—5

it s ° E o E I

OS

O

O ' <O

25

<CJ

Pn

S3Ww

COzow

02u

QWI—I

FnWCMI o

5SW

PCH

14-20
Showa Bridge in Ni igata Ci ty ( Bor ing Nol ) Niigata Railway Hospital

x—x D 50 (mm)

0.1 0 2 0 5 1.0

o o (after earthquake) < 0 10 20 30 40 v 50 0 0

20

-i 1 r

Fl

I 0 2 0

k s

-- 0 17

rd M-0 0I5Z

>)< It is estimated from the damage to piles

x x D 50 (mm)

0 1 0 2 0 5 I 0

0 10 20 30 40 50 0 0

Q

10

20

Fl

i o 2 0

ks = 0 17

rd= i -o oi5Z

^ It is reported that the sand at the depth of

10 ^ 14 m were found at thg ground surface

'IGURE 20: RELATIONSHIPS BETWEEN F

L AND Z AT THE LIQUEFIED SITE

FIGURE 21: RELATIONSHIPS BETWEEN FL AND Z AT THE LIQUEFIED SITE

14

14-21
,0 0

Q 10

Fl

I 0 R/L

2.0

• • •

•

I

k 5 = 0 17

0 •

• o

O i % • •» U = 1 -0 015 Z

a 0 O

o o

- •

••

•

># o

• o

• .° < • o

°

• * • .

• •

•

0°

o „Oo

0

-

• o ° < ° o I

- Q o o 0 °

•

* ° ° oc

» °° «b o o

-

8 On

O °J OO o o o o

o

o

• o 3 O Qo C O is o o >

O %'8 lO

) o o 3 ° o o

0o o DO O-

; • Liquefaction

O

o < o o 5> o

00

o _ No — Liqnefaction 0 o o o

° 0 o o

2 o

O 0

CDO ° 0

o

O

o a>° ° o

0 Note) The range of liquefied zone was estimated by the seueral observations

.

FIGURE 22: RELATIONSHIP BETWEEN F L AND LIQUEFACTION PHENOMENON

(THE NIIGATA EARTHQUAKE OF 1964)

14-22
15

WIND- RESISTANT DESIGN OF CABLE-STAYED BRIDGES

IN JAPAN

TadayoshI Okubo

Nobuyuki N. Narita

Masataka M. Katsuragi

Public Works Research Institute

Ministry of Construction
r
Wind-Resistant Design of Cable-Stayed Bridges in Japan

cable-stayed bridge, spanning from about 200 to 400 meters, can constructed even at a site with bad soil conditions. Though the

:chanical principle of the bridge is different from that of the

ispension bridge, both of the bridges are flexible structures,

it with the trend towards lighter and more flexible structures,

oblems of dynamic instability are becoming increasingly prevalent.

The purpose of this paper is to introduce the Japanese approach

solving instability problems. Countermeasures may consist either

removing the cause of the aerodynamic excitation by installing

tachments or of reducing the response to the excitation attachments

of reducing the response to the excitation by increasing the

gidity or damping capacity. In this context field observation

indispensable to supplement the wind-tunnel testing with laminar

ief Description

Suehiro Bridge

The Suehiro Bridge is a three-span cable-stayed bridge with

e center span of 250 m and the side spans of 110 m. The

idge deck carrying four lane tracks is composed of a trapezoidal

no-box girder with cantilevers on both edges. The dead weight

the girder is 9.3 t/m/br. The fundamental natural frequencies

e 0.483 Hz in vertical bending and 1.817 Hz in torsional oscillaon. The construction site is fairly flat and open to the Kii

raits. The prevailing wind direction is nearly perpendicular to

e bridge axis and a strong seasonal wind is expected in winter

igure 2).

Suigo Bridge

The Suigo Bridge is a two-span cable-stayed bridge with a in span of 178.9 m and a side span of 111.6 m. The bridge

ck carrying four-lane highway tracks and sideways of 2.5 m

de on both sides is composed of a trapezoidal mono-box girder

th cantilevers. The dead weight of the girder is 13.4 t/m/br.

e natural frequencies were calculated as 0.448 Hz in vertical

nding and 1.859 Hz in torsional oscillation (see Figure 4).

nd-Resistant Design Procedure

In the preliminary design, three types of deck girders were

oposed for each bridge:

Suehiro Bridge

Trapezoidal mono-box girder with cantilevers

Rectangular twin-box girder with cantilevers IXZJ

Rectangular mono-box girder with cantilevers

Suigo Bridge

Trapezoidal mono-box girder with cantilevers

15

15-1
Rectangular twin-box girder with cantilevers

Rectangular mono-box girder n a

The stability test on these six different configurations of

deck girder showed that the trapezoidal mono-box girder with cantilevers had excellent aerodynamic characteristics: no susceptibility

was observed in torsional flutter at high wind speed, but vertical

bending oscillation was observed at comparatively low wind speeds

(around 15m/s for Suehiro Bridge and around 25 m/s for Suigo

Bridge) when the wind flow approached the models with a large angle

of attack.

This type of oscillation — vortex induced oscillation —

should either be suppressed to a limited amplitude or be eliminated

to avoid discomfort to pedestrains or passengers.

Two types of coutermeasures were proposed: reduce the vibrational response to the excitational by increasing either rigidity

or damping capacity of the girder, and reduce the excitation

itself by installing special attachments. However, the former measure was abandoned for economic and technical reasons.

The original model of Suehiro Bridge showed instability at 20 m/s (a ^ 2 deg). To suppress the vortex-induced oscillation

the handrail openings were closed. But this did not show remarkable improvement. The effect of height of curb (road parapet)

was examined in the second test stage, and it was found that the

thinner curb improved the stability at a = 0 deg. However, it did not work well for larger angle of attack.

Most effective was the installation of a fairing at the top

of the handrail as shown in Figure 3. The sharpness of deck edge,

which is achieved by installation of triangular fairing or by

diminution of curb height, helped the effeciency of the fairing.

The effect of fairing was confirmed not only by a stability

test but by non-steady aerodynamic force measurement. Test result is given in Figure 6.

A series of tests were also carried out for the Suigo Bridge.

Prior to the construction of the bridge, data on wind inclination

was gathered at the construction site. The results of field

observation are shown in Figures 8 to 11, from which it can be

concluded that the wind inclination varies violently with time in

random manner and the average inclination angle decreases rapidly

with the increase of average wind speed as well as the average

time of wind inclination. It was concluded that no vortex-induced

oscillation might occur at the bridge. In Figure 5 the countermeasures for bridge vibration due to wind action are summarized.

Wind Response of Real Bridges

Wind tunnel tests have so far been carried out under laminar

flow, but no attemps have been made on turbulent effects. The

authors will continue their long-term observation after the com- pletion of the bridges to complete this study (see Figure 1).

Figure 7 introduces the instrumentation array at the Suehiro

Bridge, and Figures 12 to 15 show the field observation results.

These will be completed in the near future.

15-2
References

1. Okubo, T. , Narita, N. , and Yokoyama, K. , "Some Approaches for

Improving Wind Stability of Cable Stayed Bridges," Fourth

Int. Conf. on Wind Effects on Buildings and Structures , 1975,

Heathrow, London.

2. Okubo, T. and Narita, N. , "A Comparative Study on Aerodynamic

Forces Acting on Cable Stayed Bridge Girders," U. S. -Japan

Res. Sem. on Wind Effects on Structures , 1974, Kyoto, Japan.

3. Narita, N. , "On the Wind Resistant Design of Cable Stayed

Bridges with Solid Girders," February 1978.

_

Structural effects

static effects

wind force

distribution of wind pressure

total wind loading

stability

lateral buckling

divergence

dynamic effects

forced oscillation

random oscillation

self-excited oscillation

galloping

torsional flutter

bending-torsional flutter

vortex-induced oscillation

Functional effects

wind stream

concentration of wind stream

noise

dynamic problem

vortex-induced oscillation

FIGURE 1. WIND EFFECTS ON BRIDGES

15-3
470 000

110 000

750,, 50 000 , 55 000 50 000 . 50 000 , 50 000 , 50 000 . 50 000 T 55 000 , 50 000 ,,750

250 000 1 1 0 000 -

2 50H

FIGURE 2. GENERAL DRAWING OF SUEHIRO BRIDGE

15-4
15-5
type; original design modified design examples

Ishikari Br.

5

Long's Creeks

Long's Creek Br.

Lillebaeltfg

Lillebaelt Br.

St.Nazairefg

St. Nazaire Br.

\ /

Suehiro Br.

Kamome Br

.

\ I

\ / \ /

£teJilfg(ftli3S)

Aj igawa Br

.

(Proposed)

Eisai Br.

r 7 X?iHljfjj (truss stiffened)

Severn}^ Severn Br.

\ /

1 T

TZZT

Suigo Br.

Lower Yarrafli Lower Yarra Br. ^ T

Arakawa Br

.

(discrete)

Stromsteinflj

Stromstein Br,

Onomichi Br.

Humbert

Humber Br

.

FIGURE 5. COUNTERMEASURE FOR SUPPRESSING

VORTEX-INDUCED OSCILLATION

15-6
5 V (m/«)
apxs ^ub

q 3 3 a

T

w * m

-H

aut

m

apis )(UEq aqg-ta

Mo

0) u et

01

u E 6 0) O c 4J e u "J CD

0) £ c

rH 0 cfl 0J e 0 aj OJ o c

Cfl cfl 4J u a z

cx

H t-i o O H

tJ0 o

toM

ell RVAT > 0) w M a. o CO

0) 3

PQO

w CO

Saw

ft) 1 o E3

RlTTTTl ON

T? tfnQ [&! 1 ft ita <s OR

4- ln/1 En

-r r< to < tjm PiW

Eg c_> s§ < co n

CO • 4 s Z3HZoozowuoH

15-8
western wind

perpend icular

to bridge axis

mum fts*-f> mm.

degree

southern wind

in bridge axis

northern wind

in bridge axis

FIGURE 8. AVERAGED WIND DIRECTION AND VERTICAL

INCIDENTAL ANGLE

+

Its

n s

Vtcc,a a <Dm% = 3 5 0°~10'

FIGURE 9. V, 5 , AND a

a

( 6 = 350-10 DEG.)

15-9
s : aver ag ing time

<•—

.

<u

0)M

0) T>cH

0J rHM at C

OJ "O

rCCO

•H

JJ

to H

H

o

cH

X

: **

*

1 0.0

5 0.0

16 m/S

FIGURE 10. V, a , AND S a

<5:excess time ratio

s:averaging time

1-4.1 7 ffF ffi s# RSJ t m ii s# K o M 1fc FIGURE 11. s AND q

15-10
v'W/Vk

0.30

4-1

01

0.10L

10 20 Vk(m/s)

FIGURE 12. AVERAGE WIND SPEED AND INTENSITY OF TURBULENCE

15-11
((S/D/2* * (1V0)) A1ISN3Q H3M0d

15-12

> o

oo

S> oo

0° „.

l>

-FT c

6

_L_ =y —

•(ml ig -B —

CO CM (UIO) 3UaUI33F"[dSTp ];B0I3I3A

3

0 1 (jEg) U0iqEJ3J300B

XBOJ^aBA JO UOI JET A9p paBpUBlS

15-14
16

AERODYNAMIC STABILITY OF PROPOSED OHIO RIVER.

CABLE-STAYED BRIDGE

Lloyd R. Cayes

Federal Highway Administration

Washington, D.C.
I

1
Aerodynamic Stability of Proposed

Ohio River Cable-Stayed Bridge

I This report presents the results of wind tunnel tests on a model of

j the proposed cable-stayed bridge across the Ohio River at Huntington,

West Virginia, to determine the aerodynamic stability of the

suspended span. There is also a discussion of the background of

this type of section model testing and the procedure used by the

i Federal Highway Administration in carrying out such an investigaI tion. The test was conducted for the West Virginia Department of

Highways. The object of the investigation is the determination of

the bridge dynamic deflection under vortex shedding and the critical

wind velocity for flutter.

Background

1 There are several accounts of suspension bridges that were severely

damaged or collapsed in wind storms during the nineteenth century.

However, most of these failures were of bridges of relatively short

span length. In 1889, the 1,280-foot (390 m) span Niagara Clifton

Bridge failed from wind action, g iving the engineering world its

last severe warning about the problem of aerodynamic stability until

the collapse of the Takoma Narrows Bridge in 1940. During this

period, suspension bridge spans were built increasingly longer and

1 with lighter stiffening struture's. Investigations into the causes

of the Takoma Narrows disaster were immediately undertaken at the

University of Washington [1], where wind tunnel tests on a scale

model reproduced quite faithfully the observed behavior of the

bridge during its brief life. The stability of the proposed design

! for the new bridge was then established by wind tunnel testing.

Further research revealed that a section model representing a limited

length of the bridge and mounted on springs to reproduce to scale the

frequency of a significant mode of motion could be used to determine

the response of the bridge in a normal wind at various vertical angles

of attack.

During the past decade there has been considerable interest in

using wind tunnel techniques for the determination of the aerodynamic

stability derivatives of suspension bridge models. These derivatives

represent purely aerodynamic data — as against overall dynamic

response observations — which are the consequence of just bridge

cross-section geometry.

In earlier wind tunnel tests, bridge-deck section models were elastically suspended, with properly scaled vertical and torsional

' natural frequencies, and observed primarily for their tendencies

j' to flutter at some air speed. The present practice is to extract

] the purely aerodynamic characteristics of the model, then employ

these in the dynamic response calculations relative to full size or prototype bridge stability. This method has the advantage of divorcing

test results from all model structural properties except geometric form.

There are two types of wind effects on structures — static and

dynamic. The static effect is the pressure that the moving air

mass exerts against the structure, and it varies in direct proportion

to the square of the wind velocity. The dynamic effects of wind

are the forces created by flow separation and turbulence of the moving

air as it passes around the structure or by changes in the wind

speed such as gusting. This report will deal only with the dynamic

instability, of which there are two main types. The first is due

to the regular formation of vortices in the wake of the bridge. The

vortices form alternately on the upper and lower lee-side corners

of a body with a pronounced degree of periodicity. This phenomenon

16-1
is known as vortex shedding. The vortices give rise to fluctuating

forces, and the frequency is proportional to the wind speed. If

the vortex shedding is equal to one of the natural frequencies of

the structure, then a resonant condition can occur, if the structural damping is low, resulting in a buildup of vibration amplitude

either in torsion or vertical bending. The amplitudes of vibration

caused by vortex shedding are usually small, although they may cause human discomfort they are not normally catastrophic.

The second type of instability observed on bridges is much more

serious and is referred to as flutter [2, 3] because of the aeroelastic studies on aircraft in the past. Unlike vortex shedding, once flutter is entered into, a higher wind velocity will not cause the

structure to pass beyond the phenomenon and suppress it so that the

structure again becomes stable. In flutter, the higher the velocity

beyond onset of instability, the more violent the phenomenon becomes.

Basic Theory for Model Tests

The susceptibility of the prototype structure to aerodynamic instability is determined by wind tunnel tests. These tests are conducted

on a section model of a bridge and subjected to laminar flow.

Although the natural wind is turbulent there are some substantial

reasons for testing under laminar conditions. The level of turbulence

in the natural wind appears to decrease as wind velocity increases,

and most serious wind problems of bridges occur at high wind velocities.

Laminar testing appears to be more conservative for problems of

instability such as flutter and vortex excitation. Laminar testing

procedures and subsequent data reduction are considerably simpler

than their counterparts in turbulent flow.

Section models, representing only a portion of a long-span bridge

deck, are simple, cheap, and relatively easy to construct and modify.

Their prime basic requirement is geometric similarity to the

prototype.

The vortex shedding response was determined by References 3, 4,

and 5 and is as follows:

Considering purely vertical motion of the section model, the

homogeneous Equation (6) for the self-excited case of flutter is

provided with a special forcing function applicable to the region

of vortex shedding response:

m[h + 2 c, <*> h + 2

2 hi - 1 (-„ h] -- pv

2 (2B [KH

O y) + C, sin u t] Lo s

where

is an aerodynamic damping coefficient for decay

to a "resonant" response amplitude

is an "oscillating lift" coefficient

s is the Strouhal circular frequency

* are experimentally determined where the vortex

excitation frequency coincides with the natural

frequency of the model to produce maximum

response amplitude.

H O'

CLo'

16-2
The equation for the prototype deck station is:

m( x ) tn + 2T s w

s

n + o) g n] i|)<x) = P

s sin ws t

and the maximum half amplitude n is:

imax.

Mais * hs where

Ys PB2 H*

C _ o 2m

o P = pv B C Lq /

S p an t(x) dx = generalized force

amplitude

2 M = mJ' S p an

( x ) dx = generalized mass span

S = structural damping ratio

p = mass density of air

B = deck width

m = section mass per foot of span

v = wind speed

\))(x) for the prototype was computed by the design engineers for the

main span deflected shape with i = 900 feet and obtained:

f

l ip(x) dx = 468.515

o/* >p 2 (x) dx = 348.128

o For a prototype frequency of oscillation N^, the maximum acceleration of the deck section is:

Smax = (2, N

p

)2

Scanlan with others (6) developed the following equations to predict flutter behavior of a rigid section model.

Assuming the typical bridge deck to be symmetrical about the

vertical plane of the roadway centerline, equations of motion for

the sectional structure may be written:

2 (1) m [ft + 2 5 hWhh + h] = L

(2) I [5 + 2s a

(o ai + u

2a J

= M

where m and I are, respectively, mass and mass moment of inertia

per unit span; h and a are vertical and torsional deflections of

the given deck section (assumed uniform over the span); ^ and c a are damping ratios-to-critical in the respective h- and a-degrees

of freedom; and u are the respective associated natural circular frequencies; and L and M are the respective dynamic lift

and moment per unit span.

16-3
For the self-excited (flutter) problem that will be discussed

here, the forms of L and M are intentionally taken as linearized,

homogeneous functions of h, a, and their first derivatives only.

Clearly, much more complex models can be assumed, but observations

of model and full-scale structures in which aerodynamical ly induced

responses at small structural amplitudes are dominated by structuralinertial vibrational characteristics suggest that linear models can quite adequately represent incipient aerodynamic instability.

Thus, L and M take on the simple forms:

(3) L = mtHjfi + H

2 a +H3 «]

(4) M = l[A,n + A

2 « + A3 a]

where and A^ (i = 1, 2, 3) are dimensional aerodynamic derivative

coefficients to be determined by experiment, and the inertial factors m and I have been introduced for convenience.

To render such coefficients nondimensional , it is noted that

they are essentially functions of the dimensionless parameter

K = Bu/U, where B is bridge deck width, w is oscillation circular

frequency, and U is horizontal wind velocity. They are then

replaced by dimensionless coefficients and A^ so that L and

M can be shown to take the forms:

(5) L = — pU

2 (2B) [KH*" £ + ml ^ + K

2 H~]

2

1 U

1 U 3

(6) M =~ PU

2 (2B 2

) [KaT £ + KAo ^ + K

2 A^ a

] 2

1 U

1 U

J where p is air density.

Reference 6 describes test and data reduction methods for

obtaining and A^. Essentially, test observations produce time

histories of h and a; then with all quantities on the left-b.and

sides of equations (1) and (2) known, coefficients and A^ are matched by system identification techniques to the test observations

as functions of K.

The model damping was measured in terms of the logarithmic

decrement ft = 1/n In ~ , in which x and x are double amplitudes

xnn

of oscillation separated by n cycles. For relatively low levels of

damping such as exhibited by bridges, the damping ratios stated as

a percent of critical can be determined by dividing the log decrement by 2 it and multiplying by 100.

Damping can have a significant effect on the overall vibration

of the structure. When the wind blows over a structure, it is inputting energy that is being dissipated by the movement of the

structure. When the wind is removed, the vibration of the structure will decay or damp out as a result of the dissipation of

energy. The rate of decay depends on the magnitude of the damping.

The determination of the actual damping of any long-span capablesuspended structure can only be approximated because of the lack

of data available. The design engineers have suggested a value

of .08 log decrement for the proposed Ohio River Bridge prototype

structure.

16-4
Design and Construction of Model

A selection model of a suspended bridge is a geometrically similar

copy of a typical length of the suspended structure only. Any cables

or hangers are not modeled because their aerodynamic effects are negligible. The section model is, rigidly constructed so that all

of the elastic behavior takes place in the springs from which it is suspended. The vertical and rotational movement of the section

model supported by its springs simulates the vertical bending and

torsional movement of the prototype structure.

The section model to be investigated for this study was designed to a scale -of 1/45 to obtain a representative section of

the bridge. The model is 60 inches (1.524 m) inside to inside of

end plates, representing 225 feet (68.58 m) of the prototype. The

width of the model is 9.6 inches (.244 m) , resulting in an aspect

ratio of 6.25. End plates are employed so that the air flow near

the end of the model is two-dimensional. The model was scaled by

the Froude number criteria for mass property and vertical natural

frequency. Due to practical limitations of simulating the relatively high prototype torsional characteristics, values of polar mass moment of inertia and torsional natural frequency do not achieve Froude scaling. These scaling discrepancies in the

torsional characteristics of the model were adjusted mathematically so that the prototype predicted results are proper.

The model was built from a variety of material to maintain

the proper weight. The bottom and side panels are balsa wood, diaphragms and guard rails are aluminum, deck is plexiglass, and end

plates are of balsa and aluminum sheets glued together. The model

was fabricated by hand and assembled with double-sided tape and

stainless steel machine screws. Aluminum extension brackets were

used to provide crossbars 90 inches (2.286 m) apart for the attachment of the supporting spring connections, placing them completely

out of the wind stream of the 6-foot by 6-foot (1.829 m) nozzle.

The weight and center of gravity of the finished model were adjusted

by adding brass weights to the end plates. The dynamic properties

of the bridge section and deck configurations are shown in Figure 1. Photographs of the model are shown in Figures 8 through 11.

Tests

The tests were conducted in the George S. Vincent Memorial Wind

Tunnel, which is a low-velocity wind tunnel designed particularly

for testing bridge section models. A blower forces air through a diffusing section with a series of stainless steel wire screens into

a cylindrical pressure chamber and. through the nozzle from which the

air is discharged onto the model. The velocity over the cross section of the wind stream is very uniform to a point 6 feet (1.829 m)

beyond the end of the nozzle.

A constant wind velocity, with a range of 0 to 40 feet per

second (17.88 m/sec), is maintained with a micro-control rheostat

in the blower circuit. The wind velocity is computed from the

dynamic pressure head of the wind stream obtained by direct measurement with a pitot-static tube and an inclined manometer.

The nozzle can be rotated about the horizontal axis of the

pressure chamber, allowing tests at various wind angles of attack.

The angle of attack is the vertical angle between the wind velocity

vector and a horizontal plane; a positive angle of attack indicates

an upward wind.

16-5
The section model was suspended in front of the wind tunnel

nozzle by means of four supports with coil springs. The vertical

length of the springs was adjusted to set the proper vertical

frequency of oscilation. The torsional frequency was set by adjusting the spacing between the springs at each end of the model.

Horizontal restraining wires were attached to the end bracked

crossbars at the center of gravity of the model to prohibit longitudinal or lateral movement during the tests. The restraining

wires, which were tensioned with 2-pound (.903 kg) weights, were sufficient length that vertical force components would have no significant effect upon the vertical or torsional motion of the model.

Electrical resistance strain gages were attached to beryllium

copper flexure plates above the coil springs. The strain gages at one end of the model were wired to indicate the vertical component

of oscillation, and those at the other end were wired to indicate

the torsional component. The two components were recorded side by

side on an oscillograph.

The model is tested in smooth air flow at wind angles of attack

varying from +6 degrees to -4 degrees. While maintaining a steady

wind velocity at each of a series of increasing values, the model

is displaced (if necessary) in the respective vertical and torsional

degree of freedom, and the oscillatory dynamic response is recorded.

The damping values of the model setup were determined by displacing, then releasing the model in a no-wind condition and

recording the rate of decay. The log decrement varied slightly from

test to test but always around .01 for both vertical and torsional

motion. Preliminary tests showed the model to be active due to vortex shedding with low levels of model damping. Therefore, it was deemed necessary to test the models at damping values that would more closely correspond to what may be expected for the prototype.

Variable damping was achieved by attaching a small copper plate to each corner of the model and bringing a fixed permanent magnet in

close proximity to the plate. Varying the small air gap between the

magnets and plates gave the desired levels of model damping.

Discussion of Test Results

The different test configurations of the deck with their

dynamic properties are shown in Figure 1.

Figure 2 is a summary of all the model tests that were conducted. A section model designation number was used throughout

the investigation for easy identification and was determined as follows: take as an example model 3-2. 4-. 08; the first number,

3, refers to the prototype deck configuration; the second number,

2.4, refers to the model vertical frequency of oscillation; and

the third number, .08, refers to the approximate log decrement of

model damping at which the test was conducted. The x's in the

columns for wind angles tested in Figure 2 signify tests that were conducted

.

Figures 3, 4, and 5 summarize the results of all the tests conducted. The critical wind velocities for vortex shedding and

flutter, and the amplitudes given in these figures are prototype

values

.

The critical flutter velocities shown in the summary Figures

3, 4, and 5 are for single-degree of freedom torsion and were determined from plots of the aerodynamic coefficients and computations

16-6
of the critical An coefficient. As can be seen, the flutter velocities are all well above the design wind velocity of 125 mph

(201 kph) when any amount of damping is present. Flutter instability is not a threat to this bridge; therefore, no further discussion will be made.

Vortex shedding response in the torsional mode was only

slightly evident in configurations 3 and 3H at the lowest level of

damping, that is, without any added damping only model damping of

approximately .01 log decrement, and occurred at prototype wind

velocities around 75 mph (121 kph). But because the model torsional

frequency was only about one-half of the proper scaled frequency,

this motion, if possible at all with higher prototype damping, could

only occur at much higher wind velocities. Therefore, torsional

vortex shedding excitation is not a threat to this structure.

Vortex shedding response in the bending or vertical mode was evident in all tests of prototype configurations 3, 3H , 2, and 2H,

and in many of the tests of configuration 1.

Configuration 1 represents a construction stage of the bridge

deck system before the asphalt pavement wearing surface and curb

j barriers are installed. This model was tested at three different

vertical frequencies of oscillation because of the uncertainty of

| the actual frequency as the bridge is being erected. The lowest

: equivalent prototype frequency tested was .247 cps. No vortex shedding response was present, apparently because the wind velocity

required to excite vortex motion was so low that not enough energy

was present to cause motion for the levels of damping tested. When

the vertical frequency was increased to .29 cps, vortex response

did occur between 8 (12.9) and 12 mph (19.3 kph) but very weak and

slow to respond. When tested at frequency of .39 cps, approximately

the same as used for 3, 3H, 2, and 2H, it showed no more response

than the previous frequency tested.

Models 2 and 3 were of the same geometric shape, as well as

2H and 3H, the only difference being in the weight of the models

due to the lack of a wearing surface on 2 and 2H. The weight of

the wearing surface was added to 3 and 3H by attaching brass

weights to the end plates inside the model. The "H" designation

j signifies vent holes in the curb barriers.

Configurations 2 and 2H were tested at one level of damping of

approximately .08 log decrement. (See Figure 4.) There is insignificant difference in vortex response between these two tests

indicating that the vented barriers have no effect. Models 2 and

2H are more active than 3 and 3H with the same level of damping.

This indicates that the addition of weight to the bridge deck system,

if the natural frequencies remain constant, tends to decrease vortex

mot ion.

Model 3H was tested at model damping of .01 log decrement and

with added damping to bring the total closer to the estimated protoj; type damping. (See Figure 5.) Model 3 was tested at model damping

as well as three levels of added damping. Figure 6 is a curve showing the effect of damping on the maximum accelerations of the

J bridge deck for a wind angle of zero degrees. Positive wind angles

of attack cause greater vortex response than negative angles for all

tests and models. For the higher levels of damping, the vent holes

in the curb barriers have little effect on vortex motion. It is

also evident that this shape box section is slightly more stable

without curb barriers than with curb barriers as a result of vortex

shedding. The vent holes in the barrier rail evidently were not

16-7
large enough to be effective, but a more open type of barrier rail

may prove to be beneficial.

Summary and Conclusions

The proposed steel box cross section of the Ohio River Bridge

at Huntington, West Virginia, has critical wind velocities for

flutter above 280 mph (450 kph) for wind angles of attack from

-4 degrees to +6 degrees. This velocity is well above any expected

winds at the bridge site.

Vertical motion caused by vortex shedding took place at all wind

angles of attack with plus angles causing more deck motion than minus

angles. With assumed prototype damping, maximum calculated vertical

deck accelerations ranged between 4 percent g and 8 percent g. However, it is unlikely that the wind over the full span of the

prototype would be steady and turbulent-free enough to cause motion

of quite this magnitude.

The following conclusions are present:

(1) The prototype bridge will not experience any aerodynamic

instability due to flutter.

(2) Vortex-induced oscillations are not a threat to bridge

safety; however, pedestrains or passengers in stationary

vehicles may feel some vertical motion if wind conditions

are right.

(3) The vent holes in the curb barriers, as designed for this

bridge, have no significant aerodynamic effect.

References

1. "Aerodynamic Stability of Suspension Bridges with Special Reference to the Takoma Narrows Bridge." University of Washington

Engineering Experiment Station Bulletin No. 116.

2. "The Mathemetical Theory of Vibration in Suspension Bridges"

by Freidrich Bleigh, C. B. McCullough, Richard Rosecrans, and

G. S. Vincent, U. S. Bureau of Public Roads, Government

Printing Office.

3. "Recent Methods in the Application of Test Results to the Wind

Design of Long, Suspended-Span Bridges" by R. H. Scanlan, Report

No. FHWA-RD-75-115 , October 1975, Federal Highway Administration,

Washington, D. C.

4. Scanlan, R. H. , "Vortex-Shedding Response," Private Communication, R. Scanlan to R. Gade, 1974.

5. Scanlan, R. H. , "Theory of the Wind Analysis of Long-Span

Bridges Based on Data Obtainable from Section Model Tests."

Proc . 4 International Conference on Wind Effects . London,

England, September 1975.

6. Scanlan, R. H. and Tomko, J. J., "Airfoil and Bridge Deck

Flutter Derivatives," Journal of the Structural Division, ASCE , Vol. 97, No. EM6, December 1971.

16-8
Acknowledgment s

This investigation was carried out at the request of the West Virginia Department of Highways. The bridge was designed by E. Lionel Pavlo Engineering Company of New York, New York. Thanks are

due them for computing the modes of motion and supplying information

needed to design the scaled model.

The author wishes to thank Harold Bosch, Structural Research

Engineer, for designing the model, Rick Ablaza, draftsman, who did

the shop and detail drawings for the section model of the bridge,

and Robert K. Taylor, machinist, who machined the many parts and

fabricated the model. Special thanks are due Ronald Nelson, Technician, who assisted in conducting the tests and reducing the data.

DECK CROSS SECTION

AND

SECTION IDENTIFICATION

VERTICAL TORSIONAL VERTICAL WEIGHT WEICHT MASS/ FT NATURAL NATURAL LOCATION PER FOOT 5-FOOT OF SPAN DESIGN FREQUENCY FREQUENCY OF C. C. MODEL

(CPS) (CPS) FROM BOX (lb) BOTTOM

MODEL lbs-s ec

(lb) ft*

POLAR MASS MOMENT OF INERTIA/FT

OF SPAN lbs-sec

ft

I Box section without

curbs and wearing

surface.

36' (Prototype)

PROTO .36 1.97 5.613 ft 3445 106.99

0.8' 1/45 Model

MODEL

FROUDE 2.415

SCALE

13.215 1.50 In 1.7013 8.506 .05284

ACTUAL 2.63 MODEL 1.96

1.62

6. 20 4.29

3.42

1.50 in 1.463

1.485

1.506

7.315 .04640

7.427 .04981 7.531 .05042

11017

.002687

.024 18 .03041 .02866

Box section without

uearing surface.

H is with vented

curb barriers.

PROTO .36 1.97 5.613 ft 3807 118.23

MODEL

FROUDE 2.415 SCALE

13.215 1.50 in 1.880 9.400 .05838

ACTUAL MODEL 2.48 5.95 1.50 in 1.631 J. 157 .05127

14218

.003467

JHJ

Box section with wearing surface.

H is with vented

curb barriers.

PROTO .36 1.97 5.613 ft 4677 145.25

MODEL

FROUDE 2.415

SCALE

13.215 1.50 in 2.310 11.548 .07172

ACTUAL 2.51 MODEL 2.43

2.54

16107

.003928

6.80 2.009 10.044 .06809 .02692 6.50 1.50 in 2.014 10.073 .07552 .03100

7.76 2.044 10.220 .06934 .02200

1.0 FOOT = .3048 METRES 1.0 POUND - .4536 KILOGRAMS FIGURE 1. DECK CONFIGURATION AND DYNAMIC PROPERTIES.

16-9
PROTOTYPE DESIGNATION

FOR DECK CROSS SECTION

PROTOTYPE VERTICAL

NATURAL

FREQUENCY

(CPS)

SCALED MODEL

VERTICAL

FREQUENCY

(CPS)

APPROXIMATE MODEL

DAMPING

LOG DEC

SECTION MODEL DESIGNATIONWIND ANGLES TESTED

DEGREES

+4 +2 -2

,36 ,30

.30

.30

.24

.24 .24

2.41 2.00

2.00

2.00

1.61

1.61

1.61

.08 .08

.05

.02 .08

.05

.02

1-2. 4-. 08 1-2.0-.08

1-2.0-

1-2.0-

1-1.6-

1-1.6- 1-1.6-

.05

.02

.08 .05

.02

2 2H

.36

.36 .08 .08 2-2.4- 2H-2.4-

,08

,08

3

3

33 3H

3H

.36

.36

.36

.36

.36

.36

.08

.05

.02

.01 .08

.01

3-2,

3-2,

3-2,

3-2, 3U-2, 3H-2,

4-. 08

4-. 05

4-. 02

4-. 01 4-. 08

4-. 01

X

X

X X X X

X X

FIGURE 2. SUMMARY OF WIND TUNNEL TESTS CONDUCTED.

MODEL NO WIND NO WIND VORTEX SHEDDINC RESPONSE FLUTTER RESPONSE TESTED DAMPING VERTICAL

AND WIND LOG FREQUENCY VERTICAL MAXIMUM CRITICAL PROTOTYPE ANCLE OF DECREMENT OF PROTOTYPE MOVEMENT ACC. OF WIND VELOCITY MPH ATTACK PROTOTYPE WIND VEL. OF DECK DECK ZERO WITH

CPS MPH D.A. FT. q's DAMPING DAMPINC

1-2. 4-. 08 +6* .094 .39 15.7 .645 .051 55 * +4* .102 .39 16.0 .352 .028 100 * +2* .099 .39 14. 1 .195 .016 105 * 0' .103 .39 15.6 .188 .015 270 * -2* .096 .39 0 0 400 * -4* .105 .39 0 0 425 * 1-2.0-.08

+4* .080 .29 11.9 .519 .029 270 440

0' -4* 1-2.0-.05

+4"

0' -4' 1-2.0-.02

0"

1-1 .6-. 08 +4"

o- -4' 1-1.6-.05

+4'

0' -4" l-l.6-.02

0*

.081 .076

.053

.051 .053

.025

.082

.074 .079

.056

.057 .055

.025

.29

.29

.29

.29 .29

.29

.24

.24 .24

.24

.24

.24

.24

10.9

11.8 9.2

8.3

11.5

.210

0

.747 .219

.222

.698

0

00

0

0

0

.012

0

.041 .012

.012

.038

0

0

0

0

0

0

295 325

235 255

290

255

140

150 285

155 190 240

200

470

460

350 400

430

320

320

380 450

280

360 400

280

* Wind velocity was well above the range of the test. 1.0 FOOT - .3048 METRES; 1.0 MPH = 1 .6093 KP1

FIGURE 3, SUMMARY OF RESULTS.

16-10
HODEL NO WIND NO WIND VORTEX SHEDDING RESPONSE FLUTTER RESPONSE

TESTED DAMPING VERTICAL AND WIND LOG FREQUENCY VERTICAL MAXIMUM CRITICAL PROTOTYPE

ANCLE OF DECREMENT OF PROTOTYPE MOVEMENT ACC. OF WIND VELOCITY MPH ATTACK PROTOTYPE WIND VEL. OF DECK DECK ZERO WITH

CPS MPH D.A. FT. q's DAMPING DAMPING

2-2.4-.09

+6' .083 .37 14.8 1. 130 .090 110 390 +4* .086 .37 14.9 .933 .074 120 350 +2 % .085 .37 15.0 1.020 .081 135 350

0* .084 .37 13.6 .715 .057 120 350 -2' .084 .37 13.1 .676 .054 190 335

-4* .084 .37 13.1 .676 .054 150 300 2H-2.4-.08

+6* .082 .37 15.0 1.160 .092 110 * +4* .088 .37 15.4 1.060 .084 175 335 +2* .084 .37 15.4 .968 .077 120 370

0* .085 .37 13.4 .654 .052 200 390

-2* .089 .37 13.0 .659 .052 210 390

-4* .087 .37 13.0 .678 .054 210 * 3H-2.4-.08

.099 .38 14.4 .684 .054 A * +4' .099 .38 13.5 .586 .046 * * +2* .106 .38 14.2 .618 .049 * * 0* .101 .38 13.4 .521 .042 A A -2* .103 .38 13.9 .473 .038 * A -4* . 103 .38 14.4 .558 .044 * A

* Wind velocity was well of. the teat. 1.0 FOOT - .3048 METRES 1.0 MPH " 1.6093 KPH

FIGURE 4. SUMMARY OF RESULTS.

MODEL NO WIND NO WIND VORTEX SHEDDING RESPONSE FLUTTER RESPONSE

TESTED DAMPING VERTICAL

AND WIND LOG FREQUENCY VERTICAL MAXIMUM CRITICAL PROTOTYPE

ANGLE OF DECREMENT OF PROTOTYPE MOVEMENT ACC. OF WIND VELOCITY MPH ATTACK PROTOTYPE WIND VEL. OF DECK DECK ZERO WITH

CPS MPH D.A. FT. q's DAMPING DAMPING

3-2. 4-. 08

+6" .076 .38 12.8 .992 .079 A A +4* .083 .37 13.9 .819 .065 A

/ A +2* .083 .37 14.

1

.684 .054 A A 0* .082 .37 13.4 .567 .045 A A -2 s .096 .37 13.8 .564 .045 A A

\ ft -4' .093 .37 13.2 .494 .039 A * 3-2. 4-. 05

+4 S .065 .36 15.0 1.530 .122 120 425

0' .057 .36 15.0 .958 .076 175 375 -4* .056 .36 13.0 .867 .069 120 425

3-2. 4-. 02

0

s .023 .36 17.2 2.48 .197 A A 3-2. 4-. 01 +6' .010 .38 21.0 5.41 .430 135 A +4* .012 .38 20.6 3.30 .262 120 A +2" .007 .38 16.9 3.68 .293 120 A 0

s .009 .38 15.2 3.00 .239 120 A -2* .010 .38 16.

1

2.63 .209 115 A -4' .009 .38 17.4 1.86 • .148 115 A 3H-2.4-.01

+6* .008 .38 18.2 5.74 .456 125 A +4* .010 .38 19.7 5.63 .447 120 A +2* .006 .38 21.5 5.02 .399 125 A 0* .010 .38 18.8 4.99 .396 125 A -2* .008 .38 17.2 2.80 .223 120 A -4' .009 .38 14.8 1.80 .143 110 A

* Wind velocity was well above the range of the test. 1.0 FOOT - .3048 METRES 1.0 MPH - 1.6093 KPH

FIGURE 5. SUMMARY OF RESULTS.

16

16-11
ACCELERATION OF BRIDGE DECK - G's

FIGURE 6. VERTICAL VORTEX RESPONSE VS. DAMPING.

Continuous Main Box Girder 1723 1 -0" (525.17 m)

Tower FIGURE 7. GENERAL ELEVATION OF PROTOTYPE.

16-12
16

16-13
16-14

16-16
17

SOME AERODYNAMIC CONSIDERATIONS IN THE DESIGN OF THE RUCK-A-CHUCKY BRIDGE

Robert H. Scanlan

Department of Civil Engineering

Princeton University

Some Aerodynamic Considerations in the

Design of the Ruck-A-Chucky Bridge

The design of the Ruck-a-Chucky bridge over the American River,

upstream of the proposed Auburn dam near Sacramento, California

is unique in several respects. The bridge, spanning 1200 feet,

is cable-strayed, suspended by a spectacular "web" of cables from

the rock of the adjoining hillsides, and curved in plan. It was designed by T. Y. Lin International of San Francisco. Ref. 9 presents a few of the interesting details of this bridge.

From an aerodynamics viewpoint the bridge was studied primarily with regard to the following questions:

(1) Maximum expected winds at the site and possible

local terrain effects.

(2) Possible aerodynamic instabilities.

(3) Expected response to wind buffeting.

(4) Erection safety under wind.

(5) Possible wind-induced cable oscillations.

Steady Winds at the Bridge Site

Basic maximum winds for 50- and 100-year return periods in

the region of the bridge were obtained from Thorn's isotach maps

[1], which show the maximum fastest-mile speeds in the region to

be about 70 mph to 80 mph at 30 ft above ground in open terrain.

It was convenient in the subsequent calculations to work with

wind speeds assumed to hold at bridge height, about 60 ft. above

the expected water surface. The wind speed U at any elevation

Z was calculated from the formula

U(Z) = 2.5u* In— (1)

Z

o where

u^ = friction velocity, calculated from eq. 1 for given

reference data, such as the Thorn charts with

Z

Q = 0.02 meter. (Z

Q

= fetch roughness length)

U(Z) = wind velocity at height Z

Table 1 below lists representative wind velocities U at bridge height and corresponding velocities Umet at the

meteorological reference height of 10m.

Table 1

U(mph) 40 60 60 70 80 90 100 110

Umet (mPh) 36 45 54 63 72 81 90 99

17-1
Study of Terrain Effects

To study the possible orographic and local topographic effects

at the bridge site, a small-scale terrain model was built and

tested in the boundary-layer wind tunnel at Colorado State University, Fort Collins, under the direction of Professors J. E. Cermak

and J. Peterka. This model revealed no unusual channeling, wind

intensification, or other local factors meriting special attention

in the design of the bridge. Typical wind longitudinal gust spectra obtained at the bridge location with this wind tunnel model

exhibited normal characteristics.

Study of Possible Aerodynamic Instabilities

The bridge deck is a box girder, curved in plan. The outline

of the final external cross-sectional shape arrived at for this

deck is shown in Figure 1. On the right side of the sketch is a grid-covered equestrian path, an unusual feature of the design. The

basic section shape was arrived at by interaction among the T. Y. Lin design team, notably H. K. Lu , Alan Firmage, and the writer,

with additional important contributions by J. E. Cermak of Colorado

State. The shapes of the outermost edges were strongly influenced

by other studies previously conducted by R. L. Wardlaw [2].

For suspended spans the shape of the deck cross-section lies

at the heart of the wind stability problem. And the final crosssection in this study, indicated in Figure 1, had very stable

tendencies against wind-induced flutter.

The method chosen for examination of the bridge aerodynamics

was centered on the experimental study of a section model of the

bridge deck. From this, basic aerodynamic force coefficients and

self-excitation derivatives [3] were obtained in the wind tunnel

at Colorado State University. This work was carried out by

B. Bienkiewicz under the direction of Professor Peterka. Following

the basic analog study in the wind tunnel, the aerodynamic data were employed in analytical studies by the writer to describe the

expected responses of the full three-dimensional bridge.

The sectional test model proved to be economical and convenient

and allowed rapid modifications of form leading to improved stability

A noteworthy item is that susceptibility to vortex-induced oscillation was found to be very low with the final cross-section.

Space does not permit a full discussion of the stability derivatives determined for this bridge deck, but it is of interest to comment on the key derivatives H-^ and A2 [cf. Ref. 3], which

govern aerodynamic damping in vertical motion and in torsion,

respectively. These coefficients remained wholly negative

indicating increasingly safe conditions —for all reduced velocities

U/NB where U = wind velocity across the deck, N = oscillation frequency, and B = 58 ft is the deck width.

The possibility of flutter was thus ruled out up to a wind

velocity at bridge elevation of 150 mph or more.

Response to Wind Buffeting

The response to buffeting by gustiness was analyzed by an extension of methods described in Ref. 4 by Scanlan and Gade.

Because of the curved plan form of the Ruck-a-Chucky bridge its

vibration modes are more complex than with straight bridges, and

17-2
analytical extensions of the methods of Ref. 4 were needed,

employing with the more complex modes the same basic aerodynamic

derivatives and force coefficients obtained from the sectional

model, plus additional information on natural wind spectra.

The following wind spectra were employed [5]

Longitudinal

200u^.

S u(n) =

2 i

U

Vertical

:

Sw< n >

[1 + 50ip] 5/3

2 3.36u* —

U

i + ioe_z

) 5/3

where u^ is the friction velocity, Z the altitude, U the wind velocity, and n frequency in H .

As an example of calculated buffeting results, the maximum

expected vertical excursions of the bridge at center span are given

in Table 2.

Table 2

U(mph) 40 50 60 70 80 90

Max. Vert. defl. (ft.) 0.48 0.71 0.99 1.41 1.73 2.07

Because of the stiffness and curved span form of the bridge

buffeting deflections in torsion are negligible.

A full account of the generalized equations of linear buffeting

theory, which grew out of the present study, is given in Ref. 6.

Erection Safety Under Wind

The inherent stability of the bridge cross-section is a strong

Indicator of primary safety from catastrophic oscillation during

construction. However, some considerations were also given to the

effects of temporary tie-downs and eventual dampers should the need

arise. The detailed configuration of these will not be considered

antil the construction of the bridge is actively planned.

leasures Against Wind-Induced Cable Oscillations

Although a number of highly effective, external preventative

neasures against wind-induced cable oscillation, such as strakes,

:uned-mass dampers, shock absorbers, etc., are well-known and docunented in the literature [7] [8], full details of such devices have

tot been elaborated with respect to the present design, in view

17-3
of the following very important preventive measures already

considered

:

(1) Special study was undertaken by the T. Y. Lin design

team, notably H. K. Lu, C. Redfield, and the writer,

to provide extensively for relief from possible oscillatory stresses at the cable terminations. These latter were provided with long transitions of gradually

increasing stiffness before conducting the individual

cable wires into hard points.

(2) The cables are to be grouted within their outer sheaths

along their full lengths, providing greatly increased

damping

.

(3) The widely varying cable geometric pattern throughout the

overall design renders very unlikely the possibility of

wake-induced oscillations of one cable by another.

Cone lusion

This paper has sketched the preventive measures taken against

wind-induced oscillations related to the Ruck-a-Chucky Bridge.

The Primary concern—of removing the possibility of catastrophic

flutter—was fully met by careful attention to the design of the

bridge cross-section so that both vertical and torsional aerodynamic

dampings became such as to increase the stability of the design.

The wind tunnel section model was used as the basic study tool.

Results from this, in the form of non-dimensional coefficients, were then incorporated into both flutter and buffeting analyses using

methods developed by the writer.

Attention was given to avoiding or alleviating the vortexshedding problem, both as regards the deck cross-section itself,

and the possible effects on the cables. Peripheral aerodynamic

eventualities were also given consideration.

References

1. Thorn, H. C. S.: "New Distributions of Extreme Wind Speeds

in the United States" J. Struct. Div. , ASCE, 94, No. ST7, Proc. Paper 6038 (July 1968) 1787-1801.

2. Wardlaw, R. L. : "A Preliminary Wind Tunnel Study of the

Aerodynamic Stability of Four Bridge Sections for the

Proposed New Burrard Inlet Crossing" Report No. LTR-LA31, NAE, National Research Council, Ottawa, Canada, 1969.

3. Scanlan, R. H. and Tomko, J. J.: "Airfoil and Bridge Deck

Flutter Derivatives" J. Eng. Mech. Div. , ASCE, 97, No. EM6, Proc. Paper 8609, (Dec. 1971), 1717-1737.

4. Scanlan, R. H. and Gade, R. H. : "Motion of Suspended Bridge

Spans Under Gusty Wind" J. Struct. Div. , ASCE, 103, No. ST9

,

Proc. Paper 13222 (Sept. 1977), 1867-1883.

5. Simiu, E. and Scanlan, R. H. : Wind Effects on Structures

J. Wiley & Sons, New York, 1978.

17-4
6. Scanlan, R. H. : "The Action of Flexible, Bridges Under Wind:

I. Flutter Theory; II. Buffeting Theory Jnl. Sound and

Vibration , Southampton, U. K. (in press, 1978).

7. Scanlan, R. H. and Wardlaw, R. L. : "Reduction of Flow-Induced

Structural Vibrations" in Isolation of Mechanical Vibration , Impact, and Noise AMD Vol. 1, Sect. 2, ASME, New York, 1973,

pp. 33-63.

8. Scanlan, R. H. and Wardlaw, R. L. : "Aerodynamic Stability

of Bridge Decks and Structural Members" Proc . Engrg. Symposium on Cable-Stayed Bridges , Pasco Wash. , FHWA and ASCE,

Dec. 6-7, 1977 (in press).

9. . "Get Ready for Ruck-a-Chucky" , Western Construction , Mar. 1977, pp. 34-35.

I 1

FIGURE 1: MODIFIED DECK SHAPE USED IN WIND

STABILITY CALCULATIONS

17-5
I
18

A SUMMARY OF WIND-TUNNEL TEST RESULTS FOR

THE LULING, LOUISIANA, CABLE-STAYED BRIDGE

Harold R. Bosch

Federal Highway Administration

Washington, D.C.

A Summary of Wind Tunnel Test Results for

the Luling, Louisiana, Cable-Stayed Bridge

Ihis paper summarizes the major results of a comprehensive aerodynamic investigation of the proposed cable-stayed bridge across

the Mississippi River at Luling, Louisiana. The bridge, as illustrated in Figure 1, has a main span of 1,235 feet (376.4 m) comprised of double trapezoidal steel box-girders with an orthotropic

ieck. The supporting towers are 350 feet (106.7 m) in hei'ght with

sloped, tapered steel box legs and upper and lower cross-struts

tfhich form an A-frame shape. It is important to note here that

although cable-stayed bridge concepts have been used in Europe for

nany years, they are only recently receiving popularity by designers

in the United States. The Sitka, Alaska Bridge was the first such

jridge to carry vehicular traffic in this country.

This study was conducted in the George S. Vincent Wind Tunnel

)y the Federal Highway Administration (FHWA) for the Louisiana

)epartment of Highways. The objectives of the study may be sumlarized as selection of an optimum superstructure cross-section,

•.valuation of the erection stability for the preferred section,

valuation of the erection stability for the main towers, and

letermination of flow field characteristics for purposes of preiminary design and development of field wind instrumentation.

Structural response to vortex shedding and critical wind velocity

'or flutter together with aerodynamic data in the form of flutter

:oefficients are presented.

listorical Background

t has been well know for many years that serious vibration probems can arise with suspended bridge structures as a result of

ind action. Perhaps one of the earliest well-documented examples

s the failure of the Brighton Chain Pier Bridge in England in

836 which was witnessed and reported on by Lt. Col. William Reid

f the British Army. In fact, as noted by Farquharson [1], ten

iuspension bridges suffered major damage or collapsed during the

eriod 1818 to 1889. Most of these failures were of bridges of

elatively short spans.

The half-century following the 1889 failure of the 1,260-foot

384.1 m) Niagara-Clifton Bridge marked a period of rapid advance

p design and construction techniques as well as materials and led

Jo the appearance of lighter structures with significantly increased

pan lengths. The original Tacoma Narrows Bridge was the product

if such progress and its dramatic failure in 1940 is, perhaps,

|!he best-known example of an aeroelastic bridge deck instability,

riis event sparked vigorous examination of the aerodynamic stability

f bridge superstructures. The University of Washington immediately

onducted a series of analytical and experimental investigations [1]

ip determine the causes of the Tacoma Narrows disaster and establish

esign criteria for the new structure. Shortly thereafter, similar

bt more extensive studies [2] were conducted on the Golden Gate

ridge. Since these early endeavors, many wind tunnel research

tudies have been performed to evaluate proposed designs or deterLne appropriate modifications for existing "problem" structures

: id have made invaluable contributions to the understanding of the

ridge stability problem.

During the past decade there has been considerable interest

i employing wind tunnel techniques for the determination of the

^rodynamic stability derivatives of suspended bridge models [3].

18-1
These derivatives represent purely aerodynamic data, as opposed to overall dynamic response observations, which are the consequence,

uniquely, of bridge cross-section geometry.

In earlier wind tunnel tests, bridge deck section models were elastically suspended, with properly scaled vertical and torsional

natural frequencies, and observed primarily for their tendencies

to flutter at some air speed. The present practice is to extract

the purely aerodynamic characteristics of the model first and then

employ these in the dynamic response calculations for the full

bridge. This practice has the inherent advantage of divorcing

test results from all model structural properties, such as inertia,

damping, and frequency, except for geometry.

Wind Environment

The rational treatment of interactions between wind in the atmospheric boundary layer and man and his works on the surface of

the earth is currently termed "wind engineering." During the last

decade, intense interest in the field of wind engineering has led

to considerable advances in defining the wind environment likely

to affect a given structure. The wind engineer's task is not only

to minimize loss of life and damage to property that result from

adverse wind, but also to maximize human comfort and economy by

using beneficial wind effects.

Before conducting a bridge wind stability investigation, it is important to determine the overall characteristics of the wind

environment at the proposed bridge site. Such site studies usually

include the gathering of applicable strong-wind data for a number

of years, with emphasis on wind velocity, direction, and frequency

of occurrence. The U.S. Weather Bureau and similar records have

historically been the primary source of this information. From

these data, plots of high wind speeds versus probability of occurrence may be developed. For areas within the United States, Thorn [4]

has provided extremely valuable wind data in the form of maps of the

distribution of extreme "steady" winds for different anticipated

recurrence intervals ranging from 2 to 100 years. The data given

are for mean high winds at a nominal height of 30 feet (9.1 m).

Local plots, called "wind rows", may also be developed showing the

probability of exceeding different velocity levels at each value

of horizontal azimuth.

Among the important elements in wind environment studies is

attention to the upstream conditions, or fetch, which strongly

affect the vertical wind profile and degree of turbulence in the

wind. For relatively small surface roughness, such as over a long

fetch of open sea, the velocity profile flattens and turbulence

level decreases, resulting in potentially higher ground level

velocities and a possible enhancement of bridge vortex shedding

response or flutter. High surface roughness, as encountered in

urban areas, generates strong turbulence in the oncoming wind which

is a principal factor in bridge buffeting.

When the bridge site is unique or subject to severe stcrms

such as hurricanes or tornadoes, the aforementioned information

should be supplemented by topographic wind tunnel studies or detailed field investigations. These provide a better understanding of the turbulence structure, maximum velocities, and angles of

attack to be expected at the bridge site.

18-2
Wind Effects on Suspended Bridges

The designer has before him many possibilities when selecting the

type of superstructure for a modern bridge. Whatever the choice,

he must ensure that the aerodynamic stability of the section is

satisfactory. To this end, it is necessary to understand the

character of wind forces acting on the structure.

There are two basic types of wind effects on structures,

static and dynamic. The static effect is the pressure that the

moving air mass exerts against the structure, and it varies in

direct proportion to the square of the mean velocity. The dynamic

effects are the forces generated by flow separation and turbulence

of the moving air as it passes around the structure or by changes

in wind speed such as gusting.

In consideration of the static stability of bridge structures

under steady wind, the critical wind direction is normally taken to be perpendicular to the longitudinal axis of the bridge. Lift,

drag, and moment coefficients are assumed or measured in the wind

tunnel for use in calculating the respective uplift, side thrust,

or twist forces on the structure. When lift or drag forces become

excessive, an unstable bridge may undergo lateral buckling or torsional divergence.

As a result of interaction with the earth's surface, the natural

wind is not smooth but turbulent and gusty, with velocity fluctuations

in both the vertical and horizontal directions. These fluctuations

are random in nature, occurring over a wide range of frequencies.

When the dominant frequencies of the wind spectrum are at or near

the natural vibration frequencies of the structure and the winds are strong, buffeting can become a problem and lead to overstressing of

structural elements. Fortunately for a majority of cases, the

peak in the wind energy spectrum occurs at frequencies somewhat

lower than those expected for large bridges.

Another form of wind-induced vibration of bridges is that of

vortex shedding response. When wind blows across a slender bluff

object, such as a bridge superstructure, the large wake that is

formed consists of an array of vortices which are spaced in a characteristic pattern. The vortices form alternately on either

side of the body in a periodic fashion, with the shedding frequency

proportional to wind speed. The fluctuating nature of the wake

causes an oscillatory force on the bridge and as shedding frequency

approaches the natural frequency of the structure, it is possible

for vibrations to result. This response occurs over a range of

wind velocities in the vicinity of the resonant value, with maximum

amplitude obtained when the "lock-in" condition is achieved. Both

the velocity range and maximum amplitude are functions of structural

damping and degree of streamlining. Since vortex shedding response

is amplitude limited, structural problems arising from this mechanism

are seldom catastrophic, although local fatigue damage and human

discomfort are possible and warrant consideration.

The final class of instability, and perhaps the most important,

is commonly referred to as flutter, following terminology developed

in early work on aeroelastic instabilities of aircraft. In this

case, wind forces interact with bridge deck motions to create a self-excited oscillation of growing amplitude, possibly leading to destruction. Unlike the vortex shedding problem, the flutter phenomenon is not limited to a velocity range beyond which structural

stability returns. Increasing velocity beyond the threshold value

18-3

18
causes oscillations to buildup more rapidly and leads to more violent action.

Mathematical Modeling

To extract detailed aerodynamic derivative data from the wind

tunnel section model studies, it is necessary to provide an analytical model of the physical situation. For this purpose, a symmetrical two-degree-of-freedom system is assumed. As discussed by

Scanlan [5], the equations of motion for the section structure may

be written in the form:

i[h + 2C to h + u

2h] = L (1)

h h h

I[a + 2? u> a + w

2 a] = M (2)

a a a

where I and m are mass moment of inertia and mass per unit span

of model; h and a are vertical and torsional deflections of the

given deck section (assumed uniform; c^ and C are the damping

ratios-to-critical in the h and a degree-of-freedom; and

i» a are the circular frequencies in the h and a degrees-offreedom; L and M are the dynamic lift and moment per unit span.

For the self-excited flutter problem, L and M are taken as linear

and may be written in the form:

L = m[H h + H a + H^a] (3)

M = itA^h + + A a] (4)

where and A^ (i = 1, 2, 3) are dimensional aerodynamic derivatives determined by experiment, with inertial factors introduced

for convenience. By noting the fact that L and M are essentially

functions of the "reduced frequency" dimensionless parameter

K = Bw/U, nondimensional derivatives or coefficients may be obtained

as

:

L = l/2pU 2 (2B) [KH* + KH* + K

2 H*a] (5)

1 U 2 U 3

M = l/2pU 2 (2B) [KA* — + KA* — + K

2 A*a] (6)

1 U 2 U 3

where and A^ (i = 1, 2, 3) are nondimensional aerodynamic coefficients and U is oncoming wind velocity; p is air density. These

equations have been reduced to experimental form to enable direct

use, and the details of this reduction are included in Reference 5. Essentially, test observations produce time histories of h and a; then with all quantities on the left-hand side of equations (1)

18-4
•k ~k and (2) known, coefficients H£ and are matched by system indentification techniques to the test observations as functions of K.

The coefficients are important in evaluating the stability

characteristics of a section. The A2 coefficient measures aerodynamic damping in the torsion mode and th^s is an indicator of

single-degree flutter in that mode. The A3 coefficient is usually

small and is sometimes assumed zero in analysis. In general, it is

a^easure of the aerodynamic stiffening effect upon frequency. The

coefficient measures the aerodynamic damping in the vertical

mode

.

When vortex shedding response is exhibited either indirectly

in the aerodynamic coefficients, namely , or directly in observations of the model test, the aforementioned mathematical model

must be augmented to be applicable to the region of response [6]

.

This can be done for the case of purely vertical motion by providing

equation 5 with a special forcing function:

L(t) = ~ P U

2 (2B)C Lo sin Ug t (7)

The resulting combined equation takes on the form:

m[h + 2r o^h + u^h] = ~ pU

2 (2B) [KH* ~ + C Lq sin uigt] (8)

o

where H

Q is an aerodynamic damping coefficient for decay to a resonant response amplitude; is an oscillating lift coefficient

and Ug is the Strouhal circular frequency. It should be noted

here that a completely analogous , model could be developed for the

torsional case. The values of , > snd ojg are determined

experimentally from records of oscillations at the vortex "lock-in"

condition and enable the ultimate prediction of prototype response

amplitude.

The "strip" equation for the prototype deck section is expressed

as

2 m(x) [ n + 2 y (o n + u>_ n] f(x) = P sin u t (9)

[land the maximum mid-span amplitude is computed as:

(10)

"l^max 9 1M

where

Ys

2M

P B

2H*

o

ws Ys

2m

2 P = pU BCt . Y (x) dx = generalized force amplitude,

LiO span

M = m

S pan

,C (x)dx = generalized mass,

fCx) = normalized span deflection

) 18-5
f(x) is approximated by assuming that the main span deflected shape

is represented by a half wave sine curve.

Design and Construction of Models

A section model of a suspended bridge superstructure is a geometrically similar copy of a typical length of the suspended

structure only. Any cables or hangers are not modeled because

their aerodynamic effects are considered negligible. The

section model is rigidly constructed to ensure that all elastic

behavior takes place in the springs from which it is suspended.

The vertical and rotational movement of the section model supported by its spring suspension simulates the vertical bending

and torsional twist of the prototype structure.

Four basic section models with 15 potential configurations

were designed and constructed for this test program. Design was performed by Carroll Associates, Inc., of Maryland, while fabrication and assembly were performed by Frazier Precision Instrument

Co., Inc., of Virginia, and the FHWA Mechanical Design and Experimental Fabrication Group, respectively. All models were designed

to a geometric length scale of 1:60 to ensure adequate reproduction

of significant structural details. Model length was set at 60-inches

(1.524 m) inside to inside of end plates representing a 300-foot

(91.46 m) section of prototype span. Model widths vary from

16.85-inches (.4281 m) to 19.95-inches (.5070 m), resulting in

aspect ratios between 3.56 and 3.00. End plates were employed

so that two-dimensional flow could be maintained near the model ends.

The models were scaled according to Froude number criteria for mass properties and vertical natural frequency. Due to practical limitations of simulating the relatively high prototype torsional characteristics, values of polar mass moment of inertia and torsional

natural frequency do not, however, achieve Froude scaling.

To maintain proper weight and weight distribution, develop sharp

edges, and provide necessary model strength and rigidity, a variety

of lightweight materials were incorporated into the model designs.

Girder webs, floor beams, and asphalt decks are constructed of

aircraft birch plywood. Girder and beam flanges, edge strips, and

railing lips are aluminum. Balsa wood was used for diaphrams,

ribs, traffic barriers, stiffeners, and concrete decks. The end

plates are of "sandwich" construction, with balsa wood between

aluminum sheets. Edges were tapered to reduce drag effects, and an elliptical shape was chosen to minimize weight and polar mass moment

of inertia. Trusses with tubular aluminum members were extended from

the end plates to provide crossbars 90-inches (2.286 m) apart for

attachment of the supporting springs. Traffic barriers are attached

with machine screws to facilitate removal for modification of the

cross-section. Details of the seven Luling configurations tested

are illustrated in Figure 2 and Table 1.

Test Procedure and Instrumentation

The tests were conducted in the FHWA open-circuit, low-velocity

wind tunnel, which was designed particularly for testing bridge

section models. A blower forces air through a diffusing section

with a series of stainless steel wire screens into a cylindrical

pressure chamber and through the nozzle from which the air is

discharged onto the model. The velocity over the cross-section

of the wind stream is very uniform to a point 6 feet (1.8 m)

beyond the end of the nozzle.

18-6
A steady wind velocity, with a range of 0- to 30-feet per

second (9.1 mps), is maintained with a micro-control rheostat

in the blower circuit. The wind velocity is computed from

the dynamic pressure head of the wind stream obtained by direct

measurement with a pitot-static tube and an inclined manometer.

Corrections are made for changes in air or fluid temperature

and barometric pressure.

The nozzle can be rotated about the horizontal axis of the

pressure chamber, allowing tests at various wind angles of attack.

The angle of attack is the vertical angle between the wind velocity

vector and a horizontal plane; a positive angle of attack indicates

an upward wind.

The section models were suspended in front of the wind tunnel

nozzle by means of four supports with upper and lower coil springs

at each support, as illustrated in Figure 3. The vertical length

of the springs was adjusted to set the proper vertical frequency

of oscillation. The torsional frequency was set by adjusting the

spacing between the springs at each end of the model. Although

it was not possible to attain the relatively high torsional frequency desired, attempts were made to obtain as high a frequency

as possible. The total mass of the section model, end plates,

support trusses, end bracket crossbars, and "moving" springs was

tuned to the desired scaled quantity by the addition of small

weights to the end plates at the model center of gravity.

Horizontal restraining wires were attached to the end bracket

crossbars at the center of gravity of the model to prohibit longitudinal or lateral movement during the tests. The restraining

wires, which were tensioned with 2-pound (0.9 kg) weights, were

of sufficient length that their vertical components would have no significant effect upon the vertical or torsional motion of the

nodel

.

Electrical resistance strain gages were attached to beryllium

copper flexure plates above the coil springs. The strain gages at

one end of the model were wired to indicate the vertical component

Df oscillation, and those at the other end were wired to indicate

the torsional component. Strain gage outputs were amplified and

the two components were recorded side-by-side on an oscillograph.

The models were tested only under laminar flow conditions.

The tests were conducted at wind angles of attack varying from

*6° to -4°. While maintaining a steady wind velocity at each of

a series of increasing values, the models were displaced in the

respective vertical and torsional degrees-of-freedom, released, and

he oscillatory dynamic response recorded. For the vertical mode,

:he initial displacement was accomplished manually. For torsion,

lowever, it was necessary to "hold" the model at a 4° displacement

and release it using solenoids. The velocity and response data

jtfere processed on the FHWA computer system to obtain the aerodynamic

derivatives and vortex response information discussed earlier.

discussion of Results

Test results for each of the seven Luling configurations are sumjarized in Figures 4 through 9.

No vertical flutter (self-excited, divergent, vertical oscil-

.ations) was exhibited by any of the model configurations during

Laboratory testing. This observation is verified by the computer

18-7
plots of the aerodynamic derivatives. The Hj derivative, the function for aerodynamic damping of vertical oscillations, shows positive damping over the full test range. A typical plot of this

derivative is presented in Figure 4. Torsional critical flutter

velocities at zero damping and 1 percent damping are illustrated

for prototype wind speed and wind angle (angle of attack) in Figures 5 and 6. Torsional flutter (self-excited, divergent, torsional oscillations) was exhibited at horizontal wind speeds

greater than equivalent prototype design requirements of 150-miles

per hour (240 km/h). The greatest flutter stability was demonstrated by the most "streamlined" models, C-6-C-A, C-6-C-B, and

C-2C-A, with critical velocities in excess of 250-miles per hour

(400 km/h). As angle of attack increases in the positive direction,

flutter stability generally decreases.

Since high wind velocities at high angles of attack are not considered likely in view of data obtained at the Severn and

Newport Bridges, critical flutter velocities of less than design

wind speed at high angles of attack are not considered a point of concern. Again, these observations are verified by computer

plots of the A2 derivative, indicative of torsional aerodynamic

damping. When A2 becomes positive, self-excitation tendencies are restrained only by the structural damping. A typical plot of this

derivative is presented in Figure 7.

An excitation of the bridge deck in vertical and torsional

modes by the resonant action of the periodic shedding of vortices

in the wake of the structure was exhibited by all models. The

amplitude-limited oscillation can be an unacceptable characteristic of the design when it occurs at moderate wind speeds, and

resulting accelerations are disturbing to the user. Generally,

vertical vortex shedding response was exhibited at various

velocities or ranges of velocities between 15- and 30-miles per

hour (24 and 48 km/h), while torsional vortex shedding response

was exhibited between 60- and 100-miles per hour (96 and 160 km/h).

Since torsional response occurs at relatively high velocities,

its effect on the user may be disregarded. For the vertical

response which occurs at moderate wind speeds, deck accelerations

of 2 percent of "G" are considered acceptable [7], Estimates

of prototype accelerations are illustrated in Figure 8 for

vertical response in a horizontal wind. Models C-2C-C and C-2C-A

demonstrated the least overall vertical response, while Model

C-2-C-2 exhibited the most. These estimates are considered conservative since it is probable the real wind over the bridge span will

be gusty enough to cause poor spanwise correlation and decrease the

likelihood of "lock-in" to occur.

Drag coefficients were obtained for the preferred Luling crosssection, C-2C-A, and its three erection stages as illustrated in

Figure 9. For stage Al , the median barrier was removed; A2, median

barrier and fairings removed; A3, median barrier, fairings and railings removed. Horizontal pressures were computed from the" coefficients for a moderate wind angle of +2° and were found to be

somewhat less than recommended AASHTO values for configuration

C-2C-A and about equal to the AASHTO values for stage A2.

Summary and Conclusions

This paper summarizes the major results of a comprehensive wind

tunnel study of a major suspended bridge structure. The bridge

studied is an example of the cable-stayed, box-girder type which,

until recently, has not often been constructed in the United States.

In general, all the cross-sections tested in a horizontal wind have

18-8
critical velocities for flutter in excess of the appropriate design

wind speeds. At other wind inclinations, the critical velocities

remain high enough to exceed expected winds. Varying degrees of

vertical motion are caused by vortex shedding at speeds between

15- and 30-miles per hour (24 and 48 km/h). Vertical vortex shedding response amplitudes and accelerations are minimal and considered

neither damaging to the structure nor discomforting to the user.

Drag forces measured in the wind tunnel appear reasonably relative

to recommended design values. Further details regarding results

presented as well as results for other parts of the test program

may be found in the recent study report: "Aerodynamic Investigations of the Luling, Louisiana, Cable-Stayed Bridge," FHWA-RD-77-161

.

This report will be available through the National Technical Information Service, 5285 Port Royal Road, Springfield, Virginia 22161.

Ac knowledgments

This project was carried out at the request of the Louisiana

Department of Highways. The author gratefully acknowledges the

technical support and guideance provided by Mr. Richard H. Gade,

Project Manager for FCP Project 5A. Despite Mr. Gade's failing

health in recent years, he never ceased to serve as a guiding

force on this and other projects in the Wind Engineering field.

Mr. Gade died on September 25, 1977, and he will be deeply missed

by all those who knew him.

References

1. Farquharson, F. B., et_ a_l. , "Aerodynamic Stability of Suspension Bridges with Special Reference to the Tacoma Narrows

Bridge," University of Washington Engineering Experiment

Station, Bulletin No. 116.

2. Vincent, G. S., "Golden Gate Bridge Vibration Studies,"

Journal of the Structural Division , ASCE, October 1958.

3. Gade, R. H. and Scanlan, R. H. , "Experimental Measurement

and Interpretation of Aerodynamic Stability Coefficients for

the Decks of Two Cable-Stayed Bridges," Proceedings of ASCE

and EMD Symposium on Dynamic Response of Structures , UCLA,

March 1976.

4. Thorn, H. C. S. , "New Distributions of Extreme Winds in the

United States," Journal of the Structural Division , ASCE,

Volume 94, pp. 1787-1801, July 1968.

5. Scanlan, R. H. , and Tomko, J. J., "Airfoil and Bridge Deck

Flutter Derivatives," Journal of Engr. Mech. , Volume 97,

ASCE, December 1971.

6. Scanlan, R. H. , Recent Methods in the Application of Test

Results to the Wind Design of Long, Suspended-Span Bridges , Federal Highway Administration, Report No. FHWA-RD-75-1 15,

October 1975.

7. Wardlaw, R. L. , and Buckland, P. G. , "Some Aerodynamic

Considerations in Bridge Design," The Journal of the

Engineering Institute of Canada , April 1972.

18-9
TABLE 1: CROSS-SECTIONAL DATA

Vertical Torsional

Prototype natural natural

Sect ion section frequency, frequency. model depth. Roadway Railing in cycles in cycles

designation in feet surface type per second per second

(1) (2) (3) (4) (5) (6) C-2-C-1 14 Concrete Basic 2.62 5.29

C-2-C-2 14 Concrete Alternate 2.56 5.20

C-2C-C 14 2 . 56 5. 35

C-2C-A 14 Asphalt Alternate 2.80 5.25

C-2A-C-1 12 Concrete Basic 2.53 5.24

C-6-C-B 14 Concrete Basic 2.49 5.33

C-6-C-A 14 Concrete Alternate 2.48 5.24

1 ft - 0.3048m

Continuous Steel Twin Box Girder 2745' (836.9m)

Section B-B

FIGURE 1: BRIDGE ELEVATION AND CROSS-SECTIONS

18-10
Anrnn

16. 867"

C-2-C-1

18.87"

C-2-C-2 A

18.537"

C-2C-C and C-2C-A

C-6-C-B

19.958"

C-6-C-A

3A7"

F 3.51"

L

1

I 3.U7"

3-33'

A

16.852"

nrr/ T

3-06"

C-2A-C-1

FIGURE 2: MODEL CONFIGURATIONS

18-11

-7.0

-6.0

m -5.0 s

+ 1 '° 0 1.0 2.0 3.0 4.0 5.0 6.0

Reduced Velocity - V/NB

FIGURE 4: FLUTTER DERIVATIVE FOR VERTICAL - ANGLE OF ATTACK = 0°

18-13
Angle of Attack - Degi-ee*

FIGURE 6: CRITICAL FLUTTER VELOCITY -

1% CRITICAL DAMPING

18-14
025

0 2.5 5.0 7.5 10.0 12.5

R. M. S. Aeo«l«ration - XG

FIGURE 8: ESTIMATE OF PROTOTYPE ACCELERATIONS

RESULTING FROM VERTICAL VORTEX

EXCITATION - ANGLE OF ATTACK = 0

. 6

. 1

C-2C-A

12 0 +4

Angl« Of Attook - •gneea

FIGURE 9: DRAG COEFFICIENTS

18-15

19

A TECHNIQUE FOR MEASURING FLUCTUATING WIND LOADS

ON A TALL-BUILDING MODEL IRRESPECTIVE OF MODEL

MOTION

Timothy A. Reinhold

National Bureau of Standards

Washington, D.C.

A Technique for Measuring Fluctuating Wind Loads on a Tall-Building Model Irrespective of Model Motion

Models used to determine resultant fluctuating loads on structures

can be divided into two categories. The first category includes

aeroelastic and elastic models for which loads are determined from

the movement of the model. The second category involves direct

measurement of wind pressures. Direct measurements of wind pressures have been made since the earliest wind tunnel model tests

and involve a model with a number of pressure taps on its exterior

surface. Pressure measurements made at each tap are usually

analyzed to determine mean values, root mean-square (RMS) values,

and possibly spectra. Cross-correlation betwen pressure measurements at two points may also be obtained to determine the correlation of the pressure fluctuations over the model surface. However,

it has been extremely difficult to use pressure measurements to obtain information about resultant fluctuating loads in a form

that is useful to a design engineer. Consequently, aeroelastic

or elastic models, which require measurement of the motion of the

model, are used in practically all tests where information on the

fluctuating motion or the resultant loads is sought.

Aeroelastic and Elastic Models

Aeroelastic models constitute the most widely recognized type of

nodel for use in determining the motion of a building and in

determining resultant fluctuating loads acting on the building as

a whole (base bending moments and base shear). The requirements

for simulation of building response to wind loads using aeroelastic

nodels have been well established and are discussed by Whitbread [1].

The nondimensional parameters which must be modeled have been listed

often in the literature, for example, see references [1 and 2],

and will not be reproduced here. The important observation relative

to these parameters is that a number of structural properties of

the prototype building must be modeled to simulate the dynamic

response of the structure to wind loading. The problem associated

'with attempts to satisfy the similarity requirements are discussed

Dy Whitbread [1] and Scruton [2], who have shown that construction

of an aeroelastic model which satisfies rigorously all the similarity requirements is generally impossible. A common simplification

I involves use of a semi-rigid model, which incorporates a gimbal

j pnd a set of springs at its base, to simulate the motion of tall

juildings where the predominant mode of vibration is linear [3 and 4]

.

Aeroelastic models can be used to measure both mean and

dynamic components of the building deflection or base moment along

j:wo orthogonal axes (usually chosen perpendicular to the faces of

i rectangular prismatic building). However, even with expensive

ind elaborate aeroelastic models, it is difficult to account for

: :orsional movements which have been shown to be an important compolent of the dynamic response of a building, even for symmetrical

mildings [5]. Tests must be conducted for a number of different

find speeds at various angles of attack, and structural parameters

lust be varied. The structural properties, including stiffness,

lamping, and mass distribution, must be varied in the test to

oracket the possible values that may ultimately exist in the prototype. Measurements of the dynamic characteristics of existing

mildings [6] have shown that estimates of stiffness and damping

based on design calculations and judgment) may be extremely

,.naccurate. Furthermore, application of the results is usually

'.united to the particular building being designed because of specific modeling of the range of frequency and damping parameters [7].

19-1
A second type of model that relies on measurements of motion

to determine the loading is the elastic model. These models are rigid members mounted on elastic foundations and allow measurement

of the so-called quasi-static force spectrum (i.e., the spectrum

of actual wind loads), which is independent of the natural frequency and damping of the model [7]. The spectrum is quasi-static

because it does not reflect the amplification of the loads due to resonance. For example, while aeroelastic models may have natural

frequencies of a few Hertz, the model used by Whitbread [7] had

a natural frequency of 450 Hz and the spectrum of loads was determined for the frequency of 0 - 250 Hz. Models such as this are too stiff in the torsional mode to allow measurements of

twisting moments.

Determination of the spectra of wind loads has been attempted

for aeroelastic models by measuring the displacement spectrum and

determining the mechanical admittance function for the model. The

mechanical admittance is a function of frequency and represents

the ratio between the spectra of the amplitude of deflection and

load. For a building, a modal approach is used and the response

of the building in each mode is expressed as a single degree-offreedom system [8]. Normally first-mode (linear) response is

assumed and the corresponding first-mode mechanical admittance

function is determined. Saunders and Melbourne [9] used this

technique to determine normal force spectra for several models.

All of the techniques described can be used to obtain only

the resultant fluctuating loads on the building as a whole and it

is not possible to determine the manner in which the wind loads

are distributed over the surface of the building. A more complex

model, which yields information on the nature and distribution of

the wind loads throughout the building height, has been investigated by Ellis [10]. The technique used a multi-force transducer

composed of 6 elements to measure dynamic strains and acceleration at various levels throughout the building. Dynamic bending

moments are found from the output of calibrated strain gages; and

inertia forces are found from the measured accelerations and known

mass distribution of the model. Random loads on each element are determined from these strains and accelerations. The model can

not be used to measure fluctuating twisting moments.

Direct Pressure Measurements

Mean forces and moments on buildings have often been determined

by integrating mean pressure measurements made at numerous points over the entire building [11]. However, this procedure has not been used previously to determine resultant fluctuating loads.

Ellis [10] noted that fluctuating loads at various levels throughout a building could be determined by using numerous pressure

transducers. However, he rejected the method because of the high

cost of the large number of transducers needed and because subsequent data gathering and integration would be expensive and

cumbersome

.

The analog manipulation of the data performed in the present

investigation streamlined the data gathering and integration

procedures. Three analog signals were produced to represent the

resultant fluctuating load components on a discrete segment of the

building. Consequently, analysis of the load records to determine

load spectra was no more difficult than analysis of turbulent

velocity records. The transducers did represent a significant

initial investment but the cost of all parts, including printed

19-2
circuit boards, used in the electronic network for signal condition

ing and integration was less than the cost of three transducers.

The transducers and electronics can be readily used in future model

tests

.

The technique can be used to determine fluctuating forces

simultaneously at several levels throughout the building height.

Twisting moments can be determined in addition to normal forces

for each level. Consequently, it is possible to determine both

the nature and distribution of fluctuating normal forces and

twisting moments throughout the height of the model. Furthermore,

since the loads are measured simultaneously, it is possible to sum the loads at various levels, weighted by a mode shape, to determine the modal forces on the building. These modal forces

can then be used together witn an appropriate admittance function

to calculate the building response.

Test Program

Environment

A series of model tests, which employed numerous transducers to measure directly resultant fluctuating pressures, were carried

out at Virginia Polytechnic Institute and State University.

Square prisms with sharp edges were used in the wind tunnel

model tests. The models fulfilled the requirements of having

simple geometric shape and of having relative dimensions (heightto-width ratio) similar to realistic structures and to previously

tested aeroelastic models. Experiments were conducted in a 1.8 m

square by 8.0 m long test section of a low-speed wind tunnel.

Models were placed in a turbulent shear layer which was designed

to simulate the atmospheric boundary layer over an urban area.

The turbulent shear layer was developed over the test section

floor using a combination of spires at the entrance of the test section and 6.1 m fetch of 102 mm cubic roughness elements.

A detailed description of the mean flow and the three turbulence components for the wind tunnel shear flow, together with a comparison of the flow properties with those of the atmospheric

boundary layer over urban terrain, are contained in references [12

and 13] . It is sufficient for the purposes of this discussion to note that the mean velocity profile followed the power law with

a power index of 0.37 (corresponding to flow over urban terrain)

and that the turbulence properties were found to compare reasonably

well with full-scale estimates at a geometric scale ratio ranging

from 1/400 to 1/600 [12 and 13].

Models

Two models of identical geometry (76.2 mm x 76.2 mm x 635.0 mm)

J were used. One model was fitted with 96 pressure taps, connected

by flexible tubing to two Scanivalve systems. This arrangement

was used to obtain mean pressure measurements because the flush

mounting pressure transducers (Kulite model COH-125-5) often

exhibited a slow voltage drift. The second model was fitted with

|! the flush mounting pressure transducers, having an exposed

diaphram of 2.16 mm diameter. The transducers were used to measure

fluctuating pressures at the same locations as the pressure taps.

The pressure taps and transducers were located in groups of 16 about the perimeter of the model, with 4 transducers on a side,

at 6 elevations throughout the model height.

19-3
Selection of the locations for pressure taps and transducers

was based on results of experiments conducted using a sectionmodel of square cross-section placed in a two-dimensional uniform turbulent flow. Tests were conducted for four different

orientations of transducers on a single model face. The configurations used are illustrated in Figure 1 and range from the use of seven transducers to the use of three. The signals from

the transducers were summed to obtain the resultant normal force on the model face and the resultant twisting moment about the

center of the model face. The objective was to determine the

positioning across the face of no more than four transducers on a side, which would closely reproduce the loads measured using seven transducers on a side. A comparison was made of both RMS values

and autocorrelation functions for the normal forces and twisting

moments measured with the four configurations of transducers shown

in Figure 1. Measurements were made at a number of orientations

of the model face to the flow described by the angle of attack,

a, in Figure 1.

The positions of the transducers adopted for the boundary

layer model tests are shown in Figure 2. This configuration

reflects the observation that the outer transducers had to be as close to the edge as possible. It was necessary to offset the

outer transducers to accomplish this. The maximum estimated error obtained by using the four tranducers on a side as opposed to seven is less than 5 percent for the normal force and less than

15 percent for the twisting moment associated with any wind

direction. More detailed information concerning the placement of

transducers can be found in reference [2].

Determination of Resultant Fluctuating Loads

Accumulation of a large mass of data representing simultaneous

instantaneous pressures on each transducer was avoided by performing an analog summation of the transducer output signals to obtain

single signals representing the resultant fluctuating loads. The

electronics necessary to perform this task consisted of two units.

The transducer output signals had to be amplified to obtain uniform pressure sensitivity weighted by the area factor associated

with the transducer location. Finally, these signals had to be

added and subtracted in proper combinations. Moment arm factors

had to be introduced in order to obtain a signal representing the

resultant fluctuating moment. Figure 3 outlines the network of

electronics used to obtain the resultant fluctuating loads.

Each transducer had its own pressure sensitivity and surface

pressure fluctuations produced output voltage fluctuation of less

than one millivolt peak to peak. Consequently, a differential

amplifier capable of producing gains of between 500 and 2,500 was connected to the output of each transducer. A schematic diagram

of the amplifiers is shown in Figure 4. Design helps for such

items are available in books on operational amplifiers such as the one by Graeme, Tobey and Huelsman [14]. The amplifiers contained

two stages because it is difficult to obtain both high common

model rejection and high gain using a single operational amplifier.

The first stage provided a gain of 10 and the common mode rejection

necessary to remove noise, primarily 60 Hz generated by fluorescent

lights and equipment which was picked up by the wires carrying the

signal from the transducer to the amplifier. Two conductor shielded

cables were used in transmitting signals to ensure that most noise

picked up would be common to both leads. The second stage provided

a variable gain of between 50 and 250. A variable DC voltage

offset feature was available on the first stage of the amplifier.

19-4
The offset adjustment feature allowed any DC voltage produced by

the transducer to be removed so that the average output voltage

of the amplifier could be set to zero.

The amplifier gain was set so that the output signals of all

transducers located at one level represented identical pressure

sensitivities weighted by the proper area factor. For example,

if identical differential pressures were applied to transducers 1 and 2 in Figure 2 then the ratio of the changes in output voltages

for the transducers would equal the ratio of the area factors.

Signals were added and subtracted using inverting summing

amplifiers. A schematic of the network used to sum the signals

is shown in Figure 5. Precision resistors were used in these networks to ensure accuracy since the equation for determining gain

of these amplifiers involves only the input resistance and the

feedback resistance. Offset of the operational amplifier output

voltage was reduced by matching the resistor between the noninverting input and ground to the parallel equivalent resistance

of the input resistors and the feedback resistor.

Four summing amplifier networks were used to produce the

resultant fluctuating loads at each level. One network was used

to produce each of the normal loads. Two networks were needed to obtain the moment. By referring to Figure 2, the details of the

procedure can be easily explained. F^ was produced by connecting

tranducers 1, 2, 3, and 4 to inputs A, while transducers 9, 10, 11,

and 12 were connected to inputs B. Fy was produced in a similar

manner using a second network and connecting transducers 13, 14,

15, and 16 to inputs A, and transducers 5, 6, 7, and 8 to inputs B. The moment was produced in two segments. The first segment, called

Ml, was obtained by connecting transducers 1, 5, 9, and 13 to inputs A, and transducers 4, 8, 12, and 16 to inputs B of the third

network. The second segment, M2, was produced by connecting transducers 2, 6, 10, and 14 to inputs A, and transducers 3, 7, 11, and

15 to inputs B of the fourth network. The signal from segment Ml was multiplied by a factor of 1.0 and the signal from segment M2

was multiplied by a factor of 0.384 to express the relative lengths

of the associated moment arms. These products were then added

using an Electronics Associates, Inc., Mini AC analog computer to obtain the resultant moment. The output voltages were multiplied

by appropriate factors to obtain the loads in engineering units.

The actual length of the moment arm was incorporated in the factor

for the moment.

Sample Results

Resultant normal forces and twisting moments were determined at each of the 6 levels at which instrumentation was placed. The

levels were located at 0.16, 0.32, 0.48, 0.64, 0.80, and 0.96 of

the model height, measured from the tunnel floor upward. Because

of a limited number of transducers, it was only possible to instrument three levels simultaneously. Consequently, three tests were conducted. Test 1 had transducers at 0.64, 0.80, and 0.96 of the

model height (levels 4, 5, and 6). Test 2 had transducers at 0.32,

0.64, and 0.96 of the height (levels 2, 4, and 6), and test 3 had

transducers at 0.16, 0.48, and 0.64 of the height (levels 1, 3,

and 4). This combination of tests allowed measurements of the

fluctuating loads to be made a level 4 (0.64 of the height) in each

test. Consequently it was possible to determine the correlations

of resultant fluctuating loads between all levels and level 4. RMS

values of the resultant fluctuating loads were measured as the

tests were conducted and the signals representing the loads were

19-5
recorded on a 14 channel FM tape recorder. The records were subsequently analyzed to determine the power spectra of the loads and

the cross-spectra between loads at different levels.

I Examples of the detailed information concerning the nature and

distribution of the loads which could be obtained from analysis of

the wind load measurements are illustrated in Figures 6 through 9. The loading illustrated is the cross-wind component for a single

tall building with one face normal to the mean flow. Normalized

power spectra obtained at each of the six instrumented levels are shown in Figure 6. The shift in frequency of the vortex shedding

with height is clearly shown. Figures 7 and 8 together illustrate

the relationship between the loads at level 4 and each of the other

levels. The coherence functions, Figure 7, indicate the degree * of correlation between the loads while the phase shift, Figure 8,

contains the information on the relationship of the timing of the

loads. From observations of the cross-correlation functions computed between the load components, it was determined that the maximum cross-correlation generally occurred when the signal from the

upper of the two levels was delayed in time with respect to the

signal from the lower level. This shows up as a negative phase

shift between level 4 and level 5 or level 6 and as a positive

phase shift between level 4 and the lower levels, Figure 8. Consequently, the load fluctuations applied on the upper levels of

the model propagated downward. The double lines in Figure 8 correspond to the range of frequencies where the coherence function

was greater than 0.2.

Estimates of the modal forces were also obtained by using data

from test 1. The signals representing the forces at levels 4, 5,

and 6 were scaled to represent their relative magnitude, the areas

over which they were assumed to act and a linear mode shape factor

to represent the first mode of vibration of a tall building. The

signals were summed by passing them through an analog computer

circuit. The resultant signal represented the modal force and was analyzed to obtain the modal force spectrum and the RMS value of i the modal force was also determined. These modal force estimates

only incorporated forces which acted over roughly the top half of

the model. It is estimated that the maximum error obtained from

neglecting forces at the lower levels in determining the modal

forces and torques is less than 30 percent. The nondimensional

modal force spectrum for the cross-wind forces on the single model J with one face normal to the mean wind direction is shown in S Figure 9. The magnitude of the RMS value of the modal force can I be determined from the following equation:

( -2 1 F

y

= 1/2 p UH

BH C£ (1)

where is the RMS modal force coefficient and was found to be

y ; 0.118 for the cross-wind force.

i Comparison of Results

The true test of the validity of model test results is a comparison with results of full-scale measurements. In the absence of i full-scale measurements for a similar building, results are compared with those obtained from aeroelastic tests where models

of similar geometric shape were used. Tests of an aeroelastic

model with reasonably similar geometry were conducted by Rosati at the University of Western Ontario [15]. The only difference in

the geometric shape of the models is that the Rosati model was <

19-6
shorter (aspect ratio of 7.0 instead of 8.33). Measurements of the

response of the aeroelastic model were made in a flow with similar

properties to the one used in this study for a wide range of

velocities.

It is possible to use the modal force spectrum shown in

Figure 9 together with an appropriate mechanical admittance function to predict the RMS level of response of the Rosati model.

The necessary equations are presented in reference 16. Assuming

that the response is first mode with a linear mode shape, the equation for the response spectrum is given by:

Gy(n)

=H^ G- („) (2)

k

F y where

and

H

2 (n) = I <3> 21 , 2 (1 - + (hR)

n

2 n

0

H K = (2irn 0

) 2

\ m p 2 dz. (4) r n J 0

1

The RMS level of response, o^, is given by:

fJ 0

a

2 = G (n) dn. (5)

Predictions of the response of the Rosati model were made using

these equations and are compared with the actual response measurements reported by Rosati [15]. The comparison is shown in Figure 10. The open symbols represent measured response and the solid

symbols represent the predicted response. The agreement is within

25 percent for all cases up to reduced velocities of about 15. At high values of reduced velocity the predictions become considerably

lower than the measured values for low values of structural damping.

The differences may be attributable to differences in the properties

of the shear layers used in the tests or to aeroelastic effects,

i.e., aerodynamic damping. Negative aerodynamic damping would cause

greater building response and would be especially noticeable when

the structural damping is low. There are some indications that use

of more levels of transducers in determining the modal force would

lead to a reduction in the difference between the measured and predicted levels of building response, at least up to values of reduced

velocity of about 15.

Conclusions

The multiple transducer technique described in this paper provides

a method for directly determining the resultant dynamic wind loads

on a building. The technique can be used to obtain detailed information concerning the distribution of the loads over the surface

of the model as well as the characteristics of the loads at various

levels. Twisting moments can be readily obtained.

Forces at various levels can be integrated to determine modal

forces. Results of response predictions for first mode (linear)

19-7
agree reasonably well with response measurements of similar aero- elastic model test for values of the reduced velocity of practical

interest in design. Other mode shapes could also be used since

information concerning the distribution of the loads over the surface of the structure is available.

The experimental work reported here shows that it was possible

to use analog computer circuits to reduce the data handling to a manageable scale with very little expense.

Since the transducers provide measurements of wind loads which

are obtained independently from measurements of model motion, the

transducers can also be used in conjunction with aeroelastic models.

Use of transducers on an aeroelastic model could lead to a better

understanding of body-motion related forces.

Acknowledgments

The experimental work and some of the data analysis reported in

this paper were supported by the National Science Foundation,

Grant No. ENG-75-05327 A01. Further data analysis was conducted by

the author under the NRC/NBS Postdoctoral Research Associateship.

The author also wishes to acknowledge the guidance of Professor

F. J. Maher and H. W. Tieleman of Virginia Polytechnic Institute

and State University during the course of experimental work.

References

1. Whitbread, R. E. , "Model Simulation of Wind Effects on Structures," Proceedings of the Conference on "Wind Effects

on Buildings and Structures ," National Physical Laboratory,

England, 1963, pp. 284-302.

2. Scruton, C, "Aerodynamics of Structures," Proceedings of

International Research Seminar, "Wind Effects on Buildings

and Structures ," Ottawa, Canada, 1967, pp. 115-161.

3. Davenport, A. G. and Isyumov, N. , "The Application of the

Boundary Layer Wind Tunnel to the Prediction of Wind Loading,"

Proceedings of International Research Seminar, "Wind Effects

on Buildings and Structures ," Ottawa, Canada, 1967, pp. 201-230.

4. Robertson, L. E. and Chen, P. W. , "Application of Design of

Research and Wind Effects," Proceedings of International

Research Seminar, "Wind Effects on Buildings and Structures ,

Ottawa, Canada, 1967, pp. 83-113.

5. Jeary, A. P. and Sparks, P. R. , "Some Observations on the

Dynamic Sway Characteristics of Concrete Structures," paper

presented at ACI Symposium on Vibration of Concrete Structures , New Orleans, October 1977.

6. Jeary, A. P., Lee, B. E. , Paquet, J., De Souza, V. C. M. and

Sparks, P. R. , "The Dynamic Behaviour of the Arts Tower,

University of Sheffield," paper presented at the Internat ional

Conference on Slender Structures , City University, London,

September 1977.

7. Whitbread, R. E. , "The Measurement of Non-Steady Wind Forces

on Small-Scale Building Models," Proceedings of the Fourth

international Conference on "Wind Effects on Building and

Structures ," Heathrow, England, 1975, pp. 567-574.

19-8
8. Wyatt, T. A., "The Calculation of Structural Response," Proceedings of the Seminar "The Modern Design of Wind-Sensitive

Structures," Construction Industry Research and Information

Association, London, 1971, pp. 83-93.

9. Saunders, J. W. and Melbourne, W. H. , "Tall Rectangular Building Response to Cross-Wind Excitation," Proceedings of the

Fourth International Conference on "Wind Effects on Buildings

and Structures ," Heathrow, England, 1975, pp. 369-379.

LO. Ellis, N. , "A New Technique for Evaluating the Fluctuating

Lift and Drag Force Distributions on Building Structures,"

Proceedings of the Fourth International Conference on "Wind

Effects on Buildings and Structures ," Heathrow, England, 1975,

pp. 527-536.

.1. Cermak, J. E., "Wind Tunnel Testing of Structures," Proceedings ASCE/EMD Specialty Conference, "Dynamic Response of

Structures: Instrumentation, Testing Methods and System

Ident i f icat ion , " March 30 and 31 , 1976, University of

California Los Angeles.

2. Reinhold, T. A., Measurement of Simultaneous Fluctuating Loads

at Multiple Levels on a Model of a Tall Building in a Simulated

Urban Boundary Layer , Thesis (Ph.D.) Virginia Polytechnic

Institute and State University, November 1977.

3. Reinhold, T. A., Tieleman, H. W. and Maher F. J., Wind Tunnel

Simulation of Urban Winds for the Study of Tall Buildings , Report No. VPI-E-77-12, College of Engineers, Virginia Polytechnic Institute and State University, April 1977.

4. Graeme, J. G. , Tobey, G. E. and Huelsman, L. P., Operat ional

Amplifier Design and Applications , McGraw Hill, New York,

(1971).

5. Rosati, P. A., The Response of a Square Prism to Wind Load , Report No. BLWT-II-68, Faculty of Engineering Science, The

University of Western Ontario, London, Canada, March 1968.

5. Engineering Sciences Data Unit, "The Response of Flexible

Structures to Atmospheric Turbulence," ESD Item No. 76001,

London, England, September 1976.

otation

Width or depth of square prismatic model - mm

RMS modal force coefficient for the cross-wind

force component

Fluctuating cross-wind force component - N

RMS modal force for cross-wind force component - N

Variance of fluctuating cross-wind force component - N

2 Cross-wind displacement spectrum - m s

2 Cross-wind modal force spectrum - N s

: y (n)

: (n)

19-9
H Height of model - mm

H (n) Mechanical admittance function

K Modal stiffness - N/m

m Mass per unit length of model - Kg/mm

n Frequency - Hz

Hq Natural frequency for a particular mode of vibration - Hz

Ug Mean wind velocity at the gradient height - m/s

UH Mean wind velocity at the top of the model - m/s

Z Height above ground level - mm

a Angle of attack - deg.

y Structural damping, expressed as fraction of critical damping

Mode shape function for first mode = Z/H

3

p Density of air - Kg/m

Oy RMS displacement of top of model - mm

19-10
5 O iJW H O 2 O W 2 I Z

-1 § W M Q Q O I 2 en

In OS O O zoMHUw

CO—W CO H OS z w w u 2 3 5 Q

OS CO

OSH Z H

OS Z- (E- U-i * O I

CO

CO OS H W

CO CJ W 3 H Q

CO ^ z W <£ o od O H 2

SB

O

OH

H

H Z U W W 2

CO

w- <

CO

M Z 2 w os 2 w M H Q W I Q W

OSO

19

19-11
OUTPUT

SIGNALS

FROM

16 TRANSDUCERS

14 CHANNEL

FM RECORDER

Record Slgnoli lor Futurt Analylii

HIGH GAIN DIFFERENTIAL

AMPLIFIERS

Uniform Stnlltlvlty Arto Foclori

SUMMING

AMPLIFIERS

Signal ln«*rtlon Summation In Vorlout Configuration!

M'

OSCILLOSCOPE

Monitor Signolt

1 Original!

2 Topsd

ANALOG

COM - PUTER

Mom*

Factors

— RESULTANT

"""* DYNAMIC

LOADS

SOLARTRON

REAL TIME ANALYZER

Stabl* Tim! A«*rog«i

of RMS Valuit

FIGURE 3: DETERMINATION OF RESULTANT DYNAMIC LOADS FOR ONE LEVEL

FIGURE 4: SCHEMATIC OF HIGH GAIN DIFFERENTIAL AMPLIFIER - GAIN

500 - 2500

19-12
FIGURE 5: SCHEMATIC OF SUMMING AMPLIFIER NETWORK FOR ONE UNIT

OF TWELVE

.12 r

WIND DIRECTION

30 40 50

FREQUENCY - HERTZ

60 70 80

FIGURE 6: NORMALIZED POWER SPECTRA FOR F' AT 6 ELEVATIONS,

CONFIGURATION 0

19-13
— 4-1 — 4-5

4 -2 4 - 6 4 - 3

1 1

I 0 K> 20 30 40 50

FREQUENCY - HERTZ

FIGURE 8: PHASES ASSOCIATED WITH COMPLEX CROSS-SPECTRA FUNCTIONS

AND COHERENCE BETWEEN LEVELS FOR F* CONFIGURATION 0

19-14
500

20.0

100

5 0

2 0

I 0

0.5

0.2

0 I

MEASURED RESPONSE

ROSATI REF [15] 0 28%

0 92%

4 54%

12 9%

PREDICTED RESPONSE FROM PRESSURE MEASUREMENTS

10 20 30

FIGURE 10: COMPARISON OF PREDICTED AND MEASURED RESPONSE OF AN AEROELASTIC MODEL

19-15

20

A STUDY OF BUILDING DAMAGE CAUSED BY WIND

James R. McDonald

Patrick A. Lea

Institute for Disaster Research

Texas Tech University

Lubbock, Texas

Prepared for

The Veterans Administration

NOTE: TABLES NOT AVAILABLE FOR THIS PAPER
!i3IEy
A Study of Building Damage Caused by Wind

Each year more than $800 million in property damage is caused by

wind forces from tornadoes, hurricanes, and extratropical storms in

the United States. In addition, there are thousands of injuries

and the loss of hundreds of lives. Much of the property loss and

many of the deaths and injuries might be mitigated by changes in

building design for resistance to wind-induced forces. Recognizing the potential of property damage and the need for protection

against injury and loss of life, the Veterans Administration (VA)

has implemented a program to evaluate its present design standards,

and, where necessary, to develop new design criteria and construction standards for use by A/E's and VA personnel in the design of

VA facilities. The need for new design criteria can best be established from a historical survey of wind damage and casualties to hospitals and major buildings in the United States. From the

historical survey those construction subsystems which are most

susceptible to wind damage can be identified. With the knowledge

of how construction subsystems behave in windstorms, a methodology

also can be developed for evaluating the wind resistance potential

of existing VA hospital facilities.

The purpose of this report is to present the results of a historical survey on the kinds of significant damage caused by

wind forces to hospitals and major buildings in the United States.

The survey was conducted by members of the Texas Tech University

Institute for Disaster Research Team, which, in the past seven

years, has surveyed building damage in more than 30 windstorm

events. The survey team is made up of structural engineers who

are familiar with wind design criteria and standard construction

pract ices

.

From results of the historical survey, evaluations of structural, architectural, mechanical, and electrical subsystems are presented relative to their performance in wind storms and the

consequences of their failure on hospital operations. Based on the

data compiled during the historical survey, a questionnaire was developed for surveying existing VA hospital buildings to assist in

establishing potential for wind damage. By making an inspection of

a faculty and completing the questionnaire, a trained engineer can make a rational judgment as to how the facility will perform in a .windstorm. An evaluation of a small private hospital in West Texas

is presented as an example of the methodology and as an illustration

of the use of the questionnaire.

Windspeed data for each of 185 VA hospital sites were assembled

from storm records available from the National Oceanic and Atmospheric Administration. Tabulated in Table 5 for each VA site is

the maximum annual extreme fastest-mile windspeed and the windspeeds

associated with 50- and 100-year mean recurrence intervals.

The information from the historical survey, evaluations of the

instruction subsystems, and use of the questionnaire provides the

Information needed by VA personnel for developing new design criteria, if needed, and for assessing the condition of existing facilicies with regard to their resistance to windstorms.

listorical Survey

[he purpose of the historical survey is to make an assessment of

he performance of hospital buildings in the United States that

;iave been subjected to severe windstorms, including tornadoes,

20-1
hurricanes, and straight-line winds. Tabulated in the survey are five basic types of information:

(1) Type of damage to architectural, structural, electrical, and

mechanical subsystems

(2) Outages caused by windstorms, including power, water, sanitary

and gas

(3) Associated damage by flooding, rain, hail, slides, and storm

surge

(4) Fire, if related to the windstorm incident

(5) Number of casualties, including cause of injury and death

(6) Impact on building operation.

Because relatively few hospitals have suffered major damage in

windstorms and in order to consider a variety of construction features, major buildings, other than hospitals, are included in the

survey

.

Four basic data sources were used:

(1) Windstorm damage documentation files at Texas Tech University

and other institutes engaged in storm research

(2) Storm Data , a publication of the National Oceanic and Atmosphere Administration, Environmental Data Service, Asheville,

NC

(3) Technical literature, including published papers and reports

from both the meteorological and engineering professions

(4) Magazine and newspaper accounts of storm characteristics and

damage

.

The searches have been thorough within the limits of the

proposed efforts. However, no guarantee can be made that every

significant hospital or major building damage has been reported.

Windstorm damage documentation files were searched for hospitals and major buildings which either experienced windstorm damage

or which survived an extreme windstorm without significant damage.

Questionnaires were sent to individuals and organizations engaged

in windstorm damage research, requesting information on hospital

damage. Table 1 includes a list of those contacted. Written or oral responses were received from eight of the nine groups. Except

for Bishop Bergen-Mercy Hospital in Omaha, NB, no other major hospital damage could be recalled by the groups.

The windstorm records in Storm Data were searched for references to hospital damage. This monthly publication lists each

windstorm incident in the United States, along with a verbal

description of damage. The search proved to be arduous and time

consuming. A total of eleven hospitals and three nursing homes

were identified from the Storm Data listings in the period 1973-

1976. Individuals at the hospitals were contacted by phone and

a questionnaire was sent requesting specific information regarding

the storm characteristics, the building description, and the extent

of damage. Unfortunately, the questionnaire did not prove to be

20-2
an effective means of getting information. Follow-up telephone

calls were needed to get action and to get the specific information

reques ted

.

Specific geographic areas known to have been hit by tornadoes,

hurricanes, or straight-line wind events were examined for possible

damage to hospitals. Once the affected hospitals were identified,

authorities at the hospitals were contacted by telephone to obtain

detailed information. This approach was very effective in obtaining information on hurricanes.

Once all of the available information on hospitals was assembled, other major buildings were selected for study, because of

the need to complete the windstorm damage data matrix with respect

to type of windstorm and type of structural system. Information

similar to that tabulated for hospitals was obtained for each

building. Preference was given to convalenscence homes, special

activities buildings, and nurses' quarters because of the association or similarity to hospital buildings.

The findings of the historical survey are summarized in

Table 2. This table is referred to as the Windstorm Damage Data

Matrix. The information furnished is divided into four categories:

(1) Building Identification and Description

(2) Windstorm Identification and Description

(3) Building Damage Information, which includes damage to architectural, structural, electrical, and mechanical subystems

(4) Ancillary Information, which includes outages, associated

damage, fire, casualties and impact on operations.

Sixteen separate buildings or additions are included in the

survey. Eight buildings were hit by tornadoes, five were affected

by hurricanes, and three were victims of straight-line winds.

Tornado intensities ranged from weak to one of the most intense

on record. The two specific hurricanes (Camille and Celia) were

chosen because the damage attributed to the buildings was primarily

due to wind. The three buildings affected by straight-line winds

are typical of the type of damage caused by straight-line winds.

Substantial structures that have received adequate engineering

attention do not generally suffer major damage from straight-line

rfinds. The windstorms cited had windloads within plus or minus

20 percent of local building code windload criteria.

Among the damage to architectural subsystems, which includes

windows, nonload bearing exterior walls, interior partitions, paraoet walls, canopies, covered walkways, fascia and roofing material,

yindow damage appears to be the most significant. Window damage

is caused by a combination of wind pressure and debris impact.

Thus, upwind conditions, such as gravel roofs, and window glass

lesign considerations affect the performance of window glass in

windstorms. Broken windows in windward walls allow the wind to get

.nside a building. This situation creates additional outward act-

.ng pressures and creates havoc with interior partitions and the

:ontents of the building.

Little damage to electrical subsystems was observed except

in the case of complete collapse of the building. Mechanical

quipment located on the roof is particularly susceptible to wind
damage. Cooling towers, compressors, vents and intakes should be

protected by some type of substantial wind screens or a penthouse.

The most common outage was electrical power. Power failure

normally occurs outside of the hospital facility. Power failure

to pumps can result in reduction or elimination of water pressure

and the failure of sanitary systems. This study emphasized the

importance of adequate auxiliary electrical generating equipment.

The most common type of associated damage was rain. Carpets,

draperies, furnishings, and medical equipment are damaged or ruined by water. The point should be emphasized that if the

roof or exterior cladding is breached, significant damage can occur to the interior of a building.

Fire was not a factor in any of the buildings studied. And

the number of casualties was small -- chiefly because most hospitals

are tied into some type of windstorm warning system, emergency plans

are formulated and practiced, and interior rooms, hallways, or basements provide occupant safety. The impact on building operations

depends on utility outages and extent of damage.

The historical survey shows that adequate engineering attention to architectural, structural, electrical, and mechanical

subsystems result in minimum damage and interference with building

operations. If deficiencies in existing buildings can be identified prior to a windstorm, retrofit or redesign may be possible.

The ultimate objective of this research project is to develop a questionnaire for surveying VA building sites in order to identify

wind damage potential. The findings of the survey and its subsequent evaluation will be used by the VA as a basis for making necessary corrections to existing buildings and to design improvements

into new buildings to reduce the probability of injury or property

loss caused by windstorms.

Assessment of Performance

In this section the performance of the various construction subsystems identified in the historical survey are assessed. Included

in the evaluations are discussions relating to strength, performance

criteria, and consequences of failure.

For purposes of this report construction subsystems are divided into two classifications: structural and architectural.

The structural subsystem includes all load carrying elements of the

building. All nonmechanical and nonelectrical components that do not function as load carrying members are considered to be part of

the architectural subsystem. The architectural subsystem classification, thus, goes beyond what is normally considered to be architectural elements to include windows, exterior nonload bearing

walls, interior partitions, parapet walls, canopies, covered walkways, facia details, and roofing materials.

The strength of a structural subsystem is defined in terms of

subsystem's ability to withstand windloads without compromising the

integrity of the structure. The structural subsystems of the

buildings included in the historical survey can be placed in five

categories. Table 3 lists the categories and the names of the

buildings or additions in each one. Each structural subsystem is

discussed in the paragraphs below.

20-4
Cast-In-Place Reinforced Concrete Frame

The components of this structural system include the columns, the

beams, and the floor and roof slabs. The primary advantage of

this type of system is its inherent rigidity, which is produced

by strong connections and diaphragm action of the floor or roof

slabs. Lateral deformations due to windloads are extemely small.

None of the buildings surveyed suffered any structural damage to

the concrete frame itself. Where extensive damage to a building

with a reinforced concrete frame did occur, it was because of

failures of wall cladding and/or removal of roofing material.

Steel Frame with Nonload Bearing Walls

This type of structural subsystem has considerable reserve

strength, although lateral deflections may be excessive. The

Great Plains Life (GPL) Building survived the Lubbock, Texas

tornado of 1970 even though the average wind pressures were more

than double the 20 psf uniform wind pressure used in the original

design. In this case the building used additional stiffness from

[the exterior masonry walls and the rather stiff interior partitions. Older buildings such as the GPL have greater inherent

resistance from the nonstructural elements than a modern steel

framed building with metal curtain walls and lightweight moveable

interior partitions. Particular attention should be paid to lateral load resistance. If not provided for, excessive wind

drift will cause damage to interior partitions and plaster ceilings.

If the reserve strength of a moment-resistant frame is to be

realized, it must have adequate connections. Standard connections

bf riveted, bolted, or welded types have evolved over a long period

)f years that perform well under extreme loads. Such connections,

lowever, must have proper engineering attention, correct fabrica-

:ion (including quality control), and timely inspection. None of

:he connections or splices in the GPL Building failed.

The weak links in a building with a steel framed structural

iystem are the attachments of the nonstructural elements, specifially the curtain walls and the roof rather than the steel frame

tself. Three of the four buildings in this category suffered

xtensive damage because of wall or roof failure, even though

he steel frames proved to be adequate.

|

Steel Frame with Load Bearing Walls

his type of construction is typical of 1-story wings and additions to hospitals, or to support facilities such as maintenance

uildings, or storage facilities. Three examples were considered

n the historical survey (see Table 2). A typical 1-story strucure consists of metal deck on open web steel joists supported

m steel beams or load bearing masonry walls. Interior supports

ire furnished by steel wide-flange sections or pipe columns.

it>of overhangs of 2-4 ft. are not uncommon.

Two modes of failure were observed for this type of construcLon. In the case of the Ambulatory Care Unit, Bishop Bergen-Mercy

aspital, the windward wall collapsed because of inadequate vertial reinforcing steel in the wall. Uplift forces on the roof tend

) negate the axial compressive stresses in the wall due to dead

5ad, thus making the flexural tensile stresses in the wall due to

Lnd more critical. The other mode of failure, observed in Ward

imorial Hospital, occurred because of insufficient anchorage of the

20-5
the roof to the top of the wall. Roof uplift forces arose from

wind flowing over the top of the building, from stagnation pressures that develop underneath the overhang and from wind getting

inside the building. Roof members were anchored securely to a con- tinuous bond beam at the top of the wall. However, the bond beam

itself was not anchored. The only uplift resistance was provided

by the mortar joint underneath the bond beam. The joists remained

anchored to the bond beam, but the bond beam itself was lifted from

the top of the wall. In addition to the roof failure, pieces of the

bond beam became windborne missiles.

Precast-Prestressed Concrete Structural Subsystem with Load

bearing Walls

The observations concerning load bearing walls stated above also

applies to this structural subsystem. The primary cause of failure of the Xenia High School Addition "C" was collapse of the

load bearing walls on the windward side of the building. Precast

concrete members provide considerably more dead weight on the load

bearing wall than a roof consisting of steel deck. The additional

axial stresses are beneficial to the wall in resisting flexural

tensile stresses due to wind. However, if the wall does collapse,

the subsequent roof collapse is catastrophic.

Wood Frame Construction

This type of construction is more frequently used for apartment and

motel applications rather than hospital facilities. Some retirement

and convalescent facilities use this type of construction. The systems generally receive little or no engineering attention and as a result suffer considerable damage from straight-line winds, which

are little, if any, above building code requirements.

Roof overhangs and inadequate roof-to-wall anchorages were the

primary causes of failure in the two buildings studied. Tornado

damage to this type of construction often involves removal of an entire second story of the building. The incidence of occupant

injury or death is extremely high for this type of structure.

Strength of Architectural Subsystems

Strength as implied here can mean material strength or strength

with regard to anchorage of the architectural subsystem to the

structure.

Windows

As was pointed out in the conclusions of the historical survey,

window breakage is the major cause of damage to buildings in

windstorms. Associated damage from rain and the wind getting

inside the building is the main factor in the large dollar value

of damage.

Window glass has traditionally been designed only for wind

pressures. Recent investigations have revealed that impacts from

small pieces of debris such as roof gravel may be the major cause

of window glass breakage. Two buildings in the survey emphasize

the effects of debris on window glass and breakage. The Great

Plains Life building, located in a downtown area adjacent to num- erous gravel roofs, had extensive window glass breakage at the

lower floors. The Mississippi Power Company building is located

on an isolated beach front area. Very few windows were broken

20-6
in this building because of screens on the first floor windows and

because of the absence of small debris around the building.

Exterior Nonload Bearing Walls

Metal curtain walls and nonload bearing masonry walls are included

in this discussion. Metal curtain walls generally perform well

under windloads because most curtain wall systems must pass

certain wind and water leakage tests. Problems generally occur

with curtain walls because of inadequate or improper anchorages

to the main structural frame. Localized wind pressures at wall corners due to turbulence and oblique wind incidence angles place

additional requirements on wall anchors and reinforcement. Navarro

County Memorial Hospital was an example of this type of failure,

although in that case the consequence of failure was negligible.

Nonload bearing masonry walls must have adequate support on all

four sides. Parkwood High School is an example where support

details can be overlooked in the design and construction of such a wall. Control joints serve the purpose of retarding cracks in the

wall, but they also can be the weak link in the support system.

Interior Partitions

Interior partitions, being nonload bearing, do not always receive

the design attention required. Because wind can get inside a building, permanent partitions should have some type of positive anchorage on all four edges. Interior hallway partitions should receive

special attention, because the hallways may be used for occupant

protection. The consequence of interior partition failure, as was pointed out in the survey, is to make hospital room unuseable.

Parapet Walls

Parapet walls were damaged by wind forces and by impact from flying

debris (GPL building and Ward Memorial Hospital, respectively).

Sometimes the coping material, which can be either stone or metal,

is ripped off the wall and becomes flying debris. The wall itself

may collapse and cause damage to the roof.

Canopies and Covered Walkways

The consequences of these failures are minimal, although the repairs

may be expensive. The major danger is that the canopy material may

become airborne and strike other vulnerable areas.

Facia Details

Facia damage is usually of little consequence, with one possible

exception. If facia damage allows wind to get underneath the roof

deck, additional uplift pressures can cause the roof deck to pull

up and subsequently an entire roof section can be ripped off. The

same wind forces also will cause damage to suspended ceilings and

fluorescent light fixtures.

Built-up Tar and Gravel Roofs

Aside from the roof gravel becoming flying debris, the primary

problem with roofing material is inadequate anchorage to the roof

deck. In all too many instances there is no bonding of the felt

paper to the roof. The only thing that keeps the roof in place

is diaphragm action of the layers of roofing material. Removal

of roofing material occurred at almost all structures that were

included in the historical survey. Next to glass breakage, this

20-7
may be the single most severe problem with hosptials and other

buidings in windstorms.

Summary

Regarding structural subsystems, the following observations are made

based on the results of the historical survey:

(1) Cast-in-place reinforced concrete is the most suitable structural subsystem for wind resistance, because of its inherent

rigidity and strength.

(2) Steel frames perform satisfactorily, if wind drift is limited

and if adequate engineering attention is paid to momentresistant connections.

(3) Load bearing walls are dangerous, if vertical reinforcing steel

is not provided to resist tensile stresses due to flexure.

(4) Precast-Prestressed concrete subsystems used with loadbearing

walls have the same potential problems as those walls used

with steel structural subsystems.

(5) Wood frame construction rarely receives the engineering attention required for it to adequately resist extreme windloads.

With respect to architectural subsystems, two major observations are summarized:

(1) Window glass breakage from wind pressure and flying debris

is the most serious wind failure because of the associated

damage to the interior of a building.

(2) Inadequate bonding of built-up tar and gravel roofs to the

roof deck results in excessive damage to interior of buildings from rain.

One other observation that should be made is that from the

appearance of damage to a specific building, one cannot distinguish

the type of windstorm that caused the damage. Conversely, a welldesigned structure will adequately resist wind forces whether they

are created by tornadoes, hurricanes, or straight-line winds.

20-8
TABLE 1: ORGANIZATIONS AND INDIVIDUALS ENGAGED IN WINDSTORM

DAMAGE RESEARCH

TABLE 2: WINDSTORM DAMAGE DATA MATRIX

TABLE 2: WINDSTORM DAMAGE DATA MATRIX (CONTINUED)

TABLE 2: WINDSTORM DAMAGE DATA MATRIX (CONTINUED)

TABLE 2: WINDSTORM DAMAGE DATA MATRIX (CONTINUED)

TABLE 2: WINDSTORM DAMAGE DATA MATRIX (CONTINUED)

TABLE 2: WINDSTORM DAMAGE DATA MATRIX (CONTINUED)

TABLE 2: WINDSTORM DAMAGE DATA MATRIX (CONTINUED)

TABLE 2: WINDSTORM DAMAGE DATA MATRIX (CONTINUED)

TABLE 2: WINDSTORM DAMAGE DATA MATRIX (CONTINUED)

TABLE 2: WINDSTORM DAMAGE DATA MATRIX (CONTINUED)

TABLE 3: CLASSIFICATION OF STRUCTURAL SUBSYSTEMS

20-9

21

RECENT EARTHQUAKE-RESISTANT DESIGN METHODS FOR

DIFFERENT TYPES OF BRIDGE FOUNDATIONS IN JAPAN

Yukitake Shioi

Toshio Furuya

Michio Okahara

Yasuo Mitsuie

Public Works Research Institute

Ministry of Construction

Recent Earthquake-Resistant Design Methods for

Different Types of Bridge Foundations in Japan

This paper presents a survey of 31 examples of bridges built in

Japan. Their foundations are classified into four types: spread,

pile, caisson, and a multi-column. For each type, the authors

surveyed design methods and the methods used to set up models to study the interaction of soil and foundation.

Spread Foundations

Earthquake Calculation Method

Even in the case of a large-scale structure, model analysis by a simple system with two degrees of freedom is possible because the

earthquake behavior of the entire foundation can be examined by

assuming it as a rigid foundation. Calculations are made by

response spectra, treating the maximum response alone.

Modeling

For the purpose of maintaining the continuity between superstructure and substructure, a part of the weight of superstructure is

treated as a weight added to substructure and superstructure and

substructure are separated. When substructure is under water,

dynamic water pressure during earthquake is to be calculated in terms of virtual mass.

0.6pAhw Mw= (a/h) 2/3 w

where

p = bulk density of water

A = section area

a = width in the direction

of vibration

h w

= depth of water (In the iirec f ion

(0.8 S a/h

w i 4.0) of length b)

The soil surrounding substructure is to be calculated in terms of virtual mass, assuming it to vibrate together with the

structure during earthquake.

M

s = l/2(Y

s /g)a 2 d

where

Y

g = weight of soil

g = acceleration of gravity

d = thickness of soil layer

21-1
Earthquake force is to be calculated by the use of response

accleration spectrum (statistical value obtained from observations

made in the vicinity).

Related Constants

As for the dynamic spring constant of the subgrade, the lowest

value of the modulus of deformation of the original position

measured by the pressionmeter is to be used. The spring constants in the horizontal, vertical, rotational, and shearing

directions are to be determined as instructed in the "Specification for Highway Bridge Substructure Design (Caisson)."

KH

= kH' (b/D)" 3/4

, k H

= H E

p

, b: width (b = 'VV

Kv

= kv (a/D) -3/4 k = L0 K E v -q fi p»

a: length (a =

Flowchart of earthquake resistant design calculation ;

Subgrade

measurement

Design

calculation

Arrangement of

earthquake record

Determination

of subgrade

spring constant

I

Weight of substructure

Part of superstructure

weight calculated in term3

of weight added to substructure

Added weight of earth and

water

I

Earthquake

response

spectrum

(acceleration)

Building up of

vibration model

Response value

obtained by the

seismic coefficient method

Maximum value of response

obtained from response

spectrum

MO

Completion

21-2
K

R

= KVI, I

moment of inertia of the bottom

1/3

Or Blot's equation

2( l + v)bk

E R

E : Voigt's elastic module, v: Poisson's ratio

K. Evaluation

The values to be obtained by calculations are natural period,

center of rotation, displacement of gravity center and angle

of rotation of the model. The above values are used to evaluate

the subgrade reaction, distance of eccentricity, hearing power,

sliding and tipping.

Pile Foundations

Earthquake Resistant Design Objectives

When the soft soil stratum extends deep beneath the surface, the

ground movement is estimated to be large. Since the structural

type and dimensions are fixed, it is necessary for the bearings

to be flexible to reduce the section force.

Earthquake Calculation Method

When the piles are likely to vibrate, coupled with the surrounding soils, the ground is to be replaced by a multiple-layer

shearing model. When there are no seismic wave records available

for the same stratum of soil, the incident wave to the pile bearing stratum is to be estimated from the estimated acceleration

of the seismic foundation, using the shearing wave theory. Since

the coefficient of subgrade reaction cannot be obtained, the

dynamic reaction coefficient is to be obtained by modeling the

subgrade by the finite element method.

The ground is replaced by shearing models, stratum by stratum.

In the case of superstructure, only damping is to be considered,

neglecting the reaction characteristic. In the footings, the

righting moment of rotation and rotational inertia are to be

:onsidered.

Modeling

21-3
Original

for* c -4:1

c

'.W J 1 Bedrock

1' Bear log atratua)

Shearing

aodel

Z]

^.1

{4t

The incident wave to the bearing stratum is to be calculated

by means of the following equation, using the laws of reflection

and refraction.

uk (t,z) - fk (t --£•) + g(t

k

The stratum where the response (t,z) changes little is to be regarded as input bedrock (bearing stratum).

Re 1 a ted Constants

The spring constant of subgrade is to be estimated, using FEM in

plane strain.

K = - = k-S K S x

As for the added mass, it is assumed that footing or piles

have kinetic energy equal to that of the subgrade.

m- = J— X

TMX a

2 x

Evaluation

In the case of soft ground, there is so great a difference in the

acceleration of seismic waves between the bedrock and ground

surface that it is necessary to make investigations on the ground

responses. The bending moment in a pile is greater in its lower

part because the seismic waves are rapidly amplified.

There may be a limit to the application of the theory of

sheering wave (multiple-stratum ground, great earthquake). The

effects of a group of piles on the shearing model, as well as the

damping mechanism and nonl inear ity , must also be considered. A flowchart of this calculation follows:

21-4
Flowhcart of earthquake resistant design calculation ;

Subgrade

measurement T

l

i

l

Estimation of subgrade reaction by

FEU and calculation

of added mass of

earth

Design

calculation

Seismic wave

recordsT

I

i

Weight and rigidity

of superstructure,

footing and pile

Selection and

determination of

representative

incident wave

Building of multiplelayer shearing model

Setting up of

bedrock from the

theory of wave motion and determination of input

acceleration

Seismic response calculation by modal analysis

Caisson Foundation

Design Objectives

In the case of soft ground, there are problems with the caisson

behavior during earthquake. The type of superstructure can also

have special effects.

Earthquake Calculation Method

As for the ground modeling, the model of spring-dashpot system is

to be used from a practical veiwpoint, although the wave motion

theory and finite element method are also available. Response

spectrum analysis is done on a model with two degrees-of-freedom,

"orizontal translation, and rocking motion.

Modeling

Effects of superstructure and water are to be treated as added mass,

0.6y

(b/h

w

)

w^wW 1 , 2

j

2 / 3

» Ms = - y s gb d

21-5
When the side ground is very soft, it is to be treated as a plastic state if the horizontal reaction exceeds the allowable

bearing capacity. In the case of a large caisson, the response

spectrum of earthquake records in the vicinity is to be used. The

maximum design acceleration is 180 to 250 gal.

Original form Add*! mmmm

Dynamic Superstructure aodel

- 4 -

IT

2 Wl

Related Constants

It is necessry to examine the subgrade spring constant by a variety

of Boussinesq's equation, as given below, and other methods (Pauw,

Taj imi , Richart )

:

K„ =

2, 3/4

l p (1-v )B V

where

Ground deformation modulus = 6E *r 5E

o o Pressiometer deformation modulus

B„ = Foundation conversion width = -/Ay

Poisson's ratio

Shape factor

Evaluation

Although there are some problems about the conversion of superstructure by added mass, in general there is not much difference

from the results of analysis by the whole system model. The

response is governed by the evaluation of the subgrade spring

constant. Particularly, the evaluation of the effects of caisson

shape

.

The values to be calculated are the natural period, the horizontal and rotational displacement of gravity center, and maximum

acceleration response. Since the substructure response is influenced by the type of superstructure, it is necessary to make

response analysis at the basic design stage. A flowchart follows:

21-6
Flowchart of earthquake resistant design calculation :

Subgrade

measurement

Design

calculation

l

I Determination of

subgrade spring

constant

Arrangement of

seismic records

1

Added mass of

supers true ture

Added mass of earth

and water

Preparation of

response spectrum

Preparation of

vibration model

-tJ Determination of

subgrade constant

Maximum response value

by response spectrum

NO

valuation^

OK

< Completion

Multi-Column Foundation

Design Objectives

Since this is a flexible structure, the elastic behavior of the

foundation and footings are to be investigated. The coupled vibra-

:ion with the superstructure must also be considered.

Earthquake Calculation Method

It is necessary to investigate the three-dimensional behavior for

reasons of topographic inclination, arbitrarity of earthquake input

lirection, eccentricity of superstructure reaction, and asymetry

for the foundation structure. Since it is impossible to estimate

:he condition in which the foundation and ground are bound together

.the spring constant is not uniform), added mass of ground and

lamping constant, there is the necessity for analysis of threelimensional vibration. To make the calculation in terms of a mltiple system of material points, modal analysis is to be used.

21-7
Modeling

Considering the three-dimensional vibration, the deformations are

to be taken as unknown quantities.

Pier: Horizontal x and y displacements (2)

Footing: Horizontal x and y displacements, torsional angle

and rotational angle (10)

Superstructure: Horizontal x and y displacements and torsional (4)

MglMl

HulUpla

point mo4*\natarlal 0

o 0 o

0 o 0

o o o

Sround support nodal

lap plat*

Italtleolonn ry Motion

' \ Tartloal and borlaoatal * (two dlxootiona) Rotational aprlng

Evaluation

Static and dynamic calculation results must be compared. Likewise,

calculated values must be compared with vibration experimental

results. It is also necessary to check the responses by varying

the subgrade reaction coefficient. Since the foundation has so great an influence on the superstructure, its coupling to the

superstructure is also important. A flowchart of this analysis

follows

:

21-8
Flowchart of earthquake resistant design calculation .

Subgrade

measurement

Design

calculation

Arrangement of

seismic records

Determination

of ground

spring constant

Conclusion

NO

Calculation of natural

period as a continuous

elastic body

Calculation of natural

period of the various

parts by the system of

material points

Change of

material point

conversion factor *

Coincidence of^.

latural period^*"

Added mass Determination

of typical

seismic wave

Building up of

vibration model

Assumption of subgrade

spring constant and

damping constant

Maximum response

analysis by modal

analysis

i Completion

As seen from the examples, the objectives of earthquake resistant

design mainly concern two cases: Either when the bridge is very

long, or the ground is poor. The details of these two items are found by studying:

The effects of the seismic responses of substructure on

the superstructure due to the long span of the bridge.

'Dynamic characteristics of the substructure.

21-9
The ground behavior during earthquake.

The behavior of the foundation structure itself during

earthquake. The current trend is obviously toward evaluating the behavior during earthquake due to seismic force

characteristics, and the response characteristics (natural

period) of ground and structures. However, the behavior

during earthquake has not yet been fully clarified in

the analysis of seismic response.

Many studies involving actual observations and analysis of

experimental results have been made about the way the subgrade

spring constant, damping -constant , and added mass are to be treated.

However, from the methodological standpoint, the earthquake resistant

design calculation has now reached a state in which safe bridges

are becoming the norm.

21-10
RECENT DEVELOPMENTS IN SEISMIC DESIGN CODES

Roland L. Sharpe

Applied Technology Council

Palo Alto, California

Recent Developments in Seismic Design Codes

The need for a coordinated effort to review existing requirements

and state of knowledge and to develop comprehensive seismic design

provisions applicable to all of the country was recognized several

years ago. It was realized by design professionals and government

representatives that the effort would take many years to complete

if performed by volunteer committees as codes have been developed

in the past. After numerous discussions and careful planning, ATC

started work in November 1974 on a detailed study comprising:

A comprehensive review of seismic provisions

An evaluation of the latest state of knowledge and state

of practice

Development of a coordinated set of provisions which

could be used as a resource document in developing

seismic design provisions for all areas of the United

States.

The 85 participants involved in the development of the provisions were organized into 5 working groups composed of 14 task

committees, 2 advisory groups, 1 independent review group, and a project executive panel; the author was Project Director.

Two working drafts were submitted to several hundred outside

reviewers. These reviewers represented practice, industry, profesj sional organizations, government agencies, code promulgating groups,

and universities. The numerous comments received for each draft

were reviewed by the task committees, and changes and clarifications

were made as deemed appropriate.

The basic concepts and approaches followed in the review and

evaluation are described and a summary of some of the provisions

developed by the 85-member group assembled and retained by the

Applied Technology Council (ATC) are presented. The development

of seismic risk maps, and analysis and design provisions for

structural and nonstructural systems are discussed.

Basic Concepts

The primary basis for development of the seismic design provisions

for buildings was to protect life safety and to ensure continued

functioning of essential facilities needed during and after a catastrophe. Another major premise was that provisions should be

developed so they could be applicable to all areas of the United

States

.

In the development of the U.S. codes currently in use, it was recognized that the specified seismic design forces might be considerably smaller than those encountered in moderate or major earthquakes. Primary consideration is given to seismic design for the

main structural framing system. Consideration is given either

explicitly or implicitly to the effects (both good and bad) of

interior partitions, exterior covering, different types of materials,

and damping. However, this approach has caused misunderstandings

resulting in building designs with serious deficiencies such as inadequate tying together of building components and inadequate

provisions for the occurrence of building deformations in excess of

those calculated for the design forces.

22-1
The ATC-3 tentative provisions are intended to be logically

based with explicit consideration given to factors that are generally implicit in present code design provisions. There are a number of new concepts that are significant departures from existing seismic building codes, such as:

1. More realistic ground motions.

2. Consideration of the effects of distant earthquakes on long

period buildings.

3. Response modification coefficients (reduction factors) which

are based on consideration of the inherent capacity for energy

absorption and damping associated with inelastic response, and

observed past performance of various types of framing systems.

4. Complexity of analysis and design dependent on seismic motion

intensity, building importance or use, and assigned building

seismic performance.

5. Simplified seismic response coefficient formulas related to fundamental period of building but with certain restrictions.

6. Detailed requirements for architectural, electrical, and

mechanical sytems, and components.

7. Material design stresses approaching yield.

8. Guidelines for assessment and systematic abatement of seismic

hazards in buildings.

9. Guidelines for assessment of earthquake damage, and strengthening or repair of damaged buildings and potential seismic hazards

in existing buildings.

Items 1 through 6 are discussed in this paper. The remaining items

are contained in the ATC-3 project report [2],

It should be noted that the provisions are intended to apply

only to buildings and do not contain design requirements for special

structures such as bridges, transmission towers, offshore structures,

piers and wharves, industrial towers and equipment and nuclear reactors.

Ground Motions

In developing the seismic risk maps, four basic factors were considered: (1) a realistic appraisal should be made of the expected

ground motion intensity, (2) the probability of the design ground

shaking being exceeded should be approximately the same in all

parts of the country, (3) the effects of distant earthquakes on long period buildings should be considered, and (4) areas should

not be microzoned or actual faults shown on the maps and variations

of ground shaking over short distances of about ten miles or less

should not be delineated.

The recommended seismic design regionalizat ion maps in Figures 1 and 2 are based on an evaluation of historical seismicity, frequency

of occurrence of the earthquake motions and underlying geology where

possible. They reflect the collective judgment of several committees

based upon the best data available in 1976, adjusted and tempered by

22-2
experience and judgment. It is probable that the maps and coefficients will change with time as more knowledge is gained about earthquakes and their ground motions and as society better understands

the establishments of acceptable risk.

Seismic Design Coefficients

The effective peak acceleration (A in Figure 1) and effective peak

velocity-related acceleration (A m Figure 2) should be considered

as normalizing factors for construction of smoothed elastic response

spectra for ground motions of normal duration. Motions of high

seismic areas near active faults could exceed these values, especially in locations inside the 0.4g contour.

The development of the map was facilitated by the work of

Algermission and Perkins of the United States Geological Survey

(USGS) and is in may ways quite similar to their maps [3]. The

A

v was constructed to consider the effects of distant earthquakes.

< The velocity attenuation study by McGuire and the attenuation of

v modified Mercalli intensity by Bollinger in conjunction with review

of other data from large earthquakes were used to develop the A

v

! contours [4, 5]. It is felt that considerable further study should

be given to this area. The maps were literally drawn by a committee

after considering input from many sources, but they are judged to provide the best current estimate of the geographical variations

of A and A^ for purposes of design of buildings.

After the contour maps in Figures 1 and 2 were prepared, the

country was divided into seven map areas. Each map area, for

purposes of design, is assigned a seismicity index as shown in

Table 1. To facilitate administration by regulatory agencies, maps

were then prepared showing the A

g and Ay values on a county-bycounty basis. A

g and A

v values are given for each of the over

4,000 counties in the United States.

Applicability to All Areas of the United States

The seismic hazard varies from very low to high across the United

States. In order to make the provisions adaptable to all parts

of the United States, four seismic performance categories were established to which buildings would be assigned based on the

seismicity index of site and the seismic hazard exposure (occupancy or use) of the buildings, see Table 2.

Buildings in Seismic Hazard Exposure (SHE) Group III include

critcal or essential facilities that should function during and

after an earthquake such as hospitals, communication and control

centers, and buildings with a high density of occupants. SHE

Group II includes buildings, housing a large number of occupants

or buildings in which the occupants' movements are restricted or their mobility is impaired such as public assembly for 100 or more

persons, schools, large retail stores, hospitals, and ambulatory

health facilities. SHE Group I includes all other buildings not

classified in Group III or II.

For Seismicity Index I (the lowest risk), SHE Group III buildings are required to have limited analysis and design for

seismic forces. All other buildings do not have to be designed

for seismic forces; however, a number of basic details are specified such as connections of walls to roofs and floors, and

interconnections of other building elements. Experience has

shown that the tying together of parts of buildings can be of

material assistance in resisting earthquake forces.

22-3
All buildings in Seismicity Index 2 must have limited seismic

analysis and design. For Seismicity Index 3, all SHE Group III and II buildings must be designed to requirements roughly com- parable to the 1973 Edition of the Uniform Building Code for

California conditions of seismicity. SHE Group I buildings need

only limited seismic analysis and design.

Buildings in Seismicity Index 4 area, assigned to SHE Groups II and I, must meet the seismic analysis and design requirements for

critical or essential facilities in Seismicity Index 3. SHE

Group III buildings must also meet restricted height limitations

if they are constructed with shear walls or braced frames and

certain additional restrictions on use of materials are specified.

Design Steps: Structural

The seismic analysis and design procedures to be used in the design

of a building and its components are dependent on the Seismic

Performance Category (Table 2) to which the building is assigned.

The seismic design forces are established considering the values

of A

g and A^, the effects of site conditions, the type of framing

employed, and the fundamental period of the building based on the

elastic properties of the seismic resisting system.

The internal forces in the members of the building shall be

determined using a linearly elastic model. Individual members shall

be sized for the shears, axial forces, and moments determined in

accordance with the analysis and design provisions. The deformation

of the building when subjected to the design seismic forces shall

not exceed prescribed limits.

Analysis and design of the structural system would proceed as follows

:

1. Locate site A

g and A

y maps; determine the map area number.

2. Using the map area number, select the coefficients A

g and A

v and the Seismicity Index from Table 1.

3. Considering the use or importance of the building, establish

the Seismic Hazard Exposure Group and determine the Seismic

Performance Category from Table 2.

4. The Seismic Performance Category to which a building is assigned determines the analysis and design requirements.

Category A

The structural frame components in Category A buildings are required to be tied together, but no overall seismic design

is required. Category A buildings can be used in areas of

low seismicity.

Category B

Must meet the Category A requirements and in addition have

minimum seismic design such as collector elements, diaphragm

design, design of bearing walls for for seismic forces normal

to the flat surface of the wall, reinforcement of openings

in shear walls, or diaphragms, and certain minimum pile foundation design requirements. Category B buildings are used

22-4
for essential facilities in areas of low seismicity and for

all buildings in areas of moderate seismicity.

Category C

Must meet the requirements for Category B and also must conform

to certain framing limitations for buildings over 160-feet (49

meters) in height. The seismic resisting system for buildings

over 160-feet in height must be one of the following:

A moment resisting frame system with Special Moment

Frames (Special Moment Frames are ductile-moment

resisting frames).

A dual system; see Table 3.

A braced frame (steel or concrete) or shear wall

system with the braced frames or shear walls arranged

so that the frames or walls in any plane do not resist more than 33 percent of the seismic design

force including torsional effects; this system is

limited to buildings less than 240-feet (74 meters)

in height.

Moment resisting space frames which are enclosed by or

adjoined by more rigid elements not considered to be part of

the seismic resisting system shall be designed so that the

action or failure of those elements will not impair the vertical

Load and seismic force resisting capability of the space frame.

Hie design shall consider and provide for the effect of these

igid elements on the structural system at building deformations

:orresponding to a total story design drift A where:

A

s = C

d

A a

(1 + a d> (1)

^ is the deflection amplification factor from Table 3, A

a is

he story drift calculated using the seismic design forces

ind the elastic stiffness of the seismic resisting system,

nd a^ is the amplification coefficient related to P-Delta

ffets (P-Delta effects in a given story are the secondary

ffects due to eccentricity of the gravity load above that

tory). If the P-Delta effects are small, a^ can be taken

s zero.

All structural elements not considered in the design to

e part of the seismic resisting system in Category C buildings

hall be investigated and shown to be adequate for the vertical

oad-carrying capacity and the induced moments resulting from

he total story design drift A

g

.

tegory D

ust conform to Category C requirements and are also subject

3 more restrictive height requirements; the 160-feet and 240-

bet height limitations given above are reduced to 100-feet

31 meters) and 160-feet (49 meters), respectively.

It was also recognized that there can be potential adverse

ffects when the ratio of the strength provided in any story to

e strength required is significantly less than that ratio for

e story immediately above. No specific provisions were devel-

>ed; however, a proviso is made that the story strengths shall

22

22-5
be adjusted to compensate for this effect. Possible methods

for evaluating these ratios are presented in the Commentary

included in the ATC-3-06 document.

Similarly, the requirement is stipulated that the design

of a building shall consider the potentially adverse effect

that the failure of a single member, connection, or component

would have on the stability of the building and appropriate

design modifications shall be made to mitigate this effect.

5. The building should be classified as regular or irregular

considering both the plan and vertical configuration.

6. The analysis and design procedures are determined based on the Seismic Performance Category (SPC) and whether the building is regular or irregular. SPC B buildings, both regular

and irregular, must be analyzed as a minimum using Equivalent

Lateral Force (ELF) procedure. SPC C and D buildings that

are regular shall, as a minimum, be analyzed using the ELF

procedure. Irregular buildings shall be analyzed considering

their dynamic characteristics. For vertical irregularities

only, a modal analysis procedure is specified.

7. Assuming an SPC B, C, or D building, the following steps 8 through 16 apply.

8. Determine the site soil profile and select the soil factor S. The factor S varies from 1.0 for rock sites to 1.5 for soft

to medium -stiff clays and sands.

9. Select the Response Modification Coefficient R and the deflection amplification factor C^ from Table 3. R is dependent on

the type of structural system and the vertical seismic resisting system. The deflection amplification factor C^ , is used

to determine the total story drift.

10. For the ELF procedure, the total seismic shear at the base

V = C

g W. Where W equals total gravity load plus portions of

other loads such as permanent equipment, effective snow loads,

and 25% of the floor live load for warehouses.

1.2 A S n - v (2)

2/3 RT

The fundamental building period T

g equals

T a = C T

li 3/4

a In

where

C T

= 0.035 for steel frames

Cp = 0.025 for concrete frames

h = height in feet above the base.

For all other buildings T

fl = 0.5 h n

/V"~L, where L equals the

overall length of the building in the direction being considered,

22-6
For lower buildings, the fundamental period need not be

calculated. In such cases, the minimum value of C

g shall

be 2.5 A

a /R, except for soft-soil sites when A

a is equal to or greater than 0.3 the minimum value of C

g =£Aa /R.

The total lateral shear force at the base, V, may be

reduced for soil-structure interaction effects. Soilstructure interaction should reduce the shear force but

because of rotation about the base the total building

deflection would be increased.

11. The total lateral shear force at the base, V, is distributed

the height of the building as follows:

w h k (3)

n

1 w ihik

i = l

where k is an exponent related to building period and has the

following value:

For buildings having a period of 0.5 seconds or less,

k = 1

For buildings having a period of 2.5 seconds or more,

k = 2

For buildings having periods between 0.5 and 2.5

seconds, k may be taken as 2 or may be determined

by linear interpolation between 1 and 2.

F x is the seismic force applied to level x, w^ or w

x is the portion of W located at or assigned to level i or x, and h^ or h

x

is the height above the base to level i or x.

12. The seismic shear and torsion in any horizontal plane is distributed to the vertical resisting elements in accordance with

the relative stiffness of the elements and the horizontal

diaphragm. Torsional moments due to building irregularities

plus a 5 percent "accidental" torsion are considered.

13. A reduction in overturning moment is allowed for buildings

ten stories or higher as follows:

F

i

< h i~h x>

(4)

where

B = 1.0 for the top ten stories

B = 0.8 for the twentieth story from the top and below

B = a value between 1.0 and 0.8 determined by a linear

interpolation for stories between the twentieth and

tenth story below the top.

22-7
The foundation of buildings, except inverted pendulumtype structures, may be designed for an overturning moment

at the foundation-soil interface determined using B = 0.75

in the above formula for all building heights providing the

resultant of the seismic forces and vertical loads at the

foundation-soil interface does not fall outside the middle

one-half of the base of the component(s) resisting the overturning .

14. For buildings requiring a modal analysis, a lumped mass model

may be used. The design value for each of the story-shear

moments and drifts and the floor deflection quantities shall

be determined by combining their modal values by taking the

square root of the sum of the squares of each of the modal

values. The base shear thus determined shall not exceed a base shear calculation using Equation 2 with a value T = 1.4

The design story shears, moments, and drifts and the floor

deflections shall be adjusted accordingly.

15. The story drifts should be calculated to ensure they do not exceed the values in Table 4.

16. Using the analytical results and material design stresses

approaching yield, the framing system is resized from the

preliminary model.

17. The design should be reviewed to confirm that final member

sizes are in conformance with those assumed in the analysis.

Revisions and reanalysis should be made as appropriate.

18. The design details should conform to general requirements for

tying together of parts of buildings, concrete or masonry

wall anchorage, collector elements, diaphragms, bearing walls

inverted pendulum-type structures, vertical seismic motions,

and deflection and drift limits.

Design Steps: Nonstructural

Architectural systems and components are required to be designed

to resist seismic forces as follows:

F = A C P W

p v c c where

Fp = The seismic force applied to a component of a building

or item of equipment at its center of gravity

C

c = The seismic coefficient for components of architectural

systems as given in Table 6 (dimensionless

)

W

c = The weight of a component of a building or item of

equi pment

P = Performance criteria factor as given in Table 5 (d imens ion less )

.

Attachments of exterior wall panels to the building seismic

resisting system are required to have sufficient ductility and

rotational capacity to accomodate the total story drift.

22-8
Mechanical and electrical systems are required to be designed

to resist seismic forces as follows:

F = A 7 C P a„ a v

p v c c x c

where

C

c = The seismic coefficient for components of mechanical and

electrical systems in Table 7 (dimensionless

)

a

c = The amplification factor related to the response of a system or component as affected by the type of attachment (varies from 1 to 2).

ax

= The amplification factor at level x related to the

variation of the response with height of the building.

ax

= 1.0 + (hx

/h

n

)

where

h

x

= the height above the base to level x

hn

= the height above the base to level n.

Conclusions

The ATC-3 document is predicated on the assumption that the analysis

and design provisions must be used in conjunction with the material

provisions covering wood, steel, concrete, and masonry because the

design details and requirements are based on strength design. The

A

fl and Av values, the response modification coefficients (R), the

defletion amplication factors (C^), and the T

a formulas were

determined by considering their interdependence with the materials

requirements.

Because of the new concepts and approach, it is strongly

recommended that the ATC-3 provisions be tested by making numerous

building designs and comparing the results with the same buildings

designed using current seismic codes. It is also recommended that

further detailed study be given to the A and A maps and their

values, the values of R and C^ , the method of the limitations on determination of building periods, design requirements for torsional

response, the effects of and design for irregularities in framing

and discontinuities in strength, the design of foundations for

overturning forces, and design for the effects of orthogonal

ground motions.

References

1. Sharpe, R. L. , "Evolution of Codes and Standards for

Earthquake-Resistant Reinforced Concrete Building Construction,"

Proceedings of Workshop on ERCBC, University of California,

Berkeley, California, July 11-15, 1977.

2. Applied Technology Council, "Tentative Provisions for Development of Seismic Design Regulations for Buildings," (ATC-3-06),

prepared for the National Bureau of Standards - National Science

Foundation, Washington, D.C., 1978.

22-9
3. Algermissen, S. T. , and Perkins, D. M. , "A Probabilistic

Estimate of Maximum Accleration in Rock in the Contiguous

United States," United States Geological Survey Open-File

Report 76-416, 1976.

4. McGuire, R. K. , "Seismic Structural Response Risk Analysis,

Incorporating Peak Response Progressions on Earthquake

Magnitude and Distance," Department of Civil Engineering,

Massachusetts Institute of Technology, Cambridge,

Massachusetts, Report R-74-51, 1975.

5. Bollinger, G. , "Reinterpretation of the Intensity Data for the

1886 Charleston, South Carolina Earthquake," Bulletin of

Seismological Society of America (in publication)

.

22-10


TABLE 1: COEFFICIENTS A

fl

, A

y

, AND SEISMICITY INDEX

Coefficient Aa Map Area Coefficient Av Seismicity

Figure 1 Number Figure 2 Index

0.40 7 0.40 4

0.30 6 0.30 4

0.20 5 0.20 4

0.15 4 0.15 3

0.10 3 0.10 2

Q.05 2 0.05 2

0.05 1 0.05 1

TABLE 2: SEISMIC PERFORMANCE CATEGORY

Seismicity Seismic Hazard Exposure Group

Index III II I

4 D C C

3 C C B

2 B B B

1 BAA

22-13
TABLE 3: RESPONSE MODIFICATION COEFFICIENTS

Vertical Seismic Coefficients

Type of Structural System Resistinq System R

BEARItIG 1,'ALL SYSTEM: A structural system with

bearing walls providing support for all, or Light framed walls

with shear panels 6H 4 rnajor portions of, the vertical loads. Seismic force resistance is provided

by shear walls or braced frames.

Shear walls

Reinforced concrete

Reinforced masonry 3%

43

Braced frames 4 % . Unreinforced masonry

shear walls l»s

EUILDING FRAME SYSTEM: A structural system

with an essentially complete Space Frame Light framed walls

with Shear panels 7 AH

providing support for vertical loads. Seismic force resistance is provided by

shear walls or braced frames.

Shear walls

Reinforced concrete

Reinforced masonry

5h 5

4

Braced frames 5 4»s Unreinforced masonry

Shear walls 1%

I-BMEMT RESISTING FRAME SYSTEM: A structural

system vri th an essentially complete Space

Frame providing support for vertical loads.

Special moment frames

Steel

Reinforced concrete

8

7 6 Seismic force resistance is provided by

Ordinary or Special Moment Frames capable

of resisting the total prescribed forces.

Ordinary moment frames

Steel 4%

Reinforced concrete 2

4

2

DUAL SYSTEM: A structural system with an essentially complete Space Frame providing

support for vertical loads.

Shear walls

Reinforced concrete

Reinforced masonry

0

A Special Moment Frame shall be provided

which shall be capable of resisting at least

25 percent of th? prescribed seismic forces.

Wood sheathed shear

panels 8 5 The total seismic force resistance is provided

by the combination of the Special Moment Frame

end shear walls or braced frames in proportion

to their relative riqidities

Braced frames 6 5

INVERTED PENDULUM STRUCTURES. Structures where the framing resisting the total prescribed

seismic forces acts essentially as isolated

Special Moment Frames

Structural steel

Reinforced concrete.

?.h 2h

2%

cantilevers and provides support for vertical

load. Ordinary Moment Frames

Structural steel Us Vi

22-14
TABLE 4: STORY DRIFT A g

Seismic Hazard Exposure Group

III II

A

a 0.010 hsx 0.015 h sx 0.015hsx

h is the story height below level x.

1 Where there are no brittle-type finishes in buildings three stories or

less in height, these limits may be increased one-third.

TABLE 5: PERFORMANCE CRITERIA

Performance

Designation Characteristic Level

S Superior 1.5

G Good 1.0

I Low 0.5

22-15
TABLE 6: SEISMIC FORCE AND PERFORMANCE CHARACTERISTIC LEVELS

REQUIRED FOR ARCHITECTURAL SYSTEMS OR COMPONENTS

Architectural Systems or Components

Appendages

Exterior Nonbearing Walls

Wall Attachments

Veneers

Roofing Units

Containers and Miscellaneous

Elements (free standing)

Parititions

Stairs and Shafts

Elevators and Shafts

Vertical Shafts

Horizontal Exits Including Ceilings

Public Corridors

Private Corridors

Full-Height Separation Partitions

Full-Height Other Partitions

Partial-Height Partitions

Structural Fireproof ing

Ceilings — Fire-Rated Membrane — Nonf ire-Rated Membrane

Required Levels

Cc Seismic Haz ard Exposure' Group

Factor ITI II I

0.9

3.0

3.0

0.6

1.5

1.5

1.5

0.9

1.5

0.9

0.6

0.9

0.6

0.6

0.9

0.9

0.6

G

G

G

G

G

L

LS

G

IG

LL

L

LL

NR

NR

G

L

LG

L

NR

G

L

NR

Architectural Equipment — Ceilings,

Wall, or Floor Mounted 0.9

KR~ - Not required

22-16
TABLE 7: SEISMIC FORCE AND PERFORMANCE CHARACTERISTIC LEVELS

REQUIRED FOR MECHANICAL/ELECTRICAL COMPONENTS

Seismic Hazard

Cc Exposure Group

Mechanical /Electrical Components Factor III II I

Emergency Electrical Systems (code required)

Fire and Smoke Detection System (code required) 2.00 S S S

Fire Suppression Systems (code required)

Life Safety System Components

Boilers, Furnaces, Incinerators, Water Heaters, and

Other Equipment Using Combustible Energy Sources or High Temperature Energy Sources, Chimneys, Flues,

Smokestacks, and Vents

Communication Systems

Electrical Bus Ducts and Primary Cable Systems

Electrical Motor Control Centers, Motor Control Devices,

Switchgear, Transformers, and Unit Substations 2.00 S G L

Reciprocating or Rotating Equipment

Tanks, Heat- Exchangers, and Pressure Vessels

Utility and Service Interfaces

Machinery (manufacturing and process) 0.67 S G L

Lighting Fixtures 0.67 S G L

Ducts and Piping Distribution Systems — Resiliently Supported 2.00

Ducts and Piping Distribution Systems — Rigidly Supported 0.67 S G NR

Electrical Panelboards and Dimmers 0.67

Conveyor Systems (non-personnel) 0.67 s NR NR

22-17

LOGICAL ANALYSIS OF SEISMIC DESIGN PROVISIONS

James Harris

National Bureau of Standards

Washington, D.C.

Logical Analysis of Seismic Design Provisions

In early 1976, the National Bureau of Standards (NBS) and CarnegieMellon University (CMU) undertook a collaborative project to apply

logical and systematic techniques of analysis and information display to the earthquake resistant design criteria then being developed by the Applied Technology (ATC). The project is funded by

the National Science Foundation (NSF) and has three objectives:

1. To aid the participants in ATC's project in their task of

developing the new provisions by applying existing techniques

of analysis to detect ambiguities, omissions and inconsistencies.

2. To document our analysis and ATC's provisions by preparing

and publishing the provisions expressed tabular, logical

format

.

3. To prepare alternative arrangements of ATC's provisions that

would meet the needs of some special audiences more directly

than the orginal arrangement.

ATC has produced a large and impressive set of provisions.

Because these provisions will now be considered by the building

codes and standards community for adaption into many codes and

standards, it is imperative that they be correctly understood. Yet,

like nearly all technical provisions, they will be subject to misinterpretation. This need for precise communication is exactly

what the techniques used in this study are design to bolster.

Overview of the Techniques

A set of provisions can be analyzed to determine that they are

clear, complete, consistent and, to some extent, correct. The

analysis is conducted in an objective fashion at several levels

of detail. The primary benefit to the user of the analysis is

that it raises questions about clarity, completeness, etc. The

technique does not provide corrective answers, for that generally involves actual change in the wording of the provisions.

There are three principal tools used in the analysis:

1. The decision table is used to represent the meaning of

individual provisions. It is simply an orderly presentation

of the reasoning controlling a set of decisions. It is easily analyzed to assure that the reasoning process will lead

to a unique result and that no possibility exists for encountering a situation not defined. Decision tables present an overall analysis of situations involving parallel thought processes,

whereas written text and, to some extent, flow charts both

describe more of a sequential thought pattern.

2. The information network is used to represent the precedence

relations among the provisions. Each decision table occupies

one node in the network. The nodes are connected to their

ingredient nodes by branches that represent the flow of

information through a set of provisions from the input data

to the terminal criteria. The ingredients of any node are simply all those nodes that may be required for the direct

evaluation of the node.

3. The outline/classified index is used to represent the arrangement and scope of the provisions. The subset of the overall

23-1
set of provisions which contains the likely points of entry

by users of the provisions is selected as a set of basic

provisions. Each of these basic provisions is then classified using key words that define the scope of the provision.

The classifiers allow outlines and indexes to be constructed,

and the basic provisions are entered at the appropriate points.

Several different arrangements of the basic provisions can be

generated, allowing one ?o select the best arrangement for any

given use.

Examples of the use of each of these tools in the analysis of

ATC's provisions are presented in the following sections. The complex documentation will be available in a few months. Further

examples and explanations of the techniques are available in

several publications [1, 2, 3].

Decision Table

A decision table is a tabular arrangement of conditions, actions,

and rules. A condition is a logical statement that must have one

of only two values: true or false. An action is any operation;

generally, in the context of this study it is the assignment of

a value to a variable. A rule is a statement that prescribes a set of condition values in order that a specified set of actions

can be performed. Thus, the simple provision, "No single floor area shall exceed that permitted for a one-story building," is

shown in decision table form as follows: _ , Rule Rule

1 2

Condition 1

Actual story area < allowable one-story

building area

T F

Action 1 Single story area

X

Action 2

Single story area check - violated X

The table is read rule by rule, "If the actual story area

is less than the allowable one-story building area, then the

single-story area check is satisfied; if it is not, then the

check is violated"

The following text is taken from an intermediate draft of

ATC's provisions. The subsequent decision table represents those

same provisions, except that the three rules at the right edge of

the table represent situations for which the text gives no action

Site Effects

Site effects on building response shall be established based on three soil profile factors defined as follows:

23-2
SOIL PROFILE TYPE A is a profile with:

1. Rock of any characteristic, either shale-like or crystalline

in nature. Such material may be characterized by a shear

wave velocity greater than 2,500-feet per second, or

2. Stiff soil conditions where the soil depth is less than 200-feet

and the soil types overlying rock are stable deposits of sands,

gravels, or stiff clay.

SOIL PROFILE B is a profile with deep cohesionless or stiff

clay conditions, including sites where the soil depth exceeds 200-

feet and the soil types overlying rock are stable deposits of sands,

gravels, or stiff clay.

SOIL PROFILE C is a profile with soft- to medium-stiff clays

and sands, characterized by 30-feet or more of soft- to medium-stiff

clay with or without intervening layers of sand or other cohesionless soils.

In locations where the soil profile type is not known in sufficient detail to determine the soil profile type and where foundations are supported without the use of piles, Soil Profile B or C shall be used whichever produces the larger base shear.

1 2 3 4 5

Soil type - rock T F F

Soil type - stiff T T F F

Soil depth 200' + T F

Soil type - soft clay T F F

Depth of soft clay 30* T F

Soil type known + + + + F F T T

Piles support foundation F T

Soil Profile Type - A X X

Soil Profile Type - B X

Soil Profile Type - C X ?»

Soil Profile Type - B/C X

4

Mote: + means true predetermined by another condition value

- means false predetermined by another condition value

* means either true or false is acceptable

23-3
The analysis of the provisions for completeness and conflicts

begins with the construction of the decision table. For the preceding provisions, the result was the decision table shown without

the three rules at the right edge. The upper right quadrant of

the table is then analyzed to see that no two rules could be matched

simultaneously and that all possible rules are included. The easiest

and most reliable way to do this is by decomposing the table into

a decision tree. This analysis is easily performed with a computer

[4, 5] . The decision tree for the preceding table is shown in

Figure 1. Note that there are three nodes labeled "Else" in that

tree. They each represent possible rules that are not in the original table, and it was from them that the three rules at the right

edge of the table were constructed. These rules do represent possible situations for which the provisions, as written, provide no answer. Thus, the analysis has raised a question.

Also note that none of the nodes in Figure 1 contain more than

one of the rules in the original table. If that had been the case,

a situation would have existed in which the provisions contained

a contraction, at worst, or were redundant, at best.

The logical problem in the provision for soil profile type was solved by ATC by rewriting the last paragraph as follows:

In locations where the soil properties are not known in

sufficient detail to determine the soil profile type or where the profile does not fit any of the three types,

Soil Profile B shall be used.

Information Network

The information network represents the flow of information through

the decision points in a set of provisions. The entire network

can be assembled once each of the nodes and their direct ingredients are known. The assembly is easily performed with a com- puter program [4, 5], The complete information network can then

be used for two general operations: to trace global ingredients

of a particular node (that is, all the nodes that have any possible influence on the mode in question) and to trace the global

dependence of a particular node (that is, all the nodes that might

be influenced by the node in question). These two operations are useful to those actually complying with a set of provisions (particularly in the use of automatic data processing) because they

provide all the necessary cross references. They are also useful

in the development of a set of provisions; the global ingredients,

in particular, can be used to guide the ordering and written expression of provisions. The network can also be used to detect loops

in the precedence and detached (uncrossreferenced) sets of provisions.

In the analysis of one of ATC's intermediate drafts, it was

found through examination of the information network that a provision for the effects of vertical motion was misplaced. It was originally located in the chapter that described the "Equivalent

Lateral Force" procedure of analysis, although it was not functionally related to any other provisions in that chapter. The

information network clearly showed that it was functionally related

to the load combination provisions for component design (see Figure 2) and thus a recommendation was made to ATC that the provision

be moved to a more appropriate location in the section on component

design. This move provided for a more direct agreement between

the written expression and actual use of the provisions.

23-4
Outline/Classified Index

The classification of provisions in a systematic fashion offers

considerable insight into the overall organization of a set of

provisions for better access and retrieval. Algorithms exist to automatically generate outlines of headings once a set of provisions has been classified [5, 7] and guidelines exist for the

classification of provisions [8, 9]. Two brief examples of the use of classification in analyzing the organization of ATC's provisions are offered here.

In the first example, a chapter contained several sections,

two of which were "Framing Systems," which dealt with parameters

and restrictions for certain framing systems, and "Site Effects,"

which included the provisions for soil profile type analysed

earlier and certain restrictions on the use of some sites. The

section on framing systems was the first of the two. These two sections were not functionally related through the information network, so that had no influence on their order. On the basis of

classifying provisions by their positions in the process of design

and construction, it was recommended that their order be reversed,

for the site selection process almost always occurs before the

decision on a framing system.

In the second example, the intermediate draft of the provisions contained a chapter and an appendix dealing with wood construction. The original section headings are shown on the left

in the following, while the recommended reorganization is shown

on the right. The recommendation came about as a result of classifying each of the provisions on several different bases and noting

that there was a significant division between the provisions for

those buildings that were analyzed for seismic loads ("Engineered")

and those for buildings that were not analyzed ("Non-engineered").

This allowed a clear decision as to the applicability of each provision and eliminated the need for the appendix.

Comparative Outlines, Chapter 10

ATC-3-05 Recommended

10. Wood 10 Wood and Related Materials

10.1 Reference Documents 10 .1 Reference Documents

10.2 Strength of Members 10 .2 Non-engineered Construction

and Connections 10 .2 .1 Wall

10.3 Category A 10 .2 .2 Wall Sheathing Requirements

10.4 Category B 10 .2 .3 Acceptable Types of Wall

10.5 Category C Sheathing

10.6 Category D 10 .3 Engineered Construction

10A Wood Construction 10 .3 .1 Strength of Members

10A.1 General and Connection

10A.1.1 Diagonally Sheathed 10 .3 .2 Framing Requirements

10A.2 Wood Diaphragms 10 .3 .3 Requirements for All Shear

10A.2.1 Diagonally Sheathed Panels

Diaphragms 10 .3 .4 Diagonally Sheathed

10A2.2 Plywood Diaphragms Diaphragms

10A.3 Shear Panels Sheathed 10 .3 .4 Diagonally Sheathed

with Various Materials Diaphragms

10 .3 .5 Plywood Diaphragm

10 .3 .6 Shear Panels Sheathed with

Various Materials

10 .4 Category A

10 .5 Category B 10 .6 Category C 10 .7 Category D

23-5
Conclusion

The joint NBS-CMU project to aid in the development of new seismic

design provisions for building codes by applying systematic, logical analyses to the provisions is nearly complete. During the

first phase several recommendations were made to ATC as questions

were raised by the decision table, information network, and

classification/organization analyses of their interim drafts. The

final phase of the project will conclude with a report documenting

the analysis of the provisions and containing decision tables for

the final version of ATC's provisions.

References

1. Fenves, S.J., "Tabular Decision Logic for Structural Design,"

Journal of the Structural Division , American Society of Civil

Engineers, Vol 92, No. ST6, December 1966.

2. Fenves, S.J., and Wright, R. N. , The Representation and Use of

Design Specifications , Technical Note 940, National Bureau of

Standards, Washington, D.C. , June 1977.

3. Harris, J.R., Melin, J. W. , Tavis, R. L. , and Wright, R. N. "Technology for the Formulation and Expression of Specifications,

Volume I: Final Report, " Civil Engineering Studies , SRS 423,

University of Illinois, Urbana, 111., December 1975.

4. Harris, J.R., Melin, J. W. , and Albarran, C, "Technology for the

Formulation and Expression of Specifications, Volume II: Program

User's Manual" Civil Engineering Studies, SRS 424, University of

Illinois, Urbana, 111., December 1975.

5. Wright, R.N., Harris, J. R. , Melin, J. W. , and Albarran, C. "Technology for the Formulation and Expression of Specifications,

Volume III: Technical Reference Manual," Civil Engineering Studies , SRS 425, University of Illinois, Urbana, 111., December

1975.

6. Wright, R.N. , Boyer, J. T. , and Melin, J. W. "Constraint Processing

in Design," Journal of the Structural Division , American Society

of Civil Engineers, Vol. 98, No. ST1, January 1971.

7. Nyman, D.J. and Fenves, S. J., "An Organizational Model for Design

Specifications," Report R73-4, Department of Civil Engineering,

Carnegie-Mellon University, Pittsburg, PA, September 1973,

8. Fenves, S.J., Rankin, K. and Tejuja, H. , The Structure of Building

Speci f ications , Building Science Series 90, National Bureau of

Standards, Washington, D.C, September 1976.

9. Harris, J.R. , and Wright, R. N. , "Organization of Design Standards"

(to be published) National Bureau of Standards, Washington, D.C.

23-6
FIGURE 1: DECISION TREE GENERATED FROM DECISION TABLE BY TESTING

THE CONDITIONS TO SUBDIVIDE THE TABLE

NOTE: the branches marked ? show possible rules that are missing

from the original table, thus it is logically incomplete

FIGURE 2: PORTION OF THE INFORMATION NETWORK FOR STRUCTURAL DESIGN

23-7

24

HIGHLIGHTS OF CALIFORNIA SCHOOL AND HOSPITAL

BUILDING REGULATIONS

John F. Meehan

Office of the State Architect

California

Highlights of California School

and Hospital Building Regulations

California public school buildings have performed very well in all

earthquakes since the 1933 Long Beach earthquake. In that earthquake a great many public school buildings were badly damaged and

innumerable deaths would have occurred if the earthquake would have

occurred during school hours. Within one month following the earth-,

quake, the California Legislature adopted the Field Act, which

olaced public school buildings under the jurisdiction of the State,

similarly , the 1971 San Fernando earthquake caused a great deal of

damage to several hospital buildings, including one that was just

:ompleted a few months before the earthquake. Over 80 percent of

;he deaths in that earthquake occurred to hospital occupants. A

/ear after that earthquake the Legislature enacted the Hospital Act,

?hich also placed hospital buildings under State jurisdiction.

It is apparent that the schools performed well in past earth-

|uakes because of the provisions of the Field Act. The act requires

:hat the designers, from the private sector, must be registered

irchitects or structural engineers; an independent review of the

:ontract documents must be performed by the State; construction

urveillance by all project designers and the State; and continuous

nspection by an approved project inspector. Further, as the contruction progresses and at the completion of the project, verified

eports must be filed by the designers, inspector, and contractor

ndicating complete conformance with the approved construction

ocuments. False reports carry a felony penalty. Regulations may

e adopted and a fee may be collected to accomplish the intent of

he Field Act. The complete Field Act is given in the Education

ode, Section 39140-39156.

In 1967, the Legislature enacted statutes to require a eologic study of school sites.

Hospital statutes are somewhat similar except that the level

f performance is higher; that is, they require that the hospital

uilding "must be completely functional to perform all necessary

ervices to the public after a disaster...." They require that a tructural engineer perform the structural design; that a geologic

avest igat ion be made of the site; an independent plan review must

fe made by the State; construction surveillance must be provided by

pe State and all project designers; and the project must be continously inspected by an approved inspector. Verified reports by all

IE those involved in the construction supervision, which are the

esigners, inspector and contractor, must be filed. The complete

hospital Act is given in the Health and Safety Code, Section 15000-

Through delegation, the Field Act is enforced by the Structural

ifety Section (SSS) of the Office of the State Architect in the

:partment of General Services. However, the Hospital Act is

iforced by the Department of Health but the statutes require that

lie State structural plan review and construction supervision must

I; performed by the SSS under contract to the Department of Health.

iblic School Regulations

1 regulations applicable to public school buildings cannot be

esented here. Only a few of those which may be considered someat unique to California schools will be discussed.

5023.

24-1
The basic building regulations adopted by California are those

of the Uniform Building Code. Additions and deletions to it are adopted by various State agencies which relate to their specific

interest or legislative assignment. Currently the 1973 edition of

the Uniform Building Code (73 UBC) published by International

Conference of Building Officials is the effective basic code.

Procedures are underway at this moment to adopt the 1976 edition of

the Uniform Building Code (76 UBC).

The Uniform Building Code is a widely distributed code and

therefore those basic regulations will not be discussed here.

The regulations discussed in this paper will be only those

proposed by the SSS as additions or deletions to the 76 UBC.

Snow Load

The 76 UBC Section 2305(d), permitting a reduction on snow load due

to slope of roof was deleted because frequent measurements of actual

snow loads on roofs of many buildings by this author did not justify

any load reduction due to slope of the roof. Full load on half the

span or the ridge of the roof and no load on the remaining portion

of the span should also be used as a criteria in snow load design

because of the likelihood of having that loading condition when snow is being removed from the roof.

Veneered Walls

An addition was made to 76 UBC Section 2307(c) to require that the

perpendicular to the wall deflection of metal or wood stud walls

providing lateral support to masonry veneer be limited to L/600

when resisting seismic loads. All masonry over 1/2" in thickness

must be mechanically anchored to the structure with anchors capable

of carrying, at working stress, in both tension and compression a load equal to twice the weight of the veneer. This is usually

accomplished with a piece of 18ga galvanized strap cast in the

horizontal mortar joint and fastened to the wall by means of dovetail anchor slots which are nailed to the wood studs. Horizontal

wire reinforcing must be placed continuously in mortar joints

containing the anchors. Masonry veneer is not permitted over exitways or more than 25' above grade when the veneer is supported

by wood framing.

Drift

An addition is made to 76 UBC Section 2307 to limit the story drift

to 0.005 times the story height. The calculated story drift due to code level forces shall be multiplied by 1.00/K but this ratio need

not exceed 1.0. Also the drift or the displacement in the plane of

the wall from head to sill at a glazed opening shall not exceed

0.0025 times the height of the opening unless the glass therein is prevented from resisting shear or taking distortion, or tempered,

safety or wire glass is used. Ordinary glass embedded in hard

bedding material will fail at about 0.005 times the glass height.

It is apparent that if a window is framed such that the head of the

opening is anchored to the floor above and the sill is framed and

anchored to the floor below, the permitted story drift is limited

to 0.0025 times the opening height if ordinary glass is used. That

is, if the window is 4 feet high and the glass is bedded into the

frame the story drift limit would be 0.12 inches which is very

stiff, indeed.

24-2
In this same section, 75 UBC Section 2307, an addition has

been made to require a vertical resisting element in each resisting

plane for every 100 feet of building length when the construction

consists of a wood horizontal diaphragm or horizontal rod bracing

and the walls are of wood. In buildings having horizontal wood

diaphragms or rod bracing systems and with continuous structural

steel reinforced concrete framing of continuous reinforced grouted

masonry walls the element may then be at 125 feet of building

length.

Unreinforced masonry is not permitted in school construction.

Requirements have also been added to limit the span-depth ratio

of various diaphragm materials as related to the stiffness of the

iwall materials. For example, a flexible roof diaphragm material

such as diagonal sheathing may have a 2:1 span-depth ratio if the

walls are of wood or light steel but it cannot be used if the walls

re of concrete masonry. If the roof diaphragm is plywood nailed

t all edges of the panel, the span-depth ratio is 3:1 for any kind

f wall materials.

Earthquake Regulations

he lateral force provision in the 76 UBC closely follows the urrent Structural Engineers Association of California Recommended

ateral Force Requirements and Commentary.

The seismic force due to soils acting against walls is a newly

proposed regulation. Where buildings provide lateral support for

i*alls retaining earth, and the exterior grades on opposite sides of

he building differ by more than 6 feet, the load combination of the

ynamic increment of active earth pressure due to earthquake acting

n the higher side, as determined by a civil engineer qualified in

oils engineering, plus the difference in active earth pressures,

hall be added to the lateral forces provided in this Section. In

ieu of a determination by a civil engineer qualified in soils engieering, the dynamic increment of active earth pressure for soils

f low cohesion may be considered to be a constant uniformly distriuted pressures, , determined in accordance with the following

ormula:

P

d = Z(0.036h)

lere

= Dynamic increment of active earth pressure due to earthquake in pounds per square foot

6 = Soil density in pounds per cubic foot

h = Height of retained earth above the bottom lateral

support for the wall

In the base shear equation V = ZIKCSW the influence of snow

oad as related to elevation has been included in the factor W.

lere the design snow load is 30 psf or less, no part need be

lcluded in the value of "W". Where the snow is greater than 30

tef, the snow load shall be included; however, the snow load may be

iduced to 25 percent of the design snow load for sites below 5,000

set in elevation and may be reduced to 33 percent of the design

low load for sites above 5,000 feet in elevation.

24-3
It is quite apparent that heavy snow load at the higher

elevations will have a strong influence on the seismic design.

The maximum measured snow load made by this author to date in

California was 323 pounds per square foot, the snow depth was 141

inches, on a pitched roof of a structure located at 6900 foot

elevation.

When considering the seismic design of single-mass structures

laterally supported by cantilevered columns, the value of K has been

increased to be not less than 2.0. The columns for such structures

shall meet all the requirements for columns in ductible moment

resisting space frames. Broadly speaking, single-mass structures

are one-story structures whose columns have no moment resistance at the top and the moment resistance at the base is provided solely by

the lateral soil pressure on the column foundation.

Overturning

When calculating the internal stresses in a structural system due

to overturning effects produced by horizontal earthquake forces,

the maximum tensile stresses may be reduced by not more than threefourths of the vertical dead load stresses because of the effects

of possible vertical earthquake acceleration. However, the total

dead load resisting moment may be used in computing overturning

forces at the foundation/soil interface.

Building Separations

The required separation distance between buildings has been

increased. It must be determined by multiplying the calculated

lateral force drift or deflection by a factor of 2.0'/K but not less

than 2.0. The total separation shall be not less than the sum of

the displacements calculated for each building at that level. Any

building elements, mechanical or electrical systems which cross or enclose the building separating space shall be designed and detailed

to accomodate these calculated displacements.

Structural Systems. Ductility Requirements

The ductility requirements for school buildings have been raised to require that all moment-resisting space frames used to resist design

seismic forces in any height school building shall meet ductile

moment-resisting space frame requirements.

Exception: Moment-resisting space frames not meeting

ductile frame requirements may be used for one or two story structures provided their design is based on the

basic lateral force formula using a K not less than 2.0.

The columns for such structures shall meet all the

requirements for columns in ductile moment-resisting

space except that columns for single-story steel frames

supporting tributary dead loads, including the weight of

the frame, not exceeding five pounds per square foot need

not comply with the requirements of the steel ductile

moment-resisting space frames.

Braced Frames

The 76 UBC requires that all members in braced frames shall be

designed for 1.25 times the force determined in accordance with the

basic lateral force formula. Connections for all members in braced

frames shall be designed to develop the full tensile or compression

capacity of the member. However, an exception for school buildings

24-4
has been inserted to require that the connection need not develop

the full capacity of the member if the connection is designed for

2.0 times the force determined in accordance with the basic lateral

force formula. Also, all tension members shall be installed in a manner to remove any sag.

Flexible Horizontal Diaphragm Anchorage

Where diaphragms of wood or other materials of similar flexibility

are used to laterally support concrete or masonry walls, the anchorage shall provide a positive direct connection capable of resisting

the forces required by the lateral force formula or a minimum of 200

pounds per foot of wall length. Such connections shall be in

addition to the diaphragm to wall connections provided for shear

transfer.

Anchorage of framing members shall not be provided by use of

toe nails or nails subjected to withdrawal; nor shall wood framing

be used in cross-grain bending or cross-grain tension.

Lateral support of the wall may be provided by a 1/2" or thicker plywood diaphragm fastened with a double row of nails

directly to a 3 x 6 minimum wood plate bolted on the top of the

wall. The nails shall be 1" minimum from each edge of the plate

and 1" minimum from the edge of the plywood.

Lateral Force on Non-Structural Components

Supporting structural elements and anchorage of architectural,

mechanical and electrical components shall be designed for lateral

forces in accordance with the following formula:

F = ZC W

P P P The values of C for determining anchorage forces for architectural, mechanical and electrical components shall be as set forth in

Table 1. Where the provisions of these tables do not specify Cp

values for the anchorage of particular components which in the

opinion of the Office of the State Architect should be designed and

anchored to resist lateral forces for the safety of the occupants,

the office may assign Cp coefficients with the advice of the

architect or engineer based on coefficients specified for similar

components listed in these provisions.

Seismic Hazard Zones

To establish clearly defined enforceable boundaries of the seismic

hazard zones in California, the county lines, location of highways,

latitudes and longitudes are employed. The zones shown in Figure 1 show these proposed boundaries. As a guide to determine these zones

the suggestions given in the Commentary of the Recommended Lateral

Force Requirements and Commentary were used to locate Zone 4 boundaries. It was recommended that Zone 4 should extend 25 miles from

both sides and ends of a fault which could produce a Richter 7.0

jor greater magnitude and 15 miles from a fault which can produce

a 6.0 to 7.0 Richter magnitude earthquake.

Lay-in Panel T-Bar Ceilings

Portions of many lay-in panel T-bar ceilings have failed in earthquakes of magnitude of about 5.6 and greater. Following the 1971

San Fernando earthquake, the ceiling industry developed recommendation now found in the UBC Standards No. 47.18. Briefly, this

24-5
requires that the main and cross tees be anchored at the wall or a continuous member be provided which will prevent the bars from

spreading and dropping the panels, a wire hanger shall be provided

within 8" of the wall, and restraints consisting of four No. 12 gage wires secured to the main runner within 2" of the cross runner intersection and splayed 90° from the plane of the ceiling. These

horizontal restraint points shall be within 4 feet of the wall and

not over 12 feet on center each way. The splices of the main runners and the cross tees must be capable of resisting an ultimate

test load of not less than 180 pounds in tension and compression

provided that only the ceiling load is to be resisted. Where the

ceiling system must resist additional loads from partitions,

shelving, and light fixtures, the test load of at least twice

the design load must be resisted. Also, it must be demonstrated

that this load can be resisted in tension with a 5° misalignment of

the members. In lieu of these arbitrary requirements, a ceiling

may be designed to resist the lateral loads given in Table 1.

Hospital Regulations

Essentially all of the regulations adopted or proposed for school

buildings also apply to hospital buildings. There are, however, a few differences due to the increased level of performance required

by legislative intent — "to be completely functional...." The basic

code will be the 1976 UBC and will follow the "essential facilities"

with the importance factor requirements; i.e., the value in the base

shear equation will be 1.5.

The regulations do, however, stress the requirement that the

static analysis as proposed by the 76 UBC is applicable only to regular buildings. Dynamic analysis will definitely be required for

any structure which has highly irregular shape, larger differences

in lateral resistance or stiffness between adjacent stories or other

unusual structural features which would significantly affect the

dynamic analysis shall be based upon the ground motion prescribed

for the site in a geotechnic report. The report shall consider the

seismic event which may be postulated with a reasonable confidence

level within a 100-year period. The analysis may be based upon

appropriate time-histories or response spectra with percentages of

critical damping consistent with the strain levels in the structural

materials. All natural modes of vibration with periods of vibration

greater than 0.05 second shall be considered in the analysis. The

base shear resulting from the dynamic analysis shall be not less

than 80 percent of the base shear calculated from the basic lateral

force equation. If the base shear, as determined by the dynamic

analysis, must be increased to meet this requirement, the design

response shall be normalized to a proportionately higher value.

Lateral Force on Non-Structural Components

The design and detailing of mechanical or electrical systems and

non-structural components required to be completely functional to perform all necessary services after a disaster shall be based upon

the formula

when Cp is obtained from Table 2. The State review will be made

for the anchorage of the item to the structure and not for the

design or construction of the item.

24-6
Steinbrugge , K. V. and Moran, D. F., "An Engineering Study of

the Southern California Earthquake of July 21, 1952, and of Its Aftershocks," Bulletin of the Seismological Society of America , April 1954, p. 244 and p. 255.

Steinbrugge, K. V., Cloud, W. K ., and Scott, N. H. , "The Santa

Rosa California Earthquakes of October 1, 1969," Coast and

Geodetic Survey , 1970, p. 38 and p. 10.

National Academy of Sciences, National Academy of Engineering,

"The San Fernando Earthquake of February 9, 1971," p. 13.

Jennings, P. C, "Engineering Features of the San Fernando

Earthquake," California Institute of Technology EERC 71-02,

p. 276.

National Bureau of Standards, "Engineering Aspects of the 1971

San Fernando Earthquake," Building Science Series 40, 1971,

p. 311.

Steinbrugge, K. V., Schader, E. E. , "Earthquake Damage and

Related Statistics, San Fernando Earthquake February 9, 1971,"

U. S. Department of Commerce, 1973, Vol. I, Part B, p. 713.

U. S. Department of Commerce, National Oceanic and Atmospheric

Administration, "San Fernando, California, Earthquake of

February 9, 1971," Vol. I, Parts A and B.

Steinbrugge, K. V., Schader, E. E., Bigglestone, H. C, and

Weers, C. A., "San Fernando Earthquake of February 9, 1971,"

Pacific Fire Rating Bureau, p. 2.

24-7
TABLE 1: HORIZONTAL FORCE FACTOR "C " FOR ANCHORAGE OF SCHOOL

NON-STRUCTURAL COMPONENTS*

CATEGORY DIRECTION

OF FORCE

VALUE OF s

1. When connected to, part of, or housed within a building:

<, All mechanical and electrical

equipment or machinery

Any direct ion 0.30 2,3

b. Tanks (plus contents) and support systems

2. Emergency power supply systems^ Any direction 0.75 2 - 3

3. a. Storage racks with the upper storage

level more than 5 feet in height

(rilus content?] ^ Any direction 0.30 3 > 7 b. Floor supported cabinets and bookstacks more than 5 feet in height (plus contents )

4. Kail hung cabinets and storage shelving

(plus contents)

Any direction 0.30

5, Suspended ceiling framing systems

(See Section T21-4701(e))

Any direction 0. 30 8

o

ft ^ii^npnrl^fl 1 i oh f f i Ttnrpc Any direction 1 .00

i 7. Power-cable driven elevators or hydraulic

elevators with lifts over 5 feet:

a. Car and counterweight guides, guide

rails, and supporting brackets and fraTung'O

See Part 7, Title 24

See Part

7, Title

24

b. Driving machinery, operating devices

and control equipment

Wc I Jc d , bolted, or other Intermittent connections such ns inserts shall not be o 1 1 awe J the one third IncreiK in allowable stress permitted,

'c_ may bo two-thirds of vnluc shown for compononts mounted on foundations at prailc or on floor slabs

on earth subgidde. C„ shall be not less than the ratio of Fx/iVx for floor or roof level und._ r consideration. Where n dynamic analysis is used in tho design of the building, the forces so determined

nay be used in the design of die elements or components with appropriate resistance criteria.

^For flexible and flexibly mounted equipment and mjch J nc ry (fundamental period of vibration g re i te r than 0.0S secondsl-the appropriate vn lues of C, t slinll be determined with consideration g t vi*n to both

the dynamic properties of the equipment and machinery and to the building or structure In which it is placed but shu 11 not be less than the listed values. Tut closoly res inn nci! flexible mountings with

resilient snubbcrs, a factor of two times the vnlues shown shall be used except where more appropriate

values can be deter mined by a dynamic analysis.

'"The component anchorage shall be designed for the horizontal force, F p , acting simultaneously with a vertical seismic force equal to one-third of tho horiiontal force. F_. ^Equipment or machinery shall Include such Items as tonVs, boilers, chillers, pumps , no tors, air handling

units, coo 1 i ng towers, transformers, switch gear, control panels, etc. *r«erxcney equipment should be locnted where there Is tho least likelihood of damage due to earthquake.

Such equipment should be located at ground level and whrro It can bo oaslly maintained to assure Its operation du r 1 ng an emergency.

'floor supported storage racks, cabinets or bookstacls not more than 8 feet in height need not be anchored If the width of tho supporting base or width between tho exterior legs is equal to or greater

than two-thirds tho height.

'in lieu of the tabulated values, steel storage racks may be designed In accordance with UPC Standa rd No. 2 7-11.

^Celling weight shall Include all 1 i j*ht ftxtuies and other oqulpment or compononts which are laterally

supported ty the celling. lor purposes of determining the lateral force, a celling weight of not less

than four pound* per square foot shall be lived , "Suspension systems for light f Ixtui cs which h.ive passe J shaking table tests approved by the Office of

the State Atrhiw-ct or which, is i n> I a 1 led, nro fruo to swing a minimum of 45° from the vcitical in all directions shall he assumed to comply with the latum) force requirements of Section TM Mli(l), Unlen of the coblo typo, free swinging suspension systems shall have a safety-wire or coble attached

to the fixture and structure at each support en pab 1 o of supporting four times tho support load. 1°W_ for elevator cars shall bo tho weight of the fur plm 0.4 tlmos its rntcd load. Tho Literal forces acting on ruidr ra 1 1 1 shall be u s ii.m'd to I ft distributed une - third to the top guide rollers anil two- thirds to thu bottom guide rollers of elevator cars and counter weights unless other substantiating

data li provided.

24-8
24-9
TABLE 2: HORIZONTAL FORCE FACTOR "C " FOR ANCHORAGE OF HOSPITAL

NON-STRUCTURAL COMPONENTS*

13

CATEGORY DIRECTION

OF FORCE

VALUE OF Cp 1

• 1

1 = 1.0 1 = 1.5

1. Mechanical and electrical components

a. Equipment or machinery 5 not

required for life safety systems or for continued operations of

essential facilities

Any

direction

U « Si) * V . 50 » b. Equipment or machinery 3 required for

life safety systems or for continued

operation of essential facilities^ 0 . 75 3 »* 0.75 3 '* c. Essential communication equipment

and emergency power equipment such

as engine generators, battery racks

and fuel tanks necessary for

operation of such equipment 0 . 75 3'

1* 0. 7S 3

• " d. Piping, electrical conduit, cable

trays, and air handling ducts 0.50 7 O.SO 7

2. Hospital equipment 6 when permanently

attached to building utility services

such 3.s '. surgical, morgue and recovery

room fixtures, radiology equipment, food

service fixtures, essential laboratory

equipMcni-, i v buppuitb, etc.

Any

direction 0 . 30 3 0.50 3 '"

3. a. Storage racks with the upper

storage level more than 5 feet in

height (plus contents) 8 Anydirection 0 .30 3 0 . 50 3

.

1' b. Floor supported cabinets and bookstacks more than 5 feet in height

(plus contents) 8 0.30 3 '* O.SO 3 '"

4. Wall hung cabinets and storage shelving

(plus contents)

Any

direction 0 . 30 0 . 50

5. Suspended ceiling framing systems

(See Section Tl 7 - 4 701 (e )

)

Any

direction 0.30 9 0.50 9

6. Suspended or surface mounted light

fixtures 10 Any

di rection 1 .00 1 .00

7. Power-cable driven elevators or

II) U I dlii IL Civ Vu LUl j n 1 L i 1 J. 1 I L U V L I J feet

:

a. Car and counterweight guides, guide

rails, and supporting brackets and

framing 11

Sec Part 7, Title 24 Sec Part 7,

Title 24

b. Driving machinery, operating,

devices and control equipment

24-10
Welded, bolted or other intermittent connections such as inserts shall

not be allowed the 1/3 increase in allowable stress.

'Cp shall be not less than the ratio of Fx/V.'x for floor

or roof level under consideration. Where a dynamic

analysis is used in the design of the building, the

forces so determined may be used in the design of the

elements or components with appropriate resistance

criteria.

2por flexible and flexibly mounted equipment and machinery (Fundamental

period of vibration greater than 0.0S seconds), the appropriate values

of C p shall be determined with consideration given to both the dynamic

properties of the equipment and machinery and to the building or structure in which it is placed but shall not be less than the listed

values. For closely restrained flexible mountings with resilient

snubbers a factor of 2 times the values shown may be used except

where more appropriate values can be determined. Where a dynamic

analysis is not furnished the value Cp shall be doubled.

'For equipment and machinery mounted on foundations at grade or floor

slabs on earth subgrade the value of Cp may be reduced 1/3.

The component anchorage shall be designed for the horizontal force, Fp,

acting simultaneously with a vertical seismic force equal to one-third

of the horizontal force, Fp

.

s Equipinent or machinery shall include such items as tanks, boilers,

chillers, pumps, motors, air handling units, cooling towers, transformers,

switch gear, control panels, etc.

6 Emergcncy equipment should be located where there is the least likelihood

of damage due to earthquake. Such equipment should be located at ground level and where it can be easily maintained to assure its

operation during an emergency.

^Seismic restraints may be omitted from the following installations:

(a) Gas piping less than 1 incli inside diameter.

(b) Piping in boiler and mechanical equipment rooms less than

Ik inch inside diameter.

(c) All other piping less than 2h inch inside diameter.

(d) All piping suspended by individual hangers 12 inches or less in length from the top of pipe to the bottom of the

support for the hanger.

(e) All electrical conduit less than 2h inch inside diameter.

(f) All rectangular air handling ducts less than 6 square feet in cross sectional area.

(g) All round air handling ducts less than 28 inches in

diameter.

(h) All ducts suspended by hangers 12 inches or less in length

from the top of the duct to the bottom of the support for

the hanger.

24-11
8 In lieu of the tabulated values steel storage racks

may be designed in accordance with UBC Standard

No. 27-11.

9 Ceiling weight shall include all light fixtures and other equipment which

are laterally supported by the ceiling. For purposes of determining

the lateral force, a ceiling weight of not less than 4 pounds per square

foot shall be used.

"Suspension systems for light fixtures which have passed shaking

table tests approved by the Office of the State Architect or which, as installed, are free to swing a minimum of 45° from the

vertical in all directions shall be assumed to comply with the

lateral force requirements of Section T17-2512(l).

Unless of the cable type, free swinging suspension systems shall

have a safety wire or cable attached to the fixture and structure

at each support capable of supporting 4 times the support load.

*Wp for elevator cars shall be the weight of the car plus 0.4 times

its rated load. The lateral forces acting on guide rails shall

be assumed to be distributed 1/3 to the top guide rollers and

2/3 to the bottom guide rollers of elevator cars and counter

weights.

24-12
25

SUMMARY OF RESEARCH PROJECTS IN THE LARGESTRUCTURES TESTING LABORATORY FROM 1967-1977

Masaya Hirosawa

Yuji Ishiyama

Tetsuro Goto

Ministry of Construction

Summary of Research Projects in the

Large-Structures Testing Laboratory

from 1967-1977

Structural design should always be carried out on the basis of

theory and test results. Test results obtained by using a smallscale model of a structure or a full-scale test of structural

components give us invaluable information on the behavior of

structures, but they do not always satisfy the practical engineer's

requirements. The problem of dimensional similarity between the

prototype and model structures frequently occurs.

To solve such problems, the Ministry of Construction suggested

the construction of a Large-Structures Testing Laboratory and the

budget was allocated to the Japan Housing Corporation in the fiscal

year of 1966. The planning and basic design of the Laboratory were

performed by the staffs of the Building Research Institute and the

Japan Housing Corporation, and the Laboratory was built in 1967.

This Laboratory consists of a L-shaped testing bed and vertical

reaction wall of prestressed concrete and the shed of reinforced

concrete with several testing rooms. The height of the reaction

wall is 16 meters and the size of the testing bed is 9.5 meters by

20 meters. Horizontal loads can be applied both in pushing and

pulling directions against the wall and the vertical loads can be

applied simultaneously keeping the vertical loads constant by the

accumulation of oil. The capacity of the reaction wall is 1,000

tons in shear and 10,000 ton-meters in bending moment. Accordingly,

five storied full-scale reinforced concrete buildings with two housing units on each floor can be tested in this Laboratory (see

fig. 1).

Since the completion of the Laboratory, research projects testling full-scale or large-scale specimens of the buildings have been

carried out. Summaries of the main projects are reported in the

following chapters. 1977 will be its last year because the

Laboratory will be demolished in 1978 or 1979 with the removal of

'Building Research Institute to Tsukuba Science City.

1967: Test on a Five-Story Apartment House of Precast Reinforced

Concrete Panels

The test specimen was a full-scale five-story building with two

jliving units on each floor designed by Japan Housing Corporation

'iaccording to the general design method. The plan of the specimen

|was 14.3 meters long and 6.1 meters wide, the height was 12.8

peters, and the weight was 430 tons including the live load, which

|was not actually loaded in the test. The wall thickness was 15. cm

land the wall ratio was 28.3 cm/m^ at each floor. The vertical

joints were the so-called wet joints that use cement mortar; the

horizontal joints were dry joints that connect the embedded steel

^plates or bars with welding (see fig. 2).

The same amount of horizontal load was applied at each floor

level in the longitudinal direction; these loads were alternated by

bulling and pushing with three oil-jacks at each floor; 15 in total.

Bending cracks appeared in the beams at the design load, which

can be calculated multiplying the sum of the dead load and the live

load by the seismic design coefficient that is generally 0.2

according to the Japanese Building Standard. As the load increased

bending cracks in beams and shear cracks in walls progressed. The

naximum load carrying capacity was 651 tons, which was 7.6 times

25-1
the design load of 86 tons. The horizontal displacements at the maximum load were 4.7 cm at the top of the specimen and 1.9 cm

at the second floor, which corresponded to 1/270 and 1/131 of the

respective heights. As the deflection increased after this maximum

point, the load-carrying capacity decreased gradually to 0.77 of

the maximum load at a displacement of 16.7 cm at the top, which

was 1/77 of the height. During this loading, shear cracks in walls

progressed significantly and many new shear cracks appeared in the

beams. The final failure mechanism was shear failure of the first

and the second story walls. It is significant that the specimen

could endure large deflection after this maximum load without

appreciable strength deterioration (figs. 3 and 4).

Several conclusions were drawn from this work:

(1) This specimen would have enough reserve strength

against severe earthquakes. Accordingly, it would

be possible to reduce the wall ratio.

(2) It would be possible to reduce the number of reinforcing bars in the wall panels and vertical reinforcement in the vertical joints.

(3) The following assumptions could be used in design

calculations: (a) the rigidity of the floor slabs

could be assumed as infinite, and (b) the vertical

joint could be considered as monolithic.

A recommendation for the structural design of such precast

concrete buildings was completed in 1971 and thousands of apartment

houses have been constructed by Japan Housing Corporation with this

construction method.

1968: Test on a Five-Story Apartment House of Monolithic Reinforced

Concrete

The plan of the specimen was 13 meters long and six meters wide.

The height was 13 meters and the weight was 450 tons including the

live load, which was not actually loaded in the test. The wall

thickness was 15 cm and the wall ratio was 12 cm/m^ at each floor.

The specimen was designed to have a wall ratio and wall thickness

that were smaller than the requirements of the Japan Housing

Corporation (see fig. 5).

The horizontal loads were applied in the longitudinal direction

same as in the 1967 project.

Bending cracks appeared in the beams at the design load and

shear cracks in the walls gradually appeared as the load increased.

The maximum load-carrying capacity was 390 tons, which was 4.1

times the design load of 90 tons. The displacements at the maximum

were 4.9 cm at the top of the specimen and 1.8 cm at the second

floor. These values were 1/266 and 1/140 of the respective heights.

Thereafter the displacement at the top reached the maximum of 7.3 cm, was 1/180 of the height with a decreased strength to 0.74 of the

maximum value. Shear failure in the first-story walls was critical

to the load-carrying capacity (see figs. 6 and 7).

This work resulted in the following conclusions:

(1) This specimen was provided only with the minimum

load capacity and ductility. According to the

25-2
dynamic analysis, this structure would be damaged

heavily by a strong earthquake.

(2) Positions and sizes of openings in the walls should

be changed for uniform distribution of wall rigidit ies

.

(3) The wall in the first story should be reinforced

with more steel bars to improve the ductility of the

building.

From the results of this test, it became clear that the structural regulations that had been used for buildings up to four

stories could also be used for the five-storied ones. The Japan

Housing Corporation issued the structural design specifications of

five-storied buildings of reinforced concrete wall construction.

Wall ratios and wall thicknesses specified by the Japan Housing

Corporation differ from the specimen of this test as follows:

Specimen Specification

Wall thicknesses

Wall ratios

15 cm

12 cm/m

18 cm (in the longitudinal direction)

15 cm (in the transverse direction)

15 cm/m

1969: Test on an Eight-Story Apartment House of Precast Prestressed

I Concrete

j The project was to test a eight-story apartment house of precast

prestressed concrete. This structural system had been developed

in 1965 at the Building Research Institute. The main purpose

was to obtain data to design and construct building that use this

structural system.

The test specimen was the full-scale four-storied building

; corresponding to the lower four stories of an eight-story apartment

house. Each floor had two of the half living units. This structure

I consisted of precast reinforced concrete structural elements, i.e.,

f L, +, T, and I shaped wall columns, girders and slabs. The vertical

. connections between columns and girders are made by a posttensioning method. The prestressing bars anchored in the foundation

are extended through the column and girder coupling at each floor

level. These bars are tensioned and grouted after completion of

|| assembly. The joints between floor slabs and girders are made by

' welding the steel plates anchored in each element. The plan of the

specimen was 11.7 meters long and 4.6 meters wide; the height was

10.8 meters and the weight was 600 tons including the upper four

stories dead load and the total live load. The wall thickness was

2

18 cm., and the wall ratio per unit floor area was 13.7 cm/m (see

fig. 8).

The horizontal loads were applied in the longitudinal direction

by eight hydraulic jacks at the top of the specimen, four at the

j fourth floor level and two at the third and second floor levels.

The vertical loads that corresponded to upper four-story dead load

and the total live load of an eight-story actual building were also

25-3
applied by 12 hydraulic jacks connected to the accumulator to keep

the vertical load constant during the application of the horizontal

load

.

Fairly linear relations between the loads and the displacements

were observed up to 1.2 times the design load of 120 tons. No cracks in the columns but slight bending cracks in the beams were observed at this load. The bending cracks in the beams developed

to shear cracks at 2.9 times the design load, while the initial

shear cracks in the columns appeared at 3.2 times the design load.

The maximum load-carrying capacity was 3.9 times the design load.

The displacements at the maximum load were 10.5 cm at the top of

the specimen and 3.3 cm at the second floor, which were 1/103 and

1/82 of the respective heights. The final failure mechanism was the shear failure of second and third floor beams; further repetition of the load did not cause any significant decreasing of the

load-carrying capacity. At the final load the displacements reached

to 1/62 at the top of the specimen and 1/47 at the second floor of

the respective heights (see figs. 9 and 10).

The following conclusions were drawn:

(1) This specimen had a considerable ductility compared

to the shear wall structures.

(2) In the actual eight-story buildings, the final

failure would be caused by the failure of the

first story columns and the shear reinforcement of

the columns should be increased.

(3) Because of the early shear failure of the beams,

shear reinforcement of the beams should be increased

by twice of that of the specimen.

(4) The closely spaced reinforcement at the corner of

the columns was very effective in preventing the

development of shear cracks and the crushing of the

enclosed concrete.

(5) From the dynamic analysis the specimen had a sufficient capacity to endure severe earthquakes with the

maximum acceleration 300 gals.

In 1970, an eight-story apartment house with 104 units was built in Tokyo with this structural system. Since then, two thousand such housing units have been constructed.

1970: Test on a Five-Story Apartment House of Monolithic Reinforced

Concrete with Staggered Plan and Elevation

The test specimen was the full-scale five-story building and had

five living units on one side and four on the other. The floor

levels of both sides were staggered by half of a story and the

longitudinal walls of both sides were also staggered by the half a wall. The plan of the specimen was 13 meters long and 5.4 meters

wide, the height was 14 meters and the weight was 450 tons including

the live loads, which were not actually loaded in the test. The

wall thicknesses were 18 cm from first to the fourth story and

15 cm at at the fifth story. The wall ratio was 15 cra/m' at each

floor (see fig. 11). The horizontal loads were applied in the

longitudinal direction same as the 1967 project.

25-4
At the design load there were only few hair cracks at the corner of openings. At three times the design load, diagonal shear

cracks occurred. The maximum load was 452 tons, which was 5.0

times the design load. The displacements at the maximum load were

8.8 cm at the top of the specimen and 2.8 cm at the second floor,

which were 1/150 and 1/100 of the respective height. At the maximum load, the first story walls failed in shear severely and bars

in the walls were exposed.

The load-carrying capacity decreased after the maximum load

with increased displacement and at the maximum displacement the load

capacity was 0.65 of the maximum load. At this load the displacements were 1/100 at the top and 1/35 at the second floor of the

respective heights and the shear failure progressed continuously.

A great many broken pieces of concrete fell down (see Figs. 12 and

13).

The following conclusions were drawn:

(1) This specimen had enough strength to withstand

severe earthquakes.

(2) The effect of staggered plan and elevation would not cause any structural defects.

(3) The first story, which had higher floor level than the

other, would be damaged most heavily.

(4) The cracks around the small openings of the first

story walls would be 20 mm wide at the maximum load,

so more ductility around the openings would be

advisable.

1971: Test on a Ten-Story Apartment House of Precase Prestressed

Concrete

The test specimen was the full-scale four-storied frame model

corresponding to the lower four stories of a ten-storied apartment

house. The plan of the specimen was 7.2 meters long and two meters

wide, the height was 10.8 meters and the weight was 518 tons including the upper six stories dead load and the total live load.

! The wall thickness was 18 cm and the wall ratio per unit floor

area was 17.4 cm/m^ (see fig. 14).

A horizontal load was applied in the transverse direction of

j the building, i.e. in the longitudinal direction of the specimen,

, by eight hydraulic jacks at the top of the specimen. The vertical

load, which corresponded to the upper six stories dead load and the

total live load of a ten-story actual building, was intermittently

applied by 12 hydraulic jacks.

Fairly linear relations between the loads and the displacements

were observed up to two times the design load of 104 tons. No cracks appeared at the design load. Bending cracks in the beams and

ij vertical cracks along the vertical joints appeared at two times the

design load. At three times the design load, the bending cracks in

the beams and the columns developed into shear cracks. The maximum

load carrying capacity was 400 tons, 3.85 times the design load.

The displacements at the maximum load were 4.1 cm at the top of the

specimen and 1.4 cm at the second floor, which were 1/261 and 1/188

of the respective heights. The final failure mechanism was the

shear failure of the second floor beams. At the final load, 2.4

times the design load and 0.61 of the maximum load, the displacements

25-5
reached 18.3 cm at the top and 6.0 cm at the second floor, which were 1/59 and 1/45 of the respective heights (see Figures 15 and

16).

This test underscored the following:

(1) The final failure mechanism would be the shear

failure of the beams both in the longitudinal

direction and transverse direction.

(2) At three times the design load, cracks would be

s 1 i gh t

.

(3) The maximum load-carrying capacity would be four

times the design load.

About two thousand housing units have been constructed with

this structural system in Tokyo and Kanagawa Prefecture. The number

of stories are five or eight. A ten-story apartment house like this

has been constructed in Shizuoka Prefecture; it has 70 living units.

1972: Test on an Eight-Story Apartment House of Reinforced Concrete

The specimen was a full-scale four-story building corresponding to

the lower four stories of an eight-story apartment house and each

story had two living units. The plan of the specimen was 11.2

meters long and 7.0 meters wide, the height was 10.4 meters and the

weight was 1,090 tons including the upper four stories dead load

and the total live load. The wall column thickness was 30 cm and

2 the wall ratio per unit floor area was 6.6 cm/m in the longitudinal

direction, while in the transverse direction the wall thickness was

18 cm and the wall ratio was 22.8 cm/m (see fig. 17).

The horizontal load was applied in the longitudinal direction

by eight hydraulic jacks at the top of the specimen, four or eight at the fourth floor level and two at the third and second floor levels.

The vertical load, which corresponded to the upper four stories dead

load and the total live load of a eight-storied actual building, was also applied by 12 hydraulic jacks keeping the vertical load constant

Fairly linear relations between the loads and the displacements

were observed up to two times the design load of 218 tons. Bending

cracks appeared in the beams at half the design load. After that

the bending cracks in the beams and the wall columns progressed

gradually and at two times the design load shear cracks in the wall

columns were observed from the first story to the fourth story.

Cracks did not concentrate but spread over the beams and the wall

columns. As the load increased, the bending cracks at the end of

the beams became severe and at the final loading the first story

wall columns were crushed. No deterioration of strength was seen

at the final loading. The final load was 990 tons, 4.5 times the

design load. The displacements were 15 cm at the top of the speci- men and 7.0 cm at the second floor, which were 1/69 and 1/37 of the

respective heights (see Figs. 18 and 19).

The maximum displacement at the top was limited by the strokes

of the hydraulic jacks.

The following points were concluded:

(1) The final failure mechanism would be the bending

failure of the beams.

25-6
(2) The maximum load carrying capacity was 4.5 times

the design load. Accordingly, the strength could be

minimized within limits not spoil the ductility of

the structure.

(3) The minimum required ductility was satisfied, which

was considered to be 1/50 of the height.

1973: Test on a Eleven-Story Apartment House of Precast Concrete

The test specimen in the longitudinal direction was the full-scale

four-story frame with two spans, which corresponded to the lower

four of an eleven-story 14-span apartment house. The basic precast

concrete members were beams with halves of upper and lower columns

on both ends. The specimen was 11.1 meters long, the height was

10.8 meters, and the weight was 650 tons including the dead load of

the upper seven stories and the total live load. The columns were

87 cm wide and 55 cm thick. The beams were 40 cm thick and 55 cm

high. The special coupler joints were used to connect the beams.

Mortar-filled sleeve joints were adopted to connect the columns (see

fig. 20).

The test specimen in the transverse direction was a half-scale

eight-story shear wall with one span, which corresponded to the

lower eight stories of an eleven-story with one span apartment

house. The main precast concrete members were the columns with

beams, which also composed the longitudinal rigid frame, and the

shear walls with the beams on the tops. The specimen was 4.5 meters

long, the height was 10.8 meters, and the weight was 125 tons including the dead load of upper three stories and the live load.

The cross section size of the columns was 27.5 cm by 43.5 cm and the

walls were 7.5 cm thick with 15 cm by 25 cm beams at the tops. To connect between the columns, mortar filled sleeve joints were used

and the embedded bars were welded to connect the columns to the

beam. To connect the walls to the columns and the beams, so-called

wet joints were adopted using the spiral reinforcements (see fig.

21).

In the longitudinal direction, the horizontal load was applied

at the top of the specimen by eight hydraulic jacks. The vertical

load was also applied by 12 hydraulic jacks.

In the transverse direction, the horizontal load was applied

at the top of the specimen by two oil jacks. The vertical load was

also applied by four oil jacks.

In the longitudinal direction, bending cracks appeared in the

third floor beams at the load of 40 tons, which was 0.31 of the

design load of 130 tons. Fairly linear relations between the loads

and the displacements were observed up to the design load. At the

design load, cracks appeared at the connections between the beams

along the main bars of the beams, diagonal cracks appeared at the

panel zones of the columns and the beams, and the displacement at

the top of the specimen was 3.7 cm which was 1/292 of the height.

At 1.2 times the design load, the compression bars in the columns

began to yield and at 1.4 times the design load the tension bars

in the beams began to yield, causing a gradual decreasing of the

stiffness. At 1.7 times the design load, the outside first story

column began to crush and the displacement at the top was 11.4 cm, which was 1/94 of the height. After this load, the displacement

increased without an obvious increase in load. At the maximum

load, 1.8 times the design load, the displacement at the top became

21 cm, which was 1/51 of the height. At the final stage, the

25-7
displacement at the top became 43 cm, which was 1/25 of the height

without any evident decreasing of the load. The final failure

mechanism was the bending failure of the columns and the beams

(see figs. 22 and 23).

In the transverse direction, fairly linear relations between

the loads and the displacements were observed up to the design

load of 25 tons. At the design load bending shear cracks appeared

in the first story wall, bending cracks also appeared at the joint

between the first story columns, and the displacement at the top of

the specimen was 0.63 cm, which was 1/1690 of the height. At 1.3

times the design load, bending cracks in the foot of the first

columns appeared and at 1.4 times the design load, shear cracks

appeared in the walls from the fourth to the sixth story, causing

a gradual decreasing of the stiffness. The maximum load was 1.9

times the design load. At this load the displacement at the top

was 4.2 cm, which was 1/256 of the height. At the final stage,

the load was 0.85 of the maximum load and the displacement at the

top was 7.2 cm, which was 1/147 of the height. The final failure

mechanism was the yielding of the main tension bars of the firststory columns, which caused a slippage between the foundation and

the first-story shear wall (see figs. 24 and 25).

This test led to the following conclusions:

(1) Since the final failure mechanism of the rigid frame

in the longitudinal direction was a bending failure

of the columns and beams, it was felt that this frame

had a considerable ductility to resist a strong

earthquake

.

(2) The final failure mechanism of the shear wall in the

transverse direction was the yielding of the main

tension bars of the first story columns; there was no obvious destruction in the other parts. The connections between the main bars of the columns should be

improved

.

1974: Test on a Eight-Story Apartment House of Precast Concrete

The test specimen was the full-scale five-story shear wall that

corresponded to the lower five stories of an eight-story apartment

house in the transverse direction. The wall was 18 cm thick precast concrete from the first to th fourth story; the fifth-story

wall was cast-in-place reinforced concrete. The test specimen was

8.3 meters long, the height was 13 meters, and the weight was 283

tons, including the upper three stories dead load and the total

live load (see Fig. 26).

In the longitudinal direction the test specimens were the fullscale precast members of the beams with columns; test results are

not discussed here.

The horizontal load was applied at the top of the specimen by

six hydraulic jacks. The vertical load was applied by eight oil

jacks

.

Fairly linear relations between the loads and the displacements

were observed up to 1.5 times the design load of 57 tons. At the

design load, bending cracks appeared in the wall beams that faced

the openings. Shear cracks occurred at 2.5 times the design load.

The maximum load was five times the design load. At this load,

the displacement on top of the specimen was 7.0 cm, which was 1/200

25-8
of the height. At the final stage the displacement at the top

reached 30 cm, which was 1/46 of the height and the load became

0.67 of the maximum load. The final failure mechanism was the shear

failure of the wall beams (see figs. 27 and 28).

This test resulted in the following conclusions:

(1) The specimen had enough strengtH7fK7resist strong

earthquakes

.

(2) The allowable maximum displacement at the top of the

specimen would be 7.0 cm, which was 1/200 of the

height

.

(3) Shear failure in the wall beams should be avoided.

1975: Test on an Eleven-Story Apartment Hous e of Steel-Encased

Reinforced Concrete

The test specimens were a half-scale five story shear wall corresponding to the lower five stories of an eight-story apartment house

in the transverse direction. The first specimen (A) was a cast-inplace steel-encased reinforced concrete wall without openings; the

second one (B) was a cast-in-place steel-encased reinforced concrete

wall with openings; and the last one (C) was a precast steel-encased

reinforced concrete wall. Each specimen was 4.3 meters long, the

height was 6.8 meters, and the weight was 180 tons, including the

upper six stories dead load and the total live load. The wall

thicknesses were 7.5 cm for the one with no openings (A) and 9.0 cm

for the ones with openings (B) and (C). The size of the columns was 42.5 cm by 30 cm and the beams were 20 cm by 37.5 cm (see fig. 29).

The horizontal load was applied at the top of the specimen by

two oil jacks. The vertical load was also applied by eight oil

jacks not only to compensate for the upper six stories dead load and

the total live load of the eleven-story building but to give the

same shear span ratio as to the actual shear wall applying more load

at the compression side than the tension side.

Specimen (A) (no openings)

Fairly linear relations between the loads and the displacements were

observed up to the design load of 26 tons. At this load, bending

cracks appeared in the foot of the first-story column and the

displacement at the top of the specimen was 0.4 cm, which was

1/1680 of the height. At 1.2 times the design load, shear cracks

appeared in the walls. The maximum load was 3.5 times the design

load and the displacement at the top was 1/100 of the height. At the final stage, the load decreased to 0.1 of the maximum load.

After a triple repetition the displacement was 1/100 of the height.

The final failure mechanism was the shear failure of the wall and

the compression side column (see Figs. 30 and 31).

Specimen (B) (with openings)

Bending cracks appeared in the foot of the first-story column at 0.8

times the design load of 36 tons. Fairly linear relations between

the loads and displacements were observed up to the design load.

The displacement at the top of the specimen was 0.7 cm, which was

1/960 of the height. At 1.2 times the design load, shear cracks

appeared in the walls. The maximum load was 3.2 times the design

load; the displacement at the top was 1/100 of the height. At the

final stage, the displacement at the top was 1/60 of the height and

25-9
the load became 0.6 of the maximum load. The final failure mechanism was almost the same as specimen (A). See Figures 32 and 33.

Specimen (C) (precast concrete with openings)

Bending cracks appeared in the foot of the lst-story column and in

the 2nd and 3rd floor beams facing to the openings at 0.7 times the

design load of 26 tons. At the design load, shear cracks appeared

in the walls from first to fourth story and the displacement was

0.7 cm, which was 1/960 of the height. Fairly linear relations

were observed up to the design load. As the load increased, shear

cracks in the walls progressed to bending shear cracks and bending

cracks in the beams progressed to shear cracks. The maximum load was 3.4 times the design load and the displacement at the top was 1/100 of the height. Diagonal bars in the compression side firststory shear wall buckled at the maximum load. At the final stage,

the displacement at the top reached 1/50 of the height and the

load decreased to 0.7 of the maximum load. The final failure

mechanism was almost the same as specimens (A) and (B). See Figs. 34 and 35.

This test resulted in the following conclusions:

(1) Bending failure occured before the shear failure in

all three specimens.

(2) After the maximum load, the stiffness decreased

because of the shear failure in the first-story

walls and columns.

(3) Allowable displacements would mainly depend on whether the shear walls had openings or not.

1976: Test on an Eleven-Story Apartment House of Steel-Encased

Reinforced Concrete

The test specimen was a half-scale five-story building corresponding

to the lower five stories of an eleven-story apartment house. The

plan of the specimen was 4.3 meters long and 2.7 meters wide, the

height was 6.3 meters, and the weight was 240 tons, including the

upper six stories dead load and the total live load. The thickness

of the shear walls was 9.0 cm and the thickness of the partition

walls was 6.0 cm (see Fig. 36).

The horizontal load was applied at the top of the specimen by

four hydraulic jacks. The vertical load was also applied by eight

hydraulic jacks not only to compensate for the upper six stories

dead load and the total live load but to give the same effect

as that of an eleven-story building, changing the tension and the

compression side vertical loads according to the overturning moment

caused by the horizontal load.

Fairly linear relations between the loads and the displacements

were observed up to the design load of 48 tons. At 1.3 times the

design load, bending cracks appeared in the lst-story columns and

partition walls. At two times the design load, shear cracks

appeared in the shear walls and at three times the design load,

shear cracks appeared in the partition walls. At 4.7 times the

design load, the displacement at the top of the specimen reached

1/200 of the height, shear cracks spread over the shear walls and

partition walls, and the reinforcements yielded in the tension side

of the first story columns. The maximum load was 5.7 times the

design load and the displacement at the top was 1/100 of the height.

25-10
At the final stage, the displacement at the top reached 1/50 of the

height and the load decreased to 0.6 of the maximum load. The final

failure mechanism was the shear failure of the first-story shear

walls and columns (see Figs. 37 and 38).

This test resulted in the following conclusions:

(1) The partition walls bear 0.3 of the horizontal load

at the early stage.

(2) At the final stage the partition walls could not resist the horizontal load because of the slippage

between the foundation and the walls.

(3) The final failure was the shear failure of the firststory shear walls and compression side columns.

1977: Test on an Eleven-Story Apartment House of Steel-Encased

Reinforced Concrete

The test specimens were a half-scale three-story frame corresponding

to the lower three stories of eleven-story apartment house in the

longitudinal direction. Specimen (D) was a cast-in-place steelencased reinforced concrete frame without walls, specimen (E) had

walls along the columns, and specimen (F) had walls with openings

in the frame. Each specimen was 6.5 meters long, the height was

4.2 meters, and the weight was 250 tons including the upper eight

stories dead load and the total live load. The columns were 30 cm

thick and 42.5 cm wide. The beams were 22.5 cm thick and 37.5 cm

deep. The wall thickness was 6.0 cm (see fig. 39).

The horizontal load was applied at the top of the specimen by

four hydraulic jacks. The vertical load was also applied by six

hydraulic jacks corresponding to the upper eight stories dead load

and the total live load.

Although precise data are not yet available, the following

conclusions could be drawn:

(1) The effect of the nonbearing walls would be 0.2 of

the horizontal stiffness and strength at the early

stage.

(2) At the final stage, nonbearing walls would bear no load

.

(3) The nonbearing wall would cause no significant

structural defects.

References

1. Koizumi, Y. and Others, "Lateral Load Test on Multi-storied

Full-size Building of Reinforced Concrete Wall Construction for

Earthquake Resistant Design", BRI Research Paper No. 49, May

1971, Building Research Institute, Ministry of Construction.

2. Annual Reports of BRI, from 1967 to 1976, Building Research

Institute, Ministry of Construction.

25-11
FIGURE 1. LARGE SIZE STRUCTURES TESTING LABORATORY (Unic:mm)

25-12

...
Plan Plan Plan

Skeleton(A)

Skeleton(B)

Skeleton(C)

FIGURE 2 (1967)

\AV\ PH

PI PI

Pi pi !!pp

13,000

Skeleton(A)

Skeleton(B)

LSI

Skcleton(C)

FIGURE 5 (1968)

LO

11,770

Skeleton(A)

Skeleton(B)

FIGURE 8 (1969)

Plans and Elevations of Test Specimens (from 1967 to 1969) (Unitimm)

24-13
Plan Plan

00

sua

0 00

13,000

Skeleton(A)

0^L

Skeleton(B)

CZI B ?

EZI*f ru

EL

33

01

J21

11.160

Skeleton(A)

01

03 rr—

i

01

01

Skeleton(B)

FIGURE 17 (1972)

Skeleton(C)

FIGURE 11 (1970)

Plans and Elevations of Test Specimens (1970 and 1972) (Unit -mm)

' - 27 -

25-14
! r—

x * V

\ \

\, / /N X

/ N

i

i X

i X X

1 3600 1 3Gm 1

FIGURE 14: TEST SPECIMEN (1971)

Prototype Building (1973)

Ji

-t*

-

•4- i Z1

I"

J

ir

t = 75

\-«312.5~\ (Unit :mm)

FIGURE 20: TEST SPECIMEN (1973) FIGURE 21: TEST SPECIMEN (1973)

25-15
I *«0- i _i » » » FIGURE 26 TEST SPECIMEN (1974)

Specimen(A) Specimen(B) Specimen(C)

FIGURE 29 TEST SPECIMEN (1975)

Specimen(D) Specimen(E) Specimen(F)

FIGURE 39 TEST SPECIMEN (1977)

25-16
A

Aon

-P/5

Load Displacement Curves (from 1967 to 1970 and 1972) R^(X10'J)

25-17
25-18
FIGURE 30 LOAD DISPLACEMENT CURVE OF SPECIMEN(A) (1975)

•IGURE 32 LOAD DISPLACEMENT CURVE OF SPECIMEN(B) (1975)

5 :~T

25-19
25-20
Skeleton(C)

llEMAtKS P. P.W. Rupluro of woldmg bll.lln ulluig plold.

P- R.W. Ruptur* of welding b*lw««n tailing ploio ond onchorod bo

P.O. Delor modon of tolling ptoto. P i-.D. Stvoro daformolion of tolling plolo. R.B. Rupiuro of onchorod bor. W R.B. Rvplwro of roinforcomont m ponol. S.A. Shoo, crock (|.,, ,hon Imm)

S.6. Shoo, crock (I ,„ j m„)

"ock I J 10 |Omm) 5.D. Shoo, „ock Uorgo, Ihon 10 mm ) OA. Biding crock ( 0. 5 lo t mm ) B.B. Bonding crock ( I lo 3 mm ) fUnding c„ck ( (org., ,Ko„ 3mm ) M r

k octo ""K""»'' <> y cr„,h of control. Cock occompom.d b, .polling- off of cone. oi.

FIGURE 4 FINAL CRACK PATTERNS (1967)

25-21
: r*Vf IF

la

i • -B,£-

.f ' -

'-<'. - ^

I <i ^ I ft ^

* v V • ;•••« i

2 sszze

a X3

Skelton (A)

3

2

EE CZL

Skelton ( B

)

iff'

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Skelton (C)

"Si'

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^'IGURE 7 - FINAL CRACK PATTERNS .( 1968)

mih i

2 :/

25-22
25-23

FINAL CRACK PATTERNS (1971, and from 1973 to 1976)

25-25

26

DYNAMIC BEHAVIOR OF RECTANGULAR WATER TANKS

ASSEMBLED WITH PANELS

Keiichi Ohtani

Chikahiro Minowa

National Research Center for Disaster Prevention

Science and Technology Agency

i

-

Dynamic Behavior of Rectangular

Water Tanks Assembled with Panels

The dynamic behavior of rectangular tanks were studied by W. G. Housner and Fumiki Kito. In the formula developed by Housner, the

elasticity of the side walls was not considered. Dr. Kito analyzed

the natural frequencies of elastic side walls, by the energy method,

assuming simply supported edges, clamped edges, and so on.

This paper describes theoretical and experimental studies concerning rectangular water storage tanks. In the theoretical study,

the Kito's method is employed. The bending displacements of the

walls perpendicular to an excitational direction (which are called

the pressure walls), and the shear displacements of the walls

parallel to this direction (which are called the side walls) are

assumed-. The water in the tank is regarded to have the velocity

potential that satisfies the conditions of free surfaces, the continual conditions of velocities on the pressure walls, and the

three-dimensional Laplace's equation. From these assumptions, the

potential energies and the kinematic energies would be obtained.

The vibrational equations of this system would be expressed by substituting these energies into the Eular's equation. This analysis

is concluded by solving multi-degrees vibrational equations.

The dynamic tests concerned two tanks with dimensions of

lm x lm, one was made from fiber-reinforced plastic, and the other

was made from steel. The F.R.P. tank with a capacity of 52.5 tons

(5m x 3m x 3.5m) was strengthened by bracings. The steel tank with

a capacity of 38 tons (4m x 3m x 3.5m) was investigated in two

cases. In the first case, the steel tank was not strengthened (the

standard test). In the second case, the steel tank was braced.

The large-scale shaking table of National Research Center for

Disaster prevention in Tsukuba was used in these experiments.

Frequency properties of displacements and hydro-dynamic pressures

were measured in these tests.

The mathematical simulations were carried out in the frequency

domains, in which the sloshing would be negligible. Natural frequencies, the sinusoidal responses of displacements, and hydrodynamic pressures were calculated. The same elastic analysis for

cylindrical liquid tanks has already reported

Analysis

In this section, the vibrational equation for the rectangular water

storage tanks supported by base frames is obtained by RayleighRitz's methods. The coordinates and dimensions of rectangular water

storage tanks are shown in Figure 1. The vibrational direction is

only x-axis. The bending deformations of pressure walls and the

shear deformations of side walls are considered, and other deforma-

! tions are neglected in this analysis. So, the absolute displacements of the pressure wall u(y,z) are assumed as follow.

U(y,z) = l l £mn (t) sin(5^) skS + E ck

(t) sinfe)

m=l,3 n=l,2 k=l,3

. z + n(t)^-+ u^t) + u

g

(t)

The first term means the bending deflection of the pressure

wall with the simply supported edges. The second and third terms

represent the shear deflections of the side wall. Because of the

26-1
roof mass, a third term is added to this virtual displacement.

u

Q (t) is the relative displacement of the base frame, and ug(t) is

the ground displacement.

The water would be regarded^ J^o be incompressible, non-viscous

fluid. The velocity potential ' of the water in a rectangular

tank is assumed as follows:

4> = - ( u (t) + ua

(t)) ( x - § ) + Z i|> s

(t) cos(5Bi) oosh(^)

O y /. , _ a a

s=l, 3

+ Z Z 6ij(t) sinh(Tr/(i)2 + (i)2 (x - |) ) oos(^) cos(^

i=0,2 j=l,3

This potential <t> satisfies the three-dimensional Laplace's equation,

and the continuous conditions of velocity on the pressure wall are expressed as follows: = - u(y,z)

x=0 , a

To calculate any particular coefficient 6— (t) of the velocity

potential, both sides of this equation are multiplied by cos (—y)

jlr . b cos(^pjz) and integrated from 0 to b with respect to y and from 0 to

h with respect to z. The orthogonality 6-.(t) may be given by the

following expression:

if, i=0

6

j (t) *" " h-U a

( 1 1 C mn (t)H 5 (j,m,n)

COSh(-^pj-) m = l,3 n=*l,2

+ I i, k {t)H + nH (j))

k=l,3

if, i?"0

&

i .(t) = "h

I I tmn(t) H ct (i, j,m,n)

3 ^,Tra /.2hi. 2 . I2 ,m=l,3 n = l,2 cosh( 4h/ < -b") + ^ >

where

,1 1 v 8 , d 2 d 2 . H (] ,m,n) = ( j- t ) f

" jn3 m((n£)2 - (I) 2

)

, h i ,kiTh, . ,jir, „ .... 4 , k 2d -

2 sln( ^d-} Sln( V> H (],k) = ( r- t ) W <^>2 - ^

H (j, = (sin(^) - Jp)

, h j . . nnh. , jit. , » < ^ 16 , m(n

d

~

2 S1 "'-d~) sin(-L-))

H ;f ( 1 , 3 ,m, n) = —( r- 3 ) " „i/t 2*1

)* 1 j~* (n>

2 - i

J )(n2

( ^)

2 -

The next procedure is to derive the potential and kinematic

energies of this system. The potential (strain) energy of one pressure wall u^ is given by the following expression:

26-2
,,1* +

J

) 2 - 2(l-v) ( 3^ IF " ( 3y~^) }

> dy dz

The potential energy of one side wall U

s and the potential energy

of a base frame are given by the following expressions:

. 3u 2

) dz

y = 0

U _=-JL U

2 f 2 o

The potential energy of the water surface (U ) is given by the

following

P. .9 rh ra w

0 0 w

2

I

1 , dx dy z=h

In these energy expressions, D is the flexual rigidity of the pressure walls, G is the shear rigidity of the side walls, is the

stiffness of the base frame, g is the gravity acceleration, and w

is the up-down displacement of water.

The kinematic energies of one pressure wall (T^), one side

wall (T ), a roof (T ), and a bottom plate (Tj) are given by the

following expressions:

V u(y,

0

J 0

z)^y dz , u (0, z)7dz

T

r

= u(0,h)z

,

m. u(0,0)2

IWhere is the unit area mass of the pressure wall, p g is the

unit height mass of the side wall, m

r is the roof mass, and is

Ithe bottom plate mass. The kinematic energy of the tank water Tw

is given by the following expression:

r a

2

j 0' 0-

.3 <{) 3<t> 3<J>,

( (ktt) + (^V +

(ff)' )dx dy dz

0 3x 3y

Because the tank displacements except x-direction are neglected,

the above expression may yield the following:

-h T =w

Pw

i:

rb

0 dx dy + p z=h 0 <H ^ ti(y,z)dx dy

The Lagragian L, that is, the difference between total potential energies and total kinematic energies, is defined by

L = U f + 2U, + 2U + U,,

r b s w ( Tr + Tf + 2Tb

+ 2TS )

Substituting the Lagragian L into Eular's equation, will give the

vibrational performance of this rectangular water storage tank.

jEular's equation is defined by

3L

3v,

d , 3L

) - 0

'r " v r

«/here v

f is the generized coordinate corresponding to t . 1. u

p

, *

s -0 time t,

4) g would be obtained by integrating \Jj s with respect

26-3
After all, the vibrational equations of this system can be

written in the matrix form as follows:

[M]{v} + [K]{v} =

where

{v} T

= ( £ n. K 12, ? 13, * * • £mn , C l, C 3/

[M]

SYM mrr

[K] =

k k 11 *12

22

SYM

{m

Q

} is the column vector of matrix [M] , corresponding to u

Q

. The

result of this procedure, the dynamic behavior of rectangular

storage tanks, would be represented by the similar vibrational

equations for the general structures. But, the mass matrix has no components of zero, because of the influence of the water. Vector

{v } should be obtained by direct computation for any input Ug.

Regarding the right side of the matrix equation as zero, the natural

frequencies and the eigenvectors can be derived by using the modalanalysis method. Accordingly, the vector {v} may be shown as follows:

{v} - Ui(v,}qi(t) + y2{v 2 }q 2 (t) + + Vpiv }q (t)

The following equation will be attained by using the orthogonalities

of eigenvectors {Vp}

:

qp

(t) + w£q

p

(t) = -U

g

(t)

Adding the damping factor to this equation, one has

qp

(t) + 2h

pup

q

p

(t> + U,p

2 q

p

(t)= -iig(t)

P is the participation factor which is given by the following

expression

:

{vp

}T {mQ } Pp

{vp

} T [M]{v

p

}

Using this method, the deformation and hydrodynamic pressure of

the pressure wall would be obtained like follows.

u(y,z) = E pp qp

(t)( £ E p

£ mn sin (5^) sin (^-)

p=l,2 m=l,3 n=l,2

k 7T Z Z + Z p c k sin (-2^-) + p

n + Pu 0

) + u g

k=l , 3 26-4
P(y,z)=pw§£ - P w ( I y p q (t)( I Y s cosh(^) + uQ| x=0 p=l,2 s=l,3

+ h( E I pC mn Z Z (m, n , i , j ) tanh (A^)2 + f z|> m=l,3 n=l,2 i=2,4 j=l,3

cos(—^)cos(^-) + I (m,n, j ) tanh (^^-) cos (^^-) ) + E 3=1 » 3 k=l ,

3

I H C (j,k) tanh cos (^) +

p

n Z Hn(j)tanh(^)cos(^-)))

j = l , 3 j = l , 3

where

{vP

jlT " fpCll.pC 13 pWpSl • •••p5k' P

n 'pH'i'pV3

p

V s> In the computational analysis, the sloshing effects of free

surface are assumed to be negligible. Thus, vibrational equations

are obtained by neglecting the first terms of Uwand T^.

The sloshing equation in this case becomes, as generally known,

?. f g

- -

, 4a

„

( So ii 9

) a

S TT COSh(-^—

)

The circular frequency u

g is

In the high frequency range, where the circular frequency u is greater than w

?

, the second term in the left hand side in the

sloshing equations may be negligible. So the sloshing equation

would be rewritten in the following form:

= ~

- STTh,

( + V s ir cosh (

) a

Substituting this t|i into the matrix equation, the approximate equation for the dynamic behavior of rectangular water storage tanks

was obtained.

Shaking Table Experiments and Numerical Computations

In the first shaking-table experiment, the rectangular water storage

tank lm x 1m and lm x 0.5m F.R.P. panels with bolts and sealing

materails was investigated. This test tank was fixed to the steel

base frame, and set on the shaking table. To gain the bending

rigidities, the F.R.P. panels have the convex patterns and the

ribs along the edges as shown in Picture 1. Because of hydrostatic

pressures, the panels' thickness increases with depth. Furthermore,

this tank was strengthened by the inside braces as shown in Picture

2. In practical use the tanks should have roofs. In this experiment, the roof was taken off, and the upper panels were replaced

with transparent acrylic plates so the behavior of the free surface

could be seen. The dimensions of this tank were 5m in parallel with

the excitational direction, 3m perpendicular to this direction, and

3.5m in height. The water levels employed in this experiment were

0m, 1.25m and 2.5m. The relative displacements to the shaking table

foundation, the accelerations, and the hydrodynamic pressures actinf

on the walls were measured. The natural frequency for the empty

26-5
tank was recognized at about 2.65 Hz, the natural frequency for the

water level of 1.25m was about 12.5 Hz, and the natural frequency

for 2.5m was about 7.0 Hz. The case with a water level of 2.5m

will be discussed in detail. The deformation shapes at natural

frequency 7.0 Hz, as shown in Figures 2, 3, 4, 5, look like the

shear deformation mode for the side wall, and also look like the

superpositional mode of the side wall shear deformation and the

plate bending deformation for the pressure wall. The peak of hydrodynamic pressure was recognized at 7 Hz as well as the deformations.

Peak value of hydrodynamic pressure was about 1.5g/cra^ for 1 gal

input amplitude, and the profiles of the hydrodynamic pressure

distributions are shown in Figures 6 and 7.

In the numerical analysis for this tank, one bending mode

(sin^Z sin^2

-) of the pressure wall, two shear modes (z/d, sin^2—

)

b d 2d

of the side wall, and the deformation on the base frame, the

4 virtual displacements are employed. The constants of tank are taken as Table 1» Assuming the damping ratios as 5 percent and

4 percent the theoretical results would fairly agree to the experimental data.

In the second shaking-table experiment, the rectangular water

storage tank assembled with lm x lm steel panels with bolts and

sealing materials was investigated. This tank was only placed on

the steel base frame, and held with frail devices. The dimensions

of this tank were 4m in parallel with the excitational direction,

3m in perpendicular with the excitational direction, and 4m in

height. For this test tank, a roof was installed. To examine the

effects of the inside braces, both braced and unbraced tanks were

obs erved

.

In the standard test, the natural frequencies of the tank were recognized at about 9.5 Hz for the empty stage, 3.5 Hz for the water

level of 2m, and 2.0 Hz for the water level of 3.7m. In the test of

water level 3.7m, 5.5 Hz and 10 Hz were observed. Deformations were

found in the pressure wall, but scarcely in the side wall. The

deformations of the pressure wall were similar to the bending modes

of the plate. The amplification ratios decrease with the increase

of water level as shown in Figure 10. The amplif icational ratios

for higher natural modes wouldn't be always larger than for the

first mode, as shown in Figure 11. Using the natural frequency

9.5 Hz for the empty tank and the average unit area weight 44kg of

a panel, the flexual rigidity for the pressure wall would be evaluated about 5500~6000kg*m.

In the numerical analysis for this standard case, 6 bending

modes of the pressure wall, 3 shear modes of the side wall, and the

deformation of the base frame were used. The constants are shown

in Table 1. In the numerical computations, the dampings are considered by 2 percent for the first mode, 10 percent for the second

mode, and 10 percent or 5 percent for the third mode. The numerical

and experimental results are shown in Figures 10, 11, 12, 13, and 14.

In the strengthened case, the natural frequency was recognized

at about 6.5 Hz for the water level 3.7m. The base frame gave the

large deformations in comparison with other experiments. The bending deformations of the pressure wall were measured smaller than the case of the F.R.P. tank. The hydrodynamic pressures at the natural

frequency were about 2g/cm^ for 1 gal input amplitude, as illustrated in Figure 16. Increasing the input amplitudes at the

natural frequency, a slip occurred between the tank and the base

26-6
frame. The same numerical analysis for the F.R.P. tank was conducted. The numerical and experimental results are shown in

Figures 15, 16, 17, and 18. The constants are given in Table 1.

It is reasonable that the flexual rigidity of the pressure

wall strengthened by bracings is larger than the standard case.

The reason that the shear rigidity of the side walls is estimated

to be smaller than the standard case, are considered to be given by

the bending deformations of the side wall which are not assumed in

the theoretical analyzed.

i Earthquake Damage to Water Storage Tanks

In the Near Izu-Oshima earthquake on January 14, 1978, and in the

Off Miyagi Prefecture earthquake on February 20, 1977, many water

storage tanks were damaged. Distortions of base frames, tank slips,

and breaks at pipe connections occurred at many buildings. The wall

of one F.R.P. tank had broken out because of the increase of hydrodynamic pressure. A water tank that was set on the penthouse of a seven-story hotel survived because it was held by eight bolts rather

then the two that are ordinarily seen; one bolt was completely cut

away by the earthquake force. In Izu Inatori, we estimate that half

the water tanks set on the roofs would probably suffer some damage

in some future earthquake.

Conclusion

In this study, a method for calculating the natural frequencies of

rectangular water tanks was developed and compared with the experimental data. The pressures and deformations of rectangular water

storage tanks, obtained by the shaking table experiments and the

1 numerical analysis, indicated that conventional estimates for these

tanks would be small.

Acknowledgments

The authors are grateful to Mr. Harada and Mr. Wakana, who are the

members of Bridge-Stone Tire Co., Ltd., and Mr. Hirata and

Mr. Takeuchi, who are the members of Mitsubishi Plastics Industries

Ltd., for their cooperation in the experiments. Special thanks are

due to the staff of the maintenance section of National Research

Center for Disaster Prevention.

References

1. Housner, W. G. , (1961) Shock and Vibration Handbook , Vol. 3, 50 26-50, 29, Mac Graw-Hill.

2. Kito, F., (1959-1964) On Vibration of a Rectangular Tank Filled

With Water, I-VIII, Journal of the Society of Naval Architects

of Japan , (in Japanese).

3. Simizu, N. , Yamamoto, S., (1974) The Influences That the Long

Period Components of Earthquake Waves Give On the Long Period

Structures, Kenchiku-Gijutsu, No. 274, 381-388, (in Japanese).

4. Hori, N. , Tani, S., Tanaka, Y. , (1977) Dynamic Analysis of

Cylindrical Shells Containing Liquid, 6th. WCEE.

26-7

INPUT DATA FOR COMPUTATIONS

ITEMS SYMBOL F.R.P TANK

WITH BRACES

STEEL TANK STEEL TANK

WITH BRACES

DIMENSIONS BREADTH b 3m 3m 3m

LENGTH 5m 4m 4m

HEIGHT d 3.5u 4m 4m

AVERAGE UNIT AREA WEIGTH

p 16kg/m* 44kg/m?, 44kg/m?

FLEXURAL RIGIDITY OF

THE PRESSURE WALL 65000kg.

m

6000kg.

m

1 50000kg .mSHEAR RIGIDITY

OF THE SIDE WALL

G 3xl06 kg 9.72xl0 7 k9

... _ 6.48xl06

kg

STIFFNESS OF

THE BASE FRAME

k f I0'° kg/m 108 kg/m 108 kg/m

SI|)E WAI I

is. m i >m

BASE FRAW

WATER LFVFL

PRESSURF WALL

EXCITATION DIRECTION

TABLE 1: INPUT DATA FOR COMPUTATIONS FIGURE 1: ANALTYICAL MODEL FOR

A RECTANGULAR WATER TANK

KEY PLAN

AMPLITUDE RATIO

FIGURE 2: MEASURED DEFORMATION PROFILES FIGURE 3: THEORETICAL DEFORMATION

OF PRESSURE WALL (F.R.P. TANK PROFILES OF THE PRESSURE WALL

REINFORCED BY BRACES) (F.R.P. TANK REINFORCED BY BRACES)

26-9
AMPLITUDE RATIO

FIGURE 4: MEASURED DEFORMATION PROFILES

OF SIDE WALL (F.R.P. TANK

REINFORCED BY BRACES)

FIGURE 5: THEORETICAL DEFORMATION PROFILES

OF THE SIDE WALL (F.R.P. TANK

REINFORCED BY BRACES)

2 5m

2 0m

S*r*or •

t ' \

• sw

!

I \\

If

N

1

o 10Hz

i

\ 2 HY0ROOVNAMIC PRESSURE IN kg/cmVlG

FIGURE 6: MEASURED HYDRODYNAMIC PRESSURE FIGURE 7: HYRODYNAMIC PRESSURE PROFILES

PROFILES CAUSED BY SINU SOIDAL (F.R.P. TANK REINFORCED BY BRACES)

EXCITATION (F.R.P. TANK REINFORCED

BY BRACES)

FIGURE 8 1: FIGURE 8-2: FIGURE 9: THEORETICAL HYDRODYNAMIC

PRESSURE AT THE MIDHEIGHT

THEORETICAL FIRST MODES WITH WATER LEVEL CHANGES OF WATER LEVEL (F.R.P. TANK)

(F.R.P. TANK REINFORCED BY BRACES)

26-10
20

i.

brace

5 10 FREQUENCY IN Hz

I \ WATER LEVEL = 37ir

10 0 10 20 AMPLI TUOE RAT 10 (TEST TANK /TAB^E

)

FIGURE 10: MEASURED RESONANCE CURVES FOR

THE FIRST NATURAL MODE (STEEL TANK)

FIGURE 11: MEASURED MODES OF PRESSURE WALL

(STEEL TANK)

FIGURE 12: THEORETICAL FIRST NATURAL

FREQUENCY (STEEL TANK)

FIGURE 13: THEORETICAL RESONANCE CURVE OF

PRESSURE WALL DEFORMATION AT THE MIDHEIGHT (STEEL TANK)

26

'«fOU»«C ' la <u

FIGURE 14: THEORETICAL RESONANCE CURVE OF

HYDRODYNAMIC PRESSURE AT THE MIDHEIGHT

OF WATER LEVEL (STEEL TANK)

26-11
5

-

1 2 HYWOOYNAMIC PRESSURE IN kg/cm2

/ 1

G

rj?, MTtl LlVTl UUO in COPVUTIV J to

"».

i

a \ WU

• » S«j

f ou

» f IMf

FIGURE 17: MEASURED HYDRODYNAMIC PRESSURE

PROFILES CAUSED BY SINUISOIDAL

EXCITATION (STEEL TANK REINFORCED

BY BRACES)

FIGURE 18: THEORETICAL HYDRODYNAMIC

PRESSURE PROFILES (STEEL

REINFORCED BY BRACES)

26-12
27

INELASTIC BEHAVIOR OF NON-BEARING WALLS IN 11-STORY STEEL-REINFORCED CONCRETE FRAME

AND

ASEISMATIC SAFETY OF EXTERNAL SURFACE

FINISHES AND COATINGS

Makoto Watabe

Toshiyuki Kutoba

Akio Baba

Toshibumi Fukuta

Hiroshi Ito

Building Research Institute

Ministry of Construction

Inelastic Behavior of Non-Bearing Walls in an 11-Story Steel-Reinforced Concrete Frame

Shear walls and shear walls with openings have been designed for

many high-rise steel-reinforced concrete buildings. Many studies

have been carried out on the inelastic behavior of such bearing

walls and their success has been accepted as a part of the seismic

design procedure. In recent years, there have been many requests

to enlarge the floor area per unit of public high-rise apartment

buildings. To answer these requests, the structure shown in

Figure 1 was proposed. To estimate the seismic safety of this

structure, which consists of shear walls, shear walls with openings, and non-bearing walls, lateral loading tests were performed

in 19 75 1 . To clarify the effect of the non-bearing wall on the

structural behavior as a whole, a lateral loading test on the halfscale test specimen simulating an eleven-story apartment building

was carried out in 1976. In this paper, the inelastic behavior

of a test specimen is summarized.

Lateral Loading Test

A high-rise steel-reinforced concrete apartment building has been

designed according to the design provisions in the Architectural

Institute of Japan Standards. The distribution of the seismic

coefficient along the height is inverse triangular and the seismic

coefficient at the eleventh floor is 0.401. The non-bearing wall

is designed under the Provision of the minimum reinforcement ratio

of shear walls. The test specimen shown in Figure 1 is the halfscale space frame consisting of one non-bearing wall and two shear

walls with an opening surrounded by steel-reinforced concrete frames

and is four stories high. To apply lateral and axial forces, walls

and beams are assembled on the top of the test frame. The lateral

force is transferred to the non-bearing wall through the floor slabs

and the beams in the ridge direction. Axial forces shown in

Figure 2 are applied to the four columns around the shear walls, so that the stress distribution of the test frame may be simulated to those of the actual eleven-story apartment building in the elastic

range. Figure 2 also shows the relation of axial force N^, N2 , and

lateral force Q. Figure 3 shows the loading program. Materials

used are normal concrete, SS41 steel plate, SD30 and SR24 reinforcements, and mesh. Properties of these materials are described in

Tables 1 and 2. The test frame details are described in Figures 4 through 7

.

Test Results

I Figure 8 shows the lateral force-displacement relation, measured at the top of the test frame, 7.055 m high from the first-floor face.

Table 3 is the list of the forces when flexural cracks, flexural

shear cracks, and shear cracks are observed at each elements and

; the list of the forces at reinforcements or steel frame yielding.

I Figures 9 through 14 describe the crack distribution at three

j| stages, R = 1/200, 1/100, and 1/50 where R is the rotation angle

" measured over total height.

The lateral load-displacement curve is nearly straight prior

to the load, which is a little higher than the design load (48

tons). The test-frame load at the maximum displacement of the

sixth loading cycle is about 85 percent of that of the second loading cycle. After this, the test frame load increases only 9.2

L percent of the load at the maximum displacement of the second loading cycle prior to failure. This shows that the test frame yielding

occurs at the load of 224 tons. At 265 tons in the seventh loading

27-1
cycle, crushing occurs in compression near the foot of the first

floor of the non-bearing wall. At the load 274 tons in the same

cycle, rupture of some vertical reinforcements occurs at the first

floor of the non-bearing wall. Before this loading, the breadth

of the flexure shear crack spreads to about 14 mm at the rupture

point. Rupture of some vertical reinforcements occurs in the

tensile regions of the first floor of shear walls with openings

at the stage of the maximum displacement in the seventh cycle. The

maximum test frame load is 274 tons, 5.7 times higher than the

design load (48 tons). The rotation angle measured over total

height is 1/114. The test frame load at the maximum displacement

of the ninth loading cycle is about 85 percent of that of the

seventh cycle. Restoring force of the test frame subjected to lateral load reversals is stable. After this loading stage, the test frame load decreases gradually and failure of the test frame occurs at the stage when the rotation angle measured over total

height is 1/51. Damage to the test frame is concentrated on the

first-floor elements; the rotation angle measured at the first floor

is 1/16. The failed test frame showed crushed columns and shear

failure of the shear walls with opening. Reinforced by full-web

steel plates and mesh, spandrels of the shear wall with opening

were damaged only a little.

Analysi s

The following are the methods of the analyses for the elastic stiffness and the cracked loads of the test frame. The shear with opening is idealized to the model with rigid zones shown in Figure 14 and is analyzed by the slope-deflection method. The deflections

caused by bending moment and shear force are considered in this

method. The non-bearing wall is idealized to a cantilever whose

fixed end is the first-floor face of the wall and is analyzed by

the method considered alone with deflections by bending moment and

shear force. The stiffness effects of floor slabs and beams in

the ridge direction and the cracked loads are neglected here.

The elastic stiffness of the test frame is given as the sum of

the elastic stiffnesses of each wall.

In case of the flexure cracked load, the following equation is

applied to the section of the each wall's foot at the first floor and

the flexure cracked moments of the section given are substituted in

the lateral load-bending moment relation: _ z (ref. 2)

cMb = 1.8VFc Ze + — N Ae

Fc = compressive strength of concrete

Ze = section modulus (includes reinforcement)

Ae = sectional area (includes reinforcement)

N = Axial load

In case of the shear cracked load, the following equation is applied to the section of the each wall at the first floor:

cQ = r*fL*(7.5 + 0.015 Vfc) (ref * 3)

r = 1 - (equivalent opening peripheral ratio)

t = thickness of wall

L = distance bewteen columns

The flexure strength is calculated by applying the following

equation to the walls:

27-2
uMb = A • a «L + 0.5z(A « a )»L + 0.5N-L (ref - 4)

c y w wy

At = total section areas of axial reinforcement of

tention side column

a = yield strength of above reinforcement

A„ = total sectional areas of reinforcement of wall w

= yield strength of above reinforcement

The shear wall with opening is assumed to be a cantilever as a whole. The following equation is the relation between the lateral

load and the ultimate flexure moment of the wall:

Q = (uMb - N«L/2)/H

The non-bearing wall is assumed to be a cantilever whose fixed end

is the first floor face 0"" this wall. The flexure strength of the

test frame is given as the sum of the flexure strength of each

wall.

The shear strength of each wall is calculated by the following

equations and the shear strength of the test frame is given in the

same way as the flexure strength of the test frame:

Q = min.CuQSj, uQs 2

) (ref. 4)

uQs

1 = (0.068Pte (Fc + 180)/ V M/Q* D + 0.12 + 2.7 Vpwe CT wy

) be *j + sQ

UQS2 = max. (r»t »L. (7.5 + 0.015 VfC) r(Qw + iQc))

b

g = equivalent depth of wall

= (total section areas of wall and column)/L

L = end-to-end distance between columns

P

te = 100 x At /b

?

»L A

t = total sectional areas of axial reinforcemnt of

tension side column

Pwe

= equivalent lateral reinforcement ratio of wall

aw

= yield strength of above reinforcement

s$ = shear strength of steel frame within wall

Results given from the foregoing analytical methods are listed

in Table 4; loading test results are shown too. These analytical

methods make a good estimate of cracked loads and ultimate strengths

of the test specimen.

Inelastic Behavior of Non-Bearing Wall

Load-Displacement Curve

The load-displacement curve of the non-bearing wall is calculated

from the test results of the Reference 1. The properties of the

materials used for this test are almost the same as those of the

materials used for the Reference 1. But the compressive strength

of concrete used for this test is about 18 percent higher than that

of concrete used in Reference 1. After flexure yielding, both test specimens collapsed at shear failure. These points lead to the

assumption that the behavior of the shear wall with opening in

Reference 1 is the same as the shear walls with opening here. If

the effect of floor slabs and beams in the ridge direction on the

restoring force are neglected, then the restoring force on the nonbearing wall is given as the difference between the restoring force

of this space frame and two times of the restoring force of the

shear wall in the Reference 1.

27-3
This eleven-story apartment building has been modeled as a one-mass system with a single degree of freedom. Responses to several earthquakes are calculated to clarify the effect of the non-bearing wall

on the structural behavior as a whole.

The restoring force of this one-mass system is obtained through

the following process. The height of the loading point of both tests

in the reference 1 and this report is almost identical to the height

of the center of gravity of the half scale eleven-story apartment

building considered. Therefore load-displacement relations of walls

shown in Figure 16 are adopted as restoring forces of walls constituting this one-mass system. The restoring force of this system is

assumed to be the sum of those of walls in this system. The

restoring force of the non-bearing wall is taken into account in case A and put out of account in case B. Shown in Figure 17, the

whole restoring force of the system is approximated to several segments. Dimensions of the system adopted to response analysis are listed in Table 5. The elastic stiffness and the maximum strength

of the system in case A are 1.18 times and 1.11 times larger than

those of the system in case B, respectively. The displacement at the ultimate stage of the system in case A is nearly equal to that in case B. The undamped natural frequency of the system in case A is a little shorter than that in case B.

Earthquakes for response analysis are listed in Table 6. Responses described there are part of the results of the analysis.

The maximum displacements subjected to earthquake loadings in case A

are about 60 percent or 90 percent of those in case B. The maximum

velocities subject to earthquake loadings in both cases are nearly

equal. The maximum acceleration subjected to earthquake loadings

in case A are 30 percent of those in case B.

In this analysis, the effects of damping and basement rocking

are not considered.

Cone lusion

Loading test results clearly show the effect of the non-bearing wall

on the flexure-cracked load, the shear-cracked load, and the maximum

strength of the test specimen. These loads and maximum strength

increase, compared with those of only shear walls with openings.

The rotation angle measured over total height of the non- bearing wall is considerably small (R = 1/284) at the stage of its

maximum strength, but at the ultimate stage the rotation angle

measured over total height is nearly equal to that of the shear

wall in-filled in Reference 1. It appears that non-bearing walls

may indeed absorb the input energy under earthquake loadings.

Despite the above results however, it is recommended that non- bearing walls be neglected in the seismic design of the structures.

References

1. Endo, T. and Goto, T. "Experimental Study of Steel-Reinforced

Concrete Shear Wall with Opening," (November 1976).

2. Hirosawa, M. "Past Tests Results of Reinforced Concrete Shear

Wall and Analyses," (March 1975).

27-4
The Architectural Institute of Japan, "The Design Provisions

the Reinforced Concrete Structure and Their Explanations."

The Architectural Institute of Japan, "Earthquake Load and

Seismic Safety of the Structure - 1976," (January 1977).

27-5
i i i! i

1

1

1

1

1

1

1

1 II

1

1 hp

I

1

i

1 !

I jlji

i III!

jl

II

78 1 78

Yk I j 78m

FIGURE 1: PLAN OF HIGH

RISE FLATS

A ; TEST FRAME

B : SHEAR WALL

C . SHEAR WALL WITH OPENING D NONBEAftING WALL

694

Q<ton> FIGURE 2: AXIAL LOADING - LATERAL LOADING RELATION

column flange

web

main j D!0<S030>

reinforc-tIOl3t$D30> hoop (SR24) beam L shape

web

lattice iSR24)

wall mesh

3070

Ev'P 1

1490

! !

205 \ 4407

3407 1 1620 ,212 ; 4549

t 1 3503

f \ 2116 1 90 5103

4- 3963 2146 190 5680

4073 2221 188

I 5810 3283 1580 1 210 4356

3330 1570 209 4630

3858 2089 180 4718 4925 2661 203 5173

"T

1/100 .

[

cycles

FIGURE 3: LOADING PROGRAM

r

base 2666 ! 2190

poissomi

ratin ' EiiK/h^tnT) ratio

2.15

lfl-4fl 263 1 j 2440 I 98 021

TABLE 1: PROPERTIES OF MATERIALS: STEEL TABLE 2: PROPERTIES OF MATERIALS: CONCRETE

* -LT

soo . no .

950 . 1150

thmat wall R/C column, beam . StMl-R/C

200. FIGURE 5: REINFORCEMENT: SHEARWALL WITH OPENING

FIGURE 4: SHEAR WALL WITH OPENING

27-6
27-7
y t -

1 r * !

. -v 1 i *-i ; 1 1^1 tI/I" lift

FIGURE 12: R = 1/200 FIGURE 13: R = 1/100 FIGURE 14: R = 1/50

27-8
shear wall with opening

column beam

T

nonbearing wall wall

J-+ T

" cycl* tbrft

M. — _ - n A I O C A flexure crack t o4 I 40 9

4" J

9 *o 9£

T

i

• 9 L

1

I 9

t Anou

flexure shear crack

1 2 96 2 112 2 96 2 80 2 176 ! 2.

f 144

shear crack 2 80 2 112 2 96 2 80 2 96

t 2 80

yield point 1 2 208 2 208 2 144 2 176 2 208 2 112

yield point 2 2 176 2 176 2 240 7 235

yield point I yield point 2

reinforcing bars yield H or L shaped steel members yield

—

I

TABLE 3: TEST RESULTS

stiffness m! flexural shear

elastic range cracked loadjcracked load ultimate flexural strength ultmate shear strengnthj

experiment 224 4d cm) 64 0 t 96 0 t 274 0 t 274 0 t

' case 1 J case 2 I _case 1

T case~2 ' [theoretical 391 3 (. cm) 58 2 t | 77 5 t 248 1 t 228 1 t] 184.8 t' 164.4 T

case 1 this includes nonbearing wall strength case 2 this doesr t mcrude nonbearing wall strengh

TABLE 4: MAXIMUM STRENGTH OF TEST FLAME

Ke(t/cm)J^Ki it/cmi K2 it/cmi! K3 lt/cmi , K4 (t/cm)

| K5 't/cmi case A 2055

1 case B 2055

15790

13330

6137 3538

case A 4737

case B 3999

Kg K7(t/cmiX] Icml X2 icmi

041 244

042 3 36

392

684

-23033 -776

1 -50950 -776

X3 (cm) 1 X4 (cm) Xt,<cm> 690 730 12 8

716 730 128

15790

1333.0

h00

00

case A 6 shear walls in filled + 10 shear walls with openings 5 NONBEARING WALL

case B 6 shear walls in filled+ 10 shear walls with opening

TABLE 5: DIMENSIONS OF MODEL FOR RESPONSE ANALYSIS

27-9
Ae (cm''! le (cm*

I FL 3388 I 390

column I t H

j 2~4FL 3 324 I 369 I

b « a i 596 I 497

FIGURE 15: MOMENT DISTRIBUTION OF

SHEAR WALL WITH OPENING

Shear wall with opening

120 Ximmi

FIGURE 16: EXPERIMENTAL ENVELOPE CURVES OF WALLS

x, yield

x< collapse

k. stiffness

in elastic range

Restoring Force C.N

envelope curve

Xj X4 x$ Deflection

- ,x cx example

C

*T" C c* ,xc'* ab-c-d-e-a-f-g-hh

K-LH-M-C-N

FIGURE 17: MODEL OF RESTORING FORCE CHARACTERISTICS

El Cent ro

A

x (cm) I 20

x (kin)J 23 8 ^5,?^*vH 55,2

76

I5000 1 (gal);

X (cm)

x (kin)

X-t-y (gal)

320

719 9

B

I 33

210

428 2

3 05 42 0

Ta f t

A

1 30

23 6 580 9

2 I 0 352

8

I 92

24 5 532 8

3 32

34 5

771 2

Vernon

A

I 18

21 4 545 3

I 68

267

694 5

8

I 72 250

495 9

2 90 38 8 699 9

Hiroo

A

1 86

32 3 747 4

2 52 46 3 725 9 922 9

I 820 8

TABLE 6: MAXIMUM RESPONSE TO SEVERAL EARTHQUAKES

_B

1 56

29 8

-i ----

I 469 0

2 21 42 3

[

581.

1

Hachinohe

A

1.37

19 6 6028

2 49 322

922 2

B

2 33

t 254

600 2

4__ 4 40

T 427

831 3

- —I

27-10
27

INELASTIC BEHAVIOR OF NON-BEARING WALLS IN AN 11-STORY STEEL-REINFORCED CONCRETE FRAME

AND

ASEISMATIC SAFETY OF EXTERNAL SURFACE FINISHES

AND COATINGS

Makoto Watabe

Toshiyuki Kutoba

Akio Baba

Toshibumi Fukuta

Hiroshi Ito

Building Research Institute

Ministry of Construction
i
Aseismatic Safety of External

Surface Finishes and Coatings

Though few structural members are destroyed, many non-structural

elements and components are often destroyed in earthquakes. Worse,

these sometimes threaten human life. In view of the fact that relatively little previous work has been conducted on the safety aspects

of non-structural elements and components, this study evaluates the

seismic safety of non-structural elements such as wall tiling,

terrazzo block, brick masonry, and cement mortar rendering, etc.

Experiment

Specimens were the- external surface finishing (Table 1, Figure 1) applied on the face of 20 load bearing walls with openings. The

clearance of the boundary of each non-structural elements was 10 mm.

The loading method and program were explained in the previous paper.

The measuring method for each non-structural element was as follows:

cracking: visual observation

de laminat ion : visual observation by tapping the surface

of specimens with a steel bar

shear movement of joints: contact gage

adhesive strength: adhesion testing apparatus

Evaluation Method

First, each surface is divided into a cross-sectional square grid.

Then the rate of cracking and rate of delamination can be defined as follows

Rate of cracking _ number of grids having any crack

(ROC) total number of grids

Rate of delamination _ number of grids having any delamination

(ROD) total number of grids

Evaluation of Density of Cracks by Rate of Cracking (ROC)

? When the size of grids is a cm x a cm, density of cracks D cm/cm

can be expressed by rate of cracking as follows:

D = (0 £ 0 <1>*/2)

VT a sin (6 + 'FT/ 4)

6 = angle between horizontal line and cracking line.

This equation is based on the following assumptions:

A cracking line is straight with a certain gradient.

The distance of adjacent cracking lines is greater than the

size of grid.

Cracking is expected to be caused at any part of wall.

As for the relation between density of cracks and ROC, the comparison between the theoretical value and the experimental value is shown in Figure 2. As shown there, ROC is proportional to density

of cracks or the length of cracking. Therefore the concept of ROC

is used instead of density of cracks or the length of cracking.

27-11
Evaluation of Delaminated Area by Rate of Delaminat ion (ROD)

Since it is hard to identify the boundary of delaminat ion, the

following method (Grid Method) is proposed here. In the case of

tiling wall the grid is the same as the dimension of the tile;

with cement mortar and sprayed finishing, it is 10 cm x 10 cm.

In this method, ROD is varied by the size of grid. Therefore in

order to exclude the size effect, initial ROD is converted to normalized ROD as follows. Supposing initial ROD is P, ROD on

the enlarged size of grid by n-times is P . If the delamination

2 area is small and distributed randomly, P = 1 - (1 - P) n

(n: natural number). If each delamination area is larger than

the enlarged grid, Pn

= P. As the actual delamination consists

of those two phases, the relationship between P and Pn can be

provided as follows:

Pn

= {1 - (1 - P) n2

} a + P(l - a)

a = weight coefficient

which can be derived from the experimental data

(Figure 3)

In case of n = rational number, PQ can also be provided by

this equation. Normalized ROD can be based to compare certain

delamination areas with other different-size grid areas.

Results

Some results of the cracking and the delamination in non-structural

elements are shown in Figures 4, 5, and 6. In these figures, a filled-in square shows an element having any cracking or any

delamination up to the horizontal displacement of structure reached

at R = 1/200.

The cracking and the delamination of non-structural elements

occur along with the cracking in the load-bearing walls, while the

delamination in the stoneware tile on the first story appears to be

caused by relatively large compressive stress.

Generally more cracking in load bearing walls and more cracking and delamination in non-structural elements can be recognized

on lower stories than on uppers.

At first cycle (design load), the cracking and the delamination

in non-structural elements was scarcely recorded. At second cycle

(horizontal displacement R = 1/200), a great amount of cracking and

delamination occurred, while no falling out was observed. At seventh cycle (R = 1/100), a few fell out of the first story.

Characteristics of Cracking and Delamination in Various Surface

Finishings

The relationship among ROC in load bearing walls and ROC or ROD in

non-structural elements is shown in Figures 7 and 8. Because the

behavior and the degree of destruction is different in each loadbearing wall, the characteristics of various non-structural elements

are expressed by applying ROC in load bearing wall as the parameter.

There is a limit to the length of crack propagation in the plane of

load bearing wall even with a stepwise increase in loading. Within

this limit, ROC in non-structural elements is relatively proportional to ROC in the load-bearing wall, and its proportional coefficient depends upon the individual non-structural element. Elements

27-12
having large proportional coefficient, such as cement mortar and

and epoxy-based sprayed finishing, are liable to be cracked. The

cracking on porcelain mosaic tile wall was only within the joints

in a stepwise pattern.

The relationshup between ROC in load bearing walls and ROD in

nonstructural elements is shown in Figures 8 and 9. This result

represents the following:

Specimen -4, -5 (porcelain mosaic tile)

ROD in non-structural elements was relatively proportional

to ROC in a load-bearing wall.

Specimen -1, -2 (stoneware tile), Specimen -6 (cement mortar

rendering)

When the length of cracking in load bearing wall reached its

limit, phenomenon of delamination in non-structural elements

was started.

Specimen -3 (stoneware tile), Specimen -7, -8 (sprayed

finishing )

No delamination occurred.

The pattern of the delamination on Specimen -1 (IF) appears to

be a combination of the first two shown above. In brick masonry,

there appears only a slight cracking, while in terrazzo block slight

cracking and the falling out of several blocks were observed.

Results of Adhesion Test and Measuring of Joint Movement

Figure 10 shows result of adhesion test. At R - 1/200, the shearing

movement of joints in Specimen -9 (brick masonry) was very little,

and in Specimen -1 (stoneware tile fixed by ordinary cement mortar

method) and Specimen -10 (terrazzo block) it was about 0.25 and 0.75

mm, respectively.

Conclusion

On the basis of this experiment, the following can be pointed out:

Our "Grid Method" for evaluating the cracking and delamination of external surface finishing, is simple and

rational

.

More work needs to be done to clarify the fundamental

mechanism of the delamination—rigidity, strength, thickness, strain, adhesive strength.

"Specimen -3" (stoneware tile embedded at concrete) and

"Specimen -7, -8" (sprayed finishing) were concluded to have sufficient aseismatic performance; the rest were relatively safe provided that no structural members were destroyed.

External forces, such as the relative displacement between

the stories and the strain at the time of the collapse of

non-structural elements, also need more study.

27-13
mn tor inl ^^/TrV. H 1 rTT II 1 Ulll L/Cr 1 ufjj-fl iwj aiui y

1 1 3

2 1 3 60x1 08x20 mm 3 2 5 nnrrolnin mno/iir ttlo \j \j i tr'Lii 1 1 ii iuau \ \r ill 2 4 LI * LI1 x L mm

rpmpnt mnrtnr 6 3 5 ©po xy based sprayed finishing 7 2 3 urethan based sprayed finishing 2 5

brick masonry 50x50x2t0mm 9 1 4

terrazzo block 500x580x28mm 10 1 L

TABLE 1: SPECIMENS

cement mortar cement mortar including methyl- epoxy based

Speamen-1 Specimen- 2 Specimen- 3 Specimen- u Specimen5

27-14
Speamen-3 | I 5pecimen-5

| Specimen-3 I I Specimen-5 n n— 1 —3—

:^ H • 1— • —= si 1 j J^Specimen-i ^Specimen-"5j|J! Tl " Speclmen~-Z TgecTmen-5J "

Lll. Specimen -1 ^"Specimen -2 JlP ~7ii Specimen -1 l_ Specimen -T}J

FIGURE 4: DISTRIBUTION OF

CRACKING IN STRUCTURES

FIGURE 6: DISTRIBUTION OF

DELAMI NATIONS IN OBJECTED

FINISHINGS

FIGURE 5: DISTRIBUTION OF

CRACKING IN OBJECTED

FINISHINGS

27

27-15
Sr

ROC in Structure (%)

FIGURE 7: RELATION BETWEEN ROC IN STRUCTURE AND ROC IN FINISHING

o

0 20 40 60 80

ROC in Structure 1%)

FIGURE 8: RELATION BETWEEN ROC IN STRUCTURE AND ROD IN FINISHING

Specimen-

£

Specimen- 5 /

Specimen -

1

/

Specimen - ? Specimen- 6

H

Specimen- 3 Specimen- 7

! Specimen- 8

Critirpl ROC

ROC in Structure

cn o

.

CJ) c

f o

m

>

i/i o<

FIGURE 9: DELAMINATION PATTERN FIGURE 10: RESULTS OF ADHESION TEST

27-16
28

SEISMIC EVALUATION OF EXISTING MULTISTORY

RESIDENTIAL BUILDINGS

G. Robert Fuller

Department of Housing and Urban Development

Washington, D.C.

Seismic Evaluation of Existing

Multistory Residential Buildings

There has been a recent demand, caused by increased construction

costs, for rehabilitation and restoration of existing multistory

buildings, originally constructed for other occupancies such as commercial offices or hotels. A complex problem then arose for HUD's technical personnel and consultants to evaluate the structural resistance of existing buildings to earthquake loads.

The response of an existing building to earthquakes reflects

the performance level inherent in codes, standards, and construction practices existent at the time of design and construction of

the building. Building practices generally improve during the life

of a building, reflecting an improved state of knowledge. Thus,

implied margins of safety also change, depending on whether a comparison is made with the code in force at the time of design or with the current code. In addition, deterioration and alterations

during the service life of a building affect the actual margin of

safety provided by the structure. The need exists, therefore, to evaluate the potential seismic hazard of buildings. Following this

evaluation, the cost of appropriate strengthening or retrofit may

be determined so that the feasibility of various schemes to mitigate

unacceptable hazards may be established.

The evaluation procedure discussed here is based on a feasibility and cost study of alternative strengthening techniques. It

is intended to provide a method for upgrading existing buildings by

identifying elements needing strengthening or repair and by verifying the adequacy of rehabilitation. The methodology can be applied

nat ional ly

.

Methodology

The Methodology is based on the "Approximate Analytical Evaluation

Method" developed by S. B. Barnes and Associates for the National

Bureau of Standards. This analysis technique is contained in

Building Science Series 61 (BSS 61), titled "Natural Hazards Evaluation of Existing Buildings" (National Bureau of Standards, 1975).

The standard for determining structural integrity is based on compliance with the 1973 Edition of the Uniform Building Code

((International Conference of Building Officials, 1973).

Data collection forms are included to provide a guide for compiling the type of information necessary to evaluate the building.

Guidance is also given on methods available to verify properties of

existing structural elements. If structural drawings of the

original building are available they are very helpful, but it may

jstill be necessary to verify how well the drawings define the actual

building. Structural calculations for the original structure can

also be of help, particularly in understanding the basis of the

original design.

At the completion of the data collection, a first decision

point is reached. Information in the data collection must be

[[sufficiently complete and detailed to decide whether or not further

analysis is desirable. The data collection information may result

in termination of retrofit studies if the magnitude of renovation

is obviously excessive.

fl . To analyze a building as a whole it is necessary to assume an

[analytical model of the framing system. This is true whether a static or dynamic analysis is used and whether analysis is

28-1
accomplished by hand or computer methods. The model is determined

by the building layout, physical properties of materials in structural frame or walls, and by their size, shape and location.

Once the model is established, the elastic fundamental period

of the building can estimated. This may be done by hand calculation using the simplified formulas contained in the 1973 Uniform

Building Code (UBC) based on building layout only, or by computer

analysis to any degree of analytical complexity desired. Once this

period is determined, the minimum design base shear can be obtained

from formulas in UBC. The distribution of forces on the building

may be accomplished either by simplified methods in the Code or by modal analysis of the lateral force resisting system.

The distribution of base shear as given by the Code should be

modified if the building is highly irregular. This modification

may be accomplished by approximation or by use of modal analysis,

which is solely dependent upon the user's preference.

As part of the development of the structural model, it is

necessary to determine paths through which loads are transmitted.

This identifies the methods provided by the building frame for

resisting lateral forces. In most buildings within the scope of

this methodology, the resisting system will be structural bearing

or filler walls. Some buildings may have lateral force resisting

systems composed of moment frames or braced frames.

Once the forces on the building have been determined, an analysis of the structural model is performed in which the imposed

loads are translated into moments, shears, and axial loads on the

individual members composing the structural model. This analysis

may be performed by using hand methods or the computer program

included as part of the Manual.

From the basic data used to construct the structural model,

the strengths of the individual members are determined by using

UBC '73 criteria. These are the same methods provided by material

specifications generally available for various systems such as structural steel; reinforced concrete; reinforced, unreinforced

,

or partially reinforced masonry and wood.

Also discussed are methods for determining appropriate basic

strength of materials in existing construction, as distinct from

methods in the Code required for new construction. This particular

facet of evaluation is usually one of the most difficult to establish with any consistency or realism.

The basic design equation states that a member's strength

shall not be less than load effects on that member, with an appropriate specified or implied factor of safety. The ratio of stress

from load effects to nominal code stress limit is called the Stress

Ratio. However, it is only necessary to determine the stress ratios

from those members of the structural frame that can be seen to be

most critical to the stability of the building.

The steps in the analysis discussed to this point are capable

of being completed by either hand calculations or by using a computer program called the "ACE Program." It is emphasized that this

computer program was included in our original research only on the

surmise that it would be of help. Hand calculation methods are usually adequate for the types of buildings generally proposed for

residential use.

28-2
Once critical stress ratios are determined, another decision

point in the evaluation is reached. That is, to determine whether

the building is adequate without retrofit or whether some form of

structural strengthening is required.

In our study, various methods used to strengthen existing

structures are discussed in a general manner. Specific methods

of strenghening for individual conditions are left open to the

designer. Methods not included may also be used when appropriate

for the particular building being strengthened.

Some of the more general considerations on strengthening methods,

are as follows:

(1) Increased shear resistance can be provided by adding new shear walls, adding encased reinforcement or by

increasing thickness of shear walls upon replacement.

However, added weight should be considered in its

effect upon foundations, foundation pressures, and

the building's inertial response to earthquake

mot ions

.

(2) Increased capacity of structural steel frames can be

provided by adding material, reinforcing joints, or by additional bracing of members.

(3) Concrete frames can be strengthened by adding reinforced shotcrete.

(4) Infill walls can be isolated to be compatible with

building motions or they can be strengthened by

adding reinforced shotcrete.

(5) Removal of the upper stories of a building may reduce

the overall hazard to an acceptable level.

During strengthening of the building, the status of existing

stresses should not be ignored. Shoring may be required. In addition, the resulting effect on the stiffness of the building should

be considered when selecting the method of strengthening. The

joinery of strengthening elements to the rest of structure should

also be considered in the design.

After the methods of strengthening are determined, a detailed

cost of the strengthening should be made so that the cost of retrofit can be made a part of the decisionmaking process.

Finally, a decision at this stage of the evaluation should

conclude that either retrofit strengthening is cost-effective,

further iteration of strengthening is required, or retrofit is

uneconomical. If further iteration is required. The process is

repeated until the desired level of strengthening is reached.

Computer Analysis

Basic steps of the computer program are as follows:

Step 1: Definition of properties of building.

Steps 2-4: Calculation of 1973 UBC loads.

Step 5: Definition of critical elements.

Step 6: Calculation of moments, shears, and axial

forces in critical element.

Steps 7-10: Calculation of critical stress ratios.

28-3
The computer program is constructed in modular form so that

the user can modify any one of the modules with a minimum of inconvenience. It 'is very important to note that the computer program

can serve multiple uses. The program can be used to calculate UBC

loads on the building, or additionally, it can be used to calculate

moment and shear forces or element critical stress ratios for the

UBC loading. The computer program is capable of modeling buildings

with flexible or rigid diaphragms or buildings for which torsional

response is significant.

To demonstrate the versatility of the program let us consider

the UBC loading module of the ACE program. In the calculation of

lateral forces due to earthquakes, the UBC requires the specification of building period. The user may directly input the desired

building period, or he may input building properties and calculate

a building period and then multiply that calculated building period

by a scale factor based on experience, or he may calculate the

period from formulas given in the 1973 UBC.

Once lateral forces are calculated they must be distributed

over the building height. The program enables the user to either

use the UBC formula for distributing forces, or alternately to distribute forces using the dynamic properties of the building and

the calculated fundamental mode shape of vibration, and should be

used when the building's lateral stiffness is not uniform over the

building height.

The module of the computer program which performs the structural analysis is referred to as ETABS. This is a structural analysis program especially created for multistory buildings by

Professor E. Wilson and his associates at University of California

at Berkeley. This building analysis program was selected over the

other potential building analysis programs for the following reasons

(Wilson, et al. , 1975)

:

(1) It is a computer program which has been used by the

professional structural engineering community to calculate building response for comparison with San

Fernando earthquake building response records.

(2) The computer program enables us to do a complete

three-dimensional model of the building when existenance of either mass or stiffness eccentricity indicates that torsional response may be significant.

(3) There are many lateral force resistance systems

available within the program for use in modeling.

For example, shear walls, brace frames, and infill

walls can all be modeled.

(4) It has the flexibility of adding equivalent springs

to the stiffness of each story. These equivalent

springs have been found to be useful in representing

the stiffness characteristics of structural members

which are exceedingly difficult to model in an exact

fashion.

(5) The program has considerable flexibility in the

treatment of loads. The user may calculate the

stress ratios for only lateral loads or alternately

the stress ratios can be calculated for any combination of dead, live or lateral loads.

28-4
As previously noted, the computer program is designed to be

modular in form. In the masonry component, for example, the critical stress ratios can be calculated for axial and bending or just

for shear stress.

The concrete critical element portion of the program branches

off into separate modules for shear walls, beams or columns.

The steel critical element portion of the program has options

for beams, columns or members that only carry axial loads.

To help the user in learning the program, our original research

contained a hypothetical example with several different types of

lateral forces resistance systems. The example is a two-bay by

three-bay frame, two stories high. Floor diaphragms are rigid.

Properties of one-bay are varied in the example in order to demonstrate how to input data for different lateral force resistance

systems

.

Example One

The first building is a six-story apartment building with basement.

The building is 39' x 112' in plan and 63' high. It has unreinforced brick masonry exterior walls with wood interior framing.

A 'K' factor of 1.33 is used for determining the base shear.

The diaphragms are flexible. After an examination of the masonry,

an allowable shear stress of 7.5 psi was chosen as being appopriate.

The building is rectangular in plan, providing diaphragms having a span-depth ratio of almost three.

Elevations of exterior walls were drawn indicating that first

story strength and stiffness would be a problem. Critical elements

were chosen for specific analysis. A wall analysis similar to those

used for new buildings was made. The members of the building frame were modeled and the elastic distribution of the shear on the wall

was determined. Shears and moments resulting from this analysis

were calculated.

Critical stress ratios and results for 100 percent compliance

with UBC 73, as well as for 75 percent, 50 percent, and 25 percent

are tabulated.

Various potential strengthening methods are discussed, and the

following were chosen for a detailed cost analysis:

(1) Diaphragm span reduction by providing interior transverse resisting elements.

(2) Reinforcement of exterior walls with shotcrete.

(3) Transverse interior shotcrete shear walls.

(4) Transverse interior steel moment frame.

Steel frames are placed on one side of the building corridor,

and are used to provide flexibility for later room modifications.

Consideration is also given to frame deformation and problems of

erect ion.

Shear walls are located in positions to minimize room layout

problems. First story flexibility and weakness is corrected by

filling in walls with windows consistent with room layout.

28-5
A summary of the cost of structural modiifcation is provided to assist in decisionmaking. Very little difference in cost was found

in providing for 25 percent or 100 percent of UBC 73 forces. This

indicates that in some cases design forces have little effect on

the cost of providing earthquake resistance.

Example Two

This example is a six-story building with basement, having a con- crete frame and brick masonry filler walls. The base of the building

is 110' x 200' in plan and 81' high. A

1 K' factor of 1.00 was used

to determine the base shear. Rigid diaphragms were assumed. The

masonry was considered somewhat better than that in Example 1 so an allowable shear of 10 psf was used.

A complexity offered to hand calculations is the fact that the

west edge of the building is skewed with respect to the other side.

Upper floors in the "L" shaped tower provided an eccentricity of mass to the base structure. Elevations again indicate that the

first story flexibility would require analysis. The wall adjacent

to the building next door is concrete and solid except for a vent

shaft.

A force diagram on the building layout indicates the severity

of first floor eccentricity. Distribution of the base shear and

story shears were calculated and tabulated. A compatible design

level for diaphragms was also determined.

This building, having rigid diaphragms and having a large

torsional moment required a torsional analysis to be made so that

the appopriate design shears on each wall could be determined. Once

the forces on each wall were determined, a wall analysis was performed similar to that in Example 1. Critical stress ratios were tabulated to assist in determining the method of strengthening.

Again, tables were assembled for 100 percent compliance with

UBC 73, Zone 3, and for 75 percent, 50 percent, and 25 percent to assist in the evaluation.

Interior shear walls were added. Exterior masonry walls were strengthened using 6" reinforced shotcrete in place of one wythe

of brick. After providing what was deemed to be the appropriate

strengthening, analysis was again performed. The final critical

element stress ratio table indicated that further strengthening was

not required.

As part of the required strengthening, such detail items as struts to shear walls were included. First-story archways were partially filled, with consideration given to continued functioning

of first-story shops.

Cost analysis was tabulated to provide the data necessary for

final decision making.

Conclusions

This methodology was developed for use by HUD technical personnel

to determine the feasibility of rehabilitating existing property for

residential use under HUD programs. It should also be a valuable

tool for consulting engineers, under contract with HUD, to evaluate

proposed rehabilitation projects.

28-6
The computer program is available through the National Information Service, Eathquake Engineering, Computer Applications at the

University of California at Berkeley. Copies of our three-volume

manual may be purchased from the Superintendent of Documents, U.S.

Government Printing Office, Washington, D.C. 20402.

Ac knowledgment

s

The information in this paper was primarily obtained from a taped

slide presentation prepared by Clarkson W. Pinkham of S. B. Barnes

and Associates and Gary C. Hart of J. H. Wiggins Co. for HUD technical personnel. Special thanks goes to Andrei Gerich, HUD Structural Engineer, for reviewing the paper and Sheila Lockwood for

typing.

References

1. International Conference of Building Officials, 1973, Uni form

Building Code , 1973 Edition, Chapters 23-28.

2. National Bureau of Standards, January 1975, Natural Hazards

Evaluation of Existing Buildings , Building Science Series 61

(BSS 61), "Approximate Analytical Evaluation Method."

3. Wilson, E. L., Hollings, J. P. and Dovey, H. H. , ThreeDimensional Analysis of Building Systems (Extended Version) , April 1975, Earthquake Engineering Research Center Report

No. 75-13.

28-7

29

DAMAGE TO ENGINEERING STRUCTURES DURING THE NEAR IZU-OHSHIMA EARTHQUAKE OF JANUARY 1978

Kazuto Nakazawa

Toshio Iwasaki

Kazuhiko Kawashima

Makoto Watabe

Hiroyuki Yamanouchi

Yutaka Yamazaki

Ministry of Construction

Damage to Engineering Structures During the

Near Izu-Ohshima Earthquake of January 1978

l The details of this January 14, 1978, earthquake have been presented in Paper 12 of this volume. Seismic intensity on the JMA

Scale at various sites is shown in Figure 1

Many fore-shocks were observed near Izu-Oshima Island; aftershocks, including a shock registering a magnitude of 5.8, occurred

: in the middle part of Izu Peninsula (see Figure 2). The fault line

of the main shock was directed from the near Izu-Ohshima to the

middle part of Izu Peninsula, in an east-west direction. Other

parameters about the fault are as follows:

Length of the fault line — 20 km

Dip angle — 90°

Dislocation — 1 m

Topography and Geology of the Izu Peninsula

! The central part of Izu Peninsula is a mountainous area with steep

hills. Volcanic topography is seen in this area, with famous

volcanic Mt . Amagi near the center of the peninsula. Mountain ridges

have a nearly constant height.

As shown in Figure 3, a map of the peninsula from the Shizuoka

Prefecture, the geology of this area is represented by the Yugashima

Group and the Shirahama Group consisting of Neocene volcanic ejecta,

volcanic lavas, and volcanic mud flows and pyroclastics . The

Yugashima Group and the Shirahama Group are surface-soil strata

made from tuffaceous sandstones, siltstones, breccia, and volcanic

conglomerates

.

In the areas of Inatori, Higashi Izu Town and Mitaka, Kawazu

Town, terraces with moderate slopes are made of volcanic mud flows

including site gravels. In the areas of Naramoto, Higashi Izu Town,

and Mitaka Iriya Nanamagari, Kawazu Town, pyroclastics cover the

surface. These are loose layers made of scoria, loam, and light

yellow volcanic ash, with moderate slopes of about 25°.

As in the Off-Izu Peninsula earthquake of May 9, 1974, active

j faults were disclosed after the recent earthquake. Surface dislocaj tions were found along a line connecting Inatori and Neginota.

Right lateral dislocations of 18 cm were seen on the pavement

of National Highway Route 135 (see Figure 4) located in front of

; Inatori and Neginota. Considerable damage was concentrated close

to the active faults.

Earthquake Ground Motions

The maximum values of the observed earthquake accelerations at the

sites are shown in Figure 5. The accelerograms closest to the

fault were on a basement floor of Itoh Denden Office, the epicentral

distance to which is nearly 30 km. Time histories and response

spectra of the accelerograms are shown in Figures 6 and 7. The

maximum vlaues of acclerations , velocities, and displacements of

the earthquake motions were:

Accelerations — 101.1 gal (EW) , 71.9 gal (NS), 31.9 gal (UD)

29-1
Velocities -- 16.2 kine (EW), 10.0 kine (NS), 5.1 kine (UD)

Displacements — 10.5 cm (EW), 2.8 cm (NS), 2.4 cm (UD)

The maximum values of the EW component are considerably larger

than those of the NS component. On the other hand there are some differences in frequency characteristics between the spectra, shown

in Figure 7 in EW and NS components. The differences on the maxi- mum values of the motions and the spectra between both of the

components are explained by the fact that the earthquake records were obtained near the fault and that the direction of the fault

line was EW.

The following are pointed out in Figures 6 and 7:

The duration of the main earthquake motion is 15 to 20

seconds

.

The predominant periods in EW component are 0.4 and 0.7

seconds. The maximum amplification factor 3.07 for

5 percent of critical damping at the former peak period.

The predominant period in NS component is 0.5 seconds,

with the maximum amplification factor 3.9 for 5 percent

critical damping.

The ratio Ay/A^ of the maximum acceleration of UD component to that of horizontal components is 0.32 and 0.44

for EW and NS components, respectively. According to the

maximum values observed at the various sites as shown in

Figure 5, the ratio, A/A^ is 0.45 on an average.

Figure 8 shows the distribution of overturned tombstones in

graveyards in Izu Peninsula (171 graveyards were investigated).

In the 3 areas ,Higashi-Izu Town to Kawazu Town, Amagi-Yugash ima

Town, and Nishi-Izu Town, ratios of overturned tombstones to total

number of ones in each yard exceeds 50 percent. However, it may

be that the overturned tombstones in the latter two areas were due

to after-shocks rather than the main shock. Epicenters of major

after-shocks are plotted in Figure 8.

Statistics of Damage

Regions hit by the earthquake cover the administrative districts of

Shizuko Prefecture and the Tolyo Metropolis (Izu-Ohshima Island).

Twenty-five persons were killed in all by slope failures and landslides. Earthquake damage in Shizuoka Prefecture is summarized in

Table 1. The total amount of the damage is estimated at 31 billion

yen (104 million U.S. dollars) — far greater than the damage from

the Off-Izu Peninsula Earthquake of 1974 by a factor of 12. Approximately 40 percent of this amount was damage to roads. Damage to public works is summarized in Table 2.

Damage to Civil Engineering Structures

The major damage mode to roads in the 1978 earthquake was landslides

occurring on natural slopes, while major damage in the Off-Izu

Peninsula Earthquake of 1974 resulted from settlements in road

embankments and cracks in pavements. Nevertheless, several failures

took place in embankments, retaining walls, and tunnels but highway

bridges and pedestrian overcrossings suffered little damage.

29-2
Road damage seemed isolated to Route 135 (national road), the

Higashi-Izu Toll Road, and Shuzenji-Shimoda Route (principal prefectural road) as shown in Figure 10. The areas heavily damaged are considered to be in good agreement with the areas where after-shocks

took place (shown in Figure 2).

Landslides

The field observations revealed that the slope failures could be

classified into three categories: failures of soil slopes, failures of rocky slopes, and rockfalls.

Typical soil failure could be observed in Naramoto Area (HigashiIzu Town), Mitaka-Iraya-Nanamagari Area (Kawazu Town), and Nashimoto

Area (Kawazu Town). In the former two areas, surface soils with a depth of 1 to 2 meters were underlaid by scoria with a depth of

2 to 3 meters and loam. The sliding of soils generaly took place

in the scoria stratum or between scoria and loam strata. The amount

of soil that fell was estimated to be from several thousand to

two hundred thousand cubic meters in this type of failure. In the

case of rocky and weathered surface soils, the amount of falling

soils was estimated to be from several thousand to ten thousand

cubic meters. Figure 11 shows a representative failure of this type

at Nashimoto Area (Kawazu Town). It was noted the failure of this

sort most likely happened at convex parts of slopes as shown in

Figure 12.

The falling stones consisted chiefly of rocks with few joints

and volcanic and site gravels, such as Yugashima Group, in both

Kawazu and Amagi-Yugashima Towns. The amount of falling stones was

on the order of several hundreds of cubic meters.

In addition to the slope failures described above, failures of

cutting slopes were also observed. However, later investigation

showed that the cutting slopes themselves did not generally fail.

The natural slopes above them sometimes fell on the cutting slopes.

It was also observed that slopes retained by concrete walls, concrete

frames, or concrete or mortar covering could survive the earthquake.

Embankment and Retaining Walls

I In the case of embankments constructed by both cutting and filling,

| filled parts generally settled, resulting in cracked pavements. In the Yugano Area along the Shuzenji-Shimoda Route, several embankments failed. Figure 13 shows a reprsentat ive failure 3.5 m in

width, 60 m in length, and maximum settlement of about 60 cm.

In the area hit by the earthquake there were a number of retaining walls for cutting and embanking slopes. The most common retain-

| ing walls were concrete or masonry. Although some of them were

i damaged as shown in Figure 14, they could survive safely with minor

j non-structural damages such as cracks, gaps, or small settlements.

' There were several high embankments with a height of more than 10

to 20 meters in the east Izu area as shown in Figures 15 and 16.

|: It should be noted that these high embankments could perform quite

I well against the earthquake when adequately retained.

Tunnels and Bridges 2

In the 425 m-long Tomoro Tunnel (Higashi-Izu Toll Road), a piece

of lining concrete near the arch crown and side wall fell down as shown in Figure 17. Large amounts of soil also fell near the tunnel

entrances, blocking transportaion in the Shirata, Kitoh, and Kurone

29-3
Tunnels on Route 135. In the Inatori Tunnel (Izu-Kyuko Railway)

lining concrete was crushed by fault movements, and uplifts shifted

the railbeds.

On the Shin-Shimoda Bridge (Shimoda City), Figure 18 shows that

the free end of a prestressed concrete girder moved horizontally

about 7 cm. This lateral movement crushed concrete handrails at both sides since they were rigidly connected to the abutments to set the mermaid statues. It was also observed that several pieces

of concrete at parapets and piers fell down in Shimoda Bridge on Route 135 and that a pier of the Minato Bridge in Shimoda City

settled approximately 0.5 meter.

Two pedestrian overcrosses survived with a few minor cracks,

although they were located only 20 meters away from faults near Inatori Junior High School (Figure 19) . Ground motion at the site was estimated to be very severe since several wooden residential

houses and most gravestones around the site were overturned.

Other Damage

Two tailing dams for storing mining wastes collapsed as shown in

Figures 20 and 21. The slimes that overflowed polluted the

Mochikoshi River, Kano River, and Suruga Bay.

Damage to Building Structures

Damages to buildings were classified into 4 groups:

Damage due to vibrational response of structures

Damage due to cracks caused by faults

Damage due to landslides

Damage due to instability in road-shoulders

Wooden Structures

Roof tiles fell off around roof ridges in many places (Figure 22).

Mortar finishing exterior walls were cracked and some fell off

because of underlying dry rot (Figure 23). Moreover, some wooden

structures collapsed due to heavy roof weights or weak ground

(Figures 24, 25, and 26).

Continuous footings of wood structures lying on faults were deformed and cracked, and by these deformations the structures

were distorted. Mortar concrete covering dirt floors had many

and large cracks (Figures 27, 28, and 29).

Landslides took the greatest toll of property and lives

(Figure 30). Many houses were swept away and crushed by earth

and sand (Figure 31).

Structures standing close to roads began to lean over because

of the instability of road-shoulders or failure of the retaining

walls on the opposite sides of the roads (Figures 32 and 33).

All damage in East-Izu seemed to be caused by the main shock,

while those in West-Izu were caused by the after-shock.

29-4
Reinforced Concrete Structures

In general, no reinforced concrete structure had severe structural

damage. The 3-story school building (Inatori lower secondary

school), which was about 100 meters from the fault line, had many

great cracks in the joint between the stair hall and the schoolrooms .

Most buildings built near slopes settled unevenly. Figures 34

through 37 show this typical damage and Table 3 classifies reinforced concrete buildings according to the types of damage they

suffered.

Damage to Steel Structures

It is said that steel buildings were more severely damaged than

reinforced concrete and wooden ones. In fact, the damage that

actually occurred to steel buildings was limited to non-structural

members. This fact gave the impression that the steel buildings

were especially hard hit.

No buildings with severe damage to main structural members could

be observed outside of an apartment house and hotel. Damages to non-structural members are typified by the damage to the exterior

walls of a hotel in the Inatori region (Figure 38). In the Izu district, mortar finish on metal lathing, similar to this hotel, is

widely used for exterior walls of steel buildings. In many cases, connectors — usually steel wires used to tie the exterior walls

to furring strips — were rusted by sea breezes.

As for structural problems, some buckling of diagonal braces

and fracture of their connections were observed, especially in the

gymnasiums of schools. However, the fabrication and execution of

these works were poor, particularly in welded or bolted connections

to fasten braces to gusset plates and to fasten gusset plates to beams or columns. Since this kind of damage to connections around

braces has also been seen in past earthquakes, such as the TokachiOki earthquake, this shows that recent seismic prevention recommendation have not been put to practical use.

The restaurant that collapsed (Figure 39) was result of poor

building practices in column supports, end connections, and gusset

plate connections of diagonal braces and beam-to-column connections.

By contrast, the apartment house (Figure 40) and the hotel

(Figure 41) failed because of poor structural planning. For the

apartment house, the first-story columns were too slender. As for

the hotel, the plan of the building was so complicated and each

H-shaped column was arranged so irregularly with respect to the

sectional axis that the details of beam-to-column connections were neglected.

In the Izu district, most steel buildings are a kind of mixed

construction — the first story is often reinforced concrete and

the stories over it are made of structural steel. In general, steel

in this type of building tended to whip from the earthquake excitations, because of the large difference among ratios of story rigidity to story weight. Because of this, a great deal of damage

occurred to exterior walls and interior partitions that were not deformable enough to follow the large deflection of the steel frames.

29-5
Acknowledgments

The Public Works Research Institute and Building Research Institute

received a great deal of assistance from the Headquarters of the

Ministry of Construction, the Chubu Regional Construction Bureau,

and Shizuoka Prefectural Government. Seismic data and strongmotion records were collected from the Earthquake Research

Institute of the University of Tokyo, the Seismology Division

of the Japan Meteorological Agency, the National Research Center

•for Disaster Prevention of the Science and Technology Agency,

the Port and Harbour Research Institute of the Ministry of

Transport, the Kanto Regional Construction Bureau, Japanese

National Railways, Nippon Telegraph and Telephone Public

Corporation, and Japan Highway Public Corporation. The authors

express their sincere thanks to the staff members of these above

organizat ions

.

References

1. Public Works Research Insitute, "Progress Report on Earthquake

Damages Due to the Near-Ohshima Earthquake in 1978," Journal

of Public Works Technics 20-4, April 1978.

2. Public Works Research Institute, "Earthquake Damages Due to the

Near-Ohshima Earthquake in 1978, Technical Note No. 1346, Ministry of Construction, March 1978.

3. Shizuko Prefecture, Geological Map of Shizuoka Prefecture , Naigai

Map Ltd. , March 1973.

4. Public Works Research Institute, "Progress Report on Earthquake

Damage Due to the Off-Izu Peninsula Earthquakes," Journal of

Public Works Technics 16-7, July 1974.

5. Assoication for Promoting Strong-Motion Acceleration Measurements

Progress Report No. 13 — Near Izu-Ohshima Earthquake in 1978,

National Research Center for Disaster Prevention, Science and

Technology Agency, March 1978.

6. Public Works Research Institute, Final Report for Research and

Development on Earthquake Engineering Technics, Technical Note

No. 1250, Ministry of Construction, March 1977.

7. Building Research Institute, Progress Report on Earthquake Damage

Due to the Near-Ohshima Earthquake 1978, Ministry of Construction

January 1978.

8. Makoto Watabe, "Research on Design Earthquake," B R I Research

Paper No. 67, September 1976.

29-6
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FIGURE 3: GEOLOGICAL MAP OF IZU PENINSULA

29

jlGURE 4: TRACES OF EARTHQUAKE FAULT ON ROUTE 135 IN FRONT

OF INATORI JUNIOR HIGH SCHOOL (LIGHT LATERAL

DISPLACEMENT WITH A MAGNITUDE OF 18 cm)

29-11
FIGURE 5: THE MAXIMUM GROUND ACCELERATIONS

FIGURE 6: TIME HISTORIES: ITOH DENDEN OFFICE

29-12
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FIGURE 7: RESPONSE SPECTRA: ITOH DENDEN OFFICE

29-13
29-14


29-17
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29-18
FIGURE 17: FAILURE OF LINING CONCRETE NEAR ARCH CROWN AT TOMORO

TUNNEL (HIGASHI-IZU TOLL ROAD)

FIGURE 18: FAILURE OF CONCRETE HANDRAIL AND LATERAL MOVEMENT OF

CONCRETE GIRDER ( SHIN-SHIMODA BRIDGE, SHIMODA CITY)

FIGURE 19: OVERCROSSING FOR PEDESTRIAN LOCATED ONLY 20 m FROM FAULTS NEAR INATORI JUNIOR HIGH SCHOOL AT INATORI AREA IN HIGASH-IZU TOWN, WHICH SURVIVED

WITH MINOR CRACKS AT PARAPETS

29-19
FIGURE 20: OVERALL VIEW OF COLLAPSED HOHZUKI-ZAWA TAILINGS DAMS

FOR STORING MINING WASTE - FOREGROUND SHOWS THE FIRST

DAM COLLAPSED JUST AFTER THE MAIN SHOCKS (JANUARY 14,

1978); RIGHT SHOWS SECOND DAM COLLAPSED FROM AFTER-SHOCK

AT ABOUT NOON IN JANUARY 15; LEFT SHOWS THE THIRD DAM

WITH NO DAMAGES.

FIGURE 21: COLLAPSED DAM AND TEMPORARY DAM

29-20
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29-24
FIGURE 38: DAMAGE OF MORTAR FINISH FIGURE 40: DEFORMED COLUMN (ALREADY

ON METAL LATHING (HOTEL, REPAIRED) (KAWAZU-REGION)

INATORI-REGION)

FIGURE 41: DEFORMED COLUMN AND DAMAGED WALLS (INATORI-REGION)

29-25
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FIGURE 39 b: RESTAURANT STRUCTURE

29-26
APPENDIX - COMPARISON OF SEVERAL SEISMIC INTENSITY SCALES

29-27

30

REPAIR AND RETROFIT OF BUILDINGS

J. K. Wight

R. D. Hanson

Civil Engineering Deparment

University of Michigan

Repair and Retrofit of Buildings

Damage surveys after recent earthquakes in Alaska, Venezuela, California, Nicaragua, and Guatemala clearly demonstrate the need for two different, but closely associated technologies. First, because

it is economically unfeasible to replace all "older" buildings, an ability to retrofit (increase earthquake resistance) buildings is required. Also, because no building new or old can ever be "earthquake proof," an ability to repair (restore original strength and

stiffness) is required. In many instances, repair and retrofit

would be carried out simultaneously in buildings damaged during a recent earthquake. However, extra care must be exercised for both

repair and retrofit situations. If poorly planned or poorly executed, the earthquake resistance of a building may actually be

diminished rather than improved by a repair and/or retrofit procedure. Interest in repair and retrofit of buildings has increased

dramatically in the U.S. since the San Fernando earthquake (1970)

and a summary of research needs was formulated during a workshop

[1] sponsored by the National Science Foundation.

Repair of Reinforced Concrete Buildings

In general, many reinforced concrete structures will survive a medium or large earthquake with only minor or moderate damage. For

such structures, a combination of the following two repair techniques has been used in the past. If the damage is minor, consisting

of narrow cracks with no significant spalling or loss of concrete,

an epoxy injection technique is used. More extensive damage will

require a removal of all unsound concrete and filling of the resulting void with a concrete-type material that has strength and stiffness properties approximately equal to those of the existing

concrete. In either case, it is desireable to use a high-earlystrength material so the building can be repaired and reoccupied

in the shortest possible time.

Experimental investigations into the behavior of reinforced

concrete members and subassemblages repaired by one of the two techniques described above have been carried out at the University

of California [2, 3], the University of Michigan [4], and the

Portland Cement Association [5]. A summary of the results from

the University of Michigan tests are given below.

Epoxy Injection

Epoxy injection has been widely used as a post-earthquake repair

procedure, but the success of the repair is strongly influenced

by a skill of the operator and the quality of the epoxy resin. In

a laboratory situation where a high-quality material and a highly

skilled operator is used, the epoxy injection technique has been

shown to be quite successful.

In the University of Michigan tests, beam to column subassemblages (Figure 1) were subjected to cyclic load reversals of varying intensity to produce different levels of damage. Specimens with

only a small amount of damage, consisting primarily of narrow

flexural and shear cracks in the region of the beam adjacent to the column, were repaired by the epoxy-in ject ion technique. The

repaired specimens were then subjected to the same loading routine

and the results were compared with the original test results. The

beam load vs. deflection relationships for a typical original and

repaired specimen are given in Figures 2(a) and 2(b), respectively.

The behavior of the repaired specimens generally compared favorably

with that of the original specimens. Strength and energy dissipation

30-1
capacities were restored with only a minor loss in initial stiffness. The inability of the epoxy injection technique to fully

restore bond between reinforcement and concrete is assumed to be

primarily responsible for the loss of stiffness.

Removal and Replacement

The only effective method of repairing severely damaged structural

elements is to remove all unsound concrete and replace it with a new material. Before the new material is placed, some modification

of the reinforcement will probably be needed. Longitudinal bars

that are buckled or displaced will need to be straightened or replaced. Also, additional transverse reinforcement should be added

to satisfy the latest code requirement [6, 7]. Finally, all of the

reinforcement should be sand-blasted to remove any loose material

that may adversely effect the bond between the reinforcement and

the repair material.

In the test series conducted at the University of Michigan,

various high-early-strength repair materials were used. In general,

the best results were obtained when a standard concrete mix with

high-early-strength was used. Specimens repaired with this material

generally had the same strength and energy dissipation capabilities

as the original specimens. When other high-strength repair materials

were used, two general problems caused the behavior of the repair

specimens to be considerably poorer than that of the original

specimen. First, some of the stronger material tended to be brittle

and had little or no energy dissipation potential. Second, the use

of a higher strength material tended to alter the location and/or

type of inelastic behavior for the repaired specimen. In all of the

original specimens the majority of inelastic activity could be classified as flexural behavior and occurred in the region of the beam

adjacent to the column. In some of the specimens repaired with

higher strength material, the inelastic activity shifted into the

joint and led to less ductile shear and anchorage failures.

Conclusions on Repair

(1) If well executed, epoxy injection is a useful method for restoring strength and energy dissipation capacity to a lightly damaged

reinforced concrete member. (2) In more heavily damaged members,

a removal and replacement technique using a concrete with properties

equivalent to the existing concrete in the area of the repair region

should be sufficient. Reinforcement in the damaged region may also

need to be modified. (3) Brittle materials should be avoided in

all repair situations. (4) To be successful a repair procedure must

be well planned and should include a study of how the repair will

effect members and connections adjacent to the repaired region and

how the repair will effect the overall structural strength, stiffness, and probable response to a future earthquake.

Strengthening and Stiffening Existing Buildings

As mentioned in the introduction, there are numerous buildings in

seismic zones that do not have adequate lateral resistance and are

a potential life hazard in the event of an earthquake. However,

instead of replacing these buildings it is economically more feasible

to strengthen and stiffen the basic structural system. Unfortunately,

a well-defined procedure for strengthening and stiffening cannot be

developed because each building is unique and therefore requires a unique solution. Also the degree of strengthening and stiffening

30-2
required will vary from building to building depending on the existing lateral resistance of the building, the importance of the building, and the design earthquake intensity. It is generally accepted

that emergency facilities must be kept operational after an earthquake. However, the degree of strengthening required for other

buildings will probably be a social-political-economical decision

made through a government agency acting on information supplied by

professionals from all concerned fields.

Some strengthening and stiffening techniques that may be

used alone or in various combinations are enlargement of existing

frame members, addition of shear walls or other structural members,

uncoupling of non-structural elements that may hamper the structure's

response to lateral loads, and the strengthening of unreinforced

masonry walls by the application of shotcrete. Examples of each

of these techniques were reported at the previously mentioned

workshop [ 1 ]

.

A recent investigation at the University of Michigan on the use

of walls to strengthen and stiffen a single-story, single-bay reinforced concrete frame was reported by Kahn and Hanson [8], The

basic frame used in this investigation is shown in Figure 3. Various infilling techniques were used (Figure 3) and were compared

with the behavior of a frame with a monolithical ly cast wall. All

of the units were subjected to lateral load reversals applied at the top floor slab.

The results of this investigation indicated the general adequacy of each of the infilling techniques for strengthening and

stiffening existing frames. The cast-in-place infill developed the

same ultimate strength as the monolithic wall, but it was not as ductile because of problems with the horizontal dry-pack joint at the top of the infill. The single precast infill panel did not obtain the same ultimate strength as the monolithic wall, but it did exhibit a more ductile type of behavior. The multipanel infill

with vertical openings between the panels obtained about one-half

of the ultimate strength of the monolithic wall; but it was the

most ductile of the specimens tested. This study clearly indicated

that different strengthening techniques can result in significantly

different stiffnesses, lateral load capacities, energy dissipation

capacities and ductilities. Results such as this should be used

in analytical studies of building behavior before deciding which

infilling technique would be most appropriate in a given situation.

References

1. Hanson, R. D. , "Repair, Strengthening and Rehabilitation of

Buildings - Recommendations for Needed Research," Report .to Workshop sponsored by the National Science Foundation, Department of Civil Engineering Report No. 77R4, The University of

Michigan, October 1977.

2. Mahin, S., Bertero, V., Atalay, M. , and Rea, D. , "Rate of Loading Effects on Uncracked and Repaired Reinforced Concrete Members,

Earthquake Engineering 'Research Center, University of California,

Berkeley, California, Report No. EERC 72-9, December 1972.

3. Celebi, M. and Penzien, J., "Hysteretic Behavior of Epoxy-Repaired

Reinforced Concrete Beams," Earthquake Engineer Research Center,

University of California, Berkeley, Calfiornia, Report No. EERC 73-5, February 1973.

30-3
4. Lee, D. L. N. , Wright, J. K. , and Hanson, R. D. , "Original and

Repaired Reinforced Concrete Beam-Column Subassemblages Subjected to Earthquake Type Loading," Department of Civil Engineering Report No. UMEE 76S4, The University of Michigan,

April 1976.

5. Oesterle, R. G. , Fiorato, A. E., Johal, L. S., Carpenter,

J. E., Russell, H. G. , and Corley, W. G. , "Earthquake Resistant

Structural Walls - Tests of Isolated Walls," Report to the

National Science Foundation, Portland Cement Association,

November 1977.

6. American Concrete Insitute, Committee 318, "Building Code

Requirements for Reinforced Concrete (318-77)," Detroit,

Michigan, 1977.

7. International Conference of Building Officials, "Uniform

Building Code," 1976 Edition, Whittier, California, 1976.

8. Kahn, L. F., and Hanson, R. D. , "Reinforced Concrete Infilled

Shear Walls for a Seismic Strengthening," Department of Civil

Engineering Report No. 76S2, The University of Michigan,

January 1976.

I N (APPLIEO)

»V (REACTION)

60" 2

_lO

o

P,A (APPLIEO)

BEAM

a = 41 or 47 in.

•V (REACTION)

N (APPLIED)

FIGURE 1: BEAM TO COLUMN SPECIMEN USED IN REPAIR INVESTIGATION

(1 in. = 25.4 mm)

30-4
FIGURE 2: BEAM SHEAR FORCE VS. DEFLECTION RELATIONSHIPS

30-5
Basic Frame Dimensions

Infilling Techniques: 1. Cast-in-Place (not shown)

Mechanical Connectors r

in. gap

a | i t i | | t n im i » t u u t

3

g in. gap

in. gap -

]

SKIT"

T

LicrJo r. IH I tttI

2. Single Precast Panel 3. Multiple Precast Panel

FIGURE 3: BASIC FRAME AND INFILLING TECHNIQUES (1 in. = 25.4 mm)

30-6
31

THE DISTRIBUTION OF PROPERTY LOSSES

CAUSED BY HISTORICAL EARTHQUAKES

Eiichi Kuribayashi

Tadayuki Tazaki

Public Works Research Institute

Ministry of Construction

The Distribution of Property Losses

Caused by Historical Earthquakes

A number of quantitative estimations of earthquake disasters have

been carried out in the past. References ^ jl^deal with the damage

to existing wooden houses. References and investigate the

damage to bridges and underground pipes, respectively. To express

the amount of losses, various kinds of indexes are used such as the

ratio of razed houses to standing ones or damaged spots per service

line for underground pipes. However, a common index is necessary

to evaluate the total loss from disasters and to compare them. The

property loss valued in money meets with this requirement. With

this in mind, the authors have investigated the evaluation method

of property loss distribution by analyzing eighteen historical

earthquakes in Japan.

Methodology

The authors assumed a ratio of property losses caused by earthquakes

as an index of earthquake disasters. The property loss ratio can be

defined as a ratio between property losses [1] and existing assets

(w) valued in money and expressed as a function of earthquake

magnitudes, epicentral distances, and subground conditions as follows

Dj = G

I 10 aM + BA (1)

D n

= G n 10 aM + BA (2)

where the suffixes I and II mean older subgrounds than diluvial

deposits and younger than alluviam, respectively, and

'IIM

constant for the subground types, I and II,

earthquake magnitudes in Richter scale,

epicentral distances in km

constants

.

The ratio is to be evaluated in accordance with the subground

condition as shown in Figure 1(A). The existing assets distribute

independently from it like in Figure 1(B). The property loss caused

by an earthquake can be evaluated by multiplying the existing assets

and the ratio as shown in Figure 1(C).

The constants were determined by correlating actual property

losses and the estimation of the losses of eighteen earthquakes in

Japan since 1923.

A total property loss (L) in each earthquake can be estimated

by summing up the losses of each mesh (1|) which is appropriately

divided into

£ (6

i (DI

) i + (1 6 i>( DII>i } w

i (3)

where

w

i (DO.

< D

II>i

existing assets.

property loss ratio in subground type I, property loss ratio in subground type II,

loss in mesh i, identification index {° for subground type I in mesh i_

1 for subground type II in mesh i,

31-1
The assets at each mesh were estimated from the relationships

between the assets and local populations. Table 1 shows the relationships between them in nine districts of Japan, from the National

Wealth Survey and National Census as of 1970. It can be said that

the assets correlate with the population in each district,

AT

= 2.77 • P - 0.891 (4)

A,j, = Gross tangible fixed assets in trillion yen,

P = Population in million.

In Equation (4), the regression constant of 0.891 is negligible

compared to the assets. Then Equation (3) becomes:

P

L = I {6i(Di)i + (1 - 6i)(Dn)i} W -i- (5)

P

where

W = National wealth in yen,

P = Population of Japan,

P£ = Population of mesh i.

The population of mesh i(p> ) was estimated from the average population density of quake-damaged area, which was obtained from the sum

of the population and area of the municipalities suffering an earthquake . W Pn L = --f- I {6i (D^i + (1 - 6i)(D II ) i } a£

F A

D i

= l

?»

z {6iGl + (1 -

6i )G Il} 10«M+ BAi a . (6)

v A

D i where

Pp = Population of the quake-damaged area An

= Area of the quake-damaged area in km^.

a£ = Area of mesh i in km^,

A£ = Epicentral distance of the center of the mesh (km).

Figure 3 shows the procedure to estimate the losses caused by

an earthquake as stated above. The mesh used in this analysis was made to coincide with that of a hundred times of the National Land

Information Mesh Data of National Land Agency 1 . The size of mesh

of National Land Agency varies in accordance with the latitude.

Therefore the mesh in this analysis was determined to be the same size as shown in Table 2.

The loss was summed within the area where damage possibly

occurred as shown in Equation (7).

log R = 0.5M - 1.5 (7)

where

R = Radius of the area where damage possibly occurred in

km (See Figure 4),

M = Magnitude in Richter scale.

31-2
The unknown constant G-,, G T ^

, a and 6 were determined by

minimizing the difference or the sum of square of log / D\ and ^ ^ 1.7 D „ LPA

its estimate log

^ D)

.

WP

D

a'

LPAn LPAn ,

? tlog(

J)j

- log( °>j>

2 * Min (8)

j D D

j = Suffix to denote earthquake j

Data Used in this Analysis

National Wealth Survey has been conducted twelve times until 1970

in Japan as shown in Table . Usually it has been conducted

every five years. However there is a twenty-year blank before and

after World War II. To interpolate the national wealth during the

blank years, it was estimated from the gross national product (GNP).

Figure 5 shows the relationship between the gross national product

and national wealth in nine districts of Japan. In the relationship, the definitions of the national wealth before and after World

War II were quite different. For example, before the war it involved land and natural resources. After the war, it only

consists of reproducible tangible assets. Moreoever, the definition

of each survey year was slightly different. Figure 5 was drawn by

roughly modifying the data to be uniform as for the definition of

the national wealth.

log lQ W = 1.03 + 0.937 log 1Q

N (unit: million yen) (9)

N = Gross National Product.

Population and Area of the Damaged Area

The population and area of the damaged area to be used in Equation

(6) were the sum of those of municipalities which were involved

within the radius (R 1 ) given by Equation (10). The radius R' corresponds to the area within which the seismic intensity of Japan

Meterorological Agency is expected to be greater than V

log 10

R' = 0.5M - 1.85 (10)

M = Magnitude in Richter scale.

[131 The data of population and area were as of 1970 . Although

the local population must have changed, the ratio of the population

of Japan and that of the damaged area, which was used in Equation

(6), is considered to have changed little. Figure 6a-r show the

seismic intensity map and the area denoted by Equation (10).

Table 5 shows the population and area of the damaged area used in

the analysis.

Subground Condition

Each mesh was classified into subground type I or II according to Reference^ . Figure 7 shows an example of the classification

of subground in the northern part of Japan.

Earthquakes Used in the Analysis

Eighteen earthquakes, whose recorded property loss are listed in

Table 6, ^* were used in this analysis. The influence of

31-3
inflation was not taken into consideration because both the loss

and existing assets were values in current price.

Results of the Analysis

The regression formulas for Equations (1) and (2) were introduced

as follows:

Dj - 7.9 x lO" 7 x i 0 0- 62M-°-°° 28 (11)

Dn

= 1.9 x 10"6 x 100- 62M-°- 002 9 (12 ) 6.15M1 8.1

The multiple correlation coefficient by Equation (13) was 0.82,

r

? {( —~)\ - ( ?•);}

j WP

D

a

J wpD

a

J h _ —- (13)

! LPAn LPArs 2

f

j WPD

a

J WP

D

a

where

( LPA LPA

D

) = Average of for all earthquakes,

WP

D

a 6 WP

D

a LPA

/ D-j _ Regression estimate calculated by Equation (11) and

WP^a J (12) for earthquake j.

Figures 8 and 9 show the value of Dj and Djt> Th e relationship

LPA • . ^LPA~~ between the actual value ^ D) and its estimate (__D_) is shown in

WPD

a WPDa in Figure 10. As far as the multiple correlation coefficient and

Figure 10 are concerned, the regressional formulas can be considered

to estimate the property losses caused by an earthquake. An arrow

at Kanto Earthquake shows the band of losses of various estimators.

One reason that the loss value and its estimate of Fukui earthquake

differ might be that the earthquake occurred just after the World

War II and damage was not surveyed in detail.

Figure 11 illustrates the equivalent ratio of razed houses

given in Reference*^ . Comparing Figure 11 to Figures 8 and 9,

the following can be said:

(1) Both property loss ratio and equivalent ratio of

razed houses increase with the magnitude of earthquake, and decrease with the epicentral distance.

(2) In general, equivalent ratio of razed houses is

greater than property loss ratio.

(3) Decreasing rate with the epicentral distance for

equivalent ratio of razed houses is greater than that

for property loss ratio. Decreasing rate with the

magnitude' is almost the same for both.

* Equivalent ratio of razed houses

_ Number of razed houses + 0.5 x Number of half-razed houses

Total number of razed houses

31-4
Comparison of Property Loss

Figure 12 shows the breakdown of the social capital according to the National Wealth Survey of 1970. If the composition of social

capital can be assumed to be constant for a decade before 1970,

the ratio calculated by Equation (14) of various kinds of lifeline

facilities for eight earthquakes from 1961 till 1970 represents the

individual loss ratio of facilities.

L.P

Individual loss ratio = (14) W

k

P D L^ = Loss valued in money of facility k

W. = Existing assets of facility k

= National Wealth as of the year of the earthquake

Assets of facility K as of 1970

National wealth as of 1970

Table 7 shows the result of calculation. Figures 13 to 20 show

the relationships between the individual loss ratio of various kinds

of facilities and the total property loss.

The procedure to evaluate the loss value is not always the same

in each facility. Therefore it might not be accurate to compare

the ratio of Equation (14) for different facilities. However, it tends to be that the loss ratios of buildings and highways are almost the same as that of the total property loss. The loss ratios

of agriculture, river, and harbor facilities are higher than that of

the total. The loss ratios of electric power, telecommunication,

and water supply facilities are lower than that of the total, but

they gradually reach the same level as the ratio itself comes to a greater value (see Figure 21).

Conclusions

Property loss can be used as the index to the scale of the earthquake disaster.

Property loss can be estimated from the magnitude of the

earthquake, epicentral distance, and subground condition.

The property loss ratio is in general smaller than the

equivalent ratio of razed houses.

Buildings and highway facilities are likely to suffer the

average property loss. Agricultural, river, and harbor facilities

are likely to suffer greater loss than the average. Electric power,

telecommunication, and water supply facilities are likely to suffer

less property loss.

References

All references listed below are written in Japanese.

1. Kawazumi, H. , Damage Distribution and Earthquake in Tokyo,

Architecture Magazine , 773, 1951.

2. Omote, S. and Miyamura, S. , Relationship between Damage Distribution in Earthquake in Yokohama and Nagoya City, and Subground

Conditions, Architecture Magazine , 773, 1951.

31-5

31

iL
3. Kur ibayashi , E., et al., Investigation on the Damage of Fukui

,

Yebino and Izuhanto-oki Earthquakes, Technical Memorandum of

PWRI No. 1106, March 1976.

4. Kuribayashi, E., et al., Investigation on the Estimation Method

or Razed Houses and Fire Distribution Caused by Earthquakes,

Technical Memorandum of PWRI No. 1256, May 1977.

5. Katayama, T. , A Methodology to Estimate the Bridge Capability

Against Earthquakes, Journal "Column".

6. Kubo, K. , et al., Quantitative Analysis of Seismic Damage to Buried Pipelines, Proceedings of the 4th Japan Earthquake Engi - neering Symposium , 1975.

7. Economic Planning Agency: 1970 National Wealth Survey of

Japan .

8. National Land Agency, Regulation Bureau, General Affairs

Section: Outline of National Land Data Collection Project .

9. Katsumata, M. , Earthquake Magnitude and the Area Possibly to Suffer Damage. Proceedings of the Spring Session of Seismolog - ical Society of Japan , 1977.

10. Economic Planning Agency, Economic Institute: National Wealth

Survey in Japan , March 1976.

11. Economic Yearbook, 1975, Toyo-Keizai-Shimpo-Sha.

12. Usami, T. , List of the Earthquakes that Caused Damage in Japan , Tokyo University Press.

13. New Japanese Atlas of Each Prefecture, 1972, Kokusai-ChigakuKyokai

.

14. Economic Planning Agency and National Land Agency: Classified

Maps of Subground.

15. Reports of the Imperial Earthquake Investigation Committee,

No. 100 C, p. 57.

16. Reports of the Imperial Earthquake Investigation Committee,

No. 101.

17. Kyoto Prefecture, Report of Oku-t ango Earthquake, March 1927.

18. Central Meteorological Observatory, Report of Kita-izu Earthquake of November 26, 1930; December 1930.

19. Nishi, et al. , Report of Shizuoka Earthquake, Journal of Japan

Society of Civil Engineers , Vol. 21, 1935.

20. Akita Prefecture, Report of Oga Earthquake , April 1942.

21. Tottori Prefecture, Report of Tottori Earthquake , p. 9, 1944.

22. Construction Authority, Report of Technology Institute No. 16,

Report of Nankai Earthquake, June 1948.

23. Kochi Prefecture, Report of Nankai Earthquake.

24. Report of PWRI, No. 78, Report of Fukui Earthquake, March 1949.

31-6
I 25. Okamoto, S. , and Kubo, K. , Damage of Civi lengineer ing Facilities Caused by Imaichi Earthquake of December 26, 1949,

Proceedings of Japan Society of Civil Engineers , No. 10,

December 1951.

26. Report of PWRI No. 113, Report of the Damage of Hyuganda Earthquake, October 1961.

27. Tada, H. , et al., Investigation of the Damage of Miyagikenhokubu Earthquake (I), Journal of PWRI , Vol. 4, No. 10, 1962.

28. Niigata Prefecture, Documents of Niigata Earthquake , June 1965.

29. Miyazaki Prefecture, Documents of Yebino Earthquake , April

1969.

j 30. Research Committee of Tokachi-oki Earthquake of 1968: Report

of Tokachi-oki Earthquake of 1968, January 1969.

31. Report of Nemuro-hanto-oki Earthquake of June 17, 1973,

Research Project of Natural Disaster, Ministry of Education.

32. Oyo Corporation, Damage of Ohitaken-chubu Earthquake , June

1970.

33. Kawaguchi, M. , Introduction of Multivariate Statistical

Analysis, Morikita Shuppan.

Appendix: Evaluation Method

The property loss valued in money used in this analysis is mostly

issued by the disaster organizations of each administration. Its evaluation method is divided in two. The first is expressed as

I follows:

Property Loss = Current price x Loss ratio

Current price is evaluated by reducing the depreciation from

the purchase price. Privately owned properties, which are not to

be qualified for the national subsidy for reconstruction, belong to

this category. Buildings (excluding public ones such as school

I buildings), agricultural products, and hatching products are the

I examples.

Second category is that the loss is evaluated as the price

necessary to reconstruct the damaged facilities. It involves

public facilities, which get the national subsidy. The assessed

amount of subsidy by the upper administration becomes the loss.

Water supply, education, agricultural, civil engineering, forestry,

fishery, electric power, telecommunication, and railroad facilities

belong to this category.

In each category, the method of evaluating the loss is not equal. For example, in the first category, the current price, which

I is very important in the loss evaluation, depends on the judgment

of the personnel. In second category, the property loss is likely

greater because the former does not. take the depreciation into account

.

31-7
31-8...J
Table 1 Relationship between Gross

Tangible Fixed Assets and

Population

District

Population

(thousand)*

Tangible Fixed

Assets (Bil.) Hokkaido 5,392 14,569

Tohoku 11,774 26,559

Kanto 28,174 79,782

Hokuriku 2,785 9,444

Tokai 13,408 38,808

Kinki 16,498 46,469

Chugoku 7,129 20,621

Shikoku 4.059 12,032

Kyushu 12,770 25,875

Total 101,989 274,159

* As of Oct. 1, 1970.

31-9
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31-10
Table 3 National Wealth Survey in Japan

Surveyed

Year

Published

i car anu Month

W 1 i^d 1 11 Zcl I IU I 1 INOLc

1905 Unidentified Bank of Japan

1910 1912 »

1913 Oct. 1921 Census Authority

1917 Unidentified Bank of Japan

uct. Census Authority

1924 Feb. 1928 Statistics Bureau of the Cabinet

1930 Nov. 1933 Surveyed actually in Dec.

1 yoL

1

1 03^ 7JJ ^11 juiTT\7vcycu £*\T£*/~\ dciudiiy Q Pf"l 1 Q 1 1 If inin uct« ll C 1935

1955 March 1958 Economic Planning Agency >

1960 Dec. 1964 „ Surveyed reproducible

1965 Aug. 1967 it 1

1

'tangible assets only.

1970 Feb. 1975

Table 4 Relationship between GNP

and National Wealth

Year GNP : N

Unit:

million yen

National Wealth

(gross) : W

Unit:

million yen

1905 2,000 9,030

1910 2,757 12,904

1913 3,030 15,020

1917 3,527 28,605

1919 6,384 42,044

1924 12,904 63,824

1930 10,636 105,211

1935 14,532 132,073

1955 8,633,600 37,987,000

1960 15,499,200 57,602,000

1965 31,956,400 68,420,000

1970 70,730,900 302,219,600

31

31-11
Table 5 Population and Area of the Damaged Area.

Earthquake Year

Population of the Damaged Area*

Unit: thousand

Area of the Damaged

Area

Unit: Km 2

Kanto 1923 28,545 28,397

Kita-Tajima 1925 141 943

Kita- Tango 1927 1,247 10,130

Kita-Izu 1930 1,744 4,005

Shizuoka 1935 661 1,426

Oga 1939 225 1,392

Tottori 1943 1,035 9,264

Nankai (Kochi Pref.) 1946 787 7,105

Fukui 1948 1,310 6,115

Imaichi 1949 590 2,139

Tokachi-oki 1952 713 25,528

Hyuganada 1961 27 295

Miyagike n- ho kubu 1962 535 1,955

Niigata (Niigata Pref.) 1964 1,203 4,886

Yebino 1968 102 970

Tokachi-oki 1968 2,040 23,410

Tokachi-oki (Hokkaido Pref.) 1968 277 6,880

Tokachi-oki (Aomori Pref.) 1968 953 6,504

Nemuro-hanto-oki 1973 66 846

Ohitaken-chubu 1975 144 1,149

* As of Oct. 1, 1970,

31-12
Table 6 Property Loss Cuased by Earthquakes

Earthquake Year Magnitude Property Loss

(Unit: million

yen)

Estimated

National Wealth

(Unit: million

yen)

Kanto 1923 7.9 5,500 77,000

Kita-Tajima 1925 6.5 89 78,000

Kita- Tango 1927 7.5 82.18 68,000

Kita-Izu 1930 7.0 25.46 63,000

Shizuoka 1935 6.3 10.32 96,000

Oga 1939 7.0 7.73 182,000

Tottori 1943 7.4 160 337,000

Nankai (Kochi Pref.) 1946 8.1 2,792 2,206,000

Fukui 1948 7.3 305,000 11,120,000

Imaichi 1949 6.7 3,500 13,860,000

Tokachi-oki 1952 8.1 15,183 24,200,000

Hyuganada 1961 7.0 163 70,380,000

Miyagiken-hokubu 1962 6.5 4,049 77,500,000

Niigata (Niigata Pref.) 1964 7.5 130,000 103,700,000

Yebino 1968 6.1 8,876 178,200,000

Tokachi-oki 1968 7.9 58,395 178,200,000

Tokachi-oki (Hokkaido Pref.) 1968 7.9 11,356 178,200,000

Tokachi-oki (Aomori Pref.) 1968 7.9 47,039 178,200,000

Nemuro-hanto-oki 1973 7.4 3,925 365,400,000

Ohitaken-chubu 1975 6.4 4,493 482,700,000

31-13
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31-14
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Harbor

Facility

Assets:

w

P

712,562

1,224,839 1,224,839 1,224,839 2,510,651

Harbor

Facility

Loss:

L

P

12,177 2,247

465.2

1,781.8 OsOn

vO

cuJ

a

0*

0.002097

0.1257

0.02646 0.01443 0.01025 0.02790 0.01562

0.002472

Water Supply

Facility Assets:W

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1,155,000 1,545,000 2,656,000 2,656,000 5,443,000 7,192,000

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Facility

Assets:

W

E

1,894,000 2,791,000 4,798,000 4,798,000 9,835,000

12,994,000

Electric

Power

Facility

Loss:

Le m

2,011

00CO

-<t<

CNm

CO

*o

CO

CN "V

Earthquake

Hyuganada

M

iyagiken-hokubu

Niigata

(Niigata

Pref.)

Yebino (Miyagi

Pref.)

Tokachi-oki

(Hokkaido) (Aomori

Pref.) 1

Nemuro-hanto-oki

Ohitaken-chubu

31-15
31-16
Tangible fixed assets (A-j-)

Fig. 2 Relationship between Gross

Tangible Fixed Assets and

Population

Dimension of Earthquakes

Magnitude M

Division of the Damaged

Area into Meshes

Epicentral Distance of Mesh i Ai

Damaged AreaNational Wealth W

National Population P"

-Subground Type of Mesh i

"Area of Mesh i aj

Population and Area of Damaged Area Pq, Aq

Property Loss Ratio (D[)i (Dij)i

— Property Loss L

Fig. 3 Flow Chart to Evaluate Property Loss

31-17
Mesh

Fig. 4 Area of Loss

Evaluation

31-
31-19
31-20
31-21
6g Seismic Intensity Distribution Fig. 6h Seismic Intensity Distribution

of Tottori (1943) of Nankai Eq. (1946)

31-22
31

31-23
Ii 31-24

Fig. 60 Seismic Intensity Distribution

of Yebino Eq. (1968)

31-26
Fig. 6q Seismic Intensity Distribution

of Nemuro-hanto-oki Eq. (1973)

Fig. 6r Seismic Intensity Distribution

of Ohitaken-chubu Eq. (1975)
* o * + +

• • + * •

100 Km

<9 Epicenter

lOOKm

• Subground type I

it Subground type II

•

'm

'

t ' *

| \

* * * *

J I \

' " Tokachi-oki Earthquake (1952)

!

Note : The scales of venical and

* horizontal are not equal.

Fig. 7 An Example of Subground Types

31-28
^operty Loss Ratio (Dj) 0.10

0.0724

Fig. 8 Relationship between Magnitude, Epicentral Distance

and Property Loss Ratio (Subground Type I)

,0.1739

Property Loss Ratio (Djj)

7.0

. 20

0.0072 40

60

70 80

90

0 0053 \^100

Fig. 9 Relationship between Magnitude, Epicentral Distance

and Property Loss Ratio (Subground Type II)

31-29
I LPAD \ VWP D a'

Tokachi-oki (1952)

Kita- tango / Kanto

Fukui

.—Hyuganada

O

Tokachi-oki (1968)

I ^ /

Tokachi-oki -Tottori (1968 Hokkaido Pref .)'Nngata (Niigata Pref.) Nemuro-hanto-oki Kita-izu" /

o »Oga ol Tokachi-oki 1 Nankai (Kochi Pref.) (1968 (Aomori Pref.) »

01 , - Imaichi /I

Miyagiken-hokubu

0 / o / Ohitaken-chubu Kita-tajima ShizuokaYebino

Fig. 10 (-^?

Ap

) and its Estimation

Subground Type I DijCJ. M)(90

100 , LPAD

\ ^WPD

(Z /

Fig. 11 Relationship between Magnitude, Epicentral

Distance and Equivalent Ratio of Razed House

31-30
Social Capital (82,715 billion yen)

32,876.2

bil . yen

21,827

bil. yen

20,986.3

bil. yen

7,769.8

bil. yen

Assets of Public Corporations

(39.7)

TVansnortat ion (44 2^

Electric Power (24 7^ Telecommunication (14.8)

Water Supply (13.7)

Gas Supply (2.6)

Public Service Assets

(25.5)

Education (49 4^

Public Organizations (16.6)

Relieion (12 4)

Medical Service (11.0)

Cooperation (8.1)

Public Insurance (2.5)

Public Assets

(25.4)

Highway (38.7)

Agriculture, Fishery, Forestry (26.5)

River (19.9)

Harbor (9.9)

Others (5.0)

Governmental Assets

(9.4)

Central (29.8)

Local (70.2)

( ): percentage.

Fig. 12 Breakdown of Social Capital in 1970 National Wealth Survey

31-31
Individual Loss Ratio for Building

Lr,P wd

p d

0 1

ii 01

0.00.

Yebino O

Niigata

(Niigata Pref.) 0/

Tokachi-oki (Aomori Pref.) Miyagiken-hokubu/' °

bu ^Tokachi-oki /

ONemuro-hanto-oki

Tokachi-oki

(Hokkaido Pref.) O

0.01 0.1 Loss Ratio for Property Loss LP WPn

Fig. 13 Relationship between Loss Ratios of Property Loss and of Building

Individual Loss Ratio for Highway Facility lr

0.01

0,001

Yebino NiiS?ata (Miyazaki Pref.) (N "gata Pref ->

Tokachi-oki

(Hokkaido Pref.)_ °OTokachi-oki

Tokachi-oki /(Aomori Pref.) Ohitaken-chubu ° // O ^3Nemuro-hanto-oki

OMiyagiken-hokubu

0.01 LP

Loss Ratio for Property Loss

Fig. 14 Relationship between Loss Ratios of Property Loss and of Highway Facility

31-32

i
Individual Loss Ratio for Agricultural Facility

1

0.1

0.01

0.001

Yebino Niigata Tokachl-°oki (Ni

igata Pref-) (Aomori Pref.) o

Tokachi-oki ° ©Tokachi-okt

(Hokkaido Pref.), Ohitaken-chubu O

Miyagiken-hokubu0Nemuro-hanto-oki

Fig. 15 Relationship between Loss Ratios of Property Loss and of Agricultural

Facility

Individual Loss

1

Ratios for River Facility

0.1 wrPd

0.01 -

0.001

Yebino

(Miyazaki Pref.) O

Niigata

(Niigata Pref.)'

Tokachi-oki Miyagiken-hokubu (Hokkaido Pref.)

Tokachi-oki Ohitaken-chubu ° O

Tokachi-oki

(Aomori Pref.)

Loss Ratio for Property Loss

Fig. 16 Relationship between Loss Ratios of

Property Loss and of River Facility

31-33
Individual Loss Ratio of Telecommunication Facility

1 r

L T P

T rD

0.1

0.01

0.001

0.0001

Niigata

(Niigata Pref.) O

Tokachi-oki (Hokkaido Pref.) O

Nemuro-hanto-oki O J* Tokachi-oki (Aomori Pref.)

Loss Ratio for Property Loss

Fig. 17 Relationship between Loss Ratios of Property

Loss and of Telecommunication Facility

I r

Loss Ratio for Property Loss p Fig. 18 Relationship between Loss Ratios of Property Loss and of Electric Power Facility

31-34
Individual Loss Ratio of Water Supply Facility

1 r

0.! JLffiL

0.01

0.001

Ntigata

(Nllgata Pref.)

Tokachi-oki (Aomorl Pref.) yP OYebino Nemuro- hanto - o kl >Tokachi-oki OTokachi-okl

(Hokkaido Pref.)

Ohitaken-chubu O OM iyagiken- hokubu

0.01 3.1 LP

Fig.

Loss Ratio for Property Loss

19 Relationship between Loss Ratios of

Property Loss and of Water Supply Facility

Individual Loss Ratio of Harbor Facility 10r

Wp P D

0.J -

0.01

0.001

Niigata

(Niigata Pref.) O

Nemuro-hanto-oki O

Tokachi-oki (Aomori Pref.) O

OTokachi-oki

Tokachi-oki (Hokkaido Pref.)

Loss Ratio for Property Loss

Fig. 20 Relationship between Loss Ratios of Property Loss and of Harbor Facility

31-35
Loss Ratio of Property Loss

Fig. 21 Characteristics of Individual Loss

Ratios of Various Kinds of Facilities

31-36
32

THE DISASTER-RESISTANCE OF CITIES AND THEIR

LIFELINES

Kazuto Nakazawa

Eiichi Kuribayashi

Public Works Research Institute

Ministry of Construction
i

j
The Disaster-Resistance of Cities and Their Lifelines

Until recently, the phrase "earthquakes countermeasures" largely

meant the design and construction of structures to withstand

earthquakes. Today, however, it means much more than that, and

typically includes studies of when earthquakes are likely to occur and assessment of the preparedness of cities, neighborhoods,

and escalities to motion, fire, or tsunamis brought about by earthquakes. Such research is certainly justified by the awesome power

of earthquakes. Among all natural disasters, earthquakes have

taken the greatest toll of human life and property (see Table 1).

If the highly populated central Pacific coast of Japan were to

be hit today with a quake like the infamous Kanto quake of 1923,

losses are estimated to be in the neighborhood of several tens

of trillion yen ($200 billion).

Earthquake Prediction

The subject of earthquake prediction has always been shrouded

' in controversy. Early in the twentieth century, a debate began

to grow up around two early practitioners of seismology —

Professor Omori and Professor Imamura. Imamura had predicted a large quake in the Tokyo area and insisted on countermeasures

to protect the city. Professor Omori on the other hand belittled

this prediction as being totally without scientific evidence. It is also said that Omori was anxious about needlessly panicking

the public.

Unfortunately, the Kanto earthquake occurred in 1923 with

120,000 deaths and 5 billion yen in property losses. Based on

this prediction, Imamura became the father of Japanese earthquake

prediction. He followed this success with accurate predictions

of the Tonankaido earthquake of 1944 and the Nankaido quake of

1946.

But Professor Imamura' s predictions began to fail after

World War II and the prediction of earthquakes as a scientific

discipline began to fall into disuse. Not until 1962 was earthquake prediction study seriously undertaken again. In that year, three scientists, Tsuboi, Wadachi, and Hagiwara,

outlined a new program, called the "blueprint", that would

set earthquake research back on its feet and would require an investment of several billion yen over ten years. The program's

budget rapidly increased — in 1965 to 175 million yen; and

in 1976 to 2,213 million yen. The organizational structure of

this program is shown in Figure 1.

Today, earthquake prediction depends on many types of

observations — horizontal ground strain, vertical upheavals

and subsidences, velocity changes in seismic waves, changes

in the electromagnetic properties of rocks, and changes in

ground water. Only synthesis of all this information can

result in a useable prediction. Nevertheless, as Professor

Imamura believed, "without prediction, there can be little

prevention.

Disaster Resistance of Urban Districts

The Kanto earthquake of 1923 resulted in the following losses;

where five can be seen as the predominant threat:

32-1
Burnt 447,128 houses

Razed 128,266 houses

Half Razed 126,231 houses

Lost by Floods 868 houses

By contrast, in the Tottori earthquake of 1943 and the Fuki earthquake of 1948, shaking was the culprit:

Tottori Earthquake Fukui Earthquake

Burnt 254 houses 3,960 houses

Razed 7,485 houses 35,420 houses

Half Razed 6,158 houses 11,449 houses

Lost by Floods none none

These facts suggest that while ground motion is the major threat

in small towns, fire is what consumes the cities.

But fire and shaking are not the only enemies. Just as important are the economic and social disruptions that occur

after the quakes. Indeed, Professor Toshihiko Kato of the

Social Science Institute, University of Tokyo, suggests that the

Kanto earthquake of 1923 led to a severe economic depression,

which in turn led to militarism and expansion in Manchuria and

China, and eventually to the disaster of World War II.

As a first step in ranking a city's resistance, a number of

factors dealing with the city's layout, housing stock, and

population must be investigated. They are:

Geographical and Natural Elements

Density of population and its concentration

Kinds of industry and their distribution

Vacant areas, aquatic areas Topography

Meteorological and Hydrological phenomena

Physical Properties

Earthquake Frequencies and Intensities

Geology and Soil Conditions

Aseismic and Fire Resistance of Buildings

Water and Energy Supply

Dynamic Elements

Population, Day and Night

Dangerous Facilities

Disaster-Prevention Facilities

Traffic Volume

Regional Redevelopment

Social and Cultural Elements

Communication Networks

Governmental Control Organizations

Using this kind of information, the disaster-risk potential

of each ward in a city can be profiled, as the Tokyo Metropolitan

Government did in 1974. Their first step was to find, for each of

the items shown below, the disaster risk for all 23 wards:

32-2
Accumulated risk potential for wooden buildings

Risk potentials due to local population, open

spaces, and traffic volume

Accumulated fire risk

Risk potential of spreading fires

Risk potentials for refugees

Based on this assessment, the five wards of the city at greatest

risk are as follows:

With Fire Fighting Without Fire Fighting

1st Sumida Sumida

2nd Arakawa Arakawa

3rd Taito. Taito

4th Toshima Toshima

5th Koto Shinagawa

The five safest wards were the following:

1st Minato Minato

2nd Bunkyo Nerima

3rd Nerima Setagaya

4th Chuo Chiyoda

5th Setagaya Chuo

Distribution of Earthquake Damage to Wooden Houses

Shaking Damage

How is earthquake damage distributed? The distribution of

damage in the Fukui Earthquake of 1948 is shown in Figure 2. The circles in the figure indicate places where ground

conditions are relatively sound, and the black spots indicate

places where the ground is not sound. They suffered from the

earthquake damage in ratios larger than 1%.

Fukui Earthquake of 1948 (M = 7.3, H = 20 km) Izu Peninsula Earthquake of 1974 (M = 6.8, H = 20 km)

Ebino Earthquake of 1968 (M = 6.1, H = 0 km)

In the above medium-scale earthquakes, with magnitude of

6.1 through 7.3 and depths of hypocenters of 20 km or less, the

distribution of ratios of earthquake damage to wooden houses

were statistically as follows:

D

1 1Q (0.62.5M + 0.025M«A - 0.272)

( (1)

164

DlII = 1 x 10 (0.576M + 0.026M«A - 0.248)^ (2)

where

D^j : ratios of earthquake damage to wooden houses

(in %) in relatively sound ground conditions

( I region)

,

T^lII : ratios of earthquake damage to wooden houses

(in %) in bad ground condition (II region)

32-3
. i , i . Number of razed houses + 0. The earthquake damage ratios = Total number of

x Number of half razed houses

existing houses

x 100, in %

M : Magnitude of earthquakes

A : epicentral distances (in km)

Research on the distribution of earthquake damage ratios has

been done by Professors Omote, Minakami, Uchibori, and all of these

researches find the ratio as being proportional to the exponent

of epicentral distance. In Equations (1 and 2) the magnitude

and ground conditions were taken into account. Ground condition I means layers of diluvium and tertiary; ground condition II,

alluvium.

From the equations, it can be seen that in region I the

earthquake damage ratio is smaller than in region II. Larger

magnitude brings larger earthquake damage ratios and larger

epicentral distances bring smaller earthquake damage ratios.

These relationships are illustrated in Figure 3.

When these relationships are applied to actual earthquakes,

comprehensive evaluations are possible. For example, the AnseiEdo Earthquake of 1855 had its epicenter at the northern part of

Tokyo at a location of N 35.8° and E 139.8°.

In each ward of Tokyo, the earthquake damage ratios

corresponding to earthquakes of 6.0, 6.5, and 7.0 magnitudes

were obtained as shown in Figure 4. In this figure great earthquake damage ratios are shown in the four wards of Adachi

,

Katsushika, Kita, and Arakawa near the epicenter and also the

wards of Sumida, Taito and Koto located in the delta.

In Figure 5, the distribution of the earthquake damage

ratios for the Ansei Edo Earthquake is shown. Although the

structural characteristics of houses 100 years ago are different, we can assume the thickness of the alluvium layers

as being the same as today. The earthquake damage ratios of

every ward in the area of the central part of Tokyo are compared with the actual ratios that occurred in Table 2. The three most dangerous wards are the same in both cases.

Fire Damage

At first glance, it seems that there is a correlation between

earthquake damage ratio and the fire damage of earthquakes.

However, in case of the Fukui Earthquake of 1948 there was no

such similarity, (see Figure 6).

Damage by fires is affected by the possibility of outbreaks of fires, the possibility of expanding fires, the

meteorological conditions, and fire-fighting ability. In macroscopic points of view, however, an equation of this

evaluation was developed by Prof. Kawazumi as follows:

B = 10~ 5 x N

2

, (3)

where B = number of burnt houses,

N = number of razed houses,

N = 0.2 x 10 2(M" 5)

, and M = earthquake magnitude.

32-4
In Equation (3), it is known that about one house is burnt

down at M = 6.5. However, at M = 8.0 a million houses are burnt

down. The fact that 450 thousand houses were burnt down in the Kanto Earthquake with M = 7.9 supports Prof. Kawazumi's equation.

Lifeline Systems

Relations between various facilities and governmental agencies in

charge in Japan are shown in Table 3. From this table, the

facilities related to lifelines are as follows:

Facilities Elements

Transportation - - - Road, street, railway, airport and harbour

Communication- - - - Telegram, telephone, mail, broadcasting and

newspaper

Sanitary ------ Watersupply, sewerage and solid waste disposal

Energy ------- Electricity, gas and liquid fuel Vacant area- - - - - River, lake park, greenzone and open space

All of them are indeed the indispensable facilities in and

after earthquakes. Gases and liquid fuels are needed as heat

sources, but on the other hand, they can help expand fires.

Here, a consideration will be presented about the improvement

of the disaster-resistivity of lifelines by making use of a concept

known as their residual lives. The residual life is the length of

time obtained by subtracting the already-used time from the durable

lifetime of a structure. The following is an example

relating to the earthquake resistance of bridges in the Kanto

district

.

Assumpt ions

Probability of earthquake damage (U*). Namely, the

probability of earthquake damage (complete losses of

the functions) of bridges that have existed since tb^e Kanto Earthquake of 1923 is assumed to be 2.1 x 10

(The probability of a human death during a earthquake

is supposed to be 10""^ to 10~ for reference.)

The earthquake-resistance of bridges (R) is an exponential function of time.

Durable life-time of a bridge is assumed to be 50 years.

Examples of the analyses are shown in Figures 7 and 8. Meanings of symbols in the figures are as follows,

Required capacities of bridge earthquake resistance.

Ratios between capacities retrofitted and required.

Changes of the existing capacity without any

retrofitting.

Changes of the retrofitted capacity to maintain the

level of required capacities corresponding to the

durable life of 50 years at any moment.

Changes of the retrofitted capacity to maintain the

level of the required capacities corresponding to the residual life at any moment, but with reductions

of risks being proportional to ratios of the residual

life-time and durable life-time.

li.

RoAR

Curve A

Curve B

Curve C

32-5
In Figure 7 retrofitting ratios (AR: B-A and AR 1 : C-A) are

shown in the case of satisfying initially 96 percent of required

capacity. In Figure 8 retrofitting ratios ( AR and AR' ) are shown

in the case of satisfying the full required capacity, but with

reductions of probabilities of risks (1.05 x 10 ) to half of

the assumed probability before.

Although some technical specialists and planners may reject

such a statistical approach and rely instead on their intuition,

it is clear that the approach outlined there — or one similar to

it — must underlie any future analysis that aims at thoroughness

and dependability.

References

1. Okamoto, Shynzo: "Introduction to Earthquake Engineering,"

University of Tokyo Press, 1973

2. Ishibashi, Katsuhiko: "Re-examination of a Large Earthquake

Expected in Tokai District," Autumn 1976 Meeting of Seismological Society, Draft Manuscripts of Presentations, No. 2,

October, 1976 (in Japanese)

3. Sato, Yutaka: "Problems of Earthquake Prediction Research,"

Kensetsu Geppo (Public Relations Magazine of Ministry of

Construction), Vol. 30, No. 2, Feb., 1977 (in Japanese)

4. "Planning of Earthquake Disaster Mitigation of U.S.A.

Geologic Survey," Earthquake Prediction Data 2, Science

and Technology Agency, Research and Coordination Bureau,

Living Science and Technology Section, March 1976 (in

Japanese)

5. Kuribayashi, Eiichi: "Introduction to Earthquake Engineering"

(1) through (15) The Bridge and Foundation Engineering," Vol 4, No. 2 to Vol. 5, No. 5 Feb., 1969 through May, 1970 (in

Japanese)

6. Yoneda, Shoji: "Catastrophic Circumstances Produced by Great

Earthquake," Economist Magazine , Extraordinary Number, July 7,

1975, Mainichi Newspaper Co., Ltd (in Japanese)

7. "Earthquakes" University of tokyo, University Extension, No.

24, University of Tokyo Press, Nov. 15, 1976 (in Japanese)

8. "Rikanenpyo (Science Yearbook)," Published in 1960, edited by

Tokyo Astronomical Observatory (in Japanese)

9. Kato, Toshihiko: "Chapter VI, Earthquakes and Economy," refer

to reference book 7 above (in Japanese)

10. "Promotion of Urban Seismic Disaster Countermeasures , " Ministry

of Construction, City Bureau, Kensetsu Geppo, Vol. 28, No. 11, Nov. 1975 (in Japanese)

11. "A Guide-line for Drafting Projects on Mitigation of Earthquake

Disaster in Three Major Metropolitan Areas, Tokyo, Nagoya and

Osaka (draft)," Ministry of Construction, Aug., 1975 (in

Japanese

)

12. Kuribayashi, Eiichi, jet jal. : "Distribution Characteristics of

Seismic Damage," Technical Journal of Civil Engineering , Vol. 18, No. 9, Sept., 1976 (in Japanese)

32-6
13. Umemura, Hajime, et_ _£].•: "Survey in Regard to Evaluations of

Earthquake Damage in Building" Report of Disaster Prevention

Committee of Tokyo Metropolitan Government, 1970 (in

Japanese)

14. Kawazumi , Hiroshi: "On the Seismic Damage Distribution and

Earthquakes in Tokyo," Journal of Architecture and Building

Science, No. 775, 1951 (in Japanese)

15. "Review on Disasters of the Ansei-Edo Earthquake," Tokyo

Metropolitan Government 1973 (in Japanese)

16. Kawazumi, Hiroshi: "Intensity and Magnitude of Shallow

Earthquakes" Bureau Central Seism. Intern. Ser. A, Tran.

Sci. 19, 99-114

17. Kuribayashi, Eiichi, et_ al_ . : "Stochastic Analysis on Professional Judgements on the Importance Factors in

Aseismic Design," Technical Memorandom of Public Works

Research Institute, Sept., 1976 (in Japanese)

18. Kuribayashi, Eiichi, et_ al_. : "Retrofitting Vulnerable

Structures to Earthquakes," Technical Journal Civil

Engineering , Vol. 19, No. 9, Spet. 1976 (in Japanese)

32-7
TABLE 1: HISTORICAL RECORD OF THE DEATH BY NATURAL DISASTERS

IN THE WORLD

D i sasters

Year of

Christian Era Country Number of Death

Earthquake 1556 China 830,000

Flood

and

Tidal wave

1642

Ch i na ,Break on coastal

\

embankment

300,000

Storm 1876 India 200,000

Volcanic

eruption 1693 Italy 60,000

Fi re 1871 U.S.A.

(Wood f i re)

1 ,152

TABLE 2: THE RATIOS OF HOUSES DAMAGED BY THE ANSEI-EDO

EARTHQUAKE OF 1855

ratio of houses

damaged by earthquake

place name Evaluated ratios* Observed ratio

(1) Sumida-ward 100 % or more 80

(2) Koto 8k % 76 %

(3) Taito 5A % 51 %

(k) Bunkyo kS % 68 %

(5) Chiyoda 35 % 46 %

(6) Minato Ik % 80 %

(7) Chuo 26 % 8 %

(note) * As to the structural characteristics of houses and the

thickness of Alluvium layers in subgrounds, those of

the present days are used.

32-8
32-9
32-10
Legend

o Ground , I area • Ground , 1 I area Boundary 1 ine

of Prefectures

R i ver ® Epicenter

^National Railways

Faults

DISTRIBUTION MAPS OF TOWNS AND VILLAGES

DAMAGED BY THE FUKUI EARTHQUAKE (WITH THE

RATIO OF DAMAGED HOUSES OF 1% AND OVER)

,

CLASSIFIED BY THE NATURE OF THE GROUND

(note) As to the faults, their curves are drawn

according to the report of Mr. Yoshikatsu

Ogasawara (19*»9 ): "On the Damage by the

Fukui Earthquake and the changes in the ground,

especially on the earthquake and the fault

movement", Geodetic Survey Institute Review,

Special Issue 2. 32

32-11
32-12
'itabashi W. S Adachi KiTa T"' 18 2

| W. 43. 3 25 v 14.8^ , 100.0

W.

<0. 8

\ 2. 5 Nerima W.

\ 8.4

. 7- 0 < 36. 2

\

20- 0 \ 88. 6

I _ -J ~ 18 5 H.3 v" v 43. 6 Arakwa W,

1 ^.lToshima.W.^-^.oo.o ,

I Katsushika V>' 14 8

. r"K 13.1 \ —1 — / \ Nak&no W.

s ^-i Sueinami Ws 13

Nil /

/' ?f. '43.1/

I 1. 5 <

jl - J 5.5

j" n |- ^Shibuya xJ N - N 0. 8 CMinato \ \ 8 6 ; 0 9 \ v 2. 7

I "IV 8.7 Meguro —'0. 3

' I. 1

*>

I 4 0 / Shinagawa W. v

-'

JOO. 0_.

<? 2|;lT.ito V

36. 2

Setagaya

0- 2

0. 8

3. 1

r-'4.5

i__ C 11.9

J2.9 /Chivoda.

4 5 vi^°;

L/ vv?. 1 .86. 1

Tama «. _ I * o 3 O Tokyo Bay

*u W.

0"

0. 2 ^

°

2. 0. 8 / • »

j

3- 6 Tsununi

6 Saiwai W

6. A ^

A' ( %

s

N 2.7 ^ , Asah. W. i ^ oKanagawa

\

0

Hodogaya 1.0^ w

Kawasaki City

Yokohama City

e

Legend

assumed epicenter

Boundary line of cities

Boundary line of wards

( Numerals )

Upper one: Corresponding to M= 60

Middle one: M=«-5

Lowest one ? n

FIGURE 4: DISTRIBUTIONS OF THE RATIOS OF HOUSES DAMAGED

BY HYPOTHETICAL EARTHQUAKES M=6 , 6.5 AND 7, CLASSIFIED BY URBAN WARDS OF TOKYO, KAWASAKI

AND YOKOHAMA

32-13
FIGURE 5: DISTRIBUTIONS OF THE RATIOS OBSERVED OF HOUSES

DAMAGED BY THE ANSEI-EDO EARTHQUAKE OF 1855

32-14
A The city, town and village where the ratios

of burnt-down houses are more than 1 %

( )The ratios of the burst-down houses

6

-

O N ishifujishima V.

Maruoka T. A(?o%)

Kanatsu T. A (24. 72 %)

Harue T.

Morita T. O

o

o o

o

o o° Q 080080 <h

A

Harue T. (5%)

1 -

Maruoka T. A (70%)

,Q O 1

Matsuoka T. A (5. 6%)

Kanatsu T. A(24. 72 %)

3 x 10 x D,

Morita T. A (2.43%)

8

Fukui O,

"00 o o

I. 52 %

o

oQ o

rs OO

o o

10 20 30 40 50 60 70

Ratios of houses damaged (%)

80 90 100 (%)

FIGURE 6: RELATIONS AMONG DEATH, CASUALTY, FIRE AND SHAKING

DAMAGE IN THE FUKUI EARTHQUAKE OF 1948

32-15
xRo

0 10 20 30 40 50 t(yrs)

FIGURE 7: RETROFITTING REQUIRED, CASE 1

32-16
FIGURE 8: RETROFITTING REQUIRED, CASE 2

32-17

33

RESCUE AND REHABILITATION AFTER THE IZU-OHSHIMA

KINKAI EARTHQUAKE OF 1978

Eiichi Kuribayashi

Tadayuki Tazaki

Takayuki Hadate

Public Works Research Institute

Ministry of Construction

Rescue and Rehabilitation After the Izu-Ohshima Kinkai

Earthquake of 1978

General Information on the Earthquake

Date and time of the occurrence of the main shock: 1978, January

14th (Saturday), 12h 24m

Epicenter: 139°18'E, 34°48'N, depth 0 km

Location of the epicenter: Northern part of Ohshima Town, Ohshima

island, Tokyo 26 km in east from Higashi-Izu Town, located

along the east coast of the central part of the Izu Peninsula

Magnitude in Richter Scale: M=7.0 nominally, containing +1/4 of error Intensity distribution in JMA intensity scale: Shown in Figure 1-1,

but 6 is considered the upper bound.

Intensity of JMA 5 and 6 corresponds to an 8 or 9 on the Modified

Mercalli Intensity scale.

Focal Mechanism: A right lateral fault developed from the epicenter

to almost western direction; lift zone was observed in the

northern side.

Severest after-shock: 1978, Jan. 15th, 7h 32m a.m., M=5.7, the

epicenter is estimated as 138°54'E, 34°53'N and 10 km in depth

after Earthquake Research Institute, Tokyo University (see

Figure 1-2) Fore and after shock area: Fore and after shocks are distributed

in a belt from Ohshima Island to the western coast of the Izu Peninsula across the middle of the peninsula as shown in

Figure 1-3.

Fore-Shocks

From 36 to 55 minutes past 9 o'clock on January 14, four relatively

big shocks (4th degree of JMA's intensity scale) were observed at Ohshima Meteorological Station.

In response to this information, the Shizuoka Prefectural

Government sent the following wireless message to municipal governments in Shizuoka Prefecture at 11 o'clock: "The Japan Meteorological

Agency informs us that occurrences of earthquakes and some damage

are suspected in the very near future."

Weather in the Damaged Area

Before and after the earthquake, temperature changes and rainfall

at Shimoda City, adjacent to the severest damaged area of Kawazu

| Town and Higashi-Izu Town, are shown in Table 1—1. In this table,

the mean temperature is 8.0 degrees centigrade and the total rainfall is 50.5 mm during the whole of January, whose average is 7.5

and 66.1, respectively.

An Outline of the Disaster

In Kawazu Town and Higashi-Izu Town lifeline systems consisting of

roads, railways, buried pipe lines, communication lines and power

lines suffered damage from landslides and surface ruptures. Little

damage occurred in buildings and houses. Incredibly, most of the

damage occurred in two soilfill dykes for slime from a gold mine

located at a point about 40 km from the epicenter of the main

shock, along the upper stream of the Kano River.

In the limited district of Shizuoka Prefecture, the amount

of the direct losses is estimated as 39.3 billion yen or more;

33-1
the indirect is still being estimated (refer to Table 1-2).

Districts other than Shizuoka Prefecture suffered little damage.

Occurrences of Historical Earthquakes in the Izu Peninsula , their Geography and Geology

Since the seventh century A.D. , 14 earthquakes on the Izu Peninsula

and its adjacent area were recorded — 6 times in the twentieth

century, and 3 times stronger than 5th degree in this decade. On August 18, 1976, Kawazu Town was hit by an earthquake of Richter

Magnitude 5.4 and Maximum Intensity 5.

The Izu Peninsula is covered with igneous rocks, and steep

slopes with airslaked soils are seen along coasts and rivers.

The whole of the peninsula is very sensitive to atmospheric,

hydraulic, and geomechanical effects, especially heavy rain falls,

which have frequently brought disasters (see Figure 1-4 and

1-5).

Damage and Losses

Most of the damage to buildings and houses resulted from landslides

and surface ruptures; some damage occurred due to ground motions.

As for the value of articles inside houses, they have been estimated

as one third of the value of a house or three million yen for a house. Damage and losses are shown in Table 2-1.

Damage and losses to agricultural products, farm yards, forests,

fishing ports, water sources for agriculture, water ways, roads

for agriculture, canning plants for fruits and vegetables are shown

in Table 2-2. Damage and losses to schools and play grounds are shown in Table 2-3. Damage and losses to hospitals, medical stations, relief facilities, and water supply systems are shown in

Table 2-4.

Principal industries in the severest damaged area are tourism

and agriculture. Damage to industries is shown in Table 2-5. Damage to roads is shown in Figures 2-1 to 2-4 and Table 2-6.

The failure of two dykes of Mochikoshi Gold Mine resulted in

polluted river water from sodium cyanide contained in the slime.

The effect to water quality is shown in Table 2-7.

Great damage occurred in railway facilities between HigashiIzu and Kawazu railroad stations on the line operated by Izu Kyuko

Railroad Co. Ltd. The damage is shown in Table 2-8.

There was no substantial damage to telephone systems except

for cutting of an aerial wire caused by falling rocks in Kawazu

Town. However, during about a half day after the main shock in

all districts of the Izu Peninsula telphone service was not available. This problem was the result of countless receivers which had

been shaken off their cradles due to the earthquake motion.

As for electrical power systems, minor damage occurred in two power plants, two transformer stations, two channels, and 24 towers

for transmission lines, as shown in Table 2-9. In the severest

damaged area there are liquid petrolium gas cylinders for domestic

use, areal gas supply service, and kerosene storages for vehicles,

ships, and machinery. Their damage was very slight.

33-2
Slight damage occurred in facilities for police and fire

stations, but their activities were not disturbed.

The indirect losses are listed in Table 2-10. Losses due to this earthquake are compared with those of the Niigata Earthquake

of 1964 in Table 2-11.

Countermeasures for the Emergency in and after the Earthquake

Shizuoka Prefectural Government Office and municipalities set up

< the Rescue and Rehabilitation Center as shown in Table 3-1, following the Regional Disaster Prevention Plan of each body. Figure 3-2

and 3 are the structures of the Center of Shizuoka Prefectural

Government Office and Shimoda City Office.

At the same time, the prefectural center, headed by the

I Director of the Shimoda Tax Office, was established in Shimoda

(see Figure 3-2). A forward base of the Branch was organized on

Jan. 17 at Kawazu municipal office. Twenty people led by an

I; accounting superintendent (a position second only to a vice governor) were dispatched to Kawazu. They assisted with information

j collection, rescue, and rehabilitation in cooperation with the

Kawazu officials.

Collecting information about the damaged facilities overwhelmed the capacity of municipal offices and lower prefectural

organizations. Several prefectural officials were also dispatched

to Izu district. Three search groups were sent to the Prefectural

Construction Office in Shimoda.

The Governor of Shizuoka Prefecture asked for the help of the

j Self Defense Forces (SDF). Several corps, including 400 men of the

34th regiment, moved to Izu on January 14. Some of them arrived

at the damaged sites that night. They rescued unidentified and

j injured persons, and removed the fallen soil and rocks from slopes

beside roads and behind houses. Table 3-2 shows the work of SDF.

Fire self-defense squads and community systems (private organi1 zations) also contributed greatly to the rescue and rehabilitation,

i Table 3-3 shows the performance of these squads. Table 3-4 shows

j the list of other organizations related to rescue and rehabilitation.

Although the main shock took place at 12:24, 24 minutes after

business hours, people who were on their way home quickly came

back to the office and set up the emergency system.

When the Prefectural Government Office had transmitted the

information about the foreshocks, which JMA had issued, to branch

offices of prefectural government and municipal offices, Shimoda

City Office announced it to its personnel in the building. But Kawazu Town regarded the message as having very little urgency,

based on its previous experience with such warnings.

Refuge and Rescue

The damage brought by the earthquake was mainly caused by landslides. Moreover, there existed a possibility of relatively big

aftershocks in an area that was dotted with the dangerous cliffs

and slopes. Issued recommendations are listed in Table 3-5.

Some people in the districts other than those in Table 3-5

made their own decision to evacuate. They had detected the danger

33-3
of landslides. Table 3-6 shows numbers of evacuees and refuge

places. Private residences were also used as refuges.

Restorat ion

Water supply facilities consist of municipal and areal water supplies, The scale of the latter is smaller. Although the water

pipes sustained severe damage, the filtration plant did not. As shown in Table 3-7, municipal and areal water supplies were restored by Jan. 26 in Higashi-izu Town and Kawazu Town.

Drinking water shortage was not a severe problem during the

suspended period. One reason is that each municipality usually

prepares water wagons and can supply drinking water if only the

water source survives the earthquake. The other is that the

water wagon of the Ground Self Defense Force and the water

boat of Maritime Self Defense Force took drinking water to the

disaster area.

When the dykes of the sludge reservoir at Mochikoshi gold

mine gave way, the sludge with sodium cyanide held there (about

58,500 m ) flowed into the Mochikoshi river and its main river,

the Kano river. Izu-nagaoka Town Office was forced to stop

taking water from the Kano river. For several weeks they were using existing wells and narrow streams as their water source.

Telecommunication facilities of Japan Telegraph and Telephone

Public Corporation (JTTPC) were restored by Jan. 17, except for

49 circuits in Kami-kawazu, Kawazu Town. The entire circuit was back in operation by Jan. 25 (see Table 3-8).

The electric power was cut off to 15,000 residents after the

earthquake. Tokyo Electric Power Co. Ltd dispatched 888 people

from the 14th through 16th of Jan. to repair the problem. It was finished by 19:15, Jan. 15 (see Table 3-8).

Shimoda City was the only one having areal gas supply service

in Izu territory. The gas supply company in Shimoda City examined

all gas pipes before resuming the service. Most houses in the area other than Shimoda City were installed with a liquid petroleum gas (LPG) cylinder. The gas company specialists inspected whole

cylinders preceding the reopening on Jan. 17.

Resumption of Traffic and Traffic Control

The skelton of the highway in the Izu Peninsula consists of National

Highway routes 135 and 136 and a Principal Prefectural Road, Shuzenj

Shimoda Line, part of which are toll roads. The railroad network

is comprised of the Ito Line (JNR, Atami-Ito), Izu-Kyuko Line (IzuKyuko Railroad Co. Ltd, Ito-Shimoda) and Izu-Hakone Railroad Line

( Mi shima-Shuzen j i ) (see Figure 3-4).

Most of the damage to highways was generated by landslides.

The damaged sites were too concentrated in the mountainous area

to perform the rehabilitation all at once. The works were started

from both sides of the damaged route and went forward to the middle.

Table 3-9 shows the traffic control. The numbers in the table

correspond to those in Figure 3-4. The site indicated by X was

still out of service as of Feb. 12. Out of the major highways,

Route 136 first reopened at 12:00, Jan. 16. As for the route on the east coast, both the Higashi-izu Toll Road and ex-national

highway route 135 administrated by Japan Highway Public Corporation

33-4
(JHPC) and Prefectural Construction Office run in parallel (Figure

3-5).

Priority was given to the Toll Road even though the restoration

of the ex-national highway was obliged to be delayed. The ex- national highway was still suspended a month later because of the

heavy damage, as shown in Photo 1. The Higashi-izu Toll Road was

out of service from 14th through 16th of January. It opened from

17th through 29th exclusively to emergency vehicles only in the

daytime. It was shut again from Jan. 30 through Feb. 9 for

rehabilitation. Since Feb. 10 it has resumed with some one-lane

sections. The Principal Prefectural Road, Shuzenji-Shimoda Line

'sustained extensive damage, where a tourist bus was buried in a landslide and three passengers were killed. The Prefectural Construction Office and SDF made a great effort to restore the road,

Ibut it was still out of service a month later.

Most damage to railroad facilities was to the Izu-kyuko Railroad Co. Ltd. The line was stopped after the earthquake and

reopened as:

Ito-Izu-kogen 12:30, Jan 15 Izu-ogen-Katase-shirate 12:20, Jan 17 Katase-shirata-Inator i 6:55, Jan 30

Kawaxu-Shimoda 6:30, Jan 31

The section between Inatori and Kawazu is still blocked by the

tunnel whose lining was crushed due to the active fault.

Inatori and Kawazu are connected by bus service.

The facilities of Japanese National Railroad (JNR) suffered

no damage. However, according to its own regulations, JNR was

forced to stop services and examine its facilites. The resumed

time of each line was as below:

New Tokaido Line ( superexpress) 16:55, Jan 14 Tokaido Line 15:15, Jan 14 Ito Line 15:00, Jan 14 Minobu Line 12:40, Jan 14 Futamata Line 14:50, Jan 14

Collection and Transmission of Information

Collecting information on the damage itself and the demand of goods

and machinery for rehabilitation is a substantial problem in the

beginning of an emergency. Most of the information on the damage

to dwellings, water supply, electric power, and telecommunications

facilities was gathered by midnight of Jan 14 by the administration

and public organizations. Kawazu Town Office, the most heavily

damaged area, could not finish, however, until the next day.

For example, Shimoda City Office dispatched 38 people immediately after the earthquake to investigate damage. They collected

the number and scale of damage in a few hours and transmitted them

to the headquarters of the Prefectural Government. Detailed

investigation continued from the 15th through the 18th of January.

Community systems were well organized and each citizen knew

his duty. This contributed effectively to the collection of

information.

33-5

33
Collecting the information on the damage to roads was difficult,

because the traffic route itself was blocked everywhere. For the

Principal Prefectural Road, entering by automobile was impossible.

The search party of the Prefectural Construction Office investigated

the damaged sites on foot. It took 3 days to outline the damage

and 3 more days for detailed investigation.

Three people of the Prefectural Government Office flew by SDF

helicopter over the damaged territory on the afternoon of Jan 14. However, they could not see details because the pilot was not accustomed to the highway routes and his altitude was relatively

high.

Some information on the damage of highways was brought in by

the road monitors. When the tourist bus was buried in the landslide at Nashimoto, Kawazu Town was informed by one of the

monitors living beside the Principal Prefectural Road, ShuzenjiShimoda Line.

Telephone network of JTTPC was the principal means of communication in the emergency. However, for a few hours after the earthquake the demand for telephone calls exceeded the system's capacity.

During this period, emergency wireless — eleven circuits in Shizuoka

Prefecture — was used to communicate between Prefectural Government

Office and municipal offices. Prefectural Government has planned

to increase the circuits to 35 by October 1978.

An exclusive circuit was installed temporarily between the

Prefectural Construction Office and Shimoda Branch in Shimoda Tax Office after the earthquake.

Other Measures

The collapsed dykes on the Mochikoshi and Kano Rivers were repaired

temporarily with double-steel sheet piles. The toxic sludge was removed from the riverbed and deposited in abandoned mines. It was scheduled to be finished by Feb 20 (see Photo 2).

Recons t rue t ion

Law Systems for Relief

The law systems applied by the relevant administrations are listed

below

:

(1) Disaster Relief Law

Disaster Relief Law was applied in;

Higashi-izu Town 19:30, January 15 Kawazu Town 19:30, January 15

(2) Designation of Violent Disaster

Designation of Violent Disaster is under consideration

for the time being by the Cabinet. The damage to agriculture

and minor enterprises in Higashi-izu and Kawazu Towns will

possibly be designated. Only in Kawazu Town will it possibly

be designated for public works.

(3) Loans

People whose dwellings were totally or half destroyed by

the earthquake were qualified to have a disaster relief loan

from the Housing Loan Corporation. Explanation meetings were

held in Kawazu and Higashi-izu Towns on Jan 28 and 29. Tourism

33-6
enterprises received loans from the People's Finance

Corporation and Minor Enterprises Finance Corporation.

(4) Exemption of Tax

The fixed property tax of the razed houses was exempted.

Enterprises that suffered some damage were exempted from the

enterprise tax, real estate acquisition tax, auto tax, and

automobile acquisition tax. The restaurant tax was also

deferred. Compulsive car inspection was deferred for a month

in the damaged district.

(5) Aid

Shizuoka Prefectural Government decided to aid those

people whose residential land was considered suceptible to further landslides when its preventive cost would exceed

500 thousand yen. The Government assumed half this expense,

which totalled 53,900 thousand yen.

Reconstruction Work on Public Utilities

Reconstruction work on civil engineering, argicul tural , and educational facilities have been performed or are being prepared.

Table 4-1 shows the public structures qualified for the national

subsidy. Ministry of Construction urgently qualified the first

39 reconstruction sites for the subsidy from 15th through 17th

of February. Other qualifications by the ministry will follow.

Other Articles Concerning Reconstruction Measures

Izu-kyuko Railroad Co. Ltd has considered reconstruction of the

Inatori tunnel to resume service as early as possible. The

tunnel had been constructed in a circular section as illustrated

in Figure 4-1. An interior lining inside the existing one is

the most likely method of reconstruction.

As for the Mochikoshi mine, a Mochikoshi Mine Reconstruction

Countermeasures Committee headed by Prof. H. Yamaguchi of Tokyo

Institute of Technology had been established since Jan 28.

Specific Matters Referring Experiences of Past Disasters

The Izu Peninsula has suffered frequent natural disasters. The

Kano-gawa typhoon of 1958 killed 1,269 people. As for the past

five years, it has sustained:

Izu-hanto-oki earthquake

Flood

Heavy rain

Heavy rain

Kawazu earthquake

Heavy rain

May 1974

July 1974

July 1975

July 1976

Aug 1976

Oct 1976

People and administrative personnel in the district were thus well

prepared for this disaster. Installing water wagons in each municipality was a typical example of preparation.

The Fuji-Hakone National Park on the Izu Peninsula is occupied

by many tourists all year around. In the flood of July 1974,

several traffic routes were blocked and tourists could not leave

Shimoda City. They rushed into Shimoda Port causing some panic.

Remembering this experience, tourists were ordered after the earthquake to remain in their hotels until transportation was assured.

The municipal offices asked for or hired boats from SDF , Maritime

33-7
Safety Division, and private shipping companies to transport them

to Atami and Ito Cities. Table 5-1 shows the number of tourists

brought out by ships.

Because Route 135 on the east coast was damaged, the Prefectural Construction Office put a priority of reconstruction on its

by-pass, a section between Shimoda City and Kawazu Town of the

Principal Prefectural Road, Shuzenj i-Shimoda Line.

Shimoda City has experienced many tsunamis (tidal waves) as shown in Table 5-2. The City Office has designated refugee places

high up on the hills. They have also marked the anticipated water

level of tsunamis on electric power poles to draw the attention

of citizens.

Japan Meteorological Agency (JMA) issued the tsunami warning

12:31, Jan 14. However, Shimoda City is located only about 30 km

from the epicenter and the first wave of the tsunami was estimated

to reach about 10 minutes after the earthquake. There would be no

time to evacuate. As a matter of fact many citizens decided on their own to evacuate to the designated refuge places. The observed

tsunami was about 10 cm and the warning was canceled at 14:15.

Operation of Organizations

The Kawazu office of the Prefectural Government was exceptionally

organized for this disaster. The Kawazu office, headed by an accounting superintendent of the prefectural government, controlled

rescue, rehabilitation, and reconstruction throughout the damaged

area. A meeting was held every morning at Kawazu office attended

by staff from relevant municipalities, branch offices of the prefectural government, Higashi-izu Toll Road Office (JHPC), Izu-kyuko

Railroad Co. Ltd, SDF, and police.

Lessons from this Disaster

In this disaster, the quick collection of information allowed the

government to grasp its extent and severity. As for the highways,

it took three days to grasp the damage.

It is recommended in Japan, where mountains are steep and many

trunk lines of transportation are concentrated on a narrow flat

strip of land, to develop an effective and quick way to survey

the damage.

Telecommunication networks functioned satisfactorily except

during one confused period for a few hours. Expanding and

strengthening emergency networks should be encouraged.

Shizuoka Public Safety Commission (SPSC) was ordered to stop

traffic where the road was blocked by landslides. But passes were granted. The owner of a vehicle who wanted to get such a pass

applied to SPSC. When it was given, a seal had to appear on the

front windshield. But the number of vehicles qualified reached

7,395, which disturbed the rehabilitation and reconstruction of

road to some extent.

Lessons for Organizational Systems

The accounting superientendent sent to the disaster-hit area had

supervised various organizations in emergency relief. He had

visited Izu territory five times for previous disaster relief.

33-8
He knew the area in detail. This contributed a lot to the

rehabilitation and reconstruction after this disaster.

Other Lessons from the Earthquake

The organziat ional systems to cope with natural disasters have

been satisfactorily developed in Japan. Shizuoka Prefectural

Government had improved its organization system and information

network, and had stored daily goods for the emergency, especially

because a seimologist had warned of the possibility of a "Great

Tokai Earthquake" to occur within Suruga Bay [10]. The Government

had newly established its Earthquake Countermeasures Section in

1977. These measures created an awareness of what to do in earthquakes. For example, no fire broke out after this earthquake. This

is because most people turned off their gas, even though some of

them got burned doing it.

Most roads in the Izu Peninsula are constructed by cutting

and banking on a steep slope. The possibility of landslides along

the roads is considered to be higher than that of any other territory in Japan. To improve the situation the road administrator is

planning to construct a new disaster-proof highway consisting mostly

of long-span bridges and tunnels. This plan is quite reasonable

if all the disaster countermeasures are used.

References

1. Rika-Nenpyo (Science Yearbook): Tokyo Astronomical Observatory,

Maruzen Co., Ltd, 1978.

2. Preliminary Report of the Izu-Ohshima Kinkai Earthquake of

1978: Ohshima Meteorological Station, Tokyo Meteorological

Observatory, Jan 21, 1978.

3. Executive Office of the President, Office of Emergency

Preparedness: Report of the Congress, Disaster Preparedness,

Jan 1972.

4. J. Haas and R. Ayre: the Western Sicily Earthquake of 1968,

National Academy of Sciences, 1969.

5. R. Dynes, J. Haas and E. Quarantelli: Some Preliminary

Observations on Organizational Responses in the Emergency

Period after the Niigata, Japan, Earthquake of June 16, 1964,

Disaster Reserch Center, The Ohio State University, 1964.

5 and 7 A case Study on the Rescue and Reconstruction after An Unexpected Disaster - The Eruption of Mt. Usu, 1977 - : Technical Memorandum of Public Works Research Institute,

No. 1292 (in Japanese) and No. 1293 (in English), 1978

8. Classified Maps of Subground in Shizuoka Prefecture: Development Bureau, Economy and Planning Agency, Prime Minister's

Office, 197L

9. R. Wallace: Goal, Strategy and Tasks of the Earthquake Hazard

Reduction Program, Geological Survey Circular 701, 1974

10. Katsuhiko Ishibashi: Re-Examinations of Estimated Great

Earthquake in Tokai Region - About the Great Earthquake in

Suruga Bay -, Preprint of Autumnal Meeting, the Seismological

Society of Japan, 1976.

33-9
11. The Outline of Road Works, Shizuoka Highway Works Office,

Ministry of Construction, 1977.

12. Protective Treatments to the Izu-Ohshima Kinkai Earthquake,

Numazu Construction Office, Ministry of construction, 1978.

13. Lists of Density of Cyanide from Jan 14 to Feb 5 in 1978.

14. Geometrical Plane of Stock Pile for Slime of Mochikoshi

Mine in Houzuki Azawa Swamp.

15. Geometrical Plane of Sabo Works along Kano River.

16. Outline of Road and River Works, Numazu Construction

Office, 1977 [12-16: provided by Numazu Construction

Office, Chubu Regional Construction Bureau, Ministry of

Construct ion]

.

17. The Izu-Oshima Kinkai Earthquake of 1978, and Damage,

Feb 14, 1978.

18. Occurrence of Sensible Earthquakes in the Izu-Ohshima

Kinkai Earthquake of 1978.

19. The Outline of Damage and Losses to the Izu-Ohshima

Kinkai Earthquake of 1978, Feb 13, 1978.

20. Lists of Damage of Public Facilities Caused by the IzuOhshima Kinkai Earthquake of 1978, Feb 13, 1978.

21. Damage of Railroad of Izu-Kyuko Co. Ltd. Caused by the

Izu-Ohshima Kinkai Earthquake of 1978.

22. Rehabilitation of Roads, Jan 15, 1978.

23. Traffic Control after the Izu-Ohshima Kinkai Earthquake of

1978, from Jan 14 to Jan 16.

24. Record of Reports of Air Reconnaissance.

25-. Report on Enforcement of the Traffic Control Based on the

Disaster Countermeasures Basic Law.

26. Dangerous Zones for Landslide and Slope Failure.

27. Manuals for the Headquarters of Countermeasures Against

Disasters of Shizuroka Prefecture.

28. Map of Jurisdiction of Shizuoka Prefectural Government.

29. A Guide to Earthquakes

[17-29: provided by Main Office of Shizuoka Prefectural

Government ]

.

30. Record of the Izu-Ohshima Kinkai Earthquake of 1978.

31. Information Transfer Systems for Earthquake Prediction

Information in the Shimoda Branch 1977.

32. Manuals of the Shimoda Branch for Countermeasures against

Disasters

.

33-10
, 33. Manuals in Details for Generalizing Corps in Shimoda Branch.

34. Refuge and Inducement of Tourists from Jan 14 to Jan 15.

35. The Activity of the Fire-brigade

[30-35: provided by Shimoda Tax Office, Shizuoka Prefecture] .

36. Lists of Damage Caused by Disasters in Fiscal Year of 1974.

37. Information File of Roads.

38. Disposition Systems for Disasters in Emergency.

39. Designated Dangerous Zone for Landslide and Slope Failure.

40. Tidal Wave Investigation on the Ansei Earthquake of

Nov 4, 1860.

41. Map of Jurisdiction of Shimoda Construction, Shizuoka

Prefecture.

42. The Izu-Ohshima Kinkai Earthquake of Jan 14, 1978 and

Distribution Maps of Public Works Damaged, Prefectural

Works and Municipal Works

[35-41: provided by Shimoda Construction Office,

Shizuoka Prefecture]

.

43. Report of Earthquake Damage and Losses (Jan 31, 1978).

'44. Distribution of Damage in Shimoda Town.

45. Lists of Rehabilitation of Public Facilities.

46. Emergency Disposition System.

"47. Great Tidal Wave of the Ansei Earthquake of 1860.

48. Earthquake Disaster Mitigation Program of Shimoda City 1977:

Ordinance of Regional Disaster Prevention Conference and its

References

.

49. Public Information on Tidal Waves, Oct 20, 1977.

50. Explanatory Data on Budget of Municipal Gov. of Shimoda

City Jan 20, 1978

[43-50: provided by City Office of Shrimoda]

.

1 51. Outline of Water Services Managed by Amagi-Yugashima Town.

; 52. Map of Jurisdiction of the Amagi-Yugashima Town

[51-52: provided by Amagi-Yugashima Municipal Office],

!'&3. Outl ne of Damage and Losses caused by the Izu-Ohshima

Kinkai Earthquake of 1978, Feb 4, 1978.

54. Manuals of Headquarters for Countermeasures against

Disasters of Kawazu Town.

55. Outline of Kawazu Town, 1976

[53-55: provided by Municipal Office of Kawazu Town],

[

Ip6. Earthquake Frequency of Fore-shocks and After-shocks,

Jan 30, 1978.

33-11
57. Estimated Losses on the Izu-Ohshima Kinkai Earthquake

of 1978, Jan 15, 1978.

58. Manuals of Headquarters for Countermeasures against

Disasters of Higashi-Izu Town.

59. Outline of Higashi-Izu Town, 1977

[56-59: provided by Municipal Office of Higashi-Izu Town],

60. Distribution of Damage along Higashi-Izu Toll Road,

Administration Office of Higashi-Izu Toll Road, Japan

Highway Public Corporation.

61. Damage and Rehabilitation Caused by the Izu-Ohshima Kinkai

Earthquake of 1978, Feb 2, 1978.

62. Outline of Izu-Kyuko Co. Ltd.

[61-62: provided by Izu-Kyuko Co. Ltd],

63. Damage of Electric Power Facilities caused by the IzuOhshima Kinkai Earthquake of 1978

[63: provided by Tokyo Electric Power Co. Ltd.].

33-12
Questionnaire Used in Interviews with the Personnel of

Administrative Organizations

1. Outline of the Damage

2. Organization Response of the Administration

3. Damage of Urban and Housing Facilities

4. Damage of Hydraulic Facilities

5. Damage and Rehabilitation of Lifeline System

a. Highway

b. Railroad

c. Electric Power

d. Telecommunication

e. Water Quality

f. Water Supply

g. Fuel

5. Financial Aids, Loans and Tax Exemptions

7. Losses Valued in Money

B. Other Related Comments

33

33-13
Photo 1. Failure of Ex-national Highway Route 135 Caused by a Landslide in Higashi-izu Town

33-14
33-15
Table 1—1. Weather Condition before and after the Mam Shock

(after Shimoda Municipal Government)

Date Weather Temperature in atmosphere (Centidegree) Rain fall mm/day

Max. Min. Mean

Jan. 1 © 20.5 14.0 15.1 1.0

2 @ © 14.5 11.0 9.8 4.0

3 • CD 12.0 7.0 7.3 8.5 4 © 12.5 5.5 7.5 0

5 CD 11.0 5.5 7.3 0

6 0 13.5 6.5 9.8 0

7 CD 15.5 5.5 9.5 0

8 • CD 13.5 7.0 9.2 1.0

9 © 14.5 6.0 10.3 0

10 © 10.5 2.0 5.9 0

11 0 13.5 2.0 6.5 0

12 0 15.5 7.0 10.3 0

13 © 13.5 7.0 9.2 6.0

14 • 14.5 9.0 9.3 7.5

15 © 17.5 7.0 12.3 1.0

16 @ 14.5 6.0 8.8 0

17 © 9.5 6.0 6.9 0

18 • 13.0 6.0 6.8 1.0

19 © 12.0 8.0 7.8 0 20 © 13.0 3.0 7.1 0

21 © 13.0 5.5 7.1 4.0

22 © 9.5 2.0 4.5 0

23 © 12.0 0 5.1 0

24 12.5 1.8 6.8 7.0

25 © 11.0 2.0 5.9 0 26 12.5 4.0 6.8 5.5 27 © 15.0 6.0 9.8 0.5 28 ® 12.5 6.0 9.3 1.0 29 8.0 4.0 4.9 0 30 • © 11.5 2.0 5.8 2.5

31 © 10.0 3.0 5.4 0

Average 8.0°C Total 50.5 mm

Note: Weather: © fine. 6 cloud, # rainy

33-16
Table 1—2. Damage and Losses

(as of Feb. 15, 1978, Shizuoka Prefecture)

Classification CVdlUdlcU LOSS

(in 1 ,000 yen) Note

Direct Damage 38,709,501

Buildings and Houses 5,838,400

Agriculture and Fishery 6,892,490

Facility for Education 494,838

Facility for Welfare and Sanitary 972,783

Industrial Facility 8,407,582

Public Works 14,375,408

Facility for Electricity 228,000 Not Official

< Railroad 1,500,000 Not Official

Indirect Damage Unidentified Delay of Resumption, etc.

Total 38,709,501

I

33-17
33-18
I

I

CM

1 ro

CM

CMCM

c oCM

O (T>

CD

<

S

-r

B V i' j

O

33-19
33-20

33-22
Eo

[CJOl X

o~ to

nOt«nNO

r 1 M

0

00

000

-1- r 1 t

in

1 n/in^BM

OC 1 t

oc

•3 ILUBJV - - "3-

A oiuc>i

1— t—ri

OOO**

On"i — c

rNi

X nznuiBuiix

On

CnJ Or^t On

D°il

"~" 0m

0

Ofl 1,166 m

Iioi

O0 OO0

"* ONO

rj- — "3- - — mOn "t

O- 0r- — ON

X nzirqsiN

r- rr inO

00OfN

tnfN

c*mOn 0Cn|

3 Bpotuiiis

(N m

r^i

r

00

r— 00 On

CNl m

X nzEA\E)i 0 nom

NO

in

NO ONr00

On

t00 3,581

00

tX EiuiiJSESnXiSBiuv

Cn| c0m

(N

X nziriiSEgiH 0oc ocm

ON

1,627 1,739 1,760 7,001

NO >n

1

/

ca /

/ O /

-J /

/ c

/ -2 / «

/ —

/ C

/ 00 / s / 0

Mouse Family

Sufferer House Family

Sufferer House Family

<u iH

u.3

1

House 3

0

Totally

Razed

House

Half

Razed

House

Partially

Destroyed

House Public

Building

Others

Non-dwelling

Dwelling

0 0

0 00 O r-' 00"

cn) m r-0. m

33-23
Table 2-2. Damage and Losses in Agriculture and Fishery

Classification Number of Damage Loss in thousand yen Note

Agriculture 20.8 ha, 1,706 4,969,518 Crops, Agricultural land, Facility for agriculture

Forestry 19.9 ha, 101 1,623,972 Forest, Facility for forestry

Fishery 49 299,000 Facility for fishing port

Total 40.7 ha, 1,856 6,892,490

Table 2—3. Damage and Losses in Facilities for Educations

Classification Number of Damage Loss in thousand yen

Kinder Garten 12 24,972

Primary School 22 136,603

Junior High School 13 107,355

Senior High School 11 201,250

Hutte 1 1,030

Public Hall 5 1,918

Gymnastic Facility 1 10,000

Cultural Estate 5 11,610

Library 1 100

Total 71 494,838

Table 2—4. Damage and Losses in Facilities for Welfare and Sanitary

Classification Number of Damage Loss in thousand yen

Hospital 61 523,288

Water Supply Services 327,747

Solid Waste Disposal Plant 2 20,885

Hospital for Epidemic 1 200

Others 54 100,663

Total 972.7X3

33-24
Table 2-5. Damage and Losses in Industries

Classification Number of Damage Loss in thousand yen

Commercial 1,699 2,566,697

Mining and Industry 474 558,773

Tourism and Service 1,053 5,282,1 12

Total 3,226 8,407,582

Table 2—6. Damage and Losses in Roads

Classification Number of Damage Loss in thousand yen

Shizuoka Prefectural Government 283 10,632,771 Shimoda City 31 42,037 Kawazu Town 269 1,165,990

Higashi-izu Town 123 234,000

Prefectural Highway Public Corporation 16 102,000

J.H.P.C. 31 1,700,000

Total 753 13,876,798

33-25
Table 2—7 Water Pollution Due to Sodium Cyanide in Kano River

(in PPM)

Mizunuki Bashi Endo Bashi Chitose Bashi Kurose Bashi Time Density Time Density Time Density Time Density

Jan. 14 18:50 ND

18:15 ND

Jan. 15

9:45

10:30

13:25

1.34

ND

I .ZD

9:50

10:30

14:20

ND

ND

U.U 1 o

Jan. 16

9:40

13:20

0.19

0.14 .

10:25

13:55

0.13

0.14

Jan. 17

9:35

14:20

ND

ND

10:20

15:10

ND

ND

Jan. 18

10:25

14:15

153 ND

143 ND

11:45

15:10

61 ND

78 ND

Jan. 19

10:30

14:25

97 ND

76 ND

12:00

15:00

50 ND

56 ND

Jan. 20

8:00 2,230 8:00 153 9:45

14:10

100 ND

62 ND

10:35

14:50

43 ND

41 ND

Jan. 25

8:0C 615 8:00 89.3 9:30

14:05

30 ND

30 ND

10:15

14:40

14 ND

14 ND

Jan. 30

8:00 666 8:00 46.7 10:00

14:10

20 ND

15 ND

10:35

14:45

10 ND

13 ND

Feb. 4

9:00 325 9:00 21.0 9:20

13:50

22 ND

12 ND

10:00

14:30

7 ND

10 ND

Note: upper, suspended solid (SS) lower, density of sodium cyanide (CN~'

)

ND: not detected

33-26
33-27
Table 2-9 Damage and Losses in Facilities for Electric Power

Classification Number nfDamape Loss in thousand yen

Power Plant 2 Transformer Station 2 Water Channel 2 Tower for Transmission Line Several

Total 228,000

Table 2-10 Indirect Losses

Classification Indirect loss Note

Aorin il ti i re Decrease of agricultural product by

damage of agricultural land and agricultural facilities

Industry

Trade Decrease of income by

interruption of business

Industry Decrease of product by

interruption of work

Tourism Decrease of income by

interuption of work and decrease of tourist

Electric Power Decrease of income by

interruption of power supply

Railroad Decrease of income by

interruption of service

Table 2—1 1 Damage and Losses in Comparison between the Izu-Ohshima Kinkai Earthquake of 1978 and the Niigata Earthquake of 1964

^-^Classification Direct Loss (bil. yen) Indirect Loss (bil. yen) Indirect/Direct Eq.

Izu Ohshima Kinkai 39

Niigata 130 122 0.94

33-28


33-31
Legend

Higashi-izu Toll Road O Cracking on Road Surface ( 14) £ Slope Failure(]7)

Boundary ol Ciiy and Town

Higashi-izu Town

Kawazu Town

Damaged Section

i Higashi-izu Toll Road

3 km

Fig. 2-4 Damage Distribution of the Higashi-izu Toll Road

33-32
33-33
Table 3- 1 Established Date of Rescue and Rehabilitation Center in each Administration

Administration Established Date of Rescue and Rehabilitation Center Shizuoka Prefectural Government Office 1 1 n 1 1 a rv 1 ,1 j u 1 1 ud i y i *+

,

14 10 Shimoda Office, do I a ni i a rv 1 4 1

4

10 Kawazu Office, do January 1 7, Higashi-izu Town Office January 14, 13 10 Matsuzaki Town Office January 14, 12 45 Kawazu Town Office January 14, 14 10 Shimoda City Office January 14, 14 50

Table 3-2 Dispatched Self Defense Forces to Damaged Area

Destination Ground, Maritime, Air Period Mandays

Vehicle-days

National Highway Route 136

(Restoration of the Highway) Ground Jan. 14 680 Tractor Shovel (2) Bulldozer (7) Dump Truck (3

)

Power Shovel ( 7) Amagi-yugashima Town

(Restoration of the Highway) Ground Jan.l8~22 743 (49)

Mitakairiya, Kawazu Town

(Searching the Missings) Ground Jan.l5~29 1 1,364 Tractor Shovel (10)

Trailer (18) Bulldozer (77) Rnlldo7Pr fnr Soft Ground (48) Dump Truck (23

)

Power Shovel (93) Wrecker (40) Drugshovel (131) Others (535 ) Oil Pressure Shovel (2) Drain Pump

Fire Engine

Nashimoto, Kawazu Town

Onabe, Kawazu Town

(Rescue for the Bus Accident)

Ground Jan.l5~17 2,344 Wrecker (6) Water Trailer (39) Food Supply Car (2 1

)

Ambulance

Others (416) Yugano, Kawazu Town Ground Jan. 18 50 Dump Truck (3

)

Power Shovel (2) Others (1) Numanokawa. Kawazu Town

(Watching the Waterlevel of a Stream

)

Ground Jan.20~21 40 Chain Saw

Others (4) Kawazu Town

( Water Supply

)

Ground Jan.l8~22 139 Water Wagon (22)

Filtrstion Hcjuipmcnt Kawazu Town

(Information Exchange) Ground Jan. 26-29 490 Aircraft (5 )

Mochikoshi, Amagi-yugashima Town

(Rehabilitation oi the Dyke) Ground Jan. 15-18 305 Helicopter (60) Bulldo/cr (8) Trailer (6) Power Shovel (4

)

Dump Truck (8) Others ( 14) Naramoto, Higashi-izu Town

(Restoration of the Highway Rescue

at a Destroyed Hotel)

Ground Jan. 15-19 1,078 Power Shovel (13) Bulldozer (40) Wrecker (4) Trader Shovel (4

)

Dump Truck ( 64 ) Air Compressor ( 9

)

Inalori-ko, Higashi-izu Town

(VVater Supply)

Maritime Jan. 14-24 2.349 Guardboat (4) Troopship (13)

Speu.il Ship (2)

l/u (Vrritory

( 1 ran.sportation, Survey )

Air Jan. 14—15 X Aircraft (2)

33-34
r—c

Performance

Rehabilitation

Soil

Removal

from

Road

Repair

Water

Pipe

Investigation

of

Drinking

Water

Rescue

at

Landslide

Site

Reinforcement

to

Kawazu

T.

Per- sons SOOs

rJ

•t- cc

sn OsT

SOc

S3

Performance

Rehabilitation Rehabilitation

Water

supply

Rescue

at

Landslide

Site

Rehabilitation Reinforcement

to

Kawazu

T.

Survey

of

Damaged

Area

Rehabilitation

Per- sons ocs o

m oc>CI © so<N

rJan.

15

Performance

Rehabilitation Rehabilitation

Water

Supply

Rescue

at

Landslide

Site

Reinforcement

to

Kawazu

T.

Survey

of

Damaged

Area

Rehabilitation

Clearing

Up

the

Damaged

Area

Survey

of

Damaged

Area

Investigation

of

Dangerous

Site

Per- sons CO oCO Os

CI

sn

"* oSOc

oo osO

1,602

c

Performance

Dissemination

of

Tsunami

Alarm

Survey

of

Damaged

Area

Rescue

at

Takami

Hotel

Dormitory

Rescue

at

Landslide

Site

of

Higashiizu

Toll

Road

Rehabilitation

Rescue

at

the

Collapsed

Building

Rescue

at

Landslide

Site

Dissemination

of

Tsunami

Alarm

Watching

Tidal

Level

Survey

of

Damaged

Area

Rehabilitation

Per- sons omm ooo

"t oo

oo

©sO

CI

Shimoda

C. 1

Higashi-izu

T.

Kawazu

T.

Minami-izu

T.

Matsuzaki

T.

Nishi

izu

T.

KamoV. Total

33-35
Jan.

21

Performance

Water

Supply

Rehabilitation

Patrol

Rescue

Draining

Persons r-m

Os©

SOSO

Jan.

20

Performance

Water

Supply

Rescue

at

Landslide

Site

Per- sons oom

o OO

Osc

cd

Performance

Water

Supply

Rescue

at

Landslide

Site

Persons 00 ro

oo

00c

<73

Performance

Dissemination

of

Recommendation

of

Refuge

Traffic

Control

Soil

Removal

from

Road

Assistance

for

the

Removal

of

Lvacuees

Investigation

of

Drinking

Water

Rescue

at

Landslide

Site

Patrol

Per- sons w-i or-r l

00

SO C I Os

33-36
Tabic 3 4 Man-days Pcrsucd for Rescue and Rehabilitation of Organizations

" ________ Item

Organization " -_ Period Man-days Vehicle-days

Police Jan. 14 1 1.551 2,341

j Fire l ighting Squal

Fire Defense

Jan. 14 ~ Jan. 30 8,069 -

1 Local Public Official

1 Jan. 14 ~ Jan. 20 471 —

: Ground

i Jan. 14 ~ Jan. 30 23,065 4,268

Self Defense Forces j Maritime

f Jan. 14 ~ Jan. 24 2,315 23

i Air Jan. 14 - Jan. 15 8 2

Maritime Safety Division Jan. 15 - 2 JTTPC Jan. 14~ Jan. 18 1,180 303 Tokyo Electric Power Co. Jan. 14 ~ Jan. 23 2,164 644 LPG Association Jan. 14 ~ Jan. 23 381 106

Shizuoka Gas Co. Jan. 14 ~ Jan. 24 388 127

Total 49,592

Vehicles 7,789

Ships 25

Aircrafts 2

Table 3—5 Recommendations of Refuge

Municipality District Issued Date of Recommendation

Reason

Matsuzaki Town Uchino 19:30, Jan. 17 Possibility of landslide

Matsuzaki Town Ikeshiro 19:30, Jan. 17 Possibility of landslide

Shimoda City Takeyama in Kakizaki 15:00, Jan. 18 Possibility of rock fall

! 33

33-37
33-38
Table 3 --7 Rehabilitation of Water Supply Facilities

Area

Supplying

Population Capacity Restored Date bmcrgency Measures Shimoda City Water Supply

29,982 Ochiai 25,000 m-Vday

Kawauchi

2,000 m 3 /day

Jan. 17, 17:00 Water Wagon of 1.5 t(l)

Higashi-izu Town

Water Supply

14,498 Shirata River 20,000 m 3 /day

Jan. 24, 17:00 Water Boat (2),SDF

Water Wagon (22) Kawazu Town

Water Supply

6,302 Ozeki Mine

Sasahara

1 1,500 m3/day

Mitakairiya

500m 3 /day

Jan. 20, 14:00 Water Wagon (7)

Minami-izu Town

Water Supply

5,662 Ishii5,159 m

3 /day

Jan. 14, 20:00

Higashi-izu Town

Okawa Areal Water Supply

1,095 225 m

3 /day Jan. 15, 15:00

Kawazu Town

Kami Areal Water Supply

1,768 360 m

3 /day Jan. 21, 15:00 Water Wagon (1)

Kawazu Town

Onabe Area! Water Supply

245 60 m

3 /day Jan. 21, 12:00 Water Wagon (1)

Kawazu Town

Kamisagano Areal Water Supply

279 70 m

3 /day Jan. 19, 15:00 Water Pool (1)

Kawazu Town

Nashimoto Areal Water Supply

320 100 m

3 /day Jan. 26 Water Pool, 500/ (2)

Matsuzaki Town

Water Supply

519 6,200 m

3 /day Jan. 17, 18:00

Nishi-izu Town

Water Supply

8,500 7,500 m

3 /day Jan. 18, 18:00

Nishi-izu Town

Shirakawa Areal Water Supply

160 60 m

3 /day Jan. 16, 18:00

Nishi-izu Town

Miyagahara Areal Water Supply

178 45 m

3 /day Jan. 17, 17:00

Nishi-izu Town

Yoshinohata Areal Water Supply

160 85 m

3 /day Jan. 16, 18:00

Kamo Village Ukusu Areal Water Supply

1,760 635 m

3 /day Jan. 16, 15:00

Kamo Village Arori Areal Water Supply

2,311 465 m

3 /day Jan. 17, 18:00 33

33-39
Table 3—8 Restoration of Electric Power, Gas and Telephone Facilities

Municipality

Number of Residents out of Services Restoration

Electric

r oweI

Shimoda City Kawazu Town

Higashi-izu Town

}> 15,000

J

Restored temporarily by Jan. 15, 19:15

LPG

Kawazu Town 2,800 Inspected the overturned gas cylinder by

Higashi-izu Town 5,200 Jan. 17

Gas Supply Shimoda City 2,000 Inspected the gas pipes by Jan. 14

Higashi-izu Town 160

Telephone

Kawazu Town

Shimoda City

ItoCity

Nishi-izu Town

14030

30

70

Restored temporarily by Jan. 16 except Nashimoto, Kawazu Town by Jan. 25

Table 3-9 List of Traffic Control

Traffic Control Number Route Site Cause Started Date

Rescinded Date

for Emergency

Vehicles

Retrieval

1 Route 135

(Ex-national Highway)

Akazawa, Ito City Landslide Rock FallJan. 14, 13:00 X

2 Okawa, Higashi-izu Town

Landslide Jan. 14, 13:00 X

3 Atagawa, Higashi-izu Town

Landslide Jan. 14, 13:00 Jan. 14, 19:00

4 Kurone, Higashi-izu Town

Landslide Jan. 14, 13:00 X

5 . Kurone, Higashi-izu Town

Landslide Jan. 14, 13:00 X

6 Route 136 Kami-lunahara Amagi-yugash ima Town

Landslide Jan. 15, 8:00 Jan 15. 9:40 Jan. 15, 17:00

7 Yagisawa, Doi Town Landslide Jan. 15, 8:00 Jan. 15, 9:40 Jan. 16, 12:00

8 Kominc. Doi Town Landslide Jan. 15, 8:00 Jan. 15, 9:40 Jan. 16, 12:00

Fukada. Kanio Village Landslide Rock Fall

Jan. 15, 8:00 Jan. 16, 12:00

33-40
(Table 3-9 continued)

Number Route Site

T r.i ffic Control

Cause Started Date

Rescinded Data

for Emergency

Vehicles

Retrieval

10

11

12

13

14

15

16

17

18

19

20

21

22

23

24

25

26

21

P.P.R. ShuzenjiShimoda

P.P.R. ShimodaMatsuzaki

P.P.R.

Ito- Nishiizu

P.R.

Ito -

Kawana—

Yawatano

P.R. Atagawa—

Katase PR.

Yugashima —Matsuzaki

PR.

Shimodako

P.P.R. ShuzenjiShimoda

Route 135

(Higashi-izu Toll Road)

Yoichizaka, Amagiyugashima Town

200 m far from Toll Gate

more 200 m far from

Toll Gate

Kawaaino, Kawazu Town

Akebuse, Matsuzaki Town

Yokokawa, Shimoda

City

Miyagahara, Nishi-izu Town

Yaemeno, Nishi-izu Town

Mochikoshi, Nishi-izu Town

Kokushi-toge, Naka-izu Town

Futo, Ito City

Atagawa, Higashi-izu

Town

Onabe, Kawazu Town

Ikeshiro, Matsuzaki Town

Benten-bridge, Shimoda City

Sakagawa, Kawazu

Okawa, Higashi-izu

Town

Naramoto, Higashi-izu Town

Rock Fall

Landslide

Landslide

Landslide

Rock Fall

Rehabilitation Works

Landslide

Landslide

Landslide

Landslide Rock Fall

Landslide Rock Fall

Landslide

Landslide

Landslide

Subsidence

of Bridge

Rehabilitation Works

Slope Fail- ure

Slope Fail- ure

Jan. 14, 13:00

Jan. 14, 13:00

Jan. 14, 13:00

Jan. 14, 13:00

Jan. 14, 13:00

Jan. 16, 1 1:00

Jan. 14, 14:00

Jan. 15, 8:00

Jan. 14, 13:00

Jan. 15, 9:30

Jan. 14, 13:00

Jan. 14, 13:00

Jan. 15, 8:00

Jan. 15, 8:00

Jan. 14, 13:00

Jan. 17, 9:00

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33-41
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Secretariate General Affairs Division Environment Division

Social Welfare Division Sanitary Division Industry Division Agriculture Division Labor Division Public Works Division Urban Division Accounting Division Public Corporation Division Education Division Council Division Personnel Affairs Division Inspection Division Regional Labor Division Transportation Division Tokyo Office

Police Division

Branch Shimoda

Atami Numazu

Fuji Shizuoka

Fujieda Iwata Hamamatsu

Fig. 3-2 Structure of Rescue and Rehabilitation Center of Shizuoka Prefectural Office

Director

(Governor)

-Vice Director (Vice Governors Head of Prefectural

Police Headquater

Accounting Superintendent

,

Relevant Organizations

Director Vice Director (Mayer) (Vice Mayer)

I Secretary Division — General Affairs Division

Social Welfare Division — Economy Division — Agriculture Division — Sanitary Division — Construction Division

Water Supply Division — Education Division — Fire Defense Division — Accounting Division

Fire Self Defense Squad

Fig. 3 3 Structure of Rescue and Rehabilitation of Shimoda City Office

33-46

Fig. 3-5 Route of 135th National Highway

33-48
National Highway Route 135 (Toll Road)

Fig. 3—6 Abstract of the Landslide at Route 135

33-49
33-50

Table 5-1 Hmcrgency Conveyance of Tourists by Ships (as of Jan. Hand 15)

Departure Place •jinn \ |i in (l\A/HPr wwni.1 ri(*st inn lion Member

of crew

Shimoda City

Self Defense Forces Itn/

1 okai Shipping C o. A tami

Tokai Shipping Co. 1

1 n1 Idntr\ri UN 1

Izukyu Shipping Co. lnatori 121

Total 1

1 Zj**

Higashi-izu Town

l^iil/wM n n mmnn \ c\ lillKyU JMlippiilg V.U. lto 300

Self Defense Forces lto 70

TaI/'' 1 UNuli «JSninnino 1 1 1 UL7H 1^1 inV U - Atam i 450

Fishing Boat Kitagawa ~

Yawatano 50

Total Rid

Matsuzaki Town Izu-hakone Railroad Co. IN UllldZU

Nishi-izu Town Izu-hakone Railroad Co. Numazu 04U

Grand Total 3146

Table 5—2 Tsunamis Having Affected Shimoda City

Date Epicenter Magnitude

Maximum Water

Level Observed (m) Dead Carried Away by Tsunami

Half-de- stroyed Houses

Hooded

Houses SunkShips

Feb. 3, 1605 Off Nankaido Off Boso Peninsula

7.9

Dec. 31, 1703 Sagami Bay 8.2 27 332 160 81

Oct. 28, 1707 Tokaido, Nankaido 8.4 Shimoda 5.6 Okakata 4.4 Kakizaki 6.71

1

857 55 93

Dec. 23, 1854 OffTokaido 8.4 Shimoda 5.7 122 841 30

about 30

Dec. 24, 1854 Off Nankaido 8.4 Okakata 4.8 Kakizaki 6.7

Sept. 1, 1923 Sagami Bay 7.9 Kakizaki 4.6 Sotoura 4.1

2 9 334

Dec. 7, 1944 Kumano-nada 8.3 Suzaki-Town 2.1 Daiku Town 1.9 Kakizaki 2.5

186 86

Dec. 21, 1946 Off Nankaido 8.1 245 26

Nov. 5, 1952 Off Kamchatka

Peninsula

8.3 Shimoda 1.5 120

May 24, 1960 Off Chile 8.3-8.5 Shimoda 1.14 38

33-52
APPENDIX

I. Historical Earthquake Records in Shizuoka Prefecture after Documents No. 19

Date and year Epicenter Magnitude

Nov. 29, 684 A.D. OffTokaido Very large

July 4, 715 Totomi Area 6.4 818 Sagami Bay 7.9

July 30, 841 Izu Area 7.0

Aug. 22, 887 Nankaido Area 8.6

Oct. 29, 937 Izu Area -

Dec. 11, 1096 OffTokaido 8.4

1406 Suruga Area -

Sep.r 11, 1498 OffTokaido 8.6

Sep. 10, 1510 OffTokaido -

March 21, 1589 Off Ohigawa Riv. 6.7

Feb. 4 ; 1605 OffTokaido 7.9

1626 OffTokaido - M arch 1. 1633 Sagami Bay 7.1

Sep. 10 1699 -

3

1

1 703 Sagami Bay or off Boshyu -

Jan.

->z. 1 /U4 8.2

Oct. 28, 1707 Tokaido or Nankaido Area 8.4

Sep. 18, Pll

Dec. 23, 1854 (Ansei-Tokai) OffTokaido 8.4

Dec. 24, 1891 Middle of Ohigawa Riv. -

Sep. 1, 1923 (Kanto) Off Odawara 7.9 Feb. 21 ~ May 17, 1930 Off Ito (Swam of ho Ear.) Nov. 26, 1930 (Kitaizu) North of Izu Peninsula 7.3

July 11, 1935 (Shizuoka) Shizuoka City 6.7

Dec. 17, 1944 (Tonankai) Ciult or Kumano Q -3 O.J

Apr. 20, 1965 Estuary of Ohigawa Riv. 6.1 May 9, 1974 (Izuhanto oki) Off Izu Peninsula 6.9

Aug. 18. 1976 (Kawa/.u) East of Izu Peninsula 5.4

33-53
II. Questionnaire Used in t he Interviews to the Personnel of Administrative Organ i/at ions

1. Outline of the Damage

2. Organizational Response of the Administration

3. Damage of Urban and Housing Facilities

4. Damage of Hydraulic Facilities

5. Damage and Rehabilitation of Lifeline System

a. Highway

b. Railroad

c. Electric Power

d. Telecommunication

e. Water Quality

f. Water Supply

g. Fuel

6. Financial Aids, Loans and Tax Exemptions

7. Losses Valued in Money

8. Other Related Comments

33-54
34

NATIONAL SCIENCE FOUNDATION ACTIVITY

EARTHQUAKE HAZARD MITIGATION

Charles C. Thiel

J. Scalzi

National Science Foundation

Washington, D.C.

National Science Foundation Activity in

Earthquake Hazard Mitigation

This paper presents a brief review of grant ac tivity undertaken

by the Earthquake Hazards Mitigation program from January 1977

through March 1978. Information given on each activity includes

the title, principal investigator, institution, and a brief

description of the research to be undertaken. The objective

for each element of the program is briefly stated.

The NSF's Earthquake Hazards Mitigation program has expanded

substantially in the past year from an $8.8 million expenditure

in FY 77, ending September 30, 1977, to an $18 million research

program in FY 78 beginning October 1, 1977. This paper presents

a brief review of the objectives of the current program and of

sach individual award made between January 1, 1977 and March 30,

1978. Table below shows the allocation of program funds among

he three components of the program for the past two years and

he projection for FY 79

Sudcet Current

Program Subelement Actual Request Plan Estimate Di fference

FY 1977 FY 1973 FY 1978 FY 1979 FY 1979/78

Siting $3,020,412 $6,000,000 $6,856,600 $9,800,000 $2,943,400

Design - - — 4,517,390 9,000,000 8.750,000 11 ,800,000 3,050,000

Policy 1,732,492 6,500,000 2.500,000 4,800,000 2,300,000

TOTAL $9,270,294 $21 ,500,000 $18,106,600 $26,400,000 $8,293,400

The objectives of the Earthquake Hazards Mitigation program

are to develop methods and techniques that can provide effective

protection for man, his works and institutions from life loss,

personal injury, property damage, social dislocations, and

Bconomic and ecological disruption associated with potential

Dr realized earthquake hazards.

The Siting category of the Earthquake Hazards Mitigation

urogram provides research support for the elucidation of the

ihysical basis of earthquake energy generation and the

:ransmission and propagation of the generated shock waves

:hrough various geologic and soil conditions; with the impact

)f earthquake ground motion on structures; and with development

|pf criteria and guidelines for the mitigation of potential impacts

4>n the built environment. The specific objectives of the Siting

urogram are to:

0 Develop a comprehensive data base on the nature of

earthquake motion at typical construction sites and

for representative structures;

0 Establish the physical basis for characterizing the

nature of earthquake motions and the dynamic forces

generated by such motions and other natural hazards;

0 Develop capabilities for prediction of the magnitude

and frequencies of ground motion;

0 Develop a methodology for qualitative and quantitative

estimates of local or regional risk associated with

earthquakes and other types of hazards and combined

hazards

;

34-1
° Develop a comprehensive and unified program to improve

geotechnical engineering practices applicable to soil

dynamics, foundation design, failure and instability, and

other aspects of earthquake ground motion; and

0 Identify procedures for integrating informatin on natural

hazards into land use planning, urban and coastal zone planning, offshore engineering and siting procedures.

The specific objectives of the Design program are to:

° Improve the characterization of earthquake and natural

hazard loadings necessary for the economical design of

structures subject to dynamic loading;

° Develop new methods of analysis and design of buildings

and structures of all types which will take into account

nonlinear and inelastic behavior of materials;

° Develop methods to assess the hazard potential and risk

assessments applicable to existing structures and

facilities, and devise innovative methods for improving

performance within economically acceptable bounds;

° Obtain information for engineering analysis and design

by observing the damage of facilities following actual

earthquakes, and incorporate this information into

standard design practice;

0 Develop improved computational capability for dynamic

analysis of structures and facilities and improve user

access to any computer software which is developed;

° Develop model standards and design criteria for design

of structures and facilities subjected to earthquake

and natural hazard loadings; and

° Conduct detailed studies of the behavior of smaller

nonengineered structures and secondary components of

buildings to improve recommended minimum analysis and

design guidelines.

The specific objectives of the Policy program are to:

° Expand the base of knowledge on alternative social

adjustments to earthquakes;

° Identify the social, economic, political, legal, and

related factors which facilitate or hinder the

adoption of both social and technological solutions

to earthquake hazards;

° Facilitate the beneficial use of earthquake hazards

mitigation measures by devising effective techniques

for disseminating information to the public and to decision-makers at the local, State, and national

levels; and

° Investigate measures which will reduce possible negative

social, economic, and political consequences of earthquake

predictions and warnings.

If you wish to be apprised of the grant activity in this

program on a quarterly basis please notify one of the authors to be placed on our mailing list.

The following text presents the summary of awards in each of

the individual program element areas: siting, design, and policy.

34-2
iting

inderground Lifelines in a Seismic Environment; Melvin L. Barron;

Feidlinger Associates, 110 East 59th Street, New York, New York

.0022; Grant #76-80982; Supplemental Award.

Study the behavior of underground lifeline structures such as pipelines, conduits, and channels in seismic environments.

Apply the technical knowledge gained with cost-benefit and

optimization studies in order to plan, design, and construct

such structures.

larthquake Ground Motion Modeling for the Central United States;

:obert H. Hermann; Dept of Earth and Atmospheric Sciences; St.

,ouis University, St. Louis, Missouri 63103; Grant #76-20875;

lew Award

.

Conduct theoretical and observational studies to re-examine

the available data base and improve the empirical model for

predicting ground motion. Predict strong motion parameters

and time histories in the distance range of up to 500

kilometers for earthquakes with a magnitude equal to or stronger than five, based on the study.

iperation of the National Program in Strong Motion Instrumentation;

:obert B. Matthiesen; U.S. Geological Survey, 345 Middlefield Road,

lenlo Park, California 94025; Grant #76-82207; Renewal Award.

Develop and operate a national program in strong—mot ion

instrumentation and data management with respect to network

design, network operations, and data management. The data

development forms the basis for analysis, design, and

construction of structures of all types to be earthquake

resistant in order to mitigate the losses of life and

property

.

jnduced Seismicity at the Nurek Reservoir, Tajikistan, USSR;

'avid W. Simpson; Lamont-Doherty Geological Observatory, Columbia

hiversity, Palisades, New York 10027; Grant #01092; New Award.

Study the induced seismicity near a reservoir. Investigate

spatial and temporal variations in the seismicity and their

relationship to water level. Perform studies on the

frequency characteristics of earthquakes and various

geological and geophysical effects to the induced seismic

nat ion.

eismic Vulnerability, Behavior and Design of Underground Piping

(ystems; Leon R. L. Wang; Department of Civil Engineering,

ensselaer Polytechnical Institute, Troy, New York 12181; Grant

77-10390; New Award

Develop a systematic way of assessing the adequacy of

existing water/sewer distribution systems. Determine

the earthquake vulnerability of the systems and study

the cost-effectiveness of new designs to various magnitudes

of earthquakes.

artial Support of the Committee on Seismology; Joseph W. Berg;

ational Academy of Sciences, 2102 Constitution Avenue, N.W.

,

ashington D.C. 20418; Grant #76-24167; Renewal Award.

34-3
Provide Earth Sciences reserch data needs imput into RANN

program planning for its program in strong-motion

inst rumen tat ion

.

Processing and Analysis of Oroville Earthquake Aftershock Ground

Motion Records; Thomas C. Hanks; U.S. Geological Survey, 12201

Sunrise Valley Drive, Reston, Vriginia 22092; Grant #76-81816;

New Award

Process the large set of earthquake records obtained

after the Oroville earthquake of 1975 with the data

arranged into a four volume set. Carry out an analysis

of this data to develop methods for predicting potential

ground motions of specific sites taking into account

frequency band and variables which are of engineering

interest. The survey is processing 100 aftershock

accelerograms, analyzing data in terms of ground motion

parameters, body wave spectra, seismic source parameters,

and performing studies on synthesis of ground motion time

histories. This project is being pursued in conjunction

with the California Institution of Technology.

Processing and Analysis of Oroville Earthquake Aftershock Ground

Motion Records; Donald V. Helmberger; Department of Geology and

Planetary Sciences, California Institute of Technology, Pasadena,

California 91104; Grant #76-21652 ; New Award

Process the large set of aftershock records obtained after

the Oroville earthquake of 1975 with the data arranged into

a four volume set. Carry out an analysis of this data

to develop methods for predicting potential ground motions

of specific sites taking into account frequency bands and

variables which are of engineering interest. More specifically, it is processing long period data in terms of velocity

and displacement time histories of 40 seismograms, analyzing

data for the elasto-dynamic response of half space to faulting

motions and generating synthetic motion histories. This

project is in conjunction with the U.S. Geological Survey

( above )

.

United States-Republic of China Cooperative Research in Earthquake

Engineering-Part II; Le-Wu; Department of Civil Engineering, Lehigh

University, Bethlehem, Pennsylvania 18015; Grant #77-07470; New

Award

Conduct joint research on ground motion studies and seismic

response of embedded structures (siting problems) and on strengthening concrete buildings with differential

foundation settlements (design problems) which will be

beneficial to both countries. This portion of the research

will deal with earthquake resistance and strengthening of

concrete buildings with foundation settlement and partial

structural damage.

United States-Republic of China Cooperative Research in Earthquake

Engineering-Part I; Joseph Penzien, Department of Civil Engineering,

University of California, Berkeley, California 94720; Grant #76-

06006; New Award.

Conduct cooperative research on ground motion studies and

seismic response of embedded structures (siting problems)

and on strengthening concrete buildings with differential

foundation settlements (design problems) which will be

34-4
beneficial to both countries. This portion of the research

will study attenuation laws, maximum acceleration and other

ground motion characteristics in order to find common features,

correlations, and discrepancies.

^search Initiation-Behavior of Buried Conduit Structures Subjected

!> Seismic Loading; Jack Rosenfarb; Department of Civil Engineering,

jrexel University, Philadelphia, Pennsylvania 19140; Grant #77-06421;

;w Award.

Investigate the behavior of buried conduit-type structures such

as pipelines and tunnels. Assess the effects of structure

flexibilities as influenced by conduit geometry, material, and

connection restraint conditions.

Drkshop on Research Priorities for Geotechnical Earthquake Engineerlig Applications; Kenneth H. Stokoe, III; Department of Civil

igineering, University of Texas, Austin, Texas 78122; Grant #77-

9861; New Award: and Felix Y. Yokel; Center for Building Technology,

ational Bureau of Standards, Washington D.C. 20234; Grant #77-

4157; New Award.

This joint project is to organize, administer, and conduct a workshop on Geotechnical Earthquake Engineering Applications.

It is also to review needs and priorities for future research

and establish liaison among the various research groups.

arthquake and Strong-Motion Instrument Development and Array

esign; Francis T. Wu; Department of Geophysical Sciences,

tate University of New York, Binghamton, N.Y. 13901; Grant #76-

3897; New Award.

Provide strong-motion instrument coverage in Eastern United

States. Collect data for resolving the nature of strongmotion waves and for future predictive use in engineering

design. Study the dynamic seismic source mechanisms by

using near field data.

orkshop on Research Priorities for Geotechnical Earthquake

ngineering Applications; Felix Y. Yokel; Center for Building

echnology, National Bureau of Standards, Washington D.C.

0234; Grant #77-14157; New Award.

Analyze midwest convective cloud data, synthesize the

results, and assess the potential for effective modification to augment precipitation.

Strong-Motion Seismography Array in Northern Baja California

lortheastern Sonora, Mexico; James N. Brune; University of

California, San Diego, LaJolla, California 92037; Grant #75-

'2939 A01.

Install and maintain a timed strong-motion seismograph

system array in Northwest Mexico near the Cerro Prieto,

Imperial, Agua Blanca and San Miguel Faults.

Earthquake Response of Dams Including Hydrodynamic and Foundation

Interaction; Anil K. Chopra; University of California, Berkeley,

California 94720; Grant #76-80073 A01.

Develop reliable techniques, including computer programs,

for earthquake analysis of concrete gravity, concrete arch,

and earth dams.

34-5
Seismic Response of Three-Dimensional Dam Reservoir Systems;

Allen T. Chwang; California Institute of Technology, Pasadena,

California 91125; Grant #77-16085.

Determine the seismic response of a dynamically coupled,

three-dimensional dam reservoir system including the

hydrodynamic interaction effect, the flexibility effect,

the effect of phase variation and spatial attenuation

of seismic waves, and the effect of the side confinement

of the dam.

A Portable Vibrating Structure for Soils Investigations; Paul

Ibanex; Applied Nucleonics Company, Inc., Los Angeles, California

90024; Grant #77-19653.

Develop a method for high strain in situ measurement of soil

propert ies

.

International Workshop on Strong-Motion Earthquake Instrument

Arrays; Wilfred D. Iwan; California Institute of Technology,

Pasadena, California 91101; Grant #77-24726.

Organize and convene an International Workshop on Strong

Motion Earthquake Instrumentation Arrays (three dimensional)

in the spring of 1978. The workshop will provide a basis

for development of a plan for the future deployment of

dense, three-dimensional strong-motion earthquake

resistant arrays.

Distant and Local Tsunamis in Coastal Regions; Jiin Jen Lee; University of Southern California, Los Angeles, California 90007;

Grant #77-01599.

To conduct research into the behavior of tsunamis in

coastal regions with complicated topography for use by

coastal zone engineers and planners for efficient design

of coastal structures and communities.

Soil-Structure Interaction with Arbitrary Seismic Environment;

John Lysmer; University of California, Berkeley, California 94720;

Grant #76-23277 A01.

Develop a theory and associated finite element code CREAM

(Complex Response Earthquake Analysis Method) and make

this available thorough the NTIS and other appropriate

channels

.

Earthquake Response and Aseismic Design of Underground Piping

Systems; Teoman Ariman; University of Notre Dame, Notre Dame,

Indiana 46556; Grant #77-23236.

Investigate the earthquake response and earthquakeresistant design of underground piping systems

utilized in energy transport, particularly buried

gas pipelines; study the failure of pipes by fracture

and other failure mechanisms, and develop design guides

for piping systems subject to seismic hazard.

Underground Lifelines in a Seismic Environment; M.L. Baron; Weidlinger Associates, 110 E. 59th Street, New York, New York

10022; Grant #76-09838.

34-6
Study the behavior of underground lifeline structures such

as pipelines, conduits, channels, etc., in seismic

environments, and apply the technical knowledge gained in

conjunction with cost-benefit and optimization studies to

the planning, design and construction of such structures.

Seismological Analysis of Strong Motion Records; Bruce Bolt;

University of California, Berkeley, California 94720; Grant #78-

02650.

Examine the consequences of using effective peak acceleration

as a scaling parameter for response spectra rather than the

dislocation pulse; conduct wave analysis of strong-motion

records; and search for common physical features in terms

of source mechanisms, fault rupture, and so forth.

Numerical and Experimental Study of Earthquake Strong-Motion;

James N. Brune; University of California, San Diego, La Jolla,

California 92037; Grant #77-23829.

Investigate earthquake ground motion characterization in terms of fault proximity, fault type, models and laboratory

experiments; calibrate the computer methods for modeling

earthquakes, simulate ground motion under a variety of

conditions, and produce rules for characterizing ground

motion which are suited to the needs of design engineers.

Seismic Response of Three-Dimensional Dam-Reservoir Systems;

Allen T. Chwang; California Institute of Technology; Pasadena,

California 91125; Grant #77-16085

Evaluate the seismic response of a dynamically coupled,

three-dimensional dam reservoir system including hydrodynamic interaction effect, flexibility effect, effect of

phase variation and spatial attenuation of seismic waves,

and the effect of the side confinement of the dam.

Statistical Investigation of Engineering Seismology; Leon Knopoff;

University of California, Institute of Geophysical and Planetary

Physics, Los Angeles, California 90024; Grant #77-24742.

Introduce physics into the stochastic models of earthquake

occurrence, source, and faults, and develop basic scientific

procedures for the seismic risk estimation of a given region.

Dynamic Soil Structure Interaction; Jose M. Roesset; Massachusetts

Institute of Technology, Cambridge, Massachusetts 02139; Grant #77-

' 14174

Earthquake Engineering of Large Underground Structures; Roger

Scholl; John A. Blume and Associates, 130 Jessie Street, San

Francisco, CA 94105; Grant #77-06505.

Evaluate the current state-of-the-art in underground earthquake engineering practice, determine areas in which

additional research is needed, and produce a report

summarizing world-wide experience concerning underground

earthquake engineering design and construction practices.

Induced Seismicity at Nurek Reservoir, Tadjikistan, USSR; David W. Simpson; Columbia University, Palisades, N.Y. 10027; Grant #77-

01092 A01.

34-7
Study the induced seismicity near the Nurek reservoir;

investigate spatial and temporal variations in seismicity

and their relationships to water level; perform studies on

the frequency characteristics of earthquakes and various

geological and geophysical effects of the induced seismic

motion; study the earthquake spectra using digital recorders.

Strong-Motion Accelerograph Network in the Los Angeles Basin; TaLiang Teng; University of Southern California, Los Angeles,

California 90007; Grant #77-16983.

Install 100 accelerographs in a well-planned grid-shaped

array in the Los Angeles Basin to record future earthquake

data; use the data to study the distribution of strong

shaking, the attenuation patterns, the shear velocity

structure, and the effect of geological structure and

local conditions.

A Methodology for Optimal Strong-Motion Instrument Locations in

Building Structures; F.E. Udwadia; University of Southern

California, Los Angeles, California 90007; Grant #77-07903.

Develop a scientific methodology for the location of

strong-motion instruments in buildings for maximum possible

information about the structure's dynamic characteristics;

study both the number and position of instruments in relation

to the amount of information obtained.

Prediction of Earthquake Resistance of Structures; P.C. Wang;

Polytechnic Institute of New York, 333 Jay Street, Brooklyn, New

York 11202; Grant #78-09990.

Continue development of a minimum method to generate design

earthquakes, including development of critical response

spectra; establish procedures for choice of design variables,

and design computer programs in forms that can be directly

used by practicing engineers.

Earthquake Design Criteria for Water Supply and Wastewater Systems;

Leon Weinberger; Environmental Quality Systems, Inc., 1160 Rockville Pike, Rockville, MD 20852; Grant #77-22617.

Develop criteria for the design of water supply and wastewater

pollution control systems to reduce damage during earthquake

events and promote rapid post-quake recovery.

Analysis of Cape Ann Earthquake from Building Damage; Robert V. Whitman; Massachusetts Institute of Technology, Cambridge, MA

02138; Grant #77-15331.

Determine the strength of ground shaking caused by the

earthquake of 1755 at Cape Ann, Massachusetts.

Design

Stability of Structures Under Static and Dynamic Loading -- International Colloquium, May 17-19, 1977; Lynn S. Beedle;

Fritz Engineering Laboratory, Lehigh University, Bethlehem,

Pennsylvania 18105; Grant #76-20825.

34-8
Bring together the generators of research information on the

static and dynamic stability of structures and elements and

the users of this information to evaluate the latest research

results, identify further needed research and documentation

of static and dynamic stability information in a form suitable

to be incorporated into design guides.

Safety Evaluation of Buildings Exposed to Earthquakes and Other

Catastrophic Environmental Hazards; Boris Bresler; Department of

Civil Engineering; University of California, Berkeley, California

94720; Grant #76-82384.

Develop methods and criteria for assesing the damage potential

of existing structures. Develop guidelines based on damage

potential indices which would be suitable for use by

practicting engineers and building regulation departments.

Seismic Resistance of Fossil-Fuel Power Plants; John L. Bogdanoff;

School of Aeronautics and Astronautics; Purdue University, Lafayette,

Indiana 47907; Grant #77-01392.

Examine the structural dynamic behavior and response of major

components of large fossil-fuel steam power plants subjected

to seismic forces. Develop computer codes for the analysis

of the major components; develop analysis and design guidelines and criteria; and develop design recommendations for

code groups to consider for adaptation.

Structural Loads Analysis and Specification; C. Allin Cornell;

Department of Civil Engineering, Massachusetts Institute of

Technology, Cambridge, Massachusetts 02139; Grant #77-01389.

Develop a practical, unified approach to the analysis and

specification of structural loadings, including extremum

value loads and continuous loads by developing a relatively

simple, unified set of representatives from the random

characteristics of loads and by developing alternative

codifiable treatments of loading and load combination

specification.

Formulation and Expression of Seismic Design Provisions; Steven J. Fenves; Carnegie-Mellon University, Pittsburgh, Pennsylvania 15213;

Richard N. Wright; National Bureau of Standards, Gaithersburg

,

Maryland 20234; Grant #77-01497.

Facilitate the development, usefulness, and implementation

of the seismic design provisions that are currently being

developed by the National Bureau of Standards and the Applied

Technology Council.

Conference on the Repair and Rehabilitation of Buildings; Robert D. Hanson; Department of Civil Engineering, University of Michigan,

Ann Arbor, Michigan 48109; Grant #76-83884.

Conduct a conference and workshop consisting of professionals,

faculty and researchers to assess the current practices and

needs of the professionals concerned with the repair and

rehabilitation of structures, including other hazardous

buildings. Undertake a projection of future research.

Earthquake Induced Bond Deterioration of Reinforced Concrete, Neil

M. Hawkins; Department of Civil Engineering, University of Washington,

Seattle, Washington 98105; Grant #76-15366.

34-9
Study the bond deterioration between reinforcing bars and the

surrounding concrete when reinforced concrete elements are subjected to reverse earthquake loadings. Recommend design

regulations for seismically loaded reinforced concrete

structures

.

Effects of Damage from Strong Earthquakes; George W. Housner;

Department of Civil Engineering and Applied Mechanics, California

Institute of Technology, Pasadena, California 91125; Grant #76-

17137.

Develop practical techniques for the analysis of non-linear

structural response to seismic forces and apply the methods

to control the damage of structures beyond the elastic range.

Seismic Safety Design for Policy and Fire Stations; Earle W. Kennett; American Institute of Architects, Research Corporation,

1735 New York Avenue, NW. , Washington, D.C. 20006; Grant

#76-80695.

Explore, define and record seismic design considerations to be used by the architectural and public safety professions

in the design, construction, and operation of policy and

fire stations, including recommendations for specific,

prototypical architectural design alternatives.

Effects of Earthquake Motions on Reinforced Concrete Buildings,

Mete A. Sozen; Department of Civil Engineering, University of

Illinois, Urbana, Illinois 61805; Grant #77-01391.

Identify and understand the mechanisms of energy dissipation

in slender reinforced concrete structures subjected to strong ground motion in an effort to develop simple but

realistic design methods for earthquake resistance.

Earthquake and Wind Response of Segmentally Constructed Hyperbolic

Natural Draft Cooling Towers; Chi C. Tung; Department of Civil

Engineering; North Carolina State University, Raleigh, North

Carolina 27607; Grant #76-19663.

Study the dynamic response of segmentally constructed prestressed hyperbolic cooling towers using model studies and

develop linear and non-linear analytical models for predicting the dynamic response of such structures.

Seismic Behavior of Complete Structural Systems; Ray W. Clough;

Department of Civil Engineering, University of California,

Berkeley, California 94720; Grant #77-00280.

Develop computer procedures for predicting structural response

to earthquakes. Formulate mathematical models to reproduce

the observed behavior analytically. Develop and test special

structural components for improving the seismic resistance

and conducting reconnaissance surveys of earthquake damaged

structures.

Summer Institute on Architectural Design for Earthquake Disaster

Mitigation; John P. Eberhard; AIA Research Corporation, 1735 New

York Avenue, NW., Washington, D.C. 20006; Grant #76-80799.

34-10
Conduct a one-week short course for faculty of architectural

schools with respect to an understanding of the earthquake

phenomenon and the methods to mitigate the damages as a result of a seismic event.

Earthquake Induced Bond Deterioration of Reinforced Concrete;

Neil M. Hawkins; Department of Civil Engineering, University of

Washington, Seattle, Washington 98105; Grant #76-15366.

Study the bond deterioration between reinforcing bars and

the surrounding concrete when reinforced concrete elements

are subjected to reversed earthquake loadings. Recommend

design regulations for seismically loaded reinforced concrete

structures

.

Research Initiation-Strengthening of Reinforced Concrete Columns

for Earthquake Resistance; Lawrence F. Kahn; Department of Civil

Engineering, Georgia Institute of Technology, Atlanta, Georgia

30332; Grant #77-06478.

Determine the adequacy of various methods of strengthening

and repairing concrete columns for improved earthquake

resistance.

Third U.S. National Conference on Wind Engineering Research,

February 26-March 1, 1978; Bernard M. Leadon; Department of

Engineering Sciences, University of Florida, Gainesville,

Florida 32601; Grant #77-10170.

Bring together researchers and practitioners interested in

dynamic wind effects to exchange information, to improve

communication between researchers and users, to identify

areas of research which require further effort and to produce a proceedings of the conference.

Structural Response Under Random Wind Loading; Yu K. Lin;

Department of Aeronautical Engineering, University of Illinois,

Urbana, Illinois 51801; Grant #76-84171.

Develop stochastic models for wind forces in the alongwind and across wind directions and to establish the

stability criteria for structural response in the nonlinear range.

Reliability of Existing Buildings in Earthquake Zones-Part II.; Hugh D. McNiven; Earthquake Engineering Research Center,

University of California, Berkeley, California 94720; Grant

#77-06489.

Investigate the earthquake reliability of existing

buildings by establishing a non-linear mathematical

model representing the building and subjecting the

model to a large family of earthquake excitations.

This institution will develop the model.

Seismic Behavior of Multistory Masonry Structures; Hugh D. McNiven; Earthquake Engineering Research Center, University of

California, Berkeley, California 94720; Grant #77-00281.

34-11
Conduct experimental and analytical studies of the behavior

of masonry wall panels subjected to earthquake excitation.

Perform correlation studies with forced vibration test results and determine equivalent elastic constants of

masonry structures.

Seismic Behavior of Structures; Analysis and Design; Joseph

Penzien; Department of Civil Engineering, University of California,

Berkeley, California 94720; Grant #77-00186.

Apply improved seismic analysis capabilities to complete

structural systems and develop methods of design with

increased reliability of controlling seismic damage with a proper balance between cost and safety.

Seismic Behavior of Structural Components; Egor P. Popov; Department

of Civil Engineering, University of California, Berkeley, California

94720; Grant #77-00275.

Conduct experimental and analytical studies of the behavior

of structural components subjected to earthquake excitation.

Perform the experimental work on large-scale components

which extrapolate the experimental results to practical

situations.

Research Initiation-Nonproportional Damping of Interaction Systems

Subjected to Earthquake Motions; David T. Tank; Department of Civil

Engineering, State University of New York, Buffalo, New York 14214;

Grant #77-06425.

Explore the concept of employing measured damping data in the

computation of seismic response of interaction systems.

Shear Transfer in Thick-Walled Reinforced Concrete Structures

Subjected to Seismic Loading; Richard N. White; Department of

Structural Engineering, Cornell University, Ithaca, New York,

14850; Grant #76-23896.

Develop engineering and analytical models of the shear

transfer mechanism in cracked concrete reinforced with

large reinforcing bars. Test these models against

structures damaged in actual earthquakes.

Summer Institute of Multiprotection Design; Bernard Wobbeking;

American Society for Engineering Education, 1 Dupont Circle,

Washington, D.C. 20036; Grant #77-01090.

Educate and train the participants in the methods of

structural design to mitigate the losses as a result of

natural hazards, such as earthquakes, tsunamis, fire,

landslides, extreme winds, and others.

Reliability of Existing Buildings in Earthquake Zones Part I; James T. Yao; Department of Civil Engineering, Purdue University,

Lafayette, Indiana 47907; Grant #77-05290.

Investigate the earthquake reliability of existing buildings

by establishing a non-linear mathematical model representing

the building and subjecting the model to a large family

of earthquake excitations. This institution will formulate

reliability criteria and analyze the test results in terms of damage probability and other probablistic response measures

.

34-12
3uilding Configuration and Seismic Design; Christopher Arnold;

Juilding Systems Development, Inc., 120 Broadway, San Francisco,

California 94111; Grant #76-81821.

Provide a systematic body of knowledge and a useful

methodology that will enable a better understanding

between architects and engineers with respect to the

influence of building plan configuration on the seismic

resistance of the building.

implementation Planning for Seismic Design Provisions for Buildings;

Charles Culver; National Bureau of Standards, Gaithersburg , MD

!0234; Grant #77-15084.

Enlist the support and endorsement of the various organizations

involved in construction to facilitate the adoption of

comprehensive design provisions for buildings.

?he United States-Japan Cooperative Research Program on Largeicale Structural Tests: A Planning Study; Joseph Penzien; Departlent of Engineering, University of California, Berkeley, California

'4720; Grant #76-80835.

Establish a task committee under the United States-Japan

"Panel on Wind and Seismic Effects" program to make

detailed plans and recommendations for developing

effective cooperative research programs on large-scale

tests of structural systems.

robability Distribution of Extreme Wind; Emil Simiu; National

ureau of Standards, Center for Building Technology, Washington,

>.C. 20234; Grant #77-16113.

Establish a task committee under the United States-Japan

a probabilistic approach taking into account sampling errors,

data errors, use of short-term data and applicability of

procedures to different types of wind climates.

afety Evaluation of Structures to Earthquake and Other Natural

azards ; A.H-S Ang; University of Illinois at Urbana-Champaign

,

rbana, Illinois 61801; Grant #77-09090.

Coordinate analytical program to develop the basis for

structural safety evaluation and reliability design of

structures to resist natural hazards.

eismic Hardening of Unreinforced Masonry Walls Through a Surface

reatment ; James R. Cagley; Martin & Cagley; 6000 Executive Blvd

,

ockville, MD 20852; Grant #77-19888.

Develop a simple and economical method of reinforcing masonry

to resist seismic forces through the use of a coating.

eismic Behavior of Complete Structural Systems; Ray W. Clough

;

jniversity of California, Berkeley, California 94720; Grant

i76-04262 A02.

Develop computer procedures for predicting earthquake responses

through the use of mathematical models and actual buildings.

34-13
Masonry Buildings —Response of Existing Systems to Earthquake

Motions; Robert D. Ewing; Agbabian Associates, 250 N. Nash St.,

El Segundo, California 90245; Grant #77-19829.

Identify improved methods for analyzing the earthquake

response of existing unreinforced masonry structures and

develop methods for strength retrofitting.

The Use of Structural Foams to Improve Earthquake Resistance of

Buildings; B. C. Gabrielsen; Scientific Service, Inc., Redwood

City, California 94063; Grant #77-21426.

Investigate the feasibility of using polyurethane foams

to increase the shear capacity of timber stud walls and

timber joist floors and ceilings.

Earthquake Resistant Design of Braced Steel Frame Structures;

Subhash C. Goel; University of Michigan, Ann Arbor, Michigan

48109; Grant #76-82209.

Study configurations of braced multistory steel frame

systems subjected to seismic forces to determine their

advantages and disadvantages.

Engineering Design for Natural Hazards; W.J. Hall; University

of Illinois, Urbana, Illinois 01801; Grant #77-07190.

Develop simplified and improved methods of structure design

to provide protection against natural hazards.

A Rational Approach to Damage Mitigation in Existing Structures

Exposed to Earthquakes; Ben Kacyra; Earthquake Engineering Systems,

Inc., 141 Battery Street, Suite 400, San Francisco, California

94111; Grant #77-19463.

Undertake a feasibility study of developing a rational

decision analysis methodology for evaluating possible

modifications for existing buildings exposed to a predicted earthquake.

Mitigation of Seismic Hazards in Existing Unreinforced Mansonry

Wall Buildings; John Kariotis, Kariotis, Kesler and Allys; 1414

Fair Oaks Avenue, South Pasadena, California 91030; Grant #77-

19651.

Analyze performance of buildings with unreinforced masonry

walls built prior to adoption of seismic building codes;

evaluate methods of revising performance of unreinforced

elements; investigate changes in response caused by

modi f ications.

Reliability of Existing Buildings in Earthquake Zones—Part II; Hugh D. McNiven; University of California, Berkeley, California

94720; Grant #77-06489 A01.

Develop a realistic method of evaluating seismic reliability

of existing structures through development of mathematical

models, development of "failure criteria," and tests of

model structures.

34-14
Seismic Behavior of Multistory Masonry Structures; Hugh D. McNiven;

University of California, Berkeley, California 94720; Grant #76-

04265 A02.

Test the seismic shear resistance of single and double window

piers in multistory masonry buildings; correlate results with

various matematical models.

Structure-Fluid Interaction Due to Earthquakes; C.C. Mei;

Massachusetts Institute of Technology, Cambridge, Massachusetts

02139; Grant #77-10236.

Conduct theoretical and numerical investigations of the phe- nomenon of structure-fluid interaction in dam-reservoir

systems, offshore structures, and other structures of this

type.

The Dynamic of Structure with Localized Nonl inear ity ; Richard K. Miller; University of California, Santa Barbara, California 93106;

Grant #77-01096.

Develop a better understanding of earthquake behavior of

localized nonlinear systems and develop more efficient

analysis techniques for the transient and earthquake

responses of such systems.

Seismic Behavior of Structures; Analysis and Design; Joseph

Penzien; University of California, Berkeley, California 94720;

Grant #76-06264.

Develop improved seismic analysis capabilities and apply

them in structural designs for increased reliability of

controlling seismic damage and costs.

The United States-Japan Cooperative Research Program on LargeScale Structural Systems; Joseph Penzien; University of

California, Berkeley, California 94720; Grant #76-80835 A01.

Undertake a planning study for development of an effective

research program of maximum benefit to both the US and Japan

on large-scale tests of structural systems.

Methodology for Mitigation of Seismic Hazards in Existing

Unreinforced Masonry Buildings; C. W. Pinkham; S.B. Barnes & Associates; 2236 Beverly Boulevard, Los Angeles, California

90027; Grant #77-19523.

Review past work on methods of retrofitting unreinforced

masonry buildings, determine the need for hazard

mitigation, and determine methods to retrofit economically.

l>!Seismic Behavior of Structural Components; Egor P. Popov;

University of California, Berkeley, California 94720; Grant #76-

34263 A03.

Conduct advanced experiments and studies of braced steel

frames, reinforced concrete walls and filled frames,

reinforced concrete ductile frames, and reinforced beamcolumn components; use results to construct mathematical

models for analyzing inelastic behavior.

34-15
Seismic Behavior and Design of Buildings; Jose M. Roesset;

Massachusetts Institute of Technology, Cambridge, Massachusetts

02139; Grant #77-14174.

Improve present knowledge on the non-linear dynamic

behavior of buildings subject to earthquake excitation

and on the effectiveness of various design procedures.

Reliability of Existing Buildings in Earthquake Zones—Part I; James T.P. Yau; Purdue University, Lafayette, Indiana 47907;

Grant #77-05290.

Develop a realistic method of evaluating seismic reliability

of existing structures through development of mathematical

models, development of "failure criteria," and tests of model

structures

.

Conference on Stability Problems of Mixed Steel-Concrete Structures,

Boston, Massachusetts, May 1978; Lynn S. Beedle; Lehigh University,

Bethlehem, PA 18015; Grant #78-08818.

Bring together information on the static and dynamic stability

problems and performance characteristics of mixed steelconcrete composite construction and publish this information

for distribution.

Seismic Investigation and Design Criteria for Industrial Storage

Racks; John A. Blume; URS/John A. Blume & Associates, 130 Jesie

Street, San Francisco, California 94105; Grant #77-06505.

Analyze and test stacker-type industrial storage racks

including standard pallet racks and their subassemblies,

drive-in and cantilever racks to determine their behavior

under seismic forces.

Earthquake Engineering Research Facility Support; Ray W. Clough;

University of California, Berkeley, California 94720; Grant #77-

21787.

Provide continuing support for the essential maintenance and

operation of inelastic seismic performance experimental

facilities required at the University of California,

Berkeley, including the Earthquake Simulator Laboratory,

dynamic controlled test systems, high speed digital data

aquisition systems, and field vibration test systems.

Formulation and Expression of Seismic Design Provisions; Steven

J. Fenves; Carnegie Mellon University, Pittsburgh, Pennsylvania

15231; Grant #78-11300.

Facilitate the development, use and implementation of the

seismic design provisions that are currently being developed

by the National Bureau of Standards and the Applied Technology Council.

Progressive Collapse of Transmission Line Structures Due to Dynamic

Loads; John F. Fleming; University of Pittsburgh, Pittsburgh,

Pennsylvania 15231; Grant #77-23519.

Investigate the effect of partial failures induced by earthquake or other dynamic loads in transmission line structures,

with particular emphasis on predicting progressive collapse

conditions, and develop design criteria which would prevent

progressive collapse.

34-16
Scale Modeling and Testing of Structures for Reproducing Response

to Earthquake Excitation; James M. Gere; Stanford, California

94305; Grant #77-14444.

Investigate the direct replica modeling of elastic behavior

of civil engineering structures to achieve a routine

capability for reproducing, on existing small-scale test facilities, the dynamic response of large structures

subjected to prescribed earthquake motions.

Behavior of Buildings to Wind and Seismic Forces; Gary C. Hart;

University of California, Los Angeles, California 90024; Grant

#77-24062.

Study the behavior and response of full scale buildings in

Century City/Los Angeles when subjected to strong winds

and seismic forces; compare such responses with the soil

structure interaction due to these forces.

Seismic Response of Structures and Strong-Motion Instrumentation;

George W. Housner; California Institute of Technology; Pasadena,

California 97125; Grant #77-23687.

This project includes the analysis of earthquake records,

operation and upgrading of instruments and protable

instrument networks analysis of buildings in the nonlinear

range, and behavior of materials and structures.

Behavior of Reinforced Concrete Frame Elements Under Biaxial

Lateral Loadings; James 0. Jirsa; University of Texas, Austin,

Texas 78712; Grant #77-20816.

Evaluate the importance of the path of lateral deformation on sequence of application of bidirectional lateral loads on

the shear strength and response of columns and beam-column

joints; evaluate the influence of axial load variations on

the behavior of columns and beam-column joints; develop

design recommendations for such members; and develop behavioral models to be used in analysis of structures to predict

behavior for members subjected to large shear forces.

Seismic Safety Design for Policy and Fire Stations; Earle W. Bennett; American Institute of Architects Research Corporation

(AIA/RC) 1735 New York Avenue, NW, Washington, D.C. 20006;

3rant #77-28413.

Explore, define, and record seismic design considerations

to be used by the architectural and public safety

profession in the design, construction, and operation

of police and fire stations, including recommendations

for specific prototypical architectural design alternatives.

M> I Jnified Approach to the Design of Window Glass Subjected to Dynamic

..oads; Joseph E. Minor; Texas Tech University, Lubbock, Texas

'9409; Grant #77-24063.

Review the processes in which glass panels are selected for use in engineered structures to provide the practitioner

with safe, economical designs for glass windows which utilize

current knowledge of wind engineering, glass material

properties, and structural response of thin rectangular

panels

.

34-17
An Evaluation of the Earthquake Rehabilitation Building Ordinance

in the City of Long Beach, California; William J. Petak; J. H. Wiggins Company; 1650 South Pacific Coast Highway, Redondo Beach,

California 90277; Grant #77-22621.

Review, assess and evaluate the Earthquake Hazard Regulations

for Rehabilitation of existing structures adopted by the

City of Long Beach in 1971; determine the effectiveness of

the ordinance, the effectiveness of the City Government in

implementing the Ordinance, the costs and consequences of

implementation, and the transferability of this approach

to other natural hazard areas.

Fire Resistance of Epoxy Repaired Concrete Structures; Joseph M. Plecnik; California State University, Long Beach, California

90840; Grant #77-22946.

Investigate the behavior of epoxy repaired concrete structural

walls during fire exposure; determine the nature and extent of

residual strength of epoxy repaired structural components after

fire exposure; and prepare a Computerized Catalog of repaired

structures to study their behavior during future fires and/or

earthquakes

.

Feasibility of Force-Pulse Generators for Earthquake Simulators;

F.B. Safford; Agbabian Associates, 250 N. Nash Street, El Segundo,

California 90245; Grant #77-15010.

Design, construct and evaluate prototype pulse systems to simulate earthquake wave forms in full scale structures.

Improving Earthquake Resistance of Elevators; Anshel J. Schiff;

Purdue University, Lafayette, Indiana 47907; Grant #77-21268.

Investigate three specific problem areas for improvement

of the earthquake provisions of the national elevator code:

counterweight and guide rail design, the determination of

seismic loads on elevator systems, and the analysis of cable

dynamics .

Summer Institute on Design for Protection Against Natural and

Manmade Hazards; Bernard Wobbeking; American Society for Engineering

Education; 1 Dupont Circle, Washington, D.C. 20036; Grant #77-01090

A02

Conduct a Summer Institute to train selected participants in

methods of structural design to mitigate losses as a result

of natural hazards such as earthquakes, tsunamis, fire,

landslides, extreme winds, and others. Participants will be

requested to teach similar courses in their universities and

local areas. The courses for engineers and architects are on

a local basis to reach the practicing professionals who can apply the results of the course immediately.

Policy

National Information Service for Earthquakes-University of California

Berkeley; Ray W. Clough; Earthquake Engineering Research Center,

University of California, Richmond, California 94804; Grant #77-01934

34-18
Develop an effective information service for earthquake

engineering by developing a library of technical information;

by preparing and publishing programs and an abstract journal;

by disseminating information; and by providing educational

opportunities for professional practitioners.

Inflationary Changes in Construction Costs Created by Earthquake

Damage: Implications for Government Policy; Harold C. Cochrane;

Department of Economics, Colorado State University; Ft. Collins,

Colorado 80523; Grant #76-24169.

Assess the extent of disaster-related inflation following

the occurrences of hazardous events. Prepare recommendations

for legislative and administrative actions to counteract it.

National Information Service for Earthquake Engineering—California

Institute of Technology; George W. Housner; Department of Civil

Engineering and Applied Mechanics, California Institute of Technology,

Pasadena, California 91125; Grant #77-07472.

Serve as a source of earthquake engineering information with

particular emphasis on strong earthquake recording data such

as digitized acceleration, velocity and displacement.

Provide a unique earthquake engineering reference source collection and make available to researchers and practitioners

copies of reports, recorded accelerograms, digitized card

decks and tapes.

State Government Policy Options for the Utilization of Earthquake

Prediction Technology; Hirst Sutton; Council of State Governments;

1225 Connecticut Ave., NW . , Washington, D.C. 20036; Grant # 77-81112.

Examine public policy, legal, legislative, program, and

administrative issues related to the evolving technology

of earthquake prediction, with the objective of developing

options in state policies and programs.

Tall Buildings and Urban Habitat: Impact on the Urban Environment

|and Planning for Natural Disasters; Lyn S. Beedle; Fritz Engineering

laboratory, Lehigh University, Bethlehem, Pennsylvania 18015;

Jrant #77-01598; New Award.

Develop utilizat ion/implementat ion techniques to transfer

the large quantity of research information and data

collected by the Council on Tall Buildings and Urban

Habitat on tall buildings and their interaction with the

urban environment to decisionmakers, and others concerned

with the conception, planning, design, and operation of

tall buildings with particular emphasis on planning for

natural hazards.

ymposium on Structural Engineering and Mechanics; Karl S. Pister;

department of Civil Engineering, University of California, Berkeley,

alifornia 94720; Grant #77-06518; New Award.

Conduct a symposium to bring researchers, faculty and professionals together to present and discuss current knowledge

needed in several areas of dynamic behavior, and developments

in education, with respect to structural engineering and

mechanic s

.

34-19
Disaster Knowledge and Beliefs and Emergency Planning; Dennis E. Wenger; Department of Sociology, University of Delaware, Newark,

Delaware 19711; Grant #77-10202; New Award.

Analyze public and official knowledge regarding the consequences of natural disasters. Assess community disaster

preparedness

.

Building Futures Forum—The Importance of the Built Environment to the Quality of American Life; Robert M. Dillon; National Academy

of Sciences, Washington, D.C. 20418; Grant #77-13297; New Award.

Assemble a group of persons to examine some aspects of the

interaction between the man-built enviromment and individual

and societal health. Identify forces that are causing and are likely to precipitate changes.

A Longitudinal and Cross Cultural Study of the Post Impact Phases

of a Major National Disaster; Frederick L. Bates; Department of

Sociology, University of Georgia, Athens, Georgia 30602; Grant

#7 7-12721; New Award.

Identify, describe and measure the impact of a major earthquake on a nation's socio-economic units and develop programs

analyze external (U.S.) disaster relief and rehabilitation

delivery systems and their long and short-term effects; and

develop alternative approaches for programs.

Search and Rescue Missions in Natural Disasters and Remote Settings

Thomas E. Drabek; Department of Sociology, University of Denver,

Denver, Colorado 80208; Grant #77-14162; New Award.

Determine the problems rated to search and rescue in large

natural disasters in remote areas and develop policy and

procedural alternatives for dealing with them.

Community Response to Natural Hazard Warnings; Robert K. Leik;-

Department of Sociology, University of Minnesota, Minneapolis,

Minnesota 55455; Grant #77-01452; New Award.

Identify and analyze by field and laboratory procedures the

variables that influence the manner, speed, and appropriateness of the response by families, organizations, and

communities, to four major hazard warning systems. Develop

alternative app.oaches as needed.

Post-Earthquake Land Use Planning; George G. Mader; William Spangle

and Associates; 3240 Alpine Road, Portola Valley, California; Award

#76-82756.

An Interdisciplinary team will investigate post-earthquake

land use planning and reconstruction in San Fernando, Santa

Rosa, and Alaska and will develop recommendations for postearthquake planning procedures, programs and regulations.

Seismic Safety Preparedness by Local Governments in California;

Dean E. Mann; University of California, Santa Barbara, Santa

Barbara, California 93106; Award #77-03688.

Four communities in California will be examined to ascertain

sources of resistance to adequate seismic safety planning.

The study will provide officials at all levels of government

34-20
and in the private sector with improved strategies for over- coming resistance to effective seismic safety planning.

National Information Service for Earthquake Engineering (NISEE);

Ray W. Clough; University of California, Berkeley, California

94720; Award #76-20744 A02.

Provide for the transfer of earthquake and other hazards

information generated through research to the public users.

Community Response to Natural Hazard Warnings; Robert K. Leik;

University of Minnesota, Minneapolis, Minnesota 55455; Award

#78-10634.

Provide policy level officials with information on communities response to floods, hurricanes, tornadoes and earthquakes;

clarify how warning responses are related to content and timing of messages; interorganizational relation, predisaster

planning and individual decisions.

Extension and Utilization of Community Flood Model; Louis W. Miller;

University of Pennsylvania; Philadelphia, Pennsylvania 19104;

Award #77-26363.

Determine how the physical and economic characteristics of

communities influence the decision-making process for hazard

mitigation and recovery; determine the financial impact of

natural disasters on individual households; access the role

of financial institutions in disaster mitigation; and study

the impact of a natural disaster on regional economies.

Social and Psychological Factors in Evacuation in Natural Hazards;

Ronald W. Perry; Batelle Human Affairs Research Center, 4000 N.E.

41st Street, P.O. Box 5395, Seattle, Washington 98105; Award

#77-23697.

Produce a theoretically and empirically grounded model of the

factors affecting evacuation which will help isolate key issues

in the design and implementation of evacuation plans.

Natural Hazards Research and Applications Information Center;

Gilbert F. White; University of Colorado; Boulder, Colorado 80309;

Award #77-05164.

Accelerate natural hazards research and the application of

findings through an information center service which disseminates information on research findings, programs, and

conferences; and conduct annual and special natural hazards

workshops

.

34-21

35

RAPID SEISMIC ANALYSIS PROCEDURES FOR BUILDINGS

T. K. Lew

S. K. Takahashi

C. V. Chelapati

Civil Engineering Laboratory

Port Hueneme, California

Rapid Seismic Analysis Procedures for Buildings

Naval installations are not exempt from earthquake hazards. In addition to the potential monetary loss, the damage to Naval

installations from earthquakes can contribute to the loss of

operative capacity and life-hazard. Thus, it would be desirable

to obtain an overall estimate of damages to an installation when

the site is subjected to a potential earthquake. A typical

installation may contain 100 to 500 buildings constructed from

different materials and built in different years. Understandably,

it would not be economically feasible to analyze such a large number of buildings in detail. A rapid, systematic, and economical

procedure is needed.

The rapid seismic analysis procedure discussed here has been

developed for determining damage estimates from potential earthquake motions to buildings. The procedure does not consider other

potential earthquake damage to the buildings from tsunamis (seismic

sea waves) and soil failures such as liquefaction and differential

settlement

.

Based on the resulting damage estimates from the analysis,

the potential damage to the installation is quantified and the

relative seismic resistance of the buildings determined. The more

vulnerable buildings are identified and recommended for further

detailed study. Based on results from the detailed study, recomlendations are made to (1) modify and strengthen the inadequate

existing buildings against the potential earthquake, or (2) plan

ia systematic phasing-out of the unsafe buildings with new, adequately

designed buildings.

The procedure presented here was originally developed by

John Blume and Associates for their seismic study on Puget Sound

l^aval Shipyard, Bremerton, Washington to estimate the potential

iamage on the shipyard buildings from strong earthquake motions [1].

Results of the original procedure were presented at a Naval

facilities Engineering Command (NAVFAC) workshop given in San Fran-

:isco during December 1973 [2].

The procedure has been modified and adapted by the Civil

Engineering Laboratory (CEL) for seismic studies on Naval Air

Station, North Island (NASNI) , San Diego [3], Naval Construction

iattalion Center (NCBC), Port Hueneme, and Long Beach Naval Ship-

'ard (LBNSY), Long Beach, California.

The major modifications that CEL has made are systematization

f the analysis of the facility inventory assets at a Naval instalation and the development of the response spectra for the design

arthquakes. In addition, CEL has automated the computation of

hear stiffness for concrete or masonry buildings, the first mode

hape and natural period of multi-story buildings, and estimation

f damage for peak ground accelerations from 0.05g to 0.50g.

Generally, the rapid seismic analysis procedure consists of

[he following steps:

1. Selection of buildings.

2. Determination of response spectra for design earthquakes,

3. Estimation of yield and ultimate capacities.

4. Estimation of natural periods

5. Estimation of damage from the response spectra.

35-1
In the following sections, each of the steps is discussed.

Then, a brief discussion of accuracy of the results is given.

Because of the breadth and the complexity of earthquake engineering, it is not possible to discuss all the aspects of earthquake

engineering here in detail. However, the pertinent aspects of earthquake engineering relating to the rapid seismic analysis procedure

will be presented. Information on technical subjects may be found

in references [4 to 7], Reference [4] is highly recommended, for

it covers the total field of earthquake engineering and contains

many useful references for the practicing structural engineer.

Selection of Buildings

Because of the generally large number of buildings at the Naval

Installation, it is not economically feasible to analyze all the

buildings. Generally, about 40 mission-important and permanent

buildings at the installation are screened visually. However, only

the more critical buildings from this group are selected for analysis. The building selections are determined by the engineers performing the seismic study, NAVFAC personnel, the Public Works

Officer (PWO), and Planning Department representative of the Naval

installat ion.

The following procedure was used for the selection of the

buildings

:

(1) Make a site survey to visually screen the mission-important

and major permanent buildings at the installation. The purpose

of the survey is to determine the physical conditions of these

buildings and take photographs for future reference.

(2) Meet with the PWO and planning department representative

of the installation to select the mission-important and representative major permanent buildings for structural analysis.

(3) Gather structural drawings and calculations, if available,

from the Public Works Office or NAVFAC field division. Sometimes, drawings for the older buildings are not available or details on the drawings are very sketchy. In the later case, some detective work was required to determine the size of the

lateral-force resisting members from the available drawings.

With the exception of some of the newer buildings, calculation

sheets are generally not available.

(4) Copy information required for the rapid analysis onto

standard forms.

(5) Make a final visit to the installation to verify that the

buildings are built as shown on the drawings, especially the

lateral-force resisting elements.

To facilitate the selection of the buildings for the visual

screening, Facilities Systems Office (FACSO) has developed for CEL

a computer program that takes the data tape files of NAVFAC P-164,

Detailed Inventory of Naval Facilities, and reorders them according to:

(1) Replacement cost (largest to smallest)

(2) Year built (oldest to newest)

(3) Building size (largest to smallest square footage)

35-2
The program lists the buildings with the established category code

numbers (CCN's) for Navy facility assets as follows:

CN

LOO 200

iOO

f00 iOO

>00

f'00 ;oo 100

Class Category

2 Operational and Training Facilities

2 Maintenance and Production Facilities

2 Research, Development & Production Facilities

2 Supply Facilities

2 Hospital and Medical Facilities

2 Administrat ive Facilities

2 Housing and Community Facilities

2 Utilities and Ground Improvement

1 Real Estate

le output from the computer program is used as an aid to select

luildings for visual screening.

A sample output from the FACSO computer program is given in

?able Al of the Appendix.

The Program is general and can be used to analyze the facility

iventory assets of any Naval installation. The computer program

|s available at FACSO, NCBC, Port Hueneme

.

^termination of Response Spectra for Design Earthquakes

In this section, determination of the regional and local geojogy for the site of the installation is discussed first. Special

uphasis is placed on the regional geologic evidence of seismic

ptivity because it can be used to estimate the probable magnitudes

^cations, and frequencies of seismic events. Determination of the

istoric seismicity for the site is discussed. Determination of

iarthquake magnitudes on the selected faults is discussed. Then,

stimation of the maximum accelerations for the site using the empirical formula for attenuation of earthquake acceleration versus

Lstance and the Monte Carlo technique with random numbers is iscussed. Finally, determination of the response spectra using

|ie scaled actual earthquake response spectra compiled by the Cali-

>rnia Institute of Technology, Pasadena, California, is discussed.

The strength of the earthquake is measured by an intensity

j;ale of the Richter Magnitude scale. The intensity scale is a ibjective measure of the effects of an earthquake. It refers

the degree of shaking at a specified place. Intensity deterinations can be heavily influenced by the site conditions and type

: construction. Over the years, many scales were developed by

fferent individuals, notably by Mercalli and by Rossi-Forel.

ie most widely adapted one is the Modified Mercalli (MM) scale,

has twelve grades denoted by Roman numberal I-XII.

On the other hand, the Richter Magnitude (M) scale is defined

the equation

M = log (A/A

Q

) < 8.9

ere M = Richter Magnitude

(1)

A

Q = Recording on a standard Wood-Anderson Seismograph with

1/1000 mm amplitude.

35-3
A = Peak amplitude from the recording on a standard WoodAnderson Seismograph at 100 km from the earthquake center.

The Richter Magnitude scale measures the energy released from the

earthquake. Because it is a log scale, a magnitude 8 earthquake

releases about 32 times more energy than a 7 earthquake.

Regional Geology

Geologic evidence of seismic activity of a region is valuable

in the evaluation of the seismic risk for a given site. This

information can be used to estimate the probable magnitudes, locations, and frequencies of future seismic activities in the area. Moreover, by knowing the type of movement along a given fault, some characteristics of the ground motions may be predicted.

Regional earthquake geology involves the study of Tectonic

deformations, deformation of the rock structure from the deepseated crustal and sub-crustal forces in the earth. The purpose

of such study is to determine the nature, position, age, and movement history. The main geological features studied are warping,

tilting, faulting, and tectonic structure. Faulting generally

receive most of the attention. Because of its relevancy to earthquake engineering, only faulting will be discussed.

The main features of faulting pertinent to earthquake engineering are location, activity, and type.

Maps showing the location of faults can generally be obtained

from the State's Division of Mines and Geology, oil or water

company doing exploratory work, or U. S. Geode+ic Survey. However,

such maps are generally incomplete. The source of energy release

from an earthquake may be under a thick layer of soft soils. Consequently, no evidence of surface faulting can be seen from such an earthquake. Therefore, inference, interpolation, or extrapolation

from a fault map may be necessary to determine the locations of

the faults in the vicinity of the site.

A fault is considered active if it is known to be associated

with an earthquake during historic times. A potentially active

fault is one that has had no historic seismic activity, but has

shown evidence of activity during the last 35,000 years. When

in doubt about the activity of a fault, assume it to be potentially

active for conservatism.

The four types of faults that should be considered in a regional geology study are:

(1) Low-angle compressive, underthrust faults.

(2) Compressive, over thrust ing faults.

(3) Extension faults.

(4) Strike-slip faults.

Apparently, the characteristics of strong ground motion in the

vicinity of the causative fault can be influenced by the type of

faulting. Large horizontal motions are associated with strike-slip

faults. By contrast, large vertical motions are associated with the

compressive, overthrust ing faults.

A map showing portions of major faults near San Diego, California

is shown in Figure 1. Many of the larger earthquakes are not associated with surface fault breaks. In the selection of faults in the

vicinity of the site, be sure to include all those that are adjacent

35-4
to the site. Their exclusion would result in an underestimation of

the maximum ground acceleration for the site and, hence, the damage

Local Geology

As mentioned earlier, the primary effect of the local geology

is its influence on the dynamic amplification or attenuation of the

earthquake ground accelerations and its frequency content.

The depth of the soil overlaying bedrock affects the dynamic

response and natural period of the ground. At distances less than

or equal to 25 km from the earthquake source, the peak ground

accelerations for alluvium sites and intermediate sites are 0.60

and 0.75, respectively, times that for rock sites [8], By contrast

at distances between 26 km and 50 km, the peak ground accelerations

for alluvium sites and intermediate sites are 1.09 and 1.16, respec

tively, times that for rock sites [8]. Furthermore, the natural

period of the site increases with soil depth. The natural period

determines frequency of the waves filtered out by the soil and

relates to the amount of soil-structure interaction.

The effects of soil-structure interaction is illustrated by

the response spectra in Figure 2 for alluvium, intermediate, and

stiff (rock) sites. The peak ground acceleration is one. From

the curves in this figure, it can be seen that the dynamic

response of structures located in rock sites peaks at a period

of about 0.15 second and decreases with increasing periods thereafter. The location of the maximum structural response is caused

by the resonance effect of the structure with the site. The

apparent site period is about 0.15 sec.

On the other hand, the location of the maximum response for structures located in alluvium and intermediate sites is at about

0.3 sec, the approximate natural period of these sites. At periods greater than 0.3 sec, the structural response for the

alluvium and intermediate sites tends to greatly amplify the earthquake motions in the longer period range. Because it is softer,

the response of structures in alluvium sites are larger than those

jfor the rock site. This is because of the softer alluvium and

intermediate sites in this range.

The topography of the bedrock and soil deposites has various

effects on the seismic ground waves such as reflection, refraction,

focusing, and scattering. It is possible that local geological

features such as hidden irregularities in the bedrock topography

itaay have caused the unexplained difference in response observed

at two nearby sites in 1971 San Fernando earthquake. Accelerations of 0.21g and O.llg were recorded at two locations at the

California Institute of Technology campus. The local soil profiles at both locations were considered identical.

A ridge is another topographical feature that may affect the

structural response. Magnification of basic bedrock motions may

iiccur at these locations.

Soil failures which may result in liquefaction of the soil

ind/or differential settlement of the building are beyond the

xope of the rapid seismic analysis procedure and are not considered .

35-5
Historical Seismicity

Background data on the historic seismicity of the site may be

obtained from local officials, engineers, seismologists, local

building regulations, published papers, and reference books. All

earthquakes with magnitudes greater than or equal to 4 should be

included in the data compilation. The reliability of the earthquake frequency predictions increases with the amount of data

avai lable

.

A recurrence curve showing the number of earthquakes per year

for a given area versus magnitude is prepared from the historical

data. The resulting equation is expressed in log linear form,

log n = a - bM (2)

where

n = recurrence interval in no. of events of magnitude M per

year

a, b = empirical constants

For instance, the information required for Southern California

area is available in reference 9. The curve for San Diego is shown

in the middle of the top of Figure 3 and has a "b" constant of 0.92.

Incidentally, the b constant in Figure 3 is the inverse function of

the seismic activity of a region. The corresponding b constant for

the western and eastern United States are 1.14 and 1.38, respectively,

It is of interest to note on this curve that an earthquake of about

4 magnitude can be expected once a year in the San Diego area. However, it is very difficult to accurately predict the seismicity for

a particular site, especially when historic seismic data are lacking.

Determination of Magnitude on Faults

Based on the data available, the maximum credible earthquake on

the selected faults is determined from the relationship between

earthquake and the length of slipped faults, Figure 4 [10,11].

The bilinear curve in the figure can be expressed as

For L < 6.6725 miles,

(3) M = In L + 4.006

and for L = > 6.6725 miles,

M = 0.5 (In L + 10.7019) (4)

where M = earthquake magnitude

L = length of slipped fault in miles

The maximum earthquake magnitudes for the faults can be

obtained by substituting the fault lengths into Equations (3) and

(4).

Next, Equation (2) is rearranged into the following forms,

a = log n + bM (5)

Mn = a ~ log n

(6)

b

35-6
M25 =

a + 1.40

(7)

where n = earthquake frequency (events/year)

= magnitude of the 25 year earthquake

Equations 5 to 7 are used to determine the constant a and the 25 year earthquake magnitude,

An earthquake frequency n (either obtained from studies by seismologists or assumed) is assigned to each of the selected faults.

The assigned frequency is substituted into Equation 5 to determine

the constant a. The validity of a is then checked by substituting

it into Equation 6 and computing the known earthquake magnitudes

associated with the fault in the past. If the computed earthquake

magnitudes are too large or small, the assumed frequency is adjusted

and the whole process is repeated. When a satisfactory value is

obtained, it is used in Equation 7 to compute the 25-year earthquake

magnitude. For faults where seismic event data are lacking, judgment must be used to determine a reasonable value.

Example s

1. San Andreas fault, one 8.0 in 100 years,

a = log 0.1 + 0.93 (8.0)

= - 2.00 + 7.44

= 5.44

M25 5.44 + 1.40

0.93

7.35

2. Newport-Inglewood fault, one 6.2 in 100 years

a = log 0.01 + 0.93 (6.2)

= - 2.00 + 5.77

M25

= 3.77

= 3.77 + 1.40 = 0.93

5.56

Next, the 90% probability acceleration level according to the

|MVFAC criteria is computed.

Maximum Acceleration

The maximum ground acceleration from an earthquake along one

)f the nearby faults is an important parameter used in the estimation of structural response and damage at the site. In addition

0 this parameter, the duration of the earthquake motion and its

irequency content are the other determining parameters in estimating

he structural response and associated damage that may occur at he site.

The duration of the earthquake motion is a function of its maglitude while the earthquake magnitude depends on the energy release

it the earthquake source. Relationships proposed by Housner [10, 11] >etween magnitude and duration, magnitude and maximum acceleration,

md magnitude, distance to fault, and maximum acceleration are shown

35-7
in Figure 5. The upper curve in Figure 5 is used to determine the

duration of earthquake acceleration. Understandably, for a given

maximum acceleration, the potential damage to structures at a site

increases with the duration.

As the maximum earthquake acceleration propagates from its source, its amplitude attenuates with increasing distance from the source. The amount of attenuation depends on a great many factors

including the physical properties of the propagation medium and

the actual path traveled. There are many empirical formulas for

estimating the attenuation of the earthquake acceleration with

distance from the source. An illustration of the variation in re- sults that can be obtained from such a formula is given in Figure 6.

For the rapid seismic analysis procedure, the maximum ground

acceleration according to the NAVFAC criteria is determined by using

the Trifunac-Brady empirical relationship for the intermediate soil

condition and the Monte Carlo method.

The Trifunac-Brady empirical relationship [reference 12] is

based on the analysis of the voluminous earthquake ground motion

data available to the California Institute of Technology (Cal Tech).

This formula gives very large accelerations for earthquakes close

to the site. These large values are caused by the lack of data

for distances closer than 20 km and the large scatter of the data

points for the range. Moreover, only a few data points are available

for distances closer than 20 km from the source. To obtain more reasonable accelerations for earthquakes close to the site, CEL has

modified the formula to use the focal distance instead of the epicentral (horizontal) distance from the earthquake source. This modification reduces the maximum accelerations from earthquake on nearby

faults somewhat but have negligible effect on the maximum accelerations from earthquake far away.

To compute maximum accelerations at the site according to the

NAVFAC criteria, CEL has developed a computer program CEL3, that

incorporates the Trifunac-Brady empirical relationship and the

Monte Carlo method for determining the maximum acceleration with

90% probability of not being exceeded in a 25-year period.

Briefly, the range of the 25-year earthquake magnitudes (M25),

the range of horizontal fault distances to the site (shortest to longest), the range of focal depth, the soil type, and the number

of samples are input into the program. Generally, 1000 samples

are specified to assure a good statistical distribution. The

computer program varies the range of input parameters at random

for each sample with the aid of a random number generator and

computes at the maximum acceleration and keeps track of the number

of occurrences at each acceleration level.

Example output from the computer program CEL3 for the San

Andreas fault near San Diego is given in Table 3.

The 90% probability acceleration levels for a site is determined from the output. From similar tables, the controlling near

field earthquake and the far field earthquake are determined.

The reason for selecting a far and near field earthquake for

the development of the site response spectra is that the former

is generally more damaging to structures with longer natural

periods and the latter is more damaging to shorter period structures.

35-8
Furthermore, the computed 90% probability acceleration level

for near and far field earthquakes may have to be adjusted before

they can be used in the development of the site response spectra,

he maximum acceleration for the near field earthquake, for

instance, was reduced by one-third because typical earthquake

acceleration records generally contain very few peaks greater than

two-thirds times the maximum acceleration as illustrated in the

cceleration records shown in Figure 7. There are 5 and 7 peaks

bove 2/3 times the maximum value, respectively for the upper and

lower curves. Moreover, the durat ion of the near field earthquake

ay be relatively short in some instances and its damage potential

is smaller than that from the longer duration far field earthquake.

Because of the relatively long duration of the far field earthquake,

much smaller reduction was used.

A plot of the probability versus maximum acceleration for the

>5-year earthquake at a Navy site in San Diego is shown in Figure 8,

In this figure, the probability of being exceeded is plotted on the

/ertical axis and the maximum acceleration is plotted on the horizontal axis. The data for this curve were obtained by normalizing

he results from the Monte Carlo program for the San Andreas fault

Eor 10% probability of occurrance to 0.066g's. From the curve in

:he figure, it can be seen that there is 100% probability that the

itiaximum acceleration at the site will be greater than 0.022g. The

eturn period acceleration at the site will be greater than 0.066g.

!|rhe return period for this 0.066g accelertion is 250 years (25

rears/0. 10). Furthermore, the shape of the curve is essentially

identical for other faults.

Response Spectra

The response spectra determined for the site is a composite of

he spectra from the near and far field earthquakes. The response

kpectra for each of the earthquakes were obtained by scaling selected

pectra from the Cal Tech earthquake data records [13] which are

ivailable in both printed form and in magnetic tapes. The reference

ontains the following information:

Vol II - 987 corrected accelerogram records

Vol III - Response spectra for 987 corrected records of Vol.

II.

jhe printed information in Volume III was recorded on Cal Tech's

aster tapes NIS159 and NIS160. The information in these tapes

jas processed by a CEL computer program, CEL4 , to:

(1) Compute epicentral distance from coordinates of" the recording

stations and epicenters.

(2) Order all events according to the duration of the earthquake .

(3) Order all events according to the epicentral distance.

jnfortunately , the durations recorded on the tapes were the total

ecorded durations of the earthquakes and not the durations of the

trong-motion phase. Hence, the events order according to duration

jere useless.

The response spectra from the Cal Tech tape used in the develpment of the site spectra were selected according to similarity

n soils at the recording site, magnitude, and epicentral distances.

35-9
This selection criteria automatically includes the effects of strongmotion duration. With very few exceptions, a set of response spectra

that requires a scaling factor larger than 2.5 to 3.0 is rejected.

This is done to ensure the satisfaction of the similarity criteria.

The selected response spectra were scaled and plotted by program CEL5 . To minimize confusion between data for different curves, no more than six spectra curves are plotted on each plot. A sample

plot of the far field response spectra data is shown in Figure 9. A curve covering most of the data is drawn for each spectra plot

for a given damping. Then, the corresponding spectra from different

plots for the same damping ratio are combined to form the final

plot. Finally, the corresponding near field and far field spectra

are combined to form the composite spectra for the site.

The resulting composite response spectra for the San Diego Navy

site is given in Figure 10. Curves are given for 0, 2, 5, 10, and

20% of critical damping.

Estimation of Yield and Ultimate Capacities

Upon completion of building selection, notes and construction

drawings obtained were used for the analysis of the selected buildings. These materials were reviewed. Rough sketches were made of

typical plans and elevations to determine the primary lateral forceresisting system or systems of each building.

The weight of the building is approximated by assuming unit

loads for roof framing, floor framing, walls, actual live loads

(if any), and other miscellaneous items. The base shear capacities are then estimated for each direction of motion, viz.,

transverse and longitudinal. The procedures used for computing

the base shear capacities varied for the different types of

bui ldings

.

To facilitate the computation of the yield and ultimate capacities from the primary lateral force-resisting elements of a building

the following descriptions of the failure sequence of these elements

in different buildings are given below.

The yield capacity of a building is defined as the lateral

force required to cause the yielding of the most rigid lateral

force-resisting element in the building. By contrast, the ultimate

capacity of a building is defined as the lateral force required

to cause yielding initiation of the most flexible lateral forceresisting element of the building.

For example, a steel building with a lateral force-resisting

system consisting of infill brick walls and X-brace may behave

as follows. The brick walls and the X-braces may work together to resist the seismic forces until cracking of the brick walls is initiated. Then the X-braces and columns (only after the yielding

of the X-braces) will take more and more of the seismic loading

until they fail. For concrete frame buildings with shear walls,

the shear walls will resist most of the seismic loading until they

start to yield (crack). Thereafter, the frame will start to resist

a small portion of the loading. Wooden frame buildings will behave

like to concrete frame and shear wall buildings.

General descriptions of the procedures of estimating the yield

and ultimate capacities of steel, concrete, wooden, and masonry

buildings are given in reference 14.

35-10
After computing the weight and shear capacities for the

buildings, the base shear coefficients are then computed by dividing

the base shear capacity by the weight of the building. The base

shear capacity coefficient is then converted to a spectral acceleration capacity by multiplying by a constant. This constant is 1.0,

Es timation of Natural Periods

The natural period for the buildings was determined by using

the theoretical formula for a single degree-of-freedom system, using

the Uniform Building Code formula modified by a constant, assuming

a spectral displacement, or using a computer program developed by

CEL for multi-story buildings (CEL8).

The natural period for a single degree-of-freedom system at yield can be computed from the formula

2 it ( sec .

)

(8)

where m seismic mass, the mass of the building above the

midheight of the first story

k = stiffness of the building in the direction considered

by summing the contribution from the lateral forceresisting elements

For design purposes, the natural period of the building is often

from

0.05 h

r

( sec .

)

where h

n

= height of the building (ft.)

(9)

D = base dimension of the building in the direction

considered (ft.)

Note that this equation contains no stiffness or mass terms.

It was derived empirically from the small vibration data for concrete frame and steel buildings and for use with the UBC criteria.

JLt generally gives natural periods that are lower than the actual

o provide conservative design base shear coefficients according

:o the UBC seismic criteria. For the rapid seismic analysis, this

ormula is modified by a constant depending on the building con- duction to account for variation in stiffness and weight per

nit volume of the particular type of building.

The spectral displacement is, related to natural period T

jind spectral acceleration capacity S for a single-degree-ofrreedom system according to the following formula:

S d 9.79 r S a (in.) (10)

where = spectral displacement (in.)

T = natural period (sec.)

S a spectral acceleration (g's)

quation (10) is useful for checking the assumed natural period for

building. If is known, the natural period can be determined

y assuming a spectral displacement from

35-11
For multi-story buildings, CEL has developed a computer program, CEL8, for computing fundamental period and mode shape iteratively. The ductility factor, and masses and stiffness at the

various floor levels are input into the program and the program

does the rest. A sample output from the program is given as Table 5.

where V = ductibility factor, the ratio of the ultimate to yield deflections

S a ^

= spectral acceleration capacity at yield

S au

= spectral acceleration capacity at ultimate

By selecting an appropriate P and knowing S ' and S a ^, the

T

u /T ratio can be determined. However, the earthquake damage

would be underestimated if the assumed ductibility is larger than

that can be developed by the structural system.

Procedures for using the formulas and computer program presented above for steel, concrete, wood and masonry buildings are described in reference 14.

For a more precise computation of the natural period of the

building, CEL has developed a computer program CEL7 for computing

the equivalent stiffness of shear wall with cut-outs.

Briefly, the Young's modulus, thickness, configuration, and dimen

sions of the piers are input the program. The program then computes

the equivalent stiffness for the combinations specified. An example

output from the program is given in Table 4.

The individual stiffness of the shear walls in each direction

can then be added and used in program CEL8 to determine the natural

period of the building. An example output from the program for a 6-story building is given in Table 5.

Overestimating the natural period may have a completely different effect on the damage estimate, depending on the natural period

of the building and the shape of the response spectra. If the estimated natural period is less than 0.3 sec. (0.15 sec. for rock

sites), the predicted acceleration demand and hence damage on the

building would be greater than the actual. If the estimated period

is greater than 0.3 sec., the predicted damage would be less than

actual. The converse of the above holds for underestimating the

natural period.

Estimation of Damage From Response Spectra

To estimate the amount of damage a building experiences from an earthquake, damage must first be defined. Until the yield capacity

of the building is reached, damage is assumed to be zero and the

ductility factor equal to one. When the ultimate capacity is reached,

damage is assumed to be 100%. For intermediate values of capacity

the assessment of damage is assumed that the damage varies linearly

35-12
between the yield limit, S

a y, and ultimate limit, Sau as shown in Figure 11.

Another assumption that must be made to estimate damage is the

amount of damping applied to the response of the structure. Damping

Ls related to a form of internal energy absorption of the building

iuring its response to the earthquake ground motions. Damping

Is assumed to be a constant up to the yield capacity of the building.

\bove the yield capacity, the energy absorption capacity of the

)uilding increases because of inelastic response. The assumed damping is increased to reflect the increase in energy absorption. The

assumed damping values for the different types of buildings are

>iven in Table 1. Moreover, damping was assumed to vary linearly

>etween the yield and ultimate limits of the building.

The NAVFAC criteria earthquake response spectra for 0, 2, 5,

.0 , and 20% of critical damping for San Diego, California Naval

Station used for determining the damage estimates were presented

.n Figure 10.

The procedure for estimating damage is based on the reconciliition of the building in relation to the period and damping. The

>rocedure is illustrated graphically in Figure 11. The capacities

•f the building are identified by the open circles. The correspondng demands for the site are identified by the black dots. The

ntersection of the lines drawn through the two sets of points

letermines the estimated damage, 50.5%.

For each building, the damage is computed for each of the

>rincipal directions of motion, longitudinal and transverse. To letermine the combined damage for the two directions, it is assumed

hat one-third of the building depends on each direction of lateral

esistance and the remaining one-third depends on both directions

lor lateral resistance. That is, if a structural element requires

iboth directions for lateral resistance is damaged by one direction

>f motion, it is also damaged in the other direction. The proedure takes two-thirds of the damage of the more critical direcion than one-third of the damages are 60% and 30% in the two

>rincipal directions, the combined damage is 50%.

To reduce the amount of time required to determine damage

stimates, CEL has developed a computer program CEL9 for determining

he damage estimates. The response spectra is digitized and input

into the program together with the building identifications, costs,

lamping values, natural periods, and capacities. The program then

omputes the damage estimate for the NAVFAC criteria. In addition,

he program computes the damage estimates for maximum ground accelrations between 0.05 and 0.50g increments.

The replacement cost of the selected buildings was obtained

jfrom the inventory listed for the installation generated by the

omputer program at FACSO.

The damage cost is computed by multiplying the replacement

ost of the building by the precent damage. An example computer

rintout of the damage estimate of a seven-story building is hown in Table 6.

35-13
Accuracy

As mentioned earlier in the introduct ion, the purpose of the

rapid analysis procedure is to determine earthquake damage estimates. Approximations were made at each step of the analysis.

These approximations have been added and multiplied into the

results. Theoretically, large discrepancy between the predicted

and actual values may occur. However, checks were made at each

step of the procedure to minimize potential inconsistencies.

These checks were based mainly on engineering judgment and included

in comparisons between different buildings and comparisons with dat

from other studies. A discussion of the accuracy of the various

segments of the procedure follows.

Response Spectra

In our opinion, the response spectra developed for San Diego

Navy site are reasonable estimates of the ground-motion effects

from the 25-year earthquakes.

Variation in Site Characteristics

The use of about twenty earthquake response spectra records

from different sites to develop the response spectra should cover

the possible effects that the variation in the local soil characteristics would be no greater than the uncertainties associated wit

the earthquake criteria and analysis criteria.

Structural Analysis

Approximate methods were used to determine the capacity limits

of the buildings. Only the primary lateral force-resisting element were analyzed, assuming that the supporting elements and secondary

elements were adequate. Stress and deformation capacities had to be assumed, but the actual values can have a wide range of variation. Steel has the smallest range of variation, but it can be

about 20%. By contrast, wood, concrete and masonry have a large

range of variation due mainly to difference in age, workmanship

and original quality. Furthermore, capacities of connections can

have a wide range of variation. In some cases, the exact details

of the connections were not available. In addition, non-structural

elements contribute to the lateral force-resisting capability of

the building. Although estimates can be made for the capacities

of non-structure members, their range of variation is very wide.

Individually, the capacities of the elements that make up the

lateral force-resisting system of the building can have a large

range of variation. However, when combined, the conservatism or redundancy in some of the elements will balance some of the weakness and inadequacies in others. As a result, the overall building

may be represented adequately by the aproximate model. Even a rigorous analysis is subjected to many uncertainties.

The computed natural period may vary greatly, depending on the

assumptions made in the structural analysis.

The damping values used in the analysis are fairly representative of the state-of-the-art. In the inelastic range, effective

damping may be higher than the values used. However, insufficient

data is available to justify using higher values.

35-14
Damage Criteria

Actual damage costs for buildings from earthquakes are difficult to define. In addition to analytical evaluations, social and

administrative decisions must also be considered. For example,

suppose an old brick building develops some cracks during a moderate earthquake. A subsequent investigation indicated some uncertainty about the safe future use of the building. Three possible administrative decisions for the building are:

(1) Ignore the potential hazard and continue using the building

for a number of years (i.e., the cracks are not serious enough to justify repairs).

(2) Rehabilitate the building to eliminate the potential

safety hazards (repair the cracks and add some additional bracing

and reinforcements).

(3) Demolish the building and build a new one to replace it.

In case (1), the damage cost is zero. In case (2), it may be

30% of the replacement cost. In case (3), it is 100% plus the denolition cost. The results of the rapid seismic analysis procedure

do not consider such administrative decision, but would give a damage estimate of about 50%.

Other Computer Programs

The other Computer Programs developed by CEL for the rapid seisnic analysis procedure include CEL2, CEL6 , and CEL10.

Program CEL2 can be used to determine the maximum ground

accelerations from the various empirical acceleration attenuation

/ersus earthquake magnitude, site condition, and distance to fault

formulas built into the program. It can be used to get an idea of

:he range of predicted accelerations for the site from earthquake

)n the surrounding faults.

Program CEL6 may be used to compute the stiffness and

-strengths for concrete masonry shear panels (piers) with cantilever and fixed conditions for a range of panel sizes with constant thickness.

J Program CEL10 can be used to check the correctness of hand

digitized response spectra data before it is input into computer

>rogram CEL9. It plots the digitized data using the CALCOMP

>lotter so that any errors made during the digitization can be

spotted visually and appropiate corrections made.

References

1. John A. Blume and Associates, "Seismic Study of Puget Sound

Naval Shipyard, Bremerton, Washington," Phases I through VI, A-E Contract N62474-71-C-4324 , for Western Division, Naval

Facilities Engineering Command, San Bruno, California, June

1974.

2. Naval Facilities Engineering Command, "Structural Engineering

Workshop at San Francisco, CA," prepared by John A. Blume and

Associates, Engineers, September 1973.

35-15
3. Civil Engineering Laboratory, Technical Memorandum M51-78-08:

Earthquake hazard reduction program, North Island, Naval Air

Station, San Diego, CA, by C. V. Chelapati, S. K. Takahashi,

and T. K. Lew, Port Hueneme, California, April, 1978.

4. Dorwick, D. J., Earthquake Resistant Design , John Wiley & Sons, New York (1977).

5. Weigel, R. C. (Ed.), Earthquake Engineering , Prentice-Hall,

Inc., Englewood Cliffs, New Jersey 1970.

6. Newmark, N. M. , and Rosenblueth, E., Fundamentals of Earth - quake Engineering , Prentice-Hall, Inc., Englewood Cliffs,

New Jersey, 1971.

7. Blume , J. A., Newmark, N. M. , And Corning, L. H. , "Design

of Multi-story Reinforced Concrete Building for Earthquake

Building for Earthquake Motions," Published by Portland

Cement Association, Chicago, Illinois, 1961.

8. Kiremidjian, A., "Probabilistic Hazard Mapping: Development

of Site Dependent Seismic Load Parameters," Ph.D Dissertation,

Stanford University (1977). Report 77-12, 655, University

Microfilms International, Ann Arbor, Michigan.

9. Hileman, J. A., G. R. Allen, and Nordquest, J. M. , "Seismicity

of Southern California Institute of Technology, Pasadena,

California, 1974.

10. "Report on MSF-UCEER Conference on Earthquake Engineering

Research, March," University Council for Earthquake Research,

Pasadena, California, May 1969.

11. Housner, G. W. , "Strength Ground Motion" (Chapter 4), "Design

Spectrum" (Chapter 5), Earthquake Engineering , Weigel, R. L.

,

Ed., Prentice-Hall, Inc., Englewood Cliffs, New Jersey, 1970.

12. Trifunac, M. D. , and Brady, A. G. , "On the Correlation of

Peak Acceleration of Strong Motion with Earthquake Magnitude

and Epicentral Distance," Procedures of U. S. Natural Conference on Earthquake Engineering - 1975. Earthquake Engineering Research Institute, June 1975, Ann Arbor, Michigan,

pp. 42-52 .

13. California Institute of Technology, "Strong Motion Earthquake

Accelergr ams

, " Vol. II and III. Earthquake Engineering

Research Laboratory, Pasadena, California (EERL71-50) , September

1971.

14. Civil Engineering Laboratory, Technical Memorandum M51-78-02,

"Rapid Seismic Analysis Procedure," by Lew, T. K. , and Takahashi,

S. K., Port Hueneme, California, April 1978.

35-16
TABLE 1. ASSUMED DAMPING VALUES

Building Type

Percent of Critical Damping

At Yield Limit At Ultimate Limit

Steel 2 5 Concrete 5 10 Wood 10 20 Masonry 5 10

35-17
FIGURE 1. FAULTS USED FOR ESTIMATING GROUND ACCELERATIONS AT

SITE OF NORTH ISLAND NAVAL AIR STATION.

All curves are for mean + one standard deviation

0| i 1 i I i l_

O I.O 2.0 3.0

Period.T (sec.)

FIGURE 2. DYNAMIC AMPLIFICATION FACTORS (84 PERCENTILE, X = 5%)

FOR DIFFERENT SOIL CONDITIONS (AFTER REF . 8).

35-18
FIGURE 3. INTERVAL RECURRENCE CURVES FOR EACH OF THE AREAS

SHOWN IN FIGURE 3.10. NOTE THAT THE ORDINATE SCALES

ARE IDENTICAL FOR ALL CURVES EXCEPT FOR THE CURVESFOR THE SOUTHERN CALIFORNIA NETWORK AND NORTH MEXICO

(FROM REF. 9)

.

35-19
lilt

- -A

I ;

- L-2 2Jxi i

/

4-

;

-

-

1 8J.XI0

- /

IDEALIZED RELATION BETWEEN MAGNITUDE AND LENGTH OF SLIPPED FAULT

23456769 MAGNITUDEIdealized relation between length of slipped fault vs. Magnitude of earthquake.

Ma nltude Ltojth (rnitci)

S.S I0C0

8.5 530

S.0 190

7.5 70

7.0 25 65 9

6.0 5 55 3.4

5.0 2.1 4.5 40 0.SJ

3.0 0.33 20 0.14(73$") 10 0 0*270)

0 0 01*100")

INFERRED LENGTHS OF SLIPPED FAULT OF ACTUAL EARTHQUAKES

Earthquake Lcncth (nnlci)

Chile. I960 S.5-8.6 609 i

Alaska. 1964 8.4 450±

San rranci;co. 1906 6.2 250±

El C«ntro, 1910 7.1 40±

Bajt California, 1956 6.1 IS±

FIGURE 4. RELATIONSHIPS BETWEEN EARTHQUAKE MAGNITUDES AND LENGTH

OF SLIPPED FAULTS (FROM REF. 10 AND 11).

35-20
50

•4L

A

«(!«»)/

4 5 6 MAGNITUDE

I

*50

« 0,7C

9

0 ;•• o

1 0 0 O >

fa 4

0 o1

8- i

6 7

0

UJ

^ 20

I 10

o s

«* (1

. 0° X4—

o < 7*

po o coo Nsa

o ' o 0 o-J

0 |^Oj e

10 20 JO 40 50 60 70 SO

DISTANCE TO fMAT - MILIS

'IGURE 5. IDEALIZED RELATIONSHIPS BETWEEN MAGNITUDE, DURATION,

ACCELERATION AND DISTANCE FROM FAULT FOR STRONG

MOTION EARTHQUAKES (FROM REF. 10 and 11).

35-21
FIGURE 6. RELATIONSHIPS BETWEEN PEAK ACCELERATION AND DISTANCE

FROM SOURCE FOR MAGNITUDE 6.5 EARTHQUAKES (AFTER REF

12).

35-22
N - 6.S Distance - 6.2 Biles

IMPERIAL VALLEY EARTHQUAKE MAY 18. 1940 - 2037 PST

1 19001 40. 001.0 EL CENTRO SITE IMPERIAL VALLEY IRRIGATION DISTRICT COMP SOOE o PERK VALUES : ACCEL = 341 .7 CM/SEC/SEC VELOCITY = 33.4 CM/SEC OISPL = 10.9 CH -500 r

* ° G

j 1/2

5C0

m ... Ji! in

J I I I I ! 1-

0 10 20 30 40 50

TIME - SECONDS

M - 7.7 Distance • 2S.8 miles KERN COUNTY. CALIFORNIA. EARTHQUAKE JULY 21. 1952 - 0453 POT

IIAC04 52.002.0 TAFT LINCOLN SCHCCL TUNNEL COMP N21E o "E°K VALUES : ACCEL = 1S2.7 C^/SEC/SEC VELOCITY = -15.7 CM/SEC OISPL = -6.7 CM -250 r = o.ife <3

2/J 1/2

10 20 . 30 40 50

TIME - SECONDS

FIGURE 7. STRONG MOTION ACCELERATION RECORDS IN ALLUVIAL SITES

FROM LARGE EARTHQUAKES.

35-23
FIGURE 8. PROBABILITY VERSUS MAXIMUM GROUND ACCELERATION FOR

25 YEAR EARTHQUAKE, NAS , NORTH ISLAND, CALIFORNIA

(SAN ANDREAS FAULT.

35-24

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20

FIGURE 11. GRAPHICAL ILLUSTRATION OF DAMAGE ESTIMATES.

35-26
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35-28
Table 3. CONTINUED

»>ED!CT!OM OF SEISMIC SEVERITY FOR NAS.SAN OIECO SITE csriHiTiON or »iii<u« crouno acceleration it site xitm ro percent probapuit* or occurrence using monte carlo procedures EARTHQUAKES OCCURINS WITH EPICENTERS ON SAN ANDREAS FAULT SfjTEM

RANGE OF MAGNITUDE 7.0 rance or CISTiNCE TO FAULT 125.0 RANGE OF FOCAL DEPTH 6.0

CSl I. 00 CV» 0.00 PCL. ,90

t 0.00 1000 1 .000

2 .01 1000 1.000

I .02 I0O0 1 .000

S • OS 970 ,97«

" 5 .OR 7117 ,7S7

' A .05 SJT .117

... - T .06 178 .178

a .07 56 .056

• .OS 12 .012 — to • OR 0 0,000

ii .10 0 0.000 —

12 .11 0 0,000

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15 0 0.000

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22 .21 0 o.ooo — 2J .22 0 0.000

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26 .25 0 0.000 .... ?T .26 0 0.000

2S .27 0 0.000

29 -

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10% probability of occunenc* (0.066 or 0.07g)

35

35-29
Table 4. COMPUTATION OF EQUIVALENT STIFFNESS AND RIGIDITY

CHARACTERISTICS OF WALLS WITH OPENINGS

/TH = 12 in.

r»'!"tL">o3iij" - fV;.!v>i t 'it stffrKFM.or »>*

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'35-30
Table 5. COMPUTATION OF FIRST MODE SHAPE, NATURAL PERIOD,

SPECTRAL ACCELERATION CAPACITY. (CEL8)

Hit 8*N DUGO BIOS 763 BCd TRANSVERSE Cu«PUT»TIG* JF FIRST NATURAL "UD£ AND FREQUENCY AND PARTICIPATION FACTOR

M*L no.

«E!6«T

NASS (IN > STIFFNESS

IA».;«a

FFCJ« »D«ON

t

• IP) 600 600 600 too 600 600 500

UN) 1. 551 1.553 1.553 1.553 1 .553 1.553 1 ,i94 »P/IM 12I90U i ? l <»oo 121900 121900 609SO 60lbu 48200

A) .112 .214 .310 ,<I0« .544 • 6U1 .700

A»2> 1.715E-01 J. 1626-01 4.67SE-01 6.19«t-Pl 8.343E-01 9.S33E-01 9.0S8E-01

A*2) «.3Jettoo «. 166E*00 3.6S0Etdu 3. smtoo 2.72SE*00 1 .»«9E«00 9.05»t-0

1

»?) 3.SS9E-US 3.SS9E-05 3.SS9E-OS 3.559£.05 3.559E-05 3.559E-0S 3.5591-05

An?) 3.559E-0S 6.977E-05 1 .012E-0H 1 .2866-04 1.733E-04 2.043E-04 2.231E-04 »< ) .11? .214 .318 ,404 .544 .»«! .700 CIRCULAR FRt'JUENC' NATURAL F«tJu£NC»

56.02 RAJIANS/SEL

8.92 OCLES/SEC

.11 SECuNDb

3137.491

nOk«II HtV FIRST «ODt »»APfe

InDE*

iNOf » I^rf

INPE"

I'Ntf.t I InUE<

InDEA

01SfL»CE«EM

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'OTal SHE*" a«£A 20660. 000 SU In HELD STRESS .070 KSI

ILT1«AIE ST«tSS ,10o *SI

JHST FLOOR Sh£ar CAPACITY Al YIELD (*1P) IRST FLOUR SwtAR CAPACITY A 1 ULTIMATE C« I P

)

>DCTILITY FACTOR

1046.20') 2066.000

2.0

PAWTlCIRAlMn FACTOR, FIRST "ODE 2.929

TOTAL «EI&nT OF BU1LD1N6 9100.0 "IPS

FIRST "uot BASE S-EAR FACtUR 1.248 SPECTRAL YIELD CAPACITY .««0 OS PERIOD AT YIELD SPECTRAL UL T I h a T fc CAPACITY .629 S»S PERIOD AT ULUNA1E

.11 SECONDS

.13 SECONDS

35-31
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35-32
36

US-SOUTHEAST ASIA SYMPOSIUM ON ENGINEERING FOR

NATURAL HAZARDS PROTECTION

A. H-S. Ang

Civil Engineering

Illinois at Urbana-Champaign

Department

University

of

of

Prepared for

National Science Foundation

US-Southeast Asia Symposium on Engineering

for Natural Hazards Protection

he need for protection of man and manmade structures and facililes against the extreme forces of natural hazards is worldwide.

is need is particularly pronounced in the region of Southeast

sia where typhoons, monsoons, severe thunderstorms, and earthquakes

re yearly occurrences that regularly pose serious threats to lives

nd material properties. To mitigate these forces and provide adeuate protection within the developing economies of the countries

a the region are real challenges for the local engineering

ommun i t i e s

.

Some aspects of the approach to the engineering design of

tructures and systems to withstand the forces of natural hazards

re common irrespective of geographic regions. In this regard, the

;cent developments in the United States, particularly in the areas

E earthquake and wind engineering, would be of interest and benet to the engineers and research of the Southeast Asian Countries,

vertheless, each region in the world may have special problems

culiar to the region; moreover, because of the frequency

pd regularity of major hazards in Southeast Asia, the study of

jxtreme natural hazards in this region would be of special interest

d engineering researchers, particularly those working in the areas

E earthquake and strong wind. Certain countries in this region

ie directly within the circum-Pacif ic belt of active seismic

j:tivity. Also, this region is regularly subjected to severe

/phoons and heavy rainstorms, the magnitude or intensity of which

re some of the most severe in the world. There are also areas in

lis region that are exposed to tidal waves (which may be earthquakeiduced) and volcanic hazards. Flooding, as well as landslides, are

"equent results of these natural phenonema.

The Symposium provided a forum for delegates from the U.S. and

ie Southeast Asian countries to review the current state of tech-

>logies that have recently been developed for the design of

jicilities and structures to resist the extreme forces of natural

izards, with specific reference to earthquakes and strong wind,

lis Symposium also provided the opportunity to review the current

search and development efforts, both in the United States and

•untries of Southeast Asia, as well as to identify those problems

at are unique to the region.

The technical program consisted of two days of formal paper

esentations, and one and one-half days of workshop discussions.

was the purpose of the workshop to define and identify specific

pics related to natural hazards that are most promising for future

llaborative efforts between the U.S. and countries* in Southeast

ia, or among countries in Southeast Asia.

Aside from the U.S. and the Philippines, which had the largest

legations to the Symposium, there were delegations also from Japan,

stralia, New Zealand, Hong Kong, Thailand, and Singapore. Invita-

|ons to participate in the symposium were sent to Indonesia, Malaya, and South Korea; however, these countries were not able to send

legations to Manila, due largely to the short lead time available

fetween March and September 1977). In any future symposium of

|is type, every effort should be made to get representation from

ese other countries.

The participation of the U.S. delegation at the Manila Symposium

3 made possible through the support of the National Science Fountion under Grant No. INT 77-15418 to the Department of Civil

36-1
Engineering, University of Illinois at Urbana-Champaign. The

initial incentive for a symposium in Southeast Asia was developed

through discussions with Dr. S. C. Liu of NSF-RANN and Dr. A. Holt

of NSF-Internat ional Programs Division. Their personal encouragement and support during the organizational ph ase of the symposium

are gratefully acknowledged.

Much of the credit for the successful organization of the Symposium in Manila must go to the local organizers under the able

leadership of Dr. E. Tabujara of the University of Philippines and

R. Venturina of the National Science Development Board of the

Philippines. Also, the co-sponsorship and financial support of the

National Science Development Board and the University of the Philip

pines, as well as the contributions from other governmental and

private agencies in the Philippines, all served to insure the suc- cess of the symposium in Manila. To our Philippine hosts and

organizers, the foreign delegates are grateful for their kind hospi

tality, and for the magnificent and beautiful facilities provided

for the conference.

The Symposium benefited immeasureably from the participation

and personal support of the following Philippine dignitaries: Dr.

M. S. Magno, Chairman of the National Science Development Board;

Dr. 0. D. Corpuz, President of the University of the Philippines

System; and Hon. A. L. Juinio, Secretary of Public Works, Transportation and Communications, and Dean of the College of Engineering at the University of the Philippines.

The technical contents of a symposium of this type, of course, depend on the contributions of all the participants; the efforts

of all the delegates, guests, and observers are greatly appreciated

In particular, the efforts of the chairmen of the various working

groups deserve special acknowledgment; indeed, they deserve the

credit for the successful conduct of the research workshop sessions

and for their respective working group reports (reprinted here).

The Research Workshop

The Research Workshop consisted of one-and-a-half days of discussions by small groups of delegates. The objective was to identify and define specific topics for research that would be of

mutual interest and potential benefit to the countries involved.

Eventually, it is hoped that these will form the basis for future

collaborative research programs between the countries represented

at the Symposium. Indeed, the extent to which the topics defined

at this Symposium can be developed to full research projects will

eventually prove to be the real significance of the Symposium.

The discussions were divided into five working groups; the

working groups and their reports are summarized in the following

sections of this paper.

Working Group El: Earthquake Response and Design

Chairmen: Dr. William J. Hall

Prof. Felisberto G. L. Reyes

Rapporteur: Prof. Marino M. Mena

Report of Group El

With recognition that the need for protection against the

extreme forces and deformations of natural hazards, such as

36-2
arthquake, is worldwide, Working Group El undertook an examinaion of a set of study topics that we believe would help mitigate

hese effects in providing protection for the population and in

educing economic loss. In general this objective can be met hrough improved design and construction practices, especially

s they lead to more costef fective structures. To this end, it hould be noted that the primary purpose of the Symposium was

to identify and define specific study areas or topics for uture collaborative research among the countries represented

n this Symposium."

At the outset it is worth noting that the topics selected

erein for specific comments are major areas of concern not

nly to Southeast Asian countries but to other parts of the

orld as well, including the United States and Japan. Within

his context then the formulation of a comprehensive research

nd development study should involve the cooperative effort of

ill countries concerned and should have as its goal to develop

nprovements in technology for all.

After a brief review of the status of earthquake engieering design/analysis/construction practice as reflected in

|he papers presented at the Symposium and in the discussion

t the working group meeting, it was agreed that the following

en topical headings incorporate the areas of research needed

io improve our earthquake resistant design and construction

act ice

(1) Design Philosophy, Codes and Standardization

(2) Material Quality, Control, and Construction Practices

(3) Non-Engineered Structures — Low-Cost, Low-Rise

(4) Engineered Structures — Building Framing Systems,

Ductility, and Non-Structural Elements

(5) Evaluation of the Adequacy of Existing Buildings

(6) Information Base, Application and Interpretation

(7) Soil-Structure Interaction

(8) Bridges

(9) Utilities

(10) Dams and Reservoirs

The working group agreed that the first six topics deserved

oecial interpretative description since they appear to be amen-

)le to cooperative effort within a reasonable time. This was not

:ant to mean that the last four items were deemed to be unimporant , for indeed soil-structure interaction is intrinsically a part

p almost all of the first five items; since it is under intensive

!ivestigation in several countries and the research involves large

founts of time and money, it suffices to introduce synthesis and

bview at the appropriate time rather than to undertake detailed

udies at this time as a part of a cooperative study.

With regard to bridges, utilities and dams (Items 8, 9, and

>), it was noted that all of these areas are under study at the

went in such countries as Japan and the United States. In several

ars, it may be advantageous to examine these topics in further

tail for possible cooperative effort. It was noted that in the

se of bridges, for example, studies currently underway may sugst low-cost techniques for upgrading bridge resistance and

ch information should be brought to the attention of the coopating countries when available in view of the importance of

ansportation systems to the economy and to recovery.

36-3
Design Philosophy, Codes and Standardization

In recent years many new national building codes and guides

for resisting seismic effects have been developed. These codes

commonly include provisions for estimating the seismic forces,

techniques for providing seismic resistance, and specifications

pertaining to material strength, allowable stress, and deformation.

These codes differ in part as might be expected. In the interest

of providing a base for review or upgrading such codes from time

to time, it is believed that it would be desirable to prepare some position statements concerning the philosophy behind the development

of such codes or design guides including, for example, such items

as selection of the hazards, analysis and design techniques for

loading and resistance, and assessment of the associated risk and

margin of safety. It is believed that it would be desirable on

a general basis (not in detailed form) to compare and summarize

differences and similarities. It is believed that this effort

would provide a valuable basis for code modifications in the

future

.

Another ambitious program that could be of value would be the

preparation of a design guide that would include specific recommendations as to proportioning and detailing of structural elements.

Material Quality, Control, and Construction Practices

The state-of-the-art in local construction practices is usually

based on past experience and rule of thumb. The majority of severe damage sustained by low-rise, low-income housing due to earthquakes

could be mitigated by more explicit and simple design and construction guidelines.

Particular attention is drawn to material quality, control, and

construction practices. Quality assurance programs and procedures

in construction starting from the manufacture through the inspection,

and up to the installation of the materials, should be established

in a simple and systematic manner.

The development of indigenous construction materials that will

perform well during an earthquake, such as the use of bamboo fiber,

coconut coir, abaca fiber, etc., in roofing, wall and ceiling boards,

should be undertaken with the end view of increased strength and

reduced weight and cost.

Techniques of strengthening adobe and soil cement block houses

against earthquake disturbance by providing tension reinforcement,

banding or membrane continuity in the form of indigenous materials

should be developed, with particular attention given to anchorage

and connections. Connections and anchorage devices that increase

the safety in critical joints and sections in structures that

exhibit frequent failures during an earthquake should be studied

and developed. It may be possible to develop simple guides to permit installation of some of these strengthening techniques in

existing buildings and housing.

It is believed that studies of the type outlined specifically

in Topic 2, but also Topic 5 generally, could be carried out in

building research laboratories in the countries concerned or as a part of cooperative programs. The investigations should be carefully

planned so as to expedite adoption in practice.

36-4
Non-Engineered Structures — Low-Cost, Low-Rise

Many aspects of research required for improvement of the earth-

|uake resistance of a non-engineering structure are contained in

topic 2. However, in addition to the specific detailed topics

mtlined in Topic 2, it is believed that there is a need for an xamination of the structural system as a whole and the development

itnd simplification of procedures for'use by practitioners to expedite

valuation of the seismic resistance of such structural systems,

n view of the large number of such structures throughout the world,

t should be obvious that any adopted improvements should lead to

iajor economic savings and increased public safety.

Engineered Structures — Building Framing Systems, Ductility and

Non-Structural Elements

In the interest of improving design practices for engineering

tructures throughout the world, it is recommended that intensive

esearch be continued in the development of resistant framing

ystems (for all materials) with the aim of improving their

apability of maintaining strength throughout deformation without

egradation. Structural elements should be able to resist both

n-plane or out-of-plane loadings.

As a part of these studies, there should be encouragement in

11 cooperative countries to investigate new materials, and cominations of new materials in conjunction with new types of framing

ystems. One goal is to develop techniques for evaluating and

emonstrating the adequacy of new systems so that they can be acceped for use in practice.

Recent earthquake damage assessment studies have shown clearly

he need for more attention to the design and attachment of nontructural and secondary elements in buildings. This subject needs

ntensive study and the development of guidelines to permit improveent in design and construction.

Evaluation of the Adequacy of Existing Buildings

The evaluation of the adequacy of existing buildings

s composed of two general areas: (a) the evaluation of the strucjres following an earthquake, and (b) the evaluation of structures

5 part of the general assessment of adequacy at periodic intervals.

With regard to the former item, studies are need of the

rocedures for assessing physical adequacy of the structures as

;11 as of legal responsibilities involved in assessing adequacy

id/or repairability of the damaged buildings. At the same time

iere needs to be an assessment as to whether or not there should be

mandate that a building which externally does not appear damaged

lould be examined to confirm that its structural adequacy is not

apaired.

With regard to the second subtopic there needs to be a study

: methods that might be employed by engineering firms and governintal bodies to ascertain the adequacy of existing buildings within

given area to resist seismic loadings. It is envisioned that

ich investigation should be contemplated for structures that are,

i general, old or possibly constructed at a time when different

iterials and techniques were employed and considered desirable,

long the techniques that might be considered are non-destructive

^sting and certain forms of analysis coupled with selective testing.

36-5
Information Base, Application and Interpretation

In Southeast Asia there is a general lack of comprehensive and

centralized reference sources of technical information pertaining

to engineering design for natural hazards, analysis procedures,

construction techniques, and material properties. In contrast the

United States and Japanese depositories are replete, but not necessarily complete, with such information. It is recommended

that designated representatives of countries involved in cooperative programs of development make known generally to all participants major sources of information in order that they can procure

those of most interest if they so desire. In addition to journals

or proceedings, the names and addresses of cognizant organizations

should be provided so that continual updating can be maintained.

It is conceivable that in some cases at the national level, it may be found expedient to collect and collate abstracts of significant existing information for purposes of review and evaluation by

appropriate groups. A depository for such abstracts in the Philippines could be The Technology Resource Center. To save effort and

resources, if certain countries undertake this effort they should

make their collection known to other cognizant countries to prevent

duplication of effort.

To disseminate information and concepts rapidly as a result

of research and development activities, and to expidite information

exchange generally between the co-perating groups, it is suggested

that symposia be held in other cities throughout Southeast Asia.

Such symposia, if structured carefully in advance, could prove to be highly desirable and beneficial in terms of synthesizing research

and development information and in expediting technology transfer.

To insure that these major symposia will be structured properly,

it is suggested that certain investigators meet as appropriate at intermediate periods to compare and study results and to guide

preparation of topics that would be presented in the symposia.

It is recommended that each Southeast Asia country involved in

the cooperative effort urge their respective policy-making bodies to develop and systematize a plan to expedite research efforts and applications to practice.

To implement this plan, the information base described in subItem No. 1 above is necessary but there is also an implication of a more serious commitment of funds for qualified personnel and equipment to achieve the goals over some planned time period. It is i recognized that a major portion of the technology transfer should

be structured to be of use in accordance with specific engineering

requirements of any particular country.

Another item that should not be overlooked pertains to the

social aspect, such as disaster planning and rehabilitation following an earthquake. The efforts in this field need to be planned

separately but coordinated with the technical studies.

Working Group E2: Ground Motion And Seismic Risk

Chairmen: Dr.

Dr. Rapporteur: Ms.

Joseph Penzien

Salvador F. Reys

Perlita Morales

36-6
Report of Group E2

Working group E2, Ground Motion and Seismic Risk, identified

six areas where cooperation is recommended. The first four of

these areas — (1) Seismic Zoning Map of the Philippines, Indonesia,

and other Southeast Asian countries, (2) Intensity Attenuation Laws,

(3) Prediction and Characteristics of Bedrock Ground Motions,

and (4) Effects of Local Soil Conditions on Ground-Motion

Characteristics — are basic research studies of mutual value to all seismic regions of the world. The fifth area, Design Basic

Earthquakes, would bring together the results of the first four

to provide a basis for establishing design earthquakes that could

be used in designing important structures such as nuclear power

plants, high-rise buildings, and dams in the Southeast Asian countries. The sixth area, Institutional Approach to Reserach Information Dissemination, is directed towards stimulating research in the

Southeast Asian countries and towards collecting and disseminating

research results. The following is a brief outline of each of

the six areas:

Seismic Zoning Map

Statement of Research: This research would concentrate on collecting, collating, and analyzing significant data on seismic

events, such as location of epicenters, magnitudes, frequencies of occurrence, epicentral distances, focal depths, intensities, etc.,

and on correlating the results with geologic conditions.

Need for Research: There is a great need for seismic zoning

maps in the Southeast Asian developing countries that will define

seismic intensity levels for use in the design of important structures such as nuclear power plants, high-rise buildings, dams, etc.

Description of Research: Statistical analyses of basic seismological data would be made to establish seismic intensity levels

in probabilistic form for the Southeast Asian countries, with special

emphasis given to the metropolitan areas. Seismic intensity maps

would be developed for use in engineering design.

Cooperative Efforts: Since the results of this research are

basic in nature, they would be of value to all seismic regions of the

world. Therefore, cooperation among the Southeast Asian countries,

Japan, and the United States in the overall effort is recommended.

The individual efforts should be coordinated with educational and

research institutions and with professional organizations within

the participating countries. This cooperative program could be

carried out with a minimum of financial support and equipment.

Intensity Attenuation Laws

Statement of Research: This research would focus on the

development of theoretical and empirical laws for the statistical

prediction of ground motion intensities during an earthquake of

<nown magnitude and location.

Need for Research: Once the probable locations of epicenters

ire known, attenuation laws are needed to predict intensities of

notion at particular sites. Therefore, there is a need for

'1) development of attenuation laws for regions where such laws

ire not available, and (2) improvement of exisiting attenuation

Laws for regions where they have been developed previously.

36-7
Description of Research: Basic research is needed in two directions, namely (1) collection and exchange of recorded and

observed data from past earthquakes which are now available and

documented in participating countries, and (2) development of

equations relating ground motion intensity to earthquake magnitude,

distance, and other significant parameters. Information on soil

and geologic conditions at recording stations should be part of this

study. Various intensity scales, such as the MM scale, the RF scale,

and maximum ground motion scales should be explored depending on the

avaiability of data and engineering needs. Relationships between the

various intensity scales should be established to convert available

data into a form most useful for engineering design purposes. The

statistical scatter of data about the proposed attenuation laws

should be established as it is significant in determining levels of

seismic risk. Attention should also be given to: (1) the frequency

dependency of attenuation rates, (2) defining the most appropriate

distance parameter for use in attenuation laws, and (3) influence

of source mechanism parameters on attenuation rates.

Cooperative Efforts: Cooperation of participating countries

is essential in the compilation of statistical data that form the

basis for attenuation law studies. Comparison of results and coor- dination of efforts are believed to be vital for the success of

such a program.

Prediction and Characteristics of Bedrock Ground Motions

Statement of Research: This research is aimed at (1) establishing seismic risk maps in terms of ground motion parameters

of engineering interest, (2) characterizing strong ground motions

at bedrock level, and (3) developing appropriate stochastic models

for bedrock ground motions.

Need for Research: This research will generate information

much needed by engineers in establishing seismic load levels and

characterizations of seismic ground motions.

Description of Research: There is a great need for a comprehensive study aimed at establishing seismic risk maps in terms of

ground motion parameters of engineering interest. At present, all

available information is in the form of qualitative intensities.

There is a need to relate this information to such parameters as ground acceleration, ground velocity, duration, and other strong

motion parameters. These relationships should be established for

all major local seismic provinces and the similarities of data and

results for these provinces should be compared with comparable

information available in Japan, the United States, and other countries. In time, further information will become available from the

strongmotion instrumentation now being installed. Such information

will supplement existing data bases and should be incorporated into

a more comprehensive study of seismic risk and characterization

of strong earthquakes.

In characterizing strong ground motions for engineering

purposes, three parameters, namely, intensity, dominant ground

period, and duration are important as they greatly influence structural response. Duration is particularly important for non-elastic

systems. Ideally, information about the joint likelihood of occurrence of these three parameters should be developed. In addition, information on the variations of ground motion intensity

and its frequency content with time should be established. Such

information is needed in developing appropriate stochastic models

for three-dimensional ground motions.

36-8
Cooperative Efforts: Cooperative studies will be especially

isffective for regions where seismic zoning can be achieved only

jin terms of qualitative intensity scales. The results of studies

iffor these regions can be compared with results worked out in terms

pf engineering parameters for other regions having similar geological conditions. Calibration between such regions will provide

[quantitative reasoning for other regions not having instrumental

liata and will provide a more uniform international basis for seiskiic zoning. Cooperative studies will also be highly desirable

Ln the field of statistical analysis and data supply for characterizing earthquake ground motions and establishing appropriate

jnodels. Of greaest importance is the development of strong-motion

seismograph networks on a uniform technical basis as well as the

analysis of data obtained using a common methodology.

Effects of Local Soil Conditions on Ground Motion

Statement of Research: This research has two objectives:

'1) to collect subsoil information; e.g., depths to bedrock and

)hysical properties of soils, sufficient for establishing prelimiary soil factors for design purposes in urbanized areas, and

2) to provide a data base which can be utilized for studying

otions in alluvial ground overlying bedrock and for correlating

he results of such studies with available recorded strong ground

otions

.

Need for Research: The effects of earthquakes on engineering

Itructures in different locations may depend significantly on the

onditions of the underlying foundations soils; therefore, the

ffect of local soils on the characteristics of ground motion

! ;reatly needs investigation. Unfortunately, in most regions of

,outheast Asia, information on local soil conditions is severely

I acking

.

Description of Research: Information on local soil conditions

hroughout Southeast Asia should be collected and organized (work

long this line is now being carried out in Metro-Manila). This

nformation should include depths to bedrock, depths to the water

able, and the physical properties of the overlying soils as obtained

y various engineering and geophysical techniques. Since much conroversy exists regarding the effects of local soil conditions on round motion, correlation studies should be carried out comparing

redicted ground motion characteristics with measured ground motion

haracterist ics . Improved methods for assessing local soil effects

hould be developed.

Cooperative Efforts: Since this research is basic in nature,

he results would be of value for all seismic regions of the world;

herefore, cooperation in the overall effort is recommended among

tie Southeast Asian countries, Japan, and the United States.

Design Basis Earthquakes

Statement of Cooperative Effort: This effort is aimed at stablishing design basis earthquakes to be used in the design

jf structures, such as industrial facilities, nuclear power

ants, high-rise buildings, residential buildings, dams, etc.,

o be located in the Southeast Asian countries.

Need for Research: Design basis earthquakes are needed so

nat engineering designs are consistent with acceptable safety

svels, economic constraints, professional practice, and other

?rtinent local factors.

36-9
Description of Research: The development of design basis

earthquakes should follow two approaches, namely (1) establishing

appropriate values for the design lateral force coefficient roughly

comparable to the Z-factor in the Uniform Building Code, and

(2) establishing appropriate response spectrum curves for modal

analyses and accelerograms for use in time-history analyses.

The effect due to local soil conditions should be separated

from the lateral force coefficient mentioned above and should be

treated separately. Maps should be established showing the lateral

force coefficient to be used for any site in Southeast Asia.

The design basis earthquake defined through response spectrum

curves or an accelerogram, which would be used for more important

structures, should be established after full consideration of plate

techtonic theory, available geological and seismological data,

attenuation laws, local soil conditions, etc.

Public policy factors applicable to local regions (social,

economic, political, legal, etc.) should also be considered when

setting the design basis earthquake, whether it be expressed in

terms of the lateral force coefficient, response spectra, or an accelerogram of ground motion.

Cooperative Efforts: This work would be carried out primarily

by individuals and groups in the Southeast Asian countries working

cooperatively; however, assistance and cooperation from individuals

and groups in Japan and the United States are highly desirable due

to their previous experience in this area of activity. The overall

effort would be of benefit to all participating countries.

Institutional Approach to Research and Information

Statement of Problem: Answers are needed to the following

questions: (1) How can all information pertinent to natural hazards

mitigation be effectively collected and disseminated throughout the

Southeast Asia region?, (2) How can this information be effectively

used in research and engineering practice?, (3) How can research

be stimulated and engineering practice be improved in the Southeast

Asian countries?, (4) How can engineers and scientists be given

opportunities for continuing education related to natural hazards

mitigation?, and (5) How can funding be obtained for programs in

natural hazards mitigation?

Possible Solution: There appears to be little incentive to find answers to the above questions. To assist in solving this

problem, it is suggested that consideration be given to the

establishment of an institute in one of the Southeast Asian

countries where high seismicity is present (Manila has been suggested as a logical location). This institute could serve as a natural hazards data and research center for the Southeast Asia

region and as an educational center for engineers and scientists

working in this field. It could possibly publish a journal (say, the

Journal of Natural Hazards Protection in Southeast Asia ), sponsor

seminars, organize short courses, etc. Such an institute, if properly supported, might well provide some of the answers to the

questions raised above.

If favorable consideration is eventually given to the establishment of an institute of this type, every effort should be made to make the activities of the institute complementary to activities of

36-10
j other institutes, such as the Asian Institute of Technology (AIT) and

the International Institute of Seismology and Earthquake Engineering

(USEE).

Cooperative Efforts: Clearly, an institutional approach to research and information dissemination in Southeast Asia can be

successful only if all countries in the region cooperate fully,

assistance and cooperation is received from countries such as Japan and the United States, and adequate funding is made available.

Working Group Wl : Wind Response and Design

Chairmen: Dr. Richard D. Marshall

Prof. Angel A. Alejandrino

Rapporteur: Mr. Enrico G. Gregorio

Report of Group Wl

Working group Wl reviewed the general area of wind response and

design with the purpose of identifying those items or topics that are

in need of additional research and that might offer an opportunity

, for international collaboration. While the discussions covered many

topics, ranging from single-family dwellings to long-span flexible

bridges and nuclear power plants, three topics emerged as having a significant potential for protection against natural hazards. In addition, these three topics are believed to represent achievable

goals when the very real constraints of existing research capability

and available funds are applied. The topics are listed in order

of priority and are followed by comments describing the scope of the

proposed effort and means of implementation.

An International Wind Engineering Extension Service

The objective is to disseminate existing information of the

nature of wind forces, the design of buildings and other structures

to resist these forces, and recent developments in the formulation

of codes and standards. To make this dissemination effective and

to ensure its use by the engineering profession, it is proposed that

regional seminars or workshops be conducted from time to time in

appropriate countries of Southeast Asia, the courses being presented

by international experts and certain individuals from the engineering

profession in the host country. These seminars or workshops would

introduce the latest technology to the profession and would also

provide an opportunity to document current practice and experience

in the host country. Documents could be distributed and otherwise be

made available by appropriate local agencies such as the Technology

Resource Center (TRC), which has recently been established in the

Philippines

.

Post-Disaster Investigations

The objectives of this topic are: (1) to investigate and report

on disasters caused by extreme winds, (2) relate observed damage and

wind intensities to structural performance, and (3) to assess existing design criteria and provide feedback to the affected country

or countries. It is important to understand that this topic involves

far more than just the documentation of wind damage and that the

real value lies in the second and third items above. To be effective, contacts would have to be established before the fact with

participating countries so that investigative teams (to consist

of international experts and members of the engineering profession

in the affected country) could have timely and unlimited access to

the disaster area.

36-11
It is believed that the International Association for Wind

Engineering could provide the leadership and coordination for such an undertaking. Admittedly, this group is currently organized along

very informal lines without an operating budget and would require

financial assistance to carry out the objectives outlined above.

Establishment of a Boundary-Layer Wind Tunnel Facility

In addition to being an extremely useful research tool, a boundary-layer wind tunnel is an essential component in the design

of structures having unusual geometric features or dynamic characteristics which may make them sensitive to wind forces. The work

group believes that the establishment of such a facility, to be

supported by and made available to the countries of Southeast Asia,

is justified on the basis of current need by the engineering profession and on the basis of its potential for the reduction of

wind hazards. It is believed that the major cost of establishing

such a facility lies in the training of staff members who

would carry out commercial work in response to the needs of the

engineering community and who would also conduct research into the

mechanisms of wind loading. To be successful, a regional facility

such as this should be associated with an established research

center or institution.

Working Group W2 : Wind Hazards and Characteristics

Chairmen: Dr. R. Scanlan

Dir. Jesus F. Flores

Rapporteur: Prof. Evangel P. Quiwa

Report of Group W2

The work of the group was divided into four subsessions of

approximately 1-1/2 hours each, having the following centers of

focus

:

o Session 1: Objectives of the session and definition of technical problem areas

o Session 2: A discussion of the present state of the engineering art relative to wind and the natural

hazards caused by it in Southeast Asia.

o Session 3: A broad discussion of possible research and study

projects needed in the area.

o Session 4: A focussing discussion in which a limited number

of possible research and study projects needed

in the area were identified.

In session 1 the general objectives of the National Science

Foundation in this context were outlined as centering around the

international cooperation in the natural hazards area in Southeast

Asia. Particular willingness to supply travel funds to qualified

U.S. nationals acting as co-investigators in cooperative study

projects was indicated.

In the four working subsessions particular emphasis was given

to information and viewpoints supplied by representatives from

Australia, Hong Kong, Japan, New Zealand, Philippines, and the United

States of America. The absence of representatives from other Southeast Asian countries was noted and the general hope expressed that

these absences could be remedied at future meetings of similar type.

36-12
Principal Conclusions

A total of some 24 topics for research, and data-gathering were

broadly discussed. These finally were reduced to some 5 areas, of

which the following list is a highly condensed summary:

1. Needs for Continued Experimental Data-gathering relative to

o Large-scale cyclonic storms (typhoons, cyclones, hurricanes)

o tornadoes

o Boundary-layer properties of the wind at wind-chosen locations

(velocity profile, spectra and spatial correlations of turbulence, etc.)

o Personal discomfort in tall buildings and plazas

o Wind-tunnel modeling of specific problems

Needs for Storm-Disaster Management, Damage Assessment, and

Damage-Reduct ion

o Creation of a natural hazard data collection center and

clearing house for Southeast Asia (possibly based on PAGASA).

o Development of highly engineered approaches to disaster

warning and control.

o Establishment of qualified disaster evaluation teams to be

sent rapidly into areas of storm damage after an event.

o Launching of a campaign of public education emphasing

engineering measures that can improve the storm resistance

of non-engineered structures and reduce storm casualties

associated with them.

3. Needs for Wind-hazard Risk-assessment Studies Based on Stateof-the-Art Probabilistic and Analogous Approaches

o Assessment of tornado and typhoon risks to cities, agricultural resources, sensitive installations such as nuclear

plants, special laboratories, factories, etc.

o Needs to gather and compare wind codes from a wide range

of sources for structural engineering

o Studies assessing the effect of particular code applications

on representative designs.

+ . Assessment of the Full Implications and Efficiency of Typhoon

Modification and Reduction Efforts

A list of documents was submitted to Professor E. P. Quiwa,

rapporteur of panel W2 by interested parties. These documents

tiave been turned over to the documentation office of the Building

Research Center, University of the Philippines.

The Director of Typhoon Research at the Philippine Atmospheric,

peophysical, and Astronomical Services Administration offered con-

:luding remarks regarding his willingness to cooperate with structural engineers, in collecting and storing wind data.

Jorking Group H: Other Hazards

chairmen: Dr. Severino L. Koh

Dr. John Y. Lai

Rapporteur: Mr. Ronoldo I. Borja

36-13
Report of Group H

The group identified five problem areas of importance to the

Southeast Asia region in general and to the countries represented

in the group in particular. These problem areas were:

o Storm-generated Waves and Surges

o Landslides and Related Hazards

o Volcanic Hazards

o Inland Flooding

o Seismic-generated Waves (Tsunamis)

The group attempted a priority listing of the various research

topics based on three criteria: urgency, organizational support,

and availability of interested, qualified personnel to undertake

the specific work. The group subsequently determined that all

five problem areas were important to the region, and the participants at the Group discussions individually expressed interest in

certain specific topics.

The group noted that some collaborative efforts existed

between the Philippines and Japan on studies of inland flooding.

However, collaboration on the other areas had not been initiated.

Nevertheless, common interests were expressed by the participants,

and possible collaboration appeared to be most promising in the areas of storm-generated waves and surges, and in landslides and

related hazards. Possible collaboration among Hong Kong, the

Philippines, and the United States in the area of landslips of

filled slopes and mud avalanche was also expressed.

The group considered several questions relative to future

collaborative activities. Specifically, the group discussed the

need for continuing dialogues among the scientists and engineers

of the participating countries, the need to discuss research progress, to establish research institutes for concentrated work in

specified fields, and to create a data collection and information

center and repository of research findings. Based on these deliberations, the group formalized three specific recomendat ions

.

These are presented below.

Mathemetical Models for Predicting Storm Surges

A number of countries in southeast Asia are visited annually

by tropical cyclones which create substantial damage to property

and loss of lives. In terms of damage wrought, the storm-generated

waves and surges constitute the greatest hazard in the coastal areas around the Bay of Bengal, Hong Kong, Japan, and the Philippines.

Mitigation of the damage and loss of lives from these hazards could

be enhanced by the timely issuance of accurate warnings of their

impending occurrence. This could be done through the development

of suitable mathematical prediction models.

There are now some simplified models that seem to be adequate

for coastal areas which have relatively straight coastal lines and

well-behaved coastal bathymetry. At best, these models give only

rough and preliminary estimates of potential surges in most Southeast

Asian coastal areas due to the highly irregular coastal configurations and the steep bathymetry in many areas. Furthermore, unusually

large surges occur in closed or semi-closed basins or bays.

To predict storm surges in the region more accurately, there

is an urgent need for the development and formulation of more realistic mathematical models. Such models should predict flooding

36-14
azards resulting from the generation and the propagation of storm

urges and waves in the Southeast Asian coastal complexes including

pen coasts, bays, and rivers. They should: (a) include the

ffects of irregularities in the coastal boundaries and the bathymtry of the continental shelves; (b) reflect time-dependent inundaion boundary conditions along the low-lying terrains; (c) provide

n acceptable degree of numerical resolution of the non-linear

dvection terms to improve the stability criteria as well as the

onvergence capability of the model; (d) be applicable for storms

ravelling parallel as well as perpendicular to the shoreline; and

e) be reasonable in computational time.

Advances in mathematical modeling achieved by scientists from

he United States and Japan would be helpful to the Southeast Asia

egion. Specifically, the ongoing research efforts on storm surges

n the Philippines supported by the NSDB* with primary activity in

he PAGASA** could benefit from collaboration with workers from

he United States and Japan.

Development of Risk Methodology for Storm Surges

Supplementing the work described above, a quantitative risk

ethodology for predicting the hazards from storm surges could be

eveloped. This effort would involve the development of a deducive probabilistic model for evaluating the annual probabilities (or

eturn-periods ) associated with specified surge levels at a given

ite. For effective implementation, the model must necessarily

epend on and make use of available data on the major factors

hat influence storm surges as well as on the mechanics of surge

evelopment

.

The successful development of the model should serve to improve

he risk mapping of specific regions of a country. Such a model

ould also serve to identify the data required for improving risk

ssessment s

.

Instrumentation and Remote Sensing for Monitoring Storm Surges

The advancement of the present state of knowledge on prediction

f storm surges, waves, and currents suffers from a deficiency in

dequate measurement of parameters under storm conditions. Wave

nd storm surge research efforts in Southeast Asia can be boosted

ignif icantly with collaborative help from American scientists who

ave been successful in developing suitable instrumentation and emote sensing techniques for storm conditions.

A special instrumentation system for obtaining surge history

nd waves under hurricane conditions has been developed in Florida,

luch a system can be deployed in the Philippines to supplement and

upport their studies on storm surges. A possible area of collabration could be the installation and maintenance of the instrumentation system by Filipino participants at sites of their choice,

he American participants can build and furnish the finished instrumentation system, and can provide instructions and training for its

jjse, while the Filipino participants could assist in the installation

?nd maintenance of the system at preselected sites. The analysis

If data would then be conducted jointly in relation to storm surge

iodels available in the United States and in the Philippines.

NSDB - National Science Development Board of the Philippines.

* PAGASA - Philippine Atmospheric, Geophysical, and Astronomical

Services Administration.

36-15
The United States is also launching the first oceanographic

satellite in May 1978. The capabilities include the measurement

of wind speed and direction, wave height and direction, ocean surface elevation, and ocean surface temperature. Information obtained

from SEASAT-A in Southeast Asia can be provided to those countries

that have an interest in receiving the data and are willing and

able to furnish surface data measurements to calibrate the satellite data. The Philippine agnecy. PAGASA, has indicated a strong

interest in such a collaborative undertaking.

Rain-induced Landslides and Mud Avalanches

While extensive studies have been done to understand the

mechanics of earthquake-induced landslides, comparatively little

effort has been spent on landslides cause by means other than

seismic disturbances. In the Philippines, in Hong Kong, and other

parts of the Southeast Asia region, landslides triggered by heavy

rains cuased considerable damage and loss of life. The landslips

of 1972 and 1976 in Hong Kong were particularly disastrous. In the

Philippines, rockslides and landslides are common occurrences during

typhoon passages as well as during the monsoon season. Yet little

is known regarding the fundamental tie-up between the varying properties of soil and soil instability. No suitable prediction methods

for the occurrence of these hazards are available.

Collaboration between American scientists and those from the

Philippines, Hong Kong and other parts of Southeast Asia can prove

useful in the development of new knowledge on the mechanics raininduced landslides, where theoretical models can be tested in

actual situations in the Southeast Asia region. Ultimately, the

goal of such a collaborative project would be to develop methods

for analyzing slope stability and design criteria for slope structures. Manuals on slope such as that being developed for Hong Kong

would be useful to other parts of Southeast Asia.

Damage Evaluation Criteria for Soil Structures

Natural slopes and man-made cuts are subjected to such

influences as earthquakes and heavy rains to the extent that the

integrity of the soil structure is threatened, instability is induced, and landslides are caused. In some cases, only minor

relative earth motions are experienced and the integrity of the

soil structure remains intact. Nevertheless, it is necessary that

a damage evaluation system be developed for soil structures, so that

the stability of the slope can be assessed and the threat of landslides be monitored and their occurrences predicted.

Collaborative studies can be undertaken with the specific aim

of developing methods for assessing cumulative damage resulting from

successive seismic loads or instances of heavy rainfall of different

intensities and duration. Prediction models may be tested against

properly instrumented experimental models.

Insitu Testing of Soils to Determine Liquefaction Potentials

The possibilities of submerged deposits of loose cohesionless

soil liquefying during seismic loadings are always of concern to a designer in a country like the Philippines due to the frequency of

seismic activities and the rather extensive damage that soil liquefaction causes. Because soils of this type are hard to sample for

laboratory testing and because of the resulting poor correlation

between laboratory test data and field data, it is desirable to extend new knowledge on field testing procedures developed in the

36-16
jUnited States and other countries to the developing countries in

(Southeast Asia.

A Risk Methodology for the Prediction of Volcanic Eruption

A risk evaluation method should be developed for assessing

|and predicting the hazards from volcanic eruptions of an individual

volcano and/or the hazards to a given site from surrounding

volcanoes. The work should involve the development of suitable

jprobabilistic models for calculating the probabilities of volcanic

[eruptions, depth of ash falls, lava flows, mud flows, and the areas

'affected.

The purpose of the model will include the determination of the

likelihood of levels of volcanic hazards in specific regions of a

| [country. The form of the required model must necessarily depend on lavailable data of previous eruptions as well as the science and

land mechanics of volcanology.

The improve and support the information base needed in the

jjrisk methodology, additional data on the probable type of volcanic

[activity have to be obtained in terms of seismic and geothermic and

behavior and petrogenesis. For this purpose, further instrumentation and monitoring of specific volcanoes will be vital.

Inland Flooding

Inland flooding is a major problem in Southeast Asia. Heavy

hmonsoon rains and typhoons annually tax the inadequate flood control

[land drainage system in the countries resulting in considerable damage

Jto properties and loss of life. While some success has been experienced in the operation of a flood forecasting network for the

JiPampanga River Basin, existing models like the tank model give only

jjgross information on the expected behavior of the natural drainage

system in response to hydro-meteorological events. These models

prove inadequate in that they fail to take into proper account the

'physical characteristics of the specific basin. Therefore, there

j is an urgent need for the development of realistic models as indicated below.

Development of Conceptual Models for River Basins

Fundamental studies on flows and river stages which take into

consideration hydrologic models are needed to improve techniques for

flood forecasting. Conceptual models may then be developed for

specific river basins. The results of these efforts are urgently

needed to provide inputs for the development of flood control

systems and drainage designs for frequently flooded basins in the

country. The urgency of the problem is understood by the fact that

very vitual economic and developmental activities are either in

jprogress or planned for major river basins and deltas in the

Philippines as well as in other parts of Southeast Asia.

The PAGASA and the Bureau of Public works of the Philippines

have started working on flood forecasting and control for the

major river basins in the country. Very recently, the National

Flood Forecasting Office was created under PAGASA to oversee this

particular activity. Basic studies are here proposed together with

specific applications to particular regions would prove useful in

improving the flood warning capabilities of these agencies.

36-17
Seismic Generated Waves (Tsunamis)

The Intergovernmental Oceanogr aphic Commission of the UNESCO

through the Tsunami Warning System Coordinating Group is the coor- dinating body for the study of Tsunamis. The Philippines participates in this work, and the Internatinal Tsunami Information Center

in Honolulu provides some assistance in training. Locally, several

academic institutions and government agencies are involved in the

different aspects of this natural hazard ranging from basic studies

to the diseemination of information to the provision of assistance

to damage areas. Three specific needs are indicated below.

Development of Risk Methodology

Research on the risk of Tsunamic occurrences should be undertaken for the various areas of the Philippines. A form of probability map could be developed to formulate some sort of measure

for land-use planning and infrastructure construction in developing areas, particularly along the eastern coast of the Philippines.

Instrumentation and Remote Sensing for Monitoring Tsunamis

Instrumentation and remote sensing projects such as that discussed earlier could also be useful for improving the Tsunami

warning system in the Pacific. Better warning systems should be

developed to the extent tht the lead time is lengthened. An effort

should also be made to educate the public about Tsunamis, to be

able to react to such disaster, as well as for site planning and reset t lement

.

Development of Damage Evaluation System

A system for evaluating damage to both terrestrial and aquatic

life and property should be the subject of investigation. This

should consider as well the impact caused by Tsunamis on the

envi ronment

.

Additional Recommendations

The group supports unanimously and recommends strongly the

organization of future symposia similar to the Manila Symposium

every two years at different locations in Southeast Asia. The

possibility of having the next one in 1979 either in Thailand or in

Singapore was accepted by the group.

The group is of the consensus that research institutes could

be organized with specific charges in specified areas. It takes

note of the fact that a Center for Volcanology Research is now

under the consideration of UNESCO. However, the group feels that

at this point in the creation of a Southeast Asia Committee to oversee this as well as to coordinate future symposia would be more easily achieved than to establish institutes.

Participants at the meetings of the group voiced a need for more technical and scientific information, journals, books, and

other publications in the developing countries. The group feels

that it would be useful to establish an efficient mechanism for

quick interchange of information. A data information center could

be organized; however, the group feels that such an endeavor should

be approached with caution in that the organization of yet another

center might duplicate needlessly the serivces available at other

centers, e.g., the Asian Information Center for Geotechnical

Engineering

.

36-18
Summary and Conclusions

The Manila Symposium is the first of its kind in Southeast

Asia on the subject of Engineering for Natural Hazards Protection.

The theme of the symposium is clearly of great importance to the

countries in this region; therefore, the convening of the symposium

in Manila was timely and indeed was long overdue.

The conference part of the Symposium afforded an opportunity

for local participants to get an overview of the current state-ofthe-art of engineering research and developments related to engineering for natural hazards protection, with emphasis on earthquake

and wind. It served also to provide a forum for the researchers

from the different countries to review their recent research results

and approaches.

During the research workshop phase of the Symposiu, a number of

worhtwhile research topics were identified and defined as suitable

for joint collaborative support between or among the countries

represented at the Symposium. The specific topics have been well

summarized in the report of the various working groups, as presented

in Chapter 3. Besides these specific research topics, several

(recurring problems that are unique to the Southeast Asia region were emphasized by the different working groups; these are worthy of

repeating, and may be summarized as follows:

(1) There are local and special requirements that are unique

[to Southeast Asia and thus specialized research studies are needed;

however, one of the most urgent needs in the region is for the

proper dissemination and adaptation of available information and

research results already developed by other countires, especially in

the U.S. and Japan. In particular, there is currently a general

lack of comprehensive and up-to-date references on technology

related to natural hazards protection in the region. Consequently,

ithere is a critical need for an information center and repository

of research findings for this region; the establishment of an information center for the entire Southeast Asia region was suggested

.

(2) Concern was expressed also for an effective collection of

data related to natural hazards, and for the systematic analysis

and dissemination of this information aimed at stimulating the

improvement of engineering practice. The idea of establishing an institute in one of the Southeast Asian countries for these purposes was suggested. Such an institute could serve as a natural hazards

[data and research center, and as an educational center to sponsor

1'seminars, organize short courses, etc.

(3) The main problem in implementing the suggestions developed

at this symposium would be the commitment of sufficient funds for

qualified personnel and equipment necessary to achieve the goals of

liany joint program. It is the hope that symposia of this type will

ihelp to enhance the recognition of such needs by appropriate

funding agencies, arid expedite or encourage the future commitment

jof needed funds.

(4) The holding of additional future symposia on the subject

of engineering for natural hazards in Southeast Asia was strongly

endorsed by the various working groups. In this regard, there is an offer from the delegation of Thailand to organize and co-sponsor

ithe second symposium on the same subject in Bangkok, Thailand, with

a suggested (tentative) data of January 1980. In any future

symposia on this subject, wider representation and participation

36

36-19
by the countries should be encouraged. Planning of future symposia

may include a pre-conference meeting of the organizers, for the purpose of selecting conference topics and preparaton of avaialble

research material.

36-20
37

STORM SURGE

Celso S. Barrientos

National Weather Service

National Oceanic and Atmospheric Administration

Department of Commerce

Storm Surge

fhen a tropical or intense extratropical cyclone approaches or

:rosses a coastline, sea level rises anywhere fron a fraction of

meter to as much as 9 m. above the normal tide. This rise of

ea level due to the storm is called storm surge. It is defined

s the difference between the observed actual sea level and the

ea level that would have occurred without the storm; it is that

ortion of the rise of sea level due to the storm alone. Storm

urge is usually estimated from tide record by subtracting the

ormal or astronomic tide from the observed tide. This is illusrated in Figure 1, a two-day record for Atlantic City.

Storm surge occurs worldwide. In the Western Pacific it is ssociated with typhoons. In the Indian Ocean and Bay of Bengal,

torm surges are due to cyclones. Australia has been struck

ecently by two intense cyclones. The first, "Althea," struck

ihe coast of Queensland on the morning of 24 December 1971 and

ave a surge of 2.7 m. The second cyclone, "Tracy," hit Darwin m the early morning of Christmas Day 1974. Tracy had a maximum

'ind of 175 km ^ with gust of 200 km hr

-

^ , but the surge was only

.6 m.

In Japan, many typhoons bring surges over 2.0 m. along the

pastal areas. A typhoon that moved over Ise Bay on 26 September

)59 produced a peak surge of 3.6 m. near Nagoya. On August 21,

)70, a typhoon moved over Tosa Bay that gave a peak surge of 2.4

the highest water level, 3.4 m., was recorded in Ashisuri.

lable 1 shows some major surges in Japan.

The northern portions of the Bay of Bengal are susceptible

bd vulnerable to storm surges. A very intense cyclone passed over

Sangladesh on 12-13 November 1970. It generated an extremely high

arge, which was estimated to be more than 6 m. This cyclone

jaused much damaged and killed several hundred thousand inhabitants

n the off-shore island and coastal districts of Bangladesh. Storm

,-irges reaching up to 9 m. have been estimated in the Visayan

sland of the Philippines.

Storm surges are generated by wind and atmospheric pressure

adients associated with the passage of storms. However, passage

storms alone does not mean they will be accompanied by surge,

jiere are other significant factors in surge generation: path of

:orm, water transport by waves, the earth's rotation, coastal conguration, and off-shore depth field (bathymetry).

Figure 2 illustrates the direct wind effect in raising the

iter level. If the wind is blowing at an oblique angle to the

iastline, parallel current will be generated, and because of the

rth's rotation (Coriolis effect) this current will raise the

irge. This is shown in Figure 3. Wave breaking also causes the

ter to pile up on the coastline as shown in Figure 4. This paper

11 discuss how the coastal configuration and bathymetry are portant in surge generation later.

The remaining part of this paper will focus on the United

'ates. I will review the forecast methods for storm surges used

the National Weather Service (NWS).

Climatology of storm surge is important in determining future

obability of surges and in computing tidal frequency. We need

know the probability of surges for designing and constructing

37-1
coastal structures. Surge frequency is also necessary to estimate

flood insurance rates. Further, accurate and timely forecasting

of surges can save lives and prevent property damage.

On the Gulf and East coasts of the United States, there are

two types of meteorological systems that can cause storm surge.

These are the extratropical storms that occur on the east coast of

the U.S. during the winter and the tropical storms (or hurricanes)

that strike both the Gulf and East coasts from May to December.

I will discuss here these two phenomena and the forecast techniques

we use to predict surges connected with them.

Low water or negative storm surge (sea level below normal)

can also cause problems. In the western end of Lake Erie and in

the north end of Chesapeake Bay near Baltimore, negative surge can threaten navigation.

Extratropical Storm Surges

The damage caused by storm surges is most severe in the

northeast and mid-Atlantic states. These storms generally move

along the coast and may cause heavy snowfall in the eastern

seaboard. The storm is generally referred to as a "northeaster."

Although extratropical storm surges are not as high as the

hurricane surges, they linger for a considerable length of time

( 2 to 5 days), and can be superimposed on tidal maxima. The long

duration of the surges also cause severe beach erosion, thus

destroying the foundation of structures.

A typical sequence of weather maps for extratropical storms

during 18-20 February 1972 is shown in Figure 5. This storm

caused extremely high tides, plus extensive damage and beach

erosion in the northern U.S. Atlantic coasts (Figures 6 and 7).

Figure 8 is the storm track, observed tide curves, and storm

surge curves for 17-21 February 1972 for stations on the east coast. Most of the extratropical storm surge data here are from

a study by Pore, Richardson, and Perrotti [1974] on forecasting

extratropical storm surges.

The series of slides I will be showing you will further

illustrate the damage caused by extratropical storm surges.

These slides are not in my paper because they will be quite

expensive to reproduce. The slides are from the collection of

Mr. Art Pore, who is the chief of our marine techniques branch.

Probably, the most devasting east coast storm in recent times

is the one that occurred on 5-8 March 1962. This storm is commonly

known in meteorological circles as the "Ash Wednesday" storm. Extensive damage and severe beach erosion were left behind by this

"northeaster." This storm is a typical case similar to the 1972 storm. However, the difference was the intense pressure gradient

causing a strong northeast wind blowing along an unusually long

fetch. Figure 9 is the sequence map for this storm. Figure 10

is the storm surge during this storm, for some stations on the

east coast.

Tropical Storm Surges

Tropical storms form in the warm tropical ocean and move generally toward the west. The storm is a warm core system with

generally a pronounced "eye." It is most frequent during late summer, when the heating of the tropical o.ceans has peaked. When

37-2
the storms approach the land mass, they generally recurve northward,

n the northern hemisphere, tropical storms are most frequent in the

estern Pacific, followed in frequency in the western Atlantic. The

torm is known as a typhoon in the Pacific and a hurricane in the

tlantic. However, strictly speaking, typhoons or hurricanes are both

efined as storms with a maximum wind speed of 120 km hr

-

^" or more.

Since tropical storms are low-pressure systems, water bulges up

nder the storm. This is referred to as the inverted barometer

ffect. With the strong circulating wind acting on the water surface,

complicated rotating mound of water forms under the storm. In the

pen ocean, this rotating mound of water is hardly noticeable under

chaotic sea. When the storm reaches the continental shelf, the

otating mound of water will start to "feel" the bottom and coastal

oundary. The constrained water results in storm surges. That

s to say, a small amount of vorticity is converted to divergence and

lanifested as a surge.

The Gulf of Mexico and the east coasts of the United States

ave relatively wide and shallow continental shelves; additionally,

he coasts are concave in shape. These coastal characteristics

re favorable factors to high surges.

One of the most devastating hurricanes was "Camille" in the Gulf

f Mexico in 1969. Camille generated a surge of 7.4 m. near Pass

Ihristian, Miss. It is one of the highest surges recorded in the

'.S. More than 250 people lost their lives and several hundred

lillion dollars in damage was left in the wake of Camille.

igure 11 is an illustration of the damage of Camille. The bottom

licture shows what is left of some apartments after the passage of

Camille. In this apartment building alone, over 30 people died

uring a "hurricane party".

i orecasting Techniques for Storm Surges

The empirical method was the earliest technique used for preicting storm surges. This is based on correlating variations in

ea-level (surge) with the meteorological factors in surge generaion. Those meteorological factors are sea-level pressure (or

ntensity of storm) and the strength and direction of the prevailing

ind (wind forcing). The only problem with this kind of statistical

tudy, is that a large number of observations is needed to attain

meaningful level of significance for prediction. Since surge

ipisodes are caparatively rare, large numbers of observations to

Ierive empirical equations are generally not available.

On the east coast of the United States, it is fortunate that

e have enough extratropical strom surge cases to derive statistical

orecast techniques. We used the surge data and the sea-level

ressure at various points to compute statistical forecast equations,

figure 12 shows the station on the east coast where forecast equations were derived. Figure 13 shows the grid points where the sea-

?vel pressure data were used in the derivation of the forecast

quations. These forecast techniques are used routinely in the

ational Weather Service. Figure 14 is an example of a forecast

eletypewr iter message by this technique. Extratropical storm

arge forecasts are issued by the NWS Forecast Offices.

The NWS National Hurricane Center is responsible for predicting storm surges associated with tropical storms and hurricanes,

'fie forecast is generated by a numerical-dynamical model developed

/ Jelesnianski . The computer forecast model is called SPLASH

(Bpecial Program to List Amplitudes of Surges from Hurricanes).

37-3
SPLASH consists of a storm traveling across a rectangular coastal

basin with variable depth. The storm surge driving forces— wind

and pressure gradient —are derived from the storm model. Figure 15 is an example of the output of SPLASH (for hurricane Camille).

Recently, the World Meteorological Organization (WMO) , one of

the specialized agencies of the United Nations, has published

"Present Techniques of Tropical Storm Surge Predictions." The publ

cation was authored by Dr. Miyazaki from Japan, Dr. Jelesnianski

from the United States^ and Dr. Das from India.

37-4
TABLE 1. MAJOR SURGES IN JAPAN (MIYAZAKI, 1957)

Dote At tected Peak Highest Extreme Volues

area surge water Pressure Wind Location

leve 1

f e-lA Cm s ) r_\ (m) (m;

i in 1Q17 i. iu. lyi/ Tokyo Boy yDU.4 err j r\ r\ oot 4U.U T 1 1 okyo

lo./.lyoU Ariake Sea /L m O yD4. o CMC Trt A. tlNt OU.O Tomie

01 O IOTA Osaka Bay 1O 1 • 1 0<^A T <; AR A usa ka

i. . 7 • 1700 2 2 978.6 S 31.0 To kyo

o,7t lyju Re nUn nn\t NE 33.4 r\o uc 17.8.1956 Ariake Bay 2.4 4.2 968.4 SE 27.0 Soqa

26.9.1959 Ise Bay 3.6 3.9 958.5 SSE 37.0 Nagoya

16.9.1961 Osaka Bay 2.5 2.9 937.3 SSE 33.3 Osaka

25.9.1964 Osaka Bay 2.1 2.6 983.5 S 27.1 Sumoto

10.9.1965 Osaka Bay 2.2 966.0 SSE 38.8 Sumoto

21.8.1970 Tosa Bay 2.4 3.4 962.3 SW 35.8 Ashi zuri

1

1

f

37-5
Atlantic City

Source: Pore 1974

FIGURE 1. TIDE DATA

^h. cc JL T = wind stress

3 * D D = depth

h = storm surge

T

" >~

^^^^

disturbed water level due to t

h

still water level

Source: Harris 1963a FIGURE 2. EFFECT OF ONSHORE WIND ON WATER LEVEL

37-6
f

= Coriolis parameter

3l» T oh r T = wind stress

<r —• a. tV , dt D OX storm surge D = depth

Source: Harris 1963a

FIGURE 3. EFFECT OF OBLIQUE WIND ON WATER LEVEL

wave

''7?^. 1 setup breaker

2or>e

Water level including wave effects

Source: Harris 1963a

FIGURE 4. WAVE SETUP IN VERTICAL PLANE

37-7
37-8
m•

FIGURE 6. DAMAGE TO HOUSE AT WESTHAMPTON BEACH, NY, BY 18-20

FEBRUARY 1972 STORM. (COURTESY OF US ARMY CROPS OF

ENGINEERS COASTAL ENGINEERING RESEARCH CENTER)

FIGURE 7. UNDERMINED CONCRETE SLAB PARKING LOT AT JONES BEACH

NY, FROM 18-20 FEBRUARY 1972 STORM. (COURTESY OF

US ARMY CROPS OF ENGINEERS COASTAL ENGINEERING

RESEARCH CENTER)

37

37-9
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37-10
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37-14
FZUS3 KWBC 021200

EAST COAST STORM SURGE FORECAST IN FEET

12Z 18Z OOZ 06Z 12Z 18Z OOZ 06Z 12Z

PWM 0.1 O.U 0.5 0.7 0.7 0.9 0.8 0.9 0.7

BOS 0.1 0.2 0.3 O.U O.U 0.5 O.U 0.6 0.3

NWP 0.8 1.0 1.1 1.2 1.2 1.3 1.1 1.1 0.9

SFD 1.0 1.1 1.2 1.2 0.9 0.7 0.6 0.1 0.0

LGA 1.0 1.0 1.0 1.3 1.2 1.0 1.0 0.8 0.7

NYC 1.2 1.3 l.U l.U 1-3 1-2 1.0 0.9 0.7

ACY 0.8 0.9 1.0 0.9 0.9 0.9 0.8 0.8 0.7

BWH 1.1 1.1 1.2 1.0 0.9 0.9 0.7 0.7 0.6

BAL 2.U 2.5 2.6 2.8 2.8 2.6 2.U 2.0 1.9

ORF 0.9 0.8 0.9 0.8 0.8 0.8 0.7 0.6 0.9

PWM Portland, Maine

BOS Boston, Massachusetts

NWP Newport, Rhode Island

SFD Stamford, Connecticut

LGA Willets Point, New York

NYC Battery, New York

ACY Atlantic City, New Jersey

BWH Breakwater Harbor, Delaware

BAL Baltimore, Maryland

ORF Hampton Roads, Virginia

Source: Pore, Richardson, and Perrotti 1974

FIGURE 14. STORM SURGE FORECAST TELETYPE MESSAGE. VALID TIMES

INDICATED ABOVE EACH COLUMN; STATION CALL SIGNS IDENTIFIED BELOW MESSAGE

37-15
37-16
38

OPERATION OF TSUNAMI WARNING SYSTEM IN THE PACIFIC

Mark G. Spaeth

National Weather Service

National Oceanic and Atmospheric Administration

Operation of the Tsunami Warning System in the

Paci f ic

After the devastating Aleutian tsunami of April 1, 1946, military

and civilian sources criticized the U.S. Coast and Geodetic Survey

for the lack of warning in the Hawiian Islands. The critics correctly pointed out that seismic waves from this earthquake were recorded at Honolulu and other observatories within minutes after

the occurrence of the earthquake; and consequently, the tsunami

could have been predicted. The error in this criticism lay in the

[fact that seimograph records were changed only once each day, and

iuntil the film records were developed, no knowledge of an earthquake occurrence was normally available. In addition, the great majority

pf underwater earthquakes do not cause tsunamis, and no arrangements

were then in effect to verify the existence of the tsunami through

actual observation.

Another necessity for efficient operation of a warning system

as the development of a method to determine quickly and accurately

he amount of time between the occurrence of a tsunami-producing

arthquake and the arrival of the waves in the Hawaiian Islands,

hs problem was solved by the preparation of seismic sea-wave travel

ime chart for Honolulu in early 1947. This chart consists of a eries of more or less concentric lines overprinted on a chart of

he Pacific Ocean. The lines represent distances from Honolulu for

ach half hour and hour of the wave's travel time. Travel time to onolulu is obtained by plotting the epicenter of an earthquake on

(the chart and noting its position with respect to the time lines,

iven the time of the disturbance, the arrival time of the first sea

ave at the Honolulu tide station becomes immediately available. The

eed for traveltime charts based on the other tide stations became

evident early in the operation of the Tsunami Warning System (TWS).

this need was based on two factors: (1) The necessity of estimating

tfhen the existence or absence of a wave could be observed at a particular tide station, and (2) the necessity of providing accurate

estimates of tsunami arrival times to additional countries and areas.

Charts for all tide stations in the TWS were completed by 1950.

traveltime charts are prepared on a continuing basis for new tide

stations that join the Warning System from time to time.

iecause the manual preparation of traveltime charts is a tedious,

;ime-consuming task, a computer program was developed to compute

.he charts. Estimated times of arrival are now determined by table

.ook-up on a computer.

Personnel in the U.S. Coast and Geodetic Survey believed that

he technical problems preventing the establishment of a workable

eismic sea wave warning system could be overcome. Although the

loast Survey had no obligation under law to provide tsunami warning,

t saw the need for such warnings and had the scientific know-how

ind part of the organization necessary for such an undertaking,

'urthermore, responsible officials in the Coast Survey considered

hemselves morally obligated to create and organize such a system.

Inder the direction of Cdr. 'E. B. Roberts, Cdr. C. K. Green, and

Jr. W. B. Zerbe, work began.

The acquisition of suitable seismographs and visible recording

quipment was one problem that had to be solved. Photographic techiques were generally used by seismologists for recording earthquakes

ecause they were simple, practical, and precise. Visible recording

pparatus in existence in 1946 was generally unsatisfactory and of

38-1
poor accuracy due to electronic problems. Various instrumental syssystems were tried, the most promising being that designed by Fred

Keller, a New Kensington, Pa., scientist. In 1947 and 1948,

equipment following his design was built and' installed at Tucson,

Ariz.; College, Alaska; and Honolulu, Hawaii. These installations

were modified during the summer of 1950 by the addition of a highly

stable, cathode-coupled, split-beam amplifier developed by R. M. Wilson and L. R. Burgess of the U.S. Coast and Geodetic Survey.

Today j the TWS generally relies on helicorders driven by solid-state

amplifiers. When a strong earthquake is recorded by these instruments, an alarm (audible, visible, or both), located where it is

always noticeable, is tripped, thus insuring the prompt observation

of all major earthquakes.

The collaboration of the Armed Forces and the Civil Aeronautics

Administration (now the Federal Aviation Agency) was sought to establish a rapid, high-priority communications system. A meeting was held on August 12, 1948, to discuss a proposed communication plan

for the warning system. Attending this meeting were representatives

of the U.S. Army,Air Force, Navy, Civil Aeronautics Administration,

and Coast and Geodetic Survey. As the various armed services and

the CAA were well aware of the need for a tsunami warning system

in the Pacific, the tentative plan was approved. Thus, the operation

of the U.S. Coast and Geodetic Survey's Seismic Sea Wave Warning

System in the Pacific began [Tsunami Warning System since March

15, 1967].

Initially, the Warning System consisted of the U.S. Coast and

Geodetic Survey seismological observatories at College and Sitka,

Alaska; Tucson, Ariz.; and Honolulu, Hawaii; and the following tide

stations — Attu, Adak, Dutch Harbor, and Sitka, Alaska; Palmyra

Island; Midway Island; Johnston Atoll; and Hilo and Honolulu,

Hawaii. However, the Warning System began to expand almost immediately. By November 15, 1949, when the first edition of the Communcation Plan for the Seismic Sea Wave Warning System was issued by

the U.S. Coast and Geodetic Survey, seismological observatories at Berkeley and Pasadena, Calif., and Tokyo, Japan, and tide stations

at Kodiak, Alaska; San Pedro and La Jolla, Calif.; Balboa, C.Z.;

Canton Island; Apra Harbor, Guam; Koror Island; Kwajalein Atoll;

Wake Island; and Pago Pago Harbor, American Samoa, had been added.

Some seismological and tide stations have join ed the TWS and others

have left it over the years. At present, 22 seismological observatories and 53 tide stations are participating. Some of the observatories report telemetered data from extensive seismic nets. Figure 1 shows the participating stations.

Initially, the TWS was to supply tsunami warning information

to the civil authorities of the Hawaiian Islands and to the various

military headquarters in the Hawaiian Islands for dissemination to military bases throughout the Pacific and to the U.S. Trust Territory of the Pacific Islands. Beginning on October 14, 1953, the

warning information furnished to the civilian authorities of the

Hawaiian Islands also was given to the Civil Defense Agencies of

California, Oregon, and Washington. In June 1957, the Pacific

Tsunami Warning Center (PTWC) began supplying estimated arrival

times for seismic sea waves on the coasts of these three states.

The great destruction caused by the May 1960 Chilean tsunami caused

a large number of countries and territories to join the Warning

System in order to be protected from future tsunamis. Beginning in

November 1960, warnings were supplied to Canada, Alaska, and Tahiti.

Additional warning information was supplied to Japan, beginning

in February and March 1961. Taiwan started receiving warnings in

November 1961, and the Republic of the Philippines and the Fiji

38-2
slands in December 1961. During 1962, the PTWC began supplying

arnings to Chile (March), Hong Kong (June), and New Zealand (July),

estern Samoa began receiving warnings in January 1963, and American

amoa in November 1963.

The first test of the TWS came in connection with the Tonga

slands earthquake of September 8, 1948. Seismograph reports were

btained, and the epicenter of the earthquake was located. A ravel time of 6 hours and 35 minutes for the wave from the epienter to Honolulu was predicted from the traveltime chart, and

ilitary and civilian agencies were alerted to stand by for a ossible tsunami warning. As wave reports came in, it became evient that the wave would be small, and the alert was cancelled,

ater, as predicted, the Honolulu tide-gage record showed that

6-inch seismic sea wave had arrived.

Several small tsunamis were registered in various parts of

he Pacific during the years 1949, 1950, and 1951, but no warnings

ere issued to the public as a result of the action by the Tsunami

arning System. The extreme iregularity of earthquake occurrences

as demonstrated in the years 1950 and 1951. In 1950 the Pacific

'sunami Warning Center requested tide reports in connection with

6 earthquakes, but in 1951, there were only 3 such occasions.

Regular monthly tests to keep operating personnel familiar with

.he communication methods and requirements were begun early in

.949. These communication tests have been instrumental (1) in

liminating flaws that showed up in the TWS from time to time, and

2) in determining the most rapid and reliable communication routes.

The first major tsunami in the Pacific after the formation of

arning System occurred in connection with November 4, 1952, Kamhatka earthquake. Advance warnings provided to communities in the

ath of this tsunami resulted in a reduction of damage and no asualties. The March 9, 1957 Aleutian Tsunami, the second major

acific tsunami following the formation of the SSWWS, caused $3

illion damage in the Hawaiian Islands, but once again, due to imely warning, there was no loss of life. The May 1960 Chilean

usnami pointed out a weakness in the Warning System. Although

sunami warnings were broadcast to the residents of the Hawaiian

slands well in advance of the arrival of the wave, 61 people were

illed in Hilo, Hawaii, because they failed to heed the warnings.

The 1960 Chilean tsunami and the 1964 Alaskan tsunami stimuated a sizeable increase in the basic research on tsunamis, and

lso caused a number of countries to request tsunami warnings from

he PTWC. These requests necessitated the addition of tide and

eismic stations to secure the coverage necessary to provide

imely warnings.

A reorganization in the U.S. Government on September 16, 1973,

ssigned operation of the TWS to the National Weather Service.

perat ions

The operation of the Tsunami Warning begins with the detection

'f a major earthquake by one of the observatories in the System,

larms are set so that any earthquake of magnitude 6.0 anywhere in

ne Pacific will trigger at least one. The Pacific Tsunami Warning

enter (PTWC) collects and interprets the seismic data, locating

ne epicenter and determining the magnitude of the earthquake. If

38-3
the earthquake is of insufficient magnitude or is located in an area where a tsunami will not be generated, an information bulletin

is issued and the investigation is closed.

When an earthquake of sufficient magnitude occurs in an area where the generation of a tusnami is possible, the PTWC issues a watch bulletin immediately to the dissemination agencies. This

watch bulletin includes the location and magnitude of the earthquake

and the estimated arrival time of the tsunami if one has been genera

ted.

When the PTWC determines that the possibility of tsunami

generation exists, they immediately begin requesting data from the

tide stations nearest the epicenter. When confirmation of the existence of a tsunami is obtained, a warning is issued to the dissemination agencies. This warning contains the location of magnitude

of the earthquake generating the tsunami and the estimated time

of arrival of the disturbance at the locations for which the TWS

provides warnings. The dissemination agencies are then responsible

for warning the population in the areas likely to be impacted by

the tsunami and telling them what to do to avoid danger. The dissem

ination agencies also have the responsibility for educating the

public about tsunami dangers and preparing disaster plans.

Tsunami Warning System criteria call for the issuance of watch

messages within one hour of the occurrence of an earthquake meeting

magnitude and location criteria. The threshold calling for the

issuance of a tsunami watch is the occurrence of an earthquake

with a magnitude 7.0 in the Pacific. Generally, the TWS meets this

objective although epicentral locations are occasionally in error by 2 or 3 degrees due to lack of proper distribution

of seismic data.

The TWS attempts to issue necessary tsunami warnings within 2 hours of the earthquake occurrence. This criteria often cannot be

met because of poor communication links to many of the tide stations

and also because there may be no participating tide station within

2 hours travel time of the tsunami generating area.

Once a watch or warning has been issued by the TWS, supplementary bulletins are issued at intervals of one hour until the

investigation is closed. The automation of message preparation

and handling has greatly helped meeting this requirement.

Regional Tsunami Warning Systems

The National Weather Service operates regional tsunami warning

systems in Alsaka and Hawaii. Because tsunamis generated in these

regions can strike the shore almost immediately after the earthquake

occurs, the systems were set up to provide immediate response on the

basis of seismic data telemetered to the warning centers. While

data from tide stations are telemetered to the centers, these data

are used only for confirmation or cancellation of the warnings

issued as a result of the interpretation of seismic data.

Procedures adopted at the warning centers have been selected so that tsunami warnings can be issued in 15 minutes or less. The

Alaskan center limits the warning area depending on earthquake magnitude and location, but the Hawaiian center, reflecting the smaller area it serves, issues regional warnings for the entire state.

The Alaska Regional TWS issues an immediate tsunami watch for

an area within a 200 mile radius of the epicenter of a coastal quake

38-4
with a magnitude of 6.75 to 7.0. A warning is issued for the rest

of the Alaskan coast. If the magnitude of a coastal earthquake

exceeds 7.75, a warning is issued for an area within 500 miles of

the epicenter and a watch is issued for the rest of the coast.

The Hawaii Regional TWS issues a warning for the state whenever a coastal earthquake of magnitude 7 or greater occurs in the region.

Planned TWS Improvements

For the past 3 years, the Pacific Tsunami Warning Center has

i>een using a time-share computer for message generation and epicenter location. A program has been added to compute estimated time

of arrival.

A new minicomputer was delivered to the PTWC this month

(May 1978) and installation and programming are proceeding.

This minicomputer will provide sufficient capacity for on-line

processing of seismic data telemetered in from the Hawaii Regional

TWS net. It is also planned to store data on past tsunamis in the

computer memory to aid the Warning System personnel in making

judgments on the necessity for issuance of watch and warning

informat ion.

The greatest weakness in the TWS is its inability to provide

forecasts of tsunami wave heights for threatened coastlines.

Current model studies of propagation and terminal effects have

proceeded to their useful limits. Data from open ocean gages

kre necessary to provide relatively undistorted tsunami wave

parameters for input to these models. With such data available

in real time, tsunami height forecasts may eventually be included

in warning messages.

To eliminate delays in data gathering occasioned by the necessity of contacting observers to interpret tide and seismic data

and then transmitting the data over what are often slow or unreliable communication circuits, the National Weather Service

las been developing, in conjunction with the U.S. Geological

Survey, a method of telemetering tide and seismic data via geostationary satellite. The tide platforms consist of an interroga-

:able radio and a data register that accepts data from a tide gage

and stores it until the data becomes obsolete. The first platform

installed at La Jolla, California, takes data from a Bristol timeduration tide transmitter, storing one data point every 30 seconds

and holding 40 data points in its register. A second platform,

:o be built by the University of Hawaii, will accept data from a pressure transducer to be installed off Wake Island at a depth of

311 meters.

The seimic platforms consist of an interrogatable radio and a lata register that accepts data from microprocessor that in turn

jaonitors the output of a standard seismometer and the arrival time

)f primary seismic waves. A variable combination of amplitude,

period, and duration gates provides a reliability of better than

$0% on the detection earthquakes of interest to the TWS. The platform uses time braodcast by the satellite, thus eliminating the

possibility of error or drift in an external clock.

The ultimate in tsunami warning will come when we are able to "eliably predict coastal and offshore earthquakes. Knowing in advance

uhe expected magnitude and relative fault movement, the TWS should

>e able to forecast tsunami and their heights on exposed coasts.

38-5
38-6
MEMBERS OF TASK COMMITTEES

TASK COMMITTEE

Strong Motion Instrumentation

Arrays and Data

B. Large Scale Testing Program

Repair and Retrofit of

Existing Structures

Structural Performance

Evaluat ion

Land Use Programs for

Controlling Natural Hazard

Effects

j F. Life-Line Systems

High Speed Wind Data

Soil Behavior and Stability

During Earthquakes

U.S. PANEL

^Dr. Gerry Brady

Mr. James Lander

Mr. Jerry Dodd

*Dr. E. V. Leyencecker

Dr. S. C. Liu

Dr. W. Shaw

Dr. H. S. Lew

Dr. R. McGuire

*Dr. John B. Scalzi

Mr. James Cooper

Mr. John J. Healy

Dr. E. V. Leyendecker

*Mr. Robert Fuller

Mr. Jack Meehan

Dr. John B. Scalzi

Dr. W. F. Marcuson, III

*Dr. S. T. Algermissen

Dr. M. P. Gaus

Mr. Jack Meehan

*Mr. C. F. Scheffey

Mr. Jack Meehan

Mr. J. D. Cooper

*Dr. T. D. Potter

Dr. R. Marshall

Mr. M. Changery

*Dr. W. F. Marcuson,

Mr. Jerry Dodd

Mr. D. Tiedemann

III

JAPANESE PANEL

*Mr. H. Tsuch ida

Mr. T. Iwasaki

Mr. K. Ohtani

Dr. M. Watabe

Dr. M. Ot suka

*Mr. K. Ohtani

Mr. T. Murot a Mr. N. Nar it a Mr. K. Sawada

Dr. M. Watabe

Mr. H. Tsuch ida

*Dr. K. Nakano

Mr. E. Kur ibayashi

Mr. T. Asama

Mr. Y. Shioi

*Dr

.

M. Watabe

Dr. T. Kunih ir

o

*Mr

.

E. Kur ibayash i Dr. H. Sato

Dr. S. Hattori

*Dr

.

T. Okubo

Mr. S. Fu j iwhar

a

Dr. S. Hayashi

Mr. T. Iwasaki

Mr. E. Kur ibayash i Mr. T. Murot

a

*Mr

.

S. Fu j iwhar

a

Dr. K. Naito

Mr. T. Murota

Mr. N. Narita

*Dr

.

M. Ohashi

Mr. T. Iwasaki

Mr. K. Sawada

Mr. H. Tsuch ida

Mr. Y. Shioi

39-1

TASK COMMITTEE REPORTS

TASK COMMITTEE ON STRONG-MOTION INSTRUMENTAL ARRAYS

AND DATA

ATENDEES: U.S. Side - A. G. Brady, Co-Chairman

J. F. Lander

M. P. Gaus

Japan Side - K. Ohtani, Co-Chairman

The task Committee's plan to continue their respective

efforts in the exchange of data is outline in the following six points.

a) After an earthquake which has caused damage to strucures or an earthquake during which maximum acceleration exceeding about 0.1 G has been recorded, the task

committee of a country where the earthquake has taken

place will provide a list of the strong-motion earthquake records for the counterparts of the task committee. The list contains maximum component acceleration

of each record. If there is such a list compiled by

any organization, the list mentioned above may be

replaced by it.

b) Every year the task committees will exchange catalogs

of the strong-motion earthquake records recovered in

the last year. The catalog contains maximum component

acceleration and wave forms of major records. If

there is such a catalog compiled by any organization,

the catalog mentioned above may be replaced by it.

c) The task committees make appropriate arrangements to provide digitizable copies of records when they are requested. In addition, arrangements will be made to provide information on the characteristics of the site

and structures at the locations where such records are obtained

.

d) When the organizations taking part in the panel's

publish reports on the strong-motion earthquake

records, the organizations will distribute copies of

the reports to the organizations of the panels interested in them. The task committees exchange list of

the organizations which wish to receive the reports.

e) Every year at the time of the joint meeting the task

committees exchange lists of reports on the strongmotion earthquake records and on analysis of the

records published in the last year. The list will cover all the reports on the above mentioned topics

published by the organizations taking part in the

panels and as many reports as possible published by

other organizations and universities.

f) The task committees will continue to exchange lists

of digitized data on all the major strong-motion

earthquake records recovered in both countries.

Besides the digitized records in form of lists,

an exchange of digitized records in form of computer

magnetic tapes, cards, and other media will be cont inued

.

A-1
II. The task committees plan to assist and cooperate, where

possible, in the following areas.

a) Assistance and cooperation with governmental organizations in other seismic areas, in order to promote

high quality strong earthquake motion observations

in all seismically active areas of the world.

b) Assistance and cooperation with any international

effort to record strong ground motions close to the source of a large magnitude shock.

III. Furthermore, in view of the preceding paragraphs, the

following item is appended.

The task committees applaud the efforts of the International Association for Earthquake Engineering and the

International Association of Seismology and Physics of

the Earth's Interior for instigating the International

Workshop on Strong-Motion Earthquake Instrument Arrays

May 2-5, 1978, in Honolulu, Hawaii, USA, and especially

congratulates the workshop for the wording of their

resolution. The exchange of complete information on

all aspects of the program, as it develops particularly

in Japan and the United States will be carried out in

the manner of our standard exchanges when appropriate.

The task committees feel the Panel should play a major

cooperative role in the implementation of relevant parts

of the specific recommendations in that resolution.

A2
B , TASK COMMITTEE ON LARGE-SCALE TESTING PROGRAMS

ATTENDEES: U.S. Side E. V. Leyendecker, Co-Chairman

E. 0. Pfrang

C. Culver

R. Fuller

C. Thiel

S. C. Liu

H. S. Lew

C. Scheffey

S. Takahashi

Japan Side - K. Oht ani, Co— Chairman

The status of the current NSF-funded project on planning

large scale structural tests was discussed. Co-chairmen

of this project are Dr. Penzien of the University of

California at Berkeley and Dr. Unemura of the University

of Tokyo. Membership from the U.S. and Japan sides

includes member of the UJNR Panel on Wind and Seismic

Effects, as well as others.

The first meeting of this NSF-funded planning group was held in Tokyo September 5-10, 1977. Topics discussed at that meeting included descriptions of some of the testing

facilities in both the U.S. and Japan, various methods of

testing, as well as the need for testing at large scale.

As a result of that meeting the following building types were agreed to for study and ranking of importance at a later meeting.

a) Reinforced Concrete

b) Structural Steel

c) Mixed Construction (steel and reinforced concrete)

d) Precast and/or prestressed reinforced concrete

e) Masonry

f) Timber

Study groups were established for each of these construction types. The second meeting of this planning group

was held in San Francisco May 15-19, 1978. At that meeting the construction types were categorized in order of

importance into three groups.

Group A - Reinforced Concrete, Structural Steel

Group B - Mixed Construction (steel and reinforced

concrete, precast and/or prestressed

reinforced concrete)

Group C - Timber and Masonry

These construction types were ranked according to importance in both countries and do not necessarily

reflect the major concerns in each country.

Topics discussed at the San Francisco meeting included

details of specific designs, needed component tests,

and needed analyses.

The third meeting of the NSF-funded planning group has

been schedule to be held in Tokyo December 18-23, 1978.

At that meeting detailed designs for a reinforced

B-1
concrete and a structural steel building, that is the

structures in Group A, will be- presented. These details

will be for a building seven stories in height.

The fourth and final meeting of the NSF-funded planning

group has not yet been schedule, but will include the

final recommendations for a first full-scale test to be

undertaken jointly by the U.S. and Japan, dependent upon

the availability of funds.

The final details for conducting such an international

program between the U.S. and Japan on large-scale testing

have not yet been completed. Discussions between the U.S.

and Japan thus far have resulted in agreements that the

joint research program will be conducted under the framework of the UJNR agreement and that the UJNR Panel on Wind and Seismic Effects will be used as the implementation and coordination mechanism for the joint program.

II. The following resolution was agreed to:

The Task Committee on Large-Scale Testing Programs should

keep the UJNR Panel on Wind and Seismic Effects informed

of development in large-scale testing programs. Programs

proposed by the current NSF-funded planning group should

be circulated to UJNR for review and comments with appropriate recommendations returned to the planning group for

consideration.

B-2
C. TASK COMMITTEE ON REPAIRS AND RETROFIT OF EXISTING

STRUCTURES

(Report from Meeting of May 1978)

ATTENDEES: U.S. Side - Dr. J. Scalzi, Co-Chairman

Mr. J. D. Cooper

Dr. J. Wright

Japan Side - Dr. M. Watabe, Co-Chairman

I. SCOPE: Determine analytical and experimental methods

for the repair and rehabilitation of buildings

and bridges.

II. OBJECTIVES: To coordinate research activities, to exchange data and information, and to determine research needs.

III. POINTS OF DISCUSSION:

a) The chairman of Task Committees B, C and D should

maintain close contact so that activities in the

respective committees could be coordinated.

One of the first tasks should be the development of a glossary of technical terms in both languages.

Technical papers of interest to the Committee should

be translated into both languages.

SCHEDULE OF ACTIVITIES :

a) U.S. to establish membership of the Task Committee as soon as possible.

b) U.S. to outline areas of interest for repair and

retrofit of buildings and bridges.

c) Send items 1 and 2 to Japan Chairman of Task

Committee for review and comment.

d) Japan will return data with their membership and

confirmation of areas of interest and commonality, and suggestions for other areas of mutual

interest

.

e) U.S. will confirm areas of mutual interest with

suggestions of priorities for specific topics

to pursue.

f) Japan will review topics and confirm priorities

for study.

g) At 11th Joint Meeting, respective Chairmen will

discuss specifics of the projects selected for

mutual activity and cooperation.

C-1

D. TASK COMMITTEE ON STRUCTURAL PERFORMANCE EVALUATION

ATTENDEES: U.S. Side G. Robert Fuller, Co-Chairman

John F. Meehan

John B. Scalzi

Richard D. McConnell

S. K. Takahashi

James K. Wight

Roland L. Sharpe

Japan Side - M. Watabe, Co-Chairman

T. Kubota

DISCUSSION :

a) The Report of the Task Committee Meeting held in

Tokyo on May 26, 1977, as prepared by Dr. John B. Scalzi and Dr. Makoto Watabe, was reviewed. The

Task Committee concurred in the general proposals

of the Report, and considered the Report useful as

a base for discussion in order to establish specific

recommendat ions

.

b) It was proposed that a Scope for the Task Committee

be formulated in order to establish priorities for

future activities. Possible implementation procedures were also suggested.

c) General coordination of activities with the task

committees was emphasized in agreement with discussions and reports of those Task Committees.

SCOPE OF TASK COMMITTEE (not prioritized) :

a) Evaluation of Existing Structures - Analytical and

experimental programs need to be initiated by both

countries to determine resistance of existing structures to high wind and seismic forces. Present

evaluation methodologies and techniques need to be

compiled and analyzed.

b ) Stochastic Risk Analysis and Probabilistic Hazard

Determination - Further studies need to be initiated

by both countries to develop an inventory of structures subject to possible failures and damage caused

by high winds and earthquakes so that proper risk

and hazard potential can be determined.

c ) Coordination of Large-Scale Testing Program with

Analytical Studies Necessary analytical studies are needed by each industry by each country for proper

planning before the physical testing, and to correlate results of physical testing afterward. Closer

coordination with the Large-Scale Testing Group

should be maintained.

d) Non-Structural and Structural Component Interaction - Studies are needed and data compiled on the effect of

nonstructural elements on the structural performance

of buildings. It is possible that performance of

buildings should be evaluated with respect to

D-1
non-structural elements as well as structural. The

Task Committee could then be changed to "Evaluation

of Performance of Structures."

e) Post-Earthquake and Post-Storm Evaluation of Perfor - mance of Structures - Investigative and eduational

efforts need to be better coordinated, both internally within each country and between countries.

f) Analysis, Strengthening and Retrofit of Structures - In coordination with task committee (C) , an earthquake and high wind information center should be

established to compile data on present technology.

Possibly the Interagency Discussion Group on Disaster

Mitigation in Washington, DC and governmental institu'

tions of Japan can assemble an annotated bibliography

on this subject.

Benchmark Structures - A methodology is needed to establish "benchmark structures to which future

proposed codes and standards can be compared. Structures need to be tested and analyzed to determine

actual resistance to high winds and earthquakes. (The

USGS-NOAA report on the 1971 San Fernando Earthquake,

Buildings in Los Angeles was referenced).

III. PROPOSAL :

a) This suggested "scope" should be studied by other

task committees for possible coordination.

b) The Task Committee will evaluate this "scope" plus

any comments by other task committees, to prioritize

the items prior to the 11th Joint Meeting.

c) Each side will endeavor to initiate discussion on

the various topics in the "scope" so that preliminary

planning may properly incorporate the suggested

studies

.

D-2
TASK COMMITTEE ON LAND USE PROGRAMS FOR CONTROLLING NATURAL

HAZARD EFFECTS

ATTENDEES: U.S. Side - Dr. Robin McGuire, Co-Chairman

Japan Side - Dr. S. Hayashi, Co-Chairman

The Task Committee's efforts are summarized in the

following

a) The resolutions of the 8th and 9th Joint Panel

Meetings are still in effect and will continue.

b) The exchange of papers and references on land use planning has been very useful and should continue.

c) The exchange of experts is useful both for the host

country and for the visiting expert; for instance,

the visit of Dr. McGuire to Japan in 1977 and the

expected visit of Mr. Tazaki to the U.S. in 1978

and 1979. These exchanges should be promoted.

d) Research and methods of land use planning to reduce

seismic hazards should be promoted more vigorously,

in view of the importance and potential savings of

life and property.

e) The lessons which can be learned about proper land use planning from post-earthquake investigations in

all areas are great. Therefore, both countries will

cooperate in investigation of effects of damaging

earthquakes

.

f) Research and methods of evaluating and mapping

future potential damage to various types of structures should be promoted, as a method of understanding the hazard in terms of the types of constructed

facilities at risk and their locations.

E-1
Ji
TASK COMMITTEE ON LIFELINE SYSTEMS

ATTENDEES: U.S. Side - C. F. Scheffey, Co-Chairman

Japan Side - T. Okubo, Co-Chairman

The members of the task group express satisfaction at the growing level of interest in both the U.S. and Japan

for lifeline earthquake engineering activity and for the

benefits of exchange, of information accomplished in this

past year. The meeting of the Task Group generated useful discussion of three issues:

a) Arrangements by which the work of the task group

committee may obtain wider participation of both

governmental and prof fessional groups.

b) Coordination of our work with other task groups of

the Panel on Wind and Seismic Effects established

along functional lines rather than by the class of

structure involved; and

c) Lifeline components and systems for which full-scale

tests are necessary.

As a result of this discussion, the following resolutions

were adopted:

a) The present areas of information exchange on design

standards and criteria, research information and

administrative regulations with regard to lifeline

safety should continue, with increased effort to include non-govenrmental standards and research data.

b) Both co-chairmen will explore expansion of the work

of the task committee by inviting additional agencies

of their respective governments having lifeline

responsibility to become involved and will exchange

proposals within six months for matching counterpart

agencies. A specific proposal will be presented to the Panel on Wind and Seismic Effects at its 11th

joint session.

c) The task commiteee will examine and investigate the

problems of lifeline earthquake engineering which

may be appropriate for full-scale tests, particularly

where analytical and component tests are inadequate.

A list of candidate tests will be prepared for the

ll.th joint meeting, and a specific plan for one or more projects within twelve months of that date.

d) The Lifeline Disaster Prevention Methods Task Committee will suggest to the other Task Committees, which

have responsibilities related to our work, appropriate

limitations of our efforts to avoid duplication. It will make specific recommendations to such task

committees as those responsible for Strong Motion

Instrument Arrays, Full Scale Testing and Soil

Behavior and Stability in Earthquakes for inclusion

in their programs. In general, our group will

confine itself to those problems peculiar and unique

to lifelines systems, such as aeroelastic effects on bridges

.
IiU
TASK COMMITTEE ON HIGH SPEED WIND DATA

ATTENDEES: U.S. Side - M. Changery, Co-Chairman

(for T. Potter)

M. Gaus

C. Barrientos

R. Marshall

Japan Side - T. Okubo, Co-Chairman

M. Kanai

EXISTING ITEMS :

a) Exchange documentation types and periods of record

of high wind data available from the respective

nation meterological services and special observation

sites

.

b) Determine the requirements for actual wind observations (documented) in #1) in digitized form. If

required, the Task Chairmen will establish procedures

to facilitate the exchange of such data.

c) Exchange available specialized high wind data sets in

digitized form to facilitate research on high winds

on structures. The information may include both

surface and higher meteorological data sets. Documentation on instrument characteristics, exposure

and elevation above ground may be included.

d) Refine procedures to exchange on a timely basis

records of future episodes of high speed wind which

cause loss of life and extensive property damage.

e) Request U.S. and Japanese side to prepare summary

reports, for future Joint Meetings, which contain

information on high winds and associated loss of

life and extensive property damage.

f) Promote research on the characterization of high

winds on structures.

g) Exchange data on wind pressure on buildings and

structures and on profiles of the free wind field.

h) Establish standard wind test methods for boundary

layer wind tunnels.

NEW ITEMS :

a) Encourage the interaction between meteorologists

and engineers to improve the specifications of wind

data required in the future use: in establishing

extreme wind distributions; in determining wind

loadings on structures; in understanding the urban

wind climate; and in considering wind energy usage

and factor in structure design.

TASK COMMITTEE ON SOIL BEHAVIOR AND STABILITY DURING

EARTHQUAKES

ATTENDEES: U.S. Side - W. F. Marcuson, III,

Co-Chairman

Japan Side - Mr. M. Okahara, Co-Chairman

Dr. S. Hayashi

RESULTS OF DISCUSSIONS :

a) Activities and Principal Accomplishments to Date

The task committee (both U.S. and Japanese sides)

has prepared and exchanged:

i) list of documents relating to soil liquefaction during earthquakes, and,

ii) list of soil dynamics laboratories.

At our last meeting, we also agreed to exchange

existing codes and guidelines for assessing soil

liquefaction. At present, we know of no such documents in the U.S. and the Japanese side transmitted

an outline of some of the typical provisions used

in Japan to the U.S. side.

b ) Future Programs

Guidelines for assessing soil liquefaction are currently in preparation in the U.S. This work is

being sponsored by the U.S., N.R.C., and will be

transmitted to the Japan side when completed.

A discussion of the possibility for exchange of data

and field crew and equipment to establish uniformity

for penetration resistance testing as conducted in

Japan and the U.S. produced the following:

i) we will attempt to organize, plan and initiate

such a study, and,

ii) we will attempt to secure funding for such a study, and,

iii) we recognize that such a study is complex and

therefore, may be difficult to execute; however, we will attempt to overcome these

difficulties

.

Discussions were held on:

i) Soil liquefaction potential for submerged

foundat ions

ii) Foundation settlement due to earthquake motion,

and

iii) Joint visits and exchange of information on embankment dams which have suffered damage or failure during earthquakes.

H-1
Items i) and ii) above were discussed and it appears

that both the U.S. and Japan have limited data which

is applicable. Therefore, future work in this area

is postponed until more data becomes available.

Discussions of item 3 indicates that Mr. T. Uwabe

of the Port and Harbour Research Institute plans to visit the Waterways Experiment Station for a period

of about one year. During his stay he plans to conduct joint research on the response of earth embankments to dynamic loads. It was also proposed that

site visits to case histories should be conducted

at the next UJNR Plan meeting in Japan. The Japanese

have one or maybe two case histories of embankments

which have suffered damage during the Near Izu Oshima

Earthquakes

.

H-2
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fTTLE AND SUBTITLE

[find and Seismic Effects

proceedings of the Tenth Joint Panel Conference of the U.S. -Japan Cooperative Program

in Natural Resources, May 23-26, 1978, Washington, D.C.

|\UTHOR(S)

H.S. Lew, Editor

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1' ABSTRACT (A 200-word or less factual summary of most si gnificant information. If document includes a si gnificant

[oibliography or literature survey, mention it here)

iTljs volume includes thirty eight technical papers presented at the Tenth Joint

Milting of the U.S. -Japan Panel on Wind and Seismic Effects. It also includes the forI resolutions and the reports of the Panel's eight task committees. The subjects

cjjered in the Joint Meeting include: (1) natural wind characterization and extreme

Iwijd records, (2) characterization of earthquake ground motions and strong-motion

ieilthquake data, (3) engineering seismology, (4) response of hydraulic and earth

is uctures to seismic forces, (5) structural responses to wind loading, (6) recent

jdi elopments in seismic design criteria, (7) design and analysis of special structures,

(i damage evaluation, repair and retrofit, (9) earthquake hazard mitigation, and (10)

s jrm surge and tsunami

L|l|CEY WORDS (Six to twelve entries; alphabetical order; capitalize only proper names; and separate key words by semicolons)

I :elerograph; codes; design criteria; disaster; earthquakes; ground motion;

:ismology; standards; storm surge; structural engineering; structural response;

Jpunami; v/ind loads; winds

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