Wind and Seismic Effects

Proceedings of the 24th Joint Meeting

NIST SP 843

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Wind and

Seismic

Effects

NIST SP 843

PROCEEDINGS OF

THE 24TH JOINT

MEETING OF

THE U.S.-JAPAN

COOPERATIVE PROGRAM

IN NATURAL RESOURCES

PANEL ON WIND AND

SEISMIC EFFECTS

Issued September 1992

Noel J. Raufaste,

EDITOR

Building and Fire Research Laboratory

National Institute of Standards and Technology

Gaithersburg, MD 20899

U.S. DEPARTMENT OF COMMERCE

Barbara Hackman Franklin, Secretary

TECHNOLOGY ADMINISTRATION

Robert M. White, Under Secretary for Technology

National Institute of Standards and Technology

John W. Lyons, Director
National Institute of Standards and Technology Special Publication 843

Natl. Inst. Stand. Technol. Spec. Publ. 843, 594 pages (Sept. 1992)

CODEN: NSPUE2

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1992

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, DC 20402-9325
PREFACE

BACKGROUND

The U.S.-JAPAN Cooperative Program in Natural Resources (UJNR) was created in January 1964. It is part

of the U.S.-JAPAN Cooperative Science Program. The objective of UJNR is to exchange information on

research results and exchange scientists and engineers in the area of natural resources for the benefit of both

countries. The UJNR is composed of 17 Panels each responsible for specific technical subjects.

The Panel on Wind and Seismic Effects was established in 1969. Seventeen U.S. and six Japanese agencies

participated with representatives of private sector organizations, to develop and exchange technologies aimed

at reducing damage from high winds, earthquakes, storm surge, and tsunamis. This work is produced through

collaboration between U.S. and Japanese member researchers working in 11 task committees. Each committee

focuses on specific technical issues, e.g., earthquake strong motion data. The Panel provides the vehicle to exchange technical data and information on design and construction of civil engineering lifelines, buildings,

and water front structures, and to exchange high wind and seismic measurement records. Annual meetings

alternate between JAPAN and the United States (odd numbered years in JAPAN; even numbered years in the U.S.). These 1-week technical meetings provide the forum to discuss ongoing research and research results; 1-week technical study tours follow the meetings.

The National Institute of Standards and Technology (NIST) provides the U.S.-side chairman and secretariat. The Public Works Research Institute (PWRI), JAPAN, provides the JAPAN-side chairman and secretariat.

Cooperative research is performed through formal Panel Programs. In 1981, cooperative research in LargeScale Testing was started under the auspices of the Panel. Also in 1981, joint research on Reinforced Concrete

Structures was initiated. Full-scale testing was performed at the Building Research Institute (BRI), one of the

six Japanese member organizations, with supporting tests in JAPAN and in the United States. Two years later a joint research program on Steel Structures was initiated. Full-scale testing again was led by BRI with supporting tests in the United States and JAPAN. The U.S.-JAPAN coordinated program for Masoniy Building Research was started in 1985. A U.S.-JAPAN coordinated program on Precast Seismic Structural Systems was

initiated in 1991.

Task Committee meetings, exchanges of data and information through technical presentations at annual Panel

meetings, exchanges of guest researchers, visits to respective research laboratories and informal interactions

between Panel meetings, joint workshops and seminars, and joint cooperative research programs all contribute

to the development and effective delivery of knowledge that has influenced design and construction practices

in both countries.

Direct communication between counterpart country organizations is the cornerstone of the Panel. Effective

information exchanges and exchanges of personnel and equipment have strengthened domestic programs of

both countries. There are opportunities for experts in various technical fields to get to know their foreign

counterparts, conduct informal exchanges, bring their respective views to the frontiers of knowledge, and

advance knowledge of their specialties.

The Panel's results have supported improvements in practices in both countries. They have:

1. created and exchanged digitized earthquake records for use as the basis of research for Japanese and

U.S. geotechnics and structures;

2. produced data that advanced U.S. design and construction of bridge columns;

3. produced large-scale testing data that advanced seismic design standards for buildings;

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4. created a database comparing Japanese and U.S. standard penetration tests to improve seismic design

criteria for soil liquefaction;

5. created databases on storm surge and shore line interaction and on tsunamis and tsunami warning

systems for use by designers to verily mathematical models of tsunamis and storm surge.

HIGHLIGHTS OF THE 24th JOINT PANEL MEETING

Thirty-six papers were presented at the joint meeting (45 papers were authored), organized into five themes:

wind engineering, storm surge and tsunamis, joint cooperative research program, earthquake engineering, and

four reports on the Panel's Task Committee workshops of the past year.

Highlights include:

• Technical presentations:

highlighted important work by the United States and Japan. U.S. research in wind engineering is enhanced by links to the extensive research performed by Japan's Public Works Research Institute. Wind hazard mitigation efforts lag seismic in both Japan and the United States.

- important knowledge was exchanged for research and practice, e.g., wind tunnel and analytical modeling for coupled torsional and flexural response of suspension bridges, field and analytical studies of

the performance of pile groups, large scale finite element modeling of storm surges validated with

historical data.

- information was gained about extensive Japan and U.S. public works projects and civil engineering

research and their applications into practice.

• Base isolation systems and high performance concrete: Excellent work is being performed by both sides. Panel participants advance their competence by becoming more familiar with these research projects and

through direct participation in appropriate research.

• Seismic strengthening: Studies of assessment, research, design, and construction for seismic retrofit of

buildings and highway structures provide for private and public sector, U.S. and Japanese collaborations

in making seismic retrofit more cost effective.

• Recent earthquakes: Report on reconnaissance of the Erzincan Turkey Earthquake of March 13, 1992

shows again the need to address the vulnerability of nonductile concrete and masonry buildings.

• Publications:

- Each-side of the Panel published a catalog, List of Publications 1969-1991 Panel on Wind and Seismic

Effects. The NIST Special Publication 835, April 1992, contains titles, keywords, and author index of over 1200 Panel papers published during its 23-year life. Over 1000 copies will be distributed to developing countries in support of the International Decade for Natural Disaster Reduction (IDNDR),

- NIST published a two-volume series, Earthquake Resistant Constmction Using Base Isolation, NIST

SP 832-1 and 832-2, April 1992; the text was translated from Japanese.

iv
TECHNICAL SITE VISITS

The second week (23-29 May), was devoted to technical reviews of laboratories, public works projects,

and construction sites. In addition to the 36 papers presented during the first week; eight papers

continued the Panel's technical presentation during the second week. These presentations were

equally divided between the U.S. and Japan-sides. Three of the four U.S. papers are included in these

Proceedings. The four Japan-side presentations and one U.S. presentation were specially prepared for two technical site visits and are not part of the Proceedings.

Eight sites were visited:

1. LEHIGH UNIVERSITY, ADVANCED TECHNOLOGY FOR LARGE STRUCTURAL

SYSTEMS (ATLSS) MULTIDIRECTIONAL EXPERIMENTAL LABORATORY

The delegation visited, and discussed in detail, work on evaluating the yield strength of high strength

structural steel and crack propagation in ship hulls to examine fatigue crack growth behavior in large

scale box sections of high strength low-alloy grade 80 steel. Also, work was observed in fatigue strength

of HSLA 80 welded steel to evaluate its characteristics in double hulled ships. Some work is underway

in base isolation for buildings and bridges.

Presentations by ATLSS staff and Japan-side Panel members addressed a variety of topics: ATLSS's

work in developing consistent design methods and models for precast concrete connections for unified

connection design based on codification and rules for knowledge-based systems used to integrate dif- ferent countries' rules into structural models. Japanese presentations centered on reviewing their base

isolation and active controls work including applications for buildings and bridges and on Japanese

structural engineering facilities. The Japan-side identified a variety of field experiments used to re- duce seismic loads using mass damping, rubber bearings, viscous oil dampers, springs, sliding isolators,

etc.

The ATLSS laboratory can accommodate three-dimensional testing of large structures. Its main bay

is 91 m (300 ft) long with a test floor of 31 x 12 m (102 x 40 ft); its buttress wall is 15 m (50 ft) high.

The large-scale multi-directional facility has a capacity of applying 22.2 MN (5M lbs) in compression.

The pumps are located outside the facility to reduce noise.

2. STENNIS SPACE CENTER

The delegation viewed the three massive ground rocket engine propulsion test facilities. The largest

stand that houses the main rocket engines during test firing is 122 m (400 ft) high and is fastened to the ground by 49 m (160 ft)-long anchors. This impressive construction project illustrated technologies

used to solve a major civil engineering challenge. Testing of advanced solid rocket motors will begin

in 1994. The stands are linked by 12.1 km (7.5 mi) of canals used to transport the rocket stages and

propellants. High pressure water from a 242 m^ (64 M gal) tank cools the stands during rocket engine

test firings.

The Stennis Space Center's mission is to support the testing of the Space Shuttle main engine and

large propulsion systems; conduct research in land and oceanographic remote sensing, and manage

the 5.3 K ha (13 K acre) Stennis facility and provide support services to the 19 Federal and state agencies occupying the site.

3. CORPS NEW ORLEANS DISTRICT HEADQUARTERS

An introductory briefing provided background on the Corps' New Orleans District's mission to mitigate floods in the lower Mississippi Valley through flood control systems as: levees for containment;

floodway channel improvements and stabilization for passage of excess waters; tributary basin improvement for major drainage and flood control through dams and reservoirs, pumping plants, auxil- iary channels, environmental enhancements; wetland protection; and providing technical support to

V
appropriate public and private sector organizations. Since 1879 the Corps has had lead responsibilities

in flood and navigation control in this region — the largest drainage basin in the U.S. The Corps has

developed a planned defense system to contain a flow of 84 K mVs (3 M ftVs). Soon, the Corps will undertake a new mission assignment to provide program management and technologies to assist EPA

perform toxic waste cleanup in this region.

During the morning, three technical presentations were made; two by the U.S.-side and one by the

Japan-side. The first, a Corps paper discussed physical model test programs and computer models

under development that improve prediction of beach profile change during storms and cross-shore

sand transport processes. The second, also a Corps paper addressed the importance of non-linear

response to seismic loading of concrete gravity dams. The Japan-side presentation addressed mitigating wave damage in harbors using innovative wave dissipators. A variety of dissipators are being tested

and new concepts are being modeled to identify their most favorable performance attributes.

The Japan-side presentation was specially prepared for this meeting; it is not part of the Proceedings.

Several related papers on dissipating wave energy using breakwater structures are found in past Pro- ceedings of the Panel on Wind and Seismic Effects. A Corps boat provided the delegation with a river tour to allow a direct review of the technologies

and projects for flood control and safe and efficient navigation. The trip down the Mississippi River

and two connecting waterways showed many public works projects including the Algiers and Harvey

Locks (sector and miter gates), spill way projects, levees and flood control structures and pumping

stations, shallow draft channels, docks and waterfront structures. MINERAL MANAGEMENT SERVICES AND FREEPORT-MCMORAN GLOBAL

RESOURCE COMPANIES

MMS is responsible for leasing land for mineral exploration and for regulating environmental and

human safety for oil/gas/mining. MMS performs this work in the Gulf of Mexico using 60 field inspectors and 12 helicopters. MMS adopts American Petroleum Institute's Standards, if practical, and at times further strengthens API's Standards with MMS' Standards. MMS works closely with other

Federal agencies in performing its mission such as: EPA in the assessment of discharges of contamination into water and on land; the Coast Guard for ensuring safety of personnel associated with

offshore operations; and the Corps who issues permits for certain pipelines and for abandoning

platforms.

After these discussions a helicopter transported the delegation about 160 km (100 mi) to the

Freeport-McMoRan Offshore Sulfur Platform. The Complex is composed of 14 platforms extending

for 1.8 km (1.1 miles), tip-to-tip. It is the largest sulphur mine in the U.S; reserves total 68B kg (75M

tons). The design and construction costs of these huge platforms exceeded $750M. They are designed

for a 40 year life. The complex sets on top of a salt dome located about 670 m (2200 ft) below water

level. Sulphur is under the dome and natural gas and oil reserves surround the 3.2 km (2 mi) diameter

dome.

The main product of this production platform is sulphur. Sulphur is mined using superheated water

at 157 °C (315 °F) injected into the sulfur bed, forcing it out and into transport tubes to two large

storage tanks on its own platform where it is prepared for barging to the mainland. The storage tanks

hold a four-day supply of the liquid sulphur. Oil and gas also are mined at this site. They are trans- ported by underwater pipe to shore. Another platform contains the living quarters for 150 persons,

the boiler plant platform has daily production capability to generate 38 K m' (10 M gal) of fresh water

and 17 mW of electricity. The plant went into operation in April 1992; sulfur production is expected

to reach 5Mkg (5.5k tons) per day using 30 wells.

vi
The platform substructures are about 70m (230 ft) in length. They were constructed horizontally,

transported to the site, and slid off transport barges into the water. Their 2.1 m (84 in) outer legs and

1.2 m (48 in) central legs were flooded bringing the platform into a vertical position. Stability pilings,

107 m (350 ft) in length, were sunk into the sea bed through the center and outer platform legs. The

outer pilings are 1.8 m (72 in) in diameter; the center pilings are 1.5 m (60 in). Thirteen bridges

connect the 14 towers. Over 90.7 M kg (100 K tons) of fabricated steel were used to fabricate the

complex. The two sulphur production platforms will be moved every 10 years to drill for new supplies

as the seafloor will subside approximately 9.1 m (30 ft) due to removing the minerals from the earth.

The platforms are designed for 40 year life based on API RP 2A Recommended Practice. They are designed for the North Sea environment (safety index of 3.6). Its environmental criteria are based on

1000 year storm with 20 m (67 ft) waves. Even waves from a 5000-year storm will not reach the base

of the 34 m (110 ft) high platform. Subsidence consideration included designing for heavy loads which

used specially designed joints and bearings to accommodate expansions and contractions. Design for

vertical waves included probability-based design load, design for soil subsidence, 80 year fatigue life, 445 k Nm^ (100 ksi) steel in piles, nonlinear pile analysis, vertical wave fatigue analysis, post-sleeve

jacking system, adjustable boat landing and barge bumpers, and very long bridges. Their corrosion

safety factor considers 6.4 to 9.5 mm (1/4 to 3/8 in) scaling.

5. CALIFORNIA DEPARTMENT OF TRANSPORTATION (CALTRANS)

CALTRANS is reinforcing bridges designed prior to the 1971 San Fernando Earthquake to improve

their ability to withstand future shaking as experienced in 1971 and the 1989 Loma Prieta earthquakes. CALTRANS sponsors some research with the University of California at San Diego and at Berkeley. For the latter, the Civil Engineering Department is testing retrofit methods for double level bridges and column bents. After the 1987 Whittier earthquake, analysis showed that attention must

be focused on retrofitting bridge columns since many suffered shear failures. CALTRANS is retrofitting those structures that pose the greatest risk to damage from another earthquake. Their end

goal is to retrofit 750 bridges between 1989 and 1994. A site visit was made to the Southern Freeway viaduct (Route 280 near Alemany), San Francisco. This

bridge experienced damage during the Loma Prieta earthquake. This portion of the viaduct was

closed to vehicles following the earthquake. Its columns are being replaced; work involves shoring the

upper deck, cutting the column at its cap and installing a new reinforced concrete column structure. Some column footings are being replaced, others are being strengthened with new footings. Piles will be used to withstand high tension loads and compression loads in anticipation of similar future loads

as experienced during the Loma Prieta earthquake. Several techniques are being explored and five companies were contacted to learn about their new innovations and construction techniques to meet

the design load criteria for pilings. There is much public pressure to complete this job by the end of

1992. CALTRANS hopes single lane traffic will be possible by December 1992, with full highway use by May 1993.

6. LAWRENCE BERKELEY LABORATORY AND UNIVERSITY OF CALIFORNIA, BERKELEY (UCB)

The U.S.-side presented two papers: the first was on scale models of buildings to help visualize structural behavior during an earthquake; the second addressed developing a seismic safety program for research laboratory facilities. These oral presentations well illustrate the need for effective earthquake technology translation for use by practitioners. Following these presentations, several LBL and

UCB buildings were visited:

Advanced Light Source (ALS). The ALS is nearing completion for use in performing advanced studies of materials, surface sciences in atomic and molecular physics and fabrication technologies, and

characterization for micro- and nanostructures. The delegation viewed the installation of electron

storage rings, several electron beam ports, and the facility's thick shielded walls. The central radial area has been made earthquake resistant. The ALS is in a 0.8 ha (2 acre) building.

vii
UCB, Structural Test Laboratory. Here research is underway on evaluating the expected perfor- mance of bridge retrofits, design criteria, and detailing. CALTRANS sponsors research on performing proof of concept. A brief discussion on seismic retrofit of two LBL buildings was provided. A

review of this subject is published as "Safety and Economic Benefits Realized From Upgrading the

Lawrence Berkeley Laboratory," Wind and Seismic Effects Proceedings of the 23rd Joint Meeting,

NIST SP 820, September 1991, pp. 225-237.

Seismic Retrofit of University Hall. A tour was made of the seismic retrofit of the 61 x 21m (200 x 70

ft), seven-story University Hall. Symmetrically arranged X-braced steel frames were placed in the

building's bays. This approach was selected because it permitted building occupants to continue working during the construction and the cost for this approach was the lowest. The X-braced frames are attached to the columns transferring the seismic loads through an innovative system from the bracing

and vertical steel down to the strengthened concrete columns.

UCB, Richmond Field Station. This off campus facility houses the Earthquake Engineering

Research Center (EERC) and its information center that for more than 20 years has served the earthquake engineering community with technologies. The library offers on-line data base research and

rationales, EERC News , etc.

The Station is upgrading its shake table for three-dimensional accelerations. The shake-table was constructed in 1972. Its dimensions are 6.1 x 6.1 m (20x20 ft) with a 45.4 K kg (50 ton) capacity. The

Station closely works with the UC Berkeley, Department of Civil Engineering.

The delegation visited one of the Station's laboratories performing research in base isolation. This

project is a three-story 6 m x 10 m (20 x 33 ft) reinforced concrete residence hall at the Tohoku University campus, Sendai. The research budget is funded by Shimizu Corporation. Work involves per- forming vertical and horizontal 1.5 g acceleration of light models and 0.8 g for heavy models. The

models are of 1/2.5 scale reinforced concrete building. The work involves performing low-level earth- quake tests of the model using three different base isolation systems; performing high level earth- quake tests of the scale-model using one base isolation system; conducting analytical studies of

structures using linear and non-linear techniques; and performing mechanical tests of the three differ- ent base isolators.

FORELL/ELSESSER ENGINEERS, INC.

This consulting engineering firm specializes in structural and earthquake engineering. Three principals discussed their analyses, options, and recommendations for rehabilitating historic buildings damaged by the Loma Prieta earthquake: the San Francisco City Hall, Stanford University Green

Library-West, and the Oakland City Hall. A fourth building, Lawrence Livermore National Laboratory, Building 111, which was retrofitted in the late 1980s, was briefly discussed.

Analyses performed on the historic buildings and the several options for seismic strengthening were

discussed. For the first and third buildings, base isolators were selected as the prime method to protect the buildings from future earthquake loads. The second building was repaired using reinforced concrete structural walls. Each building repair and strengthening must have a complete lateral loadresisting system transferring loads into the foundations. New materials, seismic force levels, and construction must comply with the latest building standards and codes.

San Francisco City Hall. A five story building of plan dimensions about 91 x 122 m (300 x 400 ft) has

a dome about 91 m (300 ft) high. The building was constructed in 1916. There is much ornate finish within the building; light courts are on the second floor. The building's structure is steel frame with

reinforced concrete slabs. Exterior granite walls and light court walls are unreinforced brick masonry

and infill partition walls are hollow clay masonry tiles. These areas suffered the majority of the earthquake damage. The structural steel frame was undamaged; no structural instabilities or collapse were

• • • Vlll
observed. Over 500 base isolators are proposed for installation to reduce loads; some shear-resisting

elements are required to strengthen and stiffen the superstructure. A new ground floor of steel framing concrete/metal deck will be located above the isolators.

Stanford Green Library-West. The library was constructed in 1917, in the neo-classical style; a fourstory building of 70 x 55 m (230 x 180 ft). It is designed around four light courts and central rotunda.

Reinforced concrete spread footings beneath the interior steel columns and perimeter wall footings

form the foundation. The building has a moment resisting steel frame encased in concrete. Its exterior

unreinforced brick masonry walls and hollow clay tile partitions were severely damaged. The building

has little ductility; it experienced significant earthquake damage. It lost 40-60 percent of its lateral strength due to the large number of wall openings.

The design team recommended using new reinforced concrete structural walls 0.3-0.4 m (12-16 in) in thickness, new reinforced concrete floors to reduce adding new concrete walls and to complete the

load path, new foundation for the wall loads, and strengthening the masonry piers. Base isolators are not practical because the library would lose its basement.

Oakland City Hall. It has plan dimensions of about 38 x 56 m (125 x 185 ft) and a highly decorated

clocktower rising to 99 m (325 ft). When completed in 1914, this building was the tallest in the western

U.S. Large interior cavities within the steel frame were enclosed by unreinforced brick masonry, unreinforced concrete, and hollow clay tile. Its foundation is a 0.7 m (2 foot 3 in) reinforced concrete mat over the entire basement area. The earthquake loading was resisted by a combination of the

building's flexible steel moment frame encased in brick masonry, unreinforced brick walls, and hollow

clay tile walls. The City Hall suffered much damage from the Loma Prieta earthquake. Ground motion was estimated at about 10% g.

Its irregular vertical structure consisting of several setbacks contributed to whipping action producing

cracks in the clock tower. An interesting side light; information was uncovered that accelerometers

had been installed in the building providing data from the earthquakes on 22 March 1957. This was

valuable information used in the redesign of the City Hall.

The City of Oakland recognizes this building as a historical landmark; accordingly, it wanted as few

surface changes as possible to preserve the building's interior and exterior walls. Over 110 elastomeric

bearing base isolators will be installed below the first floor girders under the perimeter columns in conjunction with the concrete shearwalls. The superstructure will be strengthened to resist maximum

forces above the isolators. Top portion will be braced to resist lateral loads and shear walls and braces

in walls are expected to prevent overturning. The building will be decoupled from its base isolators to accommodate an upward 6mm (1/4 in) unloading.

Repair and strengthening of the City Hall provided the City with the opportunity to perform a much

needed exterior cleaning, spruce up its interiors, modify its HVAC systems, add new carpeting, etc.

Lawrence Livermore National Laboratory, Building 111. This building was built in the early-1960s

with a reinforced concrete non-ductile frame. After the 1980 earthquake the design team had added

two buttress towers on piers. Now forces are transferred from the building to the buttresses. Steel elements were constructed on either side of the floor beams. Elements were bolted and epoxied to the floor beams with cast-in-place end caps in buttress. SAN FRANCISCO BAY-DELTA TIDAL APPROACH MODEL

The Corps of Engineers constructed this 0.4 ha (1 acre) hydraulic model basin to simulate water levels, flow patterns, and salinity in the San Francisco Bay and Delta. The model is a scaled representation of the bay; 27 km (17 mi) into the Pacific Ocean; 27 km (17 mi) east toward Sacramento; and

51 km (32 mi) south of Stockton. The model provided similitudes for the dominant forces. It is dis- torted by a factor of 10 in the vertical scale. Copper strips are used throughout to correct distortion

of the hydraulic efficiency of the flow. The model is 98 x 122 m (320 x 400 ft).

ix
The model predicts the impact of deepening and realigning navigation channels, varying flows, small

changes from planned construction projects. Results help better determine salinity changes from

deepening navigation channels, failure of levees and delta island flooding, and proposed water transfer projects; sediment disposal sites, breakwaters, and municipal approved discharge projects; and

emergency cleanup procedures in the event of an oil spill.

Both dynamic and steady-state testing are performed. Computers control tides and check on boundary

conditions. Tides are generated by a valve control. The model was designed and constructed in conjunction with WES.

ACKNOWLEDGMENTS

The preparation of this publication was made possible by the Department of Army, Corps of

Engineers; Department of Commerce, National Oceanographic and Atmospheric Administration and

the National Institute of Standards and Technology; Department of Energy; Department of Interior,

Mineral Management Services; Department of Navy, Naval Civil Engineering Laboratory; Department of State, and the Federal Emergency Management Agency; the National Science Foundation;

and the Nuclear Regulatory Commission.

Noel J. Raufaste, Secretary

U.S.-Side Panel on Wind and Seismic Effects

X
ABSTRACT

This publication is the proceedings of the 24th Joint Meeting of the U.S.-Japan Panel on Wind and Seismic

Effects. The meeting was held at the National Institute of Standards and Technology, Gaithersburg, Maryland,

during May 19-22, 1992. The proceedings include the program, list of members, panel resolutions, task committee reports, and 45 technical papers.

The papers were presented under five themes: (I) — Wind Engineering, (II) — Storm Surge and Tsunamis,

(III)— Joint Cooperative Research Program, (IV) — Earthquake Engineering, (V) — Summaries of Task

Committee Workshop Reports (oral presentations only).

KEYWORDS: accelerograph; bridges; building technology; concrete; design criteria; disasters; disaster

reduction; earthquakes; geotechnical engineering; ground failures; lifelines; liquefaction; masonry; repair and

retrofit; risk assessment; seismic; soils; standards; storm surge; structural engineering; tsunamis; and wind

loads.

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CONTENTS

PAGE

PREFACE iii ABSTRACT xi AGENDA FOR 24TH JOINT UJNR MEETING xvii

LIST OF PANEL MEMBERS xxv

LIST OF TASK COMMITTEE MEMBERS XL!

RESOLUTIONS XLIV

IDNDR PRESENTATION

U.S. Federal Agency Role in the International Decade for Natural Disaster

Reduction and UJST Workshops for Natural Disaster Reduction 3 Paul Kilho Park

THEME I: Wind Engineering

Flutter Characteristics of Super Long-Span Suspension Bridge 13 Koichi Yokoyama, Tokida Kanazaki, and Masahiko Yasuda

Predicting Wind Forces Using Computer Simulation of Air Flow Around Bridges 25 Okey Onyemelukwe and Harold Bosch

THEME II: Storm Surge and Tsunamis

A Numerical Simulation of the Tsunami Generated by Landslide of Mt. Mayuyama

in 1792 41 Chiaki Goto and Tomotsuka Takayama

Empirical Simulation of Storm Histories for Coastal Design 45 Martin C. Miller and Robin D. Reinhard

THEME III: Joint Cooperative Research Program

The Japan-PRESSS Program -Analytical Tools in Japan 57

T. Okoshi and S. Nakata

U.S. Japan Cooperative Research in Precast Seismic Structural Systems (PRESSS)

Program — Status Report 69

Nigel Priestly and H. S. Lew

xiii
CONTENTS

PAGE

Seismic Response Control of Highway Bridges by Variable Damper 71 Kazuhiko Kawashima, Shigeki Unjoh, and Hideyuki Shimizu

THEME IV: Earthquake Engineering

Seismic Design Considerations for Jordanelle Dam 87

John A. Wilson, Perry J. Hensley, William O. Engemoen, and

Francis G. McLean

An Evaluation of Seismic Forces Acting on Embankment Dams and Its SimpiiOed Prediction Procedure 101 Tadahiko Fujisawa, Nario Yasuda, Norihisa Matsumoto, and Kenji Yamabe

Resolving Uncertainties in Seismic Hazard Models 115

J. K. Kimball

Earthquake Alert System Feasibility Study 127

Philip E. Harben

Guidelines for Seismic Design Methods of Large Underground Structures 143 Hajime Asakura, Kazuhiko Kawashima, and Hideki Sugita

Seismic Vulnerability and Impact of Disruption of Lifelines in the Conterminous

United States— An Overview 159 William S. Bivins

1 Design Criteria for Base Isolation of Buildings 165 Harry W. Shenton, III

Manual for Menshin Design of Highway Bridges 173 Kazuhiko Kawashima, Kinji Hasegawa, and Hiroyuki Nagashima

Friction Controllable Bearings for Sliding Base Isolation Systems 189

Maria Qing Feng and Masanobu Shinozuka

State-of-the-Arts on Building Structures With Structural Response-Control

System in Japan 199

Yosikazu Kitagawa and Mitsumasa Midorikawa

Status of NSF Structural Control Research Programs 215

S. C. Liu, H. J. Lagorio, and K. P. Chong

Significant Results From Standardization Work on Rubber Bearings for Base-Isolation Buildings 225

Yuji Ohashi and Hiroyuki Yamanouchi

xiv
CONTEOTS

PAGE

Effect of Temperature on the Dynamic Behavior of Base-Isolated Bearings 233

Osamu Nakano, HUdeyuki Taniguchi, and Hiroaki Nishi

Dynamic Centrifugal Model Tests of Embankments on Liquefiable Ground

(Part 2) 241

Yasuyuki Koga, Jun-ichi Koseki, and Akihiro Takahashi

Earthquake-Induced Landslides in the Santa Cruz Mountains 253

M. E. Hynes and M. Hudson

Study on Prediction of Lateral Ground Flow by Soil Liquefaction and Its Influence on Piles 271

Ken-ichi Tokida, Hideo Matsumoto, Ikuo Towhata, and Yasushi Sasaki

Horizontal Bearing Capacity of Pile Foundations Based on Horizontal

Loading Tests 285

Minoru Fujiwara, Michio Okahara, Shoichi Nakatani.Yoshitomi Kimura,

and Shitgeru Takagi

! Seismic Effects on Port and Harbour Structures 301

Hajime Tsuchida and Setsuo Noda

Impfementation of Executive Order 12699, "Seismic Safety of Federal and

Federally Assisted or Regulated New Building Construction 311

Richard N. Wright

U.S. Army Corps of Engineers Earthquake Engineering Research Program 317

W. D. Roper, M. E. Hynes, and R. F. Davidson

A Progress Report on the Worldwide Earthquake Risk Management (WWERM)

Program 329

S. T. Algermissen, Walter W. Hays, and Paul R. Rrumpe

Development of Advanced Reinforced Concrete Buildings Using High-Strength Concrete and Reinforcement Researches on Structural Performance

Conducted in 1991 337 Tatsuo Murota, Hisahiro Hiraishi, Takashi Kaminosono, Masaomi Teshigawara,

Hitoshi Shiohara, and Hideo Fujitani

Dynamic Testing of Reinforced Concrete Frames - 359

Pamalee Brady and Steven C. Sweeny

Shaking Table Collapse Tests of RC Columns * .„ 371

Chikahiro Minowa, Nobuyuki Ogawa, Tadashi Mikoshiba, and Norio Qyagi , A Hollow Clay Tile Wall Seismic Performance Program Oven/iisw 383

XV
CONTENTS

PAGE

The Valuation and Designation Systems for Structural Calculation Programs

in Building Engineering 393

Keiichi Ohtani

Highlights of the 13 March 1992 Erzincan (Turkey) Earthquake 397

Mehmet Celebi

Beach Profile Response to Storm Surge and Waves 411

Nicholas C. Kraus, Magnus Larson, Bruce A. Ebersole, and Jane McKee Smith

Developing a Seismic Safety Program 427

Donald G. Eagling

Desk-Top Model Buildings for Dynamic Earthquake Response Demonstrations 435

A. Gerald Brady

Manuscripts Authored for Panel Meeting but Not Presented Orally

Determination of Windbome Missile Design Parameters 447

James R. McDonald

Advances in Wood Engineering and Construction Research 459

Erwin L. Schaffer

The Impacts of Siting Lifelines in Close Proximity Upon Their Earthquake

Vulnerability 491

William Bivens

Nonlinear Research Needs for Concrete Gravity Dams 495

Robert L. Hall and Wayne G. Johnson

Seismic Risk Assessment Methodology for Water Systems 507

Craig E. Taylor and Stuart D. Werner

Bridge Scour/An Element of Bridge Management 513

J. Sterling Jones and Stanley R. Davis

U.S. Coordinated Program for Masonry Building Research — Seventh Year Status 521

James L. Noland

Appendix: Task Committees A-K Reports 529

xvi
AGENDA FOR 24TH JOINT UJNR MEETING

Tuesday 19 May

1000 OPENING CEREMONIES

(Lecture Room B, Administration Building)

Call to order by Noel RAUFASTE, Secretary-General U.S. Side, UJNR Panel

Opening remarks by John LYONS, Director, National Institute of Standards and Technology

Remarks by Yukihide HAYASHI, Counsellor for Science and Technology, Embassy of

JAPAN

Remarks by Richard N. WRIGHT, Chairman U.S.-Side, Panel on Wind and Seismic Effects,

Director, Building and Fire Research Laboratory

Remarks by Yukihiko SUMIYOSHI, Chairman JAPAN-Side, Panel on Wind and Seismic

Effects, Director-General, Public Works Research Institute

Introduction of U.S. Members by U.S. Panel Chairman

Introduction of JAPAN Members by Japan Panel Chairman

Elect Joint Meeting Chairman

Adopt Agenda

Adjourn ,

1115 Group Photograph

1130 Lunch: Hosted by the National Institute of Standards and Technology, John LYONS,

Director

IDNDR

1230-1300 IDNDR, U.S. Federal Agency Role in IDNDR, Kilho PARK*, NOAA

1300-1310 Discussion

THEME -WIND ENGINEERING

1310-1410 Technical Session—Wind Engineering

Chairman: Yukihiko SUMIYOSHI

1310-1330 Flutter Characteristics of a Super Long-Span Suspension Bridge, Koichi YOKOYAMA*,

Tokida KANAZAKI, and M. YASUDA, PWRI

1330-1350 Predicting Wind Forces Using Computer Simulation of Air Flow Around Bridges, Okey

U. ONYEMELUKWE* and Harold R. BOSCH, FHWA

1350-1410 Discussion

1410-1430 Break

XVll
TASK COMMITTEE MEETINGS

1430-1700 T/C Meetings

T/C A Strong Motion Instrumentation Arrarys and Data (Brady/ Noda)

Employees Lounge

T/C C«feD Repair and Retrofit of Existing Structures and Evaluation of Structural Perfor- mance (Chong- Fuller/Yamanouchi) Lec. Rm. B

T/C E Ground Motion and Seismic Design Forces (Algermissen/ Kawashima) Confer- ence Room, Bill

T/C H Soil Behavior and Stability During Earthquakes (Franklin/ Yokoyama) Confer- ence Room B113

T/C J Wind and Earthquake Engineering for Transportation Systems (Cooper/

Fujiwara) BFRL Conference Room, Bldg 226, Rm. B221

1700 Conclusion of day 1

Wednesday 20 May

THEME -STORM SURGE AND TSUNAMIS

0830-0930 Technical Session— Storm Surge and Tsunamis

Chairman: Richard WRIGHT

0830-0850 A Numerical Simulation of the Tsunami Generated by Landslide of Mt. Mayuyama in 1792, C. GOTO and Tomotsuka TAKAYAMA, PHRI (Presented by Setuo NODA)

0850-0910 Empirical Simulation of Storm Histories for Coastal Design, Martin C. MILLER* and Robin

D. REINHARD, WES

0910-0930 Discussion

0930-0950 Break

THEME -JOINT COOPERATIVE RESEARCH PROGRAM

0950-1050 Technical Session— Joint Cooperative Research

Chairman: Richard WRIGHT

0950-1010 Development of Precast Seismic Structural System — Interim report, Tastuo MUROTA,

Shinsuke NAKATA, M. TESHIGAWARA, H. SHIOHARA, BRI (Presented by Hiroyuki

YAMANOUCHI)

1010-1030 U.S. Japan Cooperative Research in Precast Seismic Structural Systems (PRESSS)

Program -Status Report, Nigel Priestly, UCSD and H. S. LEW*, NIST

Identifies oral presenters.

xviii
1030-1050 Seismic Response Control of Highway Bridges by Variable Damper, Kazuhiko

KAWASHIMA*, S. UNION, H. SHIMIZU, PWRI

1050-1110 Discussion

THEME -EARTHQUAKE ENGINEERING - PART I

1110-1230 Technical Session— Earthquake Engineering — Part I Chairman: Richard WRIGHT

1110-1130 Seismic Design Considerations for Jordanelle Dam, John WILSON*, Perry HENSLEY,

William ENGEMOEN, and Francis MCLEAN, BUREC

1130-1150 Considerations on Evaluation of Seismic Force Acting on Embankment Dams and Its Simplified Prediction Procedure, Tadahiko FUJISAWA, N. YASUDA, N. MATSUMOTO, and

K. YAMABE, PWRI (Presented by Yoshikazu YAMAGUCHI)

1150-1210 Resolving Uncertainties in Seismic Hazard Models, Jeffrey K. KIMBALL*, DOE

1210-1230 Discussion

1230-1330 Lunch: Hosted by National Science Foundation, Dr. Dov JARON, Director, Biological and

Critical Systems Division

1330-1450 Technical Session — Earthquake Engineering — Part I Cont.

Chairman: Richard WRIGHT

1330-1350 Earthquake Alert System Feasibility Study, Philip HARBEN*, LLNL-DOE

1350-1410 Guideline for Seismic Design of Large Underground Structures (Draft), Hajime

ASAKURA*, Kazuhiko KAWASHIMA, and H. SUGITA, ACTC

1410-1430 ATC-25 Vulnerability of Lifelines, William BIVINS*, FEMA

1430-1450 Discussion

1450-1500 Break

TASK COMMITTEE MEETINGS

1500-1700 Task Committee Meetings

T/C B Large Scale Testing Programs (Lew/ Ohtani) BFRL Conference Room, Bldg. 226,

Rm. B221

T/C F Disaster Prevention Methods for Lifeline Systems (Dikkers/ Kawashima) Employees

Lounge

T/CG Passive, Active, and Hybrid Control Systems (Liu/ Kitagawa) Lecture Room B

T/CI Storm Surge and Tsunamis (Meyers/Yokoyama) Conference Room B113

Identifies oral presenters.

xix
T/C K Wind and Earthquake Engineering for Marine Facilities (Smith/ Noda) Conference

Room Bill

1700 Conclusion of day 2

Thursday 21 May

THEME-EARTHQUAKE ENGINEERING -PART II

0815-1110 Technical Session — Earthquake Engineering— Part II Chairman: Yukihiko SUMIYOSHI

0815-0835 Design Criteria for Base Isolation of Buildings, H. Tripp SHENTON*, NIST

0835-0855 Menshin Design Manual for Highway Bridges (Draft), Kazuhiko KAWASHIMA*,

K. HASEGAWA, and H. NAGASHIMA, PWRI

0855-0915 Friction Controllable Bearings for Sliding Base Isolation Systems, Maria Q. FENG, Prince- ton University and Masanobu SHINOZUKA*, NCEER

0915-0935 Discussion

0935-0950 Break

0950-1010 State-of-Art and Building Structure with Seismic Response Control System in Japan,

Yosikazu KITAGAWA* and M. MIDORIKAWA, BRI

1010-1030 Status of NSF Structural Control Research Programs, S. C. LIU*, Ken CHONG, and Henry

LAGORIO, NSF

1030-1050 Significant Results from Standardization Work on Rubber Bearings for Base-Isolated

Buildings, U. OHASHI and Hiroyuki YAMANOUCHI*, BRI

1050-1110 Effect of Temperature on Dynamic Behavior of a Base-Isolated Bridge, Osamu NAKANO*,

H. TANIGUCHI, and H. NISHI, HDB

1110-1130 Discussion

1130-1230 Lunch: Hosted by U.S. Army Corps of Engineers, Dr. Robert OSWALD, Director of

Research and Development

1230-1630 Technical Session — Earthquake Engineering— Part II Cont

Chairman: Yukihiko SUMIYOSHI

1230-1250 Dynamic Centrifugal Model Tests of Embankments on Liquefiable Ground, Yoshikazu

KOGA, Jun-ichi KOSEKI*, and A. TAKAHASHI, PWRI

1250-1310 Santa Cruz Landslide, Mary Ellen HAYNES*, WES

1310-1330 Study on Estimation of Lateral Ground Flow by Soil Liquefaction and Its Influence on Piles,

Ken-ichi TOKIDA, H. MATSUMOTO, I. TOWHATA, and Yasushi SASAKI, PWRI

(Presented by Jun-ichi KOSEKI)

* Identifies oral presenters.

XX
1330-1350 Discussion

1350-1410 Break

1410-1430 Horizontal Bearing Capacity of Pile Foundations Based on Horizontal Loading Tests, Michio

OKAHARA, S. NAKATANI, Y. KIMURA, and S. TAKAGI, PWRI (Presented by Minoru

FUJIWARA)

1430-1450 Seismic Requalification of Offshore Structures, Charles SMITH*, MMS

1450-1510 Analysis of Seismic Effects on Quay Walls, S. lAI, T. KAMEOKA, and Setsuo NODA*,

PHRI

1510-1530 Discussion " .

1530-1550 Implementation of Executive Order 12699, "Seismic Safety of Federal and Federally Assisted or Regulated New Building Construction," Richard WRIGHT*, NIST

1550-1610 CORPS Earthquake Engineering Research Program, William ROPER*, CORPS

1610-1630 Worldwide Seismic Risk Management Program - Progress Report, Ted ALGERMISSEN*,

U.S.GS

1630-1650 Discussion

1650 Conclusion of day 3

Friday 22 May

THEME-EARTHQUAKE ENGINEERING -PART HI

0830-1110 Technical Session — Earthquake Engineering— Part III Chairman: Richard N. WRIGHT

0830-0850 Seismic Performance of Reinforced Concrete Structures Utilizing High-Strength Materials,

Tastuo MUROTA, H. HIRAISHI, T. KAMINOSONO, H. SHIOHARA, and H. FUJITANI, BRI (Presented by Yosikazu KITAGAWA)

0850-0910 Dynamic Testing of Nonductile Concrete Frame Systems, P. A. BRADY and S. SWEENY*,

CERL

0910-0930 Shaking Table Collapse Tests of Reinforced Concrete Columns, N. OYAGI, Nobuyuki

OGAWA*, C. MINOWA, and T. MIKOSHIBA, NIED

0930-0950 Discussion

0950-1010 A Hollow Clay Tile Wall Seismic Performance Program Overview, J. E. BEAVERS*, W. D.

JONES, and W. C. T. STODDART, MMES-DOE

1010-1030 The Valuation and Designation Systems for Structural Calculation Program in Building

Engineering, Keiichi OHTANI*, NIED

* Identifies oral presenters.

xxi
1030-1050

1050-1110

1110-1125

Preliminary Report on Reconnaissance of Turkish Earthquake, Mehmet CELEBI*, U.S.GS

Discussions

Break

THEME-T/C WORKSHOP REPORTS

1125-1220 Technical Session — Summary T/C Workshop Reports

Chairman: Yukihiko SUMIYOSHI

1 125-1 135 T/C "E" Ground Morion and Seismic Design Forces , May 27-30, 1991 , Tsukuba (Presented by

T. ALGERMISSEN)

1135-1145 T/C "F" Disaster Prevention for Lifeline Systems, August 19-21, 1991, Los Angeles (Presented

by R. DIKKERS)

1145-1155 T/C "G" Passive, Active, and Hybrid Control Systems, September 4-5, 1991, Buffalo

(Presented by K. KAWASHIMA)

1155-1205 T/C "J" 8th Bridge Workshop, May 11-15, 1992, Chicago area (Presented by J. COOPER)

1205-1220 Discussion

1220-1320 Lunch: Hosted by U.S. Geological Survey, Dr. Robert WESSON, Chief, Office of Earthquakes, Volcanoes, and Engineering

TASK COMMITTEE REPORTS AND RESOLUTIONS

1320-1510 Task Committee Reports

Chairman: Richard N. WRIGHT

T/C A Strong Motion Instrumentation Arrays and Data

T/C B Large-Scale Testing Programs

T/C C Repair and Retrofit of Existing Structures

T/C D Evaluation of Structural Performance

T/C E Ground Motion and Seismic Design Forces Programs

T/C F Disaster Prevention Methods for Lifeline Systems

T/C G Passive, Active, and Hybrid Control Systems

T/C H Soil Behavior and Stability During Earthquakes

T/C I Storm Surge and Tsunamis

T/C J Wind and Earthquake Engineering for Transportation Systems

T/C K Wind and Earthquake Engineering for Offshore and Coastal Facilities

* Identifies oral presenters.

xxii
1510-1610 Adoption of Final Resolutions

1610-1620 Break

CLOSING CEREMONIES

1620 Call to Order by Noel J. RAUFASTE, Secretary-General, U.S.-Side Panel

Closing Remarks by Yukihiko SUMIYOSHI, Chairman JAPAN-Side Panel

Closing Remarks by Richard N. WRIGHT, Chairman U.S.-Side Panel

1645 Conclusion of Joint Meeting

xxiii
I

I
LIST OF PANEL MEMBERS

UNITED STATES-SIDE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1992

Dr. Richard N. Wright

Chairman

Director, Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301.975.5900 FTS 879.5900 FAX 301.975.4032

Mr. Noel J. Raufaste

Secretary-General

Head, Cooperative Research Programs

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301.975.5905 FTS 879.5904 FAX 301.975.4032

Dr. Kharaiti L. Abrol

Acting Director

Structural Engineering Service (088C1)

Department of Veterans Affairs

810 Vermont Avenue, NW

Washington, DC 20420

202.233.2864 FAX 202.233.2115

Dr. S. T. Algermissen

Research Geophysicist

Hazard and Risk Assessment Group

MS 966, POB 25046

Denver Federal Center

Geological Survey

U.S. Department of the Interior

Denver, CO 80225

303.273.8557 FTS 776.1611 FAX 303.236.0618

Mr. Charles E. Anderson

P.O. Box 25007 (D-3130)

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303.236.9114 FTS 776.9114 FAX 303.236.6763

XXV
Dr. Celso S. Barrientos

Supervisory Physical Scientist

NOAA/NESDIS-E/RA2 8

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

5200 Auth Road

Camp Springs, MD 20746

301.763.8102 FAX 301.763.8108

Dr. Eddie N. Bernard

Director, Pacific Marine Environmental Laboratory

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

7600 Sand Point Way, NE

BIN C15700/Building 3

Seattle, WA 98115-0070

206.526.6239 FTS 392.6800 FAX 206.526.6815

Dr. Roger D. Borcherdt

Branch of Engineering Seismology and Geology

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5619 FTS 459.5619 FAX 415.329.5163

Dr. A. Gerald Brady

Research Civil Engineer

Branch of Engineering Seismology and Geology

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5664 FTS 459.5664 FAX 415.329.5163

Mr. H. Lee Butler

Chief, Research Division

Coastal Engineering Research Center

U.S. Army Engineer Waterways Experiment Station

Office CEWES-CR

P.O. Box 631

Vicksburg, MS 39180-0631

601.634.2405 FTS 542.2405 FAX 601.634.4314

Dr. Ken P. Chong

Program Director, Structural Systems and Construction

Processes

Directorate of Engineering

National Science Foundation

1800 G Street, N.W., Room 1108

Washington, DC 20550

202.357.9542 FAX 202.357.0167

xxvi
Dr. Riley Chung

Group Leader, Earthquake Engineering

Structures Division

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301.975.6056 FTS 879.5863 FAX 301.975.4032

Mr. James D. Cooper

Deputy Chief, Structures Division, HNR-10

Federal Highway Administration

U.S. Department of Transportation

6300 Georgetown Pike

McLean, VA 22101

703.285.2447 FAX 703.285.2379

Dr. A. G. Franklin

Chief, Earthquake Engineering & Geophysics Division

CEWES-GH Geotechnical Laboratory

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-3909

601.634.2658 FTS 542.2658 FAX 601.634.4134

Mr. G. Robert Fuller

Acting Director, Manufactured Housing and Construction

Standards Division

Room 9152 HSMM

U.S. Department of Housing and Urban Development

Washington, DC 20410-8000

202.708.2210 FAX 202.708.0299

Mr. James H. Gates

Caltrans - Structures

P. 0. Box 942874

Sacramento, CA 94274

916.445.1439 FAX 916.323.2259

Mr. Peter E. Gurvin

Director, Building Design and Engineering Division

Foreign Building Operations

Building SA-6, Room 335

U.S. Department of State

Washington, DC 20520

703.875.6117 FAX 703.875.6204

xxvii
Dr. Walter W. Hays

Deputy for Research Applications

Office of Earthquake, Volcanoes and Engineering

905 National Center

Geological Survey

U.S. Department of the Interior

Reston, VA 22092

703.648.6711 FTS 959.6717 FAX 703.648.6717

Mr. James R. Hill

Manager, Natural Hazards Mitigation Programs

Office of Safety, Health and Quality Assurance

U.S. Department of Energy

Washington, DC 20545

301.903.4508 FAX 301.903.5285

Mr. Larry C. Hultengren

Senior Structural Engineer

Office of Civil/Structural Engineering

Foreign Building Operations

U.S. Department of State

Washington, DC 20520

703.875.6194 FAX 703.875.6204

Dr. Dov Jaron

Director

Division of Biological and Critical Systems

National Science Foundation

1800 G Street, NW

Washington, DC 20550

202.357.9500 FAX 202.357.7636

Dr. William B, Joyner

Geophysicist

Branch of Ground Motion & Faulting

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5640 FTS 459.5640 FAX 415.329.5163

Dr. Michael G. Katona

Chief Scientist

Air Force Civil Engineering Laboratory

Headquarters Air Force Civil Engineering

Support Agency

Tyndall AFB, FL 32403-6001

904.283.6274 FAX 904.283.6064

xxviii
Mr. Roger M. Kenneally

Structural Engineer

Structural and Seismic Engineering Branch

Seismic Section

Office of Nuclear Regulatory Research

U.S. Nuclear Regulatory Commission

Washington, DC 20555

301.492.3893 FAX 301.443.7836

Dr. Henry J. Lagorio

Program Director, Architectural and Mechanical Systems

Division of Biological and Critical Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9780 FAX 202.357.9803

Dr. H.S. Lew

Chief, Structures Division

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301.975.6061 FTS 879.6061 FAX 301.975.4032

Dr. Shih-Chi Liu

Program Director, Structural Systems

Division of Biological and Critical Systems

National Science Foundation

1800 G Street, N.W., Room 1132

Washington, DC 20550

202.357.9780 FAX 202.357.7636

Mr. Robert B. MacDonald

Manager, Geology Branch

Engineering and Research, Code D-3610 A

Bureau of Reclamation

U.S. Department of the Interior

Building 67, Denver Federal Center

Denver, CO 8 0225

303.236.6904 FTS 776.8643 FAX 303.236.6763

Dr. Wayne N. Marchant

Chief, Research and Laboratory Services Division

P. O. Box 25007, Denver Federal Center

Code D-3700

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303.236.5983 FTS 776.5983 FAX 303.236.7664

xxix
Dr. Francis G. McLean

Chief, Geotechnical Engineering and Embankment

Dams Branch

P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303.236.3854 FTS 776.3854 FAX 303.236.6763

Mr. Herbert Meyers

Chief, Earth Geophysics

National Geophysical Data Center

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

325 Broadway

Boulder, CO 80303-3328

303.497.6521 FTS 320.6521 FAX 303.497.6513

Mr. Ugo Morelli

Policy Manager for Seismology

Federal Emergency Management Agency

500 C Street, SW

Washington, DC 20472

202.646.2810 FAX 202.646.3041

Mr. Howard D. Nickerson

Earthguake Engineering and Weapons Specialist

Naval Facilities Engineering Command

Hoffman Building #2, Room 12S63, Code 04B2

200 Stovall Street

Alexandria, VA 22332

703.325.0044 FAX 703.325.1916

Dr. William E. Roper

Assistant Director

Research and Development (Civil Works)

U.S. Army Corps of Engineers

20 Massachusetts Avenue, N.W

Washington, DC 20314-1000

202.272.0257 FAX 202.272.0907

Dr. Erwin L. Schaffer

Assistant Director, Wood Products Research

Forest Products Laboratory

Forest Service

U.S. Department of Agriculture

One Gifford Pinchot Drive

Madison, WI 53705-2398

608.231.9272 FTS 364.5672 FAX 608.231.9592

XXX
Mr. Charles E. Smith

Research Program Manager

Offshore Minerals Management

Technology Assessment and Research Branch

Minerals Management Service

U.S. Department of the Interior

381 Elden Street, MS 4800

Herndon, VA 22070-4817

703.787.1559 FAX 703.787.1010

Dr. T. T. Soong

Samuel P. Capen Professor

Department of Civil Engineering

National Center for Earthquake Engineering Research

State University of New York at Buffalo

212 Ketter Hall

Buffalo, NY 14260

716.636.2469 FAX 716.636.3733

xxxi
ALTERNATE MEMBERS

Dr. Clifford J. Astill

Program Director

Division of Biological and Critical Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9500 FAX 202.357.9803

Dr. Mehmet K. Celebi

Research Civil Engineer

Branch of Engineering Seismology and Geology

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5623 FTS 459.5623 FAX 415.329.5163

Mr. Michael Changery

Chief, Global Analysis Branch

National Climatic Data Center

National Oceanic and Atmospheric Administration

U. S. Department of Commerce

Federal Building

Ashville, NC 28801

704.259.0765 FTS 672.0765 FAX 704.259.0246

Mr. C.Y. Chen

Senior Civil/Geotechnical Engineer

Foreign Building Operations

A/FBO/PE/CSB, SA-6, Rm . 327

Washington, DC 20520

703.875.6207 FAX 703.875.6204

Mr. Vincent P. Chiarito

Research Structural Engineer

Structural Mechanics Division

Structures Laboratory

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601.634.2714 FAX 601.634.3412

Dr. James F. Costello

Senior Structural Engineer

Structural and Seismic Engineering Branch

Structural Section

Office of Nuclear Regulatory Research

U.S. Nuclear Regulatory Commission

Mail Stop 007NL

Washington, DC 20555

301.492.3818 FAX 301.492.3696

xxxii
Mr. Lucian G. Guthrie

Structural Engineer

Office of Chief of Engineers

HQUSACE (CEEC-ED)

U.S. Department of the Army

20 Massachusetts Avenue, N.W.

Washington, DC 20314-1000

202.272.8673 FAX 202.272.1485

Dr. James R. Houston

Chief, Coastal Engineering Research Center

US Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-6199

601.634.2000 FAX 601.634.2655

Mr. James Lander

Geophysicist

Cooperative Institute for Research in

Environmental Sciences

University of Colorado

Campus Box 449, Room 152 RL3

3100 Marine Street

Boulder, CO 80309

303.497.6446 FTS 320.6446 FAX 303-497.6513

Mr. Robert R. Ledzian

Senior Staff Assistant for Research

Office of Liaison - Engineering Research, Code 3020

Bureau of Reclamation

U.S. Department of the Interior

1849 C Street, N.W.

Washington, DC 20240

202.208.3432 FAX 202.208.6252

Mr. Tingley K. Lew

Research Structural Engineer

Structures Division, Code L51, Bldg. 560, Rm. 278B

Naval Civil Engineering Laboratory

Port Hueneme, CA 9 3 043

805.982.1234 FAX 805.982.1418

Mr. Michael Mahoney

Physical Scientist

Federal Emergency Management Administration

500 C Street, SW

Washington, DC 20472

202-646-2810 FAX 202-646-3041

xxxiii
Dr. J. Eleonora Sabadell

Director, Natural and Manmade Hazards Mitigation Program

Directorate of Engineering

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9780 FAX 202.357.9803

Dr. John B. Scalzi

Program Director, Structures and Building Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9542 FAX 202.357.7636

xxxiv
JAPAN-SIDE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

May 1992

Mr. Yukihiko Sumiyoshi

Chairman

Director-General

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Takashi lijima

Secretary-General

Assistant Director General

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tadahiko Fujisawa

Head, Fill Type Dam Division

Dam Department

Public Works Research Institute

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tokunosuke Fujitani

Chief, The Second Research Laboratory

Applied Meteorology Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1 Nagamine, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-7111

Mr. Minoru Fu j iwara

Director, Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

XXXV
Mr. Tadao Hoya

Director, Geographic Department

Geographical Survey Institute

Ministry of Construction

1, Kitazato, Tsukuba-shi

Ibaraki-ken 305

Dr. Kazuhiko Kawashima

Secretary, Japan-Side Panel

Head, Earthquake Engineering Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Yoshikazu Kitagawa

Director, International Institute of

Seismology and Earthquake Engineering

Building Research Institute

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Yasuyuki Koga

Head, Soil Dynamics Division

Construction Method and Equipment Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Hatsukazu Mizuno

Head, Large Scale Structure

Testing Division

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi

Ibaraki-ken 305

xxxvi
Mr. Tastuo Murota

Director, Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Shinsuke Nakata

Associate Director for Composite

Structures Research

Structural Engineering Department . . Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Mr. Kazuhiro Nishikawa

Head, Bridge Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Setuo Noda

Director, Structural Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka,

Kanagawa-ken 239

Tel. 0468-44-5029

Dr. Nobuyuki Ogawa

Head, Earthquake Engineering Laboratory

National Research Institute for Earth

Science and Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tsukuba-shi

Ibaraki-ken 305

xxxvii
Mr. Keiichi Ohtani

Director, Disaster Prevention Research Division

National Research Institute for Earth

Science and Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-1611

Dr. Michio Okahara

Head, Foundation Engineering Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Shin Okamoto

Director General

Building Research Institute

Ministry of Construction

1, Tatehata, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Mr. Shin Otsuka

Head, Typhoon Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-51-7111

Dr. Yasushi Sasaki

Director, Earthquake Disaster

Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

xxxviii
Dr. Masaaki Seino

Head, Seismology and Volcanology

Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-51-7111

Dr. Tomotsuka Takayama

Director, Hydraulic Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka,

Kanagawa-ken 239

Tel. 0468-41-5410

Mr. Ken-ichi Tokida

Head, Ground Vibration Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Takaaki Uda

Head, Coastal Engineering Division

River Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Tatsuo Uwabe

Chief, Disaster Prevention Laboratory

Structural Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi

Kanagawa-ken 239

Tel. 0468-44-5030

xxxix
Dr. Hiroyuki Yamanouchi

Special Researcher for International

Standard, Research Planning

and Information Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Yutaka Yamazaki

Director, Production Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Koichi Yokoyama

Secretary, Japan-Side

Head, Structure Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

XL
July 1992

LIST OF TASK COMMITTEE MEMBERS

1992

Task Committee US Side Japanese Side

A. Strong-Motion Instrumentation

Arrays and Data

B. Large-scale Testing Programs

C. Repair and Retrofit of

Existing Structures

D. Evaluation of Structural

Performance

A. G. Brady* S. Noda*

C. J. Astill Y. Kitagawa

R. D. Borcherdt K. Ohtani

M. K. Celebi M. Seino

ai\ . Lj . rranKiin K

i OK laa

W. B. Joyner H. Yamanouchi

R. B. MacDonald

F. G. McLean

H. Meyers

J. R. Hill

H. S. Lew* K. Ohtani*

V. P. Chiarito Y. Koga

J. E. Sabadell H. Mizuno

C. E. Smith S. Noda

J. Beavers Y. Yamazaki

S. Sweeney K. Yokoyama

K. P. Chong* S. Nakata*

F . A. Brady M . Fu^ iwara

G. E. Freeland S. Okamoto

J. 0. Jirsa Y. Sasaki

H. J. Lagorio

H. S . Lew

T. K. Lew

H. D. Nickerson

D. H. Oh

M. A. Phipps

J. B. Scalzi

D. Jones

D. Eagling

S. Woodson

J. B. Scalzi* Y. Yamazaki*

S. A. Asar T. Fuj itani

M. K. Celebi M. Fuj iwara

G. R. Fuller T. Murota

E. L. Schaffer M. Okahara

R. L. Hall S. Ohtsuka

J. R. Hill H. Yamanouchi

D. Eagling K. Yokoyama

J. Hunt

D. Ellsworth

XLI
Task Committee US Side Japanese Side

E. Ground Motion and Seismic S.T. Algermissen* K. Kawashima*

Design Forces Programs C. S. Barrientos T

.

Hoya

R.D. Borcherdt Y. Kitagawa

A.G. Brady Y. Sasaki

G.R. Fuller T. Uwabe

J.R. Hill H

.

Mizuno

J. Hunt K. Ohtani

W.B. Joyner

J. Kimball

T.K. Lew

F. Disaster Prevention Methods R.M. Chung* Y

.

Sasaki*

for Lifeline Systems C.J. Astill K. Kawashima

M.K. Celebi T. Murota

J.D. Cooper N. Ogawa

G. Al-Chaar K

.

Tokida

T.K. Lew T. Uwabe

J.B. Scalzi

J. S. Spencer

S. Wu

G. Passive, Active, and Hybrid S.C. Liu* Y. Kitagawa*

Control Systems V.P. Chiarito K. Kawashima

K.P. Chong N. Ogawa

J.R. Hayes H. Yamanouchi

L.C. Hultengren K

.

Yokovama

H.J. Lagorio

T.T. Soong

H. Soil Behavior and Stability A.G. Franklin* K. Tokida*

During Earthquakes C.J. Astill T. Fu j isawa

R.D. Borcherdt Y. Koga

C. Y. Chen S. Noda

F.G. McLean M. Okahara

C. E. Smith Y. Sasaki

H. Mizuno

I

.

Storm Surge and Tsunamis H. Meyers* T. Uda*

C.J. Astill S. Ohtsuka

C. S. Barrientos T. Takayama

E.N. Bernard M. Seino

L. Butler

C.F. Jelesnianski

J. Lander

W.E. Roper

XLII
Task Committee

J. Wind and Earthquake

Engineering for

Transportation Systems

K. Wind and Earthquake

Engineering for Offshore

and Coastal Facilities

*Chairman

US Side Japanese Side

J. D. Cooper* M. Fu j iwara*

A. G. Franklin T. Fuj itani

J. H. Gates K. Kazuhiko

J. B. Scalz i K. Nishikawa

H. S. Lew M. Okahara

Y. Sasaki

K. Tokida

K. Yokovama

C. E. Smith* T. Urabe*

C. S. Barrientos T. Fuj itani

w. E. Roper T. Murota

J. E. Sabadell S. Noda

M. C. Miller M. Okahara

M. K. Celebi

XLIII
RESOLUTIONS

RESOLUTIONS OF THE TWENTY-FOURTH JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS (UJNR)

National Institute of Standards and Technology

Gaithersburg, MD 20899

May 19-22, 1992

The following resolutions are hereby adopted:

L The Twenty-Fourth Joint Panel Meeting provided an opportunity to exchange valuable technical

information which was beneficial to both countries. In view of the importance of cooperative

programs on the subject of wind and seismic effects, the continuation of Joint Panel Meetings is considered essential.

2. The following activities have been conducted since the Twenty-Third Joint Meeting:

a. Researchers were exchanged from both countries to promote and conduct research that advanced the state of earthquake and wind engineering.

b. Technical documents, research reports, and proceedings of workshops were exchanged.

c. Both sides published a List of Panel Publications 1969-1991 for distribution to U.S. and

to Japan domestic organizations and to international organizations, especially those in the less developed countries.

d. Five significant conferences were held:

a) The First Workshop on Hazard/Risk Assessment and Design Earthquake

Loading, Task Committee (E), Tsukuba, May 28-29, 1991.

b) The Fourth Workshop on Earthquake Disaster Prevention for Lifeline Systems,

Task Committee (F), at Los Angeles, CA, August 19-21, 1991.

c) Workshop on Earthquake Protective Systems of Bridges, Task Committee (G),

at Buffalo, NY, September 4-5, 1991.

d) The Second U.S. -Japan Joint Technical Coordinating Committee on PRESSS,

Task Committee (B), at Tsukuba, Japan, October 1991.

e) The Eighth Bridge Workshop, Task Committee (J), at Chicago, IL, May 11-15,

1992.

3. A one day International Symposium on the International Decade for National Disaster Reduction was held in Gifu City on May 20, 1991, to increase the awareness of the people of Japan about

the disastrous effects from strong winds and earthquakes. This Panel provided three of the five presentations. More than 600 persons attended.

XLIV
4. The Panel will continue to seek methods to contribute to the International Decade of Natural

Disaster Reduction (IDNDR) such as exchanging Proceedings of joint Panel Meetings and Task

Committee Workshops with their respective Country's National Committees of IDNDR.

5. The Panel recognizes the importance of the work by both sides in the U.S. -Japan Joint Research

Program on Precast Seismic Structural Systems (PRESSS).

6. The Panel accepts each Task Committee's report developed during the Twenty-Fourth Joint

Meeting. Each report presents objectives, scope of work, accomplishments, future plans,

information exchange, impacts, and barriers.

7. The Panel endorses the following proposed Task Committee Workshops during the coming year.

a. Task Committee (B) plans a workshop on Hybrid Structural Systems, September 1992,

San Francisco, CA.

b. Task Committee (E) plans a 2nd workshop on Ground Motion and Seismic Design

Forces, May 1993, Tsukuba.

c. Task Committee (F) plans a 5th workshop on Earthquake Disaster Prevention for Lifeline

Systems, October 26-27, 1992, Tsukuba.

d. Task Committee (G) plans a 2nd workshop on Earthquake Protective Systems of Bridges,

December 1992, Tsukuba.

e. Task Committee (G) plans a workshop on Smart and High Performance Material and

Systems, May 1993, Japan.

f. Task Committee (J) plans a 9th Bridge Workshop before the 25th Joint Meeting, Japan.

g. Task Committee (K) plans a Workshop on Wind and Earthquake Engineering for Offshore and Coastal Facilities, May 1993, Yokosuka.

In addition, three Task Committees have scheduled workshops after the 25th Joint Panel Meeting:

a) Task Committee (A) has scheduled a workshop on Digital Recording Technologies, Fall 1993, Hawaii.

b) Task Committee (H) has scheduled a workshop on Remedial Treatment of Potentially

Liquefiable Soils, October 1993, Japan.

c) Task Committee (I) has scheduled the 3rd Workshop on Storm Surge and Tsunamis,

August 19-20, 1993, Japan.

Scheduling for the workshops shall be performed by the U.S. and Japan Chairmen of the

respective Task Committee with concurrence of the Joint Panel Chairmen. Results of each

activity conducted before the 25th Joint Meeting shall be presented at the 25th Joint Meeting.

8. In the spirit of celebrating the 25th Anniversary of the Panel on Wind and Seismic Effects, a special event will be conducted during the 25th Joint Panel Meeting to mark this occasion. Both

XLV
sides' Secretaries will identify an appropriate method that celebrates the Panel's activities and

facilitates the transfer of the Panel's technologies into practice, e.g., special symposium,

publication, joint activity. The Secretarys' suggestions will be discussed at the respective sides' early fall 1992 domestic meetings to begin scheduling and arranging this event during mid-fall

1992.

9. The Panel recognizes the importance of continued exchange of personnel, technical information,

research results, and recorded data that lead to mitigating losses from strong winds and

earthquakes. The Panel also recognizes the importance of using available large-scale testing

facilities in both countries. Thus, these activities should continue to be strengthened and

expanded and, as appropriate, share Task Committee activities at other meetings that have

technical interests in the Task Committee activities. To facilitate these exchanges, the Panel will

provide official endorsement.

10. The Twenty-Fifth Joint Meeting of the UJNR Panel on Wind and Seismic Effects will be held

at PWRI, Tsukuba, Japan, May 1993. Specific dates, program, and itinerary will be proposed

by the Japan-Side Panel with concurrence of the U.S. -Side Panel.

XLVI
IDNDR Presentation

U.S. Federal Agency Role in the International Decade for Natural Disaster

Reduction and UJST Workshops for Natural Disaster Reduction

by

Paul Kilho Park*

ABSTRACT

Under auspices of the UJST (US-Japan Science and

Technology) Agreement of 1988, annual UJST

workshops for natural disaster reduction have been

going on since March 1990 as part of the

International Decade for Natural Disaster Reduction

(IDNDR) activities of the U.S. and Japan. Tv^^o workshops held have resulted in formulation of

pilot projects and identification of future thrusts,

such as joint database establishment and

improvement of hazard mapping via Geographic

Information System (GIS). The third workshop is planned at Mt. Hood, 7-11 September 1992. The

first UJST panel was formed in 1991; it is the

UJST Panel on Volcanic Disaster Prevention.

KEYWORDS: earthquake; GIS (Geographic

Information System); hazard mapping; human

health; IDNDR (International Decade for Natural

Disaster Reduction); landslides; severe storms; socioeconomics; UJST (US-Japan Science and

Technology Agreement); volcano.

1. INTRODUCTION

Broad interdisciplinary cooperative agreement

between the two heads of the state in 1988

initiated the U.S.-Japan Science and Technology

Agreement (UJST). The seven main areas of

cooperation agreed on are as follows: life science,

including biotechnology; information science and

technology; manufacturing technology; automation

and process control; global geoscience and

environment; joint database development; and

advanced materials, including superconductors.

Almost at the same time the 42nd UN General

Assembly in December 1987 unanimously adopted

a resolution designating the 1990s as the

"International Decade for Natural Disaster

Reduction (IDNDR)." The 44th UN General

Assembly in December 1989 unanimously adopted

a resolution establishing the international

framework for action by the UN and proclaimed

the IDNDR beginning on 1 January 1990 ending

31 December 1999.

Within the United States, in 1988, the Federal

Government's IDNDR effort was vested on the

Federal Coordinating Council for Science,

Engineering and Technology (FCCSET), headed by

the President's Science Advisor. Within the

FCCSET, the Committee on Earth and Environment

Sciences (CEES) established the Subcommittee for Natural Disaster Reduction (SNDR) to directly deal

with and contribute to the IDNDR activities (Bergner, 1991). Concurrently, non-governmental

National Committee was established, and Professor

Walter R. Lynn, Dean of the University Faculty at Cornell University, is its current chairman. The

FCCSET, CEES, and SNDR are headed by Drs. D. Allan Bromley, Frederick Bemthal and William H. Hooke, respectively.

In Japan, Japanese Government Headquarters for the IDNDR, headed by the Prime Minister with the

Disaster Prevention Bureau of the National Land

Agency, Prime Minister's Office, acting as

secretariat, was established by a decision of the

Cabinet in May 1989. In August 1990, Japanese

National Committee for the IDNDR was organized

by the private sector, which includes the academia,

industry, the Japanese Red Cross Society. Professor

Jiro Kondo is the Chairman of Japan National

Committee for the IDNDR. Both the U.S. and

Japan have had similar organizational evolution

heretofore.

Vigorous national and international activities to enhance the IDNDR effort have commenced,

* NOAA Office of the Chief Scientist

1825 Connecticut Avenue, NW (Room 625)

Washington, DC 20235-0001

3
especially in Japan and the U.S. The IDNDR

International Conference 1990 Japan was held in Yokohama and Kagoshima, 27 September-3 October

1990. About 1300 participants from 43 countries

and 16 international organizations attended the

1990 meeting. Similar turnout was repeated at the

IDNDR Summit Conference on Earthquake and

Natural Disaster Countermeasures 1991 Japan,

Tokyo, 8-11 October 1991.

On 9 January 1992, Prime Minister Miyazawa and

President Bush jointly announced the Global

Partnership Plan of Action in Tokyo, in which the

two heads of the state agreed to "cooperate to reduce natural hazards, in accordance with the

International Decade for Natural Disaster Reduction

and the U.S-Japan Science and Technology

Agreement, by assisting developing counties to prepare disaster reduction plans, including hazard

mapping, risk assessment and the establishment of

joint databases;..." Much staff work is going on to follow up this accord, especially to get ready for Prime Minister Miyazawa's planned visit to the U.S. of summer, 1992.

Between the U.S. and Japan two UJST workshops

for natural disaster reduction have been held. The

first workshop was at Hilo and Punalu'u, Hawaii,

26 - 30 March 1990, and the second at Karuizawa,

Japan, 23 - 27 September 1991; the third

workshop is planned to be held at Mt. Hood

Timberline Lodge, 7-11 September 1992. This

report describes the details of the Karuizawa

workshop. Both the first and second workshops

have been chaired by Drs. Yukio Hagiwara,

Director-General, National Research Institute for Earth Science and Disaster Prevention (NIED),

Science and Technology Agency (STA), and William

H. Hooke, Chairman, Subcommittee for Natural

Disaster Reduction (SNDR). In Japan, Science and

Technology Agency belongs to the Prime Minister's

Office, as so does the National Land Agency.

Both Japan and the U.S. dispatch disaster relief teams. When a 635-year-dormant volcano,

Pinatubo, exploded in 1991, both Japan and the

U.S. dispatched their teams to help the Philippines

cope with the devastation. Their onsight activities were highlighted at the Karuizawa workshop to understand the casual relationship on the disaster

impact upon humanity. The U.S. effort was greatly

enhanced by the Office of Foreign Disaster

Assistance (OFDA) in the Department of State, and

the Japanese effort by Japan International

Cooperation Agency (JICA).

The JICA's fiscal year (FY) 1990 budget was 133

billion yen (about one billion dollars), some of

which was earmarked for emergency disaster relief. The JICA is in charge of dispatching Japan Disaster

Relief Team (JDR) and providing equipment and

materials needed for disaster relief and restoration

activities. The JDR carries out search and rescue, emergency medical care (as of August 1989, a total of 350 doctors and nurses are registered), and

emergency measurers (prevention of secondary

disasters) and restoration. Stock-piled relief goods

are located at Narita (Japan), Singapore, Italy and

Mexico. Japanese medical and rescue teams have

been dispatched to Iran (earthquake) on 21 June

1990 and to the Philippines (earthquake) on 16

July 1990. The JICA also sent an engineering team

for reconstruction to the Philippines (earthquake) on 16 July 1990 and to Saudi Arabia (petroleum

over Gulf) on 3 March 1991.

Although the UJST workshop for natural disaster

reduction continues to focus mainly on the cutting

edge of understanding of the physical processes

leading to natural hazards, resulting in improved

prediction, especially during the IDNDR period, 1 January 1990 - 31 December 1999, we are to cooperate with all the concurrent IDNDR activities so that we contribute not only at the scientific frontier but also to facilitate other bilateral and

multilateral activities. A prominent example is to cooperate with various U.S. -Japan Natural

Resources (UJNR) activities especially with the

UJNR Panel on Wind and Seismic Effects. Another

example is that Japan-U.S. Symposium on Snow

Avalanche, Landslide, Debris Flow Prediction and

Control (JUSSLDPC) was held at Tsukuba, Japan,

30 September - 5 October 1991. This effort has

been ongoing since 1980. The third UJST

Workshop for Natural Disaster Reduction, Mt. Hood, Oregon, 7-11 September 1992, will enhance

the JUSSLDPC effort by working together with

them. Similarly the non-governmental U.S. National Committee for IDNDR is planning to have

a U.S.-Japan earthquake symposium to formulate

pilot projects between the two countries in the near future, to be carried out by non-governmental

4
resources. It is a commendable effort, and when

there occurs much mutually enhancing confluence

between the non-govemmental organization (NGO)

effort and govemment-to-govemment bilateral pilot

project formulation (such as via UJNR and UJST),

the true spirit of the IDNDR cooperation blossoms.

I especially wish to single out that there are very

active ongoing bilateral disaster-mitigation activities under the umbrella of U.S. - Japan Natural

Resources (UJNR) accord for over 20 years. The

original agreement was signed off by Japanese

Foreign Minister and U.S. Secretary of State. Two

significant panels within UJNR for disaster

mitigation are that of earthquake and severe storm

mitigation technology as well as forest fire. Furthermore, at working level the two countries'

Geological Surveys continue to have a functioning

sister relationship; there also is much cooperation

between Japan Land Agency, Ministry of

Construction and the U.S. Federal Emergency

Management Agency. All of these ongoing

activities will be enhanced by the new thrust the

UJST pursues in consort with IDNDR in the 1990s.

2. THE FIRST UJST WORKSHOP, 26-30 MARCH

1990

Under the auspices of the Earth Science and

Environmental Task Group of the UJST Agreement,

the first workshop had a running start from initial planning in January 1990 to its execution in March

1990. The workshop was effective and successful

due to the high priority attached to the work by

both sides and the desire to make a substantive

start. Unsung heros include Chris Newhall, Will

Prescott, and Dick Janda of USGS and Shoichiro

Katayama, Yukio Hagiwara and Masayuki

Watanabe of Japan. Sixteen Japanese and

Nineteen Americans participated in the workshop.

This workshop provided an initial opportunity to define cooperative activities in disaster reduction

with emphasis on knowledge, and described

specific U.S.-Japan projects that complement the

ongoing activities. Three broad goals thus

generated are as follows:

1. Better understanding of the physical processes leading to natural hazards, resulting in improved

prediction.

2. Translation of scientific advance into improvement in the human condition, through better

application of scientific knowledge to disaster

reduction.

3. Better scientific understanding of the effect of

global change on the frequency and severity of

natural hazards.

The workshop developed specific recommendations

for cooperative projects in the following four

specialist groups:

Earthquake prediction

Volcanic eruption prediction

Typhoon prediction

Landslide prediction

The final report of the workshop has been

reprinted by the Science and Technology Agency of

Japan; I shall be delighted to supply you it upon

request.

3. THE SECOND UJST WORKSHOP

23-27 SEPTEMBER 1991

Formulation of cooperative pilot projects attainable

by 31 December 1999, the end of the IDNDR, was

one of the major objectives for the second

workshop. Also an added-on objective was the

establishment of joint database for disaster

reduction in cooperation with the Joint Database

Development Task Group of the UJST Agreement.

Thirty-six Japanese and seventeen U.S. scientists participated in the workshop at Karuizawa.

To flesh out the second goal of the general

recommendations generated by the first workshop,

which is "translation of scientific advance into

improvement in the human condition, through

better application of scientific knowledge to disaster reduction", the second workshop added

two new groups of socioeconomics and human

health, in addition to earthquakes, volcanoes,

typhoons and landslide groups. Keynote addresses were given in socioeconomics and human health as well as on climatic change and its impact on

natural disasters that requires innovative extensions

to current natural disaster research. Furthermore,

the typhoon group was renamed "severe weather

5
phenomena" group to include mecoscale storms as well as locally heavy rainfall. Four comprehensive

recommendations were generated by the workshop.

They are as follows:

1. Better understanding of the physical processes

leading to natural hazards, resulting in improved

prediction. Better understanding of the

opportunities for reducing natural disasters

through socioeconomic and medical sciences.

2. Translation of advances in science and

technology into improvement in the human

condition through exchange of information and

better application of knowledge to disaster

reduction.

3. Better scientific understanding of the effect of

global change on the frequency and intensity

of natural hazards.

4. To enhance progress toward [l]-[3] above,

endeavor to develop joint databases and

improve capabilities for information exchange.

During the intercessional period between April

1990 and August 1991, several prominent natural

disasters have occurred (earthquakes in Iran and

the Philippine in 1990; volcanic eruptions at Mt. Unzen in Japan and Mt. Pinatubo in the

Philippines in 1991, and a cyclone disaster in Bangladesh in 1991). Both the U.S. and Japan

dispatched medical, rescue, and engineering teams

for reconstruction for the Iranian and Philippines

disasters. Mt. Pinatubo volcano eruption was

scientifically studied by both countries; thus

resulting in having a special session to compare

the scientific understanding of the volcanic

eruptions at Mt. Unzen and Mt. Pinatubo at the

second workshop. Though painful each natural

disaster is a gigantic natural laboratory for us to understand the physical processes leading to the

disaster and during the disaster, and post-disaster processes.

Also during the intercessional period, the UJST

Agreement has established "volcanic panel" to specialize in volcanic hazard mitigation. The panel

will start its ovm independent deliberation in near

future; at the same time it will continue to contribute to the forum such as the joint database

establishment, hazard mapping through

geographical information system (GIS), and

advancing its close tie with landslide and

earthquake prediction groups. Furthermore, snow

avalanche and landslide group too plans to establish its ovm UJST panel to sharpen its focus on its respective specialty; it will be a cumulative

product of the group's U.S.-Japan cooperation that

began in 1980.

To facilitate the second workshop we issued a 411-

page preprint collection at the workshop, so that

the workshop does not become mere a scientific symposium where the bulk of time was spent on

oral presentations. We plan to follow similar

format to facilitate the third workshop, Mt. Hood,

Oregon, 7-11 September 1992. The workshop

preprint collection has been reprinted by the

Science and Technology Agency of Japan (Japan

international Science and Technology Exchange

Center, 1992).

Each individual group's reports are summarized

and given in Section 5. Pilot projects

recommended are listed and explained in the

workshop proceedings. The workshop recommends

that UJST secure necessary funds to carry out the

pilot projects, so that our effort does not become

a mere paper exercise. For example, two projects

have given the magnitude of the fund needed to carry out the pilot projects; they are a cultural and

cross-hazard study for earthquake and typhoon

(hurricane) in Japan and the U.S., and the

Tsunami Inundation Modeling Exchange (TIME)

project at Tohoku University. As Goethe said: "Until one is committed, there is hesitancy, the

chance to draw back, always ineffectiveness, ... boldness has genius, power and magic in it." Political commitment has been made by the two

heads of the state of Japan and the U.S. How we

would implement the pilot projects the workshop

generated by our governments can learn a lesson

from the recent successful 7-year Kuroshio Projects

Japan and China have executed under the

leadership of Science and Technology Agency with

timely funding.

4. THE THIRD UJST WORKSHOP

7-11 SEPTEMBER 1992

Mt. Hood Timberline Lodge in Oregon is the venue

of the third workshop for natural disaster

reduction; it provides a natural laboratory to study

volcanic activities, snow avalanche and landslide

countermeasurers. Much intercessional,

preparatory activities are going on to carry out the

6
workshop effectively. A core of the executive

committee has been established to prepare for the

workshop. Drs. Richard Janda of USGS Cascades

Volcano Laboratory (established in 1980 following

the eruption of Mt. St. Helens) and Fred Swanson

of the U.S. Forest Service in Oregon are the local co-chairmen. Professor Dennis Mileti of Colorado

State University and I comprise the rest of the

executive committee.

The second workshop has recommended sharing of

hazard mapping techniques used and planned to be

implemented as a common third workshop theme

among individual groups. Geographical

Information System (CIS) applied within disaster

reduction and elsewhere will be scrutinized; a plan

is made to invite an expert to give a keynote

speech on this regard. Interestingly, the IDNDR

International Conference 1990 Japan, 27 September

- 30 October 1990, too pointed out that some

major work must be done "to prepare hazard maps

and formulate land use plan which incorporates

the viev^oint of disaster prevention." On the

hazard mapping an attention is being made to

invite practioners to the UJST working groups.

An additional common theme to be surfaced at the

third workshop is the universality of hydrologic

factors that induce natural disasters. The Pinatubo

volcano calamity was magnified by an onslaught of

a typhoon over the volcano, which produced much

volcanic ash-mud flow upon the anthrosposhere

The typhoon energy was converted into damaging

mud flow of phenomenal size. Snow avalanche

and ice disasters also share common hydrologic

factors. Receding glaciers due to global warming

too can be considered an aspect of future disaster

mitigation concerns. A field trip is planned to study ecological means to combat landslides at the Andrews Experimental

Forest in Oregon. It relies on soft engineering means in contract to commonly seen hard

engineering practice where dams and concrete

walls are built to protect human habitat.

Ecological recovery at the site of Mt. St. Helens

volcanic eruption too will be visited.

Since each UJST workshop for disaster reduction

is a precious opportunity to advance our knowledge to mitigate natural disasters, pertinent

private sector participation, especially from the

academic and the user communities, is warranted

so that both Japan and the U.S. go forward to make significant progress on preparedness issues.

5. PRODUCTS OF THE SECOND WORKSHOP

For the 2nd workshop two new working groups were added based on the 1st workshop

recommendation (socioeconomics and human

health groups). Thus the workshop spent the bulk

of its time discussing progress toward the

achievement of the workshop goals by the

following six working groups:

Volcanic eruptions

Earthquakes

Severe weather phenomena

Landslides

Socioeconomics

Human health

In addition the workshop was charges by the UJST

liaison and working levels meetings, Tokyo, 16-19

July 1991, to work on the joint database

establishment between Japan and the U.S.

Three keynote addresses were given at the onset

of the workshop to broaden and integrate the

group deliberations. The topics covered are the

casual relationship between the global change and

natural hazards, socioeconomic issues in natural

disaster reduction, and the role of epidemiology in natural disaster reduction.

Data management initiatives were developed by

each of the working groups. In particular,

participants identified the need for improved CIS

capabilities.

5.1 Volcano group

The volcano group proposes 10 projects that are

current, practical to start within the next few

years. They are (1) volcanic hazard and risk mapping; (2) low-frequency volcanic earthquakes

and volcanic tremor; (3) processes leading to volcanic eruptions; (4) quantitative characterization

of eruptions in progress; (5) dynamics of dome

building and related pyroclastic flows; (6) data

acquisition system for volcano-genetic vibration; (7)

observation and monitoring of submarine

volcanoes; (8) volcanically induced melting of snow

7
and ice; (9) joint observation of selected Japanese

and U.S. volcanoes; and (10) coordinated

assistance to developing countries.

The volcano group is to become the first UJST

scientific panel entitled Volcanic Disaster

Prevention Panel; the new panel's charge includes

dissemination and application of volcanic hazard

information in addition to basic investigations of

processes, both volcanic and volcano-hydrologic,

contributing to volcanic disasters. The volcano

group featured a special session to discuss the

findings of the recent Mt. Unzen and Pinatubo

eruptions. At the 3rd workshop, the volcano group

will devote mainly on the substantive reports of

new methods to elucidate the processes toward

disaster prevention and hazard mapping by GIS.

5.2 Earthquake group

The earthquakes group covers both earthquakes

and tsunamis; the group also gives

recommendations in mitigation policy, for the

advancement of scientific frontier must put into

good use. On earthquakes, the detection of crustal movements is now greatly enhanced by space-based

geodetic systems, such as Global Positioning

System (GPS), Very Long Baseline Interferometry

(VLB!) and Satellite Laser Ranging (SLR). Also

continuous monitoring for deformation is essential

to detect short-term precursors of earthquakes and

volcanic eruptions, which must be coupled with

evaluations of anomalies in continuous

observations. Deep drilling and borehole

measurements, both geophysical and geochemical,

in seismically active areas is warranted to supplant

the ongoing surface observations.

Tsunami is a Japanese word for destructive, large ocean waves caused by underwater earthquakes or volcanic eruption and it propagates ocean-wide.

Much theoretical works, such as numerical models

for forecasting fare-field tsunami as well as the

Tsunami Inundation Modeling Exchange (TIME)

project, which Japan has been developing, are recommended to be pursued in consort with both

the field measurements of tsunami initial profile for submarine earthquakes and the development of

common database and maps of tsunami disasters; the working group strongly endorses to establish a satellite communication system, both voice and

data channels, to transmit seismic waveform data

to strategically located ground stations in an around the Pacific basin, and to use the PEACESAT

system of the U.S. for this purpose.

To mitigate the loss of life and property from

natural disasters by the judicious use of scientific knowledge emerging, a comparison of current

mitigation policy and programs between Japan and

U.S. is recommended to be carried out in the areas

of (1) education and awareness; (2) preparedness

and response; (3) seismic construction provisions

for new and existing buildings; and (4) land-use

planning practice. The earthquake group strongly

feels that both Japan and the U.S. serve the world

community through this bilateral agreement by

joint research and monitoring activities in the third

world countries, upholding the spirit of the

International Decade for Natural Disaster Reduction

(IDNDR), 1990-1999, transcending national

boundaries, for big earthquakes happen at many

places in the world and all of these events must be

scrutinized scientifically and their impacts upon

humanity minimized.

5.3 Severe storms group

The typhoon group was renamed "Severe Weather

Phenomena Including Typhoon and Local Heavy

Rains Group" (Severe storms group from now on).

The group now encompasses severe local storms

including heavy rainfall, floods, and hydrological

aspects as well as disaster preparedness.

Hydrological studies interlink volcano, landslide,

and severe storms in reference to physical process

studies. For instance when Mt. Pinatubo erupted

a typhoon passed over, thus causing much damages

to the communities nearby by volcanic ash flow. The predictive skill of the typhoon is gradually

improving on the basis of the advancement in observing systems, data processing systems and

numerical prediction models. Further improvement

will be accelerated by cooperation between two

countries through exchange of technical

information in numerical modeling, ongoing and

future satellite observations including new

measurement missions of the TRMM (Tropical

Rainfall Measurement Mission). Basic typhoon

research projects proposed includes the

development of diagnostics schemes for determining horizontal wind distribution in tropical

cyclone intensity; understanding the mechanisms

of typhoon movement in relation to asymmetric

typhoon structure and environmentally large-scale

8
flow, and developing a more sophisticated

convention parameterization scheme for moist

convection. The possibiUty of a change in the

numbers and intensities of tropical cyclones due to global climate change is of great concern to many

coastal and sea-going nations. Research should be

undertaken to investigate the effect of global

climate change on tropical cyclone activity using a sophisticated numerical model.

Severe local storms including heavy rainfall and

floods are affecting both countries continually. On

account of recent technological advances in atmospheric and hydrospheric measurements, we

can now link our emerging technological

knowledge to assess and mitigate hazardous

hydrologic disasters such as floods, flash floods,

storm surges, heavy snow, freezing rain, lightening.

The group recommends that joint efforts be

undertaken to incorporate new measurements by

satellite remote sensing, new lightening detection

system, high-resolution rain-gauge networks, and

polarization measurements of hydrometer particle

electric fields. They greatly help improve

hydrometerological model to better prediction of severe local storms and floods.

As other working groups recommend, the severe storms group considers disaster preparedness issue

essential, continual and evolutionary; the public use of and reaction to severe storms warning must

be timely and effective. The group as a first step

in the continuing dialogue and cooperation

between Japan and U.S. recommended a meeting

of severe weather experts in Norman, Oklahoma,

during the STORM-Fronts Experiment Systems Test (STORM-FEST), 1 February - 15 March 1992, to observe and work with the novel instrumentation

and experimental systems deployed, which include

NEXRAD Doppler radars, profiler network, new

automated surface observation network, research

aircraft and special soundings. This meeting was

held during 8-10 March 1992 to plan for the 3rd

UJST workshop for natural disaster reduction, Mt. Hood, Oregon, 7-11 September 1992.

5.4 Landslide group

The landslide research communities of both Japan

and the U.S. have been working together since

1980. Japan is a recognized leader in research

facility utilization, and the U.S has the unequalled

natural laboratories. Ongoing cooperation include

simulation and field observations of slope processes

of snow avalanches, landslides, debris flow and

stream channel processes. Physical process studies

jointly undertaken are slope failures and sediment

transport, rainstorms that trigger landslides and

debris flow, and that associate with extensive

volcanic cones.

The UJST separately supported a Japan-U.S.

symposium on snow avalanche management,

landslide prediction and control under the old

UJST agreement at Tsukuba, Japan, 30 September

-5 October 1991, where the snow avalanche and

landslide specialists of the two countries exchanged

their most up-to-date research efforts. Comprehensive future collaboration plans itemized

in the landslide group report encompasses (1)

landslide hazard assessment and information

exchange assisted by experimental study of

mechanism of landslide initiation and GIS

(Geographic Information System) technology to overlay stability attributes; (2) landslide hazard

mitigation by monitoring triggering mechanism and

preparing hazard maps; (3) mechanism studies on

the transformation of landslides to debris flows; (4)

debris-flow hazard assessment and mitigation by

better understanding of physics of flow and by

preparion of hazard maps; (5) initiation of

movement and mobility by snow and ice by

studying its dynamics in both laboratory and field settings; (6) research on subaqueous mass movement; (7) flood hazards arising from landslide

dams and sedimentation within by exchanging data on well-documented cases and educating public

officials and population to long-term hazards; (8)

elucidating the mechanism of earthquake-induced

landslide; and (9) impact of the climatic change on

landslides, including receding glaciers triggering

new landslides.

5.5. Socioeconomic group

The socioeconomic group recommends the adoption

of a cross-cultural approach to socioeconomic

studies of natural disasters that will enhance both

robustness and efficiency of our societal disaster

prevention system worldwide. The group

recommends a pilot project between the two

countties; it is a cross-cultural and cross-hazard

comparison for earthquake and typhoon disasters. Already considerable amount of data exists in both

countries. Since the ultimate goals of social

scientific investigations of natural disasters is to produce knowledge to enhance the adoption of

effective countermeasures, such as mitigation and

preparedness, by people at all levels of human

aggregation, including individuals, families.

9
governments, and businesses, factors considered

include elements of culture, experience, risk communication, conditions and perceptions that

collectively form risk perception. All of these

factors then are fed to the ultimate risk management for both mitigation and preparedness.

The group proposes to obtain representative

samples of 500 to 600 respondents via telephone

interviews from four study sites, Tokyo, Nobi Plain

in Aichi-ken, northern or southern California and

South Carolina. A planning meeting for the pilot

project in 1992 will construct the interview

schedule and standardized coding procedures. The

project proposed has a duration of two years.

5.6 Human health group

The human health group has had the benefit of

working together in the third world countries

providing medical and health reliefs following

natural disasters. Japan and the U.S. have shown

leadership and cooperation to improve disaster

medicine preparedness and response. Sixteen

recommendations prepared by the group fall under

the three categories of (1) improved disaster

preparedness; (2) response and operation

cooperation, and; (3) disaster medicine research.

Specific recommendations include the development

of disaster medicine and disaster epidemiology

centers at academic institutions; development of

their curricula and practical textbook on disaster

medicine based on the expertise and experience of

the two countries; development of methods of

rapid and accurate health needs assessment in disaster situations; development of standardized

triage system; conducting joint workshops on

focused topics, such as trauma surgery; tropical

medicine, and epidemiology of disasters; cooperative research in crush syndrome and

infectious diseases; joint database development;

human health technology transfer to developing

countries; taking a leadership role in carrying out

the five priority demonstration projects

recommended by the IDNDR Scientific and

Technical Committee, which include improved

hospital disaster planning in hurricane-prone

regions and health hazard mapping. It is also important to consider the linkage between climate

change and human health recognizing that climate

change would alter the ecosystems of many

disease-carrying vectors and agents, such as

viruses, bacteria, parasites, plants, mosquitoes,

snails, and increased heat-related illnesses such as heat stroke and death and injuries from coastal flooding.

6. CONCLUSIONS

Much progress has been made by the two

workshops of 1990 and 1991. As the third year of

IDNDR goes on now, execution, rather than

planning of pilot projects; specific cooperation,

rather than general; knowledge synthesis and

application of emerging knowledge to prepare for and mitigate natural disasters loom and guide the

two countries' joint endeavor. The third workshop

at Mt. Hood, Oregon, 7-11 September 1992, will focus on hazard mapping by the use of CIS

(Geographic Information System) and other means

to chart various hydrologic factors, earthquakes and

volcanic eruption prediction pattern to prepare for natural disasters yet to happen. Defense never

rests, and the best of two countries' academic and

public talent and government agencies continue to work together for the welfare of humanity; both

UJNR and UJST efforts are to enhance our

collective quest to reduce natural disaster damages

the best way we can.

7. ACKNOWLEDGEMENTS

The bulk of my text came from the executive

summary and the introduction I was asked to write

by the 2nd UJST workshop to prepare its proceedings. I am responsible for interpreting and

summarizing adequately the six group-deliberations

the 53 workshop participants at Karuizawa

laboriously developed. Mr. Keiji Doi of the Science

and Technology Agency has been my constant co- worker to issue the 2nd workshop proceedings; I am most grateful.

8. REFERENCES

1. Bergner, T.A. (1991). "United States Federal

Subcommittee for Natural Disaster Reduction,

History and Perspective." Preprints for the Second

Workshop on Natural Disaster Reduction under

U.S.-Japan Science and Technology Agreement,

Karuizawa, Nagano, Japan, 23-27 September 1991,

Science and Technology Agency, Prime Minster's

Office, Tokyo, pages 211-220.

2. Japan International Science and Technology

Exchange Center (1992). "Proceedings of the

Second Workshop for Natural Disaster Reduction,

Japan-United States Science and Technology

Agreement, Karuizawa, Nagano, Japan, 23-27

September 1992", Science and Technology Agency,

Tokyo, 520 pages.

10
Theme I

Wind Engineering
I

I
Flutter Characteristics of Super Long-Span Suspension Bridge

by

Koichi Yokoyama\ Tokida Kanazaki^ and Masahiko Yasuda^

Abstract

Since the collapse of the old Tacoraa Narrows Bridge, aerodynamic stability has

been verified through wind tunnel tests at the design stage of flexible bridge

structures. In Japan verification of

aerodynamic stability of suspension bridges

has generally been carried out through

spring-mounted rigid model tests using a section model (section model tests). As for

the test conditions of a section model test,

primary torsional oscillation is selected as well as vertical bending oscillation, whose

mode shape is similar to that of torsion.

This wind tunnel testing method is based on the fact derived by F.Bleich that oscillation

of a three-dimensional structure like a longspan suspension bridge can be treated as that

of a rigid section model under the assumption

that wind force is uniform along to the

longitudinal direction and that both mode

shapes for vertical bending and torsion are almost the same. However, the coupled flutter observed in

the full bridge model wind tunnel tests of

the Akashi Kaikyo Bridge is entirely

different from the general concept and the

mode shape of vertical bending motion was very much complicated. This phenomenon

implies that there is a limitation for the

verification method of the aerodynamic

stability through section model tests. This paper shows characteristics of the

coupled flutter observed in a smooth flow

test and discusses problems of the current

verification method.

Keywords : Aerodynamic stability,

Suspension bridge. Coupled flutter.

Full bridge model test, Wind tunnel test

1. Introduction

The Akashi Kalkyo Bridge is now under

construction between Kobe-city and the Awaji

Island. The bridge is a super long-span

suspension bridge with total length of 3,910

meters. The main span length of 1,990 meters

is far beyond that of the current world

longest suspension bridge, the Humber Bridge.

General view of the Akashi Kaikyo Bridge is shown in Figure-1.

The relation between span length of

suspension bridges and principal frequencies

or lateral deformation during a design storm

is illustrated in Figure-2. The longer span

length becomes, the lower frequencies become

and the bigger deformation becomes.

Therefore, it should be recognized that it is far more difficult to secure the aerodynamic

stability of the Akashi Kaikyo Bridge than

that of ordinary suspension bridge and that

it is important to estimate the deformation

by wind load more accurately than ever. There are three kinds of wind tunnel

testing methods to verify aerodynamic

stability of a suspension bridge, ® section

model test, ©taut strip model test, and

©full bridge model test. It is said that

full bridge model tests are most reliable if models are well simulated not only in shape

but also in dynamic properties.

Accordingly, wind tunnel tests using a full bridge model of the Akashi Kaikyo Bridge

were planed and are now in progress at the

large wind tunnel in PWRl. The main purpose

of the test is to verify the safety more

accurately. Therefore, it is required that

shape and elasticity of the model should be

similar to the prototype bridge as much as possible

.

Tests of completed bridge in a smooth

flow and in a boundary layer turbulent flow

have already been completed, and tests for

different wind direction and for an erection

stage are in progress . This paper shows that the aerodynamic

stability of super long-span bridges cannot

be estimated through spring-mounted rigid

model tests only, focusing on the coupled

flutter observed in the smooth flow test.

1) Dr.Engrg. , Head, Structure Division,

Structure and Bridge Department, Public Works

Research Institute, Ministry of Construction

2) Ms.Engrg., Senior Research Engineer, ditto

3) Ms.Engrg., Head, First Design Division,

Design Department, Honshu-Shikoku Bridge

Authority

13
A series of tests are being carried out

as "A cooperative research on wind resistant

design method for super long-span suspension

bridges" between the Public Works Research

Institute and the Honshu-Shikoku Bridge

Authority.

2. Purpose of the study

In Japan, spring-mounted rigid model

tests are generally carried out to verify the

aerodynamic stability of a long-span

suspension bridge. The fact that oscillation

of three-dimensional structures like a suspension bridge can be treated as that of

rigid section model was derived by F.Bleich"

under the assumption that wind force is uniform along to the longitudinal direction

and that both mode shapes for vertical

bending and torsion are almost the same. This

assumption is said to be adequate to completed suspension bridges except such a suspension bridge as partially installed with

sound-barrier. This assumption is a base that

the aerodynamic stability of completed

suspension bridges can be verified through

spring-mounted rigid model tests, in which

the lowest torsional and vertical bending

frequencies are simulated. The modes for both

oscillation are usually first symmetric.

In the results of full bridge model wind

tunnel tests of the Akashi Kaikyo Bridge,

however, counter-evidence against the

assumption is clearly recognized.

Here, summarized are analysis of the

results and problems of existing estimation

method through the section model test.

3. Test conditions

The general layout of the large wind

tunnel facility is shown in Figure-3.

Properties of a smooth flow in this wind

tunnel (width: 41 meters x height: 4 meters) were measured as follows,

a) Turbulence Intensity : less than 0.5 %

b) Variation in mean speed : less than 3 X Consequently, uniformity of the flow is sufficiently satisfied.

The full bridge model was fabricated

with a geometrical scale of 1/100, having a total length of 40 meters. The model is shown

in Figure-4. As design considerations were reported last year^' , Table-1 shows natural

frequencies of the model comparing with the

required values.

Froude number similitude has to be

satisfied in wind tunnel tests of the full

suspension bridge model, because

gravitational effects are dominant. This

means that ten times of wind velocity in the

wind tunnel is that for prototype bridge and

that the required natural frequencies are ten times of those of prototype. Natural

frequencies measured at dynamic tests of the

model have a good agreement with the required

values, which shows that the model was fabricated with good accuracy.

4. Results of full bridge model tests

The followings are results of full

bridge model tests of basic design with a cross section of a stiffening girder shown in

Figure-5. The results are reported with those

of section model tests.

4.1 Static torsional deformation

Relation between wind velocity and

static torsional deformation at the mid-point

of center span is shown in Figure-6. At wind

velocity of 6 m/s, that is equivalent to design wind speed of 60 m/s for the prototype

bridge, torsional deformation of -2 degrees

was observed. This deformation was far beyond

that of ordinary suspension bridge.

Aerodynamic sensitivity of this cross section

to the attack angle was well understood

through the section model tests that had

already been carried out. These facts

indicate that the first assumption by Bleich

is not yet applicable to this model.

In addition, through the tests and

analysis, it reveals that major cause of

torsional deformation is not the aerodynamic

moment but torque from inclined suspender

rope tension caused by the difference of

lateral deflections between cables and a stiffening truss (see Figure-7).

In Figure-6, there is a difference

between the torsional deformation of the

model and analytical value of the prototype

bridge. The difference is mainly due to restrictions in fabrication of the model. For

example, (D longitudinal stiffness of towers cannot be simulated, (D apparent fixing

points of suspender ropes are higher than

analytical model because of the stiffness of

fixing device, and (3) torsional stiffness

declines when V-shape springs to simulate the

stiffness of a stiffening girder are deformed

by lateral deflection.

The aerodynamic performance of the model

should be modified considering the difference

of torsional deformation in flutter analysis.

14
4.2 Flutter characteristics

1) Change of damping

Relation between wind speed and

logarithmic decrement is shown in Figure-8.

The damping of torsional mode was settled up until wind speed of 7 m/s, and it declined

drastically beyond 7 m/s. Wind speed of 8.5

m/s (85 m/s for prototype) was flutter onset velocity, where damping changed from positive

to negative. Here, the torsional oscillation

was not pure torsion observed in windless

condition, but coupled oscillation with

vertical bending mode, which was observed

from comparatively low wind speed.

Damping of pure vertical bending motion

increased drastically along wind speed.

In case of the section model test of the

attack angle of 0 degree, damping decrease

was observed from wind speed of 2 m/s, and

flutter occurred at 6 m/s.

In wind tunnel tests of suspension

bridges, onset velocity of flutter is most

noticed. The onset velocity of the section

model test and full bridge model test cannot

be compared directly because of the effect of

torsional deformation of the full bridge

model, so disagreement of onset velocity

obtained from each test is a matter of course

.

2) Change of natural frequency

Change of natural frequencies by

aerodynamic force along with wind speed is shown in Figure-9. Frequency of torsion was decreasing as wind speed became higher, and

at onset velocity the frequency became 90 %

comparing with that in windless condition. On

the other hand, frequencies of bending motion

were scarcely decreasing.

This tendency of natural frequencies is almost same as that of section model tests.

3) Movement of rotation center

On the flutter motion at the wind speed

of 8.5 m/s, movements of rotation center were calculated from amplitude of windward and

leeward cords. Figure-10 illustrates the

position of rotation center measured along

bridge axis. The rotation center lay on the

windward side at the midspan, on the leeward

side at quarter point of center span, and on the windward side again at the middle of side

spans. Movement of rotation center must be

uniform along the bridge axis according to the assumption by Bleich, but it clearly

varied.

Flutter which was observed in the

section model test was also bending-torsion

coupled one in which the rotation center lay

on the windward side, and the behavior was almost similar to that of the midspan.

4) Modes of flutter

Torsional and vertical bending response

at every 1/8 natural period are illustrated

in Figure-11. In the same figure, torsional

and vertical modes in windless condition are also indicated. As for the torsion, flutter

mode was the same as the first symmetric one

in windless condition. As for the vertical

bending, however, flutter mode did not agree

with any single mode in windless condition,

but very complicated.

Moreover, there was a phase angle

difference of 20 degrees between the vertical

and torsional component of the motion at midspan. This difference of phase angle

agrees with the fact that rotation center lay

on the windward side at the midspan or at side spans. Rotation center lay inside of girder

width at the midspan because torsional motion

was dominant, but lay outside of the girder

width at side spans because vertical motion

was dominant there.

At around quarter point of center span, however, a difference of phase angle was approaching to 180 degrees. Though this

agrees with the fact that the rotation center

lay on the leeward side, variety in movement

of rotation center is opposed to the

assumption by Bleich.

5. Comments on verification method

of aerodynamic safety

The full bridge model wind tunnel tests

of the Akashi Kaikyo Bridge provided

counter-evidence contrary to the Bleich's

assumption that aerodynamic stability of

three dimensional line-like structures such

as suspension bridges can be predicted by

spring-mounted rigid model test using a section model, which has only two-degree-of

freedom. For example, torsional mode during

flutter is the same as the first symmetric

mode in windless condition, but vertical mode

is similar to neither the torsional mode nor

the first symmetric vertical mode in windless

condition. Vertical mode during flutter is very much complicated. By this reason, it is probable that flutter onset velocity of the

prototype bridge cannot be predicted well

from section model tests.

Next, the causes of this complicated

coupled flutter are discussed referring to the assumption by Bleich.

15
(1) The assumption, that aerodynamic force

is uniform along to the bridge axis, is seriously disturbed by large torsional

deformation. In other words, relative angle

of attack is largely varied along the bridge

axis by the static deformation.

In case of ordinary suspension bridges,

which have a center span of about 1,000

meters, static torsional deformation was negligibly small to estimate the aerodynamic

stability of the bridge. By the test and

analysis, however, the possibility to induce

a bigger deformation than ever is indicated

in case of a super long-span bridge, which

has center span of about 2,000 meters. Therefore, the effects of torsional

deformation on the aerodynamic stability must

be sufficiently considered.

(2) The assumption, that the torsional and

vertical mode in windless condition are almost similar, is applicable to the center

span, but is not to side spans as shown in

Figure-11. Torsional component is small at side spans comparing to vertical component.

These two factors are thought to be causes that coupled flutter of the full

bridge model was different from that expected

through section model tests, and vertical

motion was complicated along the bridge axis.

In addition, it was major cause of

difference between test results that the

flutter of the Akashi Kaikyo Bridge was not a stall flutter ( single-degree-of -f reedom

flutter of torsion), which is observed for

ordinary truss stiffened suspension bridge,

but a coupled flutter.

Hereafter, planed is recently proposed

flutter analysis using measured flutter

derivatives and some vertical modes with

primary torsional mode in windless condition.

By analysing the cause of difference between

the test results and by understanding the

effects of each cause, limitation of the

section model test will be made clear, and it will become possible to predict aerodynamic

stability of super long-span bridge more precisely than ever.

super long-span bridges only through the

section model test. In this sense, the

meanings of full bridge model tests are significant

.

The causes of a complex mode flutter are thought to be as follows:

(1) Aerodynamic force acting on the girder

is not uniform along bridge axis because

of big torsional deformation of the

girder.

(2) The difference in primary mode shape

between torsion and vertical bending is comparatively big.

All of these factors originate in the

span length of the Akashi Kaikyo Bridge.

Hereafter, the estimation method of the

aerodynamic stability for super long-span

bridge will be proposed by establishing the

theory to explain the flutter performance of

the full bridge model.

References

1) Bleich F., et al.: The Mathematical Theory

of Vibration in Suspension Bridge, Dept. of

Commerce. U.S. Gov. Print Office, 1950

2) Yokoyama K. , Kanazaki T. and Yasuda M.,

Boundary Layer Wind Tunnel Study on Full

Aeroelastic Long Span Bridge Model, Proc. 23rd Joint Panel Meeting on Wind and Seismic

Effects. UJNR, 1991.

3) Miyata T.. Yokoyama K., Yasuda M..Hikami Y. Akashi Kaikyo Bridge:Wind effects and full

model wind tunnel tests. Proc. of the first

international symposium on aerodynamics of

large bridges, Copenhagen, 1992

6. Concluding remarks ,

In the results of wind tunnel tests using a full bridge model of the Akashi

Kaikyo Bridge, coupled flutter of complex

mode shape was observed. Such flutter has never been predicted from section model tests and onset velocities obtained from both tests

do not agree well.

The results implies that it is difficult

to predict onset velocity of flutter for

16
KOBE

960 1 990 960 AWAJI ISLAND

1A 2P 3P 4A

.36.5 lA) GENERAL VIEW

46.5

EE

35.50

2.50 10.75 10.75 2.50

ROAD LEVEL

3.50

14.^ STIFFENING GIRDER

MAIN TOWER ,g) ^^^^^^ jq^^^ stiffening GIRDER

Figure-1 General veiw of the Akashi Kaikyo Bridge (unit:in)

i N

40-

<a

t-, <u 3 20H

10-

^0.4- c

0.3

0.2

0.1-

0

-L 0

Lateral displacement at midspan

Torsional frequency (1st syram.)

Vertical bending

frequency (1st symm.)

500 1000 1500 2000 Center span length (m)

Figure—2 Structural properties of long-span bridges

17
22.500 22.500

i

> y >

) 4.000 1

0

41. 000

FAN

®-® SECTION

I ! i TEST SECTION 30,000

75.400

(D-(D SECTION

Figure—3 General layout of

large boundary layer wind tunnel (unitimm)
Table— 1 Natural frequency of the full bridge model

No Mode

f requency

r e q u i red full model

1st symm. bending

0 . 6 3 8 Hz 0 . 6 4 7 Hz

1st ant i -symm. bending

0 . 7 4 5 Hz 0 . 7 5 7 Hz

2nd anti-symm. bending

0 . 8 3 5 Hz 0 . 8 5 5 Hz

2nd symm. bending

1 . 2 1 3 Hz 1 . 2 4 5 Hz

1st symm. torsion

1 . 497 Hz 1 . 5 5 7 Hz

1st anti-symm. torsion

6 2 . 0 7 7 Hz 2 . 0 7 5 Hz

1st symm. lateral

0 . 3 8 7 Hz 0 . 3 9 1 Hz

1st anti-symm. lateral

8 0 . 7 7 5 Hz 0.781 Hz

19
35.5

Figure—5 Cross section of truss stiffening girder (unitlcm)

Wind Speod ( m/s )

Figure—6 Torsional deformation by wind (midspan)

Cable

MH'-Moment caused Ma -Aerodynamic Moment Torsional Deformation

by suspender tent ion

Figure—7 Moment caused by lateral deflection

20
Q.d r

c

(DE

<D

SiO

<Vo

E<-« -P

•f-l

(-1 CO

tji) o

1st symm. bending

2nd symm. bending

• full bridge model test

X section model test (a = 0^ )

1st symm. torsion

4 6\. 8 10

Wind speed ( m/s )

0.05 ^

Figure— 8 Change of aerodynamic damping

21
2.0 r

1st symm. torsion

1.5

•- •

2nd symm. bending

1.0

1st symm. bending

II •—

r

0.5

0 2 4 6 8. Id

Wind speed ( m/s )

Figure—9 Change of response frequency

22
(a)Torsion in windless condition

1/4 point of center span

Center of side span

(b)Torsion during flutter

Figure— 10 Movement of rotation center

23
—-T-TTmTTl] ! iTTin "ma lA 2P 3P 4A

0

" m.

®

(D

( b ) 1st symm. bending

(1) Mode shapes in windless condition

0

'A 2P 3P 4A

( a ) torsion

®

'

.

—

1 l!^L. U_! s • '

•

®

.I**** **•••-

® (5)

(D

*• • ••••.«•

'

" 1, •! •••»••«* *,•!•«<*

0

' 1 17;;• tT-;T "• ;

1 1» n» "' '~*

®

\A 2P 3P 4A

( b ) vertical bending

one cycle

(2) Response during flutter

Figure— 11 Response of girder in one cycle during flutter

24
Predicting Wind Forces Using Computer Simulation of Air Flow Around Bridges

by

Okey Onyemelukwe* and Harold Bosch**

ABSTRACT

The finite-difference procedure has been used to solve the non-conservative form of the twodimensional, unsteady, incompressible, laminar,

Navier-Stokes equations in a body-fitted

curvilinear coordinate system. In order to avoid

stringent numerical stability restrictions that lead

to excessive computation time, an implicit

solution scheme was employed. The finite- difference mesh was generated by using an

automatic grid generation technique, based on

the boundary-fitted curvilinear coordinate system

suggested by Thompson et al. (1977). The

advantage of using the boundary-fitted

coordinates being that the irregular body

geometries inherent in bridge shapes can be

handled more generically, than by the use of the more familiar rectangular cartesian coordinates.

Solving the primitive variable form of the

Navier-Stokes equations directly yields the

values of the velocity components and pressure. The resulting pressure coefficients are then

integrated along the body surface to obtain the

desired values of the force coefficients, drag,

lift, and pitching moment. Application to various

bluff shapes and bridge components, show

excellent prediction of the wind flow patterns

around the body, especially the vortex shedding

phenomenon. The fluctuating pressure

coefficients obtained are good qualitatively, but

sufficient time history record is required in order

to compute the mean values.

KEYWORDS: Numerical Simulation; Wind

Simulation; Flow Simulation; Bridges; Bluff

Bodies; Aerodynamic Stability.

1. INTRODUCTION

For many years, wind tunnel testing and full scale measurements have been the primary means of obtaining the fluctuating wind forces on bridge structures. The George Vincent

Memorial Aerodynamics Laboratory of the

Federal Highway Administration (FHWA),

Mclean, Virginia, remains one of the few

facilities in the United States dedicated to conducting wind related research studies, solely

focusing on bridge structures and its components. In recent years, however, due to the ever increasing availability of fast, efficient and relatively inexpensive computers, coupled

with the successful implementation in the

aeronautical field, numerical simulation of wind

effects on bridges has become a viable research

area for Wind Engineers. In this regard,

Computer Simulation Software is currently being

developed at the FHWA Aerodynamics

Laboratory, to numerically predict the wind flow

patterns and wind induced forces on bridge

structures. When evaluating wind forces on a bridge structure, it is generally assumed that the

critical wind direction is perpendicular to the

longitudinal axis of the bridge. In other words,

it is the 2-D Navier-Stokes equations that are solved. Although the natural wind is highly

turbulent and random in nature, this phase of the

research, as a start, solves the flow equations of

motion assuming laminar flow conditions.

Graduate Research Fellow, Federal Highway

Administration, Turner-Fairbank Research

Center, Mclean, Virginia 22101.

** Research Structural Engineer, Federal

Highway Administration, Turner-Fairbank

Research Center, Mclean, Virginia 22101.

25
The numerical scheme used, employs the finite- difference procedure in solving the governing

2-D, unsteady, incompressible, laminar, NavierStokes equations, in a body-fitted curvilinear

coordinate system. This paper discusses the grid

generation technique used, the governing

equations both in the physical and the

transformed planes, the finite-difference scheme,

and the results from application of this numerical

scheme to various bridge elements.

2. GRID GENERATION TECHNIQUE

The body-fitted curvilinear coordinate system is one in which, at least one coordinate line

coincides widi the boundary surface of the body

in the flow field. This automatically eliminates

the need for interpolation in the finite-difference

implementation of boundary conditions for an

irregularly shaped body. This grid generation

technique involves the solution of an elliptic

partial differential system in the physical plane.

To solve the system of equations, a coordinate

transformation from the physical plane (x,y), to a uniform rectangular computational plane (^,7j)

is necessary. Mathematically, this transformation

is stated as.

Poisson Equationr\ = n{x,y) (1)

where the transformed coordinates (^,t?) are the

independent variables, and the physical

coordinates (x,y) are the dependent variables.

The elliptic system of equations solved are either

of the Laplace type or the Poisson type, which

are both stated below:

Laplace Equation-

= 0

(2)

(3)

After transformation of these equations to the

computational plane by using the chain rule of

differentiation, the resulting form of the Laplace

equation that is actually solved for die grid point

values are,

+ cx

11 = 0

(4)

Dirichlet boundary conditions consisting of the

physical coordinates of the body geometry, and

the outer flow boundary must be specified,

before the solution can proceed. The coefficients

a, b, c, are transformation metrics given in terms of the following difference expressions,

C = X]

5 * yi

(5)

The functions Pd,?;) and Q(^,j;) in the Poisson

equation are chosen to enable the control of the

grid lines within the flow field. Thus the grid

lines can be clustered towards the body surface

where high flow gradients exist. Figures la. and

7a. show die finite-difference grid network

generated by this procedure, around a square

cylinder and a typical bridge deck cross-section

respectively. Note that the grid lines are packed

close to the body surface.

3. FLOW EQUATIONS

3.1 Physical Plane

In the physical plane, the governing flow

equations are the 2-dimensional, incompressible,

laminar momentum equations, and the continuity

equation. The dimensional form of the flow

equations are usually non-dimensional ized with

respect to the flow parameters of the undisturbed

flow region. The primitive variable form of the

26
non-dimensional flow equations are :

a + V = 0 X y

+ UU^ + vUy

= -p.

Re (6)

Re

Instead of solving the continuity equation, a new

poisson equation in pressure form is solved. The

need for the poisson pressure equation is to eliminate the pressure-velocity coupling that

exist between the momentum and continuity

equations. Also, the elliptic nature of this poisson type pressure equation allows for the

elliptic characteristic of the physical flow to be

introduced into the model. Again in the physical

plane, the poisson pressure equation solved is:

(7)

u-momentum-

' J J

J J

ReJ^

(8)

v-momentum-

' J J

J J

ReJ^

(9)

Pressure Equationwhere,

D = u + V

The derivation of this pressure equation ensures

the satisfaction of the continuity requirement.

However, the temporal D term is purposely not

set to zero according to Harlow and Welch

(1965), to avoid numerical instabilities that would otherwise occur.

3.2 Transformed Equations

Before the flow equations can be solved in the

computational plane, a similar transformation

procedure stated earlier must be applied. After

introducing the applicable transformation metrics

for the equations, the form of the transformed

momentum equations solved are :

= - ^, (10)

where.

_

{ap^^ - 2 bp^^ ^cp^^^ dp^ ^ ep^)

' J" J"

J

a, b, c, d, e, x^, x,, y^, y^ are all transformation

scale factors or metrics, while J is the Jacob ian

of transformation and Re is the flow Reynolds

number.

27
3.3 Boundary Conditions

The boundary conditions used for the momentum

equation are the no-slip boundary conditions i.e. u=0, v=0. For the pressure equation, a Neumann type boundary condition is used at the

wall surface dp/dn, and p=0 is specified at the outer boundary. Note that the pressure being

referred to is the non-dimensional pressure,

hence, setting the outer boundary value which

occurs far away from the body in the

undisturbed flow region to zero is satisfactory. In the computational plane, periodicity was

employed in implementing the boundary

conditions along the ^ coordinate in the

numerical solution routines.

4. FINITE-DIFFERENCE SCHEME

The fmite-difference formulation is based on an

Upwind-Implicit scheme. A backward difference

is used in time, and central difference

expressions used for the space derivatives. Due

to numerical instability resulting from using a central difference for the first derivative

convection terms, the transformed form of a

third order upwind difference expression

suggested by Kawamura and Kuwahara (1984) is used instead. For the u-velocity convection term

in the transformed momentum equations, the

finite-difference expression used is :

,du . -"<.2 ^ ^Ki - ^i-i) ".-2

M , - 4u . + 6m. - 4Ui , + u ~ 1+2 «+l I <-l «-2 4A$

(11)

I?. The refers to the transformed form of

the non-linear velocity term in the flow

equations of motion.

5. SOFTWARE APPLICATION

The simulation software developed in this study

has been applied to a square cylinder, a

rectangular cylinder with depth to width ratio of

0.5, and to a typical bridge deck cross-section. In each case, the flow was impulsively started,

using the results from a potential flow simulation

as the initial iteration guess. A Reynolds number

of 1 X 10"' was used for all the flow problems.

This value is considered typical for the physical

flow regime in which most bridge structures

exist. The outer boundary for the flow around a square cylinder was placed at 10 times the body

length. Figure 1 show a portion of the mesh in the physical plane, and also plots of die flow

streamlines, velocity vectors and pressure

contours at the Instantaneous non-dimensional

time (NDT= u.t/L) of 50. The computed

pressure coefficient Cp is given by the nondimensional expression:

P - P«

(12) 1 2

The variables with oo represent the flow

parameters of the undisturbed flow region far from the body. In Figures 2 and 3, the

fluctuating nature of the wall surface pressure

distribution is depicted for non-dimensional

times of 50 and 75. In both cases, negative

suction pressures exist on the top, bottom, and

leeward sides as expected. On the windward

side, the pressures are characteristically positive

i.e. pressures directed into the wall. An

illustration of the vortex shedding phenomenon

is shown in Figure 4, for flow around a rectangular cylinder of depth to width ratio of

0. 5. The distribution of pressure on the body

surface of this rectangular cylinder at nondimensional time of 75 is shown in Figure 5. The instantaneous flow streamlines and wall

surface pressure distribution on both a square

cylinder, and a rectangular cylinder computed by

Inamuro and Saito (1985), using the discrete

vortex method are shown in Figure 6a. In Figure 6b, they show a comparison of their

calculated mean pressures to experimental values

from the work of Otsuki et al. (1974, 1978 &

1980). Note that their pressure plots are for values of pressure coefficients times -1,

1. e (-Cp). The values of instantaneous pressure in Figures 2, 3, and 5 are in qualitative agreement

with those in Figure 6. Finally, the results from

28
application to a bridge deck section are shown in Figures 7 and 8. Figure 7 shows the clustered

grid used, the instantaneous flow streamlines,

velocity vectors, and pressure contours at a nondimensional time of 20. The wall surface

pressure distribution is plotted in Figure 8. Again these values are qualitatively acceptable.

6. CONCLUSION

The computer flow visualization plots of the

flow streamlines, show excellent prediction of

the wind flow patterns around the various

bodies. On the other hand, the resulting pressure

coefficients are qualitatively acceptable at this time. Sufficient time history record is required

in order to compute representative mean values,

that can be directly compared with mean values

from wind tunnel experiments.

7. ACKNOWLEDGMENTS

Much of this work is from the first author's

PhD dissertation. The study was supported

by a Federal Highway Administration grant.

8. REFERENCES

1. Thompson, J. F., Thames, F. C, and

Mastin, C. W., (1977). "Boundary-Fitted

Curvilinear Coordinate Systems for the

Solution of Partial Differential Equations on

fields containing any number of Arbitrary

Two-Dimensional Bodies." NASA CR-2729.

2. Harlow, F. H. and Welch, J. E., (1965).

"Numerical Calculation of Time-Dependent

Viscous Incompressible Flow of Fluid with

Free Surface." Physics of Fluids . Vol. 8, pp. 2182-2189.

3. Kawamura, T. and Kuwahara, K., (1984).

"Computation of High Reynolds Number

Flow around a Circular Cylinder with

Surface Roughness." AIAA Paper. 84-0340.

4. Inamuro, T. and Saito, T., (1985). "A

Numerical Analysis of Unsteady Separated

Flow by Discrete Vortex Method."

Proc. 5th U.S. National Conf. Wind

Engineering . Nov. 6-8, Lubbock, Texas,

pp. 5B-17 - 5B-24.

5. Otsuki, Y., Fujii, K., Washizu, K. and

Ohya, A., (1974). "A Note on the

Aeroelastic Instability of a Prismatic Bar with

Square Section." Journal of Sound and

Vibration. Vol.34, part 2, pp. 233-248.

6. Otsuki, Y., Fujii, K., Washizu, K. and

Ohya, A., (1978, 1980). "Wind Tunnel

Experiments on Aerodynamic Forces and

Pressure Distributions of Rectangular

Cylinders in a Uniform Flow (Part 1, 2) (in

Japanese)." Proc. 5th and 6th Symp. Wind

Effects on Structures . Tokyo, pp. 169-175,

1978, pp. 153-160, 1980.

29
Figure 1. Instantaneous Flow around a Square Cylinder at

NDT= u„t/L = 50

30
Figure 2. Instantaneous Body Surface Pressure Distribution around

a Sqaure Cylinder at NDT= u„t/L = 50

31
32
Figure 4. Time History of Flow around a Rectangular Cylinder

(d/w=0.5)

33
Figure 5. Instantaneous Body Surface Pressure Distribution around

a Rectangular Cylinder (d/w=0.5) at NDT= u„t/L =20

34
U t/*-29

U t/d— 20

I

Figure 6a. Instantaneous Flow Streamlines and Pressure

Distribution (-Cp) Using Discrete Vortex Method

• Cat

Figure 6b. Mean Pressure Distribution from Discrete Vortex Method

and Wind Tunnel Experiment

35
Figure 7. Instantaneous Flow around a typical Bridge Deck Section

at NDT= u„t/L = 20

36
Figure 8. Instantaneous Body Surface Pressure Distribution around

a typical Bridge Deck Section at NDT= u„t/L = 20

37

Theme II

Storm Surge and Tsunamis

A Numerical Simulation of the Tsunami Generated by Landslide of

Mt. Mayuyama in 1792

by

Chiaki Goto* and Tomotsuka Takayama**

SUMMARY

Through the numerical simulation, tsunami

generated by landslide of Mt. Mayuyama in 1792

is investigated. Remarkable characters of

landslide tsunami including concentration and

divergence of energy by tsunami reflection due

to sea topography is clarified.

Key wards : Landslide Tsunami.

Numerical simulation, Mt. Mayuyama,

Simabara Bay. Ariake Sea

1. INTRODUCTION

By a gigantic collapse of Mt. Mayuyama in

Simabara Peninsla in 1792. a grate landslide and

tsunami was generated and attacked the coastal area around Simabara bay. The tsunami killed

about 15,000 persons and washed away many houses.

Volcanic activity of Mt. Unzen started after

200 years interval since two years ago. Mt. Unzen is a generic name for seven volcanos and

was an active volcano up to the end of the 18th

century. Because Mt. Mayuyama is a volcano of

Mt. Unzen. goverment feels misgivings about

generation of landslide and tsunsami same as the event in 1792.

Historical data Near Japan Islands gives an imformation that about 95 percent of tsunamis

are generated by sea bottom deformation due to earthquakes and only 5 percent of tsunamis are generated by landeslide or vocanic activity.

Characters of landslide tsunami may be differ

from the earthquake tsunami. In this paper.

Fig. 1 Water depth of Simabara Bay

characters of the tsunami in 1792 is

investigated by use of numerical simulation

model.

* Cheif, Ocean Energy Utilization Lab.. PHRI,

Ministry of Transport

** Directer. Hydraulic Engineering Division,

PHRI, Ministry of Transport

41
Kashima

Kawasoe

Yanagawa

Oomuta

Isahaya

Nagasu

Yokosima

Kawati

Tenniei

Misumi

unit: mimutes

Kashi

Isahaya

Nagasu

Yokosima

Kawati

Tenmei

Misumi

unit: mimutes

Fig. 2 Arrival time of tsunami front Fig. 3 Arrival time of the first crest of tsunami

2. NUMERICAL SIMULATION MODEL OF TSUNAMI

The governing equations of tsunami

numerical simulation are following nonlinear

shallow water theory ( nonlinear long wave equations ) including friction terms.

dr? dM ,dN + "1— + ~— = U

d t d X

dM _^ d

d y

( 1

)

d t d X

Ml

D

+

d y

MN

D

dx D^^^

(2)

dN

d t +

d y D

+

d X

MN

D

d y D

(3)

where x and y are the plain coordinates, t the

time coordinate, M and N the discharge flux in

the X and y direction respectively, t] the

tsuanmi elevation, g the gravity, h the still water depth, D the total depth and n the

Manning's roughness coefficient. In the

computation, the finite difference method of the staggard leap frog scheme of equations (l)-(3)

is adapted.

The tsunami computation is carried out in

42
Fig. 4 Distribution of tsunmai maximun height

the whole region of the Ariake sea and Shimabara

Bay. Distribution of water depth at the high water level is shown in Fig. 1. In the Ariake sea, averaged tidal amplitude is observed as 2.5 m.

As the point view for tsunami defence, the high

tide condision is adapted in the computation.

As a grid length of numerical simulation,

400 m mesh is used. Accuracy of the numerical

simulation depends on mesh size because the

trancated error of computation is growth in case

of rough grid length. It is known that more

than 20 grid points are necessary within one

wave length in the tsunami numerical simulation.

In the case of tsunami simulation in Ariake Sea

and'Simabara Bay. averaged water depth is about

30 m. And the tsunami wave length generated in landslide of Mt. Mayuyama can be estimated about

10 km. Therefore, the ratio of wave length to

grid size is about 25 so that the condision for

accurate computation is satisfied.

As a open boundary and sea shore condisions,

the ragiative and the reflective condisions are adapted respectively in the computation.

The landslide of Mt. Mayuyama is assumed

earth and sand of 4.6x10' m^ flows into the sea during 4 minites the same as Aida's" model.

The area which earth and sand flows is also the same region as Oota's" estimation.

3. RESULTS OF SIMULATION

3. 1 Arrival time of tsunami front

Figure 2 shows arrival time of tsunami

front in computational region. The figure is

43
drawed by the assumption of time that sea surface changes more than 1 cm.

At the opposite side of sea. Tenmei,

tsunami arrives at about 20 minutes after

tsunami occured. At Misumi, it takes about 15 minutes. On the other hand, at the inner part of

sea, Kashima and Kawasoe, tsunami arrives at "iS

minutes.

Figure 3 shows the arrival time of the

first crest. The first crest of tsunami reaches

for coast at the time after 5 minutes since the

front arrive.

3. 2 tsunami height

Figure 4 shows the distribution of maximum

tsunami height. The contour lines indicate

equi-height and numerials are tsuanmi height.

The lines at right and left sides of figure also

show the maximum tsunami height along the east

and west sea shore lines of Ariake Sea and

Smabara Bay, respectively.

At tsunami source region, near Simabara,

the maximum tsunami height is computed more than

20 m. At the opposite coast, the averaged height

is about 5 m and the distribution of the maximum

height has a tendancy that the height of the

center coast is .relatively low but both sides

are high. This resion can be think that main

part of tsunami propagates both side coast

because of tsnami reflection related to sea bottom topography. And the characters of

numerical results coincide with tsunami

historical data.

3.3 Wave profile of tsunmai

Near the tsunami source area, the first

wave is the most dengerous tsunami. But the

second and third waves are almost same height. On the other hand, the first wave is the biggest

in the opposite coast. The second and third wave

are relatively small compared with the height of

the first wave. In the inner are of Ariake Sea. tsunami height is very small.

4. CONCLUSION

The paper reports a numerical simulation of

tsunami generated by landslide of Mt. Mayuyama

in 1792 and discusses a few remarkable tsunami

behaviour in Simabara Bay and Ariake Sea. In a future, more detail tsunami numerical simulation

includinig run-up on land should be carried out

for tsunami defence works.

REFERENCES

1) Aida. I. : Numerical Simulation of tsunami

generated by lindslide of Ht. Mayuyama in 1792. Earthquake n , Vol.28, pp. 449-460, 1970.

2) Oota. K. : A study of gigantic collapse of Mt. Mayuyama. Report of Simabara

Volcanic- thermal spring Institute.

No. 5. pp. 6-35. 1970.

44
Empirical Simulation of Storm Histories for Coastal Design

by

Martin C. Miller' and Robin D, Reinhard^

ABSTRACT

Coastal erosion and flooding caused by

hurricanes have tremendous potential for

economic damage and loss of life in populated

areas. The U.S. Army Corps of Engineers is frequently asked to determine the feasibility of

coastal protection projects through evaluation of

the economic benefits of project alternatives.

The development of design criteria for water

level, waves, and beach profile change require

the evaluation of data collected over long time

periods, or the generation of data using tested

computer models. A methodology has been

developed at the U.S. Army Engineer,

Waterways Experiment Station, Coastal

Engineering Research Center for determining

project feasibility by the application of several

environmental models. The economic evaluation

depends on the application of a statistical resampling technique known as the "bootstrap"

which allows calculation of expected values of

an outcome by sampling, with replacement,

from a limited historical data set. This method

has been applied to a potential project at Panama

City Beach, Florida.

KEYWORDS: modeling; water level; simulation; "bootstrap"

1. INTRODUCTION

Many coastal engineering design problems

require estimates of extreme events such as wave

height and water levels. Estimates of normal

and extreme conditions are also required to justify projects that protect coastal property.

Comparisons are made between the

consequences of the occurrence of the design

storm with and without the protection project in place. The economic benefit of a coastal

protection project during the design event must

exceed the loss that would be experienced if the

protection were not in place. In the Gulf of

Mexico and along the southeastern U.S. coast,

extreme waves and water levels are caused by

hurricanes. Along the mid-Atlantic coast,

extremes may be caused by either hurricanes or

northeasters, and along the northeast coast,

northeasters provide the storms of record. The

coastal response to these events is extreme

shoreline erosion with consequent flooding,

property damage and possible loss of life. The

U.S. Army Corps of Engineers is often

requested to design measures that will protect

the coastal property from the effects of these extremes.

The statistical methods used to estimate the

extreme value of an environmental variable are well known (Borgman and Resio, 1982; Goda,

1978; Haring, et al, 1976). Data are collected or hindcast over a long period of time and

systematic procedures are used to plot the data

according to some probability distribution.

' Chief, Coastal Oceanography Branch, Coastal

Engineering Research Center, U.S. Army

Engineer Waterways Experiment Station,

Vicksburg, MS 39180-6199

^ Hydraulic Engineer, Coastal Oceanography

Branch, Coastal Engineering Research Center,

U.S. Army Engineer Waterways Experiment

Station, Vicksburg, MS 39180-6199

45
The resulting curve may be extrapolated to some

selected return interval to determine the

magnitude of the parameter selected. Though a data record of three to five times the maximum

return interval to be estimated is usually

recommended, in most U.S. coastal areas the

historical database is not adequate to represent

the full range of possible occurrences of storm

effects at a specified site. Future storms may

pass at different coastal crossing locations than

are present in the historical database and the

storms may have different characteristics of

central pressure, radius to maximum winds,

forward speed, duration, etc. It may, therefore,

be necessary to augment the historical data set with additional storms that could have occurred

at the site given the random nature of events. This paper briefly describes a statistical procedure which may be applied to augment the

historical database of hurricanes with

hypothetical storms that could have occurred at the site.

2. BACKGROUND

A study has recently been conducted by the U.S.

Army Engineer Waterways Experiment Station

(WES), Coastal Engineering Research Center

(CERC) for the U.S. Army Corps of Engineers

Mobile District to examine the feasibility of

providing a protective sand berm along an

eighteen mile section of Panama City Beach,

Florida (Figure 1). This beach is heavily

developed over much of its length, with single

family homes and condominiums built along the waters' edge. Inlets at each end of the beach

provide access to small boat harbors and quiet

coastal lagoons. The beach is susceptible to flooding from hurricane related storm surge, which could result in great loss of life and

property. Beach and flood protection is considered necessary, but the recreational and

aesthetic value of the area must also be

preserved. The protective sand berm concept was determined to be appropriate to provide

protection from flooding and to preserve, to the

maximum extent, the natural scenic and

recreational ambience of the area.

In order to determine the economic justification

for the project and to effectively design the

berm, it is necessary to evaluate the degree of

protection it will provide during storms of

various magnitude that may impact the site. From the designer's standpoint, it is desirable to know the return period of the storm that would

be protected against for the selected design.

From the economic standpoint, project

justification requires that the benefit derived

from the selected level of protection outweigh

the cost.

3. GENERAL APPROACH

The primary goal of this study is to provide an

estimate of the economic damages which might

occur at the Panama City Beach area due to a

tropical storm or hurricane. The methodology

used to obtain these estimates consists of a many

phased approach which relies on the application

of several numerical models. Available data on

the historic storms affecting the study area were

collected. A subset of storms, called the

"training set," was chosen which is representative of the full range of possible storm

conditions. Numerical models which estimate

the winds, water levels, waves, and beach

profiles were run for each member of the

training set to produce variables of interest, or "responses." A Gaussian "nearest-neighbor

interpolation" relationship was developed which

allowed the response for a storm to be

determined from the training set storm

characteristics without having to simulate the

responses of all of the numerical models for

every storm. One hundred statistical simulations

of 50 years each were produced. For each year

of a simulation the number of storms was set using a Poisson process. The responses for each

of these storms was determined by the random

"nearest-neighbor bootstrap" technique. The

simulated sequence of storm responses were

used to determine the return period levels of the

storm response variables. The simulated

sequences were also used in the economic model

to determine the storm damages and comparisons

46
were made between conditions with and without

the planned improvement.

4. STORM SELECTION

Tropical storm data used in this study was

obtained from the National Hurricane Center

(NHC) (Jarvinen, et al, 1984) and CERC's

Wave Information Study (WIS) data set (Abel,

et al., 1989). The latter contains much of the

basic information of the former but has been

augmented with additional details in order to hindcast the waves in selected storms. The

NHC set covers the years 1886-1989. The

significance of this data set is it's completeness,

which makes it valuable for determining the

frequency of different storm characteristics. The

WIS data set covers the years 1956-1975.

Hurricanes Juan and Kate, both of which made

landfall near Panama City Beach in 1985, were

added later. Only storms which exceeded the

severity threshold of wind speed of at least 34

knots within 75 nautical miles of Panama City,

Florida were selected.

The meteorological database included 34 storms

which occurred over the past 105 years. The set of 34 historical storms was expanded by running

each of the storms at four random tide phases, to obtain response variables for a total of 136

storms. Meteorologic data for 31 additional

storms was also gathered. These 31 hypothetical

storms did not actually occur as described but

are based on the characteristics of the historical

Gulf of Mexico storms. The hypothetical storms were chosen so that the entire range of storms

which could impact the Panama City Beaches

study area were represented. The total storm set contains 167 events (136 historical and 31

hypothetical storms).

A subset of the total storm database, referred to

as the training set, was established to determine

the relationship between storm parameters and

environmental responses. The range of storm

types found in the training set should reflect the

likely range of storms to be encountered.

Storms were chosen for the training set based on

the ability of the set to cover the characteristics

of a full range of storm types. This was done

by working backward through hurricane records

and eliminating storms that were near duplicates

of each other. The training set included 24

storms from the set of 136 historical storms plus

the 31 hypothetical storms.

Storm parameters were determined both at the

time of closest approach and at the time of

landfall. The set of storm parameters which was

chosen to describe the storms statistically are:

(1) the relative phase of high tide and storm

surge maximum, (2) the relative phase of tidal change and storm surge maximum, (3)

maximum wind speed in knots, (4) central pressure deficit in millibars, (5) radius to maximum wind in nautical miles, (6) forward

speed in knots, (7) course angle of the storm

track in degrees and (8) closest point of

approach distance from the study area in nautical

miles.

5. HYDRODYNAMIC MODELING OF

STORM EVENTS

Several numerical models were used to establish

the design criteria for the Panama City Beach

area and to provide the data necessary for the

economic evaluation of project feasibility. Most

of these were developed by CERC engineers or by contractors under the direct technical

supervision of CERC.

The hurricane wind-field was produced using the

Standard Project Hurricane Model (SPH). This

is a parametric model which represents a steadystate hypothetical hurricane defined by a set of

interrelated parameters such as eye location,

central pressure deficit, track angle, forward

speed, radius to maximum winds, etc. (Cialone,

1991). The model produces wind components

and surface atmospheric pressures. This model was run for each storm on a grid covering the

Gulf of Mexico. Results of the model were used

to determine water level and waves.

Total water level was obtained from ADCIRC

47
(ADvanced CIRCulation model) (Luettich, et

al., 1991; Westerink, etal., 1991). This model

is a time-stepping finite-element model

containing the effects of tidal forcing, surface

wind stress, atmospheric pressure variation, and

bottom friction. A major advantage of the

finite-element method is that the computational

mesh can be refined to represent the details of

local topography in the area of interest. Figure

2 shows the grid for the area of the entrance to

St. Andrew's Bay. The ADCIRC model was

tested against tidal data and against the measured

storm surge produced by two hurricanes (Kate

and Juan) impacting Panama City Beach (Miller,

et al., 1991). No tuning was done to the surge

model, though some adjustments were required

for the SPH to correct for landfall effects.

The waves were determined using the SHALlow

water WaVe model, SHALWV, which is a time-dependent, spectral hurricane wave-field

model. It has been previously tested against a large number of hurricane wave measurements

and used to model sixty-five hurricanes along

the Gulf and East coasts of the U.S. (Jensen, et

al., 1983). The wave spectrum at a point is represented by an array of discrete frequency

and direction components. At each time step of

the model, the wave energy is first propagated in the grid, taking into account refraction,

diffraction and dispersion. The wave energy in each frequency-direction component at each grid

point is then recalculated, taking into account

energy input by winds, energy dissipation by

breaking, and energy transfers within the

spectrum through wave-wave interaction

processes.

The results of the wave and surge stage

frequency analysis were used as input for the

numerical simulation model Storm-Induced

Beach Change (SBEACH) (Larson and Kraus,

1989; Larson, et al. , 1990). SBEACH was used

to evaluate the cross-shore transport of beach

and dune materials exposed to varying waves

and water levels under hurricane storm

conditions. The model calculates sand transport

produced by breaking waves, simulates erosion

of dune and berm beach fills, and develops the

evolution of the equilibrium beach profile

including the offshore bar formation. Longshore

transport was not modeled. Four representative

beach profile lines along the 18 miles of beach were evaluated using the SBEACH model. The

selection of the representative surveys were

based on property structural value, beach profile

shape, dune elevation and shape, longterm

erosion rates, and grain size. Four responses (1)

maximum linear feet of recession during the

storm event measured from zero National

Geodetic Vertical Datum (NGVD), (2)

maximum water level above NGVD, (3) wave

height at a depth of 49 feet, and (4) wave period were obtained from SBEACH to be used in the

economic analysis.

6. DEVELOPMENT OF RETURN PERIOD

FOR COASTAL STORM IMPACTS

The purpose of running a training set of storms was to determine a relationship between the

storm parameters and the response variables,

which are values for the storm determined using

the numerical models for waves, water level,

and beach erosion. Determination of this relationship removes the need for further runs of

the numerical model by allowing determination

of the water level, waves, erosion, etc. for a given site directly from the storm parameters.

The responses for the 112 storms which were

not run through the numerical models are estimated from the responses of the training set storms with a Gaussian Nearest-Neighbor

Interpolation. The Gaussian interpolation

determines a weighted average of the responses

of the training set of storms. For the purposes

of determining the relationship, a measure of

similarity between storms is required. This

measure, called "distance" in this study, should

possess the quality that the responses of storms

converge as the distance between them

decreases. The n storm parameters used can be considered the coordinates of an n- dimensional space in which storm i is represented by a vector, c-^ , where k = I, n.

48
The difference, or distance, between vectors i and j , d,j is the sum of the squares of the

difference between each component of each

vector

(1)

of the i-th historic storm, not part of the training

set, are then given a weighted average value of

all training set storms

Ew..r.. ij jk (3)

w..

This is the square of the usual Cartesian

difference of two vectors.

For this study, different storm parameters are weighted differently for evaluating the distance

between two storms. For instance, storms with

similar distances and maximum wind speed will

produce effects more alike than would storms

similar only in radius to maximum wind and tide

phase. Therefore, each parameter can be

normalized by its root mean square (RMS)

value, and assigned a scaling radius, 7?^ , and

weight, , based on the importance of the

parameter. The RMS values are used as a scale

factor around the mean value to normalize the

range of the parameters due to differences in units. The distance then becomes

d.. = Ji (2)

where the tilde indicates optional norming. For

this study, all parameters were normalized, and

all were set to unity (Borgman 1991). All

W^. were set to unity for most parameters. The

were set to ten for maximum wind speed at closest approach and distance of closest

approach, set to five for relative phase of high

tide and storm surge maximum, and set to three

for relative phase of tidal change and storm

surge maximum. The values were chosen so the

parameters that better describe the similarities

between storms are given higher weights. Thus,

wind speed of the storm, its distance, and its tidal phase of arrival are considered most

important in this study.

The response values, , for the k-th response

where the jt are the j-s of training set storms

only. The weights are set by a bell-shaped

("Gaussian") curve

Wj. = exp (4)

where D, is an effective width of the bell equal

to twice the average distance of the four nearest

training set storms to storm i .

If the mean frequency of storms is known, the

Poisson distribution can be used to determine the

number of storms in a given period. The

Poisson distribution is given by

Pr(5;X) = (5)

where Pr(5;X) is the probability of having

events in a period in which X is the mean

frequency of events per period. For this study,

the interval is 1 year and X — 0.3238 year''

(34 historic storms / 105 years). As a result,

the most likely event for a given year is no

storms (Pr(0; 0.3238) = 0.7234). The number

of storms in a given year of a simulation is chosen using a random number generator

combined with a Poisson probability distribution.

Once the number of storms in a year of a simulation has been found, it is necessary to determine the characteristics of each of those

storms. This is done by a nearest-neighbor

bootstrap method developed by Borgman (1990

and 1991). Bootstrap methods are based on

resampling of observed data (Efron 1982).

Multiple sequences of events can be developed

49
from a single sequence of observed data by

drawing random samples (with replacement)

from the observed data using the program

HBOOT. The multiple sequences can then be

used to estimate various statistics of the

population of which the observed data is a sample.

To determine the responses of a simulated

storm, one historic storm is selected at random

from the total storm set. Since all historic

storms have equal historic probability (each

occurred once), storms selected at random will

have the same probability density function (pdf)-

The responses of the new simulation storm are

set to a weighted average of the selected storm

and its four nearest neighbors, with the weights

determined at random. This averaging

procedure makes it possible for the simulation to have storms which have never occurred, but

which are relatively similar to the storms which

have occurred. It also allows an almost infinite number of hypothetical storms to be generated

since storms with different responses will be

simulated using the same reference storm.

The results from the statistical model are then

used as input for the economic damages model.

Data for the determination of the cost-benefit

ratio for the proposed project were developed

from 100 simulations of 50 years each of the

model HBOOT, which is based on the nearest- neighbor bootstrap technique described above

and elaborated upon in the references. The

rationale for the use of multiple simulations is that tropical storms (including hurricanes) are

relatively rare and have varied characteristics so no single 50-year simulation would be adequate

to determine the risks, costs, and benefits

associated with the project. By determining the

cost-benefit ratios for multiple 50-year

simulations using the bootstrap, one can determine the expected value of the 50-year cost- benefit ratio, as well as obtain a measure of the

uncertainty of the calculations.

Since the economic model used must be given

the year of occurrence of each event in a 50-

year simulation to allow discounting of damages

to a reference year, the 100, 50-year simulations

give the number of storms each year and the

magnitude of the response variables for each.

For a given 50-year simulation, the economic

damages can be found for each storm,

discounted to the reference year, and summed to give a damage value for that simulation. The

hurricane and storm damages were evaluated

using a time series simulation combined with a composite damage reduction analysis. The

composite damages method computes the

damages of a storm to each structure for the

three basic categories: inundation, wave attack,

and storm recession.

Once the structure and content damages from

inundation, storm-induced recession and wave

attack were computed, the maximum, or critical,

damage was chosen for each. Beginning in the

first year of the project life, the model

determines if any storms occurred, calculates the

damages for each structure and each storm,

increments the shoreline by one year of long

term change, then begins the calculations for the

second year. When the model completes the

calculations for one 50 year period, it begins

again for the next 50 year period. Other factors

that are considered are long-term erosion,

rebuilding constraints, back bay flooding,

downdrift impacts, existing structural

impediments to erosion such as seawalls and

bulkheads, and individual actions to protect

against erosion.

7. ACKNOWLEDGEMENTS:

The research presented in this paper, unless

otherwise noted, was sponsored by the Office of

the Chief of Engineers, U. S. Army, and is published by its permission. Panama City Beach

studies were conducted for the Mobile District,

U. S. Army Corps of Engineers.

8. REFERENCES

1. Abel, C. E., B. A. Tracy, C. L. Vincent

and R. E. Jensen, "Hurricane Hindcast

50
Methodology and Wave Statistics for

Atlantic and Gulf Hurricanes from 1956-

1975," Wave Information Studies of U.S.

Coastlines, WIS Report No. 19, Coastal

Engineering Research Center, U.S. Army

Engineer Waterways Experiment Station,

Vicksburg, MS, March 1989.

Borgman, L. E., "An Algorithm for

Optimal Efficiency in Selecting Storms for

Finite Element Modeling of Surge,"

Contractor Report, Coastal Engineering

Research Center, U.S. Army Engineer

Waterways Experiment Station, Vicksburg,

MS, 18 November 1990.

Borgman, L. E., "A Method for

Calculating Statistical Reliability of

Estimates of Future Hurricane Damages

and Effects," Contractor Report, Coastal

Engineering Research Center, U.S. Army

Engineer Waterways Experiment Station,

Vicksburg, MS, 29 April 1991.

Borgman, L. E. and D. T. Resio,

"Extremal Statistics in Wave

Climatology," in A. R. Osborn and P. M.

Rizzoli, Proc. International School of

Physics (Enrico Fermi), Topics in Oceanography, North-Holland Publishing

Co., New York, 1982, pp 439-471.

Cialone, M. A., "The Coastal Modeling

System: User's Manual," IR-CERC91-1,

Coastal Engineering Research Center,

U.S. Army Engineer Waterways

Experiment Station, Vicksburg, MS, 1991.

Efron, B., "The Jackknife, The Bootstrap,

and Other Resampling Plans," Regional

Conference Series in Applied Mathematics.

No. 38 . Society for Industrial and Applied

Mathematics, Philadelphia, PA, 1982.

Goda, Y., "The Observed Joint

Distribution of Periods and Heights of Sea

Waves," in Proc. 16th Conf. on Coastal

Engineering, Hamburg, 1978.

8. Haring, R. A., A. R. Osborne and L. P. Spencer, "Extreme Wave Parameters Based on Continental Shelf Storm Wave Records,

in Proc. 15th Conf. on Coastal Engineering,

Honolulu, Hawaii, 1976.

9. Jarvinen, B. R., C. J. Neumann and M. A.

S. Davis, "A Tropical Cyclone Data Tape

for the North Atlantic Basin, 1886-1983:

Contents, Limitations, and Uses," NOAA

Technical Memorandum, NWS, NHC 22,

National Hurricane Center, Miami, FL,

March 1984.

10. Jensen, R. E., C. L. Vincent, and C. E. Abel, "A User's Guide to SHALWV:

Numerical Model for Simulation of

Shallow-Water Wave Growth, Propagation,

and Decay; Report 2, SHALWV-Hurricane

Wave Modeling and Verification,"

Instruction Report CERC-86-2, Coastal

Engineering Research Center, US Army

Engineer Waterways Experiment Station,

Vicksburg, MS, June 1983.

11. Larson, M. and N. C. Kraus, "SBEACH:

Numerical Model for Simulating StormInduced Beach Change, Report 1, Empirical

Foundation and Model Development,"

Technical Report CERC-89-9, Coastal

Engineering Research Center, US Army

Engineer Waterways Experiment Station,

Vicksburg, MS, July 1989.

12. Larson, M., N. C. Kraus, and M. R.

Byrnes, "SBEACH: Numerical Model for

Simulating Storm-Induced Beach Change,

Report 2, Numerical Formulation and

Model Tests," Technical Report CERC-89-

9, Coastal Engineering Research Center, US

Army Engineer Waterways Experiment

Station, Vicksburg, MS, May 1990.

13. Luettich, R. A., J. J. Westerink, and N. W.

Scheffner, "ADCIRC: An Advanced Three

Dimensional Circulation Model for Shelves,

Coasts and Estuaries, Report 1: Theory and

Methodology of ADCIRC-2DDI and

51
ADCIRC-3DL," Coastal Engineering

Research Center, US Army Engineer

Waterways Experiment Station, Vicksburg,

MS, In Prep.

14. Miller, M. C, W. E. Roper, L. E. Borgman and J. J. Westerink,

"Development of Water Level and Wave

Height Design Data," Proc. Panel on Wind and Seismic Effects, UJNR, 23rd

Meeting, Tsuduba, Japan, 14-23 May

1991.

15. Westerink, J. J., R. A. Luettich, and N.

W. Scheffner, "ADCIRC: An Advanced

Three Dimensional Circulation Model for

Shelves, Coasts and Estuaries, Report 2: ADCIRC-2DDI Users Manual," Coastal

Engineering Research Center, US Army

Engineer Waterways Experiment Station,

Vicksburg, MS, In Prep.
GULF OF MEXICO

Figure 1. Northern Gulf of Mexico from Louisiana to Florida

Bight. Study site is at Panama City.

Figure 2. Details of finite element grid near Panama City,

53

Theme III

Joint Cooperative Research Program

The Japan-PRESSS Program— Analytical Tools in Japan

by

T. Okoshi S. Nakata

Structural Engineering Department

Nihon Sekkei Inc. Nishi-Shinjuku, Tokyo 163

Japan

ABSTRACT

In Japan, it is required the use of nonlinear

time- hi story analysis for the dynamic response

analysis of a reinforced high-rise concrete

building. There are two models used in the

dynamic response analysis, one is the lumped mass model and the other is the frame model.

The lumped mass model is formulated

distributing the weight as a lumped mass for

each story level, and using equivalent flexureshear springs which are obtained from the

static elasto-plastic analysis, and the

appropriate damping values. In the static

elasto-plastic frame analysis, the distribution

of the lateral forces is assumed, and the

incremental analysis is used. Each member of

the frame is modelled to be elastic with rigidplastic rotational springs at its ends. The

restoring force characteristics of this model are decided considering its flexural cracking and

yielding. Usually, the skeleton curve is tri- linear, and the hysteresis rule is degrading tri- linear, like the origin-pointed model or the

continuous degrading system (Takeda model),

for the case of the restoring force

characteristics of the lumped mass model,

llie same procedure for the static elasto-plastic

analysis, and the restoring force characteristics

of the lumped mass model can be used for the

dynamic response analysis of the frame

model.

1. Introduction

The nonlinear time-history analysis is used as

a check of responsed story drift of reinforced

concrete high-rise buildings in Japan. The

modal analysis is not used. This dynamic

response analysis is based on the mathematical

model of elasto-plastic characteristics of the

frame and the reduction of the members

stiffness due to flexural cracking and yielding.

Associate Director for

Composite Su^uctures Research

Structural Engineering Department

Building Research Institute

Ministry of Construction

Tatehara 1, Tsukuba-city, Ibaraki 305

Japan

Digital acceleration time history records of the

following ground motions are used. Standard

ground motions are EL CENTRO 1940 NS

and TAFT 1952 EW. Local ground motions

likeTokyo-101 1952 NS. OSAKA-205 1963

EW, HACHINOHE 1968 NS and others. In

this group HACHINOHE 1968 NS and others are ground motions containing long period waves. Three kind of waves are to be used in the dynainic analysis as a rule.

The peak velocities for the earthquake waves

are decided according to the earthquake

activity at the site. For example, in Tokyo,

the peak velocity is greater than 25cm/sec for the ground motion level 1, and its equivalent

acceleration value is about 250cm/sec2; while

the peak velocity is greater than 50cm/sec for the ground motion level 2, and its equivalent

acceleration value is about SOOcm/sec^.

The dynamic elasto-plastic response analysis

is used for level I, considering the reduction

of the members stiffness due to flexural

cracking. For level 2, it is used considering

the reduction of the members stiffness due to flexural cracking and yielding. However

there are many ways to make the mathematical

model of restoring force characteristics and

damping.

These analysis are quite time consuming and

expensive to perform. The use of the dynamic

elasto-plastic response analysis is generally

justified only for the reinforced concrete

building with a height greater than 45m.

2. Static elasto-plastic frame analysis using

incremental analysis

The static analysis is the incremental analysis

using the stiffness matrix. 'Hie restoring force

characteristics of all members are difined

57
independently. The lateral forces are restoring force characteristics are defined

increased step by step, and the stiffness is independently,

assumed to be constant for each load

increment, which is called " moment stiffness The model where the member is divided into

". The moment stiffness is calculated for each many parts,

load increment taking into account every load- This is shown in Fig. 2.4. deflection relationship (skeleton curve). The

stress analysis is carried out using these data. The assumed flexural stiffness distribution

Then every story drift, the deformations and model.

stresses at the members ends are obtained for This is shown in Fig. 2.5. every loading step.

The model where two cantilevers are

2. 1 The distribution of lateral force connected.

This is shown in Fig. 2.6. The distribution of lateral forces is defined as

the modified Ai distribution. There are two ways of modelling a shear wall. In other cases, it is obtained from the One is to replace it witli a column which is a two - preliminary response analysis. dimensional member. The othdr is shown in Fig.

2.7.

2.2 Elasto-plastic model of a member

2.3 Mathematical models

There are two methods to make mathematical

model of beam and column^ whose flexural The mathematical model is a two-dimentional

characteristics are elasto-plastic. One is an frame model, and the degree-of-freedom in the

elastic member with plastic hinges at both lateral displacement of the same story is one. ends. And the other where the member is divided into many elements, whose restoring The stiffness is fixed at every step. All members

force characteristics are defined are divided into columns, beams and beamindependendy. column connections. The shear wall is replaced

by a column which is a 2-dimensional member.

There are three models to define plastic hinges

at both ends of a member. They are : the 2.4 Skeleton curve of a member

hybrid beam model, the model where both

ends are rigid plastic rotational springs and the Column, Beam : Tri-linear model considering

model where both ends are plastic hinge flexural cracking and

zones. yielding (Fig.2.8)

Connection panel : Bi-linear model considering

(1) The Hybrid beam model shear cracking (Fig.2.9)

In this model the member is composed of an Column (axial force) : Tri-linear model for tension

elastic part and a complete elasto-plastic part as and bi-linear model for

shown in Fig. 2.1. compression (fig.2.10)

(2) The model where both ends are rigid-plastic

rotational springs as Fing.2.2.

In this model, rigid-plastic rotational springs are used at both ends of an elastic member. There are two kinds of rigid-plastic rotational springs. One

shows the flexural plastic characteristics

(Degrading tri-linear). And the other shows an

additional deformation due to the slip of the

tension bars of a beam-column connection.

(3) The model where both ends are plastic hinge zones in Fig.2.3.

In this case the stiffness of the plastic hinge zone

is reduced after yielding.

There are three models to divide a member into

many elements with a certain length, whose

2.5 Analysis procedure

The static elasto-plastic frame analysis is carried out in the following way:

(a) Getting ready for the calculation.

(b) Change the matrix which is in terms of the

member end displacements to a matrix in term

of node displacements.

(c) Make the moment stiffness matrix of each

member.

(d) Make the moment stiffness matrix for every

story.

(e) When pan of the panels is neglected, the

moment stiffness matrix for every story has to be modified.

(0 Make the moment stiffness matrix of the

frame.

58
(g) Make the moment lateral stiffness matrix of the

frame (the frame is fixed at the base)

(h) Make the moment stiffness matrix of the

frame, considering rocking and sway at the base.

(i) The eigenvalue analysis at first circle.

(j) calculation of the lateral force increment.

(k) Stress analysis Calculation fo the

incremental lateral displacement.

(1) Calculation of the incremental displacements

and forces of the ends of the members.

(m) The displacement and force at the ends of the

member at one step are equal to the

displacement and force at the ends of the member at one step before plus the

incremental values.

(n) Deciding whether a member is elastic or plastic

according to its restoring force characteristics,

and then calculating the next moment

stiffness. (0) Go to (c) if the moment stiffness changed. Go

to (j) if the moment stiffness did not change.

2.6 Results

Flexural cracking, fiexural yielding and ductility

factor and forces at each member end are obtained

for every load increment step (Fig. 2. 11). Sheardrift relationship for every story is also obtained.

(Fig.2.12). The ultimate lateral shear strength for every story is obtained.

3. Dynamic elasto-plastic response analysis using

lumped mass model

3. 1 Fundamental mathematical model

(1) Lumped mass The distribution of weight is defined as a lumped

mass for every story level.

(2) Mathematical model

(a) Equivalent fiexural-shear model (Fig.3.1)

The shear deformation components of each story are separated from the results of the incremental

analysis. They are obtained to subtract the

fiexural deformation from the story drift. The

fiexural deformation components are obtained

from the axial deformations of the columns. The

equivalent fiexural and shear springs of each story are obtained from the flexural and shear

deformations, where the fiexural springs are assumed to be elastic.

(b) Rocking and sway model (Fig.3.2)

Ground or piles are replaced using an equivalent

rocking and spring or swaying spring. And the

superstructure is replaced usmg an equivalent

flexural and shear model.

(3) Damping

Kind: Viscous damping

Damping matrix : Proportional to the moment

stiffness of proportional to be

initial stiffness

Damping factor : 3% for the first mode

The damping factors for tiie other modes are proportional

to the frequencies.

(4) Numerical integration

Linear acceleration metliod is used

3.2 Restoring force characteristics

In case of seismic response analysis for reinforced

concrete structures, the model used to define the

restoring force characteristics is based on the

skeleton curves and hysteresis rules based on

geometrical rules.

These models can be : Masing type. Slip Type,

Degrading type and the combination of these

types. The degrading type is used mainly for

reinforced concrete structures.

3.3 Skeleton curve

(1) Definition of the skeleton curve

The shear deformation for each story is obtained

separating the story drift into shear deformation

and bending defonnation. The skeleton curve is obtained replacing the shear-deformation

relationship in the tri-linear model.

(2) Definition of ductility factor

A story ductility factor is defined as the second

yield point in the skeleton curve.

3.4 Hysteresis rule

(1) Tri-linear origin-oriented model

This is shown in Fig. 3.4.

(2) Tri-linear maximum value-oriented model

This is shown in Fig. 3.5.

59
(3) Degrading Tri-linear model

This is shown in Fig. 3.6.

The hysteresis rule for an amplitude smaller than

the yielding point is the same as the normal bi- linear mode. When the unloading starts from the

loop point which is over the yielding point, this point is assumed to be the new " yielding point ", the hysteresis rule is the same as in the bi-linear

model, but the stiffness is degrading.

(4) Takeda model

This is shown in Fing. 3.7.

(a) |5maxl < 5yi

The hysteresis rule for an amplitude smaller

than 5yi^ stiffness K is the same as the linear

model.

(b) 5yi ^ |6maxl 5y2

When 5 increases over ±5yi at the first time, or when 5 when S increases over the last maximum deformation point 6max. it goes to the

second stiffness K2. When the unloading starts from this line, it goes in the stiffness Kel or Ke2

to the horizontal axis. Then it goes to the opposite

maximum deformation point from the horizontal

axis. THen it goes to the opposite maximum

deformation point from the horizontal line.

(c) 5yJ <. ISmaxI

It is the same as (b). W^en the unloading starts towards zero, its stiffness is Kel or Ke2. Next,

when the load starts from zero, is goes to the

opposite maximum deformation point

4. Dynamic elasto-plastic response analysis using

frame model

This dynamic analysis is the incremental analysis

using stiffness matrix. The restoring force

characteristics of all members are defined

independently. The time is divided into

incremental times. The stiffness is assumed to be

constant for every time increment, which is called

" moment stiffness ". At each step increment, the

moment stiffness is calculated according to each

skeleton curve, and the dynamic analysis is carried out using it. In this way each story drift,

deformations and forces for every member are obtained successively.

This dynamic analysis requires many iterations

and therefore a larger computer capacity is necessary.

4. 1 Fundamental mathematical model

(1) Mathematical rrKxlel Two-dimensional frame models are shown in Fig.

A structure is divided into columns, beams, beamcolumn panels and shear walls. The restoring force characteristics are defined

independently.

(2) Member model

It is the same as the one used in the static elastoplastic analysis.

Usually, the elasto-plastic model for column and

beam is the model where both ends are rigidplastic rotational.

(3) Damping

Kind : Viscous damping

[c] = ^

[k]

Damping matrix : Proportional to the moment

stiffness or Proportional to the

initial stiffness

Damping factor : 3% for the first mode

The damping factors for the

other modes are proportional to the frequencies.

(4) Numerical integration

Linear acceleration method is used.

4.2 Restoring force characteristics of the members

(1) Skeleton curve

It is the same as in th ; section 2.4 " Skeleton

curve of a member "

(2) Hysteresis rule

It is the same as in the section 3.4 " Hysteresis

rule ". The case for columns in shown in Fig. 4.2.

4.3 Analysis procedure

60
It is the same as in the section 2.5 " Analysis

procedure ". Only that the item (j) is " incremental acceleration

of ground motion ", and (k) is " response

analysis"

Reference

1. Materials for Seismic Design in the Building

Engineering, AU, 1981

2. Recent Advancement of Structural Mechanics

in Building Engineering, AIJ, 1987

3. Study of Elasto-Plastic Frame Response

Characteristics and Seismic E)esign of High

Rise Reinforced Concrete Buildings, Edo

Hiroaki, 1987.

61
Cooplete elasw-plastic nOTber

FIcxural stiffness • (l-p)El.

/ ^ Plastic hlr«a —A (caiplete rleid-plastlc)

Elastic nember FlMural stiffness » p EI

Member

Complete elasto-plastlc resnber

Elastic nember

RIeld-plastIc rotational spring

A/.c; ^—^-—^

y»

Elastic pert

Elastic part RIgld-plastIc rotational spring

Fig. 2.1 Hybrid beam model Fig. 2.2 Model where both, ends are

rigid-plastic rotational springs

Plastic lone (degrading stiffness after yielding)

Elastlct part

It t I I I t ^ : : : : : ; ; ^

Divided eleinent : flexural stiffness ' £li

Fig. 2.3 Model where both ends are plastic ^'S- 2.4 Dividing element method

hinge zones

!/(£/).. l/(£/)a

Da

"^^^ Plastic rotational sprir^

(EI). » elastic flexural stiffness

Fig. 2.5 Assumed stiffness distribution Fig. 2.6 }todel where two cantilevers are method

connected

62
Rl^ld beam I

nonlinear / axial spring

Konllnear

'

axial spring

^Nonlinear

shear spring i

/^Nonl i'near rotational

•- spring

1

Fig. 2.7 Shear wall model Fig. 2.10 Load-deflection relationship

of the colwnn axial force

m p

m p c -

r P

k. : elastic stiffness (Included stab and re-ber)

a, • reduction factor of stiffneiss at yielding point

M, : flexural cracking moment M. : ultimate flexural force

m, : virtual pane! moment k, =Gc X ,V, (shear stiffness)

m,.= re X , V,

.V. = j . X J , X t , (volime) ^ = 0. 2

Fig. 2.8 Load-deflection relationship of pig. 2.9 Load-deflection relationship

beam or column of beam-column connection

63
5> ^ $ ^^-^3 ^>-ZS^ '^ ^ o^

^

I e» j o o a> oo i CO

I o I . CD M n ^ I CO

A) 'J^— —dtp—fc^-^—ew— ^) ee e&

3

-^p—

CO

COdIIaO

or

CO O

T3E

a

a

c

o

00»T

OOO'iS

f f f up — ^ 1

O •

O—

^?—tj?—^

O—K>—^—^—^

O—y7>— —

"fe—J&

^—4^—4—^^=^-4—4— —4—4

<!)

—^^^^^—fcSp— tfe wt^^^—t'

OII

or

O

64
65
Shear deformation (elasto-plastic)

El

.V

Flexural deformation (elastic) Lumped mass model

Fig. 3.1 Equivalent flexural -shear model

Fig. 3.2 Rock ing-sway model Fig. 3.3 Definition of the skeleton curve

66
Fig. 3.7 Takeda model

67
I rni_4

21 A A A

Issod^ssdseooiissdssQol

® (D (D (5) (D (D

Fig. 4. 1 Frame model

Fig. 4.2 Hysteresis rules for axial forces

68
U.S. Japan Cooperative Research in Precast Seismic Structural Systems (PRESSS)

Program — Status Report

by

Nigel Priestly' and H. S. Lew^

ABSTRACT

The Precast Seismic Structural Systems

(PRESSS) research program was initiated as the

fourth topic in the United States - Japan

coordinated program on large scale testing for seismic response. The first three topics were

reinforced concrete, steel and masonry. The

purpose of this coordinated research program is to develop recommendations for the seismic

design of buildings constructed of precast

concrete systems based on sound analytical

and experimental research.

This report describes the scope of the joint

program and the current projects being carried

out by U.S. researchers.

KEYWORDS:

buildings; concrete; earthquake; precast;

research; seismic; structures

1. INTRODUCTION

The fundamental objectives of the U.S.- PRESSS program are two fold:

To develop comprehensive and rational

design recommendations based on

fundamental and basic research data

which will emphasize the viability of

precast construction in the various

seismic zones.

To develop new materials, concepts

and technologies for precast

construction in the various seismic zones.

The PRESSS program sets out to achieve these

objectives by a process of examination and

assimilation of existing research and design

information, by development and evaluation of

new PRESSS concepts in close cooperation

with industry, by carefully planned analytical

and experimental research, and by formulation

of the results within a framework of design

recommendations based on state-of-the-art

design philosophy.

The program considers only the seismic

behavior of building structures where precast

construction forms the essence of the

structural system. Current precast seismic

structural systems can be divided into two

categories: precast frame buildings and

precast wall panel construction. Both are Included in the U.S. -PRESSS program.

2. STRUCTURE OF THE U.S. -PRESSS

PROGRAM

The U.S. -PRESSS program was conceived in three phases. Phase I, which is funded by the

U.S. National Science Foundation for a threeyear period ending in 1993, is focused on

identifying and evaluating the most promising

structural concepts for precast concrete

building systems in seismically active regions.

Phase II involves detailed experimental studies

of components and subassemblages of precast

systems selected during Phase 1 for more

detailed investigation.

Phase III involves one or more multi-story full- size "superassemblage" experiments and the

finalization of the seismic design

recommendations for incorporation into the

model building codes.

University of California, San Diego, La Jolla,

California 92093

National Institute of Standards and

Technology, Gaithersburg, Maryland 20899

69
Phase I Program:

- Concept Development

The project is focused on the

development and evaluation of new

concepts for precast concrete

construction in UBC seismic zones 2,

3, and 4, particularly for low and midrise buildings.

- Connection Classification and Modeling

The project involves a comprehensive

review of existing test data and the

development of a classification system

for PRESSS connections based on type

and function.

- Analytical Platform Development

The widely used DRAIN-2D inelastic

time-history analysis program for seismic response is being expanded to three dimensions. In addition, new

elements are being developed to model

the special characteristics of precast

members including joint deformations

resulting from slip and ductility.

- Preliminary Design Recommendations

Current seismic design codes for precast concrete structures are being

reviewed and preliminary prescriptive

design requirements for "strong

connection" precast construction are being developed. In addition, a framework for the final design

recommendations, based on structural

reliability theory, will be provided.

These provisions will include

recommendations for ductile

connections design.

- Coordination

This project provides the umbrella

under which the four above-mentioned

projected are interrelated, and also

provides coordination with the JapanPRESSS research team and the

research program at the National

Institute of Standards and technology.

Phase II Program:

- Evaluation of Ductile connections for Frame

PRESSS

This project involves the evaluation of

the behavior of connecting elements,

beam-column subassemblages, and

multi-bay subassemblages.

- Evaluation of Wall Panel Building for Moderate

Seismic Zones

This project includes the evaluation of

optimal range of connection strength

and ductility for a six story building,

strength and deformation

characteristics required for connection

elements, and tests on horizontal panel

connections.

3. U.S. -JAPAN COORDINATION

The second meeting of the Joint Technical

Coordinating Committee on Precast Seismic

Structural Systems was held in Tsukuba,

Japan, October 31 - November 2, 1991. The

meeting was attended by 21 researchers and 9

observers from Japan, 4 researchers and 3

observers from the U.S., and one observer from

New Zealand.

The need for, and importance of,

continuing experimental and analytical

research on PRESSS was reaffirmed.

Analysis of Japanese and U.S. frame

buildings have indicated that

satisfactory seismic response can be

expected from precast frame buildings

designed to meet recommendations of

both countries.

The U.S. side will carry out dynamic

analysis of the test-design building.

Both sides will continue to develop and

classify new connection concepts.

The third joint meeting is scheduled to be held in San Diego, October 25 - November 2, 1992.

70
Seismic Response Control of Highway Bridges by Variable Damper

by

Kazuhiko Kawashima^ Shigeki Unjoh^ and Hideyuki Shimizu^

ABSTRACT

This paper presents analytical and

experimental studies on the dynamic

characteristics of a variable damper developed

at the Public Works Research Institute.

Effectiveness of the variable damper for seismic

response control of highway bridges Is analytically Investigated. A small-size model of

the variable damper Is developed and the dynamic

characteristics through the dynamic loading

tests Is also presented.

KEY WORDS

Seismic Response Control, Highway Bridges,

Variable Damper, Small-Size Model, Dynamic

Loading Tests

1. INTRODUCTION

The variable damper"^' Is a viscous damper In which the viscous damping force Is variable

depending on the response of highway bridges as shown In Fig.l. That Is:

1) The damping coefficient of the variable

damper Is taken large during a small deck

vibration so that the damper has the same

function as a fixed bearing support against the

braking load of vehicles. It is movable against

actions with low-velocity such as an elongation

of deck by temperature change.

2) When the amplitude of deck vibration becomes

larger up to a certain level during an earthquake, the damping coefficient Is decreased

so that the energy dissipation be optimum and the

Inertia force of a superstructure to substructures be decreased appropriately.

3) Furthermore, when the vibration amplitude of

the deck becomes excessive, the damping

coefficient is gradually Increased In order to suppress the excessive amplitude. Therefore, the

damper has a function as a stopper with a shock

absorber. This Is to prevent the Impact response

by the sudden operation of the stopper*".

Therefore, the variable damper has the

advantages of an usual viscous damper

stopper^' , a passive energy disslpator and a stopper with a shock absorber function.

The concept of the variable damper, in which

the damping characteristics can be variable

dependent on the situation, has been effectively

adopted In a field of mechanical engineering, In particular In a suspension system for an aircraft

and an automobile. An active suspension to Improve driving comfortablllty and stability of

the automobile Is one of the examples^' ^'

.

However, It Is a new Idea to apply the variable

damper for highway bridges based on the concept

of distributing the Inertia force of a superstructure to substructures.

Although various types of the variable

damper can be made, the simplest one may be as shown in Fig. 2. The basic component is an cylinder-type viscous damper. A by-pass pipe Is Installed between the cylinder-cells divided by

the piston. The damping characteristics of the

damper can be controlled by varying the amount

of flow of viscous material which passes through

the by-pass. Since various technology for

controlling such material flow has been

developed, the variable damper Is most promising

and near to the practical use. Furthermore, the

external energy required for the operation Is significantly smaller In the variable damper than

In active control devlces'°*. This paper presents analytical and

experimental studies on the dynamic

characteristics of a variable damper developed.

The effectiveness of the variable damper for

seismic response control of highway bridges Is analytically investigated. A small-size model of

the variable damper is developed and the dynamic

characteristics through the dynamic loading

tests is also presented.

2. SEISMIC RESPONSE ANALYSES OF HIGHWAY BRIDGES

Wrm VARIABLE DAMPERS

2.1 Seismic Response Analysis Method of HultlDcgree-of-Freedom System with Variable

Dampers

The equations of motion for a linear

multl-degree-of-freedom system with the variable

dampers may be written as M x'+ (C + C:i) X + K X = - M Xo (1)

In which M, C and K represent mass, damping and

stiffness matrices of the system, respectively, x and xo denote displacement vector of the system

and an earthquake ground motion, respectively. Cv denotes a damping matrix of the variable

dampers and Is assumed as

1) Head of Earthquake Engineering Division,

Earthquake Disaster Prevention Department,

Public Works Research Institute, Ministry of

Construction

2) Research Engineer, ditto

3) First Research Division, Fukushlma

Construction Office, Tohoku Regional

Construction Bureau, Ministry of Construction

(Formerly Assistant Research Engineer of the

Earthquake Engineering Division, Public Works

Research Institute)

71
nv

Cv = X cu'^ (2) k=l

where

eM-= [.^ (3)

In which cu'' and nv represent damping matrix of

the variable damper, which Is Installed between

the I-th node and the J-th node, and a number of

the variable dampers Installed, respectively. V Is the damping coefficient of an each variable

damper and Is given as a function of relative

displacement and relative velocity between the

nodes where the variable dampers are Installed. Although there are various methods to prescribe the damping matrix C, It Is assumed

here to be written as Eq.(4) assuming that the

damping matrix C can be diagonallzed by the

modal matrix.

C= (^)-' dlag[2hKW h] C*."^)"' (4)

In which, denotes a modal matrix of the system

and dIag[2hK w k] denotes a diagonal matrix with

elements of 2hK« k (k=l,2,3, • • • ; mode number).

Since Cv Is time-varying, Eq.d) has to be

solved by a direct Integration method. According

to the above analytical method, a computer

program "VDAM" which can analyze earthquake

response of a multl-degree-of-freedom system

with the variable dampers was developed.

2.2 nighway Bridge Analyzed and Analytical

Conditions

In order to Investigate the effectiveness of

the variable damper, a simple span girder

bridge, as shown In Flg.3, with a span length of

30m Is analyzed. The response In longitudinal

direction Is considered. Although the variable

damper has better applicability for multi-span

continuous girder bridges, the simple span girder

bridge Is taken as an analytical model here for

simplicity. The superstructure of the model Is supported by elastic Isolators In longitudinal

direction and the variable dampers are assumed

to be Installed between the superstructure and

the pier top.

The model can be assumed as a slngle-degree-of-freedom oscillator If It Is assumed that the deformation of the

superstructure Is negligibly small and that the

piers are Idealized as a linear spring. Therefore,

the damping coefficient of the variable damper,

V, can be written as

V = yf ffld kp hv (5)

In which, md, kp and hv represent mass of the

superstructure, a combined spring stiffness of

elastic Isolators and piers, and required

damping ratio for the variable damper,

respectively. The weight of the superstructure

and spring stiffness of the Isolator are assumed

as 241.5tf and 640tf/m. respectively. The flexural

stiffness of piers Is assumed so that the

fundamental natural period of the model be O.Ssec when the both bearing supports are hinged. The

stiffness of Isolators result In the fundamental

natural period of 1 sec for the model with elastic

Isolators. Damping ratio of the variable damper hv

defined by Eq.(5) can be variable dependent on both relative velocity and relative displacement

between the superstructure and the pier top.

Only analytical cases In which the damping ratio

Is varied with the relative deck displacement are shown here for representing the effectiveness of

the variable damper although the

velocity-dependent variable damper was also

studied.

Six cases In total are analysed as shown In FIg.4. In Case 1, the variable damper Is not

Installed. I.e., no control. In Case 2, the damping

coefficient of the damper hi Is assumed

Independent on the relative deck displacement.

The damping ratio hi Is varied as a parameter to be Investigated. Cases 3 and 4 are cases for

Investigating the effectiveness of the stopper

with a shock absorber for preventing excessive

deck displacement. The damping ratio increased

from hi to ha dependent on relative deck

displacement. In Case 3, a weak stopper. I.e., a stopper with a shock absorber Is assumed and the

damping ratio hs is assumed to be 3. In Case 4, the strong stopper Is assumed and the damping

ratio ha Is assumed to be 500. The displacement at which the stopper Initiates to work Is assumed as

2/3 of the peak relative deck displacement du of Case 2. This displacement Is defined hereafter as

a design displacement of the variable damper. The

design displacement of the variable damper du Is 7.7cm as will be shown In Case 2. Case 5 Is a case

for Investigating the effectiveness of the

stopper for suppressing small deck displacement

caused by vehicle loads. The damping ratio Is Increased from hi to hs In the range of the

relative displacement from -l/6du to 1/Gd.j. Case

6 Is a combination of Cases 3, 4 and 5. An acceleration ground motion shown In Flg.5

is used as an Input acceleration. The

acceleration, recorded on the ground around the

Tsugaru Bridge during the NIhonkal-Chubu

Earthquake of 1983, was modified so that the

response spectrum matches with the target

spectrum. The acceleration spectrum, which Is specified in "Part V Seismic Design" of the

"Design Specifications for Highway Bridges", for

tlie check of bearing capacity of reinforced concrete piers on soft soil site (ground

condition HI), Is assumed as the target spectra.

When the variable damper Is not provided, the

modal damping ratio of the model bridge Is assumed to be 2% of critical In the above cases.

72
2.3 Effectiveness of the Variable Dampers for

Seismic Response Control of Illghway Bridges

Flg.S shows how the peak deck response and

force developed at the pier vary In accordance

with damping ratio hi for constant damping (Case

2). The peak deck displacement monotonously

decreases with increasing the damping ratio hi and become almost stable for the damping ratio

equal or greater than 2. The peak deck

acceleration takes minimum at the damping ratio

hi of about 0.5. This Is because the function as a damper stopper becomes predominant rather than an energy dlssipator. The shear force and

bending moment at the pier bottom with Increase of the damping ratio hi in a similar manner with

the peak deck acceleration.

On the other hand, as the damping ratio Increases the maximum damping force developed in the variable damper Increases, while stroke of

the piston decreases. Total energy dissipated

during the excitation takes maximum value at the

damping ratio of about 0.2. Based on the analysis, the most appropriate

damping ratio hi Is considered as 0.5, which makes

peak deck displacement and maximum bending

moment at the pier bottom decease to 31% and 37%

of those with no control, respectively. The

damping force and peak stroke of the piston

required fro this control are 34.7tf x 2 = 69.4tf and 7.7cm, respectively.

Assuming hi as 0.5, the effectiveness of the

stopper with a shock absorber was analysed.

Flg.7 compares the deck displacement and

acceleration as well as damping force of the

variable damper for Cases 2 - 4. In Case 4, the

pulse-type response developed by strong

operation of the stoppers Is seen In the damping

force of the variable damper. FIg.8 compares the

peak response and peak damping force for Case 2

- 4. Comparing Case 3 with Case 2, the peak deck

response slightly decreases while maximum forces

at the pier bottom increase only 7%. The maximum

damping force of Case 3 developed In the variable

damper becomes 1.73 times that of Case 2, and the

peak stroke decreases by about 10%. The decrease

of the stroke of the variable damper means that

the function of the stopper operates effectively. On the other hand, when the stopper operates

strongly (Case 4) the responses become larger,

and deck acceleration and forces at the pier

bottom becomes about 2 times those in Case 3. Although the stroke of the variable damper

decrease to the displacement at which the strong

stopper operates, the peak damping force

Increase to about 4.3 times that in Case 3. Flg.9 shows the effect of the stopper at the

small deck displacement. The damping ratio of the

damper, hi, Is assumed to be 0.5 and hs is varied

as a parameter. Although the peak deck

displacement decrease monotonously as the

damping ratio hs of the variable damper Increases up to 8, It Increases for the damping ratio hs

greater than 8. This Is because that the function

as a damper stopper becomes more predominant

rather tlian an energy dlssipator when the

damping ratio hs becomes larger. The peak deck

acceleration takes a minimum value when the

damping ratio hs is about 1, and It Increases monotonously for the damping ratio hs greater

than 1. The forces at the pier bottom vary with

hs In a similar manner with the deck acceleration. On the other hand, as the damping ratio ha

increases the damping force of the variable

damper increases and the stroke and the total

energy absorbed decrease. Therefore, damper at the small deck displacement which is for

suppressing the deck motion caused by the

vehicle loads has contrary effects to make

seismic response smaller. Therefore, It Is necessary to determine damping ratio ha In consideration with the function of bridges for

normal loads. Table 1 summarizes the peak responses in Case 1 to 6. If the variable damper with hi = 0.5 and ha = ha = 3 Is provided, peak deck

displacement and peak deck acceleration decrease

25% and 43% of those with no control,

respectively. The maximum bending moment at the

pier bottom decreases to 46% of that. The

required damping force and peak stroke of the

variable damper are 62.Gtf x 2 = 125tf and 5.78cm,

respectively. The variable damper with this specification can be designed within the current

scope of damper technology.

3. DEVELOPMENT OF A SMALL-SIZE MODEL OF TIIE VARL\DLE DAMPER

3.1 Outline of A Small-Size Model of the Variable

Damper

FIg.lO shows a design plan of the variable

damper developed. The model Is designed so that

the damping force be variable in the range of

20kgf to 200kgf, maximum relative displacement of

damper piston be within + 2.5cm. The control of

the damping ratio is arbitrarily made by a personal computer dependent on relative

displacement and velocity developed between

deck and substructures.

The piston-cyllnder has the total length of

394mm, the pressure area of the piston Is 12.56cm^ and the stroke of the piston Is ± 2.5cm. A steel

pipe by-pass is Installed between the

cylinder-cells divided by the piston and the oil tank tlirough two servo-valves. The servo-valve

Is called as a D.C. proportional magnetic valve,

and the width of the valve Is adjusted by the

magnetic force of the D.C. proportional solenoid.

The return spring is Installed at the spool as shown in Fig.ll. The differential transformer is installed at the servo-valve for the position

detection of the spool so that the accurate

control can be made. The input voltage from a servo amplifier to the servo valve Is set in the

range of 0 to 5V. The valve Is closed at the input

voltage of OV and the valve Is fully opened at

73
5V. The viscous material used for the variable

damper Is an usual oil for a dynamic loading

actuator.

Fig. 12 shows the flow of the viscous material

In the variable damper. The damper piston can move In the left and right directions

accompanying with the deck response. Wlien the

piston moves In the left side the viscous material

flows from the left cell of the cylinder to the

oil tank through the servo-valve J, and from the

tank to the right cell through the check valve 2 which can pass the oil only in one direction. At

this time the servo-valve 2 Is closed. When the

piston moves In the opposite direction the oil passes through from cylinder to the tank through

the servo-valve 2 and check valve 1. Therefore,

the damping characteristics of the variable

damper is controlled by the servo valve 1 when

the piston moves in the left side, and by the

servo-valve 2 when the piston moves in the right

side.

3.2 Dynamic Loading Tests of the Small-Size

Model of the Variable Damper

The variable damper can vary the damping

characteristics depending on the displacement

and/or velocity of the structures. Therefore, In order to provide the variable damper with the

required damping characteristics It is necessary

to know the basic dynamic characteristics of the

variable damper. The damping characteristics of

the variable damper is controlled by the amount

of opening of the servo-valve and it Is controlled by the Input voltage to the

servo-valve. Therefore, to know the

characteristics of the variable damper

developed, the relation between the Input

voltage to the servo-valve and damping

characteristics of the variable damper Is essential.

Fig. 13 and Photo 1 show the set-up of the

dynamic loading tests. The opening of the servo

valve is kept to be constant under the condition

that the input voltage Is set to be a constant

value, i.e., the damping ratio of the variable

damper is set to be constant, the load is applied

to the model by a dynamic actuator under the

displacement control. The input voltage to the servo amplifier, reaction force and stroke of the

piston are measured. The loading Is harmonic and

10 cycle were repeated for each loading

displacement.

Table 2 shows the experimental cases. Case 1 is to measure the friction force of the damper

piston and static loading was made with loading

frequency of 0.05Hz and loading displacement of

20mm. The Input voltage of the variable damper Is assumed to be 5V so as to open the valve fully.

Case 2 is to obtain the input voltage to the

servo-valve vs. damping force of the variable

damper relation. Loading frequency, loading

displacement and input voltage are varied.

3.3 Performance of the Small-Size Model of the

Variable Damper

(1) Friction force of the damper-piston

Fig. 14 shows a hysteresis loop of the

load-displacement relation when the damper Is loaded with the loading frequency of 0.05Hz and

loading displacement of 20mm to obtain the

friction force of the damper-piston. The

load-displacement relation shows almost

rectangular shape. This represents the typical

type energy dissipation. The hysteresis Is stable for a number of loading cycles. The

average friction force was obtained as about

7kgf.

(2) nysteresis Loops of the Load-displacement

Relation

Fig. 15 shows hysteresis loops of the

load-displacement relation. In the loading

displacement of 2mm, the time-lag was found in the peak damping force for the input voltage

lower than 1.5V. This is developed due to compressibility and friction between the pipes

and the viscous material. To compensate the

effect It Is required to make the pressure area greater for obtaining enough flow even if the

displacement Is small. In the loading displacement

of 6mm and 10mm, the hysteresis loop seems as expected.

(3) The relation between the Input voltage and

damping force

Fig. 16 shows the relation between the loading

displacement and damping force dependent on the

input voltage. The empirical equations obtained

by a least square method are also shown In Fig. IS. The relation between the damping force

and loading velocity shows a function of the

power of the loading velocity as an usual

cyilnder-type viscous damper. Therefore, the

empirical equation is assumed as

F = Fo + C(V) • v"'"^ (6)

where

F : damping force (kgf) Fo : friction force of the damper-piston

(7kgf)

C(V) : damping coefficient (function of the

Input voltage V) V : velocity of the damper-piston

n(V) : power coefficient of velocity terra (function of the Input voltage V)

In Eq.(G), damping coefficient C(V), and power

coefficient of velocity term n(V). are unknown

factors. They are computed by a least square

method based on the tests results. The damping

coefficient C(V) thus obtained are almost constant in the range of the input voltage of 0 to 5V, and averaged value of C(V) is 0.8. The

relation between the Input voltage and the power

coefficient of the velocity term obtained is shown In Fig. 17. The power coefficient Increases

with decrease of the input voltage. When the

74
Input voltage Is larger than 2.5V, the power

coefficient becomes almost constant (n(V) = 1.4), and when the Input voltage Is lower than 2.5V the

value takes larger. The relation obtained by a linear approximation Is shown In Fig. 17.

5. CONCLUDING REMARKS

The effectiveness of the variable damper

developed for seismic response control of

highway bridges was analytically Investigated.

The small-size model of the variable damper was developed and the performance was studied

through the dynamic loading tests. According to the above Investigations the following

conclusions may be deduced:

1) In order to provide functions as damper

stopper, energy dlsslpator and stopper with

shock absorber for the variable damper, the

damping ratio of hi = 0.5 and hs = ha = 3 Is the most favorable.

2) By providing the variable damper with

damping ratio shown In 1), the peak deck

displacement and the peak deck acceleration of

the model highway bridge were reduced to 25* and

43%, respectively. The peak bending moment

decreased to 46%. The required peak damping

force and peak stroke of the variable damper

were G2.6tf x 2=125tf and 5.78cm, respectively. The

peak damping force is equivalent with 52% of the

deck weight. These capability of the damper can be designed within the current scope of damper

technology.

3) Based on the dynamic loading tests of the

small-size model of the variable damper, the

load-displacement relation was obtained as expected. A proto-type model of the variable

damper Is now under development for further

Improving the performance.

REFERENCES

DTaguchl, J., Iwasaki, T.. Adachl, Y., Sasaki, Y. and Kaweishlma, K. : U.S.-Japan Cooperative

Research Program on Hybrid Control of Seismic

Response of Bridge Structures, 22nd Joint

Meeting, U.S.-Japan Panel on Wind and Seismic

Effects, UJNR, Galthersburg, Maryland, USA,

May, 1990

2) Kawashlma, K. , Unjoh, S., Nagashlma, H. and

Shlmlzu, H. : Current Research Efforts In Japan for Passive and Active Control of

Highway Bridges Against Earthquakes, 23rd

Joint Meeting, U.S.-Japan Panel on Wind and

Seismic Effects, UJNR. Tsukuba, May, 1991

3) Liu, S., C, Lagorlo, H.. J. and Chong. K., P. : Status of U.S. Research on Structural Control

Systems, 23rd Joint Meeting, U.S.-Japan Panel on Wind and Seismic Effects, UJNR, Tsukuba,

May, 1991

4) Kawashlma, K. , Hasegawa, K. and Nagashlma, H. : Experiment and Analysis on Seismic Response

of Menshln Bridges, Proc. 1st U.S.-Japan

Workshop on Earthquake Protective Systems of

Highway Bridges, National Center for

Earthquake Engineering Research, Buffalo,

N.Y., U.S.A.. September 1991

5) Fukuoka, S. : Design of Multi-span Continuous

Girder Bridge using Viscous Shear Stopper,

Bridge, pp. 22-29, February, 1980 (In Japanese)

6) Izekl. H. : Viscous Shear Stopper. Proposal to the Multi-Continuous Girder Bridge, Bridge,

pp. 30-33, February, 1980 (In Japanese)

7) Matsumura. S., Fukuoka, S, Mlzumoto, Y. and

Nakata. T. : Seismic Design of Multi-Span

Continuous Girder Bridge using Damper

Stopper, Bridge and Foundation, pp.31-37, May.

1982 (In Japanese)

8) MlUlken, Jr.. W., F. : Lotus Active Suspension

System, Proceedings of 11th International

Conference on Experimental Safety Vehicles,

pp.467-475, Washington D.C.. May, 1987

9) Sugawara, F. : Electronically Controlled

Shock Absorber System Used as a Road Sensor

which Utilizes Super Sonic Waves, Proceedings

of Society of Automotive Engineers,

pp.6. 15-6.25. 1986

10) Kawashlma, K., UnJoh. S. and Shlmlzu. H. : Optimum Seismic Response Control of Highway

Bridges by Active Mass Damper, Civil Engineering Journal, Vol.33, No. 9, pp. 61-68,

November, 1991 (In Japanese)

11) Kawashlma, K., UnJoh. S. and Shlmlzu. H. : Earthquake Response Control of Highway

Bridges by Variable Damper. Proceedings of

Colloquium on Control of Structures. Japan

Society of Civil Engineers. Part. B. pp. 221-224. July. 1991 (In Japanese)

12) Feng. Q. and Shinozuka. M. : Use of A Variable

Damper for Hybrid Control of Bridge Response

Under Earthquake, U.S. National Workshop on

Structural Control Research. University of

Southern California. CA.. U.S.A.. October, 1990

13) Shinozuka. M., Feng. Q. and Kawamura. S. : New

Perspectives for Sliding Base Isolation

Systems. Proceedings of International

Workshop on Recent Developments In Base-Isolation Techniques for Buildings.

Tokyo, Japan. 27-30 April, 1992

14) Kawashlma. K., UnJoh, S. and Shlmlzu, H. : Earthquake Response Control of Highway

Bridges by Variable Damper, Transactions of

the Japan National Ssonposlum on Active

Structural Response Control, Japan Science

Council, pp. 311-317, March, 1992 (In Japanese)

15) Kawashlma, K., UnJoh, S. and Shlmlzu, H. : Experimental Study on Dynamic Characteristics

of Variable Damper, Transactions of the Japan

National Symposium on Active Structural

Response Control. Japan Science Council,

pp. 311-317, March, 1992 (In Japanese)

75
Table 1 Peak Response of the Bridge and the Variable Damper

Oecl[ Pier Vorlotle Dompif

Anolyiicol Coses

Displacement

(cml

Velocity

lcm/51

Aceelerotion

(cm/S'l

Shear force

(til

Bendini) Moment

(tl ml

Oompinq force

(til

Retoti>e Displocement (cm)

Relotlie

Velocity (cm/SI

Tolol Enenjy

Itl-ffl)

1 He Control 33.04 189.0 1300 166.9 3 297

2 Constont Domoina h:0.5 10.28 60. 2 478 63.4 1 223 34.7 7.67 44.9 93.9

3

Disploceoienl Oepcnilent Ooi«pii>4 with Weol Stopper for Eicessive

Ded[ Response Displocement 1

h J

= 3 M=0.5/' 9.94 54.2 538 72.0 1 395 60.2 6.90 68.

1

94.4

4

Dnptocement Dependent Dompln^ _

wim SlroriQ Stopper (or Eicesslve | Deck Response Displocement 1

hi-iOO

111=0,5] 10.90 76.4 965 123.9 2 382 15 1.3 5.54 112.3 94.9

i

Dlsptocemenl Dependent Domplnq

«ith Slopwr tor Smoll /

Deeli Response Disptocenient '

la=3

\t»--0.i 8.50 51.3 567 80. 1 1540 64.7 6.34 55.8 74.5

Displocement Dependent Dompinq

(Comblnotlon of Cose 3 \^ /

nnit r.nii 41

M = 3 8.38 47.5 555 81.2 1 528 62.6 5.78 60.7 77.4

Table 2 Experlaent for Small-Size model of Variable Damper

Cose Objective

Loading

Frequency (Hz)

Loading

Displacement (mm)

Input

Voltoge(V)

I Measurement of Friction Force of Domper Piston

0.05 20 5

2

Relation between

Damping Force and

Input Voltage

1.2.3.4 2.6.10.15 0-5

76
-VISCOUS DAMPER STOPPER

''WoPPER FOR PREVENTING

EXCESSIVE RESPONSE

DECK RESPONSE RELATIVE TO PIER CREST

Flg.l Basic Concept of Variable Danpers

CONTROLLER

SUPER STRUCTURE

CYLINDER

SUBSTRUCTURE SERVO VALVE

Fig. 2 Variable Damper

7777777" 777^-

Flff.3 Highway Bridge Analyzed and

Analytical Idealization

77
DAMPING COEFFICIENT C n

RELATIVE DISPLACEMENT

(a) Cast t ' No Control

DAMPING COEFFICIENT C

Ct

RELATIVE DISPLACEMENT

(b) Cose 2 : Constant Damping

DAMPING COEFFICIENT C

Ci hi = 0.5

-I -f

RELATIVE DISPLACEMENT («du)

(c) Cose 3 : Displocement Dependent Damping with Weak Stopper

(or Excessive Deck Response Displacement (du = 7.67cm)

OAMRNG COEFFICIEN C

-I 0 I

RELATIVE DISPLACEMENT (idu)

(d) Cose 4 : Displacement Dependent Damping with Strong Stopper

for Excessive Deck Response Displacement (du = 7,67dn) DAMPING COEFFICIENT C

hi = 0.5

RELATIVE DISPLACEMENT txdu)

(•) Cass 5 : Displocement Dependent Damping with Stopper for

Smoll Deck Response Displocement (du=7.67eml

DAMPING COEFFICIENT C C2^^_h2=3

hi = 0.5

I I

C3 h3 = 3

-I --r

2 -J. 0 ^ 3 6 " 6 3 RELATIVE DISPLACEMENT (xdul

(f) Cose 6 : Displocement Dependent Damping

(Combination of Cose 3 ond Cose 5Mdu = 6.34cml Fig. 4 Analytical Cases

78
500 r

Fig. 5 Input Acceleration

40

PEAK DECK ACCELERATION

PEAK DECK DISPLACEMENT

-I I I

1 DAMPING RATIO

(o) DECK RESPONSE

2000

1500 ;7

200

1000 2

500

CO —

MAXIMUM SHEAR FORCE

AT PIER BOTTOM

4000

3000

2000

1000

o

12 3 4 DAMPING RATIO

(b) FORCES AT PIER BOTTOM

200 40

2 3 DAMPING RATIO

{ c ) RESPONSE OF VARIABLE DAMPER Fig. 6 Pealc Response of the Bridge and

the Variable Damper (Case 2)

79
Cose 2

0 5 10 15 20 25 30 35 40 45 TIME (sac)

(a) Deck Displacement

i_ 1000

s'ti 500 - ll 0 I

5-500 < -1000 g 1000 500 || 0 a I

-500 < -1000 g 1000 500 81 -500 9 10000

Case 2

Cose 3

Cose 4

10 15 20 25 30 35 40 45

TIME Isecl

g 50 e_ 25 g= 0 a 50 ^ 50 e_ 25 u>S 0 E -25 1 -50

-75

150

a.

<:

Cose 2

1 Case 3

Cose 4

10 15 20 25 30 35 40 45 50 TIME (sec)

(c) Domping Force

(b) Deck Acceleration Fig. 7 Computed Response of the Bridge and Variable Damper

20

1

15

10

2000

h3=3

ih3--506

PEAK DISPLACEMENT

/ —PEAK ACCELERATION

1500 -

E

500

CONSTANT WEAK STRONG

DAMPING STOPPER STOPPER

(o) DECK RESPONSE

200

1

150

1000 2 2 100

50 -

MAXIMUM SHEAR f/ FORCE

MAXIMUM BENDING MOMENT

4000

3000 1

200

150

- 2000 = eSlOO

e> Q. z

3 -E LU

1000 S 50 -

\__y "^'^

MAXIMUM

STROKE

TOTAL ENER6Y/V

DISSIPATED

MAXIMUM

OANPING FORCE

10

CONSTANT WEAK STRONG

DAMPING STOPPER STOPPER

(b) FORCES AT PIER BOTTOM

CONSTANT WEAK STRONG

DAMPING STOPPER STOPPER

(c) RESPONSE OF

VARIABLE DAMPER

Fig. 8 Effect of Weak Stopper (Case 3) and Strong Stopper (Case 4)

for Excessive Deck Response Displacement

80
20

15

10 PEAK DECK ACCELERATION

PEAK DECK RESPONSE

J I I l_

1000

800

in 600 J

400 tto

200 **

4 6 8 DAMPING RATIO hj

10

200

150

ocroUoe

<iUJX

in

too

50

1 1 1 r 4000

3000 E

MAXIMUM SHEAR FORCE

MAXIMUM BENDING MOMENT

J I I L

2000

1000

IDZ

o

2 4 6 8 DAMPING RATIO h2

10

lo) DECK RESPONSE (b) FORCES AT PIER BOTTOM

200

o 150

LU

100 Ode

% 50

'

\ 1——1

PEAK STROKE >

—1 =^\hi-0.5

TOTAL ENERGY

DISSIPATED

MAXIMUM DAMPING FORCE

1 1

'

10

J

CO<

o

Fig.

0 2 4 6 8 10 DAMPING RATIO h2

(c ) RESPONSE OF VARIABLE DAMPER

9 Effect of Stopper for Snail Deck

Response Displacement (Case 5)

81
(480)

SERVO ^

VALVE S CHECK VALVE

Fig. 10 Small-Size Model of Variable Damper

-O.C. PROPORTIONAL SOLENOID (—DIFFERENTIAL ARMATURE TRANSFORMER

POSITION DETECTION

SOLENOID i CURRENT I CONTROLLER Fig. 11 Structure of Servo Valve

TANK

! INPUT VOLTAGE

(0-'5V)

SERVO VALVE I SERVO VALVE 2

1 PISTON

CYLINDER VISCOUS MATERIAL Fig. 12 Flow of Viscous Material

LOAD SENSOR

CONTROLLER

ACTUATOR

DISPLACEMENT SENSOR

Fig. 13 Set Up of Loading Test

82
400 300 _ 200

•S. 100 ~ 0 I

-100

-* -200 -300

-400

-I 1 1 1 1

1 1 r

-J I L. -2.5 -2 -1.5 -I -0.5 0 0.5 I DISPLACEMENT (cm)

1.5 2 2.5

Fig. 14 Hysteresis Loops of Load-Displacement Relation

(Dlsplacenent=20iiUB, Frequency=0. OSHz)

500 400

300 _ 200

!• 1000 § - 100 3 -200

-300

-400

-500

500

400

300 200

1000

- 100 -200 300

-400

-500

1 1 11 1 1

I.I7V 1.4V . Y^I.2V

i.5v\y-jrll.3V

1 1 1 2 -1.5 -I -05 0 0.5 1 DISPLACEMENT Icm)

(a) Displocement 2mm

-0.5 0 05 1 DISPLACEMENT (cm) 2 -1.5 -I

(b) Displocement 6mm

500

400

300 200

1000 -100 200

-300 400

-500

1.5

1.5

r

1 1

1.6 V

1 1 1

1.35V ; 1.8V 2V I^;^/i-5v

^TMjp 1 1

1 1 1 1 J 1

—r— T ~i

1.8V

r t .

1.5 V

2.0V \ / i.6V 5.0V^!<V -~></l.7V

1 1 1 1 1 1 -1.5 -I -0.5 0 0.5 I 1.5 DISPLACEMENT (cm)

(c) Displocement 10 mm

Fig. 15 Hysteresis Loops of Load-Displacement Relation

(Frequency=lHz)

83
400

300

UJ

?5 200

z

< 100

0 5 10 15 20 25 30

VELOCITY (cm/s)

Fig. 16 Danpln? Force vs. Peak Velocity

Relation

2 2

LU

UJO^ 1

0

n(V) = 7.78-3.0V

n(V) = 3.92-V

n(V)=l.49-0.03V

12 3 4 5

INPUT VOLTAGE (V)

Fig. 17 Power Coefficient n(V) vs. Input Voltage Relation

Photo 1 Experimental Set Up

84
Theme IV

Earthquake Engineering

Seismic Design Considerations for Jordanelle Dam

by

John A. Wilson', Perry J. Hensley', William O. Engemoen', and Francis G. McLean'

ABSTRACT

When completed in 1993, Jordanelle Dam will

be a compacted, zoned earthfill dam having a sloping core. It is being constructed in northern

Utah, USA, an area of significant seismicity.

Maximum Credible Earthquakes (MCE) for the

damsite were determined to be a magnitude 7.5

at a distance of 19 miles, and a local magnitude

6.0 to 6.5. The dam has been designed to take

maximum advantage of the excellent

construction materials available at the site. Use

of these materials in combination with a sloping core has permitted the use of economical 1-1/2:

1

downstream and 2:1 upstream slopes. Seismic

analyses confirmed that the structure could

withstand seismic loadings with minimal

damage. This paper will summarize the design

and analysis of the structure, with particular

emphasis on seismic design considerations.

KEYWORDS: Embankment; Dam; Stability;

Earthquakes; Slopes; Seismicity; Analysis;

Design

1. INTRODUCTION

Jordanelle Dam and Reservoir, located on the

Provo River near Heber City, Utah, are an

important feature of the United States Bureau of

Reclamation's Central Utah Project. Its primary

function is to store and deliver municipal and

industrial water to the growing Salt Lake

City/Provo-Orem metropolitan areas. The

Reservoir will have a total capacity of over 395,000,000 cubic meters (320,000 acre-feet),

impounded by constructing Jordanelle Dam to a height of approximately 91 meters (300 feet)

above original ground. The dam is a compacted

embankment dam containing over 10.7 million

cubic meters (14 million cubic yards) of soil and

rock material. It has a crest length of 1,125

meters (3,700 feet) and a crest width of

12 meters (40 feet) as shown in Figure 1. Waterways include a fuse plug controlled

emergency spillway and a tunnel outlet works,

consisting of both low level and multi-level

intakes. Jordanelle Dam was designed in two

stages between 1986 and 1989, and construction

began in 1987. It is anticipated that it will be

essentially complete in early 1993, with

reservoir filling to commence later that year.

This paper describes the design concepts for

Jordanelle Dam, as related to the consideration

of seismic loadings anticipated in this location.

Additional analyses undertaken during the

construction currently in progress are also

presented.

2. DESCRIPTION OF EMBANKMENT

DESIGN

The embankment cross section is shown on

Figure 2. As shown, it consists primarily of

compacted zone 4 material, described in Table 1 . The upstream slope is 2 horizontal to

1 vertical, and the downstream slope is 1-1/2

U.S. Bureau of Reclamation, Geotechnical

Engineering and Embankment Dams Branch,

Denver, CO 80225

Wilson

87
horizontal to 1 vertical. A relatively thin

upstream sloping clay core, zone 1, is the

primary impervious element. It is protected by

a two-stage filter-drain system (zones 2 and 3) to prevent piping and control seepage. Slope

protection upstream consists of a 1 meter

(3-foot) thickness of riprap obtained from nearby

quarries. Downstream slope protection for the

relatively steep 1-1/2:1 slope consists of cobbles

and boulders (zone 5) processed from the

borrow area.

The upstream sloping core is a key feature in the

configuration of the embankment. It was made

relatively thin to minimize the amount of

foundation preparation and treatment required

beneath the zone 1. It was sloped upstream to reduce the potential for differential settlement

between the more compressible zone 1 core and

the relatively incompressible zone 4 shells. In addition, the inclusion of the core and filter- drain system in the upstream portion of the dam

leaves the vast downstream bulk of the

embankment in an unsaturated condition. This

ensures both static and seismic stability. The

filter-drain system immediately downstream of

the core serves as a defense mechanism to ensure that any cracking created by earthquake

loading is properly protected by a filter and that any resulting seepage is safely conveyed

downstream by the drainage system. The ability to keep the downstream shell unsaturated, plus

the availability of good quality, strong materials,

enabled the construction of a relatively steep

1-1/2:1 downstream slope.

A significant factor in the fmal configuration of

the embankment was the quality and quantity of

available materials. Fortunately, the reservoir area contained an abundant supply of both

impervious materials (gravelly clays) and

pervious materials (sandy gravels with cobbles

and boulders) of excellent quality. The pervious

materials, zone 4, are particularly desirable due

to their low compressibility, high strength,

durability and relatively high permeability.

Because of these nearly ideal properties, the

design utilized this material to the largest extent

possible. The high strength made relatively

steep slopes possible. This reduced the overall

embankment volume and lowered the

construction cost significantly.

Although the pervious materials are plentiful,

there was little contingency in the design for a possible underrun of available zone 4 materials.

For that reason, provision was made for a 1,700,000 cubic meter (2,200,000 cubic yards) zone 4A fill within the downstream zone 4 shell. Pervious materials were the preferred material

for the zone; however, the zone was designed to permit the use of other excavated materials, such

as gravelly clays. This provides for the use of

suitable, but unspecified, materials from the

borrow area and excavations. The impact of this zone on embankment stability was evaluated, as discussed later.

3. SEISMOTECTONIC SETTING

Jordanelle damsite is located in the Wasatch

Mountains within the Intermountain Seismic Belt

(ISB) in north-central Utah (Figure 3). The ISB

is a zone of contemporary seismicity extending

from Arizona to Montana along the eastern margin of the Basin and Range province. It is considered to have the highest level of

earthquake risk in the contiguous United States outside of California and Nevada. Within the

ISB, three historical earthquakes with

magnitudes ranging from 6.6 to 7.5 have been

accompanied by surface rupture on late Quaternary faults.

Potential earthquake sources for damsites in the

ISB are considered to be of two types: 1) largemagnitude earthquakes on faults with evidence

of late Quaternary surface rupture and 2) a local random earthquake that occurs on "blind"

subsurface faults. The most likely sources of

significant ground motions in the region that could affect Jordanelle damsite are the Wasatch

fault, faults in Round Valley, and the random

earthquake. A maximum credible earthquake

(MCE) of magnitude 7.5 is assigned to the

Wasatch fault at a distance of 30 km (18 mi)

Wilson

88
from the damsite. An MCE of magnitude 6.5 to 6.75 is assigned to the faults in Round Valley at a distance of 20 km (12 mi) from the damsite.

An MCE of magnitude 6 to 6.5 is assigned to a

local random earthquake source.

Geologic investigations have shown that late Quaternary surface displacements have not occurred on faults in the vicinity of the damsite.

We have, therefore, concluded that there is no

potential for significant coseismic surface

rupture in the dam foundation. However,

ground deformation has been associated with

recent moderate-magnitude earthquakes that have

occurred on blind faults in the ISB. Tectonic

subsidence accompanying the 1975 Pocatello

Valley earthquake of magnitude 6.0 suggests that a similar phenomenon could accompany a magnitude 6 to 6.5 earthquake on a blind fault in the vicinity of the dam. Historic data and model

simulations indicate that a tectonic displacement

of about 15 cm could occur on favorably

oriented faults in association with a moderatemagnitude earthquake.

4. MATERIALS CHARACTERIZATION

Shear strength parameters used for the principal

embankment materials in the preconstruction

stability analyses are shown in Table 2 for each

loading condition.

The strengths shown in Table 2 were determined

as follows:

Zone 1 - Core: Unconsolidated Undrained (UU)

triaxial tests were performed on specimens

compacted to 98 percent Proctor Density. These were performed in order to obtain undrained

shear strengths representative of the gravelly

clay core material during construction.

Consolidated Undrained (CU) triaxial tests (with

measured pore pressures subtracted) were used

to develop the drained strength used in the

steady seepage analysis. The same tests (without

measured pore pressure subtracted) were used to develop consolidated undrained strengths

representative of undrained strength available in

the consolidated clay core during a rapid

drawdown loading.

Zone 2 - Filter: Laboratory testing of the well

graded sand and gravel filter material was not performed. Drained strength parameters for the

well compacted filter materials were

conservatively selected from available literature

and past Reclamation experience with similar

materials.

Zone 3 - Drain: Laboratory testing was not considered practical due to the large particle size

(3.8 to 20.3 cm) of this material. Conservative

strength parameters, considering the rounded

and highly uniform gradation of the material, were also selected from available literature and

past Reclamation experience.

Zone 4 - Shell: Initially, the zone 4 material was assumed to behave dilatively during shear.

Drained strengths were, therefore, assumed to represent all loading conditions for this material.

The zone 4 was assumed to have drained

strengths with an effective (p of 44° (Marachi,

1972) below a normal stress of 690,000 N/m^

(100 Ib/in^ and a <^ of 39° with a cohesion

intercept of 107,600 N/m- (15.6 Ib/in^ above a normal stress of 690,000 N/m= (100 lb/in').

These strengths are supported by the work of

Leps (1970) with large scale triaxial tests on

Oroville Dam material.

Zone 4 shell material is placed in two-foot lift thicknesses and compacted using vibrating

rollers. Test fills prior to construction were

evaluated to establish the optimum procedures

for placing and compacting this material. At the

beginning of the stage 2 construction, samples of

this material were shipped to the University of

California, Berkeley for testing in their

30 centimeter (12-inch) diameter triaxial apparatus (Seed, 1990). Tests were performed on samples scalped to a 5-cm (2-inch) maximum

particle size and compacted at a density

corresponding to material placed in stage 1 . The

results indicated that the material might not behave as originally anticipated. The tests

Wilson

89
indicated that materials compacted during stage

1 could be slightly contractive rather than

dilative during shear. As a result of these tests, a heavier vibratory roller (1 1-ton-per-drum dualdrum) was specified for stage 2, resulting in higher densities than those achieved during stage

1 construction. In addition, a re-evaluation of

the design was made to check the impact of

these in-place stage 1 materials on embankment

performance.

Based on the test results, the zone 4 material

placed during stage 1 would be represented by a consolidated stress phi angle of 19° for a postearthquake analysis. This conclusion is based on

the assumption that the total material will contract and build up pore pressure. In reality,

though, the oversized material would likely

inhibit the increase in pore pressure. Due to the

greater densities of the zone 4 material being

placed in the upstream shell during stage 2, the

strength is conservatively represented by a phi of

30° for a post-earthquake analysis. The

downstream zone 4 material is also being placed

at the higher densities and will remain

unsaturated. An effective friction angle of 39°, as indicated by the Berkeley results (Seed, 1990) was used in the reanalysis.

Zone 4A - Fill: The strength used for the

gravelly clays allowed in the Zone 4A mass fill portion of the dam were assumed the same as the drained and undrained strengths obtained

from laboratory tests on the zone 1 core

materials. This provides for the flexibility of

placing materials of lesser strength in this zone.

5. STATIC STABILITY EVALUATION

The static stability was evaluated for three

loading conditions. These were: steady-state

seepage, rapid drawdown, and end of

construction. The analyses were performed

using the computer program SSTAB2 (Chugh,

1981), based on the Spencer method, to compute

factors of safety for various failure surfaces.

The downstream slope was evaluated for a range

of zone 4A configurations. For these analyses,

it was conservatively assumed that the zone 4A

consisted of impervious gravelly clay materials.

The end of construction case was examined

using the conservative assumption that pore

pressure developed during placement in either

the core or the zone 4A portion of the

downstream shell would not dissipate prior to completion of the dam. The gravel and cobble zone 4 material placed in the embankment was

assumed to be free draining in the analyses.

The factors of safety obtained for the end of

construction condition are shown on Table 3.

The steady seepage condition was examined with

the reservoir water surface at the maximum

elevation of 1884.3 m (6182 ft) and the phreatic

surface established by the top of joint use water

surface elevation of 1879.5 m (6166 ft). The

factor of safety was obtained for different

circular searches and wedge surfaces for the

steady seepage condition assumptions. The

minimum factor of safety obtained for a

significant failure surface is shown on Table 3 and meets the Reclamation design requirements

(USBR, 1987).

For the steady seepage condition, the internal

drainage systems were assumed fully functional.

The clayey zone 4A placed in the downstream

shell was assumed to remain unsaturated at a moisture content near optimum. Pore pressures

in the relatively thin sloping core were estimated

by assuming full hydrostatic pressures on the

upstream face of the core and atmospheric

pressures on the downstream underside of the

core.

For the rapid drawdown condition, no pore

pressure dissipation due to drainage in the zone 1 core material was assumed. The zone 4

gravel and cobble upstream shell was assumed

pervious enough to drain completely during the

reservoir drawdown. Construction pore

pressures in the clayey zone 4A materials were

assumed to have dissipated completely.

Wilson

90
The rapid drawdown condition was examined for a complete draining of the reservoir. This was

assumed to follow the establishment of steady

seepage conditions with the reservoir at elevation

1879.5 m (6166 ft). The values of the factor of

safety obtained for the different surfaces for the

rapid drawdown conditions are shown on

Table 3.

6. SEISMIC GROUND MOTIONS

The maximum credible earthquakes (MCE's)

derived for the Jordanelle damsite are shown in Table 4 (USBR, 1988). The Wasatch fault event (No. 3, Table 4) is considered to be the more

probable event. It has a recurrence interval for a magnitude 7.0 to 7.5, earthquake of 330

C± 90) years for the entire fault zone (Youngs,

1987) . The two segments of the Wasatch fault nearest to Jordanelle Dam were assigned recurrence intervals of about 2,000 years.

Three separate accelerograms were used to represent potential earthquakes for analysis. The

first was a record developed at the University of

California, Berkeley, by Professor H. Bolton

Seed. It represents a near-field magnitude 7.25

earthquake on the Hayward Fault in the San

Francisco Bay area (Seed, 1983, Pike, 1991). It was scaled to a peak acceleration of 0.67g. This was to represent a magnitude 6.5 event in the

immediate vicinity, not associated with a specific

fault. This resulted in a maximum velocity of

76 cm/s (30 in/s) and a velocity spectrum

intensity of 286 cm (113 inches). The latter value is very large in comparison with available

data from historical earthquakes (Von Thun,

1988) . The record has a total duration of

23.7 seconds and contains a long-period section

referred to as a "fling". This fling causes the

motion to have a great deal of energy in the

1.0 to 1.5 second period range. Figures 4 and

5 show the acceleration time history and the

acceleration response spectrum for this earthquake record.

The second design earthquake record is a

synthetically generated accelerogram considered

to be representative of a mean magnitude 6.5

local event. It was developed for use in this type of seismic setting, (Munoz, 1986). It is characterized by a peak acceleration of 0.55g, a peak velocity of 35 cm/s (14 in/s), a total duration of 20.5 seconds, and a velocity

spectrum intensity of 163 cm (64 in).

The third earthquake record represented a magnitude 7.5 event on the Wasatch fault at a distance of 30 kilometers (19 miles). An

accelerogram developed by H. B. Seed and I. M. Idriss to represent the far field of a magnitude 8.25 earthquake was used. It was

scaled to a peak acceleration of 0.29g. This

resulted in a record with a peak velocity of

25 cm/s (10 in/s), a velocity spectrum intensity

of 171 cm (67 inches) and a total duration of

48 seconds.

7. DYNAMIC DEFORMATION ANALYSIS

Seismically induced embankment deformations were evaluated using the computer program

DYNDSP (Von Thun, 1981). This program is based on the sliding block procedure by

Newmark (1965) for estimating permanent

displacements. It is assumed that slide planes

develop within the mass and that the material

above the slide plane behaves as a rigid body.

The DYNDSP program uses the limit

equilibrium method. It computes an incremental

permanent displacement along the sliding plane

each time the earthquake loading causes the

driving forces along the plane to exceed the

resisting forces. The effective angle of motion

is calculated as the weighted average of the

angle of sliding of each individual slice. The

computed displacement is, therefore, calculated

as the movement of the sliding mass along the

failure surface in the downslope direction.

Selective incorporation of the effects of

horizontal and vertical inertial forces, excess pore pressure development, shear strength

reduction, and dilation is included.

Wilson

I 91
The program estimates the relative displacement

of a specific failure mass given the applicable

time history of both horizontal and vertical

accelerations. The critical surfaces for a structure must be found by the examination of

many failure surfaces. The total displacement is determined by integration of the results from

these individual surfaces.

Several failure surfaces and driving acceleration

records were analyzed for determining response

to the MCE's. The results are shown in Table 4.

The shear strength parameters used in the

original, preconstruction dynamic analysis are shown in Table 2. These strengths reflected no

loss in shearing resistance due to dynamic

loading conditions. It was assumed that the

embankment would be compacted to a density

sufficient to preclude significant loss of shear

resistance associated with excess pore pressures

in the saturated zone 4 material (upstream). The

materials within the downstream shell (zone 4

and 4A) were assumed unsaturated. The

downstream failure surfaces were, therefore,

located primarily in unsaturated material, a planned result of the sloping core design.

The results of the analysis indicate that up to 1.4 m (4.6 ft) of vertical deformation is possible. This occurred on a circle which passes

through half of the crest to one-fourth of the

distance down the downstream slope (Fig. 6).

The remaining material from the crest would be

expected to slough into a more stable position.

Total loss of crest elevation would not be

expected to exceed the 1.4 m (4.6 ft). All of the

deeper failure surfaces evaluated resulted in less deformation. Hence, it was concluded that with

5.7 m (18.6 ft) of designed freeboard above

normal reservoir elevation, the embankment

would not lose its reservoir-impounding

capacity.

The first design earthquake record previously

discussed caused significantly larger

deformations than the synthetic record or the

Wasatch fault earthquake (Table 3). Even so, the displacements of up to 1.4 m (4.6 ft) on

shallow upper surfaces, if they were to occur, would require repair, but would cause no loss of

the reservoir water.

The largest computed vertical displacement from

the synthetic earthquake record was 0.55 m

(1.8 ft). This is believed to be the largest

seismic loading with a reasonable probability of

occurrence. Under this earthquake loading the

embankment is expected to retain approximately

90 percent of the design freeboard.

The magnitude 7.5 earthquake on the Wasatch

fault has a predominance of long period

vibrations in the design record. This caused

amplification of the motions within the

embankment up to a peak acceleration of 1 .2 g

near the crest of the dam. This, coupled with a

total duration of 48 seconds, yielded calculated

vertical displacements up to 0.85 m (2.8 ft). The largest displacement was located on the

shallow upstream shear surface (Table 3). Over

4.7 m (15.5) ft of freeboard would remain after

this seismically induced deformation.

As a check on the reasonableness of the results presented above, a simplified analysis was also

performed. The procedure described by Makdisi

and Seed (1977) was used to estimate the

earthquake-induced deformations of the

embankment with and without the clayey zone 4A in the downstream shell. Deformations were computed for potential failure surfaces through

the full height of the embankment, the upper

half of the embankment, and the upper quarter

of the embankment. The analysis was

performed using the computer program

SEIDAKA (Adhya, 1982), which uses the

Makdisi-Seed (1977) procedure. Maximum

estimated deformations were:

,

0.06 m (0.2 ft) through the ftill height,

0.24 m (0.8 ft) through the upper half, and

2.0 m (6.5 ft) through the upper quarter.

Wilson

92
These displacements compare reasonably with

the values obtained using the more rigorous

DYNDSP procedure, and confirm the acceptable

performance of this design.

As a further check on the probable performance

under earthquake shaking, literature reports were

reviewed. Several gravelfill or rockfill dams, of

varying size and design, which have been

subjected to severe seismic shaking suffered

minimal, or no damage (Bureau, 1986, and

Seed, 1977). Two modern, well-instrumented

rockfill dams with thin impervious zones

survived strong shaking from the 1985

magnitude 8.1 Mexico earthquake with minor

cracking and settlement (Bureau, 1986).

LaVillita Dam is located about 24 km (15 miles)

from the epicenter of the main shock. It recorded about 50 seconds of ground motion

with a peak acceleration of 0.13 g. El

Infiernillo Dam, about 74 km (46 miles) from

the epicenter recorded an estimated peak

acceleration of 0.5 g on the dam crest. These

two dams have been subjected to 5 other

earthquakes of magnitude 6.5 or larger since

1964, as well as a magnitude 7.5 aftershock to the 1985 event, and are continuing to perform

well. Jordanelle Dam has been designed to provide seismic behavior comparable to these

two dams.

Following the completion of the laboratory

testing on the zone 4 soils at the University of

California, Berkeley, the deformation analyses were repeated using the reduced shear strengths.

Only materials placed during stage 1 were

considered capable of experiencing significant

loss of effective shear resistance. It was

conservatively assumed that widespread pore

pressure increase occurred in zone 4 materials

placed during stage 1 in the saturated upstream

shell. The analysis results indicated that the

embankment would retain sufficient strength to remain stable. Seismically induced deformations

would still be considerably less than the

available freeboard.

8. CONSTRUCTION MONITORING

Construction began in 1987 as stage 1 of a twostage construction process. It included stripping

the abutments and west side of the valley to bedrock, producing zone 2 and 3 filter and drain

materials, and placing embankment material on

the west side of the valley up to the original

ground elevation. Stage 2 construction, which

includes all remaining construction activities,

began in 1989. It is scheduled to be completed

in early 1993. At the time of this writing,

construction of the outlet works tunnels is completed, the embankment is complete to approximately elevation 1838 m (6030 ft), or 40 m (130 ft) above original ground, and

construction of the emergency spillway is complete.

As anticipated, material placed in the

downstream zone 4A portion of the embankment

have varied in origin and composition. Much

of the material placed to date has been material

from required excavations, with a wide range of

gradations. In general, the material placed has

been superior to the "poorest" material

considered during design analyses.

Due to an overabundance of zone 3 drainage

material from the stage 1 material processing,

several upstream horizontal drainage layers have

been incorporated into the design of the

embankment. These layers will serve to allow

rapid drainage and dissipation of pressures in the event of rapid drawdown or seismic shaking,

providing an extra margin of stability.

9. CONCLUSIONS

Testing of the materials being placed in the

embankment confirm that the dam is being

constructed in accordance with specifications

requirements. Close monitoring and inspection

of construction activities is being carried out by

construction and design personnel. This assures

that the design is being properly adapted to site

Wilson

93
conditions, and that design concepts and

assumptions are valid for the as-built structure.

The availability of excellent materials for

embankment dam construction made construction

of Jordanelle Dam with a relatively steep 1-1/2:1

downstream slope in a seismically active area

feasible. Investigations and analyses indicate

that the structure will be stable and perform well

for all anticipated static and dynamic loading

conditions. The analyses were carried out using

commonly available methods to calculate the

static factors of safety and anticipated seismic

performance of the dam. Careful construction

monitoring and reevaluation of the design

assumptions during construction have indicated

satisfactory conformance to design concepts with

regard to both materials and densities in the

field.

10. REFERENCES

1. Adhya, K. A. (1982). "A Computer

Program for Simplified Earthquake-Induced

Deformation Analysis of Homogeneous

Embankment Dam," Draft Report, U. S. Department of the Interior, Bureau of

Reclamation, Denver, Colorado.

2. Bureau, G. and Campos-Pina, J. M. (1986).

"Performance of Mexican Dams, Earthquake

of 1985," U.S.C.O.L.D. Newsletter, Issue No. 79, Published by the U.S. Committee

on Large Dams, Boston, Massachusetts.

3. Chugh, Ashok K. (1981). "User

Information Manual, Slope Stability Analysis

Program SSTAB2," (A Modified Version of

"SSTABl" by Stephen G. Wright), U. S. Bureau of Reclamation, Denver, Colorado.

4. Chugh, Ashok K. (1988). "Static

Deformation Analysis of Proposed Maximum

Section," Technical Memorandum, Division

of Dam and Waterway Design, U. S. Department of the Interior, Bureau of

Reclamation.

5. Hensley, P. J. (1988). "Static Stability

Analysis, Jordanelle Dam, Central Utah

Project, Utah," Technical Memorandum No.

JD-3620-7, U. S. Department of the Interior,

Bureau of Reclamation, Denver, Colorado.

6. Lambe, T. W. and Whitman, R. V. (1969).

Soil Mechanics . John Wiley and Sons, Inc.

7. Leps, T. M. (1970). "Review of Shearing

Strength of RockfiU," Journal of the Soil

Mechanics and Foundations Division, ASCE.

8. Makdisi, F. I., and Seed, H. B. (1972). "A

Simplified Procedure for Estimating

Earthquake-Induced Deformations in Dams

and Embankments," Earthquake Engineering

Research Center, Report No. UCB/EERC77/19, University of California, Berkeley,

California.

9. Marachi, P. N., Chan, C. K., and

Seed, H. B. (1972). "Evaluation of

Properties of Rockfill Materials," Journal of

the Soil Mechanics and Foundations

Division, ASCE.

10. Munoz, Richard L. and Greg A. Scott (1986). "Preliminary Sttidies for Conceptual Designs of Monks Hollow Arch

Dam," Technical Memorandum No. MH221-1, U. S. Department of the Interior,

Bureau of Reclamation, Denver, Colorado.

11. Newmark, N. M. (1965). "Effects of

Earthquakes on Dams and Embankments,"

Geotechnique, Vol. 15, No. 2.

12. Pyke, Robert (1991). personal

communication.

13. Seed, H. B., Makdisi, F. I., and

DeAlba, P. (1977). "The Performance of

Earth Dams During Earthquakes,"

Earthquake Engineering Research Center,

Report No. UCB/EERC-77/20, University

of California, Berkeley, California.

Wilson

94
14. Seed, H. B. (1983). personal

communication.

15. Seed, R. B. (1990). "Final Report

Presenting Results of Large-Scale Triaxial

Tests, Zone 4, Area "G" Borrow Material,

Jordanelle Dam Project, Utah," U. S. Department of the Interior, Bureau of

Reclamation, Denver, Colorado.

16. Sherard, J. L., and Dunnigan, L. P. (1985).

"Filters and Leakage Control in Embankment Dam.," Symposium of

Seepage and Leakage from Dams and

Impoundments, American Society of Civil

Engineers, Spring Convention, Denver,

Colorado.

17. U.S. Department of the Interior, Bureau of

Reclamation (1987). "Design Standards,

Embankment Dams, No. 13, Chapter 4,

Static Stability Analyses," Denver,

Colorado.

18. U.S. Department of the Interior, Bureau of

Reclamation (1988). "Seismotectonic Study

for Jordanelle Dam, Bonneville Unit,

Central Utah Project, Utah,"

Seismotectonic Report 88-6, Seismotectonic

Section, Division of Geology, Denver,

Colorado.

19. Von Thun, J. L. and C. W. Harris (1981).

"Estimation of Displacements of Rockfill

Dams Due to Seismic Shaking,"

Proceedings of the International Conference on Recent Advances in Geotechnical

Earthquake Engineering and Soil

Dynamics, St. Louis, Missouri.

20. Von Thun, J. L., L. H. Roehm,

G. A. Scott, and J. A. Wilson, (1988).

"Earthquake Ground Motions for Design

and Analysis of Dams," Proceeding of the

ASCE Specialty Conference, Earthquake

Engineering and Soil Dynamics II - Recent

Advances in Ground Motion Evaluation,

Geotechnical Special Publication No. 20,

Park City, Utah.

21. Youngs, R. R., F. H. Swan, J. S. Power,

D. P. Schwartz, and R. K. Green (1987).

"Probabilistic Analysis of Earthquake

Ground Shaking Hazard Along the Wasatch

Front, Utah," Assessment of Regional

Earthquake Hazards and Risk Along the

Wasatch Front, Utah, U. S. Geological

Survey, Open File Report 87-585.

Wilson

95
EMBANKMENT TYPE OF MATERIAL

Zone 1 Lean clays, sandy-gravelly clays

Zone 2 Processed sands and gravels, max size 5 cm

Zone 3 Clean gravels and cobbles, max size 20 cm

Zone 4 Well graded sand, gravel, cobbles and boulders, max

size 61 cm

Zone 4A Variable, ranging from lean clays to rockfill

1 Zone 5 Cobbles and boulders, max size 61 cm

Table 1. Types of Materials in Embankment Zones

Principal Loading Conditions

tITIiJanKIIIcn Zone

End of Const. SteadySeepage Rapid Drawdown Dynamic

Zone 1 Conf. stress <275,000 N/m^

(40 psi)

Conf. stress >275,000 N/m^

(40 psi)

c=11 psi 0 = 21°

c=2.5 psi 0 = 28°

c=2.5 psi 0 = 28°

c=2.5 psi 0 = 28°

c=2.5 psi 0 = 28°

c=16 psi 0 = 13°

c=2.5 psi 0= 28°

c= 16 psi 0 = 13°

Zone 2 c=0

0 = 35° c=0

0 = 35° c=0

0 = 35° c=0

0 = 35°

Zone 3 c=0

0 = 30° c=00 = 30° c=00 = 30° c=0

0 = 30°

Zone 4 c=0

0 = 44° c=0

0 = 44° c=00 = 44° c=00 = 44°

Zone 4A

Full pore press, dissip.

c=2.5 psi 0 = 26° c=2.5 psi 0 = 26° c=2.5 psi 0 = 26°

No pore press, dissipation

a) conf. stress

< 550,000 N/m^

(80 psi)

0=11 psi 0 = ir

c=11 psi 0 = irc=11 psi 0 = 17°

b) conf. stress

> 550,000 N/m^

(80 psi) c=40 psi 0 = 0 c=40 psi 0 = 0 c=40 psi 0 = 0

Table 2. Material Strengths Used for Preconstruction Stability Analysis.

96
End of Const Steady

Seepage

Rapid

Drawdown

Downstream

Infinite slope

1.5 1.5 —

Circle 1.5 1.5

1 .4*

...

Wedge

Upstream

Infinite slope

1.9 1.9

Circle 2.1 2.0 2.0

Wedge 2.7 4.5 2.4

*No dissipation of construction pore pressure in zone 4A.

Table 3. Factors of Safety for Static Stability

EQ. Magnitude Distance

Earthquake

Record Used

Peak

Veloci ty

Maximum Vertical

displacement (m)

Clay

zone 4A

Gravel zone 4A

No. 1 M, 6.5 local Scaled

PacoimaTaft 76 cm/s0.4 u/s

0.5 d/s

1.3 u/s

1.4 d/s

No. 2 M, 6.5 local Synthetic, USBR 35 cm/s

0.3 u/s

0.3 d/s

0.5 u/s

0.5 d/s

No. 3 M, 7.5 30 km. Scaled,

Seed- Idriss 25 cm/s

0.4 u/s

0.1 d/s

0.9 u/s

0.3 d/s

Table 4. Maximum Credible Earthquakes and Seismically-Induced Deformations.

97
Figure 1 - Jordanelle Dam General Plan

98
Figure 3 - Location of Jordanelle Dam and Wasatch Fault Zone

I I I I I I I I I I I I I I I I I I I I 0.0 1.0 2.0 3.0 1.0 5.0 6.0 7.0 6.0 9.0 10.0 11.0 12.0 13.0 14.0 15.0 16.0 17.0 19.0 19.0 20.0 TIME (Seconds)

Figure 4 - Jordanelle Dam Design Accelerogram

99
2,5

PERIOD (Seconds)

Figure 5 - Jordanelle Dam Acceleration Response Spectrum

Figure 6 - Jordanelle Dam Deformation Analysis

100
An Evaluation of Seismic Forces Acting on Embankment Dams and Its Simplified Prediction Procedure

by

Tadahiko Fujisawa', Nario Yasuda^ Norihisa Matsumoto^ and Kenji Yamabe'*

Abstract

Seismic force acting on embankment dams is studied using one- and

two-dimensional dynamic analyses. In

this report, the distribution of

seismic acceleration in a typical dam

body section is computed using acceleration records of actual earthquakes.

On the basis of response computations

to different input motions' with var- ious frequency characteristics, a relationship for the maximum response

acceleration with frequency characteristics of the ground motions is established.

A simplified procedure to estimate the maximum acceleration from the

acceleration response spectra of

ground earthquake motions is also

presented.

Keywords: Embankment dam, maximum

acceleration, dynamic analysis, earthquake motion, permanent deformation

1. INTRODUCTION

This paper discusses the relationship between frequency characteristics of earthquake motions on base

rock and seismic forces acting on

embankment dams. Frequency-dependent

characteristics of the seismic force

acting on embankment dams were clarified by computing earthquake responses

of the dam body to various input

motions with different frequency

characteristics. In addition, a procedure was developed to easily

estimate seismic force from the acceleration response spectra of earthquake

ground motions.

A method to estimate the safety

of a dam body during earthquakes based on the permanent deformation of the

dam has been developed since the

prominent research of Newmark( 1965

)

and Seed(1966) for embankment dams

with banking materials which do not

suffer reduction of strength even when

subjected to cyclic loading. The

major parameters for computing the

amount and extent of such permanent

deformation are: 1) average acceleration, a, which is obtained by dividing

the force of inertia acting on a potential sliding soil mass hypothesized in a dam body by the weight of

this sliding mass; and 2) maximum

average acceleration for_the duration

of earthquake motions, a^max. If yield

acceleration, ay, is defined as the

acceleration when the safety factor of

the sliding soil mass becomes 1.0,

then the amount of deformation of an embankment dam due to an earthquake is

computed based on the concept that the

soil mass begins to slide at the

moment when average acceleration

exceeds ay. However, the deformation

can be approximated simply by the

ratio of ay to a„ax. Watanabe at

al. (1984) calculated the sliding

deformation by Newmark ' s method with 5 sinusoidal waves which periods range

from 0.125 sec. to 0.625 sec, and

concluded that the permanent deformation at crest is in propotion to the

square of the duration of a which

exceeds ay at any location of sliding

mass. Makdisi et al. (1978) depicted

the relationship between ay/amax and

permanent deformation at every magnitude of earthquake event, conducting

the calculation of sliding deformation

by Newmark 's method with real earthquakes. In addition, efmax is closely

related to the concept of seismic

coefficients pratically used in stability analyses of embankment dams

based on the seismic coefficient

method, that is, a"„ax expresses the

seismic force acting on the body of an embankment dam. Considering the

strength characteristics of soil, strain in soil materials does not

greatly increase when the loading

occurs during a short time period of one cycle; it does, however, increase

when the soil is subjected to cyclic loading over_a certain number of

cycles. When a, which is greater than

a specified value, occurs repeatedly

more than a certain number of cycles,

this a" is defined as equivalent acceleration, "aeq, and is regarded by Seed

et al.(1966) as a seismic coefficient

for the seismic coefficient method.

Initially, Seed et al. (1966 ) and

Ambraseys et al.( 1967) clarified that

amax for a wedge-shaped potential

sliding mass obtained by using a dam

body modelled into a homogenious

triangle, the sliding surface of which

was approximated by a straight line,

varied depending on the elevation of

the mass and the natural frequency of

*1 Head, Fill-type Dam Division,

Public Works Research Institute,

Ministry of Construction.

*2 Senior Research Engineer, ditto

*3 Dr. Eng., Senior Research Engineer

of Dam Center Incorporated Foundation. (Former Head of Fill-Type Dam

Division

)

*4 1st Technical Division, the Water

and Engineering Headquarters of the

Osaka Branch, Research Institute for

Construction and Technology, Co. , Ltd.

(Former visiting researcher)

101
the dam body. In the numerical analy- ses conducted in these studies, the

procedure for a one-dimensional dynamic analysis based on the shear beam

theory proposed by Mononobe et

al. (1936) _was introduced in order to

estimate amax for the sliding mass.

Later, Chopra et al. (1967) showed that

such values for a„ax computed by the

one-dimensional shear beam theory were greater by 10 to 20 % on average than

those computed by a two-dimensional

dynamic analysis with Finite Element

Method. amax is represented by a function of y/H, where H denotes the

height of a dam and y the difference

between the elevation of the dam crest

and the minimum elevation of a s^liding arc. The curve of normalized a'max in

relation to y/H mentioned above( hereafter referred to as an acceleration

curve) can be obtained from devided

amax by the maximum response acceleration at a dam crest, Umax. In accordance with the study by Makdisi et

al.( 1979), similar acceleration curves are obtained by both one- and twodimensional analyses, however, the

curve fluctuates to some extent depending on the natural frequency of

the dam body, the upper and lower

limits of the fluctuation are within

the sphere of +10 to +20 % for the

upper elevation and +20 to +30 % for

the lower elevation, when average

acceleration values are adopted.

Thus, it is considered that average

acceleration values can be used as a design curve for practical use. When

a conservative design is dealt with

from the aspect of safety, however, it

is recommended that the values em- ployed be 10 to 30 % greater than the

average values. Umax must be determined when seismic force is computed

from an acceleration curve. Makdisi

et al. (1978) proposed a simplified

procedure for determining Umax based

on the root mean square using the

acceleration response spectra of

design earthquake motions as well as the modal analysis based on the one- dimensional shear beam theory. Luan

et al. (1990) studied the effect of the

shape of the sliding surface on an acceleration curve; the shape of the

sliding surface is determined by the

slope gradient of a dam, the damping

ratio and the shear strength of soil

materials. They demonstrated that the

acceleration curve obtained using the

procedure by Makdisi et al. (1979)

resulted in underestimation at an elevation lower than the mid-height of

a dam, and that both the slope gradient of a dam and the shear strength of

soil materials were almost irrelevant

to the curve.

As described above, studies have

already been conducted on the relation

between a„ax and analytical procedures,

dam shapes, the elevation of the

sliding soil mass, the dam height and

physical properties of dam body.

Nevertheless, very few studies have

been hitherto conducted on aa^x in

relation to the frequency characteristics of input earthquake motions. in

this paper, therefore, the relationship between a„ax and the frequency

characteristics of input earthquake

motions is discussed and a simplified

procedure of computing a^ax from an acceleration response spectra is

summarized

.

2. ANALYTICAL MODEL AND METHODS

2.1. Model Dam

The model dam used in the analysis is a two-dimensional cross section

perpendicular to the dam axis. with

63-meter in height and the slope

gradients of 1:2.6 on the upstream

side and 1:2.0 on the downstream side.

The models illustrated in Fig. 1(a)

and Fig. 1(b) are employed in onedimensional and two-dimensional dynamic analyses respectively.

2.2 Physical Properties of the Model

Dam

Physical properties of the v;hole

section of the model dam are identical

to those of rockfill materials, based

on the idea that physical properties

of the filter and core materials will

have little effect on the results of

analyses because a rockfill dam is

mostly constited of rockfill materials. Shear modulus of rockfill material was determined from results

obtained by cyclic triaxial tests in

laboratory; taking into consideration

the dependent characteristics with

confining stress. The damping ratio

for the material has a considerable

influence on the response acceleration; therefore, its determination

becomes important when an analysis is

conducted. The damping ratio is

composed of internal damping of the

material itself, as well as energy

dissipation damping which is related

to boundary conditions such as the

foundation rock and reservoir. Fig. 2 shows the strain dependent curves of

normalized shear modulus and damping

ratio of the rockfill material that

constitutes the rockfill dam (Matsumoto et al.( 1990). The internal damping

corresponding to a strain level of 10-3 is about 10 % as shown in Fig. 2. It

has already been verified experimentally that energy dissipation damping

increases with the existence of the

reservoir water (Matsumoto et al . (1983). In addition, Ohmachi et

al.(1989) pointed out, based on actual

measurement, that energy dissipation

damping reached about 10 % when the

reservoir was full. Therefore, it can

be estimated that the damping ratio

for rockfill material during large

earthquakes when the reservoir is full

102
maximum acceleration in the base rock,

gmax, except fo^ Case 1. In addition,

the value of amax is small especially

on the dam foundations in Case 5 through Case 8 where predominant

frequencies of earthquake ground

motions are relatively high. The a^ax of an arbitrary y/H in Case 1, where

the predominant frequency is 0.59 Hz,

is the greatest of 8 cases, while a^ax at the dam crest is around 3.5 times

greater than gmax. The predominant

frequency of earthquake ground motions

in Case 4 is the closest to the natural frequency of a dam body. However,

the value of aT^ax at the dam crest in

Case 4 is smaller than that in Case 1 by around 300 gals. Thus, it is

proved that the magnitude of acceleration at the dam crest can not be

estimated only by the predominant

frequency of input ground motions.

Fourier spectra for Case 1 illustrate

that components of a long period

exceeding 0.3 sec. are predominant and

the ground motion is close to sinusoidal wave. Furthermore , comparing the

acceleration response spectra in Fig.

4 with a^ax in Fig. 7, it can be seen that the values of a„ax are great in

Cases 1, 2, 3 and 5, which have long

period components exceeding the natural period of a dam body of 0.6 sec. Fig. 8 shows the average values

of Fmax due to 8 input earthquake

motions calculated by the four methods

of analysis. As shown in Fig. 8, the

values of a'max obtained by one-dimensional analyses are about 15% greater

at the upper elevation than those

obtained by two-dimensional analyses;

the difference is about 10% at the

lower elevation. Thus, it was proved

that amax from modal analysis, which

require the least computation time,

were satisfactory comparing with those

obtained by other methods. The

magnification factors of the maximum

response acceleration, u(y/H),„ax, to

gmax for 8 input earthquake motions

were computed at arbitrary y/H corre- sponding to different elevations.

Fig. 9(a) shows distributions of such

magnification factors obtained by

averaging 8 different cases. The

values of responses obtained by a one- dimensional analysis are greater than

those obtained by a two-dimensional

analysis by about 10 % ; however, the

difference in responses is entirely

small. The magnification factors to

gmax range around 1.0 for y/H > 0.5,

and gradually increase along with the

increase of elevation, when the elevation is higher than y/H=0.5; finally,

they range from 2.2 to 2.5 when y/H

reaches 0.0 (the dam crest). Fig. 9(b)

shows distributions of ii ( y /H ) max/gm.x

for 8 input earthquake motions computed by QUAD- 4.

We compare the computed values as shown in Fig. 8 and 9 with the behaviors of actual dams. Fig. 10(a) shows

the relationship between maximum

acceleration magnifications at dam

crests and gmax monitored at embankment dams under the jurisdiction of

the Ministry of Construction. in

accordance with this figure, the

magnification ranges from 3 to 6 when

gmax is equal to or smaller than 20

gals, but it decreases as gmax increases, ranging from 2.0 to 2.5

especially when g'max is equal to or greater than 100 gals. The first

reason for this can be explained by

non-lineality in that damping increases as gmax increases. The second

is based on the fact that the earthquake event with small gmax usually

has a long epicentral distance and

induces earthquake motions with a predominant frequency close to the

resonance frequency of the embankment

dam, while the earthquake event with

gmax larger than 100 gals has a short

epicentral distance and an inclination to cause earthquake motions with

high frequencies. In other words, the

frequency characteristics of earthquake motions where gmax exceeds 100

gals correspond to that of Cases 6 through 8 in Table 1. The fact that

amax for y/H = 0 in Cases 6 through 8 in

Fig. 9 range 1.5 to 2.5 times greater

than gmax agrees with results obtained

by measurement, which indicates that

the supposition of material properties

used in the dynamic analyses was

reasonable. Fig. 10(b) shows the

maximum acceleration distribution of

embankment dams for each elevation;

the figures are arranged in order

based on the measured data at embankment dams of the Ministry of Construction, together with the results given

in the reference papers (Sawada et

al. ( 1979, 1986), Annaka et al. (1990),

Hasegawa et al . (1983) ). Fig. 10(b)

includes earthquake data where the

values of gmax are around 5 gals.

Since damping is small in such earthquake responses, magnifications in

Fig. 10(b) are greater than those in

Fig. 9(a). Acceleration distribution

at each elevation in Fig. 9(b) ranges

between the lowest and average values

in Fig. 10(b). In short, if the magnification is considered to be distributed in the middle of the lowest and

average values in Fig. 10(b), due to

the fact that damping increases when

gmax is 0.2G, it can be said that the

results of analyses shown in Fig. 9 coincide with the measured results.

Based on the above examinations, the

analyses conducted in this paper are regarded as comparable with measured

results. Fig. 11 shows the acceleration curve which was derived by dividing the average values of amax by the

values of Umax with 8 input earthquake

motions in Fig. 9 . In addition. Fig.

11 also includes the results of Mak103
is about 20 % in total. In previous

analytical studies ( Ambraseys et

al. (1967), Chopra. ( 1967 ), Makdisi et

al .( 1978, 1979 ) , Luan et al. ( 1990)) a percentage of 20 % was employed as a damping ratio. In this paper, accordingly, the damping ratio is constantly

fixed at 20 %, regardless of either

the shear strain or the mean principal

stress, in consideration of both the

intensity of the earthquake expected

and the results of these past studies.

2.3. Analytical Methods

The following four methods were

used for dynamic analysis: the modal

analyses and SHAKEN (Chugh (1985)) for

one-dimensional dynamic analysis, as well as QUAD-4 ( Idriss et a2.(1973))

and FLUSH ( Lysmer et al. (1975)) for

the two-dimensional dynamic analysis;

a being computed by each method.

The modal analysis is used to

conduct earthquake response analyses

of a structure during earthquakes, in

which the vibration of a structure

with multiple degrees of freedom is

expressed by superposing several

systems of a single degree of freedom

by the aid of the orthogonality of

normal modes (Okamoto et al.( 1975)).

In adapting the modal analysis to

embankment dam , the dam was modelled

by a set of masses and stiffness

(elastic spring) as illustrated in

Fig. 3. The gross mass of a layer,

Ml, was set in the center of gravity

of each layer and represented by the mass point, while stiffness (elastic

springs ) were located between each

mass point. Subscript i denotes the

mass point number. The spring constant, ki, can be determined by Equation ( 1 ) below:

li Gi

ki= (1)

di where li = the length of layer i at

the mass point (center of gravity); di

= the distance between mass points;

Gi shear modulus of layer;

.

a for the sliding mass overlying

layer i is calculated by the following

equation:

Fi a = ..... (2)

NZm,

1-1 where Fi = the shearing force acting

on the interface between layers i and

(i-1), which is expressed by Fi=ki x

(Yi-yi-i) where yi is the displacement

of the mass point in layer;

.

SHAKEM, a computer code, is

modified version of SHAKE ( Schnabel et

al. ( 1972)) and was developed so that

it could be applied to a finite

ground. a" of the sliding mass determined by SHAKEM was computed from the

weighed average acceleration corresponding to the area of each layer

within the sliding arc. In the analyses by QUAD-4 and

FLUSH, on the other hand, a of the

sliding mass was computed from the

weighed average acceleration corresponding to the area of each element

with nodal points in common within the

sliding arc.

2.4. Input Earthquake Motions

As a rule, input ground motions

should be determined in accordance

with earthquake characteristics set

forth as a subject of design. Real

earthquake observation records on

bedrocks revealed that the frequency

characteristics of earthquake ground

motions differed with each event, even

if the magnitudes were on the same

level and the ground motions were

recorded at the same dam site. From

the viewpoint of conservative design,

different earthquake ground motions

with various frequency characteristics

should be taken into consideration

when the input earthquake motions for

analyses are selected. Therefore, in

the estimation of the seismic force

acting on the dam body, earthquake

response analyses were conducted using

different actual earthquake ground

motions with various frequency characteristics as input ground motions. In

this paper, 8 earthquake ground mo- tions with various frequency characteristics were selected from numerous actual earthquake records obtained

from the base rock of dam foundations,

based on the predominant frequencies

in the Fourier spectra of earthquake

ground motions. Table 1 summarizes

the characteristics of the earthquake

ground motions. Actual earthquake

ground motions were amplified so that

the maximum acceleration an input

earthquake motion, g„ax, becomes 0.2G

(G: gravitational acceleration). Fig.

4 shows the acceleration response

spectra for 8 input earthquake ground

motions computed at damping ratio of

20 %.

3. PRELIMINARY EXAMINATIONS

3.1. Effect of Vertical Motions

As a preliminary investigation,

the effect of the vertical component

of response acceleration, on the

calculation of a of a potential sliding mass, as well as the effect of the

vertical input ground motions was

examined. The effects of these vertical components were estimated by the

acceleration moment. Ma, of a potential sliding mass given by Equation

(3) below:

N Ma = i:{mi X (bhi X rvi + avi x rni )

}

1-1

N /Z mi ( 3 ) 1-1

104
where Bhi and avi are horizontal and

vertical accelerations respectively,

and rHi and rvi are horizontal and

vertical distances between the center

of an arc and the center of the element mass, mi. In addition, the

following 3 cases were established in

order to evaluate the influence of

vertical ground motions:

( 1 ) Ma is computed using time histories of the horizontal response acceleration obtained through dynamic

analyses, where only horizontal ground

motions are used as input ground

motions (Case (§))

.

(2) Ma is computed using time histories of the horizontal response acceleration obtained througli dynamic

analyses, where both the horizontal

and vertical ground motions are used

as input ground motions (Case®).

(3) Ma is computed using time histories of both the horizontal and vertical response acceleration obtained

through dynamic analyses, where both

the horizontal and vertical ground

motions are used as input ground

motions (Case ©)

.

Analyses of the model dam in Fig.

1(b) were carried out by QUAD-4, a two-dimensional analysis code, using

actual earthquake ground motions.

Maximum values of Ma within the time

duration of input ground motions at

the sliding mass were then determined.

The input ground motions used in the

analyses were those of Case 5 summarized in Table 1. The maximum acceleration of the vertical ground motions

was adjusted to 1/2 of the horizontal

maximum acceleration in these analyses, due to the fact that the ratio of

horizontal maximum acceleration to

vertical one ranges from 1/4 to 1/2 in

accordance with the earthquake ground

motions recorded at dam foundations.

Fig. 5 depicts the results of the

analyses. With respect to the effect

of the vertical ground motions, the

maximum average acceleration moments

obtained in Case @ are greater than

those obtained in Case (5) by about 4 %,

though the shapes of curves obtained

in both cases are similar. A comparison of the results of Case @ and Case

@ proves that the maximum average

acceleration moments in Case © are

smaller than those in Case @ by about

4.0 % at maximum for y/H < 0.6; this

relationship goes into reverse when

y/H reaches 0.7; and the maximum

average acceleration moments in Case @

become greater by about 10 % than

those in Case @ for y/H = 1.0. Judging from these results, the effects of

both the vertical input ground motions

and the vertical response acceleration

on Ma are estimated to be around 10 %

at most. Tamura and Ok amoto ( 1 9 8 7 ) conducted an evaluation of the effect

of vertical input ground motions on the stability of embankment dams

through both vibration table tests and

dynamic analyses with sinusoidal

waves. They concluded that the effect

was almost equivalent to an increase

of 10 to 20 % at horizontal ground

motions, when a dam is vibrated not

only by horizontal ground motions but

also vertical ground motions, and the

maximum acceleration of the latter

being a half of that of the former.

Based on the above facts, the

effects of vertical ground motions and

vertical response acceleration are

estimated to be negligible. In order

to simplify computation. Case @, in

which both ground motions and response

acceleration are exclusively horizontal, was selected for the following

examinations

.

3.2. Calculation Method for Average

Acceleration, ¥

In general, "a of a potential

sliding soil mass is computed from the

horizontal force acting on the soil

mass; however, it may also be computed

from the moment acting on the mass. These two methods may be compared

thus

:

( 1 ) 'a is computed so as to be equivalent to the moment of a sliding mass around the center of a circular_arc

of the mass. In other words, a is

obtained by the equation stated below

using the time history of horizontal

response acceleration, aHi, element

mass, mi, and the vertical distance

between the center of a circular arc and the center of gravity in each

element , rvi( Case (2))

:

N N a=2(mi X aHi x rvi)/I(mi x rvi )....( 4

)

1.1 1=1 (2) a^ is computed so as to be equivalent to force of a sliding mass. In

other words, a" is obtained by Equation ( 5 ) using aHi and mi of each

element (Case @) .

N N a=I(mi X aHi ) / 2mi (5)

1-1 1-1 As a result of the computation,

it was found that the values of a^ax obtained by Equations (4) and (5) were almost identical as illustrated in

Fig. 6. Thus, Equation (5) was used

to compute amax in the following examinations .

4. MAXIMUM AVERAGE ACCELERATION, amax Fig. 7 shows the relationship

between a^ax for a circular arc sliding

mass and y/H concerning each input

ground motion (Matsumoto et

al. (1991)). The more y/H increases,

i.e., the deeper the location of a circular arc becomes, the smaller the

values of "amax become, as shown in this

figure. The values of a^max are almost

equivalent to or smaller than the

105
disi et al. (1978) and Luan et

al.(1990). The maximum acceleration

curve obtained by the two-dimensional

dynamic analyses in this paper coincides with that of Luan et al. for the

most part, but differs slightly from

that of Makdisi et al . This is perhaps attributable to the difference in

the determination of the shear modulus

of the rockfill material, shear modulus was changed depending on the

confining stress in the analyses by

Luan et al . , while it was kept fixed

in the analyses by Makdisi et al . The

acceleration curve obtained by the

one-dimensional analyses is on the

left of that obtained by the twodimensional analyses, being roughly

parallel when the height of the circular arc is over 0.3. This is a relative discrepancy caused by the fact

that the values of Umax in the onedimensional analyses were great.

Makdisi et al.( 1978) and Luan et

al. (1990 ) express the acceleration

curves in the forms shown in Fig. 11

in order to estimate the sliding

displacement. However, these curves

are nothing more than average values;

the relationship between the frequency

characteristics of input earthquake

motions and the natural frequency of a dam body are omitted in their studies.

As shown in Fig. 7, an,ax greatly depends on such a relationship; when the

curve in Fig. 11 is applied to the

earthquake motions in Cases 2 and 3,

the values of a^ax at the elevations

between the middle and the bottom

become subject to underestimation.

Therefore, there is a possibility that

a noticeble error may result if amax at

respective elevations are simply

estimated using Umax without considering the facts mentioned above, despite

that the usage of an appropriate

method for applying an acceleration

curve is valid. The method to estimate the distribution of a^ax at each

elevation using an acceleration curve

is simple and convenient; however, the

above mentioned problems do exist.

Therefore, for practical use, a method

to compute the distribution of a^ax corresponding to the frequency characteristics of input earthquake motions

based on Fig. 7 normalized by gmax is

considered to work better in minimizing errors.

5. A SIMPLIFIED METHOD TO ESTIMATE

MAXIMUM AVERAGE ACCELERATION

aT^ax of a circular arc potential

sliding mass depends on the frequency

characteristics of input earthquake

motions, and the response of a mass differs according to the frequency

characteristics. A method of evaluating the frequency component, which

greatly influences a'^a, during actual

earthquakes, is examined below:

When a physical property R, such

as a displacement or stress at an arbitrary position on a structure, is

subjected to sinusoidal waves with

frequency f Hz at a unit amplitude,

R(f) is, in general, a frequency

response function, provided that R(f)

is expressed by a function of frequency, f. Accordingly, amax corresponding

to an arbitrary y/H was computed in

relation with numerous frequencies, f. Fig. 12 shows such frequency response

functions for the dam shown in Fig. 1 computed by QUAD-4 at three elevations, i.e., y/H = 0.0, 0.4 and 1.0.

The peaks of each a^ax of the three

potential sliding masses are close to

f/fo=1.0; however, the values of a^ax differ depending on the depth of the

potential sliding mass. The frequency

component which influences a'^ax is

defined to be within the range where

amax/gmax is greater than 1.0, although

it differs depending on the depth of

the potential sliding mass. In order

to simplify the computation, however,

the frequency range was determined

based on Fig. 12, which was expressed

by Equation (6), regardless of the

depth of the potential sliding mass. Because low natural frequencies up to

the 5-th or 6-th mode govern the

response of an embankment dam during

earthquakes, the upper limit of f/fo was considered to be sufficiently 2.0,

while the lower limit was set at

f/fo=0.8, where the magnification of

responses of each sliding mass is

reduced to half of that for f/fo=1.0.

0.8 < f/fo < 2.0 (6)

As described above, Makdisi

proposed a method which gives a design

value of ¥rr,ax: when an acceleration

response spectrum of the earthquake

motion is given as a load condition,

it is used to estimate u„ax, which is

further combined with the acceleration

curve shown in Fig. 11 in order to

obtain the value of a^ax. On the other

hand, when the existence of a frequency response function of a^ax along with

its characteristics as shown in Fig.

12 is taken into consideration, a^ax can be estimated directly with high

accuracy by the acceleration response

spectrum given as the load condition.

In short, when the acceleration re- sponse spectrum of input earthquake

motions is given, the mean value of

the acceleration response spectrum

within the frequency range given by

Equation (6), Sam. is computed; should

a correlation between a„ox computed by

dynamic analyses and Sam exist, ¥max can

be easily computed by Sam- The values

of Sam were computed based on Fig. 4 using the 8 earthquake motions in

Table 1 as input motions. Fig. 13

illustrates the relationship between

Sam and amax • In addition. Fig. 14 shows the

106
results obtained by modal analysis

regarding the effect of dam heights

(natural periods) on a'^ax of a potential sliding mass. This figure shows

that there is only a minor difference

in anax/Sam CO r r 6 s pond 1 ng to arbitrary

y/H, even when the height of a dam

varies

.

The above results were obtained

by analyses under the damping ratio

of 20 %. Acceleration response spectra were computed separately for

h=10%, and the relationship between

a^max/Sam and y/H were obtained and

compared with that for h = 20 % as shown in Fig, 15. In this figure, Ro.i and Ro.2 is correlation coefficient of

fluctuation of Imax / gmax dgainst re- gression line at the damping ratio of

10% and 20% respectively. The calculation is conducted for 8 earthquake

motions in Table 1. In accordance

with the results obtained by twodimensional analyses using QUAD-4

shown in this figure, there is only a slight difference in the relationships

of these cases. The following equation can be derived as an approximation:

a„ax= {2.0 - 1.35 (y/H)} x Sa. (7)

Based on the above results, anax of a potential sliding mass can be

estimated by the following methods:

(1) The acceleration response spectrum

of input earthquake motions is computed, provided that the damping ratio to

be used is that of the dam concerned.

(2) The natural frequency of the dam

is computed.

(3) The mean value of acceleration

response spectra is computed within

the range given by Equation (6).

(4) The values of a^ax corresponding to an arbitrary elevation are computed by

substituting Sam into Equation (7) or

by using Fig. 15.

6. ESTIMATION OF PERMANENT DEFORMATION

AND LOADING STRESS IN CYCLIC LOADING

TESTS

When the critical acceleration of

an arbitrary potential sliding mass is

ay and the maximum average acceleration of the sliding mass due to input

ground motions is a„ax, the amount of

permanent deformation. Up, is not

generated when ay/a„ax > 1.0, but when

ay/ama. < 1.0. There is a close corre- lation between Up and ay/a^ax. The

charts to estimate Up from ay/a^ax were proposed by Makdisi et a2.(1978) and

Luan et al. (1990). This idea is basically an extension of the method

proposed by Newmark (1965). This

paper deals with a method to estimate

the extents of permanent deformation

by laboratory tests using a„ax. Ishihara(1980) proposed a method by which

the extent of permanent deformation is

determined through laboratory tests

under a stress condition, which is

derived by the static and dynamic

shear stresses obtained from the

safety factors of a potential sliding

mass before and during earthquakes

calculated by the simplified slice

method. This paper also follows this

concept, except for the fact that

dynamic shear stresses are estimated

directly using amax. When the moment

due to a static load acting on a sliding mass before an earthquake is

Ms, the mean shear stress, Ts, acting on

a potential sliding mass due to a static load can be given by the following equation, where r is the radius

and 1 the length of the circular arc:

Ms

Ta= (8)

r X I On the other hand, the moment,

Md, of a dynamic load can be obtained

from amax/* the mean shear stress acting on a potential sliding mass,Td, is as follows

Md

Td= ( 9 ) r X I The specimen of banking material

is consolidated by the average confining stress, Om; then Ob is applied to

the specimen under a drained condition. Finally, cyclic shear stress of

+Td is applied under an undrained

condition; then the cumulative shear

strain summed up in the specimen is

obtained. This cumulative strain

enables the estimation of deformation

of embankment dam during earthquakes.

It is desirable to apply the cyclic

shear stress, the wave form of which

corresponds to the time history of a; however, aeq is used instead of a„ax when cyclic shear stress with a con- stant amplitude is applied. Fig. 16

shows the results obtained by computation regarding the relationship between amax of a potential sliding mass

on the upstream side of the dam model

shown in Fig. 2(b) and the shear

stress of the sliding surface, x(Tb +

Td ) . The value of t for amax = 0 in

Fig. 16 coincides with Ts calculated

by Equation ( 8 ) ; the range of Xd at Xd

< Xa is shown by a dotted line. From

this figure, it is recognized that a stress reverse does not occur, due to

the fact that Xa is smaller than x, in

the range where a„ax does not exceed

0.2G. By the use of Fig. 16, the load

condition for determining permanent

strain through laboratory cyclic

loading tests can be established with

high accuracy.

Fig. 17 shows the relationship

between Up related to 10th cyclic

loading and Xa obtained by the cyclic

loading tests for the rockfill material(matsumoto et al. (1991)). The permanent strain corresponding to 10 th

cyclic loading. Up, can be evaluated.

107
by calculating Xs and Xd coping to a potential sliding mass using Fig. 16, or Equations (8) and (9) and by applying these Xb and Xd to Fig. 17. However, Fig. 17 is obtained for 6.0

kgf/cm2 of mean confining stress, o„. It is necessary that curves in Fig. 17

should be drawn for several kinds of

confining stress at the laboratory

test, because shear strength of rockfill materials has an inclination to

depend on confining stress.

7. CONCLUSIONS

The effect of different earthquake ground motions with various

frequency characteristics on the

seismic force acting on embankment

dams was evaluated by 4 methods of

dynamic analysis. In addition, a simplified procedure to estimate a^ax was examined. The conclusions of this

paper are summarized below:

(1) Even if amax is computed using

only horizontal input earthquake

motions and the horizontal component

of response acceleration of a dam

body, the errors in a„ax are around 10

% at most compared with the case where

both horizontal and vertical components of motions are taken into con- sideration in input and response.

Thus, a'max can be computed by the

horizontal components alone.

( 2 ) There are two methods to compute

amax: one to compute it from the

moment of the potential sliding mass concerned and another to compute it

from the horizontal resultant force of

the sliding mass. The difference in

evaluation of "a^ax obtained by these

two methods is about 5 % at maximum.

Thus, amax can be conventionally computed by the horizontal force alone.

(3) Distributions of a^ax due to the 8

different input earthquake motions

with various frequency characteristics were computed by 4 analysis methods.

As a result, it was pointed out that

the one-dimensional analyses, such as modal analysis and SHAKEM, gave values

of "a^ax at dam crest about 15 % larger

than those computed by the two-dimensional analyses on the average of

results due to 8 input earthquake

motions. In addition, the discrepancy

of the values of a„ax due to 4 different analyses was less than 10 % when

y/H is greater than 0.3. This discrepancy, attributed to the difference in

analytical methods, was entirely

small, except for the fact that the

responses near the dam crest obtained

by one - d i mens i on a 1 analyses were

comparatively large.

(4) Even when g„ax is the same, the

value of a„ax fluctuates +50% due to

the frequency characteristics of the

input ground motions on the basis of

the average of 8 different responses.

Therefore, the frequency characteristics of earthquake motions expected to

occur at a dam site should be taken

into consideration when the average

values of the acceleration curve

proposed by Makdisi et al .

, which was normalized using Umax, is employed for

practical use.

( 5 ) When an acceleration response

spectra of input ground motions are

given, a^ax can be easily obtained by

the use of either Equation (7) or Fig.

15, based on the computation of the

average values, Sam, for the Fourier

spectrum within the frequency range

given by Equation ( 6 ) corresponding to

y/H.

(6) Laboratory cyclic loading tests

for embankment material under the

shear stresses derived from static and

seismic forces, which are calculated

by Equations (8) and (9) using "Hmax, can contribute to greater precision in

estimating the extent of permanent

deformation to a dam body due to

seismic forces.

REFERENCES

N. M. Newmark, "Effects of Earthquakes

on Dams and Embankment, " Geotechnique,

Vol.5, No. 2, pp. 139-160, 1965.

H. B. Seed, "A Method for EarthquakeResistant Design of Earth Dams, " Jour,

of SMFD, Vol.92, No. SMI, pp. 13-41,

1966.

H. B. Seed and G. R. Marin, "The

Seismic Coefficient in Earth Dam

Design," Jour, of SMFD, ASCE, Vol.92,

N0.SM3, pp. 25-58, 1966.

N. N. Ambraseys and S. K. Sarma, "The

Response of Earth Dams to Strong

Earthquakes," Geotechnique, Vol.17,

pp. 181-213, 1967.

N. Mononobe, A. Tanaka and M. Matsumura, "Seismic Stability of the Earth

Dams," 2nd ICOLD, pp. 435-444, Vol.4,

1936.

A. K. Chopra, "Earthquake Response of

Earth Dams," Jour, of SMFD, ASCE,

Vol.93, NO.SM2, pp. 65-81, 1967.

F. I. Makdisi and H. B. Seed, "Simplified Procedure for Estimating Dam and

Embankment Earthquake-Induced Deformations," Jour, of GTD, ASCE, Vol.104,

NO.GT7, pp. 849-867, 1978.

F. I. Makdisi and H. B. Seed, "Simplified Procedure for Evaluating Embankment Response, " Jour, of GTD, ASCE,

Vol. 105, NO.GT12, pp . 1427- 1434 , 1979.

M. T. Luan, C. P. Jin, G. Lin and H. Takemiya, "Parametric Studies on

Dynamic Stability and Permanent Deformations of Cohesive Embankment,"

108
Proceedings of the Eighth Japan Earthquake Engineering Symposium, Vol.2,

pp. 975-980, 1990.

N. Matsumoto, N. Yasuda, M. Ohkubo and

Y. Kinoshita, "Shear Strength and

Dynamic Deformation characteristics of

gravels", Proc. of JSCE, Vol.424, III- 14, pp. 95-104, 1990. (in Japanese)

N. Matsumoto, M. Toyota and M. Shiga,

"Model Vibration Tests of Rockfill

Dams," Civil Engineering Journal,

Vol.25, No. 6, pp. 45-50, 1983. (in

Japanese

)

T. Ohmachi and N. Nakamoto, "In-Situ

Measurement of Radiation Damping of

Existing Rockfill Dams, " Proceedings

of 9th World Conference on Earthquake

Engineering, [6], pp. 337-342, 1989.

Ashok K. Chugh, "Dynamic Response

Analysis of Embankment Dams, " International Journal for Numerical and

Analytical Methods in Geomechanics

,

Vol.9, pp. 101-124, 1985.

I. M. Idriss, J. Lysmer, R. Wang and

H. B. Seed, "QUAD-4, A Computer Program for Evaluating the Seismic Stability of Soil Structures by Variable

Damping Finite Element Procedures,"

EERC-Report No. 73-16, University of

California, 1973.

J. Lysmer, T. Udaka, C. F. Tsai and H. B. Seed, "FLUSH - A Computer Program

for Approximate 3-D Analysis of Soil

Structure Interaction Problems, " EERCReport No. 75-30, University of California, 1975.

S. Okamoto, "Dynamics for Civil Engineers (Second Edition)," Ohmupublication, pp. 57-66, 1976. (in

Japanese

)

P. B. Schnabel, J. Lysmer and H. B. Seed, "SHAKE, A Computer Program for

Earthquake Response Analysis of Horizontally Layered Sites," EERC-Report

No. 72-12, University of California,

1972.

N. Matsumoto, S. Kondo, H. Katahira

and M. Shiga, "Earthquake Ground

Motions on the bedrock of Dam Sites."

Technical memorandum of Public Works

Research Institute, No.l7B9, 1982. (in

Japanese

)

C. Tamura and S. Okamoto, " On effects

of vertical ground motions on stability of Fill-type dams, " International

symposium on earthquakes and dams,

ICOLD, pp. 486-500, 1987.

N. Matsumoto, N. Yasuda and K. Yamabe,

"Evaluation of Seismic Force Acting on Embankment Dams, " Technical memorandum

of Public Works Research Institute,

No. 2997, 1991. (in Japanese)

S. Okamoto, "Introduction to Earthquake Engineering," Ohmu-sha,

p. 385, 1971. (in Japanese)

T. Sawada, "Behavior of Fill-type Dam

during Earthquakes, Case Study of

Namioka Dam," Proc. of Japan Acad. 62,

- Ser.B, pp. 201-204, 1986.

M. Annaka, "Earthquake Observation of

Fill-type Dams and 3-D Mode-Superposition Analysis using Three-directional

Input Earthquake Motions," Large Dams,

No. 134, pp. 22-34, 1990. (in Japanese)

T. Hasegawa, K. Uchida, M. Kikuchi and

A. Murakami, "Analysis on the observed

seismic response data of Miyama Dam

and estimation of its dynamic response

behavior," Trans, of JSIDRE, No. 108,

pp. 55-63, December, 1983. (in Japanese )

T. Sawada, T. Hasegawa and M. Kikuzawa, "An application of the proposed

complex damping Model to the dynamic

analysis of an actual dam," Trans, of

JSIDRE, No. 28, pp. 58-64, August, 1979.

( in Japanese

)

K. Ishihara, "Current practice and the

problems in the earthquake resistant

design of earth structures, " Soils and

Foundations, No. 28-8, pp. 3-8, 1980.

( in Japanese

)

N. Matsumoto, N. Yasuda, M. Ohkubo and

R. Yoshioka, "Monotonies and Cyclic

Loading Tests of Rockfill Materials,

Technical memorandum of Public Works

Research Institute, No. 2996, 1991. (in

Japanese

)

H.Watanabe, S.Satoh and K.Murakami,

"Evaluation of Earthquake- Induced

Sliding in Rockfill Dams," Soils and

Foundations, Vol.24, No. 3, 1984

109
400

(a) One-dimensional

(b) Two-dinfiensional

Fig. 1 Finite element idealization used in the analysis

1.0

II>- ^5o«

_lQO

s

cc2 0.5 X

V)a

N

10

D

>

zo

33 5 §o

10-5 2 5 10-" 2

SHEAR STRAIN, y

10-3

Fig. 2 Strain dependent curves of shear modulus

and damping ratio used in response computations

Fig. 3 Shear beam model with lumped mass (Modal Ana lys i s)

<

cc

I-olU

D.V>

Ui

</) zo

D.(A

UJ

CC

<

UJ

_l

Uio

o

<

300-

200-

100

0.4 0.8

PERIOD, T (sec) Fig. 4 Acceleration response spectra for input

ground motions

0

0.2

0.4

0.6 -

0.8 -

25,000 30,000 35,000 40,000 45,000 50,000

MAXIMUM AVERAGE ACCELERATION MOMENT (gal-m)

Fig. 5 Maximum average acceleration moment 0

100

Fig.

200 300 400 500 600

"amax (gal)

6 Maximum average acceleration computed

110
0 100 200 300 400 500 600 700 800

awax (gal)

(a) Modal Analysis

0 100 200 300 400 500 600 700 800

a'max (gal)

(b) SHAKEM

0.2

0.4 -

0.6

0.8 -

CASE 1 A CASE 2 O CASE 3 3 CASE 4 CASES

CASE 6 • CASE 7 A CASES

0 100 200 300 400 500 600 700 800

a"max (gal)

(c) QUAD-4

0 100 200 300 400 500 600 700 800

a'max (gal)

(d) FLUSH

Fig. 7 Maximum average acceleration

0.2

0.4

0.6

0.8

MODAL

ANALYSIS SHAKEM

QUAD-4

FLUSH

0 100 200 300 400 500 600

a'max (gal)

Fig. 8 Relationship between y/H and Tm.x for eight input notions

111
0.2

0.4

0.6-

0.8

MOOAL

ANALYSIS SHAKEM o OUAD-4 • FLUSH

0.4

0.6 -

0.8

T

<—

r

CASE 1 A CASE 2 O CASE 3 3 CASE 4 CASES

CASE 6 • CASE 7 CASES

0.5 1 1.5 U(y/H)max/gmax

2.5 0 100 200 300 400 500 600 700 800

ij (y/H) max (gal)

(a) Normalized maximum acceleration from different (b) Maximum acceleration from QUAD-4

analytical methods

Fig. 9 Maximum acceleration at various elevations

0| r-i>- 1

H

CO

UJ

tro

s

<

o

o

zo

o

Eza

<

20 50

gmax (gal) 200

RANGE OF

ALL DATA AVERAGE OF

ALL DATA

Kuzuryu

(Sep/09/19e9) o

Miho •

Miho

(Apf/14/1981)

©

Gottho (Dec/02/ 1981) « °oiay/23/1983) H

TokacM

(M«»/2«/1983) B

0d»9awa

(Miy/56/1983)

;a"

FukKla

(JJ/23/1982) Myama

(AU9/6/1977)

Slittokl (J«i/9/t987) ©

ShitoU

(Jul/3/1983)

(a) Maximum acceleration magnification ratio

at dam crest with maximum ground acceleration

3 4 5 6 7

U(y/H)max/gtnax

(b) Maximum acceleration distribution with

elevations

Fig. 10 Acceleration response measured at rockfil

dams

112
0.2

0.4

0.6

0.8

1 1

1

1 1

1 1 ' o

MODAL

ANALYSIS A SHAKEM • OUAD-4 A FLUSH

MAKDISI et al

LUAN et al.

0 0.2 0.4 0.6 0.8 1 3max/ Umax

Fig. 11 Comparison of normalized maximum acceleration curve with depth of sliding mass

5.0

Fig. 12 Frequency response functions of a calculated by QllAD-4

2.0

Fig. 13 Relationship between acx/Qm.. and

S ../Qm.x

for three different dam-heights

0

0.2

0.4

X\

0.6

—I—f~ry—I r

Rai= 0.983' /Ro 2 =0.980

Roi=0.987^ p Ro.2=0.984

Ro.,-0.987iK j^j = 0.981

0.8

Ro.2= 0.966

Rai=0.980«Ra2=0.965

Ra.= 0.978

o h=0.1 • h=0.2

Ro2=0.973//Ro,= 0.981

IJ:

R ; CORRELATION

COEFnCIENT

1 _ 2 3 4

3 max /gmax

Fig. 15 Relationship between am.x/S.m and y/

for h = 10!^ and 20% (QUAD-4)

3.0

2.0 +

in

cc

Icn

ccX

1.0

1 1 ro y/H = 0.2 A y/H = 0.6 y/H = 1.0

NON- REVERSAL

0.0

Fig. 16 Shear stresses estimated by using a «.,

113
Table 1 Characteristics of eartliquaice notions

observed at dam foundation

Casa 0AM site Earthquake M L(KM) f(Hl) f/fo

1 Gosho Mid Japan Sea 7.7 198.0 0.59 0.35

2 Mlnaea Mlyagl offshore 7.4 167.0 0.78 0.46 3 YoKoyama Central GIfu 6.6 60.0 1.07 0.63 4 TokachI ^4orthefn HIdaka Mts. 7.0 62.0 1.61 0.95

5 Taft« Mid Japan Sea 7.7 264.0 2.39 1.41 6 Gosho Cenbal fwale 6.6 69.0 3.20 1.88 7 Mlho l2u penlneula offehofe 6.7 56.0 4.15 2.44 8 Mlho Western Kanagawa 5.8 12.0 6.59 3.88

(Note) M: Magnitude L:Eplcentral distance

fo: Natural frequency of model dam f : Predominant frequency

114
Resolving Uncertainties in Seismic Hazard Models

by

:- J. K. Kimball*

ABSTRACT

This paper describes the types of

uncertainties inherent in seismic

hazard analysis, and identifies those

steps that can be taken to resolve

these uncertainties. The focus is on the recent advances that have been

made in performing probabilistic

seismic hazard studies, with

particular emphasis on issues related

to identifying uncertainties in the

basic seismic hazard models.

Comprehensive probabilistic seismic

hazard studies have been completed

over the past five years by Lawrence

Livermore National Laboratory for the

United States Nuclear Regulatory

Commission and by the Electric Power

Research Institute for the commercial

utility industry. The United States

Department of Energy is actively using

these studies to evaluate seismic

hazard at Department of Energy sites.

Results from these studies can be used

to quantify existing uncertainties in

seismic source understanding,

earthquake occurrence rates, and

ground motion attenuation. The

resulting uncertainties can then be

evaluated to determine those that are reducible with additional study.

Additionally, a clear understanding of

modeling uncertainties can provide

insight into the attributes that

future probabilistic seismic hazard

analysis should have.

KEYWORDS: seismic hazard; seismicity;

seismic sources; uncertainty

evaluation; ground motion

1. INTRODUCTION

The purpose of this paper is to

provide a summary description of the

types of uncertainties inherent in

seismic hazard analysis, and to

identify those steps that can be taken

to resolve these uncertainties. This

purpose will be accomplished by

describing how the Department of

Energy (DOE) is utilizing

probabilistic seismic hazard results,

including the identification and

evaluation of uncertainty. The

general purpose of a probabilistic

seismic hazard analysis is to evaluate

the vibratory ground shaking at a particular site by considering the

viable sources of earthquakes, the

likelihood of earthquake occurrences

for each source, an estimate of the

particular ground motion parameter of

interest, and an analytical model for

estimating the probability of

exceeding the ground motion parameter.

These general steps are shown on Figure 1. Each of the general steps

will be discussed below, in terms of

modeling uncertainty.

The DOE currently specifies the

loading associated with the Design

Basis Earthquake (DBE) based on probabilistic acceptance criteria.

Mr. J. Hill, DOE, presented the DOE

approach at the twenty-third annual

joint meeting of the United

States/Japan Panel on Wind and Seismic

Effects, May 1991. The DOE facilities

range from office buildings to complex

* U.S. Department of Energy, Office of

Engineering and Operations Support,

Washington, D.C. 20585

115
nuclear facilities. The design basis

earthquake (DBE) acceptance criteria

represents a range of probabilistic

input levels from those associated

with standard building codes (about

1/500 or .002 per year) to those

associated with nuclear reactors

(about 1/1,000 to 1/10,000 or .001 to

.0001 per year). Because the DBE is

specified probabilistically, the

values selected are associated with mean estimates, to account for

uncertainty in seismic hazard.

Because the DBE is probabilistically

derived, particular attention has been

focused on the types of approaches and

inputs that have gone into existing

seismic hazard analyses. Separate

comprehensive studies were completed

by Lawrence Livermore National

Laboratory (LLNL) for the U.S. Nuclear

Regulatory Commission (NRC) and by the

Electric Power Research Institute

(EPRI) for the commercial utility

industry. These studies have been

used at a number of DOE Eastern United

States sites to establish the DBE

seismic hazard curves. The LLNL and

EPRI results can be used to highlight

the strength of probabilistic results

along with potential shortcomings that

must be considered by decisions makers

when using such results. Table 1 provides a brief outline of how the

LLNL and EPRI studies were developed.

For the reader who is interested in

background material regarding

probabilistic seismic hazard analysis,

two references are particular useful.

The first is a National Research

Council report from the Panel on Seismic Hazard Analysis (Committee on Seismology) on Probabilistic Seismic

Hazard Analysis, 1988. The second is

the Probabilistic Safety Analysis

Procedures Guide, NUREG/CR-2815, Rev.

1, 1985, which describes how the

various types of seismic hazard input

should be characterized, including the

treatment of uncertainties. The

summary references for the LLNL and

EPRI studies are D.L. Bernreuter et

al , 1987, and Electric Power Research

Institute, 1986. Two useful

references regarding seismic source characterization are Coppersmith,

1991, and Thenhaus et al , 1987.

2. SOURCES ON UNCERTAINTY IN SEISMIC

HAZARD ANALYSIS

It is important in any probabilistic

seismic hazard analysis to make an estimate of the confidence in hazard

results. In the LLNL and EPRI seismic

hazard studies this was accomplished

by providing constant percentile

seismic hazard curves (such as 15th,

50th and 85th percentiles) and

estimates of mean hazard. Figure 2 shows an example of such results for

the DOE Savannah River Site in South

Carolina. Several observations are noteworthy. First, the median (50th

percentile) results are similar

between the two studies. Second, the

overall estimates of uncertainty are quite different between the two

studies, which result in drastic

differences between mean estimates of

seismic hazard. Because DOE is

attempting to use results such as these to establish the DBE, it has

been important to understand why the

two studies give such divergent

estimates of uncertainty.

A convenient way of addressing seismic

hazard uncertainty is to segregate

uncertainty into two components. For

the purpose of this paper these two

components are labeled randomness and

modeling uncertainty. Randomness

represents variability that cannot be

reduced with more data or better

models. Modeling uncertainty

represent variability that can be

reduced (in concept) with better data

and analysis techniques. The

discussion provided below will

identify modeling uncertainty issues

using the LLNL and EPRI results. The

116
type of data that could be used to

address such modeling uncertainty will

also be identified. Such an observation has generic implications

for any probabilistic seismic hazard

analysi s

.

2.1 Seismic Source Uncertainty

In the LLNL and EPRI studies each of

the experts and expert teams were requested to provide seismic source maps which generally represent areas of uniform earthquake potential.

Figure 3 shows examples of such maps

from two of the LLNL experts for the

southeastern United States. The

alternatives shown on the two maps

represent very different opinions

regarding the identification of

seismic sources. Such divergent

alternatives represents the ability of

earth scientists to identify active

seismic features (modeling

uncertainty). Presently in intraplate regions such as the Eastern

United States, the identification of

specific earthquake causal structure

remains elusive. It could be expected

that in more active tectonic regions,

such as plate margin regions in

California and Japan, earth scientists

can more readily define the seismic

sources. Such an assumption, however,

may not be totally warranted given the

occurrence of potentially damaging

earthquakes such as the 1983 Coalinga,

California earthquake. Thus, the

current state of practice should

recognize the potential large

uncertainty in seismic source identification.

Since modeling uncertainty is

potentially reducible, it is important

to identify the type of data that

should be assessed to reduce

uncertainty in seismic sources. This

data includes:

0 Analysis of historic seismicity;

0 Accurate instrumental seismic

recordings

;

0 Geologic assessment of paleoseismic features (such as fault

scarps and liquefaction

features); and

0 Integration of

geologi c/sei smol ogic/geophysical

data to provide a three

dimensional picture of the

seismogenic crust.

As this type of data evolves, one can expect the uncertainty in seismic

sources such as shown on Figure 3 to

diminish.

2.2 Earthquake Occurrence Rate

Uncertainty

In the LLNL and EPRI studies each of

the experts and expert teams were

requested to provide estimates of

earthquake activity rates (commonly

termed the "a" value), slopes of the

earthquake occurrence curve (commonly

termed the "b" value), and estimates

of largest magnitude. Figure 4

displays the earthquake occurrence

curves from the LLNL and EPRI studies

for the Charleston, South Carolina

seismic source, for the best estimate

earthquake recurrence parameters ("a"

and "b" values). Similar to seismic

sources estimation, there is wide

divergence in earthquake recurrence

estimates. This divergence is even more dramatic if the uncertainty in

"a" and "b" values is reviewed.

Should such divergence be expected?

The answer to this question is not

simple considering that the earthquake

occurrence information represents both

randomness and modeling uncertainty.

In attempting to assess whether the

estimates of uncertainty on earthquake

occurrence rates are realistic, expert

117
opinion studies must include

sufficient feedback and interaction to assure that input has a scientific

basis. Clearly seismicity experts are not necessarily uncertainty experts.

To illustrate this problem, the LLNL

data for the DOE Savannah River Site

has been evaluated in detail. One of

the seismicity experts gave a range of

recurrence intervals for magnitude 5 earthquakes for the Charleston, South

Carolina, seismic source as 20 days to

over 2000 years. Such dramatic

examples may illustrate more of the

problem of expert assessment of

uncertainty than true modeling

uncertainty.

The modeling uncertainty associated

with earthquake occurrence rates, like

seismic sources, is potentially

reducible. The type of data that

should aid this process includes:

0 Quantification, on a consistent

basis, of the size of historic

earthquakes compared to

instrumental seismicity;

0 Integration of geologic slip rate

from paleo-seismic studies with

seismicity data; and

0 Assessment of short term strain

rates in active tectonic regions.

While this type of data should aid in

reducing modeling uncertainty in

earthquake occurrence rates, it may be

more important to ensure that earth

scientists better understand the

process of uncertainty assessment.

2.3 Ground Motion Attenuation

Uncertainty

In the LLNL and EPRI studies separate

efforts were completed to develop

ground motion attenuation

relationships. In the LLNL study five

attenuation experts selected

attenuation models and provided

uncertainty estimates. In the EPRI

study one expert selected three

attenuation models with one estimate

of attenuation uncertainty. Figure 5 displays a comparison of predicted

peak ground acceleration from the LLNL

and EPRI studies for magnitudes 5 and

7. Figure 5 also shows wide

divergence of opinion, similar to

other components of seismic hazard

input. In a region such as the

eastern United States where there are few available strong motion records

one might expect relatively large peak

acceleration modeling uncertainty. In reviewing Figure 5, however, one sees

that the modeling uncertainty

predicted by the experts spans over an order of magnitude at the distance of

one hundred kilometers. When such

divergence exists, the user should

require that each model selected be

clearly justified. The LLNL seismic

hazard is particularly sensitive to one of the attenuation models at rock

sites. Much effort has been expended

by other experts in an attempt to show

that this model is inappropriate.

While DOE and the NRC have not agreed

with this conclusion, decisions made

using the LLNL and EPRI results have

had to recognize the sensitivity of

this issue.

The type of data that should be used

to reduce such modeling uncertainty

includes:

0 Collection and analysis of strong

motion data near damaging

earthquakes

0 Assessment of crustal attenuation

properties both on a local and

regional basis;

0 Evaluation of earthquake source characteristics such as stress

drop; and

118
0 Assessment of the effects of local

site conditions including the

impact of soil consolidation at

deep soil sites.

2.4 Overall Summary of LLNL/EPRI

Eval uation

Based on the evaluation of the LLNL

and EPRI data DOE has reached the

following general assessment of these

two studies with respect to modeling

uncertainty. This assessment provides

insight into the types of attributes

that any probabilistic seismic hazard

should have. There is a high degree

of similarity between the LLNL and

EPRI seismic hazard studies ranging

from overlap of experts used to

general overlap in parameter input

from the experts. The key difference

between the two studies relates to the

topic of uncertainty assessment,

particularly modeling uncertainty

assessment for all input variables.

Identified issues relate to the

process of expert opinion elicitation,

particularly the issues of how and

whether experts assess and understand

uncertainty. This puts increased

emphasis on the user or decision maker

to fully understand the uncertainties

inherent in the probabilistic results.

The DOE review has also found that

there are several stable trends in the

existing LLNL and EPRI data. These

trends include:

0 General consistency in median

probabilistic estimates of seismic

hazard;

0 General consistency within each

study of the ratio of 85th

percentile hazard curves to the

median hazard curves;

0 General consistency between each

study in the ratio of 85th

percentile hazard curves to the

median hazard curves; and

0 Relative consistency in hazard

trends; studies provide similar

ranking of high seismic hazard to

low seismic hazard sites.

The DOE has used the above trends to

develop an interim position regarding

how the two studies should be used at

DOE sites. This position is built on the stable median hazard curves with a "correction factor" to account for

uncertainty. The correction factor

will be used to develop mean estimates

of seismic hazard. The "correction

factor" is based on the above

identified trends, and was estimated

to have a value of 1.65 for peak

ground acceleration at probabilities

of about 1/5000 or .0002 per year.

3. ATTRIBUTES OF AN ADEQUATE

PROBABILISTIC SEISMIC HAZARD ANALYSIS

Given the types and the order of

magnitude of modeling uncertainties

described above, it is important to

outline the attributes that any

probabilistic seismic hazard analysis

should contain. It is particularly

important that any decision maker who

is using a probabilistic seismic

hazard analysis require that these

attributes be properly addressed.

These attributes include:

0 Clear documentation of input

including modeling assumptions and

how seismicity parameters were

selected. It is important to

understand the basis for the

seismicity parameters {geologic

versus seismologic) and attenuation

model s

0 Comparison of model input to

historic seismicity. While there

should not necessarily be an over- reliance on historic seismicity,

this type of comparison can identify issues which may otherwise

be overlooked;

119
0 Incorporation of alternative models

that represent the current stateof- knowledge. Hazard analyses

without this attribute can lead to

a false sense of security by not

recognizing viable alternatives.

It is important, however, that a basis be provided for all models

used;

0 An analytical approach that allows

the user to understand dominate

seismic hazard contributors versus simply showing the bottom-line

hazard curve. For many

applications it is important to

know which sources, magnitudes and

distance most contribute to the

seismic hazards; and

0 Clear documentation and

presentation of uncertainties.

If the above attributes are addressed

by a probabilistic seismic hazard

analysis, then the hazard results are likely to provide consistent estimates

from site to site. Even considering

the large absolute differences in

results such as LLNL and EPRI, it is

at least reassuring to know that both

studies provide consistent trends of

seismic hazard.

4. CONCLUSIONS

The state-of-knowledge regarding the causes of earthquakes in much of the

United States is limited, resulting in

large uncertainties in seismic hazard

estimates. Probabilistic seismic

hazard studies must clearly describe

and document this uncertainty to be

useful for decision makers. Continued

improvements regarding the assessment,

quantification and documentation of

expert opinion is needed before the

actual probabilistic numbers can be

used in an absolute sense. At the same time, continued study regarding

the seismic and ground motion input

should aid in reducing modeling

uncertainties. Such efforts fall

within the following major areas:

0 Continued geologic, seismologic,

and tectonic study of the causes,

mechanisms and rates of crustal

deformation and the validity of

alternative seismicity models;

0 Evaluation of instrumental seismic

data including the variability of

earthquake occurrence rates; and

0 Enhanced understanding of strong

motion characteristics, regional

attenuation properties and the

effect of local surficial material.

5. REFERENCES

Bernreuter et al , 1987, Seismic Hazard

Characterization of 69 Nuclear Plant

Sites East of the Rocky Mountains,

LLNL, NUREG/CR-5250, Vols. 1-7.

Coppersmith, 1991, Seismic Source

Characterization for Engineering

Seismic Hazard Analyses, Proceeding

Fourth International Conference on Seismic Zonation, Stanford University,

Stanford, California

Electric Power Research Institute,

1986, Seismic Hazard Methodology for

the Central and Eastern United States,

EPRI NO-4726, Project PlOl-21.

McCann et al , 1985, Probabilistic

Safety Analysis Procedures Guide, BNL,

NUREG/CR-2815, Sections 8-12.

National Research Council, 1988,

Probabilistic Seismic Hazard Analysis,

Panel on Seismic Hazard Analysis

Committee on Seismology, National

Academy Press, Washington, D.C.

Thenhaus, et al , 1987, Earthquake

Hazard in the Eastern United States:

Consequences of Alternative Seismic

Source Zones, Earthquake Spectra, Vol.

3, No. 2.

120
TABLE 1

Summary of LLNL and EPRI Study

LLNL STUDY:

Used individual expert opinion to

evaluate all seismic hazard input.

1. 10 Seismicity experts provided

zonation, earthquake occurrence

rates and upper magnitude values

including estimates of uncertainty.

2. 5 Attenuation experts provided

attenuation models and model

uncertainty.

3. The LLNL used Monte Carlo methods

to evaluate overall seismic hazard

uncertainties. Each site included

2750 seismic hazard curves, which were processed into constant

percentile seismic hazard curves.

EPRI STUDY:

Used expert teams to evaluate seismic

hazard input except for attenuation

models.

1. 6 Expert teams provided zonation,

earthquake occurrence rates and

upper magnitude values including

estimates of uncertainty.

2. One expert provided attenuation

models and model uncertainty.

3. EPRI used logic trees to evaluate

overall seismic hazard uncertainty,

which were processed into constant

hazard seismic hazard curves.

121
NPP

0"

V

Sources

%

«• E ^ II

C (Q c «

'o9loN"ao-bo"ii

Magnitude, nrij

Recurrence

Eff. peak ground

acceleration, a

ition.

a

Increasing

, magnitude

w

SL

«>u

u<

Distance, R

Seismic hazard Attenuation

Figure 1 Four steps of a probabilistic seismic hazard

analysis: Step 1 Identification of seismic

sources; Step 2 Develop earthquake recurrence

relationship for each seismic source; Step 3

Develop ground motion attenuation model;

Step 4 Use an analytical model to calculate

seismic hazard curve including uncertainty.

122
A

N

N

U

A

L

P

R

O

B

A

B

IL

IT

Y

1.000E-03E

1.000E-04 =

1.000E-06 =

1.000E-06 =

1.000E-07

0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9

PEAK ACCELERATION (g)

EPRI 50TH LLNL 60TH

Figure 2a

A

N

N

U

A

L

P

R

O

B

A

B

IL

IT

Y

0.01

1.000E-03

1.000E-04

1.000E-06 s

1.000E-06

1.000E-07

0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8

PEAK ACCELERATION (g)

0.9

EPRI 16TH

LLNL 16TH

EPRI 86TH

LLNL 86TH

EPRI MEAN

LLNL MEAN

Figure 2b

- Figure 2

Probabilistic seismic hazard curves for

the DOE Savannah River, South Carolina, site.

Figure 2a displays the LLNL and EPRI median

seismic hazard curves; Figure 2b displays the

LLNL and EPRI 15th percentile, 85th percentile

and mean seismic hazard curves

123
Figure 3a

Figure 3b

Figure 3

Seismic source maps for the Southeastern

United States from the LLNL study. Figure 3a

shows a broad seismic source for the Charleston,

seismic zone. Figure 3b shows a very localized

seismic source for the Charleston, seismic zone.

124
CHARLESTON SEISMICITY RATES; LLNL EXPERTS

Magnitude (m^ ) Figure 4a

CHARLESTON SEISMICITY RATES; EPRI TEAMS

g lO S . . . .

I

. . . ,

I

. , ,

I

10° r

3. 4. 5. 6. 7. 8. Magnitude (m^ )

Figure 4b

Figure 4

Earthquake recurrence curves for the

Charleston, South Carolina seismic source.

Figure 4a displays the range of recurrence

curves for the LLNL experts; Figure 4b

displays the range of recurrence curves for

the EPRI expert teams

125
MODEL WT. MODEL WT. RV1 0.32 SE1(X14) 0.10 RV5(X2) 0.06 SE1(X2) 0.09 RV5(X3) 0.06 - - - - SE2 0.10 G16-A3 0.20 — C0MS-1A 0.07 MMMM EPRI 100 BARS, THEOR. MAGN. MOMENT

Figure 5

Attenuation relationships from the LLNL and

EPRI studies for magnitudes 5 and 7. The

figure shows the modeling uncertainty

of the combined LLNL and EPRI models which

are given weights based on the models

selected by the individual experts

126
Earthquake Alert System Feasibility Study

by

Philip E. Harben*

ABSTRACT

An Earthquake Alert System (EAS) could give

several seconds to several tens of seconds

warning before the strong motion from a large

earthquake arrives. Such a system would in- clude a large network of sensors distributed

within an earthquake-prone region. The sensors closest to the epicenter of a particular

earthquake would transmit data at the speed

of Ught to a central processing center, which

would broadcast an area-wide alarm in ad- vance of the spreading elastic wave energy

from the earthquake. This is possible because

seismic energy travels relatively slowly (3-6

km/s) when compared with the speed of

light. Utilities, public and private institutions,

businesses, and the general public could benefit from an EAS.

KEYWORDS: earthquake alert system; microseismic; real time; strong motion

1. INTRODUCTION

The concept of an EAS is not new. It was first proposed in an editorial by J. D. Cooper published in the San Francisco Daily Evening Bulletin on November 3, 1868. Although the basic concept for the system is simple, the implementation of a modem EAS requires the use of

diverse and fairly sophisticated existing technology in addition to new research and development on a number of operational issues,

such as reliability, accuracy, and survivability.

The Japanese have employed a seismic shutdown system for many of their railway lines

since the early 1960s and have recently (1990)

deployed a five-station Urgent Earthquake Detection and Alarm System (UrEDAS) as part of an earthquake disaster prevention program

(Nakamura, 1989). A temporary EAS was deployed by the United States Geological Survey

near the epicenter of the Loma Prieta earthquake to give warning to workers on the col- lapsed Cypress overpass structure in Oakland,

California (Bakun, 1990).

The EAS concept discussed here consists of a distributed network of remote seismic stations

that measure weak and strong earth motion

and transmit the data in real time to a central

facility. This facility processes the data and is- sues warning broadcasts in the form of information packets, containing estimates of earthquake location, zero time (the time the

earthquake began), magnitude, and reliability

of the predictions. Users of the warning broadcasts have a dedicated receiver that monitors

the warning broadcast frequency. The user also

has preprogrammed responses that are automatically executed when the warning information packets contain location and magnitude

estimates above a facility's tolerance.

The alert signal broadcast is an area-wide

transmission and is probably sent via FM

radio or television subcarrier communications. Ensuring the survivability of the EAS is addressed with several strategies: hardening

the remote stations for strong motion; using

redundant widely separated central stations,

communications, and warning broadcast

transmitters; and noting that strong motion is emergent (i.e., it builds up over a period of

several seconds), allowing some period of

time in the early evolution of the earthquake,

as accelerations build to damaging levels,

during which a vulnerable system continues

to function.

A low-cost prototype EAS could be developed in some areas by making use of existing

dense networks of microseismic monitoring

stations. These stations could be inexpensively

* Lawrence Livermore National Laboratory,

Livermore, California 94550

127
upgraded to measure strong motion. When

excess communications bandwidth is not

available, a strong-motion upgrade could be

inexpensively accomplished using the existing communication channel and a switching

unit that passes the strong-motion signal during strong-motion conditions and passes the

seismometer signal otherwise.

We designed a prototype switching unit and

tested it at a surface station in the near-field of an underground nuclear explosion to simulate strong ground motion from an earthquake. This successful test demonstrated that

the strong-motion record could be switched

into the data stream within 30 ms and that the

entire strong-motion record could be reliably

switched into the real-time data stream.

About 50 stations distributed relatively uniformly along the major faults of the San Francisco Bay Area could form an effective EAS

prototype network. Such a subset of stations

could be selected from the U.S. Geological Survey (USGS) CALNET microseismic monitoring network. By upgrading these stations to measure strong motion and by installing a parallel real-time processor at the microseismic

central processing facility a low cost prototype

EAS could be developed. The experience to be

gained from an operational prototype is very

important to help address issues such as false alarm rates, false alarm mechanisms, reliability,

earthquake parameter estimation accuracy,

and user alert-signal analysis.

In this report, we examine the feasibility of an EAS and attempt to outline some of the fea- tures, trade-offs, and technical considerations

in implementing such a system. The focus of

this report is on the features and design of a low-cost EAS prototype that, to the extent possible, makes use of existing seismic stations,

communications, and broadcast facilities.

2. EAS USES

There are numerous potential users who

would benefit from an EAS. Table 1 is a partial

list of EAS applications; however, automated

shutdown and activation are the most com- mon functions. Depending on the application,

these two functions vary dramatically in their

consequences if action is taken as the result of a false alarm. An activation function such as opening a fire station door has minimal conse- quences if the door is opened due to an EAS

false alarm. An automated shutdown of a nuclear power plant could disrupt the power

grid of an entire city and result in expensive

power-up procedures to reestablish full- power operation. An EAS false alarm shutdown cannot be tolerated for such an application. Although we will discuss user processing of the EAS alert signal in a later section, it

is important to recognize that users of the alert signal have different applications and conse- quences of their actions. Therefore, the EAS

must provide an optimal summary of the current "state of knowledge" for the earthquake

parameters transmitted in the alert information so that diverse users can weigh

the most recent "state of knowledge" against

the consequences of action.

The real-time location and magnitude estimation that an EAS can produce would be very

useful in the immediate aftermath of a large

damaging earthquake. Emergency services

such as fire, ambulance, and rescue could be

deployed much more effectively if magnitude

and epicenter estimates were immediately

available. The communications media would

also be much more effective in determining

unfolding events related to an earthquake af- termath if estimates of epicenter and magnitude were available. Finally, the strongmotion records obtained from the EAS after a large earthquake would help in fundamental

strong-ground-motion seismology investigations. Such investigations could help improve

the fundamental understanding of site re- sponse and soil amplification during strong

ground motion and, thereby, improve acceleration prediction, structural building codes,

and struchiral damage assessment.

128
3. EAS OVERVIEW

The EAS discussed here was first advanced by

Heaton (1985) and is shown in Fig. 1. Strongmotion and seismometer signals from a distributed network of remote seismic stations are transmitted in real time to the EAS processor at the central acquisition facility. The EAS proces- sor estimates earthquake parameters such as

location, origin time, magnitude, and the reli- ability of this information. If the earthquake

magnitude estimate exceeds a given threshold, a warning signal is broadcast over an area- wide transmitter. This warning signal contains

the earthquake parameter estimates. Each user

of the warning signal has a dedicated receiver

and processor with preprogrammed responses

based on earthquake parameter values. The user's system continuously monitors the warning-signal frequency band. When a user

receives a warning signal that has location and

magnitude estimates in excess of the user's facility tolerances, the facility's emergency re- sponses are automatically implemented.

3. 1 Alert Information Packet

The signal transmitted by an EAS to alert an

area must provide estimates of the earthquake's location and magnitude, and an esti- mate of the reliability of the information since

many users of the warning signal would incur

significant costs by taking any action based on

the alert message. In general, an EAS should

produce more reliable estimates of earthquake location and magnitude as the earthquake evolves and as more stations in the net- work transmit data that can be used in

calculating the location and magnitude esti- mates. It is therefore important for alert information packets to be rapidly updated so that

the new, more reliable estimates based on the

most recent data can be transmitted to the users. The packet's update rate, however, is limited by the amount of information each packet

contains, the communication method employed, and the central processor and alert al- gorithms used.

An initial alert signal can give the greatest lead

time between the warning and the onset of

strong motion. By applying a minimal warning

algorithm at the central station, an alert signal can be issued shortly after only a few stations

have detected strong motion from an earthquake. A very simple algorithm could be used

that requires two EAS stations to measure ground accelerations above a predetermined

threshold within a specified time window before the initial alert signal is issued.

After the initial alert signal is transmitted, the

information transmitted in the subsequent

alert packets should contain all of the salient

parameters that form the input to any user's preprogrammed response algorithm. An ex- ample of a possible alert packet is shown in Fig.

2. TTiis alert packet consists of the following in- formation:

• Update time

• Zero-time estimate

• Location estimate

• Magnitude estimate

• Maximum acceleration

• Number of stations

Including the update time in the packet al- lows the user to identify the most recent information packet and to determine the time that

has elapsed between the most recent information packet transmission and real time. The

zero-time estimate refers to the time the earthquake began and is calculated by the earthquake-location algorithm. The location

estimate gives the hypocenter coordinates for

the earthquake. The magnitude estimate can be an estimate of body wave magnitude, max- imum accelerations expected, earthquake moment, or a combination of these. The maximum acceleration gives the value of the

maximum seen by any station (or an average

of several of the stations reporting the highest

accelerations) up to the time that the alert packet is transmitted. It provides the user with a direct measurement (as opposed to processed estimates), which can be compared

to the magnitude estimate. Finally, only the

129
stations reporting strong ground motion

above a given threshold are included in the

number of stations. This provides the user with some indication of the reliability of the

estimates. The more stations reporting strong

motion (and therefore included in the esti- mates), the more likely that the event is a large

earthquake.

3.2 Central Processor and Algorithms

The central EAS station acquires the incoming

strong- and weak-motion network data for

real-time analysis by the central processor. A

detector algorithm can monitor the level of

strong-motion signals as the data are buffered

in memory. When a threshold level is ex- ceeded, an algorithm to pick the time of Pwave arrival is repeatedly applied to all weakmotion data channels. When at least four sta- tions have an established first arrival, a location algorithm is repeatedly applied to the

first arrival times. As the earthquake evolves,

more P-arrival times are available to be used

in the location algorithm and, therefore, the

estimates produced should be more accurate.

In paraUel, magnitude and reliability esti- mates can be made for the data from the

strong-motion channels. Several algorithms can be applied to magnitude estimation.

Crude estimates can be made by using the

maximum accelerations at the stations nearest

the estimated earthquake location. The

moment estimation method (Toksoz et. al., 1990) uses a relatively dense linear arrangement of sensors along faults to estimate the

fault slippage and the length of the rupture in

real time. By assuming a fault depth, the moment can then be estimated. The Japanese

UrEDAS system estimates magnitude from

the predominant period of the initial P-wave

motion (Nakamura, 1989).

To the extent possible, all magnitude estimation algorithms should be applied to the data

from the strong-motion channels to increase

the reliability of the estimate. The arrangement of stations forming an EAS could preclude the use of the moment estimation

method for earthquakes on some faults or

fault segments because of a lack of instrument

density.

3.3 User Processing

The extent and specifics of alert information

analysis that a particular user may require before executing alert or shutdown functions

depends on the user's application. Potential user applications, such as opening a fire sta- tion door, do not have a significant cost asso- ciated with a false alarm, so minimal or limited processing may be desirable to maximize

lead time. However, the shutdown of a nuclear power plant using the alert signal would

be high in cost. Consequently, maximum alert information analysis that trades off the time

left to execute a safe shutdown against the

probability that a shutdown is unnecessary

may be desired.

In general, the user's preprogrammed re- sponse algorithm can contain a table of precalculated estimates of the ground accelerations

expected at their site from earthquakes in a range of magnitudes and at a range of locations on the major faults in the area. When the

earthquake location and magnitude estimates

provided by the EAS alert packet indicate site accelerations greater than a given shutdown

or alert threshold and when the alert packet

reliability is sufficient, action is taken. The

zero-time estimate of the earthquake can be

folded into the user's decision process. A

time-of-arrival estimate of strong motion is made by the user's algorithm based on the es- timated earthquake location, estimated zero

time, and the current time. When there is more lead-time than required for the user to perform alert actions, the user may elect to wait and receive additional alert packets to in- crease assurance that the estimates are stable

and the reliability is high.

3.4 Alert Signal Transmission

A relatively new area-wide communication

scheme that has low cost potential is FM radio

and television subcarrier communication

130
(Communications Alternatives, 1991). The in- formation is broadcast on a subcarrier fre- quency of the main transmission frequency

and does not interfere with the main broadcast.

Transmission coverage can be up to 100 miles

in radius from the transmitter, depending on

local terrain. The advantages of FM radio and

television subcarrier communications are:

• Low start-up costs

• Wide area coverage

• Low user receiver costs

• Acceptable digital data transmission rate

• High reliability

The start-up equipment cost for a broadcaster

to modulate and transmit on a subcarrier is between $5000 and $20,000, depending on existing equipment. The wide-area coverage of a

single broadcast station should be reinforced

by dual-redundant widely separated broadcast stations to improve area-wide reception

owing to terrain limitations and to assure survivability of a transmitter during a

large earthquake. User receivers should cost

less than $100.

3.5 Survivability

Although a detailed analysis of HAS survivability is beyond the scope of this study, a few

comments on this issue are in order The survivability of an EAS will be driven by the

weakest link in the individual components

that make up the entire system (i.e., seismic

stations, communications system, and the

central station and warning transmitter). In general, individual free-field stations should

survive during strong motion if proper care is taken in the station design and during station

installation. A seismic station running directly on commercial ac power would be vulnerable

to the power interruptions likely during a large earthquake. Powering the station by battery and charging the battery with ac power

or solar panels assures independence from ac power interruptions. Proper bolting of sensors to a concrete pier, attachment of voltagecontrolled oscillators and amplifiers to a rigid

structure bolted to a pier, and secure mounting of batteries and charging units will help ensure proper operation during strong motion.

The survival of the EAS communications between the stations and the central processor is a potential weak link. If satellite communications were used, antenna dish alignment during strong ground motion would be a major

concern. If dedicated telephone lines or radio

telemetry were used, the gathering, repeaters, or central switching stations would be a concern. Direct line-of-sight radio telemetry without repeaters probably would be the least vulnerable, but it is only practical for stations

within range and line-of-sight of the central

station. However, it is worth noting that in the

near-field of a large earthquake, strong

ground motion is emergent (Shakal et. al., 1989). This means that there should be some

period of time after the onset of ground motion from a large earthquake, as accelerations

build to damaging levels, when a vulnerable

communications system continues to function.

The survival of the central processing center/

warning transmission station is also a potential weak link in an EAS. A large earthquake

that occurs near the central processing center

could damage or destroy the EAS before any

alert warning is transmitted. Hardening aU

components of this EAS central processing fa- cility may be difficult to accomplish. One possible strategy is to maintain two widely separated EAS centrals in a master-slave

relationship. Using identical equipment and

algorithms, the slave station issues the alert

packets only when the master station fails to do so.

4. A LOW COST PROTOTYPE EAS

Although the conceptual design, technological features, and capabilities of an EAS are important, the cost is the factor that will determine if one is ever built. A California Division

of Mines and Geology earthquake-warning

feasibility study (Holden et al, 1989) esti- mates the cost of installing a dedicated 43-station EAS in southern California at $1.4 million

for the remote stations, $1.9 million for the

central processing facility, and $1.6 million for

the annual operation and maintenance costs. Because the benefit of such a system is difficult

131
to quantify, the study concludes that an EAS

cannot be economically justified.

A more recent study by the National Research

Council (NRC, 1991) on real-time earthquake

monitoring concludes that an EAS is technically feasible and recommends the installation

of a prototype system. This recommendation

is based, among other considerations, on

forming a prototype EAS from existing microseismic monitoring stations to significantly

reduce installation and operation costs. We have examined the characteristics of existing microseismic monitoring networks and

have made progress in detailing the modifications necessary to convert them into functioning EAS prototypes. We assumed that each

station communicates analog data in real time

to a central processing station. Two essential

tasks that must be accomplished to create a prototype EAS from such a network are:

• Upgrade the remote seismic stations to measure strong motion.

• Install a real-time central processor and

alert algorithms.

4.1 Remote Station Upgrade

to Measure Strong Motion

Existing microseismic monitoring stations re- quire adding a strong-motion-measurement

instrument for each seismometer component.

The strong-motion instruments are commercially available and are typically force-balanced accelerometers. A dramatic reduction

in strong-motion instrument cost can be

realized by making use of the new silicon microstructure accelerometers as strong-motion

instruments. These instruments are now commercially available.

The individual seismic stations of a microseismic monitoring network typically consist of

high gain vertical or three-component seismometers, amplifiers, voltage-controlled oscillators (VCOs), telephone or radio telemetry

communications, and either ac power or battery power with solar panel charging.

When there is extra bandwidth on the microseismic station real-time communication

channel, a strong-motion upgrade can be

achieved by adding strong-motion sensors and VCOs: one for each component, and multiplexing these additional data channels on

the existing communications channel. The

EAS central processing facility would then re- ceive two data streams for each station component: one containing weak-motion data

from the seismometer and one containing

strong-motion data from the strong-motion sensor When spare communication bandwidth is not available, additional communication bandwidth can be added by either upgrading the existing link or by adding

additional links. Unfortunately, upgrading or

installing new communication links (radio or telephone) is expensive.

A low-cost solution to the strong-motion upgrade of microseismic stations that have no

spare bandwidth on the existing communication channel is an instrument switch that, us- ing only the existing communication channel,

rapidly switches between seismometers and

strong-motion sensors at the onset of strong

motion. The EAS central processing facility

would then receive only one data stream for

each station component. The data stream

would contain weak-motion data from the

seismometer until the onset of strong motion

at the station. After that time, the data stream

would contain data from the strong-motion sensor When strong motion at the station subsided, the weak-motion data from the seismometer would be passed to the data stream.

4.2 Prototype Instrument-Switching Unit

We designed a prototype instrument-switching unit and tested it. This switcher concept is shown in Fig. 3. The signals from the seismometer and from the strong-motion instrument

are sent through to the instrument switcher

The strong-motion instrument signal is monitored by the switcher and, under normal weakmotion conditions, the seismometer signal is passed to the communications hardware.

132
When a user-determined threshold is exceeded

by the strong-motion instrument (assumed to be set at a level well above the saturation amplitude of the seismometer signal), the strongmotion signal instead of the seismometer's signal is passed to the communication hardware.

The switcher has a timer microchip that will

perform the switching function for a fixed user- determined time period before switching back

to passing the seismometer's signal. The timer

chip will continue to extend the switch time by

one user-set time period as long as the threshold continues to be exceeded. The unit

switches back to the seismometer's signal

when one complete user-defined time period

has past since the last instant that the threshold

was exceeded. It is worth noting that monitoring the strong-motion sensor to determine a switching condition rather than monitoring the

seismometer is more reliable since the switching threshold is well within the linear regime of

the strong-motion instrument. Furthermore,

routine signals transmitted to the seismometers for calibration purposes cannot trigger the

switch and are not affected by the prototype

EAS.

We tested the prototype switcher in the near

field of an underground nuclear test. A Marks

Products Inc., model L-4 vertical component

seismometer, a Kinemetrics FBA-11 vertical

component strong-motion accelerometer, and

the switcher output signal were all digitally re- corded at 200 samples/s at unity gain. This

prototype was set to switch-over at a threshold

of 0.01 g and the switchover time period was

set at 10 s. The results are shown in Fig. 4. The

strong-motion recording is shown in the top

trace in units of g. The middle trace is the L-4

seismometer recording in units of m/s. Note

that the L-4 seismometer was physically clipping during the first 8 s or so of strong ground

motion. The bottom trace is the switcher output. Since the switcher output is a mix of velocity and acceleration records, the scale was left in raw output voltage units. Clearly, the

switcher output is capturing the full strongmotion record before switching back to the

seismometer. The initial large negative pulse at about 15 s is actually the L-4 seismometer

output. The relative scales of the two instruments is a function of the raw output voltage of

the individual instruments. The relative amplitude scales could easily be modified by adding an amplification or attenuation stage to one of

the instruments before input to the switcher

Finally, the dc offset evident on the strong-motion part of the switcher record was introduced

from an unbalanced voltage supply that was

driving the switcher A dc voltage regulator

has since been added to the switcher to eliminate this problem.

Figure 5 zooms in on the switchover time period. All traces and units are as described

above. The initial L-4 seismometer signal

spike is now evident on the switcher trace. Switchover occurs at about 15.1 s. The strongmotion signal contains a small transient associated with the switching relay but stabilizes

quickly. The time between switchover and a stable strong-motion signal is about 30 ms. The switching is fast enough to capture the

first peak of the strong ground motion from

the nuclear test (Fig. 5). It should be noted that a nuclear test is a worst case test of the

switcher performance because the most significant strong motion occurs in the first few

cycles of the record. Strong motion from large

earthquakes is generally emergent, building

to peak levels over many seconds.

4.3 Central Processing Facility Modification

Most existing microseismic monitoring net- works telemeter data from remote stations to the central processing facility in real time. The

USGS CALNET microseismic network, for ex- ample, transmits seismic data in analog form,

using radio telemetry or dedicated telephone

lines, to a central processing facility. This facil- ity digitizes the incoming data and runs an

event detector on the data streams to determine if a small earthquake has occurred. In general, the data are not continuously ar- chived; data are archived when an event is declared by the detector Although the data are communicated to the central processing facil- ity in real time, the central processor and detector do not run in real time. The data are

133
buffered in memory so that event processing

and data archival can be accomplished after

the event has occurred.

Assuming that the remote stations are modified to telemeter strong-motion information,

either on new data channels if sufficient spare

bandwidth is available, or using the instrument switcher discussed earlier, the EAS re- quires a real-time processor that performs

real-time earthquake parameter estimation on

the EAS data channels. An EAS central processing facility that is operated in parallel

with the microseismic-network central-processing facility would provide the needed

real-time analysis capability without impacting the routine operation of the microseismic

network. Figure 6 illustrates this concept. The

EAS central processing facility would be col- located with the microseismic central facility.

The EAS facility consists of a dedicated real- time central processing unit, digitizers, alert algorithms, and warning communication

hardware. The incoming data from the EAS

stations, a subnet of the full microseismic net- work, would be branched off the

microseismic central input for digitization

and real-time analysis by the central processing facility. The microseismic central proces- sor could also receive the data channels used

by the EAS and perform routine event analysis and archiving of those channels. This

would eliminate the need for archival capability from the EAS central processing facility.

Archived strong-motion events could be "re- played" through the EAS system to test new

algorithms or system enhancements.

4.4 A Prototype EAS in the San Francisco Bay Area

The San Francisco Bay Area is an example of a region that has an existing USGS CALNET

microseismic network that may be suitable,

with modifications, to also function as an

EAS. One critical element is the arrangement

of sensors; some of them must be located near any expected major earthquake epicenters in

the area. Assuming that earthquakes occur with a uniform random distribution of epicenters only on known major faults and

neglecting hypocentral depths, Heaton (1989)

showed that, for a uniform distribution of n sensors along faults of total length L, the mean

travel time for P-waves from the earthquake

epicenters on the faults to the closest station is

where v is the P-wave velocity. By assuming a

fixed hypocentral depth of h for all major

earthquakes and, if we also assume that a minimal warning algorithm that issues an initial alert signal would require strong ground

motion at two or more stations, then the aver- age time before P-wave arrival at the nearest

two stations is

V

Since the San Francisco Bay Area is known to have three major active faults: the San Andreas, Hayward, and the Calaveras Faults. We

can apply Eq. (3) by assuming that the total length of these faults (with instrumentation)

is 800 km. Assuming a P-wave velocity of 6

km/s and a hypocentral depth of 10 km (Scheimer, 1985), the mean travel time to the nearest

two stations can be calculated as a function of

the number of seismic stations along the total length for the three faults. Figure 7 shows the

results. There is little improvement in using over 50 stations. The mean P-wave travel time

using 50 stations is ~3 s. If we assume that 5 s are required to account for the small delays introduced by radio telemetry or phone line

transmission and by the time it takes the alert

algorithms to issue a warning and that the P- wave speed is 1.67 times the S-wave speed

(Scheimer et. al., 1982), then the damaging S- wave will only have reached an area of 8 km

in radius from the earthquake epicenter before the initial alert warning is received by the

entire area.

It is worth noting that an EAS is most effective

(i.e., gives the longest overall warning times

and requires the minimum number of seismic

134
stations) on shallow faults, which are the sources of most of the major earthquakes in

California. Known faults allow the use of a linear distribution of sensors instead of an ar- eal distribution, effectively reducing the number of stations required for the same warning

times by about a factor of 10. Shallow focus

earthquakes provide the seismic ray path geometry for maximum warning to an area. Ja- pan, for example, has major earthquakes that occur on a myriad of faults without surface

expression and at relatively deep focus

(Nakamura et. al., 1988), making an EAS

much more difficult to implement and of less value because of the very short warning times

that could be expected.

The uses CALNET stations in the Bay Area

are shown, along with the major fault traces,

in Fig. 8. Clearly, a subset of 50 stations that are located on or near the major faults with an

approximately uniform spacing can be se- lected. A large fraction of the USGS stations

consist of a single vertical component seismometer Some of these stations have fully

utilized communication channels and cannot

carry additional data on existing communication lines. Such stations could be upgraded to strong motion by adding the switching device

described earlier and a single vertical-component strong-motion instrument. When communications are limited to a one component

instrument, the vertical component is probably the best axis on which to add a strong-motion instrument since peak accelerations in the near field are generally larger on the vertical

component in the early evolution of an earthquake (Shakal et. al., 1987, 1989).

5. CONCLUSIONS

An EAS would provide warning to a diverse

community of users, who have applications

that range from simply opening a fire station

door to shutting down large manufacturing

or power-production facilities. Because of the

vastly different costs associated with false alarms depending on the application, an EAS

must furnish a rapidly updated information

packet with estimates of the earthquake's

location and magnitude, and an estimate of

the reliability of this information so that users can evaluate the data and make independent

decisions, depending on their specific needs.

The technology currently exists to develop

such an EAS; however, a dedicated system

built to all the desired specifications (e.g.,

number of stations, locations, communications, and hardening) would be very expensive. Furthermore, some operational experi- ence with an EAS is required to define how an optimum system should be constructed. Is- sues such as false alarm rate, false alarm

mechanisms, reliability, earthquake parameter estimation accuracy, and user analysis of

the alert information cannot be effectively addressed without the aid of information from

an operational EAS. It is therefore very important to establish some prototype systems to serve as test beds for future development.

In some areas, an EAS prototype system can be developed at low cost by upgrading existing dense networks of microseismic monitoring stations so that these stations can also measure strong ground motion; once modified, they could form the nucleus of the EAS

network. Strong-motion upgrades can be ac- complished without installing any new or ex- panded communication links. For instance,

strong-motion instruments can be added to existing communication links (when unused

bandwidth is available) or the instrument

switching device that we developed and

tested can be added (when spare bandwidth

is unavailable). Furthermore, EAS station

maintenance can be accomplished during

normally scheduled microseismic mainte- nance visits.

6. ACKNOWLEDGMENTS

First and foremost I wish to thank Bob Murray,

Tom Nelson, and Mark Eli for their strong support, encouragement, and helpful criticism. I also thank Jim Hill for supporting the overall

effort. I am indebted to Pete Rodgers and Dan

Ewert for the design, development, and field- ing of the seismic switcher prototype. Special

135
thanks to Elaine Price for editing this manuscript. Finally, I thank Paul Kasameyer for first introducing me to this topic.

This work was performed under the auspices

of the U.S. Department of Energy by

Lawrence Livermore National Laboratory under contract No. W-7405-Eng-48.

7. REFERENCES

Bakun, B. (1990), Early Warning Alert System,

U.S. Geological Survey, Menlo Park, Public

Affairs Memo.

Communications Alternatives (1991), Datapro

Reports (McGraw-Hill, New Yoit), vol. 2. Heaton, T. H. (1985), "A Model for a Seismic

Computerized Alert Network," Science 228,

987-990.

Holden, R., R. Lee, and M. Reichle (1989), Technical and Economic Feasibility of an Earthquake Warning System in California, Califomia

Division of Mines and Geology, Special Publ.

101.

Nakamura, Y, and B. E. Tucker (1988), "Japan's

Earthquake Warning System: Should it Be

Imported To Califomia?" California Geology,

pp. 33-40

Nakamura, Y. (1989), "Earthquake Alami System

for Japan Railways," Japanese Railway Engineering 28(4), 3-7.

National Research Council of the National Academy of Science (1991), Real-Time Earthquake

Monitoring, Report from the Committee on

Seismology (National Academy Press. Washington, DC).

Scheimer, J., S. R. Taylor, and M. Sharp (1982),

Seismicity of the Livermore Valley Region,

1969-1981, Lawrence Livemiore National

Laboratory, Livemiore, CA, UCRL-86969.

Scheimer, J. (1985), Depth Determinations ofSeismicity in the East San Francisco Bay Region,

Lawrence Livermore National Laboratory,

Livennorc, CA, UCID-20522.

Shakal, A. P., M. J. Huang, C E. Ventura, D. L. Parke, T. Q. Cao, R. W Sherburne, and R. Blazques (1987), CSMIP Strong-Motion

Records from the Whittier, California Earthquake of 1 October 1987, Califomia Division

of Mines and Geology OSMS 87-05.

Shakal, A., M. Huang, M. Reichle, C. Ventura, T. Cao, R. Sherbume, M. Savage, R. Darragh, and

C. Petersen (1989), CSM/P Strong-Motion

Records From the Santa Cruz Mountains

(Loma Prieta), California Earthquake of 17

October 1989, Califomia Division of Mines

and Geology OSMS 89-06.

Toksoz, M. N., A. M. Dainty and J. T. BuUitt

(1990), "A Prototype Earthquake Warning System for Strike-Slip Earthquakes," Pure and

Applied Geophysics 133(3), 475-487.

136
Table 1. A partial list of potential EAS applications.

Shutdown computers

Shutdown disk drives

Shutdown airport operations

Stop trains

Shutdown gas distribution

Open fire station doors

Stop elevators in a safe position

Issue audio alarms

Shutdown nuclear power plants

Maintain safe-state in nuclear facilities

Reroute electrical power

Shutdown high precision facilities

Shutdown manufacturing facilities

Shutdown high energy facilities

Alert hospital operating rooms

Start emergency generators

Shutoff oil pipelines

Shutdown refineries

Shutoff water pipelines

Electromechanical Shuldown/Alerl

Functions

Figure 1. The EAS consists of an area-wide strong-motion network that telemeters real-time data to a

central processor. The processor estimates earthquake parameters and issues an alert packet containing

these estimated parameters. The user facilities receive the warning packet and execute alert or shutdown

functions based on their specific preprogrammed response.

137
Electromechanical

Shutdown/Alert

Functions

User CPU and Algorithm

Figure 2. The alert information packet is broadcast and is rapidly updated for user analysis and possible

shutdown and alert functions.

Figure 3. The switching unit receives input data from the strong-motion instrument and the seismometer. The seismometer signal is passed to the station communications under normal seismic conditions.

During strong ground motion, the strong-motion instrument data, rather than the seismometer signals, are passed to the station communications.

138
30

Time (s)

Figure 4. The surface recordings in the near field of an underground nuclear test. The top trace is the

strong-motion accelerometer, the middle trace is the seismometer, and the bottom trace is the switcher

output.

139
-21 1 I I I I I I I I I I I I I I I I I I I 14.0 14.2 14.4 14.6 14.8 15.0 15.2 15.4 15.6 15.8 16.0 Time (s) Figure 5. A zoom-in on Fig. 4 at switchover. Ttie switchover time takes place at ~15.1 s, and the stabili- zation time is -30 ms. The first major peak of the strong-motion signal is captured in the switched signal.

140
Strong Motion Modified

* , Electromechanical Shutdown/Alert

Functions

Figure 6. The subnet of a microseismic network that has been modified for strong motion is event- processed by the microseismic processor as usual and is streamed through the EAS central processor

for real-time estimation of earthquake parameters and warning broadcasts.

20 f T ( I I I

1 r~~i 1 1 1 1

1 I ' I

1 1 r—~T

16 -

(A

0 20 40 60 80 100

No. of stations

Figure 7. Plot showing the mean P-wave travel

time to the nearest two stations of a uniformly distributed network of stations along the major

faults in the San Francisco Bay Area as a function

of the number of stations.

141
> NWR O NTY

<> NVE O NCF

O NAD O NAP O NVA

O NGV

O NHM

Ocos

O CMP

IGS trqZ\OHGW ^

HCA OJPl\ \ ^ . OHFE

V HCR\

,0 JEL >XHC^ ^ h!sr ,

^^^^

Figure 8. Map showing the location and identifiers of the

USGS microseismic network in the San Francisco Bay Area.

Major fault traces are also shown.

142

J
Guidelines for Seismic Design Methods of Large Underground Structures

by

Hajime Asakura\ Kazuhiko Kawashima', and Hideki Sugita^

ABSTRACT

Outline of the Guidelines for Seismic Design

Methods of Large Underground Structures is presented. This Is one of the final accomplishments of the 5-year Mlnlstrlal Research Project on Development of Utilization

of Underground Space which was made between

1987 and 1992. Description Is given to seismic

design methods proposed for the Lateral Long

Underground Structures, Laterally Wide

Underground Structures and Vertically Long

Underground Structures.

Investigations on the seismic design methods of

underground structures were compiled In the

form of "Guidelines for Seismic Design Methods

of Underground Structures"^'. Because the

seismic design methods for the underground

structures with small diameter such as the

petroleum pipes'' and water pipes'", and common utility ducts^' were already In use, the

Guidelines show the seismic design methods for larger underground structures than those

facilities. This paper outlines the Guidelines.

KEY WORDS

Underground Structure, Seismic Design,

Seismic Deformation Method, Shield Tunnel.

Disaster Prevention Systems

1 INTRODUCTION

Utilization of underground space Is becoming significant Importance In urban area

in Japan because of various restrictions for

getting open space for new construction. The

Ministry of Construction Initiated a 5-year

Research Project on "Development of

Utilization Technic of Underground Space" In 1987. The Public Works Research Institute,

Building Research Institute and Geographical

Survey Institute worked together In the

Research Projects. One of the main research

targets of the Project was to develop rational

seismic design methods of large underground

structures embedded In soft ground. For executing the Research Projects, Joint

researches were made between the Public Works

Research Institute and Advanced Construction

Technology Center. As one of such Joint

researches, the Joint Research on Development

of Seismic Design Method of Underground

Structures was conducted between the

Earthquake Engineering Division of the Public

Works Research Institute, Advanced

Construction Technology Center and 8 private

firms ( Fujlta, Kajlma, Talsel, Toblshlma,

Hazama, Okumura, kounolke and slmlzu).

Including the research results obtained

through the Joint research" with Advanced

Construction Technology Center, the final accomplishments of the 5-year Research Project was complied In March 1992. The results of

2 SCOPES AND CONTENTS OF THE GUIDELINES

The Guidelines present the seismic design

methods for large underground structures constructed In soft ground In city area. As shown In Fig. 1, underground structures may be

classified, based on their structural response

characteristics, into three groups as

a

(a) Laterally Long Underground

Structures

(c) VertlcaOy Long Underground

Structures

^am WM^^ —

lb) Laterally Wide Underground Structures

Fig. 1 Classification of Underground Structures

1) Laterally Long Underground Structures

Long underground structures embedded

laterally along ground surface. Seismic force

along their axes Is the major effect of an earthquake for such structures.

2) Laterally Wide Underground Structures

Widely spread underground structures along

DDlrecter, 1st Research Department, Advanced

Construction Technology Center

2) Head, Earthquake Engineering Division,

Public Works Research Institute

3) Research Engineer, ditto

143
the ground surface. Shear stress along the

surface of structures associated with shear

deformation of ground due to an earthquake is the major seismic effect.

3) Vertically Long Underground Structures

Long underground structure In vertical

direction. Shear stress along the surface of structures associated with shear deformation

of ground due to an earthquake is the major

seismic effect.

The seismic safety required for design

depends on the type, utilization and importance

of the structures. Therefore, seismic design

methods which are required for underground

structures are presented for the three ts^es

in general sense in the Guidelines.

Underground highways, underground parking

space and common utility ducts were imaged for

developing the Guidelines.

It should be noted that although the

description is given In the format of the

specifications, it is not the mandate

specifications. It merely describes the

research results. The table of contents of the Guidelines is as follows:

1 General

1.1 Scopes

1.2 Definition of Terms

2 Basic Principle of Seismic Design

3 Loads and Loads Combination for Seismic

Design

3.1 Loads and Loads Combination

3.2 Seismic Effects

3.3 Inertia Force

3.4 Design Ground Displacement

3.5 DjTiamic Earth Pressure

3.6 Design Shear Stress around Surface

3.7 Classification of Soil Condition

3.8 Input Ground Motion for Dynamic

Response Analysis

3.9 Soil Spring Stiffness

4 Seismic Design Method of Shield Tunnels

4.1 General

4.2 Seismic Design Method In Axial Direction

4.2.1 Basic Principle of Seismic Design

4.2.2 Seismic Design by Seismic Deformation

Method

4.2.3 Equivalent Stiffness of Shield Tunnel

without Second Lining

4.2.4 Equivalent Stiffness of Shield Tunnel

with Second Lining

4.2.5 Evaluation of Seismic Safety

4.2.6 Dynamic Response Analysis

4.2.7 Minimum Reinforcement of Second

Lining in Axial Direction

4.3 Seismic Design Method In Transverse

Direction

4.3.1 Basic Principle of Seismic Design

4.3.2 Seismic Design by Seismic Deformation

Method

4.3.3 Dynamic Response Analysis

4.3.4 Evaluation of Seismic Safety

5 Seismic Design Method of Laterally Wide

Underground Structures

5.1 General

5.2 Seismic Design by Seismic Deformation

Method

5.3DynamIc Response Analysis

6 Seismic Design of Vertically Long

Underground Structures

6.1 Basic Principle of Seismic Design

6.2 Seismic Design by Seismic Deformation

Method

6.3 Dynamic Response Analysis

6.4 Evaluation of Seismic Safety

7 Structures Expecting Reduction of Seismic

Effects

7.1 General

7.2 Flexible Joints

7.3 Structures Reducing Stiffness of Ring

Joint of Shield Tunnel

7.4 Isolation of Underground Structures

from Surround Subsoils

8 Earthquake Disaster Prevention Systems for

Underground Highways

8.1 General

8.2 Fear and Anxiety of Drivers against an Earthquake

8.3 Factors Causing Fear and Anxiety to Drivers at Underground Highways

8.4 Basic Countermeasure against Earthquake

Disaster

3 BASIC PRINCIPLE OF SEISMIC DESIGN

Seismic design of underground structures

is recommended to be made by the Seismic

Deformation Method^'. In the Seismic

Deformation Method, the displacement and

deformation developed In the subsurface

ground during an earthquake are taken as the

major seismic effects to underground

structures. The seismic response of

underground structures is distinguished by the

small mass effect and the large damping ratio

from regular structures on ground. I.e.,

144
a) mfiss effect Is small because unit weight of

underground structures Is In most cases

smaller that that of subsurface ground, and

b) damping is very large because radiatlonal

damping due to energy dissipation from

underground structures to subsoils is very

large.

Therefore, It is assumed in the Seismic

Deformation method that during an earthqualte

underground structures deform following the

deformation of subsurface ground. Modelling of

underground structures has to be made so that

structural response of the underground

structures be correctly represented. Because

the increase of stress and deformation of

underground structures due to seismic effects are evaluated by the Seismic Deformation

Method, the stress and deformation actually

developed during an earthqualte have to be

computed by adding them with those developed

due to the static loads such as the dead

weight.

Evaluation of seismic safety of

underground structures can be made by either

the Allowable Stress Design Method or the

Bearing Capacity Design Method. In the

Allowable Stress Design Method the stress and

the displacement developed In structual

components are required to be less than the

allowable stress and allowable displacement.

In the Bearing Capacity Design Method the

sectional forces developed In structural

components are required smaller that those of

dynamic bearing capacity of structural systems

considering ductility. Displacement of structures are also required to be smaller

than the critical displacement.

Table 1 Classification of Ground Condition

GROUND CONDITION DEFINITION APPROXIMATE ESTIMATION

GROUP I To<0.2SEC TERTIARY OR OLDER

GROUP II 0.2STq<0.6SEC ALLUVIUM AND DILUVIUM

GROUP in O.eSECSTo SOFT ALLUVIUM

After the design by means of the Seismic

Deformation Analysis, dynamic response

analysis is recommended for the underground

structures with complex structural response.

4 LOADS CONSIDERED FOR SEISMIC DESIGN

In seismic design of underground

structures, the dead weight, the earth

pressure, the water pressure, the uplift force,

the ground settlement and the seismic effects are to be considered. The loads have to be

combined so that the most critical stress or displacement be developed. As the seismic

effects, the following loads have to be

considered:

(1) Inertia Force

The lateral force coefficient has to be

evaluated as kh = Cz- Co • cu- kho (1) where,

kh : lateral force coefficient Cz : modification factor for zone (refer

to Fig. 2) Co : modification factor for ground

condition (refer to Table 1) ^ Cu : modification factor for depth, and Is evaluated as Cu= 1.0-0.015Z (2) z : depth from ground surface (m) kho : standard lateral force coefficient

145
05

o«» 0.4

S/Ui) 0.3

>

COEt

0.1 0.07

«>

in 0.05 e

o

o. 0.04

38 yjxjo cc

>•» 0.02

*oo?

001

I 1—1—r-nrm 1 1 1

- /(1.0.0.25)

: /(0.2.0.05)

(0.1.0.02)

1 1 1 1, t 1 _LJLJ_._. 1 1 1 O.I 0.2 0.3 0.5 1.0 2.0 3.0 5.0

Nolurol Period of Subsurfoce Ground (sec) Fig. 4 Standard Velocity Response Spectrum

{Allowable Stress Design Method)

0.1

0.7 0.5

0.4

0.3

0.2

S 0.07 I 0.05

<£ 004 ^ 0.03 I 0.02

0.01

—r —1 1—i—n III 1 1

t— /(1.6.085)

\

; /

: /(0. 18. 0097)

(0.1.0.045)

III 1 II

01 02 030405 1.0 ZO 3.04050

Noturol Period of Subsurface Ground (sec) Fig. 5 Standard Velocity Response Spectrum

(Bearing Capacity Design Method)

(=0.2)

(2) Design Ground Displacement

The design ground displacement is evaluated depending on the depth and the

distance as (refer to Fig. 3) :

1 z « X u(z.x)=—gSv Ts cos(—--) •sln(----) (3) n 2n zL

where.

u(z,x): Design ground displacement (m) at depth z (m) and at distance x (m) Sv : Design velocity response spectrum at design base rock (cm/sec), and Is determined as Sv=C2 • Svo (4) Cz : Modification factor for zone (refer

to Fig. 2) Svo : Standard velocity response spectrum

(cm/sec) at design base roclt, and is given by Fig. 4 for the Allowable

Stress Design Method and by Fig. 5 for the Bearing Capacity Design

Method

Ts : Natural period of subsurface ground

To = 4S

(5)

(6)

(7)

(sec), and is evaluated as T3=1.25

_H> Vsi

Hi : Thlcl{ness of I-th soil layer (m) Vsi : Shear wave velocity of i-th soil layer (m/sec) L : Design wave length (m) of ground

motion

^

2L.L2

Li +L2

Li, La : Design wave length (m) at subsurface ground and base rocl<, respectively, and are determined as L.=T3VgD (8) L2=TsVsDB (9) VsD : Averaged shear wave velocity of

subsurface ground (m/sec)

VsDB : Shear wave velocity at design base

rock (m/sec)

(3) Design Shear Stress around Surface

At the surface of underground structures

contacting with soils, the design shear stress around surface has to be applied to the

underground structures as shown In Fig. 6.

Ulz) Ground Surface U(z) Z"0

z-H. Subsurface Ground

iese-Rock Fig. 6 Design Shear Stress around Surface

The design shear stress around surface Is given by Eq. (10). If the design shear stress around surface by Eq. (10) Is larger than the

strength of soils around the underground

structure, the design shear stress around

surface shall be the strength of soils. ,=„„S..Tssin(----) (10)

where,

r : Design shear stress around surface

(tf/m=)

Sv : Design velocity response spectrum at design base rock (cm/sec), and is given by Eq. (4) Gd : Shear modulus of subsurface ground

(tf/m=)

Tg : Natural period of subsurface ground

146
H

(sec), and Is computed by Eq. (5) Thickness of subsurface ground (m)

(4) Dynamic Earth Pressure

Dynamic earth pressure by Eq. (11) Is applied to side wall of underground structures

as shown In Fig. 7.

u(z) Ground Surface u(z) 'W///^'

z-O

Subsurface Ground ^7M^ Bese-Rock Fig.T Dynamic Earth Pressure

p(z)=kH- {u(z)-u(zb)) (11) where,

p(z) : dynamic earth pressure per unit area (tf/m^) at depth z (m) Kh : spring stiffness of soils per unit area (tf/m^)

u(z) : Design ground displacement (m), and

Is given by Eq. (3) zb : Depth of base of underground

structure (m)

(5) Input Ground Motions for Dynamic Response

Analysis

Design acceleration response spectrum at design base rock Is given as S=CzCdSo (12) where,

S : Design acceleration response

spectrum (cm/sec) at design base rock

cz : Modification factor for zone (refer

to Fig. 2) Cd : Modification factor of damping ratio,

and is given £is 1.5 Cd =

40hi+l

hi : modal damping ratio of critical So : Standard acceleration response

spectrum at design base rock, and is given by Fig. 8 for the Allowable

Stress Design Method and by Fig. 9 for the Bearing Capacity Design

Method

Input accelerations for time response

analysis shall be either selected from strong

motion records which have the close response

spectral characteristics with S by Eq. (12) or

-+0.5 (13)

500

400

300

200

1007050

40

30

20[-

10

1 1 r- T r 1 1 ri "1 1 1

(02.200) (1.0.200)

-

(0.1.158.7)

-

lJ.U,^u/

1 1 1 1 1 1 1 1 1 1 1 1 0.1 02 03 0.5 1.0 2.0 3.0 5.0 Natural Period Ti (sec) Fig. 8 Standard Acceleration Response Spectrujn

(Allowable Stress Design Method)

1000 700 500

400

300

200

100705040

3020

10

-T r—I—I I I [ I I

1 1—

(0.18.425) (1.6.425)

(0.1.348.1)

(5.0.136)

0.1 0.2 030.40.5 1.0 2.0 30 4.05.0 Noturol Period Ti (see) Fig. 9 Standard Acceleration Response Spectrum

(Bearing Capacity Design Method)

obtained by adjusting the strong motion

records In frequency domain so that the

acceleration response spectrum be matched

with S by Eq. (12).

5 SEISMIC DESIGN l^fETDOD OF LARGE SniELD

TUNNELS

5-1 Seismic Design Method in Axial Direction

(1) Seismic Design by Seismic Deformation

Method

Shield tunnels are constructed by

assembling a number of segments, ring bolts

and segment bolts as shown In Fig. 10. Although there may be various idealizations of

shield tunnels as shown In Fig. 11, the

Idealization of shield tunnels as a beam with

equivalent stiffness may be the most appropriate for seismic design. Assuming a shield tunnel as a beam, elastlcally supported

by subsurface ground, with equivalent

stiffness as shown In Fig. 12, the seismic forces

developed In the shield tunnel are evaluated

147
SECMEWT JOINT

SEGMENT

Fig. 10 Segments and Joints of Shield Tunnel Lining

by the Seismic Deformation Method as""

P'n' a", ------ {EA)la

Mh= a

L

4 Uh

4 Uv

(EA)i<,

(EI)«.

Mv= a 3---'-5— (El)eq

Qr,= a 2—^-3-(EI),q

8n Uv Qv= or 3--|-£— (EI)«q

(14)

(15)

(16)

(17)

(18)

(19)

(20)

(21)

where,

PrT.Ph : Compression and tension force (tf), respectively, developed due to the

design ground displacement in

lateral direction

P^.Pv^: Compression and tension force (tf), respectively, developed due to the

design ground displacement In vertical direction

Mh.Mv : Bending moment In horizontal

plane and vertical plane (tf m), respectively

Qh.Qv : Shearing force In horizontal

plane and vertical plane (tf), respectively

(EA)S<,,(EA), Equivalent compression

rigidity (tf) and equivalent tension

rigidity (tf) of shield tunnel, and

is given by Eqs.(31) and (32)

(£!)•<, : Equivalent flexural rigidity (tf), and is given by Eq. (35)

Ut, : Design lateral ground displacement (m) at depth of gravity center of shield

tunnel, and Is given by Eq. (3) Uv : Design vertical ground displacement

(m) at depth of gravity center of

shield tunnel, and is given as 1/2 of

Uh

L : Design wave length (m), and is give by

Eq. (7) a 't, a T, a ^, a 3 : Coefficients to represent the degree of transmission

of response displacement from the

subsurface ground to shield tunnel,

and is given by Eqs. (22) ~ (25).

Spring Elements Spring Elements Beam Elements cz _i_ r

(a) Finite Element Model (b) Spring-Beam Model

<3=

N N

Tension 8 Compression

j Flexural Stiffness Axial Stiffness

^ ?

(c) Idealization by Beams with Equivalent Stiffness

Fig. 11 Idealization of Shield Tunnel

I I

Ring Joint Kj

Fig. 13 Axial Stiffness of Segments-Ring Joints

148
Ui =Aeoj*-jln

Fig. 12 Ground Displacement Assumed In Seismic

Deformation Method

a T

a I

a 2

a 3'

l*{2n /t

1 l+(2ff /X^

1 l+(2ff / A

1 l+{2ff / A. ID'

where,

/kV.

(eaTIc

V(EI)«<,

"(£!")«<,

Li=\/2'l

Koi, Ko2, Kci3 : soli spring In

(22)

(23)

(24)

(25)

(26)

(27)

(28)

(29)

(30)

longitudinal, transverse and

vertical direction, respectively

(tf/m=)

(2) Equivalent Stiffness of Shield Tunnel

without Second Lining

1) The equivalent stiffness of shield tunnel is different between tension and compression as shown In Fig. 13. The equivalent rigidity is computed by Eqs. (31) and (32). (EA)^.,=EsAs (31)

1 (32)

where.

(EA)S<, : Equivalent compression rigidity

(tf) (EA)I<,: Equivalent tension rigidity (tf) Es : Elastic modulus of shield segments

(tf/m=)

As : Sectional area of segment- ring (m^) Is : Length of segment (m) in axial

direction Ks : Stiffness of segment-ring (tf/m),

and is given by Eq. (33) EsAs

Rs=—

2 3

(33)

Kj Tension stiffness of total

ring-joints (tf/m), and is given by Eq.

(34) Kj = nkj (34) kj : Tension stiffness of a ring-Joint

(tf/m) n : number of ring-Joint

2) The equivalent flexural rigidity of shield

tunnel Is computed as shown in Fig. 14, and Is given by Eq. (35)^' COS^0

'135)

where.

(El) ^ . . J, " cos ^ +( » /2+ # )sin ^

M

ICll.

Fig. 14 Flexural Stiffness of Shield Tunnel

149
AXIAL DEFORMATION [mm) (a) Experimental Set-up (b) Loading Hysteresis Loop

Fig. 15 Effect of Second Lining for Ductility of Flexural Deformation

Ground Suffoce ^Model of 33^ Ventilation Model of Shield Tunnel '^^'^ Tower

1 ? ?! 5/6 7 8 9 10^,1 12 13 14 15 16 17 18 19 20 2! 22 23 24 25 26 27 28 29 30 31 ^ \ w y\ w k \

^

\ \ \ \ \ u w \ \ w \ \ w w

123456789 lO'^n 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31

Base -Rock

Fig. 16 Model of Shield Tunnel, Ventilation Tower and Subsurface ground

Model of Subsurface Ground

(EDe., : Equivalent flexural rigidity of

shield tunnel (tfm") Es : Elastic modulus of segment (tf/m^)

I3 : Moment of Inertia of segment (m")

4) : angle measured from center line to neutral line (rad), and Is given by

solving the following equation :

1 K..

(t> *cot(l> = n {- —*

) (36)

2 EsAs/lg

(3) Equivalent Stiffness of Shield Tunnel with

Second Lining

General description for evaluating the

equivalent stiffness of shield tunnel with

second lining Is difficult because It depends

on the spring action of axial reinforcements of

the second lining. Based on a series of model

loading test results at the Public Works

Research Institute"' '°' as shown In Fig. 15, It Is known that as well as the segments the

second lining resists to the compression force. However, to the tension force the Interaction

between the segments and the second lining

depends significantly on the damage developed

in the second lining and at the ring-Joints. It may be classified Into four steps as shown In Table 2.

(4) Evaluation of Seismic Safety by Dynamic

Response Analysis

In the dynamic response analysis, shield

tunnel Is generally Idealized by a elastic beam

with the equivalent stiffness as shown in Fig.

16. The subsurface ground Is Idealized by

either finite element model or spring-mass

model, and the nonllnearlty of soil stiffness Is modeled by the equivalent linear method. Input

ground motion Is prescribed at the base-rock.

Evaluation on spatial variation of input motion

at the design base-rock Is very Important.

(5) Minimum Axial Reinforcement In Second

LiningFor avoiding the concentration of the

damage at only selected ring-Joints as shown in

150
SECOND LINNING

Table 2 Damage Degree of Shield Tunnel with Second

Lining

Step Damage Degree

I All sections In 1st and 2nd linings are In elastic. No damage.

II Cracks In 2nd lining around ring only at ring Joints.

III Cracks In 2nd lining around ring occur not only at ring Joint but at other sections.

IV Contact between the 1st lining and 2nd lining Is lost. Final

failure Is to be developed.

-SEGMENT (OUTER LINNING) -SECOND (INNER) LINNING

Fig. 17, It Is required to provide the mlnlmuin reinforcement In the second lining In tunnel

axis as As=(Al- a u-Pr)/<7 ls (37) whercx

As : Minimum reinforcement required for

second lining In longitudinal direction

(cm^)

Al : Sectional area of second lining (cm^) a u : Tension strength of second

llnlng(kgf/cra=) a LS : Tension strength of reinforcement In second lining (kgf/cm^) Pr : Tension strength of ring Jolnt(kgf)

,and takes the smaller value between Pa by Eq.(38) and Pb by Eq.(39).

PA=npA (38) PB=mpB

where,P

(39)

Tension strength of anchor

relnforcement(kgf)

Pb : Tension strength of ring bolts(kgf) n : Number of anchor reinforcements per

ring Joint m : Number of ring bolts per ring Joint Pa : Tension strength of an anchor bolt(kgf)

Pb : Tension strength of a ring bolt(kgf)

5-2 Seismic Design Method In Transverse

Direction

The seismic design of shield tunnel in transverse direction is generally made as :

1) Idealize the shield tunnel as a ring with

equivalent stiffness,

2) Consider soil springs, around the tunnel

ring, which represent the soil-structure

interaction,

3) Evaluate the design ground displacement.

z

SECOND LINNINGTENSION CRACK

SEGMENT

FAILURE OF JOINT

(a) Insufficient Reinforcement

TENSION CRACK

TENSION CRACKS

PROPAGATION OF CRACKS ^^'L^ •^"S"

(b) Enough Reinforcement Fig. 17 Nlnlnum Retnforcement Required in Second Lining for Preventing Concentration

of Damage at only Selected Ring Joints

the design shear stress around surface and the

Inertia force at the center of gravity of the

shield tunnel,

4) Divide the design shear stress around

surface and the inertia force into the

components in tangential and normal

directions, and apply them to the shield

tunnel. The design ground displacement is also

divided into the components in tangential and

normal directions, and apply them to the shield

tunnel through the soil springs,

5) Compute the Increase of force, due to seismic effects, developed in the tunnel rings,

and

6) Evaluate the force, developed in the tunnel

rings, during an earthquake by adding the

force of 4) with the force developed due to

static loads

When the following assumptions can be made

for the shield tunnel and subsurface ground.

151

\
Elastic and Homogeneous

'

. Go, I'D

' Cn'= ( (43)

Reaction Force

in Tangentiol Direction

Reoction Force in Normal Direction

Displacement in Normal Direction

Fig. 18 2nd Model Reformation of Fourier Serrles

assuBed for Tunnel Ring

the force Increase, due to the seismic effects,

developed in the tunnel ring can be easily

computed.

1) The subsurface ground is elastic and

homogeneous,

2) The fundamental mode is significantly

predominant In the subsurface ground,

3) Tunnel ring is elastic,

4) The inertia force to the tunnel ring is negligibly small,

5) Depth of soil to the shield tunnel is large

enough, and the shield tunnel is not In close

proximity of the design base-rock,

6) Saturation of the design shear stress around

surface at the shear strength of soils is disregarded, and

7) The soil spring can be evaluated based on

force vs. displacement relation of an elastic

ring placed in infinite elastic homogeneous

material as shown in Fig. 18.

The assumption 7) require more explanation.

Consider an elastic ring which is placed in the

Infinite elastic material. From the displacement

of the ring by applying a distributed load, the

soil spring may be evaluated by Eqs. (40) and

(41).

l<n=--|--Cn (40)

Cn' (41) R

C„= ( ^ ^ (42) 2n+l-2 -—V D(n+1) :nS2

3-4 f D

0 :n=0,l

:n^ 2 n»2-2 V o(n*l) 3-4 V D where*

kr, : Spring stiffness associated with the

displacement In normal (or tangential)

direction due to the force In normal

(or tangential) directlon(tf/m^)

kn* : Spring stiffness associated with the

displacement in normal (or tangential)

direction due to the force in tangential (or normal) dlrection(tf/m')

Go : Elastic modulus of subsurface

solls(tf/m^) R : Radius of shield tunnel rlng(m)

u D : Polsson's ratio of subsurface ground

n : mode number In terms of Fourier series

When the soil spring can be represented by

Eqs. (40) and (41), the increase of force,

developed In shield tunnel ring, due to the

seismic effects may be represented as 3« EI

' '2R h'o Ut

7t ' H •sln(— ) C sln20

3 7t ' EI n H

Q=- yg . ------ • Uh • sin(---~) • C • cos2 6

K 'Ho Znc

(45)

(46)

y5

• sin2 e

U..sln(-----)(l.-g---. C

(47)

(45)

N

e

^_ 4(1- «/ d) Gd R^

'('3-2VV)'GD'l^^»6('3-r«/"DT"EI

where, s

1*1 : Bending moment of tunnel ring per unit

length(tf m/m)

Q : Shear force developed In tunnel ring

per unit length(tf/m)

Axial force developed In tunnel ring

per unit length(tf/m)

: Angle measured from crest of tunnel

rlng(rad)

Hg : Thickness of subsurface soils(m) H : Depth of the gravity center of shield

tunnel from the ground surface(m)

Radius of tunnel rlng(m)

: Flexural rigidity of tunnel ring per

unit length(tf m^/m)

Gd : Shear modulus of subsurface

ground(tf/m^)

u D : Polsson's ratio of subsurface ground

Uh : Design ground displacement at the

ground surface, and Is given by Eq. (3)

yS : Modification factor for higher order

modes ( = 1.3 )

R : EI

152
Fig. 20 Two Dimensional Finite Element Idealization

6 SEISMIC DESIGN METHOD OF LATERALLY WIDE UNDERGROUND STRUCTURES

The shear stress around surface Is significant In seismic design of the Laterally

Wide Underground Structures. The increase of

force, developed In the structures, due to

seismic effects are computed by applying the

design shear stress around surface, the

dynamic earth pressure and the Inertia force

as shown in Fig. 19. Dynamic response analysis Is recommended

for the Laterally Wide Underground Structures

with the following conditions :

1) For the underground structures constructed

in subsurface ground with significantly

irregular soil condition in either vertical or

lateral direction. The condition where

Irregularity of the surface or the design

base-rock Is significant is also the one recommended for the dynamic response analysis.

Two dimensional finite element Idealization as shown in Fig. 20 is recommended for such

analyses.

2) For the underground structures with

significant out-of-plane deformation.

Analytical models as shown In Fig. 21 Is recommended for such structures. In the

models, the subsurface ground Is recommended

to be Idealized by the mass-spring model with one or two layers in vertical direction and the

Laterally Wide Underground Structure Is idealized by plate elements, shell and solid

elements.

Fig. 21 Idealization by Plate Elements and

Shell Elements

7 SEISMIC DESIGN METHOD OF VERTICALLY LONG

UNDERGROUND STRUCTURES

Because the design shear stress around

surface Is predominant in the Vertically Long

Underground Structures, it is Important to evaluate it carefully. Fig. 22 represents how

the seismic forces are applied to the

structures. There are structures with large

section, which shows significant deformation in lateral direction as well. Shell elements and

solid elements are recommended for the

Direction of Excitation Direction of Excitation 17

1

\ t

la) Dynamic

Earth Pressure

lb) Shear Force around

Surfoce

(c) Inedio Force

Fig. 22 Seismic Effects to Vertically Long

Underground Structures

153
0.9 0.9

Unclear

1 Complex

Evaeuolion Evacuation Route Route

ogp

Uncleor Where

is the Evocuotion Space

0.9

Panic

0.9

TO.9

Lonj Time Required

for Evacuation

Poor Disaster Prevention

Focility ol Evocuotion Space

Q9

SPoor

jGuide- signs and

iMonitorlng ISystems for lEvocuotion

^0.9

Rush of Large Crowds of People

0,9

Poor

Evacuation Guide Systems

0.9

T

0.9

Small Number of Evacuation Route

Poor

Evacuation

Focilifies to Ground such

OS Stairs and

Elevators

109

Norrow Evacuation

Route

Small Number of Evocuotion Space

0.9

Narrow Evocuotion Space

0.9

Far Emergency

E>il

Flg.23 Structural Model predicted by FSM for Fear and Anxiety

against Evacuation

Closed Spoce Damage i« Likely to be Eiponded

to.9

To Know Damage

0.5

Ponic

Submersion

07

0.7

Osj |o.8

BreaVdown of Electriclly Smoke and Got

Vulnerobillty of Inside Facility

07

Vulnerobillty

of Cor]

0.7

0.7

Observotlon

10 07

Fire

0.7 0.7

I Inodeqvota Emergency

j Equipment of Cors

08

Traffic Accident

Collopse of Tunnel osj

Weoli Tunnel Structure

0,7

Q8

IWeok Moteriols 06

No Seismic Counlermeosure

Fig. 24 Structural Model Predicted by FSM

for Possible Influence of Damage

analysis of such underground structures.

8 EARTOQUAKE DISASTER PREVENTION SYSTEMS

FOR UNDERGROUND HIGHWAYS

8-1 General

Although seismic design Is, of course, made

for constructing underground highways, there are always considerable anxiety for seismic

safety. When an earthquake occurs at the

situation where drivers have much anxiety

about seismic safety, fear Is likely to be

amplified to cause panic. Therefore although

seismic design Is of course essential

Importance for reducing seismic damage, it is

also Important to mitigate fears and anxiety of

drivers against safety during an earthquake.

8-2 Fear and Anxiety of Drivers against an Earthquake

Fear and anxiety which are likely to be

Induced in drivers when they Imagine

"underground highways" and "earthquakes" were analyzed by the Value Engineering Method. It was found from the analysis that there are two types of fear and anxiety. The first Is the fear

and anxiety which are always possessed by

drivers against the Image of "underground

highways", and the second Is the fear and

anxiety against the accidents which are likely to be developed during an earthquake.

Although some of them are Infantile

Imaginations from the professional sense, it Is Important to know how the drivers feel against

"underground highways" and "earthquakes".

(1) Fear and Anxiety Always Possessed by

Drivers

a) Fear and anxiety from latent sense to the

underground structures Fear and anxiety come from various latent sense to the underground highways such as darkness, narrowness and dirtiness In comparison with the on-ground highways. They

also come from unexperienced driving on underground, unfamiliar sense and

insensibility to time and direction.

b) Fear and anxiety against driving Fear and anxiety are sometimes developed

even only anticipating to drive on underground highways. The typical senses are

154
the difficulty to know where they are driving,

limited visibility and possible bad manner of

drivers.

c) Fear and anxiety against environment

They come from anticipation that the

environment may be worse at underground

highway than at on-ground highways, such as Insufficient facilities for services and rest room and difficulty for access to other

facilities. Little change of scenery and

circumstance and possible Insufficient

maintenance for service facilities also cause

the fear and anxiety. Sometimes they are Induced even from Imaging that there may be

few drivers at the underground highways.

(2) Fear and Anxiety against Possible Accidents

during an Earthquake

a) Fear and anxiety associated with possible

Influence of damage

Fear and anxiety are developed against

possible Influence to drivers due to structural

damage, damage of non-structural components

and crash of cars. Submergence and fire are

the most typical scenario for such fear and

anxiety. They are also Induced by looking at the occurrence of panic, break-down of

electricity and accidents.

b) Fear and anxiety against emergency

countermeasures

Some drivers have fear and anxiety because

they do not know how to respond to emergent

situation, where disaster prevention facilities are and how to operate them due to their

Inexperience.

c) Fear and anxiety against evacuation Fear and anxiety are Induced by imaging

unsuccessful evacuation from underground

highways. For example, they have the fear and

anxiety against the possible difficulty to know

where the evacuation and refuge space are, possible long time required for evacuation,

Insufficient sign for direction and rush of

refugee.

d) Fear and anxiety for Insufficient

Information for damage and refuge

They come from the anxiety for not be able

to get enough Information on damage situation

and evacuation. Drivers at underground

highways anclpate that they could not get the

Information on what damage was developed on

ground, and that the Information may be Incorrect even If they could get the

Information.

e) Fear and anxiety against traffic accidents

This Is not necessarily unique to the

underground highways. Fear and anxiety are Induced against snake-llke driving'" and

collision of cars due to sudden brake.

8-3 Factors Causing Fear and Anxiety to Drivers at Underground Highways

There are various factors causing Tear and

anxiety to drivers during an earthquake. Because the factors are compllcatedly coupled,

their structure was analyzed by the

Interpretive Structural Modeling Method and

the Fuzzy Structural Modeling Method. Figs. 23 and 24 show some of such analyses. Based on

the analyses, the most basic factors (the

lowest factors Identified from the analyses)

causing fear and anxiety to drivers during an earthquake may be represented as :

(1) Fear and anxiety always possessed by

drivers

a) sense of closing and Isolation

b) complex routes

c) little separation between automobiles and

pedestrians

d) Insufficient education for traffic morals

e) Insufficient environmental control

d) difficulty for access to other facilities

(2) Fear and anxiety against possible accidents

during an earthquake

a) vulnerability of facilities

b) vulnerability of cars

c) unexperienced response

d) Insufficient evacuation facilities

e) insufficient evacuation control systems

f) insufficient disaster information systems

g) unexpected obstacles on highways

h) vibration of tunnel

8-4 Basic Countermeasures against Earthquake

Disaster

It is the most important to take

countermeasures against the most fundamental

factors causing fear and anxiety of drivers

during an earthquake. Because the

contribution of each factor Is different, the

countermeasure have to be taken considering

their effectiveness. The Analytical Hierarchy

Process Analyses were made to Identify the

predominant factors for causing fear and

anxiety of drivers, and It was found from such

analyses that the following measures may be

effective :

(1) Simulation of environments of on-ground

155
highways

It Is basic to simulate the environments of

on-ground highways so as to prevent drivers to have strong feeling to be driving on the

underground highways. It Is effective to make

the space large enough so that mental pressure

due to the sense of closing and Isolation be

small. It Is effective to have sun-llght at underground.

(2) Positive utilization of advantage of

underground

Because ground shaking at underground Is smaller than that at the ground surface,

positive education of this feature may be

Important. It Is also Important to educate that

the evacuation to the ground surface Is not

the only one choice for drivers during an earthquake, because underground structures

are sometimes more safe that the on-ground

structures.

(3) Disaster education and disaster prevention

training Because there are various fear and anxiety

which come from unawareness of disaster

prevention systems such as the unawareness of

where the disaster prevention facilities are, more efforts are required to educate drivers on disaster prevention systems. Distribution

of pamphlet for educating the disaster

prevention systems may be essential.

Construction of disaster prevention center,

which Is in most cases constructed at the corner of the underground space difficult to see from drivers, at the place where drivers can see easily may be an effective way for

obtaining the trust from drivers. Positive

advertisement of such disaster prevention

systems may be effective.

(4) Traffic control systems

Nice traffic control systems are Important.

It Is required to shut-down the traffic to enter the underground highways after a significant earthquake.

(5) Emergency Facilities

Reliable access to the ground surface

which can be used even after a significant

earthquake Is required. However as was pointed out In (2), It Is not the only one choice for

drivers to evacuate to the ground surface. It may be more reliable to construct stable

shelter In underground so that drivers could evacuate Into the shelter In case of emergency.

(6) Formulation of dlsjister Information system

It Is required to provide the disaster

Information system for showing what are happening on at on-ground soon after the occurrence of an earthquake. Such monitoring

system of the on-ground scenery may be also

effective at normal time to mitigate the fear

and anxiety to be in underground. It should be

noted however that the Information systems

which can provide Information by only letters

and voice are not sufficient enough, because

Insufficient amount of Information Is likely to amplify the fear and anxiety of drivers at underground. Visible monitoring systems Is the most effective way.

9 CONCLUDING REMARKS

The preceding pages presented the outline

of the Guidelines for Seismic Design Methods of

Large Underground Structures. Because it Is anticipated that more use of underground

space for various purposes such as underground highways and parking Is going to be made due to various restrictions of

utilization of on-ground land, active use of

the Guidelines for designing large underground

structures Is expected.

ACKNOWLEDGEMENTS

Various supports and guidance were

obtained for compiling the Guidelines. The

authors express their sincere appreciations to the participants of the Joint Research Program

on Development of Seismic Design Method of

Underground Structures between the Public

Works Research Institute, ACTEC and 8 private

firms. The authors also express their

appreciation to Professor Kenzo Toki, Disaster

Prevention Research Institute of the Kyoto

University for his Invaluable suggestions

throughout the research.

REFERENCES

1) Public Works research Institute and

Advanced Construction Technology Center: Investigation on Seismic Design Methods of

Large Underground Structures, 1989. 1990, 1991, 1992 (In Japanese)

2) Public Works Research Institute : Guidelines

for Seismic Design Methods of Underground

Structures, Technical Note. March 1992 (In Japanese)

3) Japan Road Association ; Technical Standard

for Petroleum Pipeline. June 1971 (In Japanese)

4) Japan Water Supply Association : Selsmln

156
Design Specification for Water Supply

Facilities, December 1979 (In Japanese)

5) Japan Road Association : Design Method of

Common Utility Ducts, March 1986 (In Japanese)

6) Ministry of Construction : New Seismic Design

Methods, Technical Note. No. 1185, Public Works

Research Institute, March 1977 (In Japanese)

7) Shlba, Y., Kawashlma, K., Oblnata, N. and

Kanoh, T. : Evaluation Procedure for Seismic Stress Developed In Shield Tunnel Based on

Seismic Deformation Method, Proc. Japan

Society of Civil Engineers, Vol. 404, April 1989

(In Japanese)

8) Shlba, Y.. Kawashlma, K., Oblnata, N. and

Kanoh, T. : An Evaluation Method of

Longitudinal Stiffness of Shield Tunnel Lining

for Application to Seismic Response Analysis, Proc. Japan Society of Civil Engineers. Vol.

398, October 1988 (In Japanese)

9) Kawashlma, K., Suglta, H. and Kanoh, T. : Loading Test of Shield Tunnel Specimens

subjected to Alternative Axial Loading, 4th

U.S. National Conference on Earthquake

Engineering, Palm Springs, CA., U.S.A., May 1990

10) Kawashlma, K., Hasegawa, K. and Nagashlma,

H. : Seismic Behavior of Burled Pipelines

through Field Observation, 3rd U.S. National

Conference on Lifeline Earthquake Engineering, Los Angeles, CA., U.S.A., August 1991 IDKawashlma, K., Suglta, H. and Kanoh. T. : Effect of Earthquake on Driving of Vehicles

Based on Questionnaire Survey. Proc. Japan

Society of Civil Engineers. Structural

Eng./Earthqueike Engineering. Vol. 6. No. 2, October 1989

157

Seismic Vulnerability and Impact of Disruption of Lifelines in the Conterminous

United States— An Overview

by

William S. Bivins

ABSTRACT

The Federal Emergency Management Agency

(FEMA) initiated a project to assess the seismic

vulnerability and impact of disruption of lifeline systems nationwide. The purpose of the project was

to develop a better understanding of the impact of

disruption of lifelines from earthquakes and to assist in the identification and prioritization of hazard

mitigation measures and policies. FEMA plans to utilize results from the project to promote national awareness of the importance of protecting lifeline systems from earthquakes, and assuring reliability

and continued serviceability of lifelines. The project was undertaken by the Applied Technology Council

(ATC) and was conducted in several phases. Phase

I, which provides the bases for this paper, provides

a national overview of lifeline seismic vulnerability

and impact of disruption. Lifelines considered

include electric systems, water systems,

transportation systems, gas and liquid fuel supply

systems, and emergency service facilities. The

vulnerability estimates and impacts developed are presented in terms of estimated direct damage losses and indirect economic losses. The losses are considered to represent afirst approximation because

of the assumptions and methodology utilized, because

several lifelines are not included, and because, in some cases, the available lifeline inventory data lack

critical information. This paper is an overview of the project drawn from the full report. Seismic

Vulnerability and Impact ofDisruption ofLifelines in the Conterminous United States — ATC-25, Applied

Technology Council, (ATC, 1991).

KEYWORDS: lifelines, earthquake, damage,

impacts

1. INTRODUCTION

Lifeline is a term denoting those systems necessary

for contemporary life and urban function, without

which large urban regions cannot exist. Lifelines

basically convey food, water, fuel, energy.

information, and other material necessary for modem

human existence from the production areas to the consuming areas. Also included are, of course, the transportation of the modem human.

With the advent of more and more advanced

technology, the United States has increasingly

become dependent on the reliable supply of lifeline- related commodities, such as electric power, fuel, and

water. Earthquakes are probably the most likely natural disaster that would lead to major lifeline dismption. Extended disruption of lifelines serving

a city or major urban region would lead to major

economic losses, deteriorated public health and

probable population migration. It clearly became

necessary to identify the potential for major

dismption to these lifelines, especially at the regional

level. FEMA initiated the study to better understand the impact of dismption of lifelines from earthquakes and

to assist in the identification and prioritization of

hazard mitigation measures and policies. The study

report, Rojahn (1991), from which this paper is extracted, will improve national awareness of the importance of protecting lifeline systems from

earthquakes, and of assuring lifeline reliability and

continued serviceability.

Lifelines of critical importance at the U.S. national

level have been analyzed to estimate overall seismic

vulnerability and to identify those lifelines having the

greatest economic impact, given large credible

earthquakes. The lifelines examined include electric systems; water, gas and oil pipelines; highways and

bridges; airports; railroads; ports; and emergency

service facilities. The vulnerability estimates and

impacts developed are presented in terms of estimated

direct damage losses and indirect economic losses. These losses are considered to represent a first approximation because of the assumptions and

'Federal Emergency Management Agency,

Washington, DC 20472

159
methodology utilized, because several lifelines are not

included, and because, in some cases, the available

lifeline inventory data lack critical capacity

information.

Four basic steps were followed to estimate lifeline damage and subsequent economic disruption for given

earthquake scenarios. The steps were:

• Develop a national lifeline inventory database,

• Develop seismic vulnerability functions for each

lifeline component/system,

• Characterize and quantify the seismic hazard

nationwide, and

• Develop direct damage estimates and indirect economic loss estimates for each scenario

earthquake.

Several problems were encountered that could not be

resolved because of technical difficulties and lack of

available data. For example, telecommunication

systems, nuclear and fossil-fuel power plants, dams,

and certain water, electric, and transportation facility

types at the regional transmission level were excluded

from consideration in this project because of the unavailability of inventory data or the need for more

in-depth studies.

Interaction effects between lifelines, secondary

economic effects, and damage resulting from

landslide were also not considered in developing the project report. These, and other limitations, tend to overestimate the losses. Lack of capacity information

for most lifelines was also a definite limitation. In aggregate, due primarily to the exclusion of certain systems, the estimates of losses are believed to be

quite conservative.

2. NATIONAL LIFELINE INVENTORY

Development of the inventory for all major lifelines in the United States was a major task. Sufficient

detail was needed for lifeline seismic vulnerability assessment and impact of disruption at the national

level. This required that the inventory be compiled

electronically in digital form and dictated the inclusion of lifelines only at the transmission level.

The investigators initially contacted a number of

government, utility, trade and professional

organizations, and individuals to identify nationwide

databases, especially electronic databases. In most

cases, these organizations referred the project back to FEMA, since they had either previously furnished the information to FEMA, or know that the data had

been furnished to FEMA by others. FEMA's

database became a major source of data for several of

the lifelines. Unfortunately, much of the inventory

data generally date from about 1966.

The network inventory contained in the database is generally at the higher transmission levels, as opposed to lower distribution levels. As an analogy,

the inventory contains only the national arterial level, and neglects the capillary system. For example, all federal and state highways are inventoried, but county

and local roads are not.

2.1 Inventory Overview. The inventory data have

been compiled into an electronic database consisting

of:

• digitized location and type of facility for singlesite lifeline facilities (e.g., ports), and

• digitized right-of-way and limited information on

facility attributes for network lifelines (e.g.,

longitude and latitude of each end of a pipeline

segment with diameter).

The inventory is only partial, in that important

information on a number of facility attributes (e.g.,

number or length of spans for highway bridges) was

unavailable in the FEMA database.

TTie inventory data include information only on the conterminous United States because lifelines in other

regions would not be affected by the scenario

earthquakes considered in this study.

2.2 Inventory Summary. The specific lifelines that have been inventoried are:

• Transportation

- Highways (489,890 km of highways; 144,785

bridges)

- Railroads (270,611 km of right-of-way)

- Airports (17,161 civil and general aviation

airports)

160
• Energy

- Electric Power Transmission (4,551

substations; 441,981 km of transmission lines)

- Gas and Liquid Fuel Transmission (77, 109 km

of crude oil pipelines; 85,461 km of refined oil pipelines; 67,898 km of natural gas pipelines)

• Emergency Service Facilities

- Emergency Broadcast Facilities (29,586

stations)

- Hospitals (6,973 medical care centers)

• Water Aqueducts and Supply (3,575 km of

aqueduct, excluding Utah for which data were

unavailable).

Telecommunication, an important lifeline, was

excluded because of the unavailability of data. In addition, data on nuclear reactors and dams are excluded because it was believed that such facilities should be the subject of special studies because of the

existing regulations relating to seismic safety in many

regions and the expected complexity of the performance and impact of these facility types. Thus

the losses established by this study will be

underestimated to the extent that these facility types

are not included.

Also excluded from the inventory, but included in the

analysis, are distribution systems at the local level. The number of facility types (water, highway, and

electrical systems and police and fire stations) with in each 25-km by 25-km grid cell was estimated on the

basis of proxy by population.

3. LIFELINE VULNERABILITY FUNCTIONS

The second step in the project was the development

of lifeline vulnerability functions. These functions

which describe the expected or assumed earthquake

performance characteristics of each lifeline as well as

the time required to restore damaged facilities to their pre-earthquake capacity. The development of the vulnerability function for each lifeline considered are too lengthy to detail here but consisted of the following components:

• General information, which comprises (1) a description of the structure and its main

components, (2) typical seismic damage in qualitative terms, and (3) seismically resistant

design characteristics for the facility and its components. This information was included to define the assumed characteristics and expected

performance of each facility and to make the functions more widely applicable.

• Direct damage information, which consists of (1) a description of its basis in terms of structure type

and quality of construction, (2) default estimates

of the quality of construction for present

conditions and corresponding motion damage

curves, (3) default estimates of the quality of

construction for upgraded conditions, and (4) restoration curves.

These functions reflect the general consensus among

practicing structural engineers that, with few

exceptions, only California and portions of Alaska

and the Puget Sound region have had seismic

requirements incorporated into the design of local

facilities for any significant period of time. For all other areas of the United States, present facilities are assumed to have seismic resistance less than or equal

to that of equivalent facilities in California.

The two key quantitative vulnerability-function

relationships developed under this projectmotiondamage curves and restoration curves—d&fmo,

expected lifeline performance for each of the study

regions and for the heart of the quantitative

vulnerability analysis. The curves are based on the data and methodology developed on the basis of

expert opinion in the ATC-13 project (ATC 1985).

The data and methodology of ATC-13 are applicable

for California structures only and the data were

revised and reformatted to reflect differences in seismic design and construction practices nationwide.

Motion-damage curves developed define estimated

lifeline direct damage as a function of seismic

intensity (i.e.. Modified Mercalli Intensity); direct damage is estimated in terms of repair costs expressed as a fraction or percentage of value. Curves are provided for each study region.

Restoration curves developed for this project define

the fraction of initial capacity of the lifeline as a function of elapsed time since the earthquake. Again,

curves are defined for each region.

161
4. SEISMIC HAZARD

Seismic hazard, as used in the study, is the expectation of earthquake effects. It is usually

defined in terms of ground shaking parameters (e.g.,

peak ground acceleration, Modified Mercalli

Intensity, peak ground velocity) but, broadly

speaking, can include fault rupture, ground failure, or

other phenomena resulting from an earthquake.

Seismic hazard is a function of the size, or magnitude

of an earthquake, distance from the earthquake, local

soils, and other factors, and is independent of the buildings or other items of value that could be

damaged.

The technical approach for evaluating the seismic

hazard of lifeline structures are defined in detail. The approach involves identifying (1) the most

appropriate descriptor (parameters) for describing the seismic hazard, (2) regions of high seismic activity,

(3) representative potentially damaging, or

catastrophic, earthquakes within each of these regions

that could be used as scenario events for the investigation of lifeline loss estimation and disruption,

and (4) a model for estimating the seismic hazard for each of these scenario events.

Following a review of available parameters for characterizing seismic hazard, the Modified Mercalli

Intensity (MMI) Scale was elected for use. The

MMI scale is subjective; it is dependent on personal

interpretations and is affected by the quality of

construction in the affected area. Even though it has

these limitations, it is still useful as a general

descriptor of damage, especially at the regional level.

The study presents an overview of the seismicity of

the United States and representative earthquakes that were selected for the study.

In order to estimate the seismic hazard of the scenario events over the affected area, a model of

earthquake magnitude, attenuation, and local site effects is required. For this study the Evemden

model (Evemden and Thomson-1985) was selected. The advantages included: (1) verification via comparison with historical events, (2) incorporation

of local soil effects and ready availability of a nationwide geologic database, and (3) ready

availability of closed-form attenuation relations. An

important additional attribute was that the Evemden

model would estimate the distribution and intensity of

seismic shaking in terms of MMI, the shaking

characterization used in the ATC-13 study and the basic parameter, in this study, for lifeline vulnerability functions.

Based on the review of representative earthquakes

scenario events were selected for this investigation.

These events, presented by Region, Event Name, and

Magnitude, are:

• Northeastem, Cape Ann, 7 • Southeastem, Charleston, 7.5 • Central, New Madrid, 7 and 8 • Westem Mountain, Wasatch Front, 7.5 • Northwestem, F*uget Sound, 7.5 • Southern California, Fort Tejon, 8 • Northern California, Hayward, 7.5.

With the exception of the Cape Ann, Charleston, and

Hayward events, all magnitudes are reflective of the representative earthquake for the region. The

scenario events for Cape Ann, Charleston, and

Hayward have magnitudes one-half unit higher than

the representative event. These magnitudes are

interpreted as maximum credible for these locations.

5. ESTIMATES OF DIRECT DAMAGE

The analysis of seismic vulnerability of lifeline systems and the economic impact of dismption is based on an assessment of three factors:

• Seismic hazard, • Lifeline inventory, and • Vulnerability functions.

In this study these factors are used to quantify

vulnerability and impact of dismption in terms of (1) direct damage and (2) economic losses resulting from

direct damage and loss of function of damaged

facilities.

Direct damage is defined as damage resulting directly

from ground shaking or other collateral loss-causes such as liquefaction. For each facility, it is expressed

in terms of cost of repair divided by replacement cost and varies from 0 to 1.0 ()% to 100%). It is estimated using (1) estimates of ground shaking

intensity provided by the seismic hazard model, (2) inventory data specifying the location and type of

facilities affected, and (3) vulnerability functions that

162
relate seismic intensity and site conditions to expected

damage.

significant. However, there is very little literature on

indirect economic losses.

The approach to estimate direct damage considers

both damage resulting from ground shaking as well as damage resulting from liquefaction. Damage due to other collateral loss causes, such landslide and fire, are not included because of the unavailability of

inventory information and the lack of available

models for estimating the losses nationwide.

The analysis approach for computing direct damage

due to ground shaking proceeded as follows. For

each earthquake scenario, MMI levels were assigned

to each 25-km grid cell in the affected region, using

the Evemden MMI model, assigned magnitude, and

assigned fault rupture location. Damage states were

then estimated for each affected lifeline component in each grid cell, using the motion-damage curves.

Dollar loss estimates are based on the available

inventory data, cost per facility assumptions, and

other models and assumptions with are described in the report. As a result, the accuracy of these estimates may vary from lifeline to lifeline. Data from the various lifelines are combined to estimate

the total direct damage dollar losses for the eight

scenario earthquakes as follows:

Earthquake

Direct Dollar Loss

(in Billions. 1991)

Cape Aim $4.2

Charleston $4.9

Fort Tejon $4.9

Hayward $4.6

New Madrid, M= 8.0 $11.8

New Madrid, M = 7.0 $3.4

Puget Sound $4.4

Wasatch Front $1.5

6. ESTIMATION OF INDIRECT ECONOMIC

EFFECTS

Earthquakes produce both direct and indirect economic effects. The direct effects, such as loss due to collapsed structures, are obvious and dramatic.

However, the indirect effects that these disruptions

have on the ability of otherwise undamaged

enterprises to conduct business may be quite

This study provides a first approximation of the

indirect economic effects of lifeline interruption due

to earthquakes. A methodology was developed to relate lifeline interruption estimates to economic

effects of lifeline interruption in each economic

sector. This required a two-step process: (1) development of estimates of interruption of lifelines as a result of direct damage and (2) development of

estimates of economic loss as a result of lifeline interruption.

Lifeline interruption resulting from direct damage is quantified in the study in residual capacity plots that define percent of function restored as a function of

time. The curves are estimated for each lifeline type

and scenario earthquake using (1) the time-torestoration curves, (2) estimates of ground shaking

intensity provided by the seismic hazard model, and

(3) inventory data specifying the location and type of

facilities affected.

For site-specific systems (i.e., lifelines consisting of

individual sited or point facilities) the time-torestoration curves are used directly. Extended

regional networks required special analyses

procedures consisting of connectivity analyses and

serviceability analyses.

The extensive analyses of the study yields the following estimates of total indict economic losses for the eight scenario earthquakes as follows:

Earthquake

Indirect Loss

(in Billions. 1991)

Cape Ann $9.1

Charleston $10.2

Fort Tejon $11.7

Hayward $11.1

New Madrid, M= 8.0 $14.6

New Madrid, M = 7.0 $4.9

Puget Sound $6.1

Wasatch Front $3.9

7. HAZARD MITIGATION OF CRITICAL

LIFELINES

Based on the combined direct and indirect economic

losses presented in the study and with consideration

163
of the assumptions and limitations expressed

throughout the report the following relative ranking

of the criticality of different lifelines in terms of the estimated impact of damage and disruption:

Rank 1 Electric Systems

Rank 2 Highways

Rank 3 Water Systems

Rank 4 Ports Rank 5 Crude Oil.

The study also provides extensive recommendations on measures to reduce the vulnerability of lifeline systems. The recommendations fall into the following categories:

• Damage reduction measures (e.g., strengthening a building, densifying the soil beneath a structure, component improvements)

• Provision of system redundancy

• Operational improvements

9. RECOMMENDATIONS FOR FURTHER WORK

The ATC-25 project has raised a number of questions

and indicated areas in which knowledge is inadequate

or nonexistent with respect to the impact of lifeline disruption due to earthquake. Following are recommendations for further research and other

efforts.

• Lifeline Inventory. Appropriate organizations are encouraged to develop standards for complete

lifeline inventories, and coordinate the acquisition

of the needed additional updated data from various

lifeline owners.

• Lifeline Component Vulnerability. A major effort should be initiated to acquire data on lifeline seismic performance and damage, and conduct

analysis towards the development of improved

component vulnerability functions. This effort should also investigate lifeline recovery data, and

incorporate the extensive experience realized during the October 17, 1989 Loma Prieta,

California, earthquake, as well as from other

damaging earthquakes.

• Seismic Hazard Data. The U.S. Geological

Survey should develop a series of digitized

soil/geologic databases.

• Economic Analysis and Impacts Data and

Methodology. Further research is recommended

in economic areas such as:

- Economic impacts associated with lifeline disruption,

- Second-order economic effects (e.g., interaction between lifelines),

- Elasticities of demand, or substitution of lesser disrupted lifeline for a more disrupted lifeline,

- Inter-regional impacts, and

- So-called "benefits," such as increased

economic activity associated with repair or replacement of older equipment with new

technology.

It should be noted that this study did not address the environmental consequences associated with lifeline disruption, especially the potential for oil spills from

broken pipelines in the nations' waterways following,

for example, a New Madrid event.

10. REFERENCES

Applied Technology Council (ATC-25), 1991,

Seismic Vulnerability and Impact ofDisruption of

Lifelines in the Conterminous United States, Report ATC-25, Redwood City, California.

Applied Technology Council (ATC-13), 1985,

Earthquake Damage Evaluation Data for

California, Report ATC-13, Redwood City, California.

Evemden, J.F., and J.M. Thomson, 1985, Predicting

Seismic Intensities, in Evaluating Hazards in the Los Angeles Region — An Earth Science

Perspective, J.I. Ziony, ed., US Geological

Survey Professional Paper 1360, U.S. GPO,

Washington D.C.

164
Design Criteria for Base Isolation of Buildings

by

Harry W. Shenton, III*

ABSTRACT

Presented are the results an investigation in

which the performance of "code-designed" fixedbase and base-isolated structures were compared.

Two frames, a steel concentrically braced frame

and a steel special moment resisting frame were

designed as both fixed-base and base-isolated in accordance with the 1991 Uniform Building

Code. To establish comparable performance, the

base-isolated frames were designed to varying

fractions of the code recommended base shear:

100%, 50% and 25%. Non-linear time history

analyses were conducted, for three ensembles of

earthquakes, to assess the effect of design force

level and base-fixity on response and damage

level. In general, the base-isolated frames designed to 100% of the code recommended base

shear out-performed the fixed-base frames.

Based on average values of a number of re- sponse quantities, comparable performance could

be achieved at design forces lower than currently

specified by code for the base-isolated frames.

KEYWORDS: base-isolation; building code;

dynamic response; relative performance; steel frame.

1. INTRODUCTION

This paper presents the results of an analytical

investigation in which the performance of "codedesigned" fixed-base and base-isolated structures are compared. The objective of the study is to determine approximately the design force level for a base-isolated structure which will result in

performance comparable to that of a fixed-base

structure.

Two simple planar frames, a steel concentrically-braced frame (SBF) and a steel special

moment-resisting frame (SMF) have been designed using the static lateral load procedure, in

accordance with the 1991 Uniform Building

Code(UBC). The frames have been designed as both fixed-base and base-isolated. Design parameters for the frames have been selected to represent a large part of the building inventory:

frames are 4-story, 3-bay structures, assumed to be located in Seismic Zone 4 (PGA =0.4^), and

founded on dense or stiff soil (UBC soil profile

S2). The base-isolated buildings are assumed to be supported on laminated elastomeric isolation

bearings, designed to provide an isolated period

of 2.5 s, with equivalent viscous damping equal

to 15% of critical.

To establish comparable performance, each baseisolated frame has been designed to 100%, 50%

and 25% of the code recommended base shear.

The response of the fixed-base design on the

isolation system has also been considered (this is equivalent to designing an isolated structure to greater than 100% of the code recommended

base shear). To facilitate reference to the various

design cases, the following nomenclature has

been adopted:

FRAME TYPE - DESIGN CASE

where FRAME TYPE is SBF or SMF, and

DESIGN CASE is FB (fixed-base), BI.O (fixedbase design on isolators), BI.l (base-isolated,

designed to 100% of the code recommended

base shear), BI.2 (base-isolated, 50% of code)

and BI.3 (base-isolated, 25% of code). Thus,

SMF-BI.2 refers to the steel moment frame

designed to 50% of the code recommended

lateral force.

To evaluate the performance of each design,

nonlinear time history analyses have been conResearch Structural Engineer

National Institute of Standards and Technology, Gaithersburg, Maryland 20899

165
ducted, for 3 ensembles of earthquakes, using

the computer program DRAIN-2D (Kanaan and

Powell, 1973). A number of response quantities

have been examined, including, the number of

frames that exhibit yielding or collapse, the

number of yielded superstructure elements, the

maximum total and relative roof displacement of

the superstructure, first story drift and the

maximum displacement of the base-isolation

bearing. Statistical analyses of these quantities

have been conducted to determine the approximate design force level for the base-isolated

structure which would provide performance

comparable to that of the fixed-base structure.

in which Z is the seismic zone factor, I is the

importance factor, is a reduction factor

which is measure of the ability of a structure to sustain inelastic deformations without collapse,

W is the seismic dead weight, and C is given

by.

c = 1.25S

•2/3 ^ 2.75 (2)

where S is the site soil coefficient, and T is the

fundamental period of the structure, in seconds.

T is approximated by

T = (hj- (3)

2. FRAME DESIGNS

2.1 Building Configuration

where h„ is the height of the building in feet and

C, = 0.02 for the braced frame and 0.035 for the moment frame.

The frames under consideration are assumed to be part of a hypothetical four-story office building, which is rectangular in plan, measuring

nominally 45.7 m by 22.9 m with 7.62 m bays

in both directions. Story heights are 3.66 m for

the first story, and 3.35 m for the second, third,

and fourth stories. The roof and floors are assumed to be rigid and to act as diaphragms in the horizontal direction. Alternating frames are designed to carry the lateral load; hence, the

tributary width for lateral load is 15.3 m and for

gravity load is 7.62 m. The plan and a typical

section of the building are shown in Figure 1.

2.2 Dead and Live Load

Frames are designed for a roof dead load of

2.75 kPa and a floor dead load of 3.92 kPa. The

corresponding live loads are 0.9 kPa and 2.40

kPa, respectively. The total seismic dead weight

(W) is 5190 kN for the fixed base frames and

6570 kN for the base-isolated frames.

For the fixed-base frames: Z=0.4, 1=1, S = 1.2,

and Rw = 12 for the moment frame and Rw =

8 for the braced frame.

2.4 Base Shear, Base-Isolated Design

The design base shear for elements above the

isolation plane is given by (UBC, 1991)

K D

R

(5)

Wl

in which D is the design displacement of the

isolation system, is the maximum effective

stiffhess of the isolation system at the design

displacement D, as normally determined by

experiment, and R^i is similar to Rw, although

not equal in value. The design displacement D,

in inches, of the isolation system is given by

D =

10 Z N S. T. B

(6)

2.3 Base Shear, Fixed-Base Design

The design base shear, V, for a fixed-base frame

is given by (UBC 1991),

Z I C V = w

R«/

(1)

in which Z is the seismic zone factor; N is a

coefficient related to the proximity of the building to the nearest active fault; S, is the site soil coefficient for isolated structures; B is a coefficient related to the effective damping of the

isolation system, and T^ is the fundamental

period of the isolated structure. The fundamental

166
period of the isolated structure is related to K^,

the minimum effective stiffness of the isolation

system, by

= 271 W(6)

In this investigation, the isolation bearings are modeled as bi-linear hysteretic elements as shown in Fig. 2. The bi-linear model is fairly

general and represents sufficiently well the

experimentally observed cyclic load-deformation

behavior for many types of bearings. The model

is defined by initial stiffness Kj, post-yield

stiffness Kj and yield force Fj. Assuming yielding of the isolation system occurs at 5%W and

the post-yield stiffness is equal to 10% of the

pre-yield stiffness W - 0M5—

D

(7)

where Kg is the effective stiffness of the isola- tion system at the design displacement D (Fig-

.2). Note that with the assumed bi-linear model,

K_=K^=Ke (UBC, 1991).

For the base-isolated frames: Z=0.4, Si=1.5,

N=l, Ti=2.5 s, B=1.35, and Rw,=3 for both

the braced frame and moment frame. This

results in a design displacement for the isolation

system of 280 mm.

2.5 Final Designs

A sunmiary of the design specified or computed

fundamental period and base shears for the

various design cases are listed in Table 1. Note,

under current guidelines, the moderate 43%

reduction in base shear for the base-isolated

braced frame, and the mere 14% reduction for

the base-isolated moment frame.

Presented in Table 2 are member sizes for the

braced frames, and in Table 3 for the special

moment frames. All members are standard

structural sections, per the American Institute of

Steel Construction. Note, due to a limitation on minimum member sizes, no reduction in member

size was achieved for the braced frame designed

to 25% of the code recommended base shear,

hence, there are only 3 distinct isolated designs

for the concentric braced system (BI.O, BI.l and

BI.2).

3. EARTHQUAKE TIME HISTORIES

For the dynamic analysis, 18 recorded earth- quakes have been selected from the original 30

used by Seed et al. (1976) to develop the Sj

design spectrum of the design provisions. The

18 have peak ground accelerations ranging from

0.11^ to 0.28^, and peak ground velocities

ranging from 120 mm/s to 360 mm/s. The

original 18 form the "As-recorded" ensemble.

To conform to the zone 4 design spectrum PGA

of 0.4^, a second ensemble has been created by

scaling each record to a peak ground acceleration of 0.4g. This is referred to as the "PGA"

ensemble. The influence of velocity is examined

by scaling each record to a peak velocity of 500

mm/sec, which is the average PGV of the en- semble after it has been scaled to 0.4^. This is referred to as the PGV ensemble.

4. METHOD OF ANALYSIS

Models were developed for the SBF and SMF

for analysis using DRAIN-2D (Kanaan and

Powell, 1973). In all cases, nodes are located at the geometric beam-column centerline intersection, seismic mass is lumped at the beam-column

joint, joints are assumed to be infinitely rigid

and a single element is used to model each story

column and each girder span.

In all cases, beams and columns are modeled

using a beam/column element with typical A36

steel properties (E = 200 GPa, Fy=248 MPa) and

nominal section properties. Beams and columns

are assumed to exhibit an elastic-plastic moment

curvature relation, with no hardening in the

SBF, and 2% hardening in the SMF. Braces are modeled using buckling truss elements (Jain and

Goel 1978) with typical A46 steel properties

(E=200 GPa, Fy=317 MPa). Isolators are modeled using bilinear, elastic-plastic springs located

at the base of each first story column. The

stiffness of the individual isolators are a fraction

of the overall isolation system properties, based on the proportion of dead load carried by each

167
column. A schematic of the DRAIN-2D model

for the SBF is shown in Fig. 3, and for the SMF

in Fig. 4.

In the time history analyses, P-A effects are considered approximately, the dynamic analysis

is performed from the deformed static equilibrium position and original elastic stiffness proportional damping, corresponding to 2% in the

fundamental mode, is assumed. Analyses have

been conducted for the first 30 seconds of each

strong motion record in the 3 ensembles, for the

9 design cases. A total of 486 time history

analyses have been conducted.

5. RESULTS

Of the 3 ensembles used in the investigation, the

PGA ensemble was the most "damaging"; only

the results for the PGA ensemble are presented

herein. A compilation of all the results may be

found in Lin and Shenton (1992). A summary of

response statistics for the steel braced frames are presented in Table 4, and for the steel moment

frames in Table 5.

5.1 Braced Frame Designs

The "code-designed" base-isolated frame (BLl)

clearly out-performed the fixed-base frame (FB)

in almost every response category, as noted in Table 4. This includes the number of frames

which exhibit yield (a measure of the incidence

of superstructure yielding for a given frame and

ensemble), average number of yielded superstructure elements (a measure of the extent of

superstructure yielding), time of first superstructure yield and average relative roof dis- placement. In only one response category did the

SBF-BI.l not perform better than the SBF-FB,

that being first story drift ductility (defined as the first story drift divided by the story drift required to cause first yield in the brace). The

average first story drift for SBF-BI.l was 31

mm, while that for the SBF-FB was 28 mm,

resulting in a comparable, but slightly larger

drift ductility for the BI.l.

As would be expected, the absolute roof dis- placement for the isolated structures is greater

than that of the fixed base design. The total roof

displacement is also seen to be relatively insensitive to design force level. Note that the average

relative roof displacement for SBF-BI.l is roughly 1/3 the total roof displacement, indicating near rigid body behavior in the superstructure.

The results in Table 4 are presented in an alter- nate format in Fig. 5. Plotted in the figure is the

isolated response quantity normalized by the

fixed-base response quantity, versus the fraction

of code recommended base shear. Comparable

performance in this figure corresponds to an

ordinate value of 1.0. It is obvious from this figure that the design force level required to provide comparable performance varies with

performance criterion. The isolated design BI.2

out-performed the fixed-base design when based on incidence and extent of yield; a design force

of approximately 85% of code would be re- quired to provide comparable performance when

based on relative roof displacement; the codedesigned fixed-base frame performed slightly

better than the code-designed base-isolated frame

when based on drift ductility.

The average peak isolator displacement for the

PGA ensemble was 224 mm, 173 mm and 138

mm, for SBF-BI.O, SBF-BI.l and SBF-BI.2,

respectively. In every case, the, average is below

the design displacement of 280 mm; however,

the design displacement was exceeded in 10 of

54 analyses conducted for the isolated braced

frame designs. The largest isolator displacement

for SBF-BI.l and the PGA ensemble was 538

mm, almost twice the design displacement.

5.2 Moment Frame Designs

The "code-designed" base-isolated moment

frame (SMF-BI.l) out-performed the fixed-base

frame (SMF-FB) in a number of categories;

however, performance was only comparable in others. The SMF-BI.l out-performed the SMFFB when based on average number of yielded

superstructure elements, relative roof displacement and first story drift ductility, as noted in Table 5. The two performed comparably when

based on incidence of collapse (defined by an

168
extreme roof displacement and excessive number

of yielded members) and extent of superstructure

yielding.

It is interesting to note that in this case the

average total roof displacement of SMF-BI.l is slightly lower than that of the fixed-base design.

Also, the average relative roof displacement for

SMF-BI. 1 is roughly 2/3 the total roof displacement. This, in conjunction with the incidence

and extent of superstructure yielding for SMFBI.l, indicates superstructure behavior which is far from rigid or purely elastic, a common

assumption in the analysis of isolated structures.

The results in Table 5 are presented in the

alternate normalized format in Fig. 6. In this case it is clear that comparable performance is achieved, for almost all response criteria with

the PGA ensemble, with an isolated moment

frame designed to between 50% and 75% of the

code recommended base shear.

The average peak isolator displacement for the

PGA ensemble was 150 mm, 139 mm, 100 mm

and 76 mm, for SMF-BI.O, SBF-BI.l, SBF-BI.2

and SMF-BI. 3, respectively. Again, the average

is below the design displacement of 280 mm;

however, the design displacement was exceeded

in 9 of 72 analyses conducted for the isolated

moment frame designs. The, largest isolator

displacement for SMF-BI. 1 and the PGA ensemble was 394 mm, just 1.4 times the design displacement.

6. CONCLUSIONS

In this study the relative performance of "codedesigned" fixed-base and base-isolated steel frames have been compared. Results presented

indicate that, based on a number of performance

criteria, code-designed base-isolated structures perform better than conventional fixed-base

structures. Comparable performance can be

achieved with an isolated structure designed to a

lateral force level less then that required by the

1991 UBC provisions.

Comparable performance in a braced frame can be achieved with a design force level between

50% and 100% of the code recommended value:

the actual level of design force varies with

- response criterion. Comparable performance in a moment frame can be achieved with a design

force level between 50% and 100% of the code

recommended value. A lateral force corresponding to 50% of the code specified value provides

comparable performance in many cases.

7. REFERENCES

1. Jain, A.K., and Goel, S.C. (1978), "Hysteresis Models for Steel Members Subjected to Cyclic Buckling or Cyclic End Moments and Buckling", University of Michigan Department of Civil Engineering,

Report No. UMEE 78R6. Dec.

2. Kanaan, A.E. and Powell, G.H. (1973),

DRAIN-2D User Guide, EERC Report 73-

6 and 73-22, University of California,

Berkeley, CA. Apr.

3. Lin, A.N. and Shenton III, H.W. (1992),

"Seismic Performance of Fixed-Base and

Base-Isolated Steel Frames," ASCE, Journal of Engineering Mechanics, Vol. 118,

No. 5, pp 921-941.

4. Seed, H.B., Ugas, C. and Lysmer, J. (1976), "Site-Dependent Spectra for Earthquake-Resistant Design," Bulletin of the

Seismological Society of America, Vol. 66,

No. 1, pp. 221-243, Feb.

5. Uniform Building Code (1991), International Conference of Building Officials,

Whittier, Calif

169
Table 4. Summary of Steel Braced Frame Results for PGA Ensemble

Design

# of

f ramps

that

exhibit

yield

Average #

of Yielded

Elements

Average

Timp nf

first yield

(s)

Average Roof

Hisnlappmpnt

v\/ith respect to

ground (mm)

Average Roof

HiQnlappmpnt

v^ith respect

to isolators

(mm)

Average

First finnr

ductility

ratio

FB 18 7.3 5.1 106 106 2.2

BI.O 3 1.1 15.4 253 38 1.4

BI.1 6 2.4 9.0 257 89 2.4

BI.2 12 5.1 6.9 246 159 9.4

Table 5. Summary of Steel Moment Frame Results for PGA Ensemble

Design

# of

frames

that

collapse

# of

frames

that

exhibit

yield

Average #

of Yielded

Elements

Average

Time of

first yield

(s)

Average Roof

displacement

with respect to

ground (mm)

Average Roof

displacement

with respect to

isolators (mm)

Average

First

floor

ductility

ratio

FB 0 17 34.1 5.1 149 149 2.1

BI.O 0 14 13.1 7.6 129 78 1.0

BI.1 0 17 21.7 5.3 145 96 1.2

BI.2 2 18 33.2 6.1 175 137 2.6

BI.3 5 18 33.5 6.1 184 163 4.3

170
II1-^

nXBD-BASE BASB-ISOLATSD

3. DfUIN-a> Model of Sad CoocaiBk Bntad Fnot

PDOBD-BASB BASS-ISOLATED

Fi«uc 4. DIIAIN-2D MoM of SmI Mom Pma

0.5 0.75 1 1.25 1.5 FRACTION OF COCC BASE SHEAR

Figure 5. Normalized Results for Steel Braced Frame and PGA Ensemble

3

Z. 9

2

^

in<

00

T

-m- RELATIVE ROOF DISP. #FRAMES THAT YIELD -e- #yiELDED MEMBERS

FIRST STORY DUCTILITY

0 0.25 0.5 0.75 1 1.25 1.5 FRACTION OF CODE BASE SHEAR

Figure 6. Normalized Results for Steel Moment Frame and PGA Ensemble

171

Manual for Menshin Design of Highway Bridges

by

Kazuhiko Kawashima', Kinji Hasegawa-, and Hiroyuki Nagashima'

ABSTRACT

A three-year research program on "Development of Menshin Design of Highway

Bridges" was made from April 1989 to March 1992. Concentrated efforts were paid to develop energy

dissipating devices and falllng-off prevention

devices for bridges. An appropriate design

method of highway bridges with energy

dissipating devices and favorable application of

the menshin design to highway bridges were

studied. Final accomplishment of the three-year

research was compiled In the form of "Manual for Menshin Design of Highway Bridges" In March

1992. This paper outlines the Manuals.

KEY WORDS

Menshin Design, Highway Bridge, Joint Research

Program, Manual for Menshin Design

1. nn-RODucnoN

A three-year Joint Research Program between

the Public Works Research Institute and 26 groups consisting of 28 private firms on "Development of Menshin Design of Highway

Bridges" was made from April 1989 to March 1992. This program Intended to develop a rational

seismic design method of highway bridges with

energy dissipating devices. The highway bridges

with span length from 20 m to 100 m were

considered BiS the major target for this Joint

Research Program.

As will be discussed later, "menshin" means

"reduction of response" In Japanese. Although

the menshin design Is close with the

base-Isolation, natural period of bridge is not

forcibly elongated In the menshin design, because

there are various restrictions for Increasing the

natural period. Instead of elongating the natural

period, emphasis Is placed In the menshin design

for increasing energy dissipating capability and

distribution of lateral force to as many

substructures as possible for decreasing lateral

force for design of substructures.

As shown In Table 1, concentrated efforts were paid to 1) development of menshin (energy

dissipating) devices for highway bridges, 2) development of falllng-off prevention devices

and expansion Joints appropriate for the menshin

bridges, 3) development of rational and simple

menshin design method and 4) favorable

application of menshin design to highway bridges.

In the research project, four high damping

rubber bearings, 2 sliding friction dampers, a steel damper, a roller menshin bearing a link bearing and a viscous damper were newly

developed for bridges (refer to Fig. 1). The

menshin devices for highway bridges have to be more compact and more weather-proof than the

base-Isolation devices for buildings since the

mensliln devices are Installed at narrow and

exposed crest of bridge columns.

The knock-off mechanism at an abutment to ease the Impact force Induced by the collision

between the deck and the abutment (refer to Photo 1), and a finger type expansion joint

(refer to Photo 2) which Is distinguished from

regular finger Joints by the transverse movement

as well as the longitudinal movement were also

developed.

In the research project, a simple but

rational menshin design method was developed. A

series of shaking table tests were made to study

the response of menshin bridges, and to provide

realistic data of the seismic response of menshin

bridges (refer to Photo 3). Analysis of strong

motion records measured at a menshin bridge was

also made. From the study to investigate the favorable

application of menshin design, it was found that

the menshin design Is effective for constructing

super-multl-span continuous bridge with total

deck length over 1 km. Application of the menshin

design for seismic retrofit of exiting bridges and

deck connection for making existing simply

supported girder bridges continuous was also

studied (refer to Photo 4). The reports of the project were compiled and

published by the Public Works Research Institute

In the form of the Joint Research Report. As well

as these reports, as the final accomplishment of

the three-year research program, the Manual for

Menshin Design of Highway Bridges was compiled.

This paper presents the outline of the

Manual.

2. SCOPES AND CONTEKTS OF THE MANUAL

The Manual presents the seismic design method

of highway bridges including design details when

"the Structures Expecting Reduction of inertia

Force", which Is specified In the "Part V Seismic

Design" of the "Specifications of Design of

Highway Bridges", is adopted. Application of the

1) Head of Earthquake Engineering Division,

Earthquake Disaster Prevention Department,

Public Works Research Institute, Ministry of

Construction

2) Assistant Manager of Planning Division, Road

Bureau, Ministry of Construction Proper

3) Technical Research Division, Kantoh Regional

Construction Bureau, Ministry of Construction

173
menshln design to seismic strengthening of

substructures and connection of adjacent decks

for making existing simply supported girder

bridges continuous Is also presented.

It should be noted that although the

description of the Manual Is given In the format

of the specifications, It merely represents the

accomplishments of the three-year research

program, and it Is not the mandate specifications.

The table of contents of the Manual Is as

follows:

1. General

1.1 Scope of Application

1.2 Definition of Terms

2. Basic Principle of Menshln Design

3. Menshln Design

3.1 General

3.2 Lateral Force Coefficient for Seismic

Coefficient Method

3.3 Lateral Force Coefficient and Equivalent

Lateral Force Coefficient for Bearing

Capacity Method

3.4 Characteristics value for Menshln Devices

3.5 Computation of Natural Period of Bridge

3.6 Computation of Modal Damping Ratio of

Bridges

4. Design of Menshln Devices

4.1 Basic Principle for Design of Menshln

Devices

4.2 Design of Laminated Rubber Bearings

4.3 Design of Lead Rubber Bearings

4.4 Design of High Damping Rubber Bearings

4.5 Design of Steel Dampers

4.6 Design of Friction Dampers

4.7 Design of Viscous Dampers

4.8 Design of Roller Dampers

5. Dynamic Response Analysis

5.1 General

5.2 Djmamlc Response Analysis Method

5.3 Analytical Models

5.4 Design Seismic Forces

5.5 Check of Safety

6. Design Details

6.1 General

6.2 Gap between Structures

6.3 Falllng-off Prevention Devices

6.4 Knock-off Devices

6.5 Details of Expansion Joints

7. Approving Tests of Menshln Devices

7.1 General

7.2 Approving Tests for Confirming

Requirements for Dynamic Loads

7.3 Approving Tests for Confirming

Requirements for Static Loads

8. Seismic Retrofit of Existing Bridges with Use of Menshln Design

9. Application of Menshln Design for Making

Existing Simply Supported Girder Bridges

Continuous

Appendices

1. Procedure of Menshln Design

2. Menshln Devices for Highway Bridges

3. References for Dynamic Response Analysis

4. Expansion Joints and Falllng-off Prevention

Devices for Highway Bridges

5. Design Example of Prestressed Concrete

Bridges

6. Design Examples of Steel Girder Bridges

7. Design Examples of Super-multl-span

Continuous Bridges

8. Design Examples for Seismic Retrofit

9. Design Examples for Making Existing Simply

Supported Girders Continuous

10. Standard Acceleration Ground Motions

Recommended for Djmamlc Response Analysis

3. BASIC PRINCIPLE OF MENSHIW DESIGN

Whether the menshln design should be adopted

or not has to be decided based on the advantages

of Increasing energy dissipating capability from

not only seismic safety but function at normal

time.

In general, menshln design can be adopted

with advantage for the following conditions:

1) bridges on stiff and stable soils

2) bridges with stiff substructures with short

natural period

It Is recommended that the natural period of the

bridge designed by the menshln design Is at least

1.5 times longer than that of the bridge without

menshln devices. Such elongation of the natural

period of the menshln bridge makes the coupling

vibration so small that the deformation of the

bridge be concentrated at the menshln devices.

The most Important decision of the menshln

design Is how much Increase of the natural period

be made. In Japan soils are generally very weak

at the bridge construction sites. Significant

earthquakes with magnitude over 8, which produce

long period ground motions, occur at shorter recurrence period than other countries. Taking account such evidences into account, the design

seismic force level Is take very high for design

of highway bridges In Japan In comparison with

the seismic force level adopted In U.S.A., New

Zealand and Italy. The natural period at which

the design acceleration decrease with Increasing

the natural period Is taken longer. I.e., 1.1 second for type I soils (stiff site), 1.3 second

for type n soils (medium site) and 1.5 second for

type in soils (soft soils). Therefore, the

advantage for reducing deck response can be

realized only by elongating the natural period

longer than 1.1 ~1.5 second.

However elongation of the natural period

brings the Increase of the deck response

displacement, and this requires the adoption of

Joints with long legs. This Is, however, a crucial

requirements difficult to adopt. In particular,

for overcrosslng In city area because It causes

noise and vibration pollution, which Is the most

serious concern In city area. Furthermore It will have the disadvantage for maintenance because

replacement of the damaged expansion Joint due

to traffic load as we,' as the pavement adjacent

174
to the Joint has to be frequently made. Based on these reasons, It Is not adopted to forcibly elongate the natural period in order to decrease the deck response acceleration. It Is therefore more appropriate to adopt the concept

of base Isolation not for elongating the natural

period but for Increasing energy dissipating

capability. Distribution of the Inertia force to

as many substructures as possible, which Is the

practice adopted In Japan for long time. Is also

significantly advantageous.

Therefore basic principle recommended for

introducing the base Isolation to highway bridges

In Japan is as follows : 1) Distribute the Inertia force to many

substructures so Intentionally that the

Inertia force be distributed equally by

selecting the spring stiffness of the

Isolators. Simultaneously, reduce deck

response by Increasing energy dissipating

capability by means of damper.

2) Do not forcible elongate the natural period,

but select the natural period so that the resonance of the bridge with soils be

avoided.

Such design criteria Is no more the

"Isolation" of the "base". Therefore It Is proposed to call the design concept as "menshln". The "menshln" means "reduction of seismic

response" In Japanese.

In menshln design, bridges can be designed by

following the standard static design method

(static frame analysis). More precise approve of

the seismic safety can be made by the dynamic

response analysis, If required. In such analyses,

the menshln devices (Isolators and dampers) are

Idealized by as a set of equivalent linear

springs. Equivalent stiffness and the equivalent

damping ratio of the Isolator and damper are the

major parameters used In the analyses. In the

static frame analysis, natural period of the

bridge can be computed for each seismic design

structural unit as :

T = 2.01»/~5~

/ w

(J =

U I / W I

• U I

(1)

(2)

where

T : natural period (sec)

Wi : dead weight (tf/m) of the seismic

design structural unit (super structure and substructure above the

ground surface assumed In seismic

design) at point "I" u t : lateral displacement (m) developed In the seismic design structural unit at point "1" when subjected to wi In the

direction considered In design

Damping ratio of the bridges Is computed as

h =

Z K B 1

• U B l' C r. I

c h I = h

1 +

h F0 H

K I K,

K 1

K K Ful K

where

h : Modal damping ratio of bridge

hBt : Damping ratio of 1-th damper

h PI : Damping ratio of 1-th pier/abutment

hp.jT : Damping ratio of l-th foundation

associated with translatlonal

movement

hF0 I : Damping ratio of I-th foundation

associated with rotation

Kpi : Equivalent stiffness of l-th pier/abutment

Kfui : Translatlonal stiffness of l-th foundation

Kf5 I ; Rotational stiffness of l-th foundation

u Bi : Design displacement of l-th menshln

device

H : Height from the bottom of pier to the

gravity center of deck

Eq.(3) give the approximate estimation of the

damping ratio of bridge assuming that translation

and rotation of the foundation are developed

only by the Inertia force of the deck. When the

effect of the Inertia force of piers and

abutments for evaluating the translation and

rotation of the foundation can not be

disregarded. It Is appropriate to compute the

modal damping ratio as

n

J=l ^ K • <t> \ (4)

Z Ki (3)

where

4> \ s

• Mode vector of J-th structural

component for l-th mode

h J : Damping ratio of J-th structural

component

K J : Stiffness matrix of J-th structural

component

(j> 1^ : Mode vector of bridge for l-th mode

K : Stiffness matrix of bridge

Table 2 shows the damping ratio recommended

for structural components for Eqs.(3) and (4).

4. DESIGN FORCE FOR MENSniN DESIGN

In menshln design, the menshln devices are designed by the Seismic Coefficient Method and

the Bearing Capacity Method. In both method, the

lateral force Is statically applied to the bridge,

and the seismic safety Is examined based on the

allowable stress design approach In the Seismic

Coefficient Method and bearing capacity basis

considering ductility In the Bearing Capacity

Method. Bridges are designed by the Seismic

175
Coefficient Method, and then the ductility Is checked for reinforced concrete piers by the

Bearing Capacity Method.

In the Seismic Coefficient Method, the design

lateral force coefficient k h Is given as

k h= cz - Co - c I

- ct- c e - k ho^O. 1

where

c T- c (5)

where

k h : Lateral force coefficient

c z : Modification factor for zone (refer to

Fig. 2) c Q : Modification factor for ground condition

(refer to Table 3) c 1 : Modification factor for Importance

(refer to Table 4) c T : Modification factor for structural

response (refer to Table 5) c E : Modification factor for energy

dissipation capability (refer to Table

6) k ho : Standard design horizontal seismic

coefficient (= 0.2) Those factors excluding c e are the ones

specified In the Specifications of Design of

Highway Bridges. The modification coefficient c e takes a value as shown In Table G depending on

the first modal damping ratio of the bridge.

Reduction of the design lateral force as large as

20 % Is proposed In the Manual.

In the Bearing Capacity Method, the lateral

force coefficient k ho and the equivalent lateral

force coefficient k h« are given as

k h< k ho ^0.3

^ 2</ -1

kho = Cz - C I

- Cr - CE' k hoO

(6)

(7)

where

k he : Equivalent lateral force coefficient

for Bearing Capacity Method

k ho : Lateral force coefficient for Bearing

Capacity Method

c z : Modification factor for zone (refer to

Fig. 2) c I : Modification factor for Importance

(refer to Table 4) c R : Modification factor for structural

response (refer to Table 7) c E : Modification factor for energy

dissipation capability

k hoo : Standard lateral force coefficient

for Bearing Capacity Method (=1.0) u : Allowable ductility factor of

reinforced concrete piers

The modification factors excluding c e are

the ones specified in the Specifications of Design

of Highway Bridges. The modification factor c e depends on the modal damping ratio of the

bridge, and takes a value of Table 8. Reduction

of the design force as large as 30 % Is proposed.

5. DESIGN OF MENSBIN DEVICES

5.1 Characteristics Value of Henshln Devices

In the menshln design, the displacement

assumed In design for the devices u b (design

displacement of menshln device), the equivalent

stiffness K b and the equivalent damping ratio h

B of the menshln device are the key factors. They

are evaluated as:

(1) Design Displacement of Menshln Device

The design displacement of menshln device Is evaluated as

U 8 =

k h- Wu

Kb

k ho- Wu

Kb

(S. C. Method) (8)

(B. C. Method) (9) where

u B : Design displacement (m) of menshln

device

k h : Lateral force coefficient by Eq.(5) for

Seismic Coefficient Method

k ho: Lateral force coefficient by Eq.(6) for

Bearing Capacity Method

Kb : Equivalent stiffness (tf/m) of menshln

device

Wu : Weight of superstructure (tf) supported by the menshln device

(2) Equivalent Stiffness and Equivalent Damping

Ratio

The equivalent stiffness and the equivalent

damping ratio of menshln device are evaluated

from the hysteresis loops as

Kb =

h B =

F ( u be) - F (-U be)

2u

AW

2;r W

(10)

(11)

where

Kb : Equivalent stiffness (tf/m) of menshln

device

h B : Equivalent damping ratio of menshln

device

u BE : Effective design displacement (m) of

menshln device, and Is given as

U BE= C B • U B (12) u B : Design displacement (m) of menshln

device

c B : Modification coefficient for evaluating

the effective design displacement (=0.7) F ( u ) : Lateral force required to produce u for the menshln device

W : Strain energy Induced In menshln device

associated with c be displacement (refer

to Fig. 3) A W : Energy dissipated In menshln device per

176
cycle (refer to Fig. 3) '

5.2 Requirements for Dynamic Load

(1) Menshin devices have to be designed and

fabricated so that their equivalent stiffness K b and equivalent damping ratio he be within + 20 * of the design values. The equivalent stiffness

K B and equivalent damping ratio h e for this requirement have to be evaluated by Eqs. (10) and

(11), and they shall be the ones averaged overKB

and h a from 4th to 10th loading hysteresis

among 10 cyclic loading reversals with harmonic

displacement of c be. This Is because deck

acceleration and displacement are within

tolerable range when the scatter of equivalent

stiffness and the equivalent damping ratio are

within ± 20 % of the design values. However, the

devices which produce unstable hysteretlc

behavior should not be adopted.

(2) Menshin devices have to be stable against 50 cycles of harmonic loading with displacement of

u B. The number of 50 was determined from the

fact that number of load reversals developed

during a major earthquake may be about 30. About

50 % tolerance was included In 50.

(3) Deck should return to the rest position even

after It was subjected to a large earthquake.

Therefore the residual displacement developed In menshin devices after smoothly releasing from

the deformed displacement of u b by Eq. (9) has

to satisfy

U BR^O.l- U B (13)

where

u BR : Residual displacement (cm) u B : Design dlsplacement(cm) of menshin

device by Eq. (9)

(4) The equivalent stiffness and the equivalent

damping ratio of menshin devices have to be

stable against the change of load condition at normal time, change of natural environment such

as the temperature change and cyclic loading

developed by an earthquake. Stability has to be

examined against the following requirements:

1) cyclic loading due to elongation and

shrinkage of deck due to the temperature

change and the active load

2) effect of loading hysteresis

3) variation of vertical loading

4) effect of loading rate

5) effect of pre-deformatlon due to creep and

shrinkage

6) direction of excitation

7) change of the equivalent stiffness and the

equivalent damping ratio depending on the

temperature change

5.3 Requirements for Static Load

(1) Materials and mechanism of menshin devices

have to give credit to long term use. They have

to be stable against cyclic elongation and

shrinkage of deck due to temperature change.

(2) Menshin devices have to be stable against

local shear strain. Check of the local shear

strain has to be made In accordance of "3.6 Design

of Rubber Bearings" of the "Design Guidelines of

Bearings".

(3) For rubber-type menshin devices, the creep of

rubber In vertical direction due to dead weight

of superstructure shall not exceed 5 % of total

thickness of the rubber.

(4) The equivalent stiffness of menshin device at

-10°C normalized by the equivalent stiffness at 40

"C shall not exceed 1.5. Because the temperatures

of -10°C and 40°C Is for moderate climate area, at cold area they have to be decided based on the

appropriate site condition.

(5) Menshin devices have to have appropriate

Initial stiffness so that harmful displacement of

deck due to non-selsmic lateral force such as wind effects be avoided.

6. APPROVING TESTS OF MKNSniN DEVICES

6.

1

General

Because the menshin device Is one of the the

important structural components of menshin

bridges, their requirements have to be well

approved by the tests. There are various types

of menshin devices for use of bridges. Based on

the mechanism for producing the energy

dissipation, the devices may be classified Into three groups as

(a) displacement dependent type,

(b) friction force type, and

(c) velocity dependent (viscous) type.

The approving tests for the menshin devices

are presented In the Manual for each of the three

types.

6.2 Approving Tests for Dynamic Loads

For approving the requirements for dynamic

loads presented In 5.2, the following approving

tests are provided In the Manual.

(1) Approving test for confirming the equivalent

stiffness and the equivalent damping ratio

(2) Approving test for confirming stableness

against 50 cycles harmonic load reversals with u

B displacement

(3) Approving test for confirming residual

displacement requirement

(4) Approving tests for confirming stableness

against variation of load condition at normal

time, change of natural environment such as the

temperature change and cyclic loading developed

by an earthquake. These tests Include : (a) Stableness against Cyclic Load Reversals

Some menshin devices have the dependence of

the equivalent stiffness and the equivalent

damping ratio on number of cyclic loading. Because considerable change of these parameters

during an earthquake produces bridge response

significantly different from the one assumed In

177
design, It Is not desirable.

(b) Stableness against Load Hysteresis

In some menshln devices, the equivalent

stiffness and the equivalent damping ratio

depend on the past experience, In particular, the

largest deformation ever experienced. Such

dependence of K b and h b on the hysteresis ever experienced Is significant. It produces different

bridge response during an earthquake. Therefore,

effect of load hysteresis has to be examined.

(c) Stableness against Change of Compression

Force

Compression force applied to the menshln

device due to dead weight of superstructure may

change associated with the error Involved in construction stage and settlement of

substructure, if the menshln device shows

significantly different stiffness and damping

properties due to such change of compression

force. It brings different structural response . Therefore, stableness of the equivalent stiffness

and the equivalent damping ratio of menshln

devices have to be checked.

(d) Stableness against Change of Loading Rate Because loading rate to the menshln devices

Is not the same during excitation, dependence of

the equivalent stiffness and the equivalent

damping ratio on the loading rate has to be

examined.

(e) Stableness against Pre-deformatlon

Menshln devices often start to deform from

the displacement drifted to the rest point. This

Is actually developed if an earthquake occurs when the menshln devices deform due to the deck

elongation associated with temperature change

and shrinkage of concrete. Because the

equivalent stiffness and the equivalent damping

ratio depend on the amount of such

pre-deformatlon, their properties have to be

checked.

(f) Stableness against Temperature Change

Menshln device have to be stable against

dally and yearly change of temperature.

6.3 Approving Tests for Static Loads

For approving the requirements for static

loads presented In 5-3, the following tests are

required.

(1) Approving test for confirming durability and

stableness against cyclic loading associated with

the dally and yearly elongation and shrinkage of

deck.

(2) Test for evaluating the stiffness of menshln

devices subjected to extremely low-rate

deformation such as the one encountered due to yearly temperature change and concrete creep. The stiffness of the menshln device during such

low-rate deformation Is quite Important In static

design for computing the lateral force developed

In substructures due to the temperature change

and shrinkage of concrete.

(3) Approving test for confirming the dependence

of the equivalent stiffness of menshln device on

temperature.

6.4 Two examples of the Approving Test

Approving test methods presented In G.2 and

6.3 are precisely described In the Manual for

each of the displacement dependent type devices,

friction force type devices and the velocity

dependent type devices. As the example of such

approving tests, the test method for confirming

stableness against cyclic loading and the test method for confirming the effect of loading rate are briefly described for the displacement

dependent type devices In the following.

(1) Approving Test for Confirming Stableness

against Cyclic Loads

At the room temperature of 20 °C apply 10 cycles of harmonic lateral load to the menshln

device with loading displacement of u b computed

by Eq. (9) or smaller and with the frequency of

0.5 Hz. The vertical load equivalent to the design

dead load shall be simultaneously applied. The

equivalent stiffness and the equivalent damping

ratio from the test have to satisfy Eqs. (14) and

(15).

Rkc =

R hc =

I K BJ-K Bm I K Bm gO.3 (J=1.2.3)

gO.l (J=4,5- • 10)

h Bj-h Bml <0.3 (J=1.2.3)

h sm gO.l (J=4.5- • 10)

(14)

(15)

where, R Kc : Variation ratio of equivalent stiffness

R hc : Variation ratio of equivalent damping

ratio

K Bj : Equivalent stiffness at J-th load

reversal(tf/m) K : Averaged equivalent stiffness by Eq. (16) h BJ : Equivalent damping ratio at J-th load

reversal

h Bm : Averaged damping ratio by Eq. (17)

1

h Bm =

10Z K BJ J=l

10

J=l

(16)

(17)

Being different with Eqs. (10) and (11), the

equivalent stiffness and the equivalent damping

ratio for the approving tests are defined as

h B =

F ( u b)- F (- u b) 2u B AW

2n W

(18)

(19)

where

Kb : Equivalent stiffness (tf/m) of menshln

device

178
h B : Equivalent damping ratio of menshln

device

u s : Design displacement (m) of menshln

device

F(u): Lateral force (tf) required to produce u displacement for the

menshln device

W : Strain energy (tfro) Induced In menshln

device associated with u b displacement

(refer to Fig. 4) A W : energy dissipated In menshln device per

cycle (refer to Fig. 4)

(2) Approving Test for Confirming the Stableness

against Loading Rate

At the temperature of 20*C apply 10 cycles of

harmonic load with the displacement equivalent

to either 100 * of the total thicl<ness of rubber

or + 15 cm. Frequency of the load reversal shall

be 0.1 Hz, 0.5 Hz and 1.0 Hz. The equivalent

stiffness and the equivalent damping ratio have

to satisfy Eqs. (20) and (21).

^^^^ I Kb^.-Kbh.,!

g Q 2 (J=1.2) (20)

Vl Bml

^^^^ I hB^.-hB»>j|

(J=1.2) (21) n Bml

where,

Rkt : Variation ratio of equivalent stiffness

R ht : Variation ratio of equivalent damping

ratio

KBmi, Kbi..2, K Bm3 : Equivalent stiffness by

Eq. (16) at frequency of

0.5 Hz, 0.1 Hz and 1.0 Hz, respectively

h Bml . h Bm3, h Btns '. Equivalent damping ratio

by Eq. (17) at frequency of

0.5 Hz. 0.1 Hz and 1,0 Hz, respectively

7. CONCLUDING REMARKS

Excluding specially long-span bridges, it is not easily adopted for highway bridges to reduce

the lateral force by forcibly elongating the

natural period, because seismic force level is very high even at long natural period, and

because the Incretise of declt response

displacement brings various problems such as environmental pollution. Therefore the menshin

design, which places emphasis not on the

forcible elongation of the natural period but on

the Increase of energy dissipating capability and

distribution of inertia force of superstructure

to as many substructures as possible, is proposed. By the menshin design, it is proposed

to decrease the lateral force as large as 20 * for

the Seismic Coefficient Method and 30 % for the

Bearing Capacity flethod.

ACKNOWLEDGEMENTS

For executing the three-year Joint research

project, various supports and encouragement

were obtained from many organizations Including

the Head-quarters of the Ministry of

Construction. Without endeavour of more than 100 participants from the 28 private firms, this project could not be successfully made. The

authors express their sincere thanks for all who

participated in this Joint Research Program.

REFERENCES

1) Public Worlts Research Institute and 28 Private

Firms : Report of the Joint Research on

Menshln Systems for Highway Bridges, No. 1. July 1990, No. 2. July 1991 and No. 3. March

1992, Joint Research Report, No. 44, No. 60 and

No. 75, Public Worl<s Research Institute (In Japanese)

2) Public Works Research Institute and 28 Private

Firms : Manual for Menshln Design of Highway

Bridges (Draft), March 1992, Joint Research

Report, Public Works Research Institute (in

Japanese)

3) Japan Road Association : Part V Seismic Design

of the Specifications of the Design of Highway

Bridges, February 1990 (in Japanese)

4) Technology Research Center for National Land

Development : Guidelines for Base Isolation of

Highway Bridges, March 1989 (in Japanese)

5) Japan Road Association : Guide-Specifications

of Design of Bearings, July 1991 (In Japanese)

6) Buckle, I. G. and Mayes, R. L. : Seismic

Isolation : History, Application and

Performance - A World View, Earthquake

Spectra. Vol. 6, No. 2, May 1990

7) AASHTO : Guide Specifications for Seismic

Isolation Design. June 1991

8) McKay. G. R.. Chapman. H. E. and Klrkcaldle. D. K. : Seismic isolation : New Zealand

Application. Earthquake Spectra, Vol. 6, No. 2, May 1990

9) Guidelines for Seismic Design of Bridges with

Isolatlon/Dlssipator Devices, Antiseismic

Protection of Structures on the Autostrade

S.p.A/ Network. October 1991

10) Kawashlma. K. , Hasegawa, K. and Nagashima, H. : A Perspective of Menshln Design for Highway

Bridges in Japan. Proc. 1st U.S.-Japan

Workshop on Earthquake Protective Systems of

Highway Bridges. National Center for

Earthquake Engineering Research. Buffalo,

N.Y., U.S.A., September 1991

11) Kawashlma, K., Hasegawa, K. and Nagashima, H. : Experiment and Analysis on Seismic Response

of Menshln Bridges, Proc. 1st U.S.-Japan

Workshop on Earthquake Protective Systems of

Highway Bridges, National Center for

Earthquake Engineering Research, Buffalo,

N.Y.. U.S.A.. September 1991

179
I
Table 2 Dasplng Ratio Reconmended for Structural Components

Structural Components Steel Concrete

Superstructures 0.02^0.03 0.03~0.05

Menshln Device Damping Ratio by Eq.(ll)

Pier/Columns 0.03--0.05 0.05--0.1

Footing 0.1~0.3

Table 3 Modification Factor for Ground Condition c g

Ground Group I II III

Co 0.8 1.0 1.2

Table 4 Modification Factor for Importance c i

Group Definition

1st class 1.0

Bridges on expressway (limited access highways),

general national road and principal prefecturai road. Important bridges on genera] prefecturai road and municipal road.

2nd class 0.8 Other than the above

Table 5 Modification Factor for Structural Response c -r

Ground Group Structural Response Coefficient

Group I T < 0.1 c^ = 2.69T"'21.00

0.1 ^ T S 1.1 c^ = 1.25

1.1 < T

c^ = 1.33T-^

Group II T < 0.2

c^ = 2.15T"'^1.00

0.2 ^ 1.3

c^ = 1.25

1.3 < T

= 1.49T^

Group III T < 0.34

Cr = 1.80T"'>1.00

0.34 < 1.5 = 1.25

1.5 < T

c^ = 1.64T-^

181
Table 6 Modification Factor for Energy Dissipation Capability c

Damping Ratio h Modification Factor c e

h < 0.1 1.0

h ^ 0.1 0.9

Table 7 Modification factor for Structural Response c r

Ground Group Structural Response Coefficient c^

Group I Teq ^ 1-4

c, = 0.7

1.4 <

c, = 0.876T,^-^

Group II

< 0.18

c„ = 1.51 T^''>0.7

0.18 ^T^%\.6

c, = 0.85

1-6 <

c, = 1.16V"

Group III < 0-29

c, = 1.511,^-" >0.7

0.29 <Tgg< 2.0

c, = 1.0

2.0 < T,^

c, = 1.59V"

Table 8 Modification Factor for Energy Dissipating Capability

Damping Ratio h Modification Factor c e

h<0.1 1.0

0.1 g h< 0.12 0.9

0.12^ h<0.15 0.8

0.15^ h 0.7

182
Fig. 1 Menshln Devices Developed

TYPE OUTLINE OF MENSHIN DEVICE

HIGH DAMPING

RUBBER BEARING

Flange

Steel Plate

High Damping Rubber ^Surface Rubber

SLIDE FRICTION

-RUBBER BEARING Slide Friction

Damper

STEEL DAMPER

Laminated Steel Damper

Rubber -—

Bearing I Pier I OkcqO

THttT

ROLLER MENSHIN

BEARING

High Damping Rubber

1^

7777,I

Roller

VISCOUS

Rod

L

H

viscous Material

183
Fig. 2 Modification Factor for Zone c z

Lateral Force

A AW

Lateral

Displacement

Fig. 3 Definition of Equivalent Stiffness and Equivalent Danplng

Ratio of Nenshln Device for Design of Nenshln Device

184
Lateral Force

4 AW

Lateral

Displacement

Fig. 4 Definition of Equivalent Stiffness and Equivalent Damping

Ratio of Menshln Device for Approving Tests

Photo 1 Knock-off Abutment

185
Photo 2 Expansion Joint Movable In Longitudinal and

Transverse Directions

186
187

Friction Controllable Bearings for Sliding Base Isolation Systems

by

Maria Qing Feng and Masanobu Shinozuka*

ABSTRACT

Recently, sliding ba^e isolation systems began to be utilized widely in civil and axcliitectural engineering as exemplified by the recent decision on

the part of US Federal Government (General Ser- vices Administration) to deploy 250-300 spherical sliders to base-isolate the 100 year old, 5- story and 60,000 ton Court of Appeals Building

in San Francisco. While the building survived

the 1906 San Francisco Earthquake, it was damaged by the Loma Prieta earthquake. Sliding

base isolation systems, it is generally perceived,

have disadvantages in that they are not effective

for small to medium earthquakes, and may not

be able to maintain the same value of friction

coefficient under a prolonged inactivity and that

they may suffer from excessive displacements under strong earthquakes. This paper presents the

results of a recent experimental and analytical

study jointly performed by Princeton University

and Taisei Corporation (Refs. 1 and 2) demonstrating that the use of friciton controllable bearings can not only neutralize the disadvantages

indicated above but also make it possible to take

full advantage of the beneficial features of sliding

base isolation; its inherent effectiveness in limiting the transfer of the ground acceleration to the

structure.

KEY WORDS

Hybrid sliding isolation. Friction controllable

sliding bearings. Earthquakes, Instantaneous optimal control. Shaking table

"Dept. of Civ. Engrg., Princeton Univ., Princeton,

NJ 08544

1 INTRODUCTION

The purpose of this research is to physically develop a sliding bearing that can actively control

the friction force on the sliding interface and to demonstrate by shaking table experiments and

numerical simulation that the difficulties associated with sliding isolation systems can be significantly alleviated, if not totally eliminated, by

such a device. More specifically, the purpose is:

L to propose and physically develop a friction

controllable sliding bearing and a hybrid

sliding isolation system using such bearings,

2. to develop control algorithms for controlling

the friction force which has nonlinear characteristics, and

3. to analytically and experimentally demonstrate that such a hybrid isolation system

can indeed intelligently control the friction

force on the sliding interface so as to confine the sliding displacement in an acceptable range, to reduce the residual displacement, and at the same time, to minimize the

treinsfer of seismic force to the structure.

The hybrid isolation system using friction controllable bearings (FCB's) is conceptually depicted in Figs. 1 with a bridge structure resting on the bearings. Each bearing has a fluid chamber which is connected to a pressure control sys- tem composed of a servo valve, an accumulator

and a computer. The friction on the interface between the bearing and the ground is controlled

by adjusting the fluid pressure in the chamber.

The computer calculates an appropriate signal to control the fluid pressure based on the observed

189
structural response, such as response accelera-- tion and sliding displacement, and sends it to the pressure control device as shown in Figs. 1. The idealized section view of the friction controllable sliding bearing is given in Fig. 2. The

bearing made of steel is of disk shape containing a fluid chamber inside which is sealed by a

rubber 0-ring around the circular perimeter just

inside the sliding surface of the bearing. A slid- ing material such as PTFE plate is placed on the

sliding surfa<;e. The word "HYBRID" is used for this system

for the following reason. The system can be a

paasive sliding isolation system as long as the

pressure of the bearing chamber, and thus the

friction is kept at a constant value. At the same

time, the system can be an active system as long

as the pressure is controlled so that the friction

is controlled. In the analytical formulatin involving such a hybrid system, the control force does

not appear in the equation of motion as an in- dependent term like the force from an actuator. The active control in this case is implemented

through the friction term in the equation, and in

that sense the system may also be interpreted as "Semi-Active". From the view point of the en- ergy requirement to control the isolation device,

the system has a significant advantage since the

pressure control does not need as much energy

cis the common active structural control by ac- tuators would require.

2 ANALYTICAL MODEL

A rigid structure supported by the friction controllable sliding bearings is considered. The motion of the structure can be modeled by a singledegree-of- freedom (SDOF) model as shown in

Fig. 3. The equations of motion of the structure under earthquake excitation can then be written

as follows.

1. Sticking Phase — Phase I

X = 0, X = const.

2. Sliding Phase — Phase II x= -z - f sgn(x).

3. Criteria for transition from Phase I to Phase

II

\z\ > f (3)

4. Criteria for transition from Phase II to Phase I s = 0 (4)

|ii|<2/ (5)

where x is the sliding displacement of the mass

relative to the ground; z is the input earthquake

acceleration; fj. is the coefficient of friction on

the sliding interface; / is the normalized friction

force defined as f = ^g.

In the sticking phase, Eq. 1 governs the motion of the structure until the Eq. 3 becomes

true. As soon as the condition in Eq. 3 is met,

the sliding phase starts and Eq. 2 governs the

sliding motion. During the sliding phase, whenever X becomes zero, the criterion, Eq. 4, is checked to determine the subsequent behavior.

Validity of the inequality given by Eq. 4 is the

condition for entering the sticking phase. That

is, if the inequality holds, the structure will stick to the ground and Eq. 1 apply. If the criterion

given by Eq. 4 is not satisfied, Eq. 1 will con- tinue to govern the the subsequent sliding motion.

On the other hand, the normalized friction

force / on the sliding interface between the struc- ture and the ground is controlled by changing the

fluid pressure in the bearing chcimber through a

pressure control system consisting of a computer,

servo valve, amplifier, etc. The dynamic characteristics of the pressure control system are as- sumed to follow the first order time delay model:

Tp-\-p=u (6)

where, p is the pressure in the fluid chamber

of bearing; u is the pressure control signal from

computer; T is the time constant

The normalized friction force / is negatively

(1) proprtional to the pressure p in the bearing

chamber

/ = -cip + 02 (7)

(2) where Ci and C2 are constants.

190
3 INSTANTANEOUS OPTIMAL CONTROL

Control theory has not been well developed for the systems in which the control force has the

nonlinearity, unique to the friction controllable

sliding isolation device. For this reason, an optimal control algorithm for control of the nonlinear

friction system is developed on the basis of the

instantaneous optimal control theory originally

proposed by Yang et al. (Ref. 3).

The optimal pressure control signal u{t) is determined by minimizing the following time dependent objective function J{t) at every time in- stant t for the entire duration of an earthquake.

+x{t - At)

+ [x(0 - x{t - At)] — (9)

At m= f{t-At)^f{t-At) AtHfit)-f{t-At)] —

(10)

Furthermore,

x{t) = a fit) s^{x{t)) + hz{t)^d 1 [t-At) (11)

f{t)= -cu{t) + d2{t- At) (12)

where

Af2 ciAt

a = 0 = —, c =

6

' 2 T-h At

(13)

J(t) = gdi2(i) + gy/2(t) + ni2(i) (8) di{t-At) = x{t - At) x{t - At)At

in which the normalized friction force / equivalently represents the amount of response acceleration and also serves as a measure of the transfer

of seismic force to the structure. The weighting

coefficient qd and g/ are non-negative and r is positive. They indicate the relative importance

in the control objectives of the sliding displacement, response acceleration and pressure control

signal, respectively. The basic objectives of the

control is to make the structure slide as much as possible within an acceptable range ajid at the

Seime time to ensure the transfer of seismic force to a minimum.

With this theory, the following control algorithm is derived under the assumption that the

structural motion is always in the sliding phase.

The equation of motion given by Eq. 2 should

be used as a constraint when minimizing the objective function J(t). The first order time delay relationship between the control singal and

the pressure described in Eq. 6, as weU as the

linear relationship between the friction and the

pressure shown in Eq. 7, are also constraints. In the present formulation, however, these equcitions will be solved numerically using the Newmcirk's P method with /3 = 1/6 as shown below

eind these numerical solutions will be used as constraints:

x{t) = x{t - At) -I- x(t - At) At

+ -x(t-At)At2 (14)

ci2(t-At) =

2 T

2 r + At ' 2T

-|-i/(t - At)At)

.Co At . . {-^ + f{t-At)

(15)

Thus, the following generalized objective function is established by introducing the Lagrangian

multipliers Ai and A2:

H{t) = qdx^{t) + qf f{t) + Tu\t)

+Ai[x(t)-a/(t)sgn(i(t))

-b z{t) - di {t- At)]

+X2[f{i) + cu{t)-d2 (t- Atp)

The necessary conditions for minimizing the objective function J(t) are: dH ^ dH ^ dH ^ dH ^ dH ^ ^='^ 17='' 1^='^ wr'' aAi=''

(17)

Substituting Eq. 16 into Eq. 17 yields the optimal pressure control signal:

u{t) = Ff f{t) + Fd x{t) sgn(x(t)) (18)

where, the control feedback gains F f and Fd are calculated by

9/ Fd =

c a qd

(19)

191
and the dispicw;ement x{t) and friction f{t) are used for feedback purpose. Again, notice that

sgn(i) can be obtained from the displacement

signal without the need to measure the velociy.

The friction is difficult to measure by a sensor, but the signal from the acceleration sensor

{x{t) + z{t)) can be used instead of/. Therefore,

the control signal becomes:

u{t) = Ff \x{t) + z{t)\ + Fd x{t) sgn(i(0) (20)

In the development of the optimal control al- gorithm shown above, the time delay of control

device shown in Eq. 6 has been incorporated. In

this case, the control is referred to as "instanta- neous optimal control with time delay".

K the response of the control device is so fast that the time delay can be ignored, a similar procedure to the above can be shown to result in the

following control signal:

u{t) = F + Fd x{t) sgn(i(<)) (21)

where

F=^l^, f,= -^

(22)

In this case, only the sliding displacement x{t)

needs to be measured and fed back.

4 SHAKING TABLE TEST

4.1 Structure Model and Isolation Device

The structure model used for experiments is shown in Fig. 4. The model, representing a

rigid structure, consists of a steel frame and steel weights. The total weight of the model is 12 tonf. The model is supported equally by four friction

controllable sliding beaxings on the shaking ta- ble. The bearing developed for the experiments

is shown in Fig. 5. Figure 6 is a photograph of

the bearing being used in the experiments. The

bearing, with a brass sheet of 1 nmi thickness

attached to be used as sliding surface, slides on

a stainless steel plate fixed on steel I-bars bolted

down on the shaking table, as shown in Fig. 4. Furthermore, a rubber 0-ring of 5.7 mm in diameter acts as seal for the fluid in the fluid chamber.

The area of the sliding surface is 86.0 cm ^, and

the vertically projected area of the fluid chamber is 57.7 cm ^. No restoring force device is used

in order to study the effect of friction force only.

A servo valve is located at the center of the ex- perimental structure from which the pressurized

fluid is distributed to each sliding bearings as described in Fig. 4.

4.2 Passive and Hybrid Isolation

Typical experimental results axe obtained under

the El Centro (NS, 1940) ground acceleration. In the passive isolation results as shown in Fig.

7(a), the pressure in the bearing chamber is kept

at 30 kgf/cm ^, corresponding to the coefficient of

friction 5.3%. A large residual sliding displacement of 62 mm after the earthquake is observed.

Figure 7(b) shows the results of hybrid iso- lation experiments using instantaneous optimal

control with time delay, in which the feedback

gains are set to: Ff = 16.5 kgf/cm^ and Fd = -16.4 kgf/cm^ with the pressure control signal confined between Umax = 45 kgf/cm^ and

Umtn = 10 kgf/cm^.

The performance of the hybrid isolation sys- tem is compcired to that of the passive isola- tion system. Figure 8 shows the maximum re- sponse acceleration, maximum sliding displacement, and the residual displacement of the struc- tural model with passive or hybrid isolation sys- tem under linearly scaled El Centro records of

different peak ground acceleration. Hybrid iso- lation results shown in these figures are those

under instantcineous optimal control algorithm

without time delay. In the passive isolation, if a small friction coefficient, for example, 1.6%, is used, a high level of

isolation performance is expected since the re- sponse acceleration is reduced to a small level. In this case, however, the maximum displacement becomes excessive very rapidly as the input

earthqualce becomes more intense. On the other

hand, if a large friction coefficient, such as 10.2%,

192
is used, the sliding displacement can be coniined

within a relatively small range. However the iso- lation performance in this case is limited in the

sense that the acceleration can not be satisfactorily reduced particularly for small to medium

earthquakes.

The hybrid isolation system can alleviate these

problems associated with passive isolation sys- tem; the response acceleration can be considerably reduced even for small to medium earthquakes, while excessive sliding displacement can be prevented for large earthquakes. Another advantage of the hybrid system is clearly seen in

Fig. 8 where the residual sliding displacement can be maintained almost at zero level. In summary, the hybrid isolation system per- forms better than the passive system in the

sense that a reduction of response acceleration is achieved for small to medium earthquakes, and

at the same time, the excessive sliding displacement under large earthquakes is prevented. Furthermore, the residual displacement is reduced

to zero in most cases.

5 CONCLUSIONS

A systematic study on a hybrid sliding seismic

isolation system using friction controllable bearings has been presented in order to significantly

improve the base isolation characteristics of slid- ing systems. For this purpose, a hybrid sliding

isolation system using friction controllable bearings has been physically developed, and shaking table experiments were performed on a rigid

structural model equipped with such a hybrid

system. A computer code has been developed for real-time on-line control implementation. The

dynamic characteristics of the control system between bearing pressure and sliding friction have

been identified. The results of hybrid sliding iso- lation experiments were compared with those of

passive isolation.

The following conclusions are obtained:

1. Significant advantage of the proposed hybrid sliding isolation system has been

demonstrated: (1) for the small to medium

earthquakes, the coefficient of friction is kept at the minimum value to reduce the

transfer of the seismic force to the structure

to a minimum; (2) As the input earthquake

becomes more intense, the friction is controlled to confine the sliding displacement of

the structure to an acceptable range, while

at the same time to keep the transfer of seis- mic force as small as possible. Such intelli- gent features of the hybrid system can not

be achieved by the passive sliding isolation

system.

2. Control algorithms developed for control of

nonlinear friction force proved to be effec- tive in achieving the desired control perfor- mances. In addition, they are practical and

easy for real-time on-ling control operations.

3. The analytical model of dynamic characteristics between the bearing pressure control signal and the friction on the slid- ing interface has been identified. Computer simulation results excellently match

the experiments. This implies that the analytical model represents the actual system

very well, showing the possibility of utiliz- ing the model to perform analytical study on other types of real structures equipped

with the hybrid isolation system under dif- ferent earthquake conditions.

4. The hybrid sliding isolation system proved

to be quite robust, demonstrating the high

potential for the application of the system

to actual structures.

Items 3 and 4 above have not been discussed

in this paper. The reader is refered to Ref. 4 for the detail.

6 ACKNOWLEDGEMENT

This work was supported partially by the National Center for Earthquake Engineering Research under Grant Number NCEER 90-2204,

and partially by Taisei Corporation, Tokyo,

Japcin.

193
References

[1] Feng, Q., Pujii, S., Shinozuka, M., and

Fujita, T., "Hybrid isolation system us- ing friction-controllable sliding bearings,"

Eighth VPI&SU Symposium on Dynamics and Large Structures, Blacksburg, VA,

1991.

[2] Feng, Q., Sliinozuka, M., Fujii, S., and Fujita, T., "A Hybrid isolation system for bridges," Proceedings the First US-Japan

Workshop on Earthquake Protective Systems for Bridges, Buffalo, NY, 1991.

[3] Yang, J.N., Long, F. X., and Wong, D.,

"Optimal control of nonlinear structure,"

Journal of Applied Mechanics, Vol. 110, pp.

931-938, 1988.

[4] Feng, Q., Fujii, S., and Shinozuka, M.,

"Experimental and analytical study of a

hybrid isolation system using friction controllable sliding bearings" Technical Report

NCEER,, NCEER, SUNY, Buffalo, NY,

1992 (to be published).

194
Qi r de

r

Pier

Rest o ring

Force Device

Displacement Sensor

Pressure Sensor

Acc e 1 e r ome t er

Pressure

Control

Device

L.

I

Con t ro

1

] S i gna 1

Compu t e r

Figure 1: Concept of Hybrid Sliding Isolation System for Bridges

Fluid Pipe

Sliding Material

Fluid

Chamber Fluid •

"

' Pressure

Steel Plate ••

Figure 2: Idealized View of Friction Controllable Sliding Bearing

Z

Figure 3: Analytical Model

195
Figure 4: Structure Model with Hybrid Sliding Isolation Device

Figure 5: Friction Controllable Sliding Bearing

196
Figure 6: Friction Controllable Sliding Bearing for Experiment

5300- |o -

S-300--

(a) Passive

Input Acceleration

1 met.o- 274 8 Oil i300

<-300 r +

(b) Hybrid

Input Acceleration mai> 308.1 0*1

5 to IS

Response Acceleration

20

1-150-

Response Acceleration msi- 79.78 Gal 5^"" I m»». 925Qal

, , , , 5-150-1 , , , ,

|150-

1 0

-

S 10 t5

Sliding Displacement mu- 119.2 mm 1

150-1

1-150

Sliding Displacement

max- 64.65 mm

10

Pressure

50

0

I 50

IS

maxm 29.07 kQVan2

10

Pressure

20

ma>» 43.85 kgl/anZ

Control Signal

10 15

Control Signal mn. 29 S3 kqf/canZ ma>- 46.09 ygl/anZ

to Tlma (sec)

10 Time (SBC.)

Figure 7: Experimental Results

197
Comparison of Response Acceleration

150 oH

u

o

'« au

100

Z

o

m

w 50

in 7 %

/ — 1 B

/

/

% A

A

/l.6% i! / ^

100 200 300 400 ( Gal )

150

B

B !w 100

O H

I-] u

ca z

u z u

3 << Z .J

l-l 04 X w

•< 1-4 z o

50

Comparison of Sliding Displacement

o

/

/

/

/A

'5.3 %

/o

1.6 % ^/

A

/

/ i //

^T'^

^ 10.2 %

100 200 300 400 ( Gal )

Comparison of Residual Displacement

u

zu

u

enu

150

100

50

1 . .,

: HYBRID mu = 1.6 - 10.2 % •

PASSIVE mu = 1.6% O

mu = 5.3 % A

mu = 10.2 %

0 100 200 300

PEAK INPTJT ACCELERATION (EL CENTRO)

Figure 8: Compaxison of Passive and Hybrid Isolation

400 ( Gal )

198
State-of-the-Arts on Building Structures With Structural Response-Control

System in Japan

by

Yosikazu Kitagawa' and Mitsumasa Midorikawa^

SUMMARY

This paper presents a brief introduction

to active/hybrid structural response-control

building in Japan. First, classification of

seismic response-control system and

relationship between control theory and

required performance are roughly reviewed.

Finally, examples of structural responsecontrol buildings and model structures are given.

KEY WORDS : Classification of Seismic

Response-Control Systems, Active and Hybrid

Structural Response-Control Building, Control

Theory, Required Performance

1. INTRODUCTION

Japan is characteristically subjected to the seismic activity. Therefore, it is a matter of concern how building structures behave

under earthquake conditions. Currently the research focus for seismic design methods has

changed from static analysis to dynamic

analysis. Building structures are also

becoming to be taller in height and longer in span with increased demands in living comfort

and safety. In response to these changes, seismic

isolation and response-control systems for

building structures are becoming important

areas in the field of earthquake engineering.

The former system includes base and floor

isolations, and the latter includes

active/passive structural response-control

buildings under earthquake and wind

excitations.

Many studies and proposals involving

active/passive structural response-control

system have been conducted. A portion of

these projects has resulted in actual and model

structures in Japan, USA, and other countries.

Recently, structural response-control buildings

have been proposed and enthusiastically put

into practice in Japan.

In this paper, the structural responsecontrol systems are roughly classified, and the

present status of building structures with

response-control systems are outlined.

2. CLASSIFICATION AND

CHARACTERISTICS OF RESPONSECONTROL STRUCTURES

A response-control structure may

generally be classified into two groups: one

that requires an active response-control system

(i.e., some form of applied external energy is required in order to control the response of a building to external turbulence); and one that only needs a passive response-control system

(i.e., no such energy is required). Active

response-control systems can be classified as

either a full-active method in which some form

of response control force is applied to resist external turbulence or a semi-active method in which the dynamic characteristics of a

structure are controlled. Passive responsecontrol systems can be classified as either a

base-isolation method or a tuned-mass damper

method. Hybrid methods use a combination of

active and passive methods.

In an active response-control structure, the vibrations of a building are reduced

through an active method using the signals

sensed in and around the building. From the

standpoint of signal processing, an active

response-control system can be classified in two ways: a forward control in which the

necessary control is introduced using signals

from the outside sensors before external

turbulence can reach the building; and a feedback control in which control is introduced

after response of a building are sensed by

measurement instruments installed in the

building. A flow chart of an active structural response-control system is shown in Fig. 1. The resultant motion of a building,

when it is subjected to external dynamic forces

1) Dr. Engrg., Director of International

Institute of Seismology & Earthquake

Engineering, Building Research Institute,

Ministry of Construction.

2) Dr. Engrg., Head of Earthquake

Engineering Div. I, USEE, BRI, MOC.

199
such as earthquakes or winds, can be

represented by the following equation: nrx + c*x + k«x = f + p (1)

where,

m = mass of the vibrational system;

c = viscous damping coefficient;

k = stiffness of the vibrational system;

f = external force acting on the vibrational

system; X = response acceleration with respect to the ground;

X = response velocity with respect to the ground;

X = response displacement with respect to the ground; and

p = response control force. Each term in this equation is expressed by a matrix or vector. The object of a responsecontrol structure is to reduce the response

acceleration x, response velocity x, and

response displacement x. The reduction is done by controlling and adjusting the mass m,

the viscous damping coefficient c, the stiffness k, and/or external force f, or by applying a control force p. According to the basic principles of

dynamics briefly outlined above, any structural response-control method can be classified as follows: 1) the method based on control and

adjustment of mass, such as rigid mass

dampers and liquid mass dampers; 2) the

method based on control and adjustment of

damping, such as variable-damping

mechanisms and building-to-building

connecting mechanisms; 3) the method based on control and adjustment of restoring force

characteristics, such as variable-stiffness

mechanisms and flexible base mechanisms;

and 4) the method based on applying control

forces using equipment such as jet or injection

devices, or using reaction walls. Theoretically, the structural responsecontrol concepts differ from conventional

techniques for earthquake or wind resistant structures in only the method introducing

control forces. The difference is that the

restoring force characteristics or energy

absorption properties depend on structural

elements such as columns, girders, walls, and

braces in conventional systems, but on

mechanical equipment or mechanisms in response-control systems.

3. CONTROL THEORIES

At present, a feedback control system

using responses such as acceleration, velocity,

and displacement is mainly adopted in an

active response-control structure, whose target

is to suppress the vibration of the structure induced by external turbulence.

Using old control theory, control effect

is attained by avoiding resonance in a structure during external turbulence or by increasing the damping capacity of a structure. In modem

control theory, the quantity-of-state is used for the feedback control, even though question

remains concerning controllability and sensing

ability. If this point is solved, then adjusting

dynamic characteristics of a modem control

system can put into practice. Optimal control

theory is the one mainly adopted in modern

control methods.

Control theories may be classified into

the following four groups: I) the self- organizing/structuring control in which the

structure of a control system is determined by

the control system itself to achieve the target

of an entire structure system; 2) the adaptive

control in which the evaluation function and

the parameters to be optimized are determined

adaptively in a given structure; 3) the optimal

control in which the operational quantities are determined so as to optimize a given

evaluation function; and 4) the direct control in which the operational quantities are directly

determined to match the process variables with

target values. These classifications are made

based on two factors: the degree of complexity

or simplicity of a system; and the degree of

certainty of information. A mapping of these

classifications is shown in Fig. 2. In the design of a stmctural responsecontrol system, it is important to identify the dynamic characteristics of the controlled

structure. It is also necessary to include fail- safe mechanisms to improve the reliability of

an entire structure system. Therefore, it is desirable to develop a fail-safe system with

adaptive control in which the control

parameters can be adjusted in response to environmental effects.

In building structures, it is generally

difficult to construct a control system since

there are many unknown variables such as the

type of structure, material, and construction,

and the deterioration of the structural

performance with passage of time. In case of

controlling the vibration of a structure subjected to earthquakes, it is critical that the response of a control system be quick enough

to include time from sensing external

turbulence to putting actuators in operation.

There are, however, uncertainties in input

signals to a control system caused by many

factors such as the direction of input ground

200
motions and the location of sensors. For

dealing with these vague and uncertain

problems, further development and application

are needed on an intelligent control system that uses fuzzy logic or neural networks. Figure 3

shows an example of the relationship between

system requirements and control theories for structural safety evaluation of a responsecontrol structure. Figure 4 shows an example

of a control system that uses a fuzzy optimal

method. The features of this system are: 1) real time prediction of earthquake ground

motions, 2) real time structural identification,

and 3) maximizing decision using fuzzy logic.

This fuzzy optimal method is basically

different from fuzzy control method that uses

the proportional, integral, and differential

(PID) control.

4. PRESENT STATUS OF STRUCTURAL

RESPONSE-CONTROL BUILDINGS

Table 1 shows an outline of active and

passive structural response-control buildings in Japan. Many studies and proposals involving

active/passive structural response-control

systems have been conducted. A portion of

these has resulted in actual buildings or model

structures in Japan. The KS Building which is actual building structure with an active

structural control system was reviewed by the Board of Technical Members of A Special

Committee in the Building Center of Japan.

This committee inspected the technical aspect

of structural safety, vulnerability, and

maintenance of the KS Building and its structural control system.

4.1 KS Building (Ref. 6. Figs 5. 6. and 7)

This eleven-story, steel structure office

building was constructed in Chuo-ku, Tokyo,

in 1989. An active mass damper (AMD)

device with an auxiliary weight of about 1% of

the total weight of the building was installed on the top floor of the building. Because the building shape is very slender and the damping

factor of this structure is small, the aim of the AMD is to control the lateral and torsional

vibrations and to satisfy the serviceability

condition during earthquakes and strong

winds. To do this, the AMD system reduces

the maximum response quantities of the

uncontrolled structure to about one-half to two-third during earthquakes and strong winds.

The AMD system is also designed to function

within the device capacity if an excessively

large load is applied. After complenon of the building, both free and forced vibration tests

were carried out. Seismographs and an anemometer have also been installed for

carrying out observations.

4.2 K Laboratory Building (Ref. 7. Fies 8. 9. and 10)

A three-story, actual, steel structure for a large-scale shaking table facility was

constructed in Chofu, Tokyo, in 1990. To

control the vibration characteristics of the

building, an active variable-stiffness (AVS)

system was installed. A variable-stiffness

device was installed on each floor at both end

frames, at the joint of the braces and beam to change the connecting condition at the joint

using control computer. The resonance with

continuously arriving earthquake motions can

be avoided, and the building response can be

suppressed with a small power energy supply.

After completion of the building, the forced

vibration test was carried out. An earthquake

observation system was also installed in this building to confirm the actual control effect.

4.3 S Building (Ref. 8. Figs 11. 12. and 13)

An eleven-story, actual, reinforced

concrete structure office building was

constructed in Shibuya-ku, Tokyo, in 1991.

The control mechanism is an AMD system

used to control translational and torsional

motion of the building, and to satisfy

serviceability for living comfort during

earthquakes and strong winds. The AMD

device was installed on the top floor of the

building with an auxiliary weight of 1.1% of

the total building weight. The purpose of

installation is to reduce the maximum response

quantities of the uncontrolled structure to about 0.4 to 0.8 times. After building

construction was completed, an earthquake

observation system was installed.

4.4 MM Tower Structure (Ref. 9. Figs 14 .15. and 16)

A seventy-story, S/SRC tower

structure, for an office and hotel, is under

construction in Yokohama, Kanagawa

Prefecture, and will be completed in 1993.

The control mechanism is a tuned active mass damper (TAMD) system with a multi-stage

pendulum to suppress structural response to an

occasional strong wind force. The TAMD

device will be installed on the first floor of the penthouse (282.3 meters above the ground),

with an auxiliary weight of 0.13% of the total building weight. The goal of the TAMD

device is to reduce the maximum response

quantities of the uncontrolled structure to

201
about 0.4 times. This control system

incorporates the following safety functions:

switching to TMD by shutdown of motor

electric current in case of abnormality or power failure of the control system; applying a damper brake by switchover of the solenoid

valve; and use of malfunction checking system

such as turning on warning lamp.

4.5 Full-Scale Model (Ref. 10. Figs 17 .18 and

191

A six-story, full-scale, steel model

structure was constructed to check the

performance of an active structural responsecontrol system in Koto-ku, Tokyo, in 1989.

The control mechanism is an AMD system

used to reduce the acceleration and

displacement response of the uncontrolled

structure to one-half to one-third during

earthquakes and strong winds. The AMD

device is installed on the top floor of the model

structure, with an auxiliary weight of 1% of

the total model structure weight. A shaking

table test to verify the damping effect of the AMD device, the micro-tremor measurements,

and a forced vibration test were carried out. An earthquake observation system was also

installed in this full-scale model structure. This structural control system was applied to an actual eleven-story, reinforced concrete

structure. (See 4.3 S Building).

4.6 Full-Scale Model (Ref. 11. Figs 20. 21 and

22}

A seven-story, full-scale, steel structure was constructed in Koto-ku, Tokyo, in 1991 to check the performance of a hybrid structural control system. The control mechanism is a hybrid mass damper (HMD) system which

makes use of its capability to suppress

vibration as a tuned mass damper (TMD)

system with greater energy efficiency than an AMD system while maintaining the same

performance. The HMD device was installed on the seventh floor, and consists of an

auxiliary mass supported by multi-stage rubber

bearings and actuators driven by AC servo- motors. An auxiliary weight of the HMD

system is about 1.2% of the total model

structure weight. A shaking table test of the

HMD device and a forced vibration test of a full-scale model were conducted to evaluate

the dynamic characteristics of the device and

the full-scale model. Earthquake and wind

observations have also been carried out. In addition, a large-scale genuine HMD system

will be installed in an actual fifty-story, steel

building (ORC 200 symbol tower for use as

office and hotel) at Benten-cho in Osaka.

4.7 Full-Scale Model (Ref. 12. Figs 23. 24 and

25)

A six-story, full-scale, steel model

structure was constructed in Tukuba, Ibaraki

Prefecture, in 1991 to determine the

performance of a structural response-control

system. The control mechanism is an activepassive mass damper (APMD) system which

exhibits great structural response-control in an

active mode during medium and small

earthquakes as well as strong winds. When

external forces exceed the capacity of the

system under strong earthquakes, the APMD

system switches over continuously from active

to passive mode as a tuned-mass damper

system. The APMD device is installed on the

top of the structure, with an auxiliary weight of

about 1% of the total model structure weight.

The control effect is to reduce the maximum

response quantities of the uncontrolled

structure to about one-half. A forced vibration

test was carried out. Earthquake and wind

observations have also been carried out.

4.8 Full-Scale Model (Ref. 13. Figs 26. 27 and

28)

A six-story, full-scale, steel model

structure was constructed in Tukuba, Ibaraki

Prefecture, in 1991 to confirm a structural control system. The control mechanism is a tuned roller-pendulum damper (TRD) system.

A free vibration test and forced vibration test were carried out. Earthquake and wind

observations have also been carried out. A

variable resistance damper (VRD) system as a

device for a semi-active structural responsecontrol system was developed, and its effect was confirmed by the shaking table tests. This

VRD will be installed inside the earthquakeresistant walls of high-rise buildings or between the base mat and the first floor of

seismic isolation building structures with no

need for a huge energy supply or backup

systems.

4.9 Reduced-Scale Model (Ref. 14. Figs 29. 30

and 31)

An eight-story, reduced-scale, steel model structure was constructed to examine

both experimentally and analytically the

performance of structural control taking into

consideration spillover, robust control, and

multi-mode control for high-rise building

structures. The control mechanism is an AMD

system used to reduce acceleration and to

202
control vibrations through a wide frequency

range that includes higher-order vibration

modes of the structure. The AMD device has a ball-screw drive system equipped with an AC

servo-motor, and is installed on the top floor of

the structure with an auxiliary weight of 0.4%

of the total weight of the model structure. The

purpose of installation is to reduce the absolute

response of the uncontrolled structure to 0.4 to 0.8 times. A plan has been made to install this system in an office and residential building

scheduled for construction, in Tokyo, in the near future.

4.10 Reduced-Scale Model (Ref. 15. Figs 32.

33 and 34)

A four-story, steel, reduced-scale

model structure was built to confirm the

capacity of AMD and ATMD system. These

systems involve an auxiliary mass floating on

a sliding base, variable friction bearing (VFB).

The auxiliary weight of AMD and ATMD

systems is about 1% of the total model

structure weight. The target of the control

effect is to reduce the maximum acceleration

and displacement of the uncontrolled structure

to about 0.6 times. A shaking table test was

carried out. An active control system using

gyrostabilizer and a hybrid base isolation

system that uses a friction-controlled sliding

bearing have been developed, and its effect was confirmed by free vibration and shaking

table tests.

4.11 Reduced-Scale Model (Ref. 16. Figs 35.

36 and 37)

A four-story, steel, reduced-scale

model was constructed to confirm the effect of

structural control. The control mechanism

suppresses the vibration of the structure as a hybrid structural control system. An active

rolling pendulum (ARP) device is designed to work with a much smaller added mass and to produce a more effective resistive force. This

ARP system is installed on the top floor of the model structure. The system works as a passive system, which is made more reliable

than an active system by switching the

electromagnetic valve when external forces

exceed the capacity of the system, and the

power supply is disconnected. A shaldng table

test was carried out.

4.12 Reduced-Scale Model (Ref. 17. Figs 38.

39 and 40)

A ten-story, reduced-scale, steel model

structure was built to verify the control effect,

the robustness, and the performance of a

hybrid structural control system. The control

mechanism is a hybrid tuned multi-mass

damper (HTMMD) system. This TMMD

system consists of three floors unified with

braces, and a control device which is composed of levers with an auxiliary mass and

actuator driven by an AC servo-motor through

a ball screw installed on the ninth floor of the

structure. The control effect reduces the

maximum response quantities of the

uncontrolled structure to about 0.5 to 0.8 times

with high robustness, even if the passive

TMMD system is not well-tuned. Shaking

table tests have been carried out.

5. CONCLUDING REMARKS

This paper presents a brief introduction

to active/hybrid structural response-control

building in Japan. Response-control devices

have been installed in actual building

structures, and have been developed, mainly

by construction companies, and realized to provide human comfort against excitation such

as small and medium earthquakes and strong

winds.

It is no exaggeration to say that the

performance of structural response-control

buildings depends on the control devices

which have been developed mainly from a

practical use view-point. When we reach a consensus on the structural safety of controlled

buildings in view of structural reliability, new

structural design systems may be developed.

When that occurs, it is important for us to ensure the reliability and to ascertain the

vulnerability of a structural response-control

system concerning the structural safety as a

total system: building structural system,

machinery system, and environmental system.

6. ACKNOWLEDGMENTS

The authors express their sincere

gratitude to the following people; Mr. M.

Sakamoto, Kajima Co.; Mr. S. Aizawa,

Takenaka Technical Research Laboratory.; Mr.

H. Abiru, Mitsubishi Heavy Industries, Ltd.; Mr. K. Shiba, Shimizu Co.; Mr. O. Chiba,

Toda Co.; Mr. I. Abe, Okumura Co.,; Mr. A. Teramura, Obayashi Co.; Mr. S. Kawamura,

Taisei Co.; Mr. N. Ogino, Kumagai Gumi Co.,

Ltd. Appreciation is also expressed to Prof. S. Ishimaru and Mr. T. Niiya, Nihon University

for offering the data concerned with Section 4

of this paper.

203
REFERENCES

1) Y. Kitagawa, M. Hirosawa, "Base-Isolated

Building Structure in Japan", 20th UJNR,

1988.

2) Y. Kitagawa, and M. Midorikawa, "BaseIsolated Building and Seismic Safety

Evaluation in Japan", 23rd UJNR, 1991.

3) Y. Kitagawa, M. Midorikawa, and et al., "Seismic Response Control System by Fuzzy

Optimal Logic (Part 1-4)", AIJ Annual

Meeting, 1992, (in Japanese).

4) "Report on R &D to Achieve Active

Seismic Control of Building Structure",

Cooperative Research with Public

Organization and Private Companies, BRI,

MOC, 1992.

5) "Seismic Isolanon and Response Control for Nuclear and Non-Nuclear Structure", Special

Issue, SMiRT 11, 1991.

6) T. Kobori, M. Sakamoto, and et al., "Study on Active Mass Driver (AMD) System (Part

1)", 4th World Congress of Council on Tall

Buildings and Urban Habitat, 1990.

7) T. Kobori, M. Takahasi, and et al., "Shaking

Table Experiment and Practical Application of

Active Variable Stiffness (AVS) System", 2nd

Conference on Tall Buildings in Seismic

Regions, 1991.

8) M. Higashino, S. Aizawa, and et al., "The

Study of Translational-Torsional Control by

the Active Mass Damper of Practical Use",

Japan National Symposium on Active

Structural Response Control, 1992.

9) S. Yamazaki, N. Nagata, and H. Abiru,

"Tuned Acrive Dampers installed in the Minato Mirai (MM) 21 Landmark Tower in Yokohama", 8th International Conference on Wind Engineering, 1991.

10) S. Aizawa, Y. Hayamizu, and et al., "Experimental Study of Dual Axis Active

Mass Damper", U.S. National Workshop on

Structural Control Research, 1990.

1 1) K. Maebashi, K. Shiba, and et al., "Hybrid

Mass Damper System for Response Control of

Building", lOWCEE, 1992.

12) T. Fijita, S. Takahashi, and et al., "Mass

Damper with Convertible Active and Passive

Modes for Response Control of Building",

lOWCEE, 1992.

13) S. Otsuka, Y. Oka, and et al., "Vibration

Control of Tall Buildings by Roller

Pendulum", AIJ Annual Meeting, 1991, (in

Japanese).

14) T. Suzuki, M. Kageyama, and et al., "Acrive Vibrarion Control for High-Rise

Building using Dynamic Vibrarion Absorber

Driven by Servo Motor", U.S. National

Workshop on Structural Control Research,

1990.

15) 1. Nagashima, B. Bhartza, and et al., "Experimental Study on Active Mass Damper

System", Japan National Symposium on

Acrive Structural Response Control, 1992.

S. Sakamoto, M. Yamada, and et al., "Test on

Active Vibration Control System using

Gyrostabilizer", Japan Narional Symposium on

Active Structural Response Control, 1992.

S. Kawamura, M. Shinozuka, and et al., "Hybrid Isolation System using Friction

Controllable Sliding Bearings", US-Italy-Japan

Workshop/Seminar on Intelligent Systems,

1991.

16) K. Hasegawa, Y. Katano, and et al., "Basic

Study on Optimal Control Algorithm for

Active Vibrarion Control System", 2nd

International Conference on Adaptive

Structures, 1991.

17) S. Ishimaru, T. Niiya, and et al., "Experimental Study on Hybrid Tuned MultiMasses Damper (HTMMD) Structure Controlled by Model-Following-Algorithm",

Japan National Symposium on Acrive

Structural Response Control, 1992.

204
Table 1 List of Seismic Response Control Building Structure (Active/ Passive Control Kethod)"

Building Name Date Coopletion

Building Structure Story

Building

Uses

Location Designer/ ConstructionType of Seisoic

(control Device

Chiba Port Toter 1986 S (1250) Observatory Chiba-shi. Chiba

Nikken Sekkei/

Takenaka

Tuned Vass Daoper (TMD)

Yokohaoa larine Toter 1987 S 30F Observatory Vokohama-shi. Kanagaia

Shiuizu/

Shioizu

Tuned Liquid Daii)per(TLD)

Sonic City 1988 SSRC

3IF4B Office Oniiya-shi. Sai tana

Nikkenn Sekkei/

Fujita

Friction Daoper

Cold To»er 1988 S (I36o) Observatory Utazu-cyo. Kagata

Mitut/

Uitui

Tuned Liquid Daoper(TLD)

Rigashiyaoa Sky Toier 1988 s 7F

(I34ni)

Observatory Nagoya-shi.

Aichi

Nagoya City 4

Nihon Sogo Arch. / Kajinja. Taisei. Tokuso

Tuned Mass Daoper (TMD)

Asahi Beer Toier 1989 s 22F2B Office Taito-ku. Tokyo

Nikken Sekkei/

Kumagai

Friction Daoper

Pukuoka Observatory 1989 s (H8ffl) Observatory Fukuoka-shi. Fukuoka

Nikken Sekkei/

Taisei. Takenaka,

Kajina. Shinizu. Obayashi

Tuned Mass Daoper (TMD)

Kajina Corp. KI-Building 1989 SRC 5F(A)

9F(B)

IB

Office Minato-ku. Tokyo

Kajina/ Kajioa

Joint Daiiiper(Bell Daoper)

Kyobashi Seiia Building

(KS Building)

1989 SSRC UFIB Office Chuo-ku. Tokyo

1 S 1 Architects

8 Associates

Kajioa/ Kajioa

Active Mass Driver(AMD)

Sky To»er 1989 S (135o) Observatory Nagasaki-shi. Nagasaki

Mitsubishi Heavy

Industry/ Mitsubishi Heavy

Industry

Tuned Mass Damper (TMD)

Crystal Tower 1990 S

SRC

RC

37F2B Office Osaka-shi. Osaka

Takenaka/ Takenaka

Tuned Mass Daoper (TMD)

Control Bldg. of Shaking Table Facility in Kajioa

Ins. of Construction Technology(K Lab Building)

1990 S 3F Office Cyofushi.

Tokyo

Kajmai/ Kajioa

Active Variable Stiffness

(AYS)

Fujita Corp. Bead Office

Bldg.

1990 SSRC

20F4B Office Shibuya-ku. Tokyo

Fujita/ Fujita

Lead Damper

Oujiseishi Beadquarters

Bldg.

1991 S 15F4B Office Chuo-ku. Tokyo

Ouji-Kajima JV Design Rooo/

Kajima4 Takenaka

Honeycofflb Daaper

Shidizu Bead Office Bldg. 1991 SSRC24F2B Office Minato-ku. Tokyo

Shioizu/ ShioizuViscous Daoper

Bibikiryokuchi Sky To*er 1991 S (135o) Observatory Kitakyushu-shi. Fukuoka

Mitsubishi Heavy

Industry/ Mitsubishi Heavy

Industry

Tuned Mass Damper (TMD)

Sendagaya INTES

(S Building)

1991 S IIFIB

(45o)

IB

Office Shibuya-ku. Tokyo

Takenaka/ Takenaka

Active Mass Driver(AMD)

MM2t Land Hark Toier

(MM Toter Building)

1993 SSRC70F

(296o)

Botel a Office

Yokohama- shi. KanagataMitsubishi Estate

Co. /constructor JV

Tuned Active Daaper

F: Floor. B: Basement. R: Height(o). — : Active Seisoic Control Device

205
Excitation

JL

Sensor

Building Structure

Control Force

Actuator Sensor

Controller

Feedforiard System

Feedback Systea

Fig. I Flofchert of An Active Structural Control Systen

Conplex

Adaptive Control

Al Control

Fuzzy Control

Certainty

Optimal Control Optimizing Control

PID Control Sequential Control

Uncertainty of Inforaation

Simple Object

Fig. 2 Mapping of Control Theory Classification

<Paraieters> <Sy»le« «eqolr»ent8> <Adaptable Control Theory)

Hon-decurrence

Non-tlnearlty

Non-St«Hon«ry

Fail-Ssfe lechaniva

Tarlation of Stracture

Characteristics

Effect between Building

and Subsoil

llde tante Excitation

Qnlck Response

ili|h Robustness

Self-Adsptabillty

High Suppleaetsry

* Real-Tiae Response

Adaptive Fuzzy.

Al. Robust Control

Adaptive Fuzzy.

Al. Robust Control

Fuzzy. Al, Robust Control

Feed forvard. Sequential. PID Control

FIf. 3 Relationship between Control Theory and leqniraul

Sensor X I (Structural Response) ZitEarthquale Input)

Coofort. Safety Econoay L

Fuzzy Rule A Fuzzy Rule B Fuzzy Rule C

I Predicted Excitation |

| Target Response ~~|

| Target Control Variable"] X,.,

[structural Identification \-~ Predictive Respottse

' -T- ' f(C,) I Equation

i I X..,'t(C ,.,)-Z,.,

-

I laxlaizing Decialon |

I Optiaal Control Variable|

I C*..i

(Actuator) Figure 4 Flo* Chart of A Fuzzy Optieal Control Systea

206
Fig. 5 Entire View of Building Fig. 6 AMD Device

Realization of Active Mass

Driver(AMD) System in 1989 (Kyobashi Seiwa Building in Tokyo)

Aiieiic3e;?r

CoBtral Cojpoter

Obser/atioi Systa

0.4 — COiT'CL a aim

Jesionss Oisplacanent at Iltn ."loor

10 1

3espon!9 iorticnal .\ngl5 it ilth ?lcor

ma. 3.1 Ual

. ... I.. ». 40. (s.c)

Hecorded Groiinii Acceleration

Epicenter : lOlo iiortli-east of izu Ctishiaa ( Cepth 25lu)

Date : Oct, M. 1989

ilagnitude : H 5. J

Fig. 7 AMD Systen and Earthquake Observation

207
208
Tig. 15 TAD Divice

Tig. 14 Zntire View of Buildias Model

:Sn>

FiS. 16 TAD System «nd Coapsrisoii of Response

Tig, n Kntire Viev of Actual Model

209
6F DISP (CONTROLLED) ANA, MAX = - 0 27 — OBS MAX = 0.30

CONTROLLED OBJECT + (X)t^ (XI

(b)

[A] =!

(fl

IX) = [A] (XI U- (f)^ (XI

(XI

(XI : STATE VECTOR

(fl : FEEDBACK COEFFICIENT VECTOR

6F ACC (UNCONTROLLED) ANA. MAX = 35 26

6F ACC (CONTROLLED) OBS. MAX --35.58

I—

• OBS. MAX = 13 39

X — direction

Ft;. 19 AMD Systen and Conparison of Response

210
FiS. 23 Entire View of Actual Model

,Y direct ion

Active-Passive lifass Damper

Fi?. 24 Active Passive Mass Damper

.1 m

30 osai m. 25 2l(lBl

riartfntiEihe Acceleration

J^n-Conlrolledfralculalioo) IttX 1. 96cffl

Roof Floor PisplacmEnt

Fis. 25 APMD System and Comparison of Response

FiS. 26 Entire View of Actual Model

Specimen ibDOF model)

Fig. 27 TMD, VRD Divices

211
^ Sensor jLf_

intfollcr

• ithoiit TBD

5al nax'IS 5 gal

30 160 ?40 320 400

TI«E ISECI

•ith TRD

gal aax«l 1 . 4 gal 20,

Varloblo RMtstanss Da«r«r

3

I ^

^1 0 80 160 2A0 320 400

TIME I SEC)

Fig. 28 Application of Seil-Active Vibration Control Systen

and Acceleration Response of THD Systen

AC Servo Molor DVA Mm,

No Coflool 200.

— Subopbofnit ConDol « -:00f

» i .),V|',K^'*'^A***'

u ,„ Robasi Control with fllref < 300,

0 0 — '*v;'/— >—.V>---'—

-:ooi

- _ No Connol MAX- 1 62cm

Sobopiloma] Control NfAX«0 738cm

Robult Control ^(h f\ltet MAX-0,679cm

I 500-

! 1 0.0 ^.'W,^-

Fig. 31 AHD Systei and CoDparisoo of Response

212
Tig. 32 Entire View of Reduced Model

Fi?. 34 AMD Systei and AMD Device using Gyroslabil izer

Hg. 35 Entire View of Reduced Model

213
214
Status of NSF Structural Control Research Programs

by

S. C. Uu\ H. J. Lagorio', and K. P. Chong>

ABSTRACT

This paper reports the status

of a major research initiative

in Structural Control launched

in 1991 by NSF. Proposals

submitted during the first

year (in Fiscal Year 1992) of

a 5-year program are analyzed.

The proposal review process is

described. Technical

implications of intelligent

control systems and smart

materials as applied to civil

infrastructure systems are

discussed.

KEYWORDS: Structural control,

infrastructure systems, smart

materials, intelligent

structures.

1. INTRODUCTION

Starting from 1989, an

intensive effort has been made

to develop a long-term

national research program in

structural control. An

important consideration of

such a program is to foster,

on a coordinated basis,

multidisciplinary research and

development of passive, active

and hybrid control technology

for application to structures

and nonstructural building

systems under dynamic loads.

Collaborative efforts among

researchers in

' Program Directors,

Engineering Directorate,

National Science Foundation,

Washington, DC 20550, USA

interdisciplinary areas

involving academics,

industrial experts and

professional practitioners are

considered extremely

essential.

The important technical events

taken place in this program

development effort are:

* Establishment of the U.S.

Panel on Structural Control

Research, 1989.

* International Workshop on

Intelligent Structures,

Taipei, Taiwan, July 1990.

* NSF-EPRI Joint Workshop on

Intelligent Control Systems,

October 1990

.

* U.S. National Workshop on

Structural Control Research,

University of Southern

California, October 1990.

* NSF Workshop on Sensors and

Signal Processing for

Structural Control , February

1991.

PROGRAM INITIATION

Following these events, a firm

research program base is

established. In the summer of

1991, NSF officially announced

the "NSF Research Initiative

on Structural Control for

Safety, Performance, and

215
Hazard Mitigation" - a

collaborative effort with the

Strategic Highway Research

Program (SHRP) of the National

Research Council (NRC)

.

The fundamental objectives of

the program initiative are:

encourage innovations in

passive, active, and

hybrid control systems.

• advance sensor and actuator

technology and signal

processing techniques and

their applications.

• develop intelligent robots,

other devices and computer

systems for structural

safety.

• develop innovative systems

for energy absorption, added

damping, and variable

stiffness

.

• study the robustness and

system reliability of

control systems.

• develop guidelines,

standards and practical

approaches and techniques

for design, fabrication and

field installation.

• investigate emerging

technology subjects in

innovative structural

systems, and smart

materials, sensors, and

devices for detection,

sensing, monitoring, and

diagnosis

.

In line with the above

objectives, the following key

research problems are

identified:

• New materials.

• Performance evaluation,

analytical and experimental

verification.

• Standardized test procedures

and performance criteria.

• Integrated structural and

architectural systems design

techniques and standards.

• Control algorithm

development, robustness, and

reliability, adaptivity, and

optimization.

• Sensors, actuators and

monitoring system

development and optimal

deployment

.

• Advanced computer technology

incorporating artificial

intelligence, neural

network, fuzzy logic, etc.

for on-line targeting,

detection, monitoring,

diagnosis, and control.

• Intelligent robots and

devices.

• Modelling, signal

processing, and computer

simulations

.

• Practical methods and

techniques for design,

manufacturing, fabrication,

and implementation.

• Laboratory and field

experiments of systems and

devices under actual or

simulated dynamic

excitations.

• Design and demonstration

projects of control systems

as applied to new or

216
existing structures.

• Innovations in base

isolation, energy

dissipation devices, and

other passive techniques.

To carry out the above

described research, - collaboration among

researchers, practicing

engineers, and manufacturers

is encouraged. It is expected

that the research will

generate new knowledge,

techniques, and a better

understanding of the working

principles as well as

practical limitations inherent

in various control systems so

that conditions of

applicability and design

standards and limits of each

class of systems can be

established and evaluated.

The ultimate goal of the

research program is to bring

about technological advances

through system integration and

development innovation so that

control technology can be

realistically implemented in a

timely fashion to provide

safety resistance to protect

structures such as buildings,

bridges, tunnels, and other

critical lifeline systems

against earthquakes, winds,

and other natural or manmade

hazard environments, in

addition to improving the

quality of life.

PROPOSAL EVALUATION

Upon announcing the

above-mentioned research

initiative, sixty seven (67)

proposals were received by the

specific proposal submission

deadline, i.e., October 30,

1990. Since these proposals

cover a wide spectrum of

diverse technical subjects,

special considerations were

given in the review process of

such proposals - following the

traditional NSF peer review

system - to ensure all

proposals are evaluated by

their technical/scientific

merits, relevant factors such

as educational and practical

values, collaborations between

academic and industrial

sectors, interfaces among

cross-disciplinary subjects,

etc

.

As a result, a 2 -stage review

process was adopted by which

(1) Each and every proposal

was sent out by mail to

peers for disciplinary

reviews

,

(2) A cross-disciplinary

review panel consisting

of experts in such areas

as control systems,

earthquake and wind

engineering, sensors and

actuator technology,

materials and

transportation systems

was convened to evaluate

and rank these proposals.

The panel, after a 1 1/2 day

deliberation ranked the

proposal into three groups:

(1) Highly Recommended,

(2) Fundable if Funds are

Available and (3) Not

Fundable.

Based on the Panel's

recommendations, NSF will

select by Summer 1992

approximately 7 to 10

proposals for support.

217
PROPOSAL ANALYSIS

Among proposals submitted, 4 0

deals with control of

earthquake responses of

structures in various modes,

vs. passive, active, or

hybrid. Twenty four (24)

proposals study the wind

response control and 15

proposals address problems in

control of general structural

responses and vibrations

including one dealing with

waves in periodic structures

having imperfection and

another dealing with random,

chaotic motions by active

techniques. Among the active

control techniques, one

proposal investigates the

application of active

structural members for

earthquake and wind

resistance

.

Several proposals (Table 2)

involve development and

application of fiber optical

sensors in structural control4 against earthquake loads,

one against winds, and 3

others for general structural

motion control. Among

proposals which contain the

study of actuators and

actuation, 6 address

earthquake problems, 5 wind

problems, and 3 general

vibration problems. Types of

actuators proposed in these

studies include active mass

dampers (or drivers) , active

bracing systems, piezoelectric

and electromagnetic actuators,

regeneratic actuators, screw

jackets, shock absorbers,

controlled friction, viscous,

or other types of dampers,

connectors, or energy

dissipators. In terms of

processors (i.e., control

algorithms, computer software

and hardware systems) , 3

proposals focus on optimal,

robust control against

earthquakes, 1 against winds,

adaptive control techniques, 6

proposals address earthquake

problems, 3 wind problems, and

2 general motion control.

Several other proposals employ

or involve neural networks

techniques in control

algorithms and models, 3 for

earthquakes and one each for

winds and general motion

control. Proposals involve

material innovations and

applications (i.e., highperformance or smart

materials) are relatively few.

Among them are investigations

in using shape memory alloys

to provide variable damping

and actuation, and selfhealing materials for

repairing of concrete crack.

Among the proposals described

in Tables 1 and 2, there are a

few dealing specifically with

the systems aspects of the

control systems themselves or

the structural systems which

are under control. These

aspects include performance

evaluation/ requirements/

standards, control systems

which are designed for

monitoring, detection, and

maintenance purposes, and

methods for system

identification, modelling, and

reliability issues involved

(Table 3)

.

These proposals can also be

looked at from the application

points of view (Table 4)

.

Almost 50% (30 proposals) of

the total proposals submitted

are attempted to applied to

earthquake protection of

218
various types of buildings

structures or their

components. Seventeen (17)

proposals involve windresistant applications and 14

others for other types of

motion or response control.

In addition, there are a few

proposals for applications to

non-building types of

structures, i.e., bridges (and

their components) , lifeline

systems, industry facilities,

and other civil

infrastructures. Among these

one proposes to investigate

control of marine vehicles

from impacting to bridge

piers, another proposes to use

passive approach to control

the earthquake damage to wharf

structures. It is also

interesting to note that while

most proposals are formulated

for engineering analysis/

design/construction

applications in general terms,

a few are prepared for more

specific purposes, for

example, to retrofit a

structure under various

environmental load conditions,

and to retrofit bridges using

a vibration isolator

controlled in real time.

ENGINEERING RESEARCH IN

INTELLIGENT CONTROL SYSTEMS

AND SMART MATERIALS

The above analysis indicates

an enthusiastic response by

the U.S. research community

for research in developing and

applying intelligent control

systems (ICS) to civil

engineering structures. It is

clear that considerable

advance can be made to improve

structural safety through

research in structural,

earthquake, and wind

engineering through use of

control technology. The

possibility of applying

control systems to mitigate

the structural damage against

earthquakes, winds, and other

loads open up a wide range of

challenging research

opportunities for

multidisciplinary

investigators in the field.

In this connection, it is

interesting to also note some

current engineering research

in smart materials under way.

Smart materials is a class of

materials which have their own

sensors (nervous system)

,

actuators (muscular systems)

and control (brain system)

,

mimicking biological systems.

Current research activities

aim at understanding,

developing and synthesizing

material systems which behave

like biological systems. The

U.S., Japan , and

industrialized Western

European countries are leading

in the fundamental research on

smart materials and smart

systems. To this point, no

smart materials, except some

that have been biologically

synthesized, have been

successfully produced beyond

laboratories. The Japanese

remains at a stage of concept

formulation. The U.S. efforts

have been focused on

development of smart

biomaterials and smart systems

for control of space

structures and mechanical

vibrations. Sensors include:

optical fibers, strain gages,

corrosion sensors (e.g., one

developed at Lehigh

Engineering Research Center)

and other environmental

sensors, sensing particles

219
(e.g., magnetic particles).

Examples of actuators are:

shape memory alloys (which

would return to their original

shape) , hydraulic systems,

piezoelectric devices (which

move in an electric field)

.

Microchips, computer software

and hardware systems

constitute the control

aspects

.

Smart materials may even heal

themselves when cracked, for

example, self-healing concrete

being investigated in

universities of Illinois and

Michigan. One idea is to

place hollow fibers filled

with crack-sealing material

into concrete, which if

cracked would break the fiber

releasing the sealant. Hollow

beads filled with sealant (for

self-healing) , and smart

paints, which will release red

dye (contained in capsules)

when cracked are also being

investigated. Concrete may

also heal its own cracks under

pressure with sufficient

humidity. Optical fibers

which change in light

transmission due to stresses

are useful sensors. They can

be embedded in concrete or

attached to existing

structures. Optical fiber

sensor systems have been

applied for on-line, and realtime monitoring of critical

components of structural

systems (such as bridges) for

detection and warning of

imminent structural failures.

Fundamentals and dynamics of

embedded optical fibers in

concrete, electro-rheological

(ER) fluids for the vibration

control of structures have

been investigated. Such ER

fluids can stiffen up very

rapidly (changing elastic and

damping properties) in an

electric field.

Several such smart materials

are studied to provide

sensoring and actuating

capabilities, i.e., shape

memory alloys, and surface

superelastic microalloying as

sensors and microactuators

and magnetostrictive active

vibration control.

CONCLUDING REMARKS '

It is clear that smart

materials, smart structures,

and control technologies, when

sufficiently developed and

broadly implemented, possess

an excellent potential to

impact the safety design and

construction of new structures

in the future. Furthermore,

the dynamic control and

intelligent technology also

offers a viable, alternative

solution to the nation's

enormous problems in the

construction of new civil

infrastructure and

rehabilitation of existing

structures which are deficient

in their functionality,

serviceability and in their

seismic, wind, or other

resistant capacities.

As such it is generally agreed

that dynamic control and

intelligence technology holds

considerable promises in the

design applications of civil

engineering structures, new

and existing. By its

interdisciplinary nature, such

technology also has the

potential impact to the future

development of a diverse field

of scientific or technical

subjects, such as material

220
science, vibration,

biotechnology, neural network,

and new devices.

No. of Proposals

Control

Technique Earthquake Wind

General

Vibration Waves

Random

Chaotic

Motion

Passive 12 8 4 1

Active 18 12 8 1

Hybrid 10 4 1

Total No.

Proposals 40 24 13 1 1

Table 1. Proposal Distribution by Control Techniques

Control

Element

No. of Proposals

Earthquake Wind General Vibration

Sensors 4 1 3

Actuators 6 5 3

Processors

(Algorithms,

Modelling,

etc.

)

13 4 3

Materials 2 1 2

Total No.

Proposals 25 11 11

Table 2 . Proposal Distribution by Control Elements

221
Earthquakes Winds

General

Vibrations

Applied to

Highway Systems

Performance

Evaluation 2 1 2 1

Monitoring

Detection

Maintenance

4 1 3 4

System

Modelling &

Identification

1 1

Reliability 3 2

Total No.

Proposals 10 2 8 5

Table 3. Proposals to Study System Characteristics, Modelling,

and Analysis
Earthquakes Winds Others

Elements/Systems

Types

of

Structures

Buildings 30 17 14

Bridges 4 1 1

Lifelines,

Infrastructures 3 1

Wharves 1

Traffic

Signal Structures 1

Claddings 2

Nonstructural

Elements,

Secondary Systems 2

Special

Purposes

Retrofit 3 2 1

Diagnosis

,

Monitoring 1 1 4

NDE 2

Table 4. Proposal Distribution by Applications

223

Significant Results From Standardization Work on Rubber Bearings for

Base-Isolation Buildings

by

' Yuji Ohashi' and Hiroyuki Yamanouchi^

SUMMARY

This paper reports the study on techniques

of effective use of the laminated rubber

bearings used for base isolation system,

the mechanical characters of the bearings

and the relation between the mechanical

characters and the effective factors of the

bearings.

KEY WORDS

base isolation system, rubber bearing,

standardization.

Comprehensive Project on New & Advanced

Materials by the Ministry of Construction.

1. PREFACE

The Ministry of Construction carries

on the Comprehensive Project on New & Advanced Materials. As one part of the

study, the laminated rubber bearings used

for base isolation system are studied as one of the new material for new structural

engineering of buildings.

In this study, some types of laminated

rubber bearings, the cracters of the rubber,

the mechanical characters of the bearings,

the methods of testing, design procedure,

production procedure of the bearings,

maintenance of the bearings and the base

isolated buildings and technical terms are summarized.

The mechanical characters of the bearings and the relation between the

mechanical characters and the effective

factors of the bearings are studied.

In this project, we intend to provide

the guideline of the effective use of the

rubber bearings.

In Japan, a lot of studies are already

carried out for base isolation systems and

rubber bearings. Many research papers which

deal with the characters of the materials of the rubber bearings, mechanical characters

of the bearings, design, analysis and response of base isolated buildings in case

of earthquakes are reported.

The Architectural Institute of Japan

issues " the Recommendation of the Base

Isolated Buildings

"

Some fifty buildings which adopt base is olation system are constructed or under

construction in Japan.

In the project, considering these circumstances, we intend to contribute to build

base isolated buildings with sufficient

safety by providing information and making

guideline of laminated rubber bearings.

*1 Senior Research Engineer

*2 Senior Research Engineer of

International standard

Structural Engineering Department, Building

Research Institute, Ministry of Construction

Tatehara-1, Tukuba-City, Ibaraki, Japan

225
In order to make the guideline, we have

been collecting much information on rubber,

rubber bearings and mechanical characters of the bearings which the structural engineers

should understand and we intend to arrange

the condition for the establishment of standards of the bearings.

We set up research committees in Japan

Rubber Association. The Ministry of Const- ruction and that association made research

corporation contract.

There have been many disastrous earthquakes in the history of Japan. The extreme forces of earthquakes have inflicted

great damage on building structures, and

it has long been a major task of Japanese

building technology to find ways of preven- ting danger of earthquakes.

Thus, it is highly necessary that buildings in Japan be able to safely withstand

attacks of earthquakes. Concept of earthquake proof design of buildings in Japan has

been to arrange structural elements appro- priately and to maintain sufficient strength, ductility and stiffness for the struc- tural elements.

Several decades of years ago, such kinds

of ideas as to isolate the basement of a building from ground were considered.

These ideas were called "base isolation

systems" , and a few researchers studied

and developed these ideas, but because of

lack of methods of analysis and devices,

these ideas were seldom realized.

After the second world war, the develop- ment of computer technology also promoted

structural analysis methods. Using computerized seismic analysis, construction techno

logy of super-high raised buildings, which use the earthquake ground motions recorded

in the United States, was developed around

1960.

With these knowledge, we have gotten the

behavior of buildings in cases of earthquakes. Laminated rubber bearings were

developed. Using the bearings, we became to

realize base isolation systems for

buildings.

These systems can be adopted not only for

buildings but also for other structures such

as bridges and machines. This research

project focus on base isolation systems of

buildings and laminated rubber bearings.

2. Base isolation structure

Advances in response analysis technology

had given the knowledge that seismic force

could be reduced by lengthening the natural

period of a building, and research was

carried on how this could be done.

Laminated rubber bearing was invented

around 1980. The bearing is made of several

alternating layers of rubber sheets and

steel plates and can withstand great forces

in a vertical direction, such as the selfweight of the building, while horizontally

it is flexible in character.

When they are installed in the foundation of a building, or the columns of its

lowest floor, they can support the selfweight of the building and at same time,

because of horizontal flexibility, can lengthen the building's natural period.

That reduces seismic force.

In Japan, it is already beyond the

experimental stage and has been introduced

gradually into actual construction. Compared

with conventional structure it is capable

of considerably reducing the effects of

seismic force.

Research has also done into other method

of absorbing or controlling the response

from an earthquake. One of them is to install a "damper". Dampers can be installed

independently in the walls or structural

braces, or can be used in combination with

laminated rubber bearings as part of a base

226
isolation structure system.

Records of seismic response of the buildings which use base isolation systems using

rubber bearings indicate that the acceleration response can be reduced to about 1/2 or

1/3 of the earthquake ground motion. Seismic response of usual buildings usually becomes

twice or three times of ground motion. We

can understand the effects of the base

isolation systems apparently from these

records.

3. KINDS AND CHARACTERS OF RUBBER BEARINGS AND DAMPERS

3.1 RUBBER BEARINGS

High strength and stiffness for vertical

direction are required for rubber bearings to sustain self weight of buildings. If the stiffness for vertical direction is not sufficient, vertical response will be larger

and rocking vibration may occur.

For lateral direction, low stiffness and large deformability are required to have

long natural period of building.

Steel plates of the bearings constrain

vertical deformation and that gives high

strength and stiffness for vertical direction. For lateral direction, deformation is not constrained, soft character of rubber

gives low stiffness and large deformability.

Shapes of the rubber bearings are usually

flat cylinders. Capacity of load sustain

ability, stiffness and deformability depends on the shape, size, thickness and number of

rubber sheets. (Fig. 1)

The lower the thickness is and the smaller

the number of rubber sheets are, the higher

the vertical stiffness of the bearings is.

The lateral stiffness of the bearings is usually about 1/1000 of the vertical

stiffness of the bearings. Shape of the

bearing relate with the character of deformability.

We use the index which shows shape, "flat

ness" , of a bearing. We call the index

second shape factor, 82, which means diameter/thickness of the bearing. Usually,

bearings S? of which are over 2.0 are used.

(Fig. 2)

When Sa is large, namely, the baring is very flat, sheer deformation becomes

dominant and the character of deformation

for lateral direction becomes stable. When

the 8^ is not so large, ratio of bending

deformation in total lateral deformation

becomes large and the character of deformation for lateral direction becomes

unstable.

The role of rubber sheets is to maintain

low stiffness and high deformability of the

bearings for lateral direction. The role of

steel plates is to maintain high strength

and stiffness of the bearings for vertical

direction by constrain of deformation of

rubber sheets.

3.2 KINDS AND CHARACTER OF RUBBER BEARINGS

In Japan, we have four kinds of rubber

bearings as showed below.

(D bearings of natural rubber

(2) bearings of high damping rubber

(3) bearings of lead and rubber

(D bearings of slip layer and rubber

(X) bearings of natural rubber

This type of bearings are used most often.

Natural rubber is used and have rarely

damping. Up to 200% of sheer deformation of

rubber itself, character of the deformation

of the bearings is usually elastic. When the

deformation becomes larger than that,

hardening will occur. The sheer deformation

of the rubber when the rubber is broken is about or over 400%.

(D bearings of high damping rubber

227
This type of bearings use synthetic rubber

which has high damping character.

@ bearings of lead and rubber

This type of bearings use lead and install

lead in the center of bearing. Plastic

deformation of the lead absorbs energy of

vibration and gives damping character.

(Fig. 3)

® bearings of slip layer and rubber

This type of bearings use slip layer of

such as fluorine and rubber bearings. In cases of small earthquakes, because of

fliction of the layer, rubber bearings act and in cases of large earthquakes because of

the slip of the layer seismic force of the

earthquake dose not become large. Springs

or elastic rubber bearings are used with

these types of bearings to return the

position of lateral deformation of the

building to original position.

(Fig. 4)

(D and (3) type bearings have damping, so these bearings do not require dampers. But,

(I) type bearings require dampers to have

damping character.

Cover rubber is used for each types of

bearings to protect bearing from corrotion

and to have durability.

Flange steel plates are installed on the

face and bottom of the rubber bearing.

3.3 DESIGN OF RUBBER BEARINGS

It is required for rubber bearings to have below mentioned mechanical characters.

(a) vertical strength to sustain the load of building

(b) lateral stiffness to give suitable

natural period

(c) lateral deformability

Design procedure is described below.

1) to settle the design condition such as vertical strength, lateral stiffness and

lateral deformability, etc.

2) to settle the kinds of rubber material

and characters of rubber

3) to settle the shape of the bearing such

as diameter and total thickness

4) to settle the structure of the bearing

such as number and thickness of rubber

sheets,

5) to certify whether the design meets the

design conditions of the bearing

3.4 KINDS OF DAMPERS USED WITH RUBBER B EARINGS

When the rubber bearings do not have

enough damping, some kinds of dampers are

used for base isolation systems. Kinds of

the dampersare followed.

1) viscous damper (Fig. 5) 2) plastic damper (Fig. 6)

3) friction damper (Fig. 7)

4. STRUCTURAL DESIGN PROCEDURE OF BASE ISOLATION SYSTEMS

Usual procedure of structural design of

base isolation is followed.

1) to settle design criteria such as allowable response and size of earthquake

2) study and survey of site and character of

earthquake which is anticipated at the site

3) to settle the natural period of the

building and characters of bearings

4) structural design based on the initial response analysis

5) certification whether the design meets

228
the design criteria

6) design of details

There is a special approval system by

the Minister of Construction for new structures which do not comply with the standards to prevent the building codes

from becoming too rigid. The system has

provided the way for a lot of new building

structure to date.

Before the approval of the Minister, the

safety of the buildings shall be evaluated

by the committee of the Building Center of

Japan under the supervision of the Minister.

Base isolation systems are the newly

developed systems and do not comply with the

existing building codes.

5. USUAL CHARACTERS OF RUBBER BEARINGS

Mechanical characters of bearings and

dampers which are already designed are as

followed.

1) vertical load sustain ability for permanent loads - 20 ~ 90 kgf/cm^

2) natural period of base isolation system

- 1.0 ~ 3.0 second

3) damping coefficient - 5 ~ lb %

These characters are shown in Fig. 8 ~

Fig. 10

6. CLOSURE

The construction of base isolation

systems requires the approval of the Minister of Construction and evaluation by

the comittee of the Building Center of

Japan before the construction work of the

building. In order to evaluate the safety

of the buildings which use base isolation

systems

We intend to provide guideline to give

effective information for use of rubber

bearings by studying above mentioned items

To evaluate the safety of such buildings,

adding the allowable stress design,

dynamic response analyses are taken using

records of the ground motions of past

earthquakes, where some typical ground

motions such as 'El Centro' and the

ground motions which were recorded near the

construction site of the building are used.

At this time, the scales of the ground

motions are normalized and 2 levels of the

ground motions are adopted.

Maximum velocity of the ground motions

of level 1 is some 25cm/sec and that of

level 2 is some 50cm/sec. Response from

level 1 ground motions shall not exceed

elastic limit of the bearings and that

of level 2 ground motions shall be within

apt value of deformation.

229
D

"•— Cover Rubber

Inside Rubber — Sleel Plate Sz = D/T

Fig. 1 Rubber bearing

(D Single Rubber (2) Laminated Rubber

Fig. 2-1 Single rubber and laminated rubber

Vertical Load Lateral Load Load and Deformation

Single

Rubber

I .P—*-

Laminated

Rubber

Completion

Shear

Fig. 2-2 Single rubber and laminated rubber

230
Lead

'^m

Fig. 3 Lead rubber bearing

231
Movement

Viscus Material xj I /

Resist^

I

Fig. 5 Viscus damper

m

//

Movement

JL -J- ~

//

Steel Bar

Plastic Deformation

Fig. 6 Plastic damper

Movement

Frictional Material Resistance

Fig. 7 Frictional damper

10 Number of Examples

5

-

15 Number of Examples

0 10 20 30 40 50 60 70 80 90 100 (<».:t<?AJ) ^

Fig. 8 Stress of bearing

1.2 1.4 16 IB 20 22 24 26 28 30 30

National Period (Second)

Fig. 9 Natural period of building

•

•

* si • •

•

• >*• > •

•

0 I 2 3 4 5 6 7 Sz (D/hn) -

Fig. 10 Sz and stress of bearing

232
Effect of Temperature on the Dynamic Behavior of Base-Isolated Bearings

by

Osamu Nakano', Hideyuki Taniguchi^, and Hiroaki Nishi^

SUMMARY

Our concern is whether the base-isolated

bearings, which have recently been introduced to the field of the bridge design in Japan, assume its performance in such cold regions as Hokkaido. For this reason, full-size base-isolated bearing models in accordance with an ongoing bridge construction project were prepared in order to pay attention to their behavior

at low temperatures through indoor experiments. As a

result, it has been shown that the base-isolated bearing

hself has been much effected by temperature, however,

it has not been understood that how the change of

performance on the base-isolated bearings could

influence the response of the entire bridge.

Therefore, further investigation has been made

by performing dynamic analysis.

As the result, it was learnt that the bending

moment at the base of pier could increase by 40-50% in a certain condition. KEY WORDS: Base-isolated bearing, Cold region.

Effect of temperature variation. Dynamic analysis

1. INTRODUCTION

Generally, the base-isolated bridge is so

designed as to decrease the seismic force by making the natural period of the entire bridge longer with a special

device called a base-isolated bearing, which supports

the superstructure softly on the bearing part in horizontal directions and has a damping characteristic. The Hokkaido Development Bureau is now

constructing the Onnetoh-Oohashi Bridge on National

Trunk Road 44 in Nemuro in which side-span on the Kushiro side was selected as such the above-mentioned

type of bridge.

As concern about the performance of baseisolated bearings was presented by a regional

construction office, that meant the performance might

be influenced by temperature drop, indoor experiments

in terms of stiffness and damping factor of rubber

material properties have been performed for the past

three years, which include 50-cycle shear tests, compression tests, shear tests and shear history tests on

various temperatures.^^ Two types of full size models of

lead rubber bearing (LRB) and high damping rubber

bearing (HDR) are used in the experiments. The models were designed and produced in accordance with the basic design of base-isolated bearings on the OnnetohOohashi Bridge. As the result, it was proved that the base-isolated bearing itself had the strong dependency on temperature with regard to the stiffness, which was

not expected before the tests. In this connection, further investigation has been made by conducting dynamic

analysis of the entire bridge from the point of view that how the above-mentioned tendency might contribute to the response of the entire bridge.

2. OUTLINE OF THE ONNETOH-OOHASHI

BRIDGE

The Hokkaido Development Bureau has

decided to plan a special bridge with base-isolated

bearings under the schedule of the reconstruction of the existing Onnetoh-Oohashi Bridge on National Trunk

Road. In Japan, eight base-isolated bridge projects were planned at the first stage, out of which only the

Onnetoh-Oohashi Bridge project is located in a cold

region like Hokkaido. The construction site is located at the bottom of the Nemuro Peninsula where the lowest mean temperature is -9.1°C in the past 30 years. The

present Onnetoh-Oohashi Bridge is the steel curvedchord Warren truss + steel composite girder bridge with

the total span of 99.5 m, built on the Onnetoh in Nemuro, Hokkaido and the new bridge (Fig. 1) after the reconstruction will be 456 m long with the side-span of

four-span continuous plate girder and the main span of

Nielsen-type Lohse girder with the span of 140 m. The

reconstruction is expected to be completed in 1995. Its base-isolated structure is adopted only in the side-span on the Kushiro side and the normal rubber bearings

(RB) will be used for both ends Al and P4. This year, the side-span bridge with baseisolated bearings is supposed to be erected and the

vibration test is scheduled to be done by our laboratory

late fall.

3. OUTLINE OF BASE-ISOLATED BEARINGS

The base-isolated device consists of two parts: the isolator softly supporting the structure and the damper with the damping effects by energy absorption.

The isolator uses the material called the layered rubber

consisting of rubber and steel alternately layered and the damper uses lead or steel bars. The base-isolated

devices utilized in bridges are often installed at the bearings part, so that they are especially called baseisolated bearings. Therefore, the base-isolated bearing

should be of such a compact design as an integration of

isolator and damper. The base-isolated bearings

currently available are roughly divided into the

following two kinds: those in which lead plugs are

inserted so as to provide the layered rubber called lead rubber bearings with damping effects and those using

the specially-designed rubber called the high-damping

1. Head, Structure Division, Civil Engineering Research

Institute, Hokkaido Development Bureau

2. Research Engineer, Structure Division, Civil

Engineering Research Institute, Hokkaido Development

Bureau

233
rubber in which the rubber material itself of the layered

rubber called high-damping rubber bearings has the damping effect. Each kind of base-isolated bearing is as shown in Fig. 2.

4. TEMPERATURE DEPENDENCY OF BASEISOLATED BEARINGS

The dynamic characteristics of base-isolated

bearings are presented by the equivalent stiffness (Ke)

and the equivalent viscosity damping factor (heq)

obtained from the shear force - shear displacement

curve approximated with the bilinear spring as shown in Fig. 3. Ke and heq can be defined as the following

expressions in general:

AE 1 AW

Ke = , heq = EC jt W

Wliere, AW: Area of parallelogram ABCD

W: Area of triangle AEC

The shear tests were performed using a fullsize model for obtaining the above Ke and heq from the actual shear force - shear displacement curve^^ The

design displacement of the model can be acquired by

the check of ultimate bearing capacity of reinforced

concrete piers for lateral force with the resulted value of

15 cm. For details such as the design conditions of the specimen, refer to Bibliography 1). The shear test was

performed under a vertical load of 92 tons (= dead-load

reaction) as the horizontal displacement is 150 mm (=

design displacement) and loading is 40 seconds. For the temperature of specimens it was changed from +30°C to -20°C. The equivalent stiffness and the equivalent

viscosity damping factor on each temperature level is as shown in Table 1. The shear force - horizontal

displacement curves of LRB, HDR measured at room

temperature and -20°C are shown in Fig.4. For the stability against the temperature change

of base-isolated bearings, it is stipulated that the ratio of the equivalent stif&iess is within 1.5 between +30°C

and -30°C in order to suppress the change of the natural period of the bridge within 20% in the "Guidelines for the base-isolated bearing design method for road bridge

(draft)"^^(liereinafter referred to as the guideline).

But, the ratio of the equivalent shear stiffness

at -20°C to that at 30°C is 1.63:1 for LRB and 1.59:1

for HDR, where the average of the values of the two

specimens were taken for LRB, HDR respectively.

Recently additional experiments have been

done at -30°C and room temperature for each bearing

under the condition that the loading cycle is 100

seconds due to the limitation of capacity of the test machine.

The following data were obtained from the

experiments.

Tlie equivalent stiffness at room temperature

are 1391 kg/cm for LRB and 1832 kg/cm for HDR, and

those at -30°C are 2089 kg/cm and 3390 kg/cm

respectively. The ratio of the stiffness between the two

temperature are 1.58, 1.85 for LRB, HDR respectively.

5. ANALYTICAL MODEL AND METHOD

In this analysis, the focal points are placed on

the relative displacements of the bearings where relative displacement is defined as the difference between the

top and the bottom of the base-isolated bearings and the maximum bending moments at the base of the piers in the case of temperatures between room temperature

(approx. 20°C) and -20°C.

For the analytical model, the side-span bridge

with the base-isolated bearings was picked up alone

due to the simplification. Then, the bridge was replaced

by the lumped mass model as shown in Fig. 5. In which,

the pile foundations were also replaced by the threetypes springs. For the modeling of the base-isolated

bearings, which is shown in Fig. 6, a linear-spring

model with the equivalent stiffness and the equivalent

viscosity damping factor was applied to the equivalent

linear analysis and the bi-linear model was adopted in the case of the non-linear analysis.

In the both analysis, the piers and base springs were treated as elastic bodies and the latter was

calculated from N value of the ground depth intended

for the calculation according to the calculation method

described in "Road bridge specifications for seismic

design edition"'*^ Tlie time history analysis was adopted

as the analytical method. For the both analyses, the computer program "NLDAP"^^ for two dimensional

analysis use that was developed in our laboratory, in which calculation was made with the direct integration

of Ncwmark-^ method and the Proportional damping was assumed for the viscosity damping. In this case, the analysis has been carried out in the longitudinal

direction only, because the Onnetoh-Oohashi bridge is allowed to move toward the above-mentioned

direction.

6. DYNAMIC ANALYSIS

6.1 Eigenvalue Analysis

First of all, the eigenvalue analysis was

performed in the respective temperatures in order to

identify the vibration characteristics of the bridge.

The equivalent stiffness and the equivalent

viscosity damping factor for each pier at room

temperature and -20°C are listed in Table 2. Fundamental value for each bearing is based on the indoor experiments where the bearing was deformed up

to the design displacement ( = 150 mm ). But, the

values in the table are get by multiplying the average of

the fundamental values (two specimens for each of LRB

and HDR) by the number of bearings (six units). The normal rubber bearings (RB) will be used

in the bearings of Al and PI, however, because the experimental results are not obtained at room

temperature, the design value^^ 363 t/m for the stiffiiess

is adopted as it is, while at -20°C that value is decided

as 472 t/m, 1.3 times larger than the design value

assuming that the change of the stiffness of normal

rubber bearing is similar to that of the base-isolated

bearing at this temperature. Table.3 represents natural periods for the first ten modes by temperatures. The table indicates that the natural period of the first mode on -20°C was decreased

234
by approx.l3 - 14 % compared with that of the room

temperature due to the increase of the base-isolated

bearing stiffness, however, those of the other modes

changed slightly. Since the horizontal displacement of

the superstructure dominates in the first mode, it might

be considered that the effect appeared strongly. Tlie first mode in the case of LRB at room temperature is shown

in Fig. 7.

6.2 Result of time history analysis

6.2.1 Equivalent linear analysis

The strong motion observation has being

conducted since 1975 when a analog-type seismograph was installed at the existing Onnetoh Oohashi Bridge.

The strong motion record used in the time

history analysis is one of recorded data from the

earthquake which took place on January 12, 1978 with

magnitude at 5.1. The waveform recorded on the strong

motion seismograph and the acceleration response

spectrum are shown in Fig. 7 and 8 respectively.

The analysis was performed under the

following conditions:

Integral time interval: At = 0.01 second

Response time: T = 10 seconds

Newmark-/3 method: /3 = 1/4

The damping factor h of the entire bridge was

calculated using the following expression which is indicated in the guideline.

" - 2 2

5; K B i u B i + ^ p^—

Where, UBi: Design displacement of the base-isolated

device

hfii: Damping factor of the base-isolated device

hpi: Damping factor of the substructure Kbi: Equivalent stiffness of the base-isolated

device

Kpi: Equivalent stiffness of the substructure For the damping factor of the first and second

modes for deciding a and ft values of proportional

damping, we adopted the damping factor h of the entire bridge for each mode. The h values acquired from the above expression are as shown in Table 5. As the result of the analysis performed under

the above conditions, the comparison between room

temperature and -20°C is as shown in Table 6 and 7 on

the relative displacement of bearing and on the bending

moment of base of pier.

From these tables, it is revealed that the relative

displacement of bearing at -20°C is a little larger than

that at room temperature in case of LRB, however, in case of HDR, the displacement value varies little even with the change of temperature. With regard to the

comparison of the bending moment, it is proved that the bending moment increases only by several percents at -20°C compared to that at room temperature in case of

LRB, however, as much as 40 - 50% increase in case of

HDR. The reason why the bending moment increased

largely according to the temperature drop might be

considered as follows

.

1, The variation of the equivalent stiffness and the

natural period was similar to each other between LRB

and HDR

.

2, Therefore, it is difficult to consider that the above

variation contributes to the increment of the bending

moment.

3, Consequently, the increment is supposed to be arose from the relationship between the natural periods of

entire bridge and the response characteristic of the

inputted seismic wave.

In conclusion, it was proved that the change of

temperature didn't always effect the response of the

entire bridge.

6.2.2 Non-linear analysis

Since the dependency on temperature of the base-isolated bearings was focused on in the previous

section, its influence on the entire bridge was

investigated by carrying out the equivalent linear

analysis on various temperature. As the results, it was

clarified that the change of temperature didn't always

contribute to the response of the entire bridge.

Tlierefore, without considering the temperature and the type of the bearing, the following hysteresis curve was

assumed, in which a bilinear spring model was adopted,

and non-linear analysis was performed in order to compare the results with those of the equivalent linear analysis.

The first stiffness, the second one and the

yield displacement of the bilinear spring model (See

Fig.3) for each pier are as follows. The first stiffness : 5550t/m

The second stiffriess : 854t/m

The yield displacement : 16.44mm

As the equivalent stiffness is the function of a

horizontal displace is calculated to be 1360tfym in this case because the design displacement is decided to be

150mm.

The bending moment at the base of pier

obtained from each analysis, in which the same seismic wave using in the previous section is inputted is listed in Table 8. As a result, the value of bending moment in the non-linear analysis is 4-5 times larger than that of the equivalent linear analysis

.

This difference might be arose from the reason why the horizontal displacement was assumed to be

150mm, In other words, the rigidity of the equivalent

linear spring was too small.

In fact, the maximum displacement, which was

calculated in the equivalent linear analysis, was about

20mm.

Therefore, two additional analysis were carried tout . In the first one, the equivalent stiffness was

assumed to be 4714tf/m, which corresponded to the horizontal displacement of 20mm. In the second one,

the equivalent stiffness was assumed to be equal to the

first stiffness. These results are shown in Table 9. However, the values in the non-linear analysis

are still 60-90% larger than those of the former, 50-

235
70% larger in the case of the latter. In this connection, it could be clarified that the time when the maximum bending moment took place were different in the three cases. Therefore, careful attention should be paid the equivalent stiffness in the case of the equivalent linear analysis.

7. CONCLUSION

The points clarified from this investigation are

listed as follows : 1) Tlie equivalent stiffness on base-isolated bearings of

the both, which were tested, increased by approx. 40%

due to the change from room temperature to -20°C. It is also shown that in spite of the above fact, the natural period of the bridge is shortened by approx. 14% for both LRB and HDR.

2) As far as natural periods of the bridge was concerned,

the influence, which was aroused from the change on

the stiffness of the base-isolated bearings due to the temperature drop, appeared only in the first mode.

3) Although the variation on the equivalent stiffness was similar to each other between LRB and HDR, for

the base-isolated bearing, with LRB the displacement was larger by around 10% at low varied little even upon

the temperature change.

4) For the bending moment at the base of pier, the result by the equivalent linear analysis was different between

LRB and HDR, showing only several % increase with

LRB, while 40-50% increase with HDR. It is supposed

that the main reason why the response was different might be arose from the relationship between the natural period of the entire bridge and the predominant period

of the strong motion, because the natural period in the

case of HDR is about 10% shorter than that of the LRB

and the predominant period of the inputted seismic wave is around 0.6sec. 5) As a result of comparison between the values of the equivalent linear analysis and the non-linear one, it was

clarified that careful attention should be paid to the

equivalent stiffness in the case of the equivalent linear analysis. It is also indicated that the response might be

seriously influenced by the assumed horizontal

displacement in the equivalent linear analysis in which

time history analysis would be made.

8. POSTSCRIPT

We would like to perform the linear and nonlinear analysis using different magnitude seismic waves

to compare the both values, to grasp more exact

behavior of the base-isolated bridge and to acquire the tendency of the change of sectional force in the future because in this analysis a single inputted seismic wave

was only adopted.

Bibliography

1) Structure section, Civil Engineering Research

Institute: Experimental Research Reports on

Base-Isolated Bearings at Low Temperatures,

1991. 3 (In Japanese)

2) H. Taniguchi, O. Nakano: On the Behavior of

Base-Isolated Bearings at Low Temperatures 7th U.S.-Japan Bridge Workshop, 1991. 5

3) Japan Institute of Construction Engineering : Guideline of Base-Isolated Bearing Design

Method for Road Bridge (Draft), 1991. 3 (In

Japanese)

4) Japan Road Association : Road Bridge

Specifications for Seismic Design, 1990. 2 (In

Japanese)

5) Structure Division, Civil Engineering Research

Institute: Work Reports No. 2 on Base-Isolated

Bearing-Included Seismic Design program, 1990. 2 (In Japanese)

6) Kushiro Construction Department: Preliminary

Report on the Dynamic Analytical Model for the Onnetoh-Oohashi on National Trunk Road 44 in Nemuro, 1990. 3 (In Japanese)

236
OL tS 78 01 12 lOOOO.O

sooao

KOOO

leoo o :

sooo p

I iilllii

O.t 0 9 1.0 Z.0 PERIOD (SECt

Fig. 9 Acceleration response spectrum

Table.5 DampinR factor of the entire bridfie:h L R B HDR

Room T. -20^ Room T -20^

h (%) 26 23 10 11

A 1 P 1 P 2 P 3 P4

L Temp. R.T -20 TC R.T -201C R.T -20 "C R.T -20 x; R.T -20'C R (5 (mm) 18.5 21.0 18.5 20.6 18.5 20.5 18.7 19.4 26.1 26.6 B Ratio 1.14 1.11 1.11 1.04 1.02 H Temp. R.T -20 x; R.T -20 1C R.T -20 x; R.T -20 1C R.T -20'C D 5 (mm) 26.7 27.7 26.3 26.8 26.6 26.4 24.5 24.5 40.1 33.1 R Ratio 1.04 1.02 0.99 1.00 0. 83

Table.7 Bending moment at the base of pier P 1 P 2 P 3 P4

L Temp. Room T. -201C Room T. -20°C Room T. -ZOV Room T. -20X: R M(tf.m) 129.67 134.38 142.00 147.34 197.54 202.68 16627.0 16587.0 B Ratio 1.04 1.04 1.03 0.998 H Temp. Room T. -20V Room T. -20t: Room T. -2ox: Room T. -2ox:

D M(tf«m) 156.94 225.07 184.37 281.14 214.36 305.24 23321.0 21248.0 R Ratio 1.43 1.52 1.42 0.91

Table.8 Comparison between linear and non-linear analysis

P 1 P 2 P 3

Non-1 (tf'm)

A

655.62

(0.870)

743.86

(0.870)

823.27

(0.890) Linear (tf«m)

B

128.93

(0.390)

147.38

(0.670)

194.14

(0.630) Ratio (A/B) 5.09 5.05 4.24 The figures in parentheses is a occurrence time.

Table.9 Bending moment at the base of pier in case of changing the equivalent stiffness P 1 P 2 P 3

20mm 1st. stiff. 20mm 1st. stiff. 20mm 1st. stiff. M (tf-m)

C

403.81

(0.600)

444.08

(0.590)

440.75

(0.610)

484.69

(0.900)

434.94

(0.590)

483.2

(0.890) Ratio(A/C) 1.62 1.48 1.69 1.53 1.89 1.70 The figures in parentheses is a occurrence time.

237
Table.3 Natural period (s) LRB HDR

Mode Room T. -20'C Room T. -20'X:

1 1.20973 1.05605 1.09633 0.94095 2 0.50958 0.50902 0.50957 0.50902

3 0.48331 0.48258 0.48278 0.48176 4 0.39794 0.39758 0.39775 0.39728

5 0.29420 0.29407 0.29414 0.29397 6 0.23620 0.23618 0.23620 0.23617

7 0.22022 0.21838 0.21863 0.21589

8 0.18659 0.18566 0.18579 0.18440

9 0.17457 0.17385 0.17393 0.17287

10 0.15261 0.15259 0.15259 0.15256

Fig. 7 Figure of First mode (LRB: Room lemperalure)

Table.4 Slronfi motion data Date 1978.1.12 Magnitude 5.1 Depth(km) 40

Epicentral distance(km) 31 Max. acc.(Ral) 120.57

ONNETOH GL LG 78.01 . 12

7 5 10 0 12 5 15 0,

GAL MAX. • 120. 57 (SMAC-Q)

Fig. 8 Inputted seismic wave

238
Table.l Equivalent stiffness at each temperature Specimen Temo CC") E<i.stiffness(kg/cm) Dsinp.fBctor

LRB 1

+ 30 13 3 2 0. 3 4 Room temp. 16 4 5 0.3 4

0 18 3 2 0. 3 1 - 1 0 2 0 7 2 0. 3 1 - 2 0 2 2 17 0. 3 1

LRB 2

+ 30 13 4 6 0.3 4 Room temp. 16 19 0.3 4 0 19 9 4 0.3 3 - 1 0 2 0 0 5 0. 3 1 - 2 0 2 15 5 0. 3 0

HDR 1

+ 30 17 8 7 0. 1 3

Room temp. 2 0 7 9 0. 1 2

0 2 2 4 9 0.12

- 1 0 2 5 8 6 0. 1 4 - 2 0 3 0 4 4 0. 1 4

HDR 2

+ 30 19 4 4 0. 1 1 Room temp. 2 13 6 0. 1 2 0 2 3 4 0 0. 1 3 - 1 0 2 5 15 0. 1 3 - 2 0 2 8 7 4 0. 1 3

: Observation Point T/f777

Superstructure

[ Bearing part

Subslniclure

Fig. 5 Analysis model Fig. 6 Magnified bearing pari

LRB HDR

Room T. -20X; Room T. -20'C Ke(t/m) 979 1312 1265 1775 heq(%) 34 31 12 14

239
To Kushiro

itmy""""" t*ooa>

P-l P-2 P-S

To Nemuro

p-S p"« P-7 P-9 P-9 P-C P-ll P-12

Fig. 1 Onneloh-Oohashi Bridge

Higli damping rubber ^/

Lead rubber bearing High damping rubber bearing

Fig. 2 Base-isolaled bearings

Shear

fo

1st. stif&iess

. c

B / 7 1 1

/ .r' 1 2nd. stiffness

-~\ °

Shear displacement

yield displacement

Fig. 3 Bilinear-type spring

Shear Force jSOt

Specimen :LRB1 Displacement: ISOmni

20t

Room Temperature

Shear Force rsot

Specimen: II DRl Displacement: 150mm •-

Horizontal

Displacement

-20*C

Room Tempera tur(

Horizontal

Displacement

Fig. 4 Hysteresis loop comparison

240
Dynamic Centrifugal Model Tests of Embankments on Liquefiable Ground (Part 2)

by

Yasuyuki Koga', Jun-ichi Koseki^, and Akihiro Takahashi^

ABSTRACT

Djmamic centrifugal model tests were performed

to study a seismic behavior of embankments on a liquefiable ground using sinusoidal waves of a different frequency and two kinds of irregular waves as a shaking acceleration. Effects of the shape of the shaking acceleration on the excess

pore pressure in the ground and on the settlement

of the embankment were investigated, and the applicability of a cumulative damage concept in equalizing the irregular shaking wave was examined

KEY WORDS ; Embankment, Liquefaction, Dynamic

Geotechnical Centrifuge, Irregular

Wave, Cumulative Damage Concept

1

INTRODUCTION

Liquefaction of a sandy ground causes a settlement of an embankment founded on the ground. A lot of model tests have been executed

to reveal the seismic behavior of the embankment,

but few of them have been done under a high

confining stress. A centrifugal model test is a test method which

can simulate the same stress condition in a scaled model ground as in an actual ground. It has been applied to a dynamic problem using an earthquake simulator" . The writers have reported dynamic centrifugal

model tests''' of a horizontal sand layer and an embankment on the layer using a sinusoidal wave

of 60Hz as a shaking acceleration. This paper

deals with continued tests using a sinusoidal wave of lOOHz and two kinds of irregular waves as well as the above mentioned tests.

2. TEST PROCEDURE

Embankment models and horizontal layer models

as shown in Fig.l were made in a rigid soil

container of 80cm in length, 30cm in height and 10cm in width. A sand layer was prepared by pouring Toyoura

sand through air. The surface of the layer was

rounded in accordance with a rotational radius.

In the embankment models a colored sand mesh was drawn to observe the deformation of the sand layer through a transparent front glass of the container.

After setting the container in a vacuum box the sand layer was saturated with silicone oil which

is 30 times as viscous as water. An embankment

was made of a mixture of Toyoura sand and a clay-sand which has a ratio of 4:1 in weight and

a water content of 15%. A horizontal shaking was conducted after

applying a centrifugal acceleration of 30g.

Response accelerations, excess pore pressures and

a settlement of the embankment were measured by

transducers located as in Fig.l. It was repeated

several times increasing an amplitude of the shaking acceleration. A summary of the test condition is shown in Table 1. Sinusoidal waves

of 60 Hz and lOOHz, and two kinds of irregular

waves were used as the shaking acceleration. The shapes of these shaking waves are shown in Fig. 2.

3

METHOD OF DATA ARRANGEMENT AND CALCULATION

The general procedure of the calculation is similar to that of the writers' former study"*'.

3 . 1 Equivalent Shak ing Acceleration

An equivalent sliaking acceleration for the cases using irregular waves was calculated based

on a cumulative damage concept. The procedure of the calculation is shown in Fig. 3.

1) Calculation of Dynamic Shear Stress Ratio

The effective overburden pressure ovo' and the dynamic shear stress ratio t<j/<Jvo' were calculated one-dimcnsionally by Eqs (1) and (2) at a point: where the pore pressure transducer was installed (PI to P6).

«»vo' = v ' • z + v» • II (1)

in which - ' = effective unit weight, z = depth

from the ground surface, yi ~ total unit weight, H = height of embankment.

== E 5^^^ Ai (2) O vo O vo I f?

in which <Jvo' = initial effective overburden

pressure, Ai = response acceleration at the 1-th

layer, h 1 = thickness of the i-th layer, v i = unit weiglit of the i-th layer, g = acceleration

of gravity.

Although the response acceleration was

'Head, Soil Dynamics Division, Construction

Equipment and Method Department, Public Works

Research Institute, Ministry of Construction,

JAPAN. ^Researcher, do. ^Assistant, do.

241
amplified or reduced due to liquefaction, It was assumed to be the same as the shaking

acceleration for simplicity.

2) Assumption of Liquefaction Resistance Curve

The liquefaction resistance curve was assumed

to be given by Eq.(3), which was determined

referring to the liquefaction resistance curve of Toyoura sand obtained from cyclic torsion shear

tests as shown in Fig. 4.

T <i/<j„o' ^ 0.3 ; ra/<J„o*- 0.3=-1.25 log (Nc/NM

Td/omo' < 0.3 : ra/omo'- 0.3=-0.11 log (Nc/N') (3)

in which Nc = number of cycles to cause liquefaction when uniform cyclic stress ratio

Td/omo' is applied, N* = parameter to locate

the liquefaction resistance curve, omo* = effective mean stress, which was calculated from

the effective overburden pressure ovo* by

assuming the coefficient of earth pressure at rest Ko = 0.5. Eq.(3) can be rewritten as

Nc/N' f ( T d / O mo ' ) (4)

in which f = equation to give the relation

between cyclic stress ratio and normalized number

of cycles.

3) Calculation of Damage Factor When the irregular train of the dynamic stress ratio is divided into every half cycle, the damage factor D up to the i-th half cycle is defined in the cumulative damage concept as

.(Nc)i (5)

in which (Nc)! = number of cycles to cause liquefaction when the i-th half cycle stress ratio (rd/<Jino')l is repeatedly applied.

The damage factor was calculated from the equation below, which was obtained by

substituting Eq.(4) into Eq.(5).

D =

1 0.5 N' f {( T d/ <J„o'),} (6)

4) Calculation of Equivalent Shaking Acceleration

If an uniform cyclic shear stress of 20 cycles

is applied to give the same cumulative damage as the irregular wave, the amplitude of the equivalent shear stress ratio ( t d/ <j mo ' )o.i can

be calculated from the equation below.

D =

1 0.5 N' I f{(Td/omo*)«q}

20 N'- f {( t d/omo')..,} (7)

After converting the effective mean stress a into the effective Dverburden pressure ovo', the equivalent shaking jicceleration A«q was

calculated by substituting the the equivalent

shear stress ratio into Eq.(2).

{ r d/ " vo ' ) sq ~ _J i • hi(8)

The coefficient of equalization C«q was defined

and calculated as

A max

(9)

in which A„ox is the maximum amplitude of the shaking acceleration before the equalization.

The relationship between the equivalent shaking

acceleration A«<, and the settlement of the embankment for the cases using irregular waves

was arranged to compare with the ones for the cases unlng the sinusoidal waves. The relationship between the equivalent shaking

acceleration A»q and the excess pore pressure in the ground was also compared in the same way.

3.2 Liquefaction Re s istance Factor

The change of the response acceleration due to liquefaction was not considered in the preceding

section, but was taken into account in this

section in order to evaluate the dynamic shear

stress ratio in the ground precisely.

Although the parameter N' to locate the liquefaction resistance curve did not affect the calculated equivalent shaking acceleration Aoq

in the preceding section, it was estimated in this section in order to evaluate the liquefaction resistance of the ground.

1) Calculation of Liquefaction Resistance

The liquefaction resistance in the free ground

was evalunted based on the damage factor I) by

the following procedure.

When the assumed value of the parameter N' is correct, a value of D ^ 1.0 indicates a liquefied condition and D < 1.0 a non-liquefied

condition. Therefore following equations are obtained for respective conditions.

Liquefaction condition:

o.r.

f {( t d/om„')l}

Non-liquefaction condition:

0.!. N ' > L"

f {( T d/<Jmo*)l}

(10)

(11)

The liquefaction condition was defined to be

the state where the measured excess pore pressure

was equal to the effective overburden

pressure ovo' obtained by Eq.(l), which means

the excess pore pressure ratio lu defined in the

242
following equation was 1.0. The number of cycles

required to cause liquefaction was counted at every pore pressure transducer for both the cases using irregular waves and the cases using

sinusoidal waves.

a vo

Substituting the dynamic shear stress ratios up

to the cycles to liquefaction into Eq.(9) and those up to one half cycle before the above

cycles into Eq.(lO), the possible range of the parameter N* which locates the liquefaction

resistance curve was estimated.

2) Calculation of Liquefaction Resistance Factor A liquefaction resistance factor Fl is defined

as

Fi.= —^ (13)

in which R is the dynamic shear resistance ratio, L is the shear stress ratio during earthquakes. An equivalent liquefaction resistance factor

Fli up to the i-th half cycles was defined as well by the following equation.

F,..=-^ (14)

in which R^o is the dynamic shear resistance

ratio corresponding to an uniform wave of 20cycle, L I is the equivalent shear stress ratio up to the i-th half cycles. The shear resistance ratio Rzo was computed by

substituting the estimated value of N ' and the number of cycles Nc = 20 into Eq.(3) as

N' ^ 20 ; R2o = 0.3-1.25 log (20/V')

N'< 20 R2o = 0.3-0.11 log (20/N') (15)

The equivalent shear stress ratio L , was computed in the same way as was used to derive

( T d/ o mo' )«.<j in the preceding section using the estimated value of N '

.

The relationship between the liquefaction

resistance factor Fui and the excess pore pressure ratio r^ was arranged for every pore pressure data to investigate the effect of the embankment on the relationship.

4. RESULTS AND DISCUSSION

4.1 Time Histories

Fig. 5 and Fig. 6 show time histories of measured

data of the embankment models using an irregular wave of a high frequency and a sinusoidal wave of lOOHz as a shaking acceleration respectively.

Compared witli the rosult of a sinusoidal shaking,

that of an irregular shaking can be characterized

as follows:

(1) An increase of an excess pore pressure (P2, P3 and P6) proceeded stepwise at a moment when

the amplitude of tlie shaking acceleration (AO) was large.

(2) A settlement rate of the embankment (DV) was

not constant, but increased while the large

amplitude of the shaking acceleration was

applied.

(3) There were some cases where a rapid response

of an acceleration in the free field (Al and A2)

occurred after the acceleration had reduced due

to liquefaction.

4 . 2 Equivalent Shak ing Acceleration

Table 2 shows the equivalent shaking

accelerations and the coefficient of equalization

for the first shaking steps of embankment models

using an irregular wave as a shaking

acceleration.

The coefficient of equalization below the embankment were mostly larger than in the free

field, and it sllp.htly increased as the calculation point below the embankment was shallower. These results may be related with the reduction of the cyclic shear stress ratio below

the embankment due to the increase of the effective overburden pressure.

The coefficients of equalization for the cases using an irregular wave of a liigh frequency were

generally larger than those for the cases using

an irregular wave of a low frequency, because the number of waves with a large amplitude was larger

in the former than in the latter.

In the following the equivalent shaking

acceleration at a middle point below the embankment (P5) is used as a representative

value.

4.3 Settlement of Embankment

Fig. 7 fihows the relationship between the settlement of embankment and the shaking

acceleration in the first shaking steps, where

the shaking acceleration was defined to be a maximum amplitude In the irregular shakings and

an average amplitude in the sinusoidal shakings.

The settlement for the same shaking acceleration

was larger for the cases using the sinusoidal

wave than for the cases using the irregular wave. On the contrary the relationship between the settlement and the equivalent shaking

acceleration was almost similar irrespective of the shaking waves as is shown in Fig. 8. This

result indicates the efficiency of the cumulative

damage concept in equalizing the irregular

shaking wave.

As for the effect of the frequency, the shaking

using a sinusoidal wave of 60Hz caused a larger

settlement than the shaking using that of lOOHz . It may be caused by a larger amplitude of a

243
cyclic displacement and a longer duration for the sinusoidal shaking of a lower frequency. A series

of shaking table tests*" conducted by one of the writers showed a similar result. Such effect of the frequency cannot be considered in the process

of the equalization based on the cumulative

damage concept except the frequency-dependent

characteristics of the responce acceleration.

4. A Excess Pore Pressure in Ground

Fig. 9 shows the relationship between the maximum excess pore pressure ratio in the ground

and the equivalent shaking acceleration in the first shaking steps of the embankment models. It was almost similar irrespective of the shaking waves both in the free ground (PI to P3) and below the embankment (P4 to P6) . This result

indicates the cumulative damage concept was effective in equalizing the irregular shaking wave not only for the settlement of the embankment, but also for the excess pore pressure

in the ground.

4.5 Liquefaction Resistance in Free Ground

Table 3 shows the calculated liquefaction

resistance of the horizontal layer models. The estimated N' to locate the liquefaction

resistance curve was almost in the range from 1 to 3, and the dynamic shear resistance ratio Kso

was almost in the range from 0.16 to 0.21. They were not affected by the shaking wave. These

results indicate the efficiency of the cumulative

damage concept in estimating the liquefaction

resistance of the ground.

4.6 Relationship between Liquefaction Resistance Factor and Excess Pore Pressure Ratio

The relationship between the liquefaction

resistance factor Fn and the excess pore pressure ratio ru of the horizontal layer models

is shown in Fig. 10, and that of the embankment

models is shown in Fig. 11. In the calculation the estimated N' was used for the horizontal layer

models, and for the embankment models N' was assumed to be 2 which is a representative value

of horizontal layer models. In the embanlanent models the point P4 just below the embankment was excluded from the data arrangement, because the excess pore pressure at the point was negative

during the shaking as shown in Figs. 5 and 6. The results can be summarized as follows.

(1) In the horizontal layer the excess pore pressure ratio at shallow points PI and P4 was similar to and the one at the other deep

points was similar to Fl""' . The horizontal

location of the points did not affect the relationships (Comparison between PI to P3 and P4 to P6 in Fig. 10).

(2) Below the embankment (P5 and P6 in Fig. 11) the excess pore pressure ratio for the same value

of Fl was smaller than in the free ground (PI to

P3 in Fig. 11).

(3) The effect of the shaking history (Comparison

between Step 1 and Step 2 in Figs. 10 and 11) and that of the shape of the shaking wave on the results of (1) and (2) were small. Concerning (1) and (2), Fig. 12 shows the result

of former study"' based on shaking table tests.

The excess pore pressure ratio at the shallow

points showed a smaller tendency than that at the deep points for the same value of Fl., and the relationship between the liquefaction resistance

factor and the excess pore pressure ratio in the free field was not so different from the one below the embankment as it was in this study. The reason for these disagreements is not known yet, but is a subject of future investigations.

5. CONCLUSIONS

(1) A generation of an excess pore pressure in a sand layer and a settlement of an embankment

proceeded while a large amplitude of an irregular

shaking acceleration was applied.

(2) The cumulative damage concept was effective

In equalizing an irregular shaking wave to evaluate the excess pore pressure in the sand layer and the settlement of the embankment.

(3) The excess pore pressure ratio for the same liquefaction resistance factor was larger in the free field than below the embankment. It was the largest near the surface of the free field.

REFERENCES

1) Koga, Y., Taniguchi, E., Kosekl, J. and Morishita, T : Sand Liquefaction Tests Using a Geotechnical Dynamic Centrifuge, Proc. of the 20th Joint Meeting of U.S. -JAPAN Panel on Wind

and Seismic Effect;), U.S. -JAPAN Conference on Development and Utilization of Natural Resources,

1988.

2) Inatomi, T. , Kazama, K., Noda, S. and Tsuchida, H. : Centrifuge Dynamic Model Tests in

PHRI, Proc. of the 21st Joint Meeting of U.S. -JAPAN Panel on Wind and Seismic Effects, UJNR,

1989.

3) Ledbetter, R. El. : Modelling of Earthquake

Induced Pore Pressure Behavior, Proc. of the 21st Joint Meeting of U.S. -JAPAN Panel on Wind and Seismic Effects, UJNR, 1989.

4) Roga, Y. and Koseki, J. : Dynamic Centrifugal

Model Tests of Embankments on Liquefiable

Grounds, Proc. of the 22nd Joint Meeting of U.S. -JAPAN Panel on Wind and Seismic Effects, UJNR,

1990.

5) Koga, Y., Koseki, J., Shlmazu, T. and Matsuo,

0. : Influence of Embankments on the Liquefaction

of Ground, Proc. of the 23rd Joint Meeting of

U.S. -JAPAN Panel on Wind and Seismic Effects,

UJNR, 1991.

6) Koga, Y. and Matsuo, 0. : Shaking Table Tests

of Embankments Resting on Liquefiable Sandy

Ground, Soils and Foundations, Vol.30, No. 4, pp. 162-174, 1990.

244
Table 1 Test Cases and Conditions

iiUUc J. r* a c o Type of

Jlla^XiXg nfaVC Shaking Acc. (g)"" Dr of Sand Model Ground 1st 2nd 3rd 4 th 5th 6 th Layer (%) 63-1 63-1-1 Embankment Sinusoidal 1.2 63-1-2 60Hz,20cycle 2.5 4.1 5.2 7.3 8.7 12.7 444 63-1-3 3.7 5.6 7.1 8.3 12.5 444 63-1-4 5.5 4*4 63-1-5 7.3 444 63-1-6 9.0 444

1-3 1-3-1 Horizontal Sinusoidal 2.3 3.3 4.5 5.6 64 1-3-2 Layer 60Hz,50cycle 3.4 4.3 5.2 61 1-3-3 4.5 5.6 59

2-1 2-1-1 Embankment Sinusoidal 4.0 5.8 9.4 57 2-1-2 100Hz,20cycle 5.4 8.5 8.6 62 2-1-3 8.9 9.1 60 2-1-4 Embankment Irregular 6.0 8.6 10.8 60 2-1-5 High Frequency 8.3 10.8 61 2-1-6 4.3 6.3 7.7 10.6 63 2-1-7 Embankment Irregular 5.4 6.9 8.4 63 2-1-8

2-1-9 Lov Frequency 7.0

8.7

8.6 65

66 2-2 2-2-1 Horizontal Sinusoidal 3.9 5.0 6.0 66 2-2-2 Layer 100Hz,50cycle 3.4 4.5 5.6 6.6 68 2-2-3 Horizontal Irregular 4.9 6.4 7.1 9.2 66 2-2-4 Layer High Frequency 8.1 8.9 11.1 65 2-2-5 Horizontal Irregular 44 6.8 8.4 63 2-2-6 Layer Low Frequency 6.7 8.4 63

* An average amplitude is shown for sinusoidal shakings, and the maximum amplitude for irregular shakings.

** The shaking acceleration was not known due to mismeasurement.

*** The relative density of the sand layer for model 63-1 was not known due to mismeasurement.

It was assumed to be almost the same as others, because the preparation method was the same.

Table 2 Equivalent Shaking Accelerations and Coefficients of Equalization

Case Shaking Wave

Equiv. Shaking Acc. Ab<j (g) / Coef. of Equalization Coq

Free Ground'' at at P5 at P6 2-1-4

2-1-5

2-1-6

Irregular

High Frequency

3.0 / 0.50

6.8 / 0.82

2.4 / 0.55

3.8 / 0.64

4.7 / 0.57

2.6 / 0.61

3.6 / 0.61

5.6 / 0.68

2.6 / 0.61

3.5 / 0.59

6.2 / 0.75

2.5 / 0.59 2-1-7

2-1-8

2-1-9

Irregular Low Frequency2.5 / 0.47

2.8 / 0.40

3.7 / 0.42

3.2 / 0.59

3.7 / 0.53

4.1 / 0.48

3.0 / 0.56

3.5 / 0.49

3.9 / 0.44

3.0 / 0.55

3.3 / 0.48 3.7 / 0.42

* In the free ground A.,, and Caq were the same irrespective of the depth, because the water level was equal to the surface of the sand layer.

245
Table 3 Liquefaction Resistance In Free Ground

Case Shaking Wave Shaking Point of Liq. Resistance N ' to Locate Liq.

Stepr Calculation Ratio I?2o Resistance Curve

1-3-1 Sinusoidal 1st PI 0.186 1.8 60Hz,50cycle

1-3-2 Sinusoidal 1st PI 0.235 5.1 60Hz,50cycle P2 0. 195 2.2

P4 0.271 11.0

P5 0.26A 9.5

P6 0.258 8.3

2nd P2 0. 166 1.2

P5 0.251 7.1

P6 0.265 9.5 3rd P5 0.261 8.8

P6 0.273 11.3 1-3-3 Sinusoidal 1st P3 0.153 0.92 60Hz,50cycle P6 0.155 0.96

2nd P3 0. 188 1 .9 P6 0. 191 2.0

2-2-1 Sinusoidal 1st PI 0.212 3.2 100Hz,50cycle P2 0.167 1.2

P3 0.204 2.7

2nd PI 0.19A 2.2

P2 0.15A 0.95

P3 0.197 2.3 2-2-2 Sinusoidal 1st PI 0.213 3.3 100Hz,50cycle P2 0.180 1.6

P3 0.187 1.9 2nd PI 0. 199 2.4

P2 0. 178 1.6

P3 0.195 2.2 2-2-4 Irregular 1st PI 0. 174 1.4 High Frequency P2 0. 196 2.3

P3 0.212 3.2

P4 0.214 3.3

P5 0. 193 2.1

P6 0.216 3.5

2nd PI 0.152 0.90

P2 0. 189 2.0

P3 0.202 2.6

PA 0.200 2.5

P5 0.218 3.6

P6 0.233 4.9 2-2-6 Irregular 1st PI 0.200 2.5 Low Frequency P2 0. 147 0.81

P3 0. 176 1.5

PA 0.266 9.9

2nd PI 0.207 2.8

P2 0.168 1.3

P3 0.194 2.2

PA 0.231 4.7

P5 0.235 5.2

P6 0.211 3.1

246
Unit ton

o- Accelerometer

o- Pore pressure transducer

Displacement transducer

R: Rotational Radius

(1) Embankment Model

R=21

Flg.l

(2) Horizontal Layer Model

Test Models and Location of Transducers

(1) Sinusoidal Wave of 60Hz (20cycle)

(2) Sinusoidal Wave of iOOHz (20c7cle)

(3) Irregular wave of High Frequency

(4) Irregular wave of Low Frequency

Fig. 2 Shape of Shaking Waves

Dynamic Centrifugal Model Tests

1000

Tlne(nisec)

Acceleration record

1) Calculation of dynamic

shear stress ratio

2) Assumption of liquefaction

resistance curve

Cumulative Damage Concept

3) Calculation of damage

factor D

4) Calculation of equivalent

shaking acceleration A,,

Fig. 3 Procedure of Equalization

0.7

0.6 05

0.4

0.3 0

0

01 H 0.2

tou

W ft M 0

CO

—' A= 1.25 \

1 1 1 t1 1

\ A-0.3)"A'(logN-logN*)

N*=3.4

A-O.ll /

05 2 5 10 20 50 100 200 500 Number of Cycles N ^ A : Cyclic torsional

shear test results

Toyoura sand, Dr= 55~60%

.Tvo*=I.O, a^o'=0.5(kgf/cra=)

Axial displacement confined

VDA=15%

Fig. 4 Assumed Liquefaction Resistance Curve

247
10

Acceleration

A 0 (g)

Settlement

I) V (ran)

10

5 r

Excess Pore Pressure

P I (kgf/cra=)

0

Excess Pore |Q

Pressure

I' 2 (kgf/cm^)

0

Excess Pore Pressure

P 3 (kgf/cm^)

0

0

.jrrf Excess Pore Pressure

P 4 (kgf/cm^)

10

Excess Pore

^ q

Pressure

P 5 (kgf/cm^)

0

Excess Pore

^ q

Pressure

P 6 (kgf/cm^)

0

10 Acceleration

A (g)

Acceleration

A 2 (g)

Acceleration

A 3 (g)-10

10 Acceleration

A 4 (g) 0

-1010 Acceleration

A 5 (g) "

-lo' ^

10| . . . _

Acceleration ^ ..... ., A 6 (g)

0 —^fVY^MV^"^

-1010 Acceleration

A 7 (g)

-10 500 Time (msec) 1000

100

-10

10

-10

10

-1010

-1010

-10

10

-10

100

-10

AflAVr500

Time (msec) 1000

Fig. 5 Measured Data of Embankment Models

(Irregular Wave of High Frequency)

Fig. 6 Measured Data of Embankment Models

(Sinusoidal Wave of lOOHz)

248
0 5 10 Shaking Acceleration ( k )

Fig. 7 Shaking Acc. vs. Crest Settlement

5 10 Equivalent Shaking

Acceleration Ao<, ( g )

Fig. 8 Equivalent Shaking Acc.

vs. Crest Settlement

° Sinusoidal wave of 60Hz, 20cycle

• Sinusoidal wave of lOOHz, 20cycle A Irregular wave of high frequency

, Irregular wave of low frequency

( !• I )

0 5 10 Equivalent Shaking

Acceleration A«q ( g )

0 5 10 Equivalent Shaking

Acceleration A«q ( g )

0 5 10 Equivalent Shaking

Acceleration Ao<, ( g )

(1) at PI " (2) at P2 (3) at P3

Fig. 9 Equivalent Shaking Acc. vs. Excess Pore Press. Ratio (Embankment Models)

249
2.0

3 1.5

PS

0)

l-i 3

1.0

O

v>S 05

0.0

o sinusoidal wave of 60Hz, 20cycle • Sinusoidal wave of lOOHz, ZOcycle A Irregular wave of high

frequency

Irregular wave of low frequency

0 5 10 Equivalent Shaking

Acceleration Aa<, ( g )

5 10 Equivalent Shaking

Acceleration Ao<, ( k )

5 10 Equivalent Shaking

Acceleration A»<, ( g )

4.0

J

S 3.0 o

oO

nl 2.0

tn paa

o

•H

4Jo

rt

O)cr J

1.0

(4) at P4 (5) at P5 (6) at P6

Fig. 9 Equivalent Shaking Acc. vs. Excess Pore Press. Ratio (Embankment Models) (continued)

Sinusoidal wave

of lOOHz

\

o

O P 1 A P 2 P 3

0 02 0.4 06 08 1.0 Excess Pore Pressure Ratio r „

4.a

o

rt

<ao

O

3.0

Sinusoidal wave

of lOOHz

2.0

1.0

0

I <^ \ o

(STEP n (STKPS) O p I • r 1 A P 2 A !• 2 P 3 P 3

: r i, = I'l :r„ - K,

I I u 0 02 04 06 08 1.0 Excess Pore Pressure Ratio r u

4.0

J

3.0

2.0

a

o

•Hu

V

1.0 3

C7*

j Irregular wave

j of high

l o frequency

\

k - \

'v - >^^^^

o p I • P * A P 2 ^ P 5 P 3 P B

: - r„ - P^-^

0 02 0.4 06 0.8 1.0 Excess Pore Pressure Ratio r u

(1) Case 2-2-1, Stepl (2) Case 2-2-2, Stepl and Step2 (3) Case 2-2-4, Stepl

Fig. 10 Liq. Resistance Factor vs. Excess Pore Press. Ratio (Horizontal Layer Models)

250
4.0

J

30

2.0

3

I Irregular wave

1 of low frequency

1^

O P I A P 2 DPS

: r t

0 0.2 0.4 0.6 0.8 1.0 Excess Pore Pressure

Ratio r u

4.0

J

3.0,

2.0

Sinusoidal wave

of lOOHz

o P 1 A P 2 A p 5 P 3 Bps

CC CO C o am

0 0.2 0.4 06 0.8 10 Excess Pore Pressure Ratio r u

4.0

J

Cl.

3.0

hi 2.0

1.0

0)

0

Irregular wave

of high frequency

O P I A P 2 APS

P 3 I' 6

0 02 04 0.6 08 10 Excess Pore Pressure

Ratio r u

(4) Case 2-2-6, Stepl (1) Case 2-1-2, Stepl (2) Case 2-1-4, Stepl

Fig. 10 Liq. Resistance Factor vs. Excess Pore Press. Ratio

(Horizontal Layer Models)

(continued)

4 0r-^

J

tJ 30

2.0

o^ 10

Irregular wave o of high

frequency

A

t

A

CP

o o

CD.-U-S- A A

"O—

O P I A P 2 APS

P 3 P (5

0 02 0.4 06 0.8 1.0 Excess Pore Pressure Ratio r u

4.0

J

3.0

2.0

<u3

1 -T~ A Irregular wave

of low frequency °o

A

o

- A

A ^ A A^ !3 no p

A

O P 1 A P 2 A P6 a P 3 P 6

1 . 1 0 0.2 04 0.6 0.8 1.0 Excess Pore Pressure

Ratio r ^

(3) Case 2-1-4, Step2 (4) Case 2-1-9, Stepl

Fig. 11 Liq. Resistance Factor vs. (Embankment Models) Excess Pore Press. Ratio

251
« 0 0 •P21 •P22* 'PlW •P2G •P27 +

0 •P12

O

0 0>15*

e •P18

O o

•P3 •P6 • •P8 •P9

• Accelerometer + Shear strain gauge

• P.W.R transducer { Displacement meter

(unit In cm)

Case Sinusoidal wave of ZOcycle

Frequency Acceleration

1 2.5Hz 190 gal

2 lO.OHz 315 gal

3 5.0Hz 208 gal

(1) Analized Model and Location of Transducers

3o

a. I Vu

o

a,

on • Pt o »>i» Pie »p» » pjj

0)u

3

(0

vo

op J 'P < . OPIJ "Plj

»P1I «PJJ os

3

0_ 05

O

a.

t1

• p«

P15 APU

> \

i:

*\

I--—— A J

Liquefaction Resistance Factor Fu Liquefaction Resistance Factor Fu

(2) Case 1 Free Field (3) Case 1 Below Crest

Liquefaction Resistance Factor Fu

Ci) Case 1 Below Slope Toe

4)U

o

a.

eP8 •P9

0P17 BptS

a P26 * Pn ^—

j

1

Fu-' '

o^.

4)

t.3

ca

cnv

Q-C

0)

Lio

a.

OPJ

opll "PIS

'PJl »P2J

iXa

\^

15 3o

<sV

10 u3

U]n

lUu

a. 03 a

t.o

Cu

• p«

»p*

A

Fl-"" A \.

V.

Liquefaction Resistance Factor Ft, Liquefaction Resistance Factor Ft

(5) Case 2 Free Field

, »'r

(6) Case 2 Below Crest

o

I.Vo

ou

n •

opj •PS

OP17 mtnt *PW APJ7

?

P.

Fc-'

V^^-:>.

•-.1 .

o

uuu

CLQJ

0)U

O

Ou

opj cpt

oPtl «PI3

APJl »PJI

! L^A

•<

*

Liquefaction Resistance Factor Fl

(7) Case 2 Below Slope Toe

. ts

o

1 0)

(>3

la

ta

0)L. QuOS V

I.o

cu

• A F

• pe

PIS aPK

Liquefaction Resistance Factor Ft, Liquefaction Resistance Factor Ft,

(8) Case 3 Free Field (9) Case 3 Below Crest

Liquefaction Resistance Factor Ft,

(10) Case 3 Below Slope Toe

Fig. 12 Liq. Resistance Factor vs. Excess Pore Press. Ratio (Shaking Table Test*")

252
Earthquake-Induced Landslides in the Santa Cruz Mountains

by

M. E. Hynes* and M. Hudson**

ABSTRACT

The magnitude 7.1 Loma Prieta earthquake which

occurred at 5:04 p.m on October 17, 1989,

ruptured a 25-inile long segment of the San

Andreas fault in the Santa Cruz Mountains,

California, and caused numerous landslides in

the epicentral vicinity. A detailed study of

the ground failure in this area is documented

by Keefer et al. 1991. This paper summarizes

slope stability and deformation analyses of

damaged slopes in the Upper Schultheis Road

and Villa Del Monte areas that were performed

to assess the potential for future movement

caused by earthquakes, intensive rainfall, or

a combination of the two. The Loma Prieta

earthquake and resulting slope movements

provided a full scale field test of the

strength of the materials in the slope. The

stability and deformation analyses were used

to back-calculate the mobilized strength of

the in situ materials during the earthquake

from the observed displacements and estimated

ground motions, sliding surface locations, and

groundwater conditions. This stability model

was then adjusted to simulate intense rainfall

conditions with higher groundwater levels, and

used to assess static (no earthquake) slope

stability for these same sliding surfaces.

The resistance to earthquake- induced sliding

deformations on postulated surfaces by future

seismic events was determined for low

groundwater and high groundwater conditions to

simulate the relatively dry conditions that

existed during the Loma Prieta event and the

relatively wet conditions that would result

from intense rainfall. It was concluded that

(1) the slopes would become marginally stable

when the water levels rise to within about 10 ft of the ground surface, and (2) if the Loma

Prieta earthquake had occurred during a period

of elevated groundwater level, then the

maximum landslide displacements could have

been three times as large or more; i.e., on

the order of at least 25 ft rather than 8 ft.

1. INTRODUCTION

The 1989 Loma Prieta earthquake which occurred

at 5:04 p.m on October 17, 1989, ruptured a 25-mile long segment of the San Andreas fault

in the Santa Cruz Mountains, California, and

caused numerous landslides in the epicentral

vicinity. A detailed study of the ground

failure in this area is documented by Keefer

et al. 1991. This paper summarizes the

engineering analyses of two landslide areas in

the Santa Cruz Mountains, the Upper Schultheis

Road landslide and the Villa Del Monte

landslide complex, both of which are located

in the epicentral region, about 1000 ft from

the trace of the San Andreas fault. Location

maps are shown in Figures 1 and 2, and mapped

observations of surface cracking are shown in

Figures 3 and 4. The geologic materials in

the slide areas consist of colluvium and

regolith on upturned and folded interbedded

shale, sandstone and siltstone that have been

complexly sheared and are variably weathered.

The Upper Schultheis Road landslide and the

Villa Del Monte landslide complex were modeled

in slope stability and deformation analyses

first to backcalculate static and dynamic

strengths of the slope materials. These

backcalculated strengths were then used in the

slope stability and deformation analyses to estimate the stability of the slopes under

higher groundwater conditions and the

displacements that could occur in future

earthquakes. In this case history,

backcalculated strengths are preferred to laboratory test results because they represent

the strength of in situ materials averaged

over the entire failure surface, which passes

through several types of materials, at fullscale stress conditions. Laboratory tests were performed on selected samples, but these

test results represent only point estimates of

strength of individual materials in this

complex geologic setting.

The slope stability and deformation analyses

were calibrated using observed past

performance of the slopes. The past

performance of the slopes against which the

models were calibrated are: Condition (a) : The slopes are generally

stable under deep groundwater conditions and

static loading as observed during periods of

drought and low rainfall before and after the

1989 Loma Prieta earthquake.

Condition (b) : Deep basal sliding surfaces of

large areal extent are stable to marginally

stable under elevated groundwater conditions

and static loading, since areally extensive

sliding of this type has not been observed

during past periods of high rainfall.

Condition (c) : Shallow sliding surfaces of

limited areal extent are marginally stable to unstable under elevated groundwater conditions

and static loading, as observed from localized

slope movements experienced in the slopes

during past periods of high rainfall.

Condition (d) : The severe earthquake ground

shaking caused by the 1989 Loma Prieta

earthquake triggered instability of deep basal

sliding surfaces of large areal extent and

resulted in downslope displacements as large

as approximately 8 ft. This occurred during a period of drought when the groundwater levels

were relatively deep.

The models were developed to reflect condition

(a) , and then slope stability for conditions

(b) and (c) were computed. Thus, it was concluded that the models provided reasonable

estimates of averaged static strengths of the

slope materials.

* U.S. Army Engineer Waterways Experiment

Station, Vicksburg, Mississippi

** Department of Civil Engineering, University

of California, Davis, California

253
The slope stability and deformation models

were then adjusted to represent the dynamic

loading case in condition (d) . The computed

dynamic strengths from this case were compared

to the estimated static strengths, and it was

found that the dynamic strengths reflected

typical levels of strength degradation

associated with dynamic loading and

displacements on the order of several feet

(Idriss 1985, Seed 1987). Thus it appears

that the analytical models provided realistic

estimates of both static and dynamic strengths

of the slope materials.

The backcalculated static and dynamic

strengths were then used to develop charts and

procedures from which slope displacements

caused by future earthquakes and groundwater

conditions could be estimated. For example,

the analyses indicated that had the

groundwater levels been significantly higher

when the 1989 Loma prieta earthquake occurred,

then the maximum displacements could have been

three times as large or more; i.e., on the

order of at least 25 ft rather than 8 ft. This paper describes the slope stability and

displacement models that were developed, and

presents the resulting charts and procedures

for use in analysis of slope stability for a range of earthquake and groundwater

conditions

.

2. STABILITY AND DEFORMATION APPROACH

In the Newmark sliding-block deformation

analyses adopted for this study, the part of a slope displaced by earthquake shaking was

idealized as a rigid block sliding on an

inclined plane, as originally proposed by

Newmark (1965). Since 1965, a number of

researchers (Goodman and Seed 1966, Ambraseys

and Sarma 1967, Sarma 1975, Franklin and Chang

1977, Hynes 1979, Makdisi and Seed 1979, Sarma

1979, Wilson and Keefer 1983, Hynes and

Franklin 1984, Idriss 1985, Wilson and Keefer

1985) have developed Newmark's conceptual

model into a widely used, practical procedure

for estimating permanent, earthquake-induced

deformation of slopes.

A conventional, limit-equilibrium slope

stability analysis with slight modifications

was used to determine the shearing resistance

along a failure surface. Because the base

motions may be amplified upon being propagated

upward through a slope, consideration of the

dynamic response of the slope was incorporated

to account for amplified accelerations in the

slope. The shearing resistance determined in

the limit-equilibrium slope stability analysis

was then used to determine the relative

displacement that will accumulate along a slide surface for a representative group of

earthquake ground-motion histories. Each of

these steps is described in greater detail

below.

3. LIMIT-EQUILIBRIUM SLOPE STABILITY ANALYSES

3 . 1 Method used

A two-dimensional, force-equilibrium solution

method using Lowe and Karafiath's side force

assumption was used for the slope stability

calculations (Lowe and Karafiath 1960) . The

programs used were UTEXAS2 (Edris and Wright

1987) and UTEXAS3 (Wright 1990). In this

method, a pseudo-static horizontal

acceleration, kj^, expressed as a fraction of

gravity, can be applied to the center of

gravity of the sliding mass. The value of kj, that causes the factor of safety against

sliding of the slope to equal unity is termed

the yield acceleration, ky. A parametric

study was performed to determine values of k^

and static (kj^ = 0) safety factors against

sliding, FSg, for a range of shear strength

values and groundwater conditions. A unit

weight of 134.5 pcf was used as an average

estimate for all materials in the slope based

on laboratory test results.

3.2 Geometry of Failure Masses

An analysis of the field evidence used to estimate the geometry of the failure masses

and locations of the shear surfaces is

described by Keefer et al. (1991). Several

sliding surface locations were assumed in the

slope stability analyses ranging from deep

surfaces (greater than approximately 100 ft) to fairly shallow surfaces (approximately 50

ft or less) . This range provided (1) insight

about the controlling mechanism of failure

acting in the slope during sliding caused by

the 1989 Loma Prieta earthquake, (2) an indication of the sensitivity of the results

to possible slip surface locations, and (3) information about potential slope stability

during future rainfall and earthqueOce events.

A single cross section, denoted SR, was

selected through the Schultheis Road landslide

area, as shown in Figures 5 (plan) and 6 (section) . Two cross sections, denoted DM and

SB, were chosen through the Villa Del Monte

landslide complex, as shown In Figures 7 (plan) and 8-9 (sections) . Estimated deepseated basal surface locations are shown in

Figures 6, 8 and 9 for sections SR, DM and SB,

respectively. Additional assumed shear

surfaces are shown in Figures 10-12 for

sections SR, DM and SB, respectively.

3 . 3 Groundwater Levels

A range of water-level conditions were assumed

in the analyses to simulate the range from the

1989-1990 drought conditions to high

groundwater levels, such as may be caused by

intense or long-duration rainfall. Monitoring

of groundwater levels is discussed In greater

detail by Keefer et al. (1991). A key step in

the stability analyses is to estimate the pore

pressures that existed in the slope just prior

to the earthquake; it is estimated that the

piezometric levels measured during the spring

of 1990 were indicative of those that existed

just prior to the earthquake. Since no significant rainfall occurred in the area for

several months prior to the earthc[uake, and

the winter and spring months of 1990 had a similar rainfall environment, it is estimated

that transient effects related to the

earthquake had passed and the groundwater

regime had again reached a steady-state

condition similar to that before the

earthquake. Estimated groundwater levels

along the three analyzed sections are shown in

Figures 6, 8 and 9. Due to the complex

stratigraphy, it was expected that the in situ

groundwater pore pressure fields are also

complex. However, for the purposes of

analysis, hydrostatic pore pressure fields

were used as an approximation of the in situ

conditions.

In the Upper Schultheis Road landslide. Figure

6, a hydrostatic groundwater level at a depth

of 50 ft was used to represent pre-earthquake

pore pressure conditions. Analyses were

254
conducted for hydrostatic groundwater levels

ranging from 0 to 50 ft below the ground

surface to represent the effects of low to high rainfall.

In the Villa Del Monte area, Figures 8-9, the

pre-earthguake groundwater level was assumed

to be at a depth of 100 ft, and analyses were

conducted for hydrostatic groundwater levels

ranging from 0 to 200 ft below the ground

surface for sensitivity analyses and to represent groundwater conditions for low to high rainfall.

3 . 4 Shear Strength

Friction angles represented the shear strength

on the stability computations. Friction

angles ranging from lOh to A0\ were used in

the parametric studies.

3.5 Parametric Results from Slope Stability

Calculations

The sliding masses bounded by deep-seated

basal shear surfaces shown in Figures 6, 8 and

9 were analyzed in a detailed pareimetric study

for a wide range of shear strength values and

groundwater levels. The resulting relations

determined between ky, mobilized friction

angle and groundwater level are shown in

Figures 13-18. Minimum values of friction

angle required for static stability for the

range of groundwater levels correspond to ky -

0 in these plots and range from ll^i to 2A\.

These figures were used to estimate the in

situ strength mobilized during earthquakeinduced sliding and critical levels of

groundwater increase, as described later in

this paper.

4. ESTIMATED GROUND MOTION AND

SLOPE AMPLIFICATION

The next step in the Newmark analysis was to

estimate the ground motions (accelerations)

that were generated at the base and crest of

the mountain by the Loma Prieta event,

particularly the peak horizontal ground

accelerations. No strong-motion instruments

were located in the immediate area. Two

California strong-motion instrument stations

were close to the epicenter of the earthquake-

-Corralitos, to the south, at an epicentral

distance of approximately 10 km, and Lexington

Dam, to the north, at an epicentral distance

of approximately 18 km (Shakal and others

1989). For the Loma Prieta event, the peak

horizontal ground accelerations at Corralitos

were 0.48 and 0.63 g; at Lexington Dam (left

abutment) , the peak horizontal ground

accelerations were 0.41 and 0.44 g. The Upper Schultheis Road landslide and Villa

Del Monte landslide complex are located

approximately 10 km north of the earthquake

epicenter. The Krinitzsky, Chang and Nuttli

(1988) attenuation chart shown in Figure 19 is

based on a large body of California data, and

indicates a peak horizontal ground

acceleration of about 0.63 g would be

estimated for a magnitude 7 earthquake

occurring at a distance of 10 km; about 0.40 g would be estimated at a distance of 18 km. Since the recorded peak acceleration (maximum

of two horizontal records) at Corralitos

located at the same distance as the slide

areas and the predicted value from a large

body of California data both result in a peak

acceleration of 0.63 g, it was concluded that

the observed values at Corralitos of 0.48 and

0.63 g were realistic estimates of the peak

motions at a rock outcrop in the study area. The possible amplification of the base motions

as they propagated upward through the slope to the ridge crest was estimated from

consideration of typical values from

analytical studies and field obsezrvations of

natural slopes and embankments, and from a simplified shear-beam numerical model

described by Makdisi and Seed (1979)

Typically, analytical studies and field

observations indicate that embankments amplify

base (toe-of-slope) accelerations, a, by

factors ranging from approximately 1 to 4 to obtain peak crest accelerations, u. In special cases, amplification factors greater

than 4 have been observed in the field (based on data from Hynes and Franklin 1984, and the

Corps of Engineers Strong-Motion network) . At high levels of acceleration, embankment

amplification factors tend to approach unity,

as shown in Figure 20 (from Harder 1991)

Simplified shear-beam models tend to yield

fairly high amplification factors; values of 4 to 5 were computed in this case. A range of

amplification values was used in the

sensitivity analysis shown in Table 1 to estimate the pseudo-static acceleration, kjjg„, most appropriate to represent the earthquake

motions in the deformation calculations.

The yield acceleration, ky, from the slope

stability analyses is compared to the mass

averaged acceleration, kj^^jj, induced by the

earthquake in the sliding mass. Both shearbeam and finite-element numerical analyses

have been used in earlier studies (Sarma 1979,

Makdisi and Seed 1979, Hynes and Franklin

1984) to develop generalized charts to compute

representative values of k^^ax for a range of

sliding mass shapes, slope stiffnesses, and

geometries (which affect amplification) , and

incoming earthquake ground motions.

The value of kj^a^ should range from about 0.75

to 1.5 times tne base motion on the basis of numerous dynamic finite element analyses of

slopes (Idriss 1968, Idriss 1991). The chart

for estimating kj^a^ embankment and slope

crest peak accelerations developed by Makdisi

and Seed (1979) is shown in Figure 21. This

chart indicates that deep-seated slides

typically have ratios of k^ax ^° crest

acceleration of about 0.20 to 0.45, depending

on the amount of amplification of the

accelerations in the mountain. Based on the

sensitivity analysis shown in Table 1, values

of kjjiajj of 0.5 to 0.6 seem to best represent

the dynamic load for the 1989 Loma Prieta

event. These values were used in the

deformation calculations described next.

5. DEFORMATION CALCULATIONS, ESTIMATED FIELD

STRENGTHS, AND CALCULATION OF STATIC STABILITY

Charts of the relative displacement, u, that

can accumulate between a sliding block and an inclined plane for a range of ratios of

ky/kj^ax have been developed by Franklin and

Chang (1977), Seed and Makdisi (1979), and

Hynes and Franklin (1984), based on many

earthquake accelerograms. The Seed and

Makdisi (1979) curves were selected for this

study. Generalized displacement curves from

Seed and Makdisi (1979) for earthquake

magnitudes of 6.5 and 7.5 were averaged to estimate a curve for magnitude 7 earthquakes.

This estimated curve used to backcalculate

mobilized strength is shown in Figure 22.

255
The deformations observed in the field were

analyzed to estimate the displacement of the

center of gravity (CG) of the sliding mass for

the various failure surfaces. Both the upper

bound and average displacements were used,

determined as a sum of observed displacements

along each cross section analyzed. The

displacements are listed in Table 2 (from

Keefer et al. 1991). The total displacement

is approximately the cumulative value of

downslope displacements observed along the

ground surface along and near the limits of a section. Because the sliding mass is

compressed during movement, it was estimated

that the CG displacement might best be

associated with about half of the total

displacement (Keefer et al. 1991). Thus, half

of the total displacement was defined as the

average displacement and represented the best

estimate for the backcalculation of strength.

The estimated CG displacements were analyzed

with the sliding block model to determine the

mobilized shear strength in the field and to

estimate the amount of groundwater rise that

would lead to static slope instability. The

computations are shown in Tables 3 and 4 for

upper and lower deep-seated basal shear

surfaces identified in Figures 6, 8 and 9. Backcalculated dynamic friction angles for the

additional assumed surfaces shown in Figures

10-12 are listed in Tables 5-6.

The computed mobilized friction angles for

deep basal sliding surfaces ranged from

approximately 16 to 19 for the Upper

Schultheis Road landslide and from about 18 to 23 for the Villa Del Monte landslide

complex. These backcalculated strengths are

from dynamic loading conditions and are

typically lower than the strength actually

available for static loading conditions. The

calculations shown in the last three columns

of Table 4 were performed to obtain

conservative estimates of the stability of the

slopes and the critical rises in groundwater

levels for triggering slope instability under

static conditions. These conservative results

indicated that a groundwater rise of 15 to 30

ft, from a depth of 50 ft to 20 to 35 ft below

the ground surface in the Upper Schultheis

Road landslide would result in renewed

landslide movement without earthquake shaking.

At villa Del Monte, the range of computed

critical groundwater levels was broader; the

results suggested that some analyzed sliding

masses would become statically unstcible with a groundwater level increase of as little as 30

ft, while others would remain stable with

rises of more than 100 ft (which would require

artesian conditions) . Computed static factors

of safety using dynamic strengths, with

estimated pre-earthquake groundwater levels,

were in the range of 1.07 to 1.16 for the

Upper Schultheis Road landslide and 1.15 to

1.62 for the Villa Del Monte landslide

complex. These results are summarized in

Table 4. These safety factors for the low

1989-1990 groundwater levels are all quite

low, but greater than unity, even when the low

dynamic strengths are used in the

computations

.

The strengths and static stability conditions

discussed above are backcalculated from a dynamic loading event. Usually, there is some degradation of strength caused by dynamic

loading of geologic materials, especially if

they are below the water table. To estimate

the amount of strength degradation, the

additional assumed sliding masses shown in

Figures 10-12 were analyzed. The surfaces

were chosen based upon estimated critical

locations of slope geometry. Two cases were

analyzed to determine (1) the friction angle

that would have been mobilized during the Loma

Prieta event, and (2) the friction angle

required to have a static factor of safety

against sliding of unity if the water table is near the surface of the slope, at a depth of

approximately 10 ft. These results are listed

in Table 5 for shallow (approximately 50 ft or

less) surfaces and Table 6 for deep

(approximately 100 ft or more) surfaces.

Localized movements have been observed in

these slopes after periods of heavy rains.

The strengths of the materials in the slope were probably at least as high as the values

calculated for static conditions with the water table near the surface and the factor of

safety against sliding equal to one. If this

were not the case, large failures would have

been expected to be seen as the water table

rose at times in the past. Instead, localized

slope movements have been observed during wet

weather, indicating a factor of safety near

unity. The strength values in Tables 5 and 6 indicate that the friction angles for the

static high-water case are larger than those

necessary to produce the observed

displacements during the 1989 Loma Prieta

event. Therefore, in order to satisfy both

conditions, the strength of the slope

materials must have degraded due to the

shaking of the earthquake.

Degradation of strength is most likely to occur in saturated conditions. The materials

along shallow surfaces were not saturated at

the time of the earthquake. All of the deep

failure surfaces had some of their lengths at

or below the water table. Therefore, it is most likely that materials along deep surfaces

experienced strength degradation. It was

therefore concluded that deep-seated sliding

surfaces were more likely to have been

responsible for the landslide activity in the

Upper Schultheis Road and Villa Del Monte

areas, with a reduction in strength occurring

in the materials along the deep surfaces due

to earthquake shaking.

Another approach was used to estimate the

strength of the materials in the slope under

static loading conditions. It is known that

the slopes were stable before the 1989 Loma

Prieta earthquake, with the groundwater teible at a depth of about 50 ft in the Upper

Schultheis Road area and about 100 ft in the

Villa Del Monte area, A factor of safety

against sliding of 1,5 was used as a reasonable estimate for the static factor of

safety of the slopes for these groundwater

conditions. A series of slope stability

calculations was performed using these

additional slip surfaces to determine the

strengths required to have a static factor of

safety of 1.5. These strengths were then used

to compute the static factor of safety with

the water table near the surface. The results

of these calculations are shown in Table 7, As can be seen from this table, the friction

angles thus generated seem reasonable in that

they provide a low factor of safety against

sliding after the water table has risen

significantly. This correlates well with the

observations of localized movements during

periods of high water table.

In summary, the strengths backcalculated from

the Loma Prieta 1989 event probably

256
underestimate the available strength of the

materials in the slope for the static case of

a rise in groundwater levels because of

strength degradation of saturated materials

due to dynamic loading. The static strengths

are likely to be greater than the

backcalculated dynamic values and

approximately equal to or less than the static

values listed in Tables 5-7. Also, if

groundwater levels rise, slope movements may

involve both shallow and deep surfaces.

6. SLIDING DURING FUTURE EARTHQUAKES WITH

HIGHER GROUNDWATER LEVELS

The Newmark sliding block calculation

procedure demonstrated in Teible 3 can be used

in conjunction with the yield accelerations in

Figures 13-18 to estimate the potential for

further sliding-block displacement in future

earthquakes with higher groundwater levels.

For example, displacement computations that

simulate occurrence of the Loma Prieta-type

event with higher groundwater levels indicate

that displacements on a deep seated shear

surface along cross section SB, in the Villa

Del Monte landslide complex, might have been

two to three times greater than they actually

were if the groundwater level had been at a depth of 50 ft rather than 100 ft during the

earthquake. This is determined by observing

in the last column in Table 3 that the average

mobilized friction angle for the slip surface

denoted SB-U was approximately 21 . One then

enters Figure 17 at a friction angle of

approximately 21 to find the value of ky

corresponding to a water table depth of 50 ft,

by intersecting this contour. The

corresponding value of ky is approximately

0.03. Then, one can compute the ratio ky/^ax

using a value of kjjgj, equal to 0.5 Tor a nearby magnitude 7 event. The resulting ratio

is 0.03/0.5 <= 0.06. One then enters the

displacement chart in Figure 22 at a ratio

value of 0.06. For a magnitude 7 event, this

yields a displacement of about 43 inches (110

cm) . This displacement value is twice as

large as the estimated average observed value

of 22 inches (55 cm) during the Loma Prieta

event when the water table was at a depth of

approximately 100 ft.

If an earthquake occurs when the water table

are higher in the slopes, more materials will

be saturated and likely to suffer some

degradation of strength due to earthquake

shaking; consequently, for these conditions,

slope instability may not be limited to areally extensive, deep-seated basal shear

surfaces, but may also include shallower

surfaces and deep surfaces with more limited

lateral extents. The occurrence of an earthquake when the water

table is near the surface of the slopes could

result in large displacements if the epicenter

is close to the landslide areas. The meaning

of "close" varies with the magnitude of the

earthquake event and the depth to the water

table at the time of the event. For the

relatively deep water tables that existed

during the Loma Prieta event, a magnitude 7 epicenter would need to be as close as approximately 10 miles (16 km) to cause

calculated slide displacement on Villa Del Monte slopes of more than a foot; for the

Upper Schultheis Road area, the magnitude 7 epicenter could be as far away as about 25

miles (40 km) and still cause this magnitude

of displacement. It should be noted that the

southern end of the Hayward-Calaveras fault is

approximately 25 miles away from the area

involved in this study.

7. SUMMARY AND CONCLUSIONS

Slope stability and deformation analyses of

the Upper Schultheis Road landslide and Villa

Del Monte landslide complex were performed to backcalculate the strengths mobilized in the

slopes during the 1989 Loma Prieta earthquake,

and to use this and other field performance

infoirmation to assess the potential for future

movement caused by earthquakes, rainfall, or a combination of the two. Both shallow and deep

block sliding were analyzed using a limitequilibrium pseudo-static slope stability

procedure (Lowe and Karafiath 1960) and a Newmark sliding-block deformation analysis

(Newmarkl965) . These analyses indicated the

following:

(a) It is likely that the saturated slope

materials suffered some degradation of shear

strength due to earthquake shaking. Strengths

backcalculated from observations of the Loma

Prieta earthquake can be used to assess future

earthquake stability, but are likely to underestimate the available strength for

static loading conditions.

(b) The analyses confirm that deep-seated

basal shear surfaces were the likely

controlling failure mechanism for these

landslide areas in the Loma Prieta event,

since these deeper materials were saturated

and subject to earthquake-induced strength

degradation. It is unlikely that the

controlling slope failure mechanism for these

landslides was shallow slip surface movement

since these materials were well above the

water table and would not be likely to have

suffered earthquake-Induced strength

degradation.

(c) If there is significant rise in

groundwater levels, slope failure mechanisms

may include shallower and deeper slip surfaces

of limited areal extent. Large slope failures

have not been observed in this area during

past heavy rainfall events, but rather have

involved small movements on sliding masses of

limited extent. This past field performance

of the slopes and the analyses indicate that

the factor of safety against sliding for such

surfaces is close to unity when the water

table is near the ground surface.

(d) If there is a significant rise in the water table and another earthquake shakes the

slopes, the slope failure mechanisms may

include deep basal shear surfaces of large

areal extent as well as shallow and deep

surfaces of limited areal extent. This is

because the shallower slope materials would

then be saturated and subject to earthquakeinduced strength degradation.

(e) If the Loma Prieta event had occurred when

the water table was higher, maximum slope

displacements might have been on the order of

25 ft rather than 8 ft. This level of

displacement was indicated for deep basal

shear surfaces as well as the shallow and deep

surfaces of limited areal extent for a nearby

magnitude 7 seismic event. Consequently, the analyses Indicated that

deep-seated basal shear slide masses could

become marginally stable when the groundwater

rises to within about 20 ft of the ground

surface at the Upper Schultheis Road landslide

257
area and to within about 30 ft of the ground

surface at the Villa Del Monte landslide

complex. If the groundwater rises to within

about 10 ft of the ground surface, the slopes

could become marginally stable against deepseated, large scale, basal shear sliding; the

computed safety factors for this case ranged

from about 1 to 1.2. The analyses also

indicated that shallow and deep surfaces of

more limited areal extent may possibly become

unstable to marginally stable if the

groundwater rises to within 10 ft of the

ground surface; the computed safety factors in

this case ranged from about 0.8 to 1.1.

However, if an earthquake occurs nearby when

the water table is high, there may be

extensive damage.

8 . REFERENCES

Ambraseys, N. N. 1960. "The Seismic Stability

of Earth Dams," Proceedings of the Second

World Conference on Earthquake Engineering,

Japan, Vol. II. Ambraseys, N. N., and Sarma, S. K. 1967. "The

Response of Earth Dams to Strong Earthquakes,"

Geotechnicme , Vol 17, No. 2, pp 181-213.

Edris, E. V., and Wright, S. G. 1987. "User's

Guide: UTEXAS2 Slope Stability Package, Vol

1," Instruction Report GL-87-1, U. S. Army

Engineer Waterways Experiment Station,

Vicksburg, Miss.

Franklin, A. G., and Chang, F. K. , 1977.

"Earthquake Resistance of Earth and Rockfill

Dams; Permanent Displacements of Earth

Embankments by Newmark Sliding Block

Analysis," Miscellaneous Paper S-71-17, Report

5, U. S. Army Engineer Waterways Experiment

Station, CE, Vicksburg, Miss.

Goodman, R. E., and Seed, H. B. 1966.

"Earthquake-Induced Displacements in Sand

Embankment," Journal of the Soil Mechanics and

Foundations Division . American Society of

Civil Engineers, Vol 92, No. SM2, pp 125-146.

Hynes, M. E. 1979. "Dynamic Analysis of Earth

Embankments for the Richard B. Russell Dam and

Lake Project," Final Report Prepared for U. S. Army Engineer District, Savannah, Georgia.

Hynes, M. E., and Franklin, A. G. 1984.

"Rationalizing the Seismic Coefficient

Method," Miscellaneous Paper GL-84-13, U. S. Army Engineer Waterways Experiment Station,

Vicksburg, Miss.

Keefer, D. et al. 1991, "Geologic Hazards in

the Summit Ridge Area of the Santa Cruz

Mountains, Santa Cruz County, California,

Evaluated in Response to the October 17, 1989, Loma Prieta Earthquake: Report of the

Technical Advisory Group," U.S. Geological

Survey Open-File Report 91-618, Menlo Park,

California.

Krinitzsky, E. L. , Chang, F. K. , and Nuttli,

0. W. 1988. "Magnitude-Related Earthquake

Ground Motions," Bulletin of the Association

of Engineering Geologists . Vol XXV, No. 4, pp

399-423

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Makdisi, F. I., and Seed, H. B. 1979.

"Simplified Procedure for Estimating Dam and

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Journal of the Geotechnical Engineering

Division . American Society of Civil Engineers,

Vol 104, No. GT7, pp 849-867.

Martin, G. R. 1965. "Response of Earth Dams

to Earthquakes," Thesis submitted in partial

satisfaction of the requirements for the

degree of Doctor of Philosophy, University of

California, Berkeley.

Newmark, N. M. 1965. "Effects of Earthquakes

on Dams and Emiiankments, " Geotechnique , Vol

15, No. 2, pp 139-160.

Sarma, S. K. 1975. "Seismic Stability of

Earth Dams and Embankments," Geotechnique. Vol

25, No. 4, pp 743-761.

. 1979. "Response and Stability of

Earth Dams During Strong Earthquakes,"

Miscellaneous Paper GL-79-13, U.S. Army

Engineer Waterways Experiment Station, CE,

Vicksburg, Miss.

Schnabel, P. B. , Lysmer, J., Seed, H. B. 1972.

"SHAKE, A Computer Program for Earthquake

Response Analysis of Horizontally Layered

Sites," Earthquake Engineering Research Center

Report No. EERC 72-12, University of

California, Berkeley.

William Cotton and Associates 1990. "Villa

Del Monte and Schultheis Road Landslides

Geotechnical Exploration, Santa Cruz County,

California," Draft Contract Report for U. S, Army Engineer District, San Francisco, June

1990.

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"Schultheis Road and Villa Del Monte Areas

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California," Final Contract Report for U. S. Army Engineer District, San Francisco,

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Wilson, R. C. , and Keefer, D. K. 1983.

"Dynamic Analysis of a Slope Failure from the

6 August 1979 Coyote Lake, California,

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Wilson, R. c, and Keefer, D. K. 1985.

"Predicting Areal Limits of Earthquake-Induced

Landsliding," Earthguake Hazards in the Los Angeles Region . Ziony, J. I., ed. , USGS

Professional Paper 1360, pp 317-494.

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Calculations," Geotechnical Engineering

Software GS90-1, Geotechnical Engineering

Center, The University of Texas at Austin.

258
Table 1 Sensicivicy Analysis of Dynamic Load, K,^

Base

a

Acceleration. ^

Cresc

u

Ratio

a/a

Dynamic Load

Ratio Dynamic Load

0.4 0.4

0.4 0.4

0.4

2.4

2.0

1.6

1.2 0.8

0.20

0.25 0.30

0.35 0.40

0.48 0.50

0.48 0.442

0.32

0.48 1.44 0.35 0.50

0.5

0.5

0.5

0.5

2.5

2.0

1.5 1.0

0.25 0.30

0.35

0.40

0.63 0.60

0.53 0.40

0.6 0.6

0.6

2.4

1.8 1.2

0.30

0.35 0.40

0.72

0.63

0.48

0.63 1.26 0.40 0.50

0.7 0.72.1 1.4

0.35 0.400.74

0.S6

Based on sensitivity analysis above,

Kn„ - 0.5 to 0.6 g selected for displacement computations

Table 2 Summary of Slope Displacements

Provided bv the T.A.G.

Displacement

Section Total

in. cm Average

in

.

£m

Villa del Monte

DM

SB

28 72 44 111

14 36

22 36

Schultheis Road

SR 96 244 48 122

259
Table 3 Displacement Computations for Villa del Monte Landslide Complex

and Upper Schulthels Road Landslide

Section

(U-upper

,

L- lower

Dynamic

Load Total Displacement

Ratio k1Yield AccelerationMobilized

Friction

Average Displacement

Ratio

K, / Yield Acceleration

Mobilized

Friction

surfaces)* E In. cm e deerees In. cm R degrees

Villa del Honce

DM-U 0 . 5 28 72 0 .11 0 ,055 17,,6 14 36 0. 21 0.,105 20.7

DM-U 0 ,6 28 72 0 .11 0,.066 18,,3 14 36 0. 21 0,,126 22 SB-U 0 .5 44 111 0 .06 0,,03 17 ,8 22 55.5 0. 15 0,,075 21 SB-U 0 .6 44 111 0 .06 0,,036 18,.3 22 55.5 0. 15 0,,09 22 SB-L 0 .5 44 111 0 .06 0..03 19 .3 22 55.5 0.,15 0.,075 22.5 SB-L 0 .6 44 111 0 .06 0..036 19,.8 22 55.5 0.,15 0,.09 23.5

Schulthels Road

SR-U 0..5 96 244 0 .03 0 .015 17,.6 48 122 0,.05 0 .025 18.3 SR-U 0,.6 96 244 0 .03 0 .015 17 .8 48 122 0,.05 0 .03 18.6 SR-L 0 .5 96 244 0 .03 0 .015 18 .6 48 122 0,.05 0 .025 19.4 SR-L 0,.6 96 244 0 .03 0 .018 18 .9 48 122 0 .05 0 .03 19.8

* Surfaces Identified in Figures Lo^Sand ^

Table 4

Sumraarv o f Mobilized Dynamic Friction. Static Stability. and Critical Ground-Water Depths Pre-Ouake Ground-Water Conditions

lection* Estimated Dynamic Displacement Mobilized Friction Factor Ground-Water Critical Ground-Water Load u Friction Required of Safety Depth That Change in Depth in. cm ob for Static Against Would Groundat Time e degree Stability Sliding** Trigger Water

of Quake FS. Static Depth**

Instability**

ft deerees ft ft DM-U 100 0.5 28 72 17.,6 14 1,,27 45 55 DM-U 100 0.6 28 72 18.,3 14 1.,33 30 70 DH-U 100 0.5 14 36 20.,7 14 1,.52 10 90 DH-U 100 0.6 14 36 22 14 1..62 >0 >100 SB-U 100 0.5 44 111 17. 8 15 .5 1..15 70 30 SB-U 100 0.6 44 111 16..3 15 .5 1,.19 60 40 SB-U 100 0.5 22 55.5 21 15 .5 1,.38 15 85 SB-U 100 0.6 22 55.5 22 15 .5 1 .46 5 95 SB-L 100 0.5 44 111 19,,3 16 .9 1 .15 55 45 SB-L 100 0.6 44 111 19,.8 16 .9 1 .19 45 55 SB-L 100 0.5 22 55.5 22.,5 16 .9 1 .36 >0 >100 SB-L 100 0.6 22 55.5 23..5 16 .9 1 .43 >0 >100

SR-U 50 0.5 96 244 17 .6 16 .5 1. 07 35 15 SR-U 50 0.6 96 244 17,.8 16 .5 1. 08 35 15 SR-U 50 0.5 48 122 18 .3 16 .5 1. 12 30 20 SR-U 50 0.6 48 122 18 .6 16 .5 1. 14 30 20 SR-L 50 0.5 96 244 18 .6 17 .2 1. 09 35 15 SR-L 50 0.6 96 244 18 .9 17 .2 1. 11 30 20 SR-L 50 0.5 48 122 19 .4 17 .2 1. 14 25 25 SR-L 50 0.6 48 122 19 .8 17 .2 1. 16 20 30

* Surfaces identified In Figures ^ , tF ar-d ^

** Based on dynamic strengths

260
Table 5 Calculated Friction Angles for Shall ow Failure

Surfaces Shown In Figures

Friction Angle Back- Friction Angle Required Calculated Froo Loma for Static F.S - 1 With

Failure Surface Prieta Earthquake Water Table Near Surface Designation (degrees) (degrees)

SR-2 19 33 DM-7 33 • >A0

DM- 8 20 22 SB-3 22 26

Table 6

Calculated Friction Angles for Deep Failure Surfaces Shown in Flpires lo^ II

,

I ZFrlcCion Angle Back- Friction Angle Required

' Calculated Fron Loma for Static F.S. - 1 With

Failure Surface Prieta Earthquake UaCer Table Near Surface Designation (deprees) (degrees)

SR-1 25 32 SR-U 18 22 SR-L 16 22 DM-1 20 29 DM-3 21 27 DM-U 20 23 SB-1 22 29 SB-2 19 22 SB-U 18 22

Table 7

Estimated Static Friction Angles for Deep Surfaces Shown in Figures IP li |ZFriction angle required

Failure surface for static F.S. - 1.5 Factor of safety

designation with a deep water table water table at ground

(<t^&l^^9) Surface

SR-1 35 1.09 SR-U 26 1.18 SR-L 24 1.10

DM-1 24 0.81

DM-3 26 0.94

DM-U 3 23 1.01

SB-1 28 0.98

SB-2 25 1.16 SB-U 23 1.03

261
Table 8 Computed Slope Displacements wlch No Scrgngth Degradation

fSurfaces Identified in Figures "j U , I 2-

Failure Surface Designation

Yield Acceleration

Based on Friction

Angles Listed

In Table 7 (g's)

Calculated

Displacement

in. cm

SR-1 0.150 0.5 0.300 24 9 0.6 0.250 32 13 SR-U 0.100 0.5 0.201 44 17 0.6 0.167 51 20 SR-L 0.100 0.5 0.199 44 17 0.6 0.166 51 20 DM-1 0.136 0.5 0.272 38 11 0.6 0.227 38 15 DM3

0.147 0.5 0.294 24 9 0.6 0.245 32 13 DM-U 0.104 0.5 0.209 44 17 0.6 0.174 51 20 SB-1 0.145 0.5 0.290 24 9 0.6 0.242 32 13 SB-2 0.138 0.5 0.276 28 11 0.6 0.230 25 10 SB-U 0.108 0.5 0.217 38 15 0.6 0.181 48 19

Table 9

Estimated Displacements for Postulated Scenarios

(Surfaces Identified in Figures lo, /

1

, I 2_

Scenario 1: Scenario 2:

Failure

Surface Designation

Water

at 100 ft In. £]B

Water

at 50 ft

ilL SSI

Water

at 50 ft Id sffi

Water

at 25 ft

in- cm

SR-1 76 194 65 165 SR-L 120 306 215 545 SR-U 120 306 194 494

DM-1 80 204 165 419

DM376 194 144 365 DM-U 109 278 241 612 SB-1 76 194 165 419 SB-2 80 204 157 398 SB-U 109 278 215 545

262
Q)

>-l (0

>i

T>3 rH

4-> Oi

rH o • a (0

03e -p 3 Jh o a) +J "4-1 C Q) O 0) u «• Q) CM +J

0) (0 u3

•H

(0

0)

H

>i(J> a a^ 3 H

-pw • rH

<4-t (0 oC tt) o

•H

4-> Q) (0 <t-l O Q) O Q) iJ «U • Q) H -P

<w

>^3H

263
264
(0GO

-H

•PU

0) (0 U 0)u c

(0 0) p

05

C

D>

O

C S O (1) A Q

<M 10 O rH

+) -H

(0U OQ 3"C • (0

0)Q 3

0» •H

fa

COco

•H

+io

0)

CO

(0 -O 0) C M

flj (0 M T3 D> «3 C O

•H « O Q) -H

0) o -P

rH

m

D

C J3 o o

•H

W

4J

(0 !-t U Q) o aD• c

If) -H

0)3

01

265
FIGURE fc Estimated Upper and Lower Locations of Basal Shear

Surfaces for Section SR, Upper Schulthels Road Landslide, from William Cotton and Associates, Inc. (1990)

FIGURE 8 Estimated Upper and Lower Locations of Basal Shear Surfaces for Section DM, Villa Del Monte Landslide Complex, from William Cotton and Associates, Inc. (1990)

FIGURE <1 Estimated Upper and Lower Locations of Basal Shear

Surfaces for Section SB, Villa Del Monte Landslide Complex, from William Cotton and Associates, Inc. (1990)

266
20C0

1500

1000

500

0

-3000

Ground Surlactr

SchuJthels RMd \

urface L (Deep) Surface U CDeep) Surface 2 (SbaUow) Ptueadc Surface

Surface 1 (Deep. UraJted)

Upper Schultheis

Cross Section SR

-2500 -2000 -1500 -1000 -500

FIGURE lo Assumed Locations of Shallow and Deep Sliding Surfaces

for Section SR, Upper Schultheis Road Landslide

u.

«1

ca

0)3

O

sz

Villa Del Monte

Sky Wrw Terrace

Phreallc Surface

Gmun<l Surface

Sunact Drtvc

Sky View Tenace

Surface 1 (Deep. Limited)

Surface 7 (Shallow)

Surface 8 (Shallow)

Surface U (Deep)

Cross Section DM

-2 -1 Thousands of Feet

FIGURE V\ Assumed Locations of Shallow and Deep Sliding Surfaces

for Section DM, Villa Del Honte Landslide Conplex

2(XM

1500

-

CD© 1000

Phreatic Surface

Ground Sur&ce

Sky View Terrace

Surface 2 (Deep, Limited)

' Surface 3 (Shallow)

^ Surface U (Deep) Surface 1 (Deep. United)

Villa Del Monte

Cross Section SB

-1500 Feet

FIGURE IZ Assumed Locations of Shallow and Deep Sliding Surfaces

for Section SB, Villa Del Monte Landslide Complex

267
DEPTH TO WATER IN FEET

0.36^

FRICTION ANGLE. DEGREES

FIGURE I 3 Relation Between Mobilized Prictlon Angle, Depth to

Water Table, and Hold Value of Horizontal Acceleration for Upper

Basal Shear Surface, Section SB. Upper Schulttiels Road Landslide

DEPTH TO WATER IN FEET

FRICTION ANGLE, DEGREES

FIGURE (4- Relation Between Mobillted Friction Angle, Depth to Water Table, and Yield Value of Horizontal Acceleration for Lower

Basal Shear Surface, Section SR, Upper Schultbels Road Landslide

UPPER SLIOINQ SURFACE

38 CROSS SECTION

VILLA DEL MONTE AREA

FRICTION ANGLE, DEGREES

FIGURE /7 Relation Between Mobilized Friction Angle, Depth to

Water Table, and Yield Value of Horizontal Acceleration for Upper

Basal Shear Surface, Section SB, Villa Del Honte Landslide

Complex

3

LOWER SLIOINQ SURFACE

SB CROSS SECTION

VILLA DEL MONTE AREA

DEPTH TO WATER IN FEET

0 e 10 16 20 2S 30 3S 40 FRICTION ANGLE. DEGREES

FIGURE /? Relation Between Hobillzed Friction Angle, Depth to water Table, and Yield Value of Horizontal Acceleration for Lower Basal Shear Surface, Section SB, Villa Bel Monta Landslide Coeplex

268
d d d o d

CMd

(6) NOIlVa31300V lS3dO 3Sa3ASNVyi >IV3d

4) „

(/I <-i u o>

0) o\ >

mc u

la <D U T>

Fh u

(A

<S) B

u o o u

4-1 •oc

« « n a

fl CO £

« Id O M

•p <M (0 o

c <v O

« 3

-H m

M m

(0 Q) as

B

o w o c

o

•H

4Jm

0 <u

0) w o 2 o 5 <

oH

6^

ALL

SITES

MEAN

PLATE

BOUNDARY

^19

KM

FOCAL

DEPTH

II o yo y >

O S

2 io

z s ^

5

o

o

o

o

o

to o

o

o

lO

,03S/IAI0 'NOIiV«31300V IVINOZIUOH

0) tT> u c

C ID

ID r:

4J o o

•H E - a in >t

4) w

+J W N

•H +J

C 10 -H

•H c 2

+) 3 _ rt o e

M CO O

V ^

9) in

«) 4-1 o « >

rt: 0- S

§0 •-• o c

1-1 n Id O 01 ^

"i^ ^ 0) O (A 04 ID ^

«-l -H o w w

n ^ 4J O O4 H < 4) •P Q

IS3 O —

C <w nj 05 0 O «

4-1 M o in

4J «) fc. H

3 -P -H U -H r-l S 4-> a) 4J ^ « 4J Z

^^ c 2 > > c 5 < w fl o

269
FIGURE il Relation between Acceleration Ratio k^iax / ^V^? (averaged dynamic load to peak crest acceleration^ and Relative

Height of Sliding Surface (y/h)

FIGURE -2 Average and Range for Relation Between Accumulated

Newmark-Sliding-Block Displacement, U, and Acceleration Ratio,

v,/k^ for Interpolated Magnitude 7 Earthquakes after Makdisi

ahd Seed (1979)

270
Study on Prediction of Lateral Ground Flow by Soil Liquefaction

and Its Influence on Piles

by

Ken-ichi Tokida', Hideo Matsumoto^, Ikuo Towhata^ and Yasushi Sasaki'* SUMMARY

An analytical method based on the ininimiim potential energy principle is proposed lor pre- dicting the permanent lateral displacement of

ground due to soil liquefaction. Furlliermoro,

a practical and simpliCiod method ohialneci tiy parametric studies witli the use oC tiie analytical method proposed is also discussed to f)re- dict a maximum lateral displac^emenl oC an in- clined ground. Tliis was applied to i^eal cas(\s

of shal<ing table tests and case studies in situ

for past ear tiiquaUes.

Finallj', experimental studies were carried out

to investigate the external force acM.ing on

piles in ground flowed by soil liquefaction.

KEY WORDS :I,iquefaction, Lateral (Ground I'low, Permanent Displacement, Prediction, Minimum I'o- tential Energy Principle, Paramoliic Study,

Model Test, External Force, Pile

1. FOREWORD

f,ateral ground flow induced by soil liquefaction toolt i)lace apparently in past earthcpiakes

[Refs.]~3] . For establishing a seismic; design method of structures considering a lateral

ground flow induced by soil liquefaction, es- sential factors relating to ground flow, a pre- diction method of permanent displacement and

an external force acting on substructures

should be investigated. The characteristics of

lateral ground flow due to soil liquefaction

have been studied by large-scale shaking table

tests In the Public Works Research Institute [ Refs.4~-6] . In tliese tests, the influence of

inclination of tlie ground surface and thickness

of the liquefiable layer were investigated.

This paper presents results of large-scale

shaking table tests for investigating the inl'lu- ence of the slope length on ground flow [Ref.7

] , and an analytical method based on the mini- mum potential energy principle to i)redict

permanent lateral displacement of inclined

ground flowed by soil liquefaction [Refs.8~ll

J . A practical and simplified procedure obtained by the parametric studies is also proposed to predict maximum lateral ground displacement. Furthermore, shaking table tests on piles in flowed ground [Ref.T] and model

tests on piles driven in liquefied soils [l?ef. 14

] were conducted to investigate the external

force acting on piles in ground flowed by soil

1 iquefaction.

2 • LARCE SCALE SHAKING TA13LE TESTS

2. 1 Ml'TllO D OF EXPERIMENTS

Shaking table tests on four ground models

(Model-9 -~ 12) were carried out to investigate

tlie influence of slojie length on ground flow

induced by soil liquefaction. The conditions

of the ground models are shown in Figure 1 and

Table 1. A rectangular container used for Model-r)~l2 was 8.0 m long, 1.0 m wide and 1.0 m

high. Six transparent; glass windows (each is 0.8 m wide and 0.9 m high) were installed at tlie one side of the square container. The ground

models fundamentally consisted of three layers

: a lowermost nonl iquefiable layer, a middle

liquefiable layer and a uppermost unsaturated

layer. However, Model -12 did not have tlie uppermost unsaturated layer. The materials

used for a lowermost nonliquefiable layer and a middle liquefiable layer were mountain sands

(Mean diameter of particles: Dsn = 0.25mm) taken

from Mt. Sengen-yama and gravcls(Dso = 7mm) for

the uppermost layer. The detailed method to prepare tlie ground model is presented in I^efer- ence 4. The thickness of liquefiable layer

(111), the thickness of uppermost unsaturated

layer (IInl) and the gradient of ground surface

( 0 ir.) were constant to be 350 mm, 150 mm and 5 %, respectively. The length of Models-9, 10, 11 and 12 was varied to be 2 m, 4 m, 8 m and 8 m, re- spectively. The water level was set equal to the upper boundary of the liquefiable layer.

In each test, a constant sinusoidal acceleration

with a frequency of 2 Hz was employed for 20 seconds (40 cycles); and several levels of maxi- mum acceleration were applied stepwise to the

shaking table as an input motion. As shown in Figure 1, pore water pressure meters, accelerometers, displac^emcnt meters and strain gage me- ters were installed in and/or on the ground

1) Head, (Ground Vibration Division, Earthquake

Disaster Prevention Department, Public works

Research Institute (PWRl), Ministry of Construction.

2) Research Engineer, ditto.

.'!) Associate Professor, University of Tokyo.

4) Director, Earthquake- Disaster Prevention

Department, PWRl.

271
m(i(i(M.s to measure the time histories of pore

walcr pressure, aeceleralion and displacemcril,/

strain of layers during exeilal ion. 13eneh mark

points were also set on the ground surface to measure manually and photograpli the movement

of lateral displacement of ground surface.

2.2 RKSULTS OF EXPI' RIMENTS

Figure 2 summarizes the relation between inpijt

maximum acceleration of the shaking table and

the accumulative lateral displace- ment. at the

center of the ground surface. As for Models-9,

10 and 11, it can be seen that the lateral

ground flow begins to go up at the lower level of input acceleration, according to the in- crease of length of tiic ground surface.

IBccause the effect of both end walls of the container is small in the case of Model-ll, the

liquefiablo layer seems to be deformed more

easily than Modcls-9 and 10. On tlie other

hand, the liquefiable ground overburdened by

an uppermost unsaturated layer such as Model-ll is less liquefiable than that without

an uppermost unsaturated such as Model-12.

3. PREDICTION OF LATERAL GROUND FLOW

sand is simply modeled by equal ion (4).

du

r = G + T 3z

(4)

where, r , G and r , are the shear stress, the

shear modulus of a liquefied la3'er and the re- mained stress. Hereupon, G = 0 and t v = 0, because liquefied soils beliave as liquid whose

flow is governed by the hydraulic gradient.

4) The surface unsaturated layer behaves as a solid bar of Young's modulus E.

Substituting equation (2) into equation (3) and

considering zero displacement at tlie surface of

the unl iqucfiable base layer {z=13), the function w is expressed in terms of F(x). By using u and

w, the potential energy of the ground in Figure

3 is calculated [Rcl".8~U] and the true F(x)

makes this energy minimum. By using the

variational principal and theory of Euler's eq nation.

dF C^x^ + C.-.X + A,

dx Cix"^ + C-^x + C3

(5)

3 . 1 M INIMU!^ POTENTIAL ENERGY PRINCI P_AL

Figure 3 illustrates a grounrl model which consists of an unliquefiable base la^'er (Thickness:

B), a liquefiable layer (II) and an unsaturated

surface layer (T) [Refs.8— ll j . The surcharge

(P) includes the weight of the surface layer,

which is estimated to be moved together with

the liquefiable layer (based on the shaking ta- ble tests). The horizontal and vertic'al displacements are denoted by u and w, respectively. The ground configuration varies lineally with the horizontal coordinate , x;

where,

Ci

Cs

47 R r bllo

, C2 = + Ec

n n

4 r Mo''

+ ETo,

n

1 27b 2eb

C^ = - — ((a>b)(e t ) + 2 n n

2 7 Il>3 2cIlo Co = r , - (a+b)(Po+-

B = Bo + ax, H = llo + bx

T = To + cx, P = Po + ox (1)

Furthermore, four assumptions were made in the

analysis;

1) The lateral displacement is approximated by a sinusoidal distribution function in the z direction;

u = F(x)sin [ n (z-B)/211J (2)

where, F is an unknown function of x and stands

for tlie displacement at the surface (z=B+ll).

2) Soils flow with a constant volume;

3u 9w

+ = 0 (3) 3x 3 z

3) The stress-strain relation of the liquefied

Ai Is a constant paramet(;r and 7 is the unit

weight of the liquefied soil. Thus, equation (.5) is solved under relevant boundary conditions

at x=0 and x=L, and the solution of Equation (.5) is described in References 9~ 11.

3.2 APPLICATION MlNllviUM POTENTIAL ENERGY

P lllN CniA I. :iXLl^ EA L_CAS_ES

The minimum potential energy method proposed

in the former section 3.1 was applied to real cases of shaking table tests conducted at the

Public Works Research Institute. Figure 4 shows the results of calculations and mea- surements of maximum lateral displacement in the case of Model- 10 (slope length: 4 m). The

analysis was conducted with the estimated val- ues of E = 5 t/m', 7 = 1.80 t/m". and the unit

weight of the surface layer o'f 1.40 t/m^. Be272
cause Lhc ground model surrac,-e rieig-liboririf; (he

end of upper side of Ihe container was settled

and not cracked, the boundary conditions of F = 0 at both X = 0 and x = I_, (fixed at bolli ends of

lower and upper sides) were em|)loyed in this

study. It can be see that the peak value of

maximum lateral displacement calculated is a- bout 16 cm at the horizontal coordinate of 3 m. Because the maximum lateral displacement mea- sured ranges from 20 to 25 cm, the calculation

is somewhat larger than the measuremetit.

The minimum potential energy method was also

applied to case studies in situ In past earthquakes. Figure 5 shows one of the examples to compare the calculation of lateral displacements at the surface and the measurement at the

Maeyama-llill area in Noshiro City damaged in the 1983 Nihon-Kai Cyubu Farthciuake. Since

the soil parameters were not known clearly, the

analysis was conducted with the fixed values of

E = 1,100 t/m^, r = 1.80 t/m". and the unit

weight of the surface layer of 1.60 t/m".

Boundary conditions of F = 0 at x = 0 (fixed at the end of the lower side) and dlVdx = 0 (free at the end of the upper side) were employed in this

study. Although overestimating, the agreement

with the observation seems satisfactory.

3 . 3 PHAniCAIJ^llOP_BX 'iAJiiAM '^T^ I C. SJlj D.I FS WITH MINIMUM ENERGY PRINCIPAL

Although the minimum potential energy method

can be applied to In situ sites according to each ground condition, a simplified [)ra(!tical

method Is necessary to be established from the

engineering view point. llamada, M. et al. proposed equation (6) empirically based on case

histories in past earthquakes in .Japan to estimate the maximum lateral displacement of an Inclined ground simply and quantitatively [ Ref.l2] .

D = 0.75/'Ii ''vr 0 (6) where, D Is the maximum lateral displacement of

the ground surface (m), II is the thickness of

liquefied layers (m) and 0 is the maximum gradi- ent of the ground surface or the lower boundary of the liquefied layer {%).

However, parametric studies were conducted to propose a simplified method to estimate the maximum lateral displacement in this study. As shown in Figure 6, according to experimental

studies [Ref.7] , four parameters: the length of

ground flow area (L), the average thickness of

the liquefied layer (11), the average tlilckness

of the unsaturated surface layer (T) and the

gradient of the ground surface ( 0 n) were em- ployed as important factors against lateral

ground d is[) lacomcn I. induced by soil liquc;fac;-

tion. The values of each parameter to calc^u- late a maximum lateral displacement were se- lected as follows;

L (m) : 10, 20, ,50, 100. 200, .500, 1000

11 (m) : 0.1. 0.2. 0.5. 1. 2, 5, 10, 20 T (m) : 1, 2, ,'5, 10, 19

0 ..(%) : 0. 1, 0.2, 0.5, 1. 2. 5. 10. 20

where, total thickness of a uiisaluratcd surface

layer and a liquefied layer was limited within 20

m. and the end of lower side and the one of up- per side were assumed to be fixed and free, re- spectively.

Equation (7) is assumed to represent the maxi- mum lateral displacement (D) in this study.

D .X X ir X T^' (7)

The coefficients a, b, c, d and e are obtained by

the multiple regression analysis for the maxi- mum displacement at tiie center of a slope and

the end of the upper side, according to the

length of the slope as follows;

[ 10 < I, g; 100]

Center of slope:

11^ T' 0 s'-

End of upper side: D= 2.07 X 10'^ l/-"" iI"-""" T~

[ 100 g 1. S 1000]

Center of slope:

(8-1)

(8-2)

(8-3)

End of upper side: D= 1.70 X 10 """ L 0

(8-4)

Equation (8) was applied to sites where lateral

ground flow due to soil liquefaction occurred

in past earthquakes. 12 sites In the 1964

Nligata Earthquake and 27 sites in the 1983 Nihon-Kai Cyubu Earthquake where lateral

ground displacement was measured and whose

ground profiles were investigated in detail [ Ref. 13] were selected. Because the ground

conditions in situ were not always clarified to decide the length of ground flow area (L), ground conditions of Interest are classified in- to four types according to the reliability to decide both boundaries of grounds as follows;

Type I : Both boundaries are reliable.

Type 11 : Only one boundary of upper side is reliable.

Type ni : Only one boundary of lower side is reliable.

273
Tj'pe IV: Botli boundaries arc nol reliable.

I'lfjiiros 7 niu] 8 presoiil relations IhMwcvmi maxi- mum displaeomoni s measured in silu and eal(;ulated ones with the use oC eijuations (8-3) and

(8-4) for ground types I, 11, 11! and IV, re- spectively. The maximum displacemenls calculated by equation (G) are also referred. It can be seen that the calculated displacement bascni on equation (8) l\as relatively good agreement

with the measured one (except for large displacements measured) especially at sites of

ground types I, II and 111. This also indicates

that Icnowing the ground boundaries is very

Important and the ground condition should t)c well understood.

4. EFFECT OF LATI'i^L GROUND FLOW ON PILi'S,

4.1 LARQI:L_SCA|'I^__ SHAKINGJlAm

LNJILOMRJ^!iQ_UNI)

Shal<ing table tests on two model grounds Model-1 and Model-12) wore also carried out to in- vestigate the influence of ground flow against

pile models set in liquefied soils. The conditions of ground models are shown in Figure 1. As shown in Figure 9, a vinyl chloride pipe ( 1 = 790 mm, <f> = 21 mm) was fixed on the bottom of a container. Strain gages were put on the pipe

model to measure the section forc;e. Figures

10 and 11 show time histories of bending moment

of the pile model at points of B2 to 13(5 and

bending moment distribution of the i)ile model

at the elapsed time (1,4, .5, 0, 8 and 10 seconds)

respectively. The bending moment of the pile

model begins to increase at tiie elapsed time of

about 5 seconds when the ground (liquefied partially) begins to flow. Although points of R2

to 134 are overscaled, the bending moment begins

to decrease after the ground licpiefied perfectly at the elapsed time of about 11 to 12 seconds. Tliis indicates tliat the maximum load

induced by ground flow due to soil liquefaction

doesn't occur in the perfect liquefaction l)ut in partial liquefaction, and the relation between

the load acting on piles and the liquefaction

potential should be investigated for design of

foundations against soil liquefaction.

4.2 TESTJ3NJ2ILFSJMyjLNJJ^

4.2. 1 MEinODJjriJiXPEILLM

Tests on piles forced by lateral ground flow in liquefied soils were conducted with the use of a test apparatus as shown In Figure 12. The container was 1.8 m long, 0.6 m wide and l.O m high.

At the bottom of the container, a dense sand

layer was prepared to arrange the height of

licjuefiable sand layer and a water supply tank was installed to boil the above layer l)y supplying water through a water suF)ply pump. In this wa3', a loose saturated layer was made easily and repeatedly. The material of the

liquefiable layer Is Toyoura sand. The

liquefiable layer was shaken and liquefied by one blow of a hammer against the side wall of

the container. Just after the occurrence of

liquefaction, a group of pile models fixed by a sliding plate wliich was supported on a support

frame were forced to displace in the liquefied

layer by a winch which was driven by hand. As

sliown in Figure 13, 10 pore water pressure me- ters (PI ~ PIO) wewe installed at two positions

with the horizontal distance of GO cm. The displacemetit of a sliding plate and/or pile models

and the total load to drive them were measured

by a displacement meter and a load cell which were set on the sliding plate, respectively.

Three values of constant sliding velocity of the

sliding plate were used: 4. .5 mm/sec, 2.6 mm/sec

and 1..S mm/sec. The pile models was driven

horizontally inside of two positions of pore water pressure meters. A vinyl chloride pipe {<}> = 18 mm) reinforced inside by a steel bar was

used as a pile model. The number of piles and

their arrangement (lattice arrangement and zigzag arrangement) were varied as shown in Figure

14. The friction forces between the support

frame and a sliding plate measured in advance

were 0.9 kgf, 0.8 kgf and 0.75 kgf according to the sliding velocity 4. .5 mm/sec, 2.6 mm/sec and

1.."5 mm/sec, respectively.

4.2.2 RESULTS OF I-XPE RIM ENTS

Figure 15 shows an example of the time histories: acceleration of the container, total

pulling load measured by the load cell, displacement of piles and excess pore water pressure. It can he seen that after sliaking by the

hammer, the condition of perfect liquefaction at points of Pi and P.5 continues for about 4 sec- onds and it takes about 20 seconds for excess pore water pressure of the whole ground to disappear. Furthermore, the total load increased

gradually until the elapsed time of about G sec- onds and rapidly after that.

Figure 16 presents relations between the total

load, excess pore water pressure at the points

of PI and P5 and the displacement of pile models. The relation between tlie total load and

the displacement can be classified into two

stages: stage I and stage n . Although the

displacement of pile models increases at the

stage I , the total load is almost constant at a low level. On the other hand, the total load

increases according to the Increase of the displacement at stage n. It should be noted that

274
the excess pore water pressure ol" IM and P5 at the turning point ( D ) between stage I and

stage II indicates about 20 gl'/cm^.

Figure 17 explains a concept of the distribution

of an effective overburden pressure ( a ' ) and/

or an excess pore water pressure ( A U ) in liquefied soils. In tliis paper, the excess pore

water pressures measured at the points of PI and P5 which are 35 cm deep are used to calculate the excess pore pressure ratio ( Ru ) at the tip of a pile model wliich is 20 cm deej).

Furthermore, the exc:ess pore water pressures ( A U ) of PI and P5 at tiie turning point ( D ) shown in Figure 16(b) are identified as tlie ini- tial effective overburden pressure at the tip of

pile models as shown in Figure 17. Because the

average of the excess pore wa1er pressure of PI and P5 at the turning point ( 1) ) for all lest cases is 20 gf/cm^ and the unit weight of soils

in the water (r ') is 0.914 gf/cm% the initial effective overburden pressure at tlie tip of

pile models is assumed to be 20 gf/cm^, in other

words, the initial depth to calculate the excess pore pressure ratio ( llu ) is converted to l)e 21.9 cm (= 20 / 0.914). The distribution of ex- cess pore water pressures in the cases of I^J = 1.00, 0.90, 0.45 and 0.00 is exemplified in Figure

17.

Sliding/Relati ve Ve locity and Fxternal Force

Actin g on Piles

Figure 18 presents relations between the total

loads and the sliding velocity of 9 piles ( 3 rows X 3 piles) latticed in the cases ol' Ru = 0.90 and 0.00 (almost perfectly liquefied condition and not liquefied condition) respectively.

It can be seen that the total load acting on 1 he

piles is almost constant in spite of the sliding

velocity of the piles when soil liquefaction

does not occur: Ru = 0.00. On the other hand,

the total load increases according to the in- crease of the sliding velocity of tlie piles when

soil liquefaction occurs partially: for example,

Ru = 0.90.

Liquefaction Poten tial anci I^xtexnaJ £orcc

Actin g on Pil es

Figure 19 presents relations between the total

load, the excess pore water pressure at the tip

of pile models, the sliding velocity of pile models (4.5 mm/sec, 2.6 mm/sec and 1.5 mm/sec), and

the arrangement of them (lattice arrangement of

5 rows X 2 piles, l'l-9, and 2 rows x 5 piles, PI -8, and zigzag arrangement of 3-2-3 piles, P2-5, and

2-1-2-1-2 piles, P2-6, see Figure 14) in the sliding direction. From these figures, it can be

noted that the total load decreases according

to the increase of excess pore water pressure

ratio. I'll rthormore, according to the inc:rease

of the sliding velocity of the pile models, the

total load increases at each excess pore water

pressure ratio. As for pile arrangement, the

total loads in tlie cases of Pl-8 and P2-6 are

larger than that in the cases of Pl-9 and P2-5,

respectively. This indicates that the total

load is related to the number of piles in tlie normal direction of a groutid flow; in otlier

words, the total load acting on piles or tlie resistance potential of piles against a lateral

ground flow goes up according to an increase in the number of piles in the normal direction of

the lateral ground flow.

5. CONCLUSION AND FUTURE REMARKS

Based on the studies in this pafier, the fol- lowing results have been obtained;

(1) I'^om shaking table tests, it can l^e indicated

that a liquefied inclined ground becomes easy

to flow laterally according to the increase of

slope length and/or the decrease of the thicl<- ness of an uppermost unsaturated layer.

(2) A method based on the concept of minimum

potential energy is i)roposed to predict the

lateral ground displacement induced t)y soil liquefaction anci applied to real cases of both

shalting table tests and case studies in situ in past eartluiuakes. Although overestimating,

the agreement with the observations seems

satisfactory.

(3) A simplified method to est;imate the maximum

lateral displacement due to soil liquefaction is proposed based on a parametric study with the

use of the minimum potential energy method proposed in this study. it can be indic:ated that

the calculated displacement has relatively good

agreement with tlie measured one when the

ground condition sucli as boundaries of both

sides, area of the slope and soil profile can t)e understood jjroperly.

(4) From shaking table 1:ests, It can be indicated

that the maximum load acting on a pile in the

soils flowing due to soil liquefaction doesn't occur in the perfectly liquefied condition but

in tlie partially liquefied one,

(5) Fundamental tests on piles forced by a lat- eral ground flow In the liquefied soils are conducted: and it can be known that the total load

acting on piles by ground flow due to soil liquefaction decreases according to the increase

of excess pore water pressure ratio and In- creases according to the increase of relative

velocity between flowing soils and piles In the

partially liquefied condition.

275
(G) The total- load acting on piles, in other

words, a resistance potential of jiilcs against

lateral ground flow, goes up ac(K)rding to tlie increase of the number of piles in tJie direction

normal to the lateral gro\ind flow.

Further investigation should be carried out in the future to establish a seismic design method

to consider the effects of lateral ground flow

due to soil liquefaction.

ACKNOWLEDGEMENT

Tlie authors wish to exin'c^ss their sinc;ere grat i- tude to Professor T. Kat ayama, University of

Tol<yo and F''rof'cssor M. Ilamada, University of

Tokai, for their Invaluable advise and discussion, and to Professor Timotliy B. D'Orazio,

San Francisco State University, who visited the

Ground Vibration Division of Public Works I?c- search Institute as an invited researcher by the

STA (Science and Technology Agenc^v) Fellowsliip

and Mr. Takuo Azuma of PWRI for their assistance to present this jiaper.

REFERENCES

1) Ilamada, M., Yasuda, S. and Wakamalsu, K., "Case Study on Liquefaction-Induced Ground

Failure During Earthquakes in .Japan", Proc. of

1st .lapan-United States Workshop on Liquefaction, Large Ground Deformation and Their Effects on Lifeline Facilities, Tokyo, November,

1988.

2) Ilamada, M.,' Yasuda, S., Isoyama, R., and

Emoto, K., "Observation of Permanent Ground

Displacement Induced by soil Liquefaction",

Proc. of JSCE, No. 376/111 -C, pp21l-220(in .Japanese).

3) Yasuda, S., Watanabe, II. and Yoshida, N., "Reconnaissance Report on the Valle dc la Estrella (Costa Rica) Earthquake of Ar)ril 22, 1991", .Journal of the .Japan Society of Soil

Mechanics and Foundation Engineering, Vol. ,39, No. 8, August 1991 (in .Japanese).

4) Sasaki, Y., Tokida, K., Mnlsumoto, II. and

Saya, S., "Experimental Study on Lateral Flow

of Ground Induced by Soil Liquefaction", 2]sl. U.S. -.Japan Panel on Wind and Seismic Effects, UJNR. Tsukuba, May. 1990.

5) Sasaki, Y., Tokida, K., Matsumoto, II. and

Saya, S., "Experimental Study on Lateral Flow

of Ground Induced by Soil Liquefaction", 8tli

Japan Earthquake Engineering Symposium 1990,

Tokyo, December, 1990.

G) Sasaki, Y., Tokida, K., Matsumoto, II. and

Saya, S., "Experimental Study on Lateral

Flow of Ground Due to Soil Liquefaction", 2nd

International Conference on Recent Advances

in Geotechnlcal Earthquake Engineering and

Soil Dynamics. St. Louis. March, 1991.

7) Matsumoto, II.. Tokida. K. and Saya, S.. "Effects of Boundary Condition on Ground Flow

Due to Soil Liquefaction", 21th JSCE Earthquake Engineering Symposium-1991, Tokyo.

.July, 1991(in Japanese).

8) Towhata. 1. and Tamari, Y., "Calculation

Method of Permanent Displacement of Liquefied

Ground by Variational Theory", 14th Lecture

Meeting of Research Society of Civil Engineering, March, 1990(in .Japanese).

9) Towhata, 1.. Tokida, K., Tamari. Y.. Matsumoto.

II. and Yamada, K., "Prediction of Permanent

Lateral Displacement of Liquefied Ground by

Means of Variation Principle". 3rd Japan-U.S.

Workshop on Earthquake Resistant Design of

Lifeline Facilities and Countermeasures for

Soil Liquefaction, San Francisco, December,

1990, Technical Report NCEER-91-0001

.

in) Towhata, I., Sasaki, Y., Tokida, K.. Matsumoto, II. and Tamari, Y., "Permanent Displacement of Liquefied Ground", 9th Asian Regional Conference on Soil Mechanics and Foundation Engineering, Bangkok. December. 1991.

11) Towhata. L, Sasaki, Y.. Tokida. K., Matsumoto, II., Tamari. Y.. Saya. S. and Yanada.

K.. "Prediction of Permanent Displacement of

Liquefied Ground by Means of Minimum Energy

Principle", submitted to Soils and Foundations.

12) ilamada, M., Yasuda, S., Isoyama. R. and

Emoto. K.. "Study on Liquefaction-Induced

Permanent Ground Displacement and Earthquake

Damage". Proceedings of Japan Society of Civil

Engineers. No. 376/ III -G. PP.221- 229. December.

198G(in .Japanese).

13) Ilamada, M. , "Large Ground Deformations and

Their I'Mfects on Lifelines/1964 Niigata Earthquake/1983 Nihon-Kai Chubu Earthquake, Technical Report NCEER-92-001, National Center for

Earthquake Engineering Research, USA. 1992.

14) Tokida, K.. Matsumoto. II. and Azuma. T., "Experimental Study on External Force Acting

on Pile Induced by Lateral Ground Flow Due to

Soil Liquefaction", submitted to the 47th Annual Meeting of Japan Society of Civil Engineering, Sendai, September. 1992(ln Japanese).

276
Table 1. Condition of ground models

G round Grad i en I of Length of Thickness of Th i ckness of

Model Ground Ground Model Mquef iable Uppermos

I

NonNo. Surface (%) /Slope (m) Layer (cm) 1, iquficd 1 ,ayer (cm

9 5 2. 0 35 15

10 5 4.

0' 35 15

11 5 8. 0 35 15

12 5 8. 0 35 0

Surfoce Loyer

Liquefiable Loyer Non- Liquefioble Layer

Model - 9

o Pore Woter Pressure Meter D AcceLeromeler

~l Displocement Meter (Surfoce)

> Displacement Meter (Underground) + Bench Mark Point I—I Stroin Goge Meter

Unit : mm

Pile Model

Model - II

P.W R Meter Disp. Meter

Model-12

Figure 1. Ground models and measurements for

large scale shaking table tests

277
90

E E

70

.5-^ 60 Q t: H'^ 50 i o ^5 40 |<5 30 |S 20 ^ 100

Length of Ground Model

L

= 8m Model - 9 0 Model - 10 A Model -

1

1

+ Model -

12

0 50 100 15 0 200 250 Maximum Acceleration at Shaking Table [gol]

300

Figure 2. Relation between lateral ground displacement at center

of ground surface, input acceleration and length of ground surface

x = 0 x= L

Figure 3. Ground model applied for minimum

potential energy method

Model-IO

E = 1l00tf/m^ y^l.60t/m^

1 1 1 1 1

0 0 Calculotion

' Measurement

1

, ! .! ! !! i ! i 11 Ill

100 200 300 400

Horizontal distance (ml

500

Figure 5. Case history study at Maeyama Area

in Noshiro City in the 1983 Nlhon-Kai Cyubu

Earthquake, Japan

600

0,1

0

-0.1

Figure 4. Case study applied for

shaking table test: Model-10

278
Figure 6. Ground model for parametric study

with use of minimum potential energy method

0.0 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0 Estimation (m)

Figure 7. Relation between measured

displacement In situ and calculated one (Types 1 . 0 . HI)

0.0 1.0 2.0 3.0 4.0 5.0 6.0 7.0 R. 0 Estimation (m.)

Figure 8. Relation between measured

displacement in situ and calculated one (Type IV)

Model - 12

Liquefied Layer

Non-Liquefied

Layer

Fixed 777777777777777T77777777

Bottom of Container

Figure 9. Condition of pile model

and measurements

279
70

-35 -70

kgf- cm B6

10 15 20 25

(sec)

kgf cm

-35 -70 ,

B5

10 15 20

. (sec) 25

70

-35 -70

kgf cm B4

10 15 20 25

(sec)

70

-35 -70

kgf cm 83

10 15

. . (sec) 20 25

B2

kgf cm

70

^a

-35 .. -70r '.""r TT" (sec) 5 0 5 10 15 20 25 Elopsed Time

Figure 10. Time histories of bending moment of pile model

Liquefied Loyer

Non- .. Liquefied Loyer

8

7- 8 6

8 5 8 4 8-3

8'

2'

8 1

-100 0 100 -100 0 100

t : 1 sec t

: 4 sec t - 5 sec t = 6 sec

Elopsed Time

! : 8 sec t = 10 sec

Unit : kgf cm

Figure 11. Time history of bending moment distribution of pile model

280
Woler Supply Pump

Support Ffome - Sliding Plate -

P 1-1

• 1 r4

Lood Cell J i?.

Water Supply Pump

" Water Tonl<

Wincti

Woter Supply Tonk

1 I

Woter

Supply Tank

Vibration

Valve

Water Tonk

Woter Supply Pump

Pipe Model Water Pump

Lood Cell , Disp

p '1 -

P 3 - •P 8 P 7 P 2 - - -P 6 PI- -P 5

Woler Su pply Tonk Dense Sond Layer

Valve

Woter Tonk

Woter Supply Pump

Figure 12. Test apparatus to drive piles in liquefied soils

( Unit mm)

1800 900 900

Pore Water Pressure Meter

p I ~ p 4 • I

/ Load Cell

-Sliding Plate

Oisplocement Meter

P 4 •

P 3 •

P 2 •

p : •

O

Accelerometer

Group of Pile Model Load Cell f^Wfp m DisplocemenI Meter

• P6

'

• P 5

• _L

Woter Supply Tonk ond Dense Sond Loyer

600 600 600

Figure 13. Ground model and measurements

281
t

Pl-2 (IRx 3P)

Pl-3 (3Rx IP)

t t

Pl-1 (IRx IP)

o • • o o o • • • • • • • <!' O • • • '-' • • • • • o o c- o • • • o • • • • •

Pl-4 Pl-7 (2Rx2P) (3RX3P)

Pl-IO (3Rx 5P) t

Pl-5 (IRx 5P)

• • • • .• o o o o o

O C' o o o

O Ci O C' o

Pl-8 (2Rx 5P) Pl-11 (5Rx 3P)

is t a

oo • o • o • o •

Pl-6 (5Rx IP)

Pl-9 (5Rx 2P)

• • • • • • • • •

P1-I2 (4Rx 4P)

: •

• •

J • c

P2-1

(1-2-1 P)

P2-2

(2-1-2 P) t

O • *

<'• o • o

O • • Ci

P2-3

(3-2-3 P)

• • •

• • >; • • •

P2-5

(1-2-1 P)

• • •

• • •:: • • •

P2-S

(2-1-2 P) t • • • • • • •

• • • •

P2-7

(2-1-2-1-2 P)

J • ; • • • •

• • • eV>

• • • •

Nolr:(«Rx ..p)=(.Rnwsx ••Pile';).

( • X • • y . • • P)-(»PI les X ••PI Ics X • • — Is slifllriK dlroclion of pllrs.

PI Ir-s) P2-4

(4-3-4 P)

r2-8

(2-1-2-1-2-1-2 P)

282
s age I — —

- S age n —

- "H" — — rJ — —

1

1

. _ i

_

1 — 1- —

1

1 — D ^

J

.

10 20 ;30 -lU 50 60 70 80 90 100

I Displacement (mm)

(a) Load ~ D isplocement

Q5 4035

CU 30

qI 25

oter 20 o 15

a> CT> 10 o

Q_

5

toQJ 0 O

X -5

) — p

P

1 ^

5 -

10 20 30 40 50 60 70 00 90 100 Displocement (mm)

(b) Excess Pore Water Pressure~Displacement

Figure 16. Example of relation between total load, excess pore water pressure and displacement of pile model

Pile ModelRu = 1.00

P4,8 •GL- 5cm

20 0cm

P3. 7 •GL-I5cm

P2,6»GL-25cm

GL-35cm Pl,5«" ^ij = 20 0gf/cm' GL-^Scm

Ru = 0,90

21 9cm

-vAlS.O gl/cm'

Ru = 0.45 Ru = 0.00

0,0 cm

j,9.0 gf/cm^^"-^|ggjM0.0 qt/cm'

Figure 17. Identification of excess pore water

pressure ratio at tip of pile model

283
6,0

40

20

00

Lattice Arrangement

3 Rows x 3 Piles Ru ^ 0 90

I o Measurement

I — Average

1.0 20 30 40

Sliding Velocity (mm/sec)

(a) Ru = 0.90

50

50 0

400

o 30 0

20,0

100

0,0

Lattice Arrangement

3 Rows X 3 Piles H

Ru = 00 0 • • • o o • • • o

°

' Meosurement

o • • • o O 0 o o o

1 — : Averoge o o o o o

10 20 30 40 5.0

Sliding Velocity (mm/sec)

(b) Ru = 0.00 Figure 18. Relation between total load acting on piles

and sliding velocity of piles In the case of Pl-7

50 0

400

o 30,0

200'^

100

o • • o o

o • o o

o • • o o

o • • o o

o • # o o

Lattice Arrangement

5 Rows X 2 Piles 4 5mm/s Measurement 4 5 mm/s Average

2 6 mrn/s Measurement

2 6 mm/s Average

1 5 mm/s Measurement

1 5 mm/s Average

00 02 04 06 08 10 Excess Pore Water Pressure Rotio (a-1) Lattice arrangement:

Pl-9 (5 Rowsx 2 Plies)

00 0 2 0,4 06 0 8

Excess Pore Woter Pressure Ratio (b-1) Zigzag arrangement:

F2-6 (3-2-3 Piles)

Lattice Arrangement

2 Rows X 5 Piles 4 5 mm/s Measurement 4 5 mm/s Average

: 2 6 mm/s Measurement

2 6 mm/s Average

5 mm/s Measurement

I 5 mm/s Average

00 02 0,4 06 08

Excess Pore Water Pressure Rotio (a-2) Lattice arrangement:

Pl-8 (2 Rowsx 5 Piles)

400

30.0

20 0

10,0

Zigzag Arrangement C2-1-2-1-2) Plies ° 4 5 mm/s Measurement 4 5 mm/s Average

" 2 6 mm/s Measurement

2 6 mm/s Average

• 1 5 mm/s Measurement

I 5 mm/s Averoge

00 02 04 06 08

. Excess Pore Water Pressure Ratio (b-2) Zigzag arrangement:

P2-5 (2-1-2-1-2 Piles)

Figure 19. Relation between pore water pressure

at tip of pile and total load acting on pile

284
Horizontal Bearing Capacity of Pile Foundations Based on

Horizontal Loading Tests

by

Minoru Fujiwara', Michio Okahara^ Shoichi Nakatani^Yoshitomi Kimura\

and Shitgeru Takagi^

SUMMARY

At present, the design standard used for structures Is shifting from the working stress design method hitherto In use to the limit states design method. In accordance with the limit states design method, a limit state Is defined so that It can be verified that a structure can

maintain required functions; then, It Is checked

that the structure possesses required safety

against external forces such as load and that It doesn't reach limit states during the design's

life span. Limit states which are taken into

consideration In stability calculations are the

elastic limit, which Is the reversible limit of the

behavior of a foundation, and the ultimate limit,

where deformation progresses and reaches a

tilting condition. In order to apply the limit states design method to pile foundations, It Is necessary to clarify these limit states that must be checked In the design and to establish models

for analyses which represent such states.

In this paper, the data of horizontal loading

tests of various plies are analyzed In order to elucidate the horizontal elastic limit, and an

estimation formula to obtain the coefficient of

horizontal ground reaction used for a stability

calculation of pile foundations within the elastic

limit Is proposed. Furthermore, In order to clarify the behavior of pile foundations during

large-scale deformation exceeding the elastic

limit, ground reaction, effect of pile group, and

decrease of flexural rigidity of the pile body

are examined by analyzing the results of

horizontal loading tests.

KEY WORDS: Pile Foundations, Horizontal Loading

Tests, Horizontal Bearing Capacity, Elastic Limit,

Coefficient of Horizontal Ground Reaction,

Ground Reaction

1. INTRODUCTION

At present the design standard of structures

In overseas countries is making a shift from the

working stress design method hitherto In use to the limit states design method or the load

resistance factor design method Triggered

by the revision of "Standard Specifications for

Concrete" (Japan Society of Civil Engineers,

1986), various institutions In Japan are also

examining adoption of a new design method. In accordance with the limit states design method, a

limit state Is defined so that It can be verified

that a structure can maintain required functions;

then, It Is checked that the structure possesses

required safety against external forces such as load and that It does not reach limit states during the period of design durability. Since

the functions that a structure must be equipped

with are expressed as the limit state in accordance with this method, design purposes can

be clear. In addition, with respect to the safety

of structures, this design method Is believed to realize more practical designs, due to the fact

that various uncertain factors In the design can be evaluated using the theory of reliability, and

that the coefficient of load as well as the

coefficient of resistance can be employed as coefficients of safety.

In order to introduce this limit states design

method to the design of pile foundations, It Is necessary to determine the limit state which must

be checked In the design stage and to establish a model for stability calculation which can represent the limit state with high precision.

Analyses on the horizontal behavior of elastic

foundations such as piles can be divided Into (1) ultimate ground reaction method, (2) elastic

ground reaction method and (3) compound ground

reaction method, depending on the treatment of

ground resistance (ground reaction, p) surrounding the piles.

(1) Ultimate ground reaction method

This method estimates horizontal resistance

only from the balance with external force

acting on the piles, supposing the distribution

of the ground reaction at the ultimate

condition of the ground. Since the given

ground reaction Is Irrelevant with the

displacement, and since the flexural

deformation of the piles Is not taken Into consideration, the deformation at the ultimate

resistance can not be estimated by this method.

(2) Elastic ground reaction method

This method expresses the ground reaction,

p, by the function, p=p(x,y), which represents

the relationship between the depth, x, and the

amount of deflection of piles, y. Depending on the treatment of p(x,y), this method can be

• 1 Director, Structure and Bridge Department,

Public Works Research Institute, Ministry of

Construction

• 2 Head, Foundation Engineering Division,

Structure and Bridge Dept., PWRI, MOC.

»3 Engineering Affairs Management Section,

Minister's Secretariat, MOC.

•4 Research Engineer, Foundation Engineering

DIv., Structure and Bridge Dept., PWRI, MOC.

»5 ditto

285
further divided Into the linear elastic ground

reaction method, the non-linear elastic ground

reaction method, and the p-y curve method.

With respect to the constant of ground spring

used In the linear elastic ground reaction

method, the so-called reverse operation value

calculated based on the data obtained from

loading tests Is used In general. The

non-linear elastic ground reaction method

unllnearly treats the relationship of ground

reaction and displacement In order to take the

non-alignment of the resistance of the ground

Into consideration. The p-y curve method Is to estimate the relationship of ground reaction

and the amount of displacement based on

examinations, etc., and to conduct numerical

analyses.

Since these methods are based on the

supposition that the ground is elastic, neither

the ground reaction at the ultimate state nor

the ultimate bearing capacity can be obtained.

(3) Compound ground reaction method

This Is a method to conduct analyses on the

supposition that the ground near the surface

which Is largely transformed is In a plastic area and that the lower ground which Is less transformed Is In an elastic area. Consideration (1) Is used for the ground resistance in the plastic area, whereas consideration (2) Is used for the ground resistance in an elastic

area. Modellzatlon of the ground Is still shelved for future examination; however, this

method Is best able to calculate the actual

behavior of foundations with high precision.

As described above, a variety of methods are

available for analyses of pile foundations In the

horizontal direction. Nevertheless. the choice of

method in the design of pile foundations Is closely related to the range of amount of

displacement that Is assumed In each design.

According to the concept of design

prescribed In "Specifications for Highway

Bridges, Part IV, Specifications for Substructures'"" (Japan Road Association, 1990) which

Is used as a technical standard for substructures

of highway bridges (herein after referred to as "Specifications"), concerning the level of design

load employed at present, a foundation should be

sufficiently safe against bearing force, and "any

residual displacement that Is harmful to the

foundation should not occur." In short, pile

foundations are designed based on an analysis. In which the pile Is converted Into linear spring

constant, so that both the displacement and the

reaction of the pile head obtained by the

analysis are controlled within the range where

the foundation behaves elastlcally.

Concerning the limit states design method, on the other hand. It Is necessary to keep trying to clarify the limit state that must be checked In the design. In addition to the concept described

above, "any residual displacement that Is harmful

to the foundation should not occur," another

concept is also proposed; "foundations should be

designed to resist against a load which acts upon

It quite rarely during the period of design's

lifetime, such as a huge earthquake, so that any

destructive damage to the structure, such as a

fall of bridge, does not occur." (see Figure

1.1)^'. Thus, an ideal model for a stability

calculation used for the design Is a model which can be applied to a large-scale deformation

exceeding the range where the foundation

behaves elastlcally, or a model which can be used

for a calculation of horizontal durability of the

foundation. In this case, the non-linearity of

the ground resistance and of the flexural

rigidity of piles can not be Ignored.

Based on these circumstances, horizontal

loading tests were conducted and their data were analyzed in this paper in order to clarify the

horizontal behavior and the horizontal bearing

capacity of pile foundations. The data obtained

from many horizontal loading tests on a single

pile were analyzed statistically In order to grasp quantitatively the displacement at the

limit where the foundation is considered to behave elastlcally from a viewpoint of design,

and an estimation formula of the coefficient of

horizontal ground reaction used for the design

within the domain of this displacement was proposed. Furthermore, in order to clarify the

behavior of pile foundations In the domain

exceeding this displacement, data obtained from

horizontal loading tests on both a single pile

and pile groups were analyzed; then the

non-linearity of the ground resistance and of the

rigidity of piles, and the Influence of the pile

group were examined.

2. LOADING TEST DATA OF FILES USED

The data used for the analyses in this paper

are horizontal loading test data filed with

prefectural authorities and other Institutions

such as the four public highway corporations,

and which the Public Works Research Institute

collected in 1988. The total number of the

collected loading test data was 415°'; however,

data used in this paper were selected from these

415 data based on the content of the loading

tests and the substantiality of the data. The

quantity of data used In this paper Is as shown In Table 2.1. In addition, depending on the content

of examination, efficient data are selected

occasionally from the data shown In Table 2.1. In order to examine the behavior of pile

foundations during large-scale deformation, the

data of loading tests which cover large-scale

deformation were selected from the above data,

and the case with a single pile was analyzed. As

for the Influence of a pile group, results

obtained from model pile tests conducted In the

test pit at the Public Works Research Institute as

286
well as from an In-sltu pile foundation test conducted by Metropolitan Expressway Public

Corporation were used.

3. HORIZONTAL ELASTIC LIMIT OF PILES

3.1. Definition of Horizontal Elastic Limit

The load-displacement curve of the pile head,

when the horizontal load acts upon the pile

foundation, shows strong non-linearity. For

example. Figure 3.1. shows results of horizontal

loading tests^' on a pile foundation using 9 steel

piles (3x3 piles) with a diameter of 318mm and

length of 1.5m. Figure 3.1. shows that the

relationship between load and displacement is non-linear from the Initial stage of loading, and

that residual displacement remains when loading

Is removed. This Is considered to be attributable

to the non-llnearlty of the deformation characteristic of the earth Itself and the consecutive

progress of the plastic area of the ground along

with the increase of displacement from the

surface toward the direction of the depth.

Therefore, It Is difficult to define the elastic

limit In a strict sense. even If the load is controlled at a low level. Figure 3.2. shows an example of the

relationship between the residual displacement of

the pile head at every loading cycle In horizontal loading tests of piles and the ratio of

displacement of the pile head against the

diameter of the pile prior to the removal of

loading (strain of the pile head. Sy/D). This

figure shows that the residual displacement

Increases rapidly from a certain level of

displacement. If the displacement of the pile

head during loading can be controlled within the

point, Sry, where the residual displacement makes

an abrupt Increase, huge residual displacement of

the foundation can be prevented. In other words.

If the foundation can be designed so that

displacement Is controlled within this level of

displacement even when a short-term loading,

such as an earthquake, acts upon the foundation,

the foundation Is usable without any repair work

after the loading Is over, since there remains no

major residual displacement. Thus, the elastic

behavior in an engineering sense, in other words,

the limit of displacement which is the amount of

residual displacement negligible in the design was defined as the horizontal elastic limit, and

its quantitative evaluation was conducted.

The evaluation Is conducted based on the

various data obtained from horizontal loading

tests Introduced In Chapter 2. In addition to the

point, Sry, where the residual displacement makes

a sudden increase, the point, Sy, which Is equivalent to the yield load on the distribution curve of the Welbull function defined In Figure

3.3. was studied. A comparison between the data obtained from

horizontal loading tests using a cast-ln-place

pile and other piles, such as a steel pipe pile,

proved that there was peculiarity In the form of

load-displacement curve. The following two

points stated below are the reasons for this:

(1) In the case with a cast-ln-place pile, a bending crack occurs on the pile body at a relatively small load level. Accordingly, the

flexural rigidity of the pile body changes and

the non-llnearlty of the pile body appears.

(2) Since a cast-in-place pile loosens the

surrounding ground at its placement, the

Influence of ground slackness appears until the

surrounding ground reaches the original stress condition of the natural ground.

Thus, when an analysis Is conducted, the

displacement Index of the Welbull curve formula,

m. Is made variable without fixing It to I, In order to Improve the conformity of the distribution curve of the Welbull function for

cast-ln-place piles (In the case of other piles, m = 1 works sufficiently to obtain approximate

values).

The point. Sry. where the residual dis- placement makes a sudden Increase and the yield

point. Sy, on the Welbull curve were compared

using the data obtained from loading tests. Figures 3.4. (a) and 3.4.(b) show the results of

this comparison In the cases of a steel pile and a cast-ln-place pile, respectively. Since they show

a comparatively favorable correlation, there may

not be any problem In considering that the point

where residual displacement suddenly Increases roughly corresponds with the yield point on the

Welbull curve.

3.2. Modulus of Deformation of Ground and Yield

Displacement

The horizontal elastic limit described above

Is considered to be related to the flexural

rigidity of piles and the modulus of deformation

of the ground, Eo. in addition to the diameter

and kind of plies. Figure 3.5. shows the

relationship between the modulus of deformation

of the ground. Eo. and the yield displacement,

Sy, which were estimated by approximating the

load-displacement curve using the distribution curve of the Welbull function. On the other

hand. Figure 3.6. shows the relationship between

the yield strain, (Sy/D) and the modulus of

deformation of the ground. The modulus of

deformation of the ground, Eo, Is estimated to be

Eo=7N based on the N value of the standard

penetration test. Thus, the following points stand:

(1) The correlations between Sy and Eo, as well as between Sy/D and Eo vary greatly across the

board.

(2) There Is a general trend that Sy and Sy/D

decrease as the modulus of deformation of the

ground increases. Nevertheless, this trend is

287
not apparent in the case of Sy/D . (3) Values of Sy and Sy/D are large In the case of

steel pipe piles. PC piles and PHC piles;

however, they are small in the case of

cast-in-place piles. This trend Is especially

remarioble with Sy/D.

3.3. Statistical Characteristics of Yield

Displacement

As described above, since Sy and Sy/D do not correlate strongly with the modulus of deformation of the ground, statistical characteristics

of Sy and Sy/D are analyzed regardless of the

modulus of deformation of the ground.

Figures 3.7. and 3.8. show the distribution of

Sy and Sy/D. Both of these figures are in favorable conformity with the logarithmic normal

distribution. Figure 3.9. show distribution of Sy

and Sy/D of various Itlnds of piles , which was approximated by the logarithmic normal distribution. Tables 3.1. and 3.2. summarize statistical

characteristics of Sy and Sy/D by type of piles.

Thus, the following points stand:

(1) Coefficients of variation of Sy and Sy/D

exceed 100%, and they scatter greatly.

(2) The mean value of Sy is 26mm In the case of

steel pipe piles and 14mm In the case of PC and

PHC piles. On the other hand, the mean value of

Sy/D is 4.3% in the case of steel pipe plies and

3.7% in the case of PC and PHC piles.

(3) With respect to cast-in-place piles, values of

Sy and Sy/D are small and greatly different

from the cases with other piles. For Instance,

the mean value of Sy/D Is 1%. This Is not due to the fact that the horizontal resistance of

cast-in-place piles is small. This is rather due

to the fact that the occurrence of a cracit in the pile body Influences the determination of

an yield point.

3.4. Horizontal Elastic Limit

When piles receive huge horizontal load

exceeding the yield load expressed by the

load-displacement curve, deformation of piles

suddenly increases and the stability of pile

foundations is damaged, as mentioned above.

Accordingly, one of the states that must be

checlted In the design is the limit state where the

deformation of pile foundations Is kept on the

side of the so-called elastic area. Nevertheless,

since the yield displacement varies a great deal,

the problem is to what extent this variation

should be talten into consideration In order to reasonably establish the horizontal elastic limit. The "Specifications" (1990) prescribes that

the allowable displacement should be 1% of the

pile diameter (minimum 15mm), based on the results

analyzed In this paper; however. It might be more

reasonable to establish the elastic limit for each one of the piles depending on its respective

characteristics. For instance, in the case of

steel pipe piles, the allowable displacement can be around 2% to 3% of the pile diameter, since

their deformation function is large, while in the case of cast-in-place piles, it can be around 0.7%

to 1.0% of the pile diameter, since, in contrast,

their deformation function Is small.

4. STABILITY CALCULATION METHOD OF PILE FOUNDATIONS WITHIN THE ELASTIC LIMIT

4.1. Model for Stability Calculation

As defined in 3.1., If the displacement of a foundation when load Is acting on It Is within the

elastic limit, the displacement of the foundation

is comparatively small and the residual

displacement is on the small level. Therefore,

within this sphere, no major error arises even If a stability calculation Is conducted by replacing

the pile with a linear spring. Therefore, the

"Specifications" prescribes that a stability

calculation of pile foundations should be

conducted In accordance with the elastic analysis

method which gives consideration to the

displacement of the footing (vertical, horizontal

and rotary displacement), under the condition

that the footing Is considered as a rigid body.

On the supposition that a pile is linear

elastic when pushed in, pulled out and bended,

each pile is modellzed into a support spring

which Is of equivalent value at the pile head as

Is the case with the analytical model Illustrated

In Figure 4.1. In short, the stability calculation

of pile foundations is replaced with the stability

calculation of a rigid body (the footing)

supported by many supporting springs which

represent a pile group. The supporting springs

consist of the spring constant In the direction of

the pile axis, Kv. and the spring constant

perpendicular to the pile axis, K1-K4, as defined

below:

Kl, K3: The perpendicular to the axis force

(tf/m) and the bending moment (tf m/m)

which should act upon the pile head, when

the pile head is made to displace by a unit

amount in the perpendicular to the pile

axis direction without causing the pile

head to rotate.

K2, K4: The perpendicular to the axis force

(tf/rad) and the bending moment (tf m/rad)

which should act upon the pile head, when

the pile head is made to rotate by a unit

amount without causing any shift to the

pile head.

These spring constants of the pile are obtained as summarized in Table 4.1., based on

the relationship between the load and dis- placement calculated using the theory of beams on the elastic foundation which employs the

coefficient of horizontal ground reaction, kH. Thus, evaluation of the coefficient of horizontal

288
ground reaction Is an Important Item In the

stability calculation of pile foundations within

the elastic limit. Estimation methods of the

coefficient of horizontal ground reaction is examined below based on the data obtained from

horizontal loading tests,

4.2. Coefficient of Horizontal Ground Reaction

Obtained by A Reverse Operation

Coefficient of horizontal ground reaction

obtained from horizontal loading tests Is referred to as the reverse operation k value.

This Is calculated In the following manner: Displacement at pile head, S, In loading tests

Is obtained by the formula of beam on the elastic

foundation, as stated below:

where,

P: Load on the pile head (kgf)

S: Horizontal displacement on the pile head (cm)

El: Bending rigidity of the pile body (kgf cm^)

0 : Characteristic value of the pile (cm~')

>9 = 4 /kH D / 4EI

D: Pile diameter (cm)

In short, 0 Is obtained by substituting P obtained from loading tests, where the standard

displacement was So, for the formula(4. 1). Then

kH can be obtained using this $

.

4.3. Coefficient of Horizontal Ground Reaction

Obtained by Estimation

The estimation formula for the coefficient of

the horizontal ground reaction of piles

prescribed In the 1980 edition of the

"Specifications" was as follows:

k = koy~'^^ (4.2) ko= a Eo D"^-'^ (4.3) where,

k: Coefficient of horizontal ground reaction of

the elastic foundation (kgf/cm^)

ko: Coefficient of horizontal ground reaction

(kgf/cm3) In the case where the displacement

on the surface of the design ground Is determined to be 1cm

Eo: Modulus of deformation of the ground

obtained by the method shown In Table 4.2. (kgf/cm^^) a : Coefficient shown In Table 4. 2. which

corresponds to the method to obtain Eo

D: Pile diameter (cm)

y: Displacement on the surface of the design

ground that must be taken Into consideration

In design (cm) As proved by the formulas (4.2) and (4.3), the

coefficient of the horizontal ground reaction Is expressed by a function which consists only of

ground strength, displacement and pile diameter,

In which the Influence of the deformation mode In

the vertical direction of piles Is ignored. As a

result, the k values were estimated to be

comparatively small In the case of pile

foundations using piles of large diameter for

their lengths, and this was found to be In adequate when applied to actual cases. Furthermore, the k values did not conform with those of

other foundations, such as rigid foundations.

Thus, In order to settle these problems, an

estimation method of the coefficient of

horizontal ground reaction of elastic foundations Is proposed as follows:

When the modulus of deformation of the

ground Is obtained from soil tests, etc., the

coefficient of horizontal ground reaction Is calculated in the following manner: kH = kilo (Bh/30)"^'"' (4.4) where,

kH: Coefficient of horizontal ground reaction

(kgf/cm^)

kHo: Coefficient of horizontal ground reaction

(kgf/cm^) which Is equivalent to the value of

plate bearing tests using a rigid circular

plate with diameter of 30cm. In the case where It Is estimated from the modulus of

deformation obtained by various soil tests and investigations, the below stated formula

Is used. kHo = 1/30 a Eo (4.5) Bh: Equivalent loading width of the foundation

(cm) which crosses at right angles with the

load direction. With respect to the ground

which is Involved with the horizontal

resistance of the elastic foundation, 1/ 0

from the design surface of the ground i s taken into consideration in general (= Jd/ 0 ). Eo: Modulus of deformation of the ground

(kgf/cm^) in the position taken into

consideration in the design, which is measured or estimated by the method shown

In Table 4.3. a : Coefficient used for the estimation of

modulus of ground reaction, which Is shown

In Table 4.3. 0 : Characteristic value of the foundation

(cm"').

Although the structure of the calculation

formula of the coefficient of horizontal ground

reaction of the elastic foundation proposed here

Is basically the same as the calculation formula

of rigid foundations hither to in use, this proposed formula takes into consideration the

Influence of elastic deformation of the pile body.

This Is due to the fact that the larger the

elastic deformation of the pile body becomes In comparison with the horizontal force acting on

the upper part of foundation, the more the

ground reaction that contributes to the

horizontal resistance concentrates in the upper

layer of the ground. Thus, based on both

experimental and empirical judgment, the area of

289
the ground resistance that contributes to the

horizontal resistance was determined to be D x

1//9 (D: width of the foundation. l/y9: length of

the ground In the direction of depth which is In relation with the horizontal resistance); then

the equivalent loadin g wid th of the foundation

was determined to be Jd/W. This Idea can settle

the contradiction concerning the calculation of

the k values and problems due to difference In the type of foundations.

4.4. Comparison Between k Value Obtained From

The Estimation Formula and Reverse Operation

k Value Obtained From Loading Test

Estimation precision of formula(4.4) Is examined based on the data obtained from loading

tests of plies. Figure 4.1. summarizes a comparison between kH which was estimated using

formula (4.3) following the 1980 edition of the

"Specifications" and the reverse operation kH

with standard displacement of 10mm. On the other

hand, Figure 4.3. summarizes reverse operation kH

obtained under the condition that the standard

displacement was IX of the pile diameter and kH

obtained using formula (4.4). In addition,

Flgures4.4. and 4.5. show the logarithmic normal

distribution of the ratio of the reverse operation kH to the estimated kH. Since more than

90% of data used is for piles with less than

1,000mm of pile diameter (about half the data are

using piles with a diameter of less than 600mm).

precision of the k value obtained using the

proposed formula Is not very Improved In comparison with that obtained in accordance with

the 1980 edition of the "Specifications." This is attributable to the fact that the proposed

formula Improved the adaptability to k values of

rigid foundations and pile foundations with large

diameter. It Is Important to recognize that an estimation formula of k values might result In some dispersion to this extent.

5. BEHAVIOR OF PILE FOUNDATIONS DURING

LARGE-SCALE DEFORMATION

5.1. Influence of The Non-Linearity of The Ground

Resistance

On the supposition that the characteristics

of ground reaction are linear elastic when the

deformation of pile foundations Is small and

within the elastic limit, as discussed In 4.1., the

characteristics of ground resistance are modellzed by using the coefficient of horizontal

ground reaction. kH. This supposition Is con- sidered to be adequate for the design In the

range where the displacement Is small; however,

the non-llnearlty of the ground reaction must be

taken Into consideration, since the errors become

great during large-scale deformation.

With respect to the relationship between the

ground reaction, p, and the displacement, y, at each depth of the ground, the p-y curve Is used

by DNV*^' as shown In Figure 5.1. it would be

Ideal to modelize accurately the characteristics

of the ground reaction in order to express the

behavior of foundations with high precision.

However, when the precision of the ground

surveys and simplification of calculations in the

stage of design are taken Into consideration, it Is not necessarily appropriate to apply this Idea

to general bridges. Thus, in this paper, the

characteristics of ground reaction are modellzed

as a bi-linear model, and the effectiveness of

this analysis method was examined based on the

data obtained from horizontal loading tests. As for the data used for analyses, data of

piles, the pile was deformed more than 30% of the

pile, were selected from the data of loading tests collected in Chapter 2. As for the constant of

ground resistance, the coefficient of horizontal

ground reaction, kHs, and the maximum ground

reaction, pus, obtained by the below-stated

formulas are used as parameters. and a combination was selected so that errors of the

load-displacement curve and the distribution of

bending moment of piles can be minimized.

kHs = a k X (kH prescribed In the

"Specifications") (5.1) pus = a p X (Passive earth pressure strength of

Coulomb) (5.2) The analyses covered the test results In which the horizontal displacement at the ground

surface was up to 10% of the pile diameter.

Variation was given to the parameters stated

above in order to obtain the difference from the

analytical results. As an example, figure 5.3. shows a contour map of errors that arise In a

test on a steel pipe pile with diameter of 61cm

and length of 21m which was placed In the sand

ground. In this case, the errors are recognized

to be the minimum, the supremum value of the

ground reaction being 3.5-foid of the passive

earth pressure strength and the coefficient of

ground reaction being 1.5-fold of the value

obtained In accordance with the "Specifications." As discussed in Chapter 4.. the coefficient of

horizontal ground reaction. kH, prescribed In the "Specifications" is determined under the

condition that the displacement of the pile head

Is IX of the pile diameter. On the contrary, the

coefficient of horizontal ground reaction used

for the elasto-plastic analyses expresses the

characteristics on a smaller level of displacement. Thus, the value of a « exceeds 1. On the

other hand, with respect to a p, the area of

ground resistance spreads three-dimensionally;

thus. It Is presumed that the plastic area became

around three times longer than the width of the

pile. Figure 5.4. Illustrates the loaddisplacement curve of this situation, while

Figure 5.5 shows the distribution of bending

moment of piles. These two figures accord with

290
each other quite favorably; therefore, It was proved that the behavior of piles can be

expressed by determining an adequate constant, even when a simplified model Is used.

The above result was commonly recognized In the case of other test piles; however, the values

of parameters of the ground resistance, a k and

a p, varied depending on each test, as shown In Table 5.1. Despite the fact that the effectiveness of this analytical method was proved,

further studies are needed for better evaluation

methods to obtain constants used for the design,

and more analytical data are needed for that.

5.2. Influence of The Pile Group

A pile foundation used as a bridge foundation

rarely consists of a single pile. In general,

many piles are put together by footing to form a

pile foundation. When the piles deform greatly

due to the horizontal force acting on this

footing, each pile Interferes with another

through the aid of the ground. Therefore, the

ground reaction of each pile differs depending of

the location of each pile. Nevertheless, the

degree of such Interference has not been

clarified yet since It differs a great deal

depending on the kind of piles, pile Intervals

and the soil condition. The "Specifications"

prescribe that the minimum central Interval of

the piles should be In principle longer than 2.5 times the pile diameter. In this case, the same value of coefficient of horizontal ground

reaction for a single pile Is used for all the

piles which form a pile group. Recently, many

Institutions have conducted experiments on these

problems in order to grasp the mechanisms of the

horizontal resistance of plies. In this paper, the

Influence of a pile group was examined based on horizontal loading tests using steel pipe piles

conducted by the Public Works Research Institute

and Metropolitan Expressway Public Corporation.

The Public Works Research Institute^' con- ducted loading tests of the pile group In a test

pit. Piles used for these tests were a single

pile, series piles, parallel piles, and a 9 piles

group aligned In a square 3»3 ( of the model steel

pipe piles Is 101.6mm) as shown In Figure 5.6. (also see Figure 5.7.). On the other hand. Metropolitan Expressway Public Corporation conducted

horizontal loading tests, establishing a 9 plies

group In situ using steel pipe piles of natural

size (with of 318.5mm) as shown In Figure 5.8. Figure 5.9. shows the curve of average

loading (average load per pile) and footing

displacement obtained from the tests conducted

by the Public Works Research Institute. The

figure shows that the load-displacement curve Is located lower than the curve of a single pile

when the displacement Is great. This Is attributable to the Influence of the mutual

Interference of piles. The degree of Influence Is

the greatest for the 9 piles group. It decreases

respectively In the order of series piles and

parallel piles, then approaches that of the

single pile. This Is due to the fact that the share

of horizontal reaction against the horizontal

force of a pile groups different depending on the

location of piles within the same footing.

Figure 5.10. shows the share ratio of

horizontal reaction carried by the front and

rear piles to that carried by the central pile in the case of the 9 piles group. Both of the tests proved that the front piles carry the largest

load, and that the load share becomes smaller

respectively In the order of the central pile and

the rear plies. This trend becomes more apparent

as the load becomes larger. A simulation concerning the difference of the

reaction share of each pile in a pile group and

the behavior during large-scale deformation was attempted for the purpose of establishing a new

design method. In the simulation, the nonlinearity of the p-y characteristics of the

ground, which becomes an important subject In the case of large-scale deformation, was dealt

with bl-llnearly as examined In 5.1. in addition,

in order to consider the difference of load share

of each pile, characteristics of the ground

resistance, (pus and klls). In which a p and a k of

different values were used as parameters, were

established on a trial basis for each position of

piles. Then the values of a p and a k were

selected so that the experimental values and

analytical values coincide with each other most favorably in terms of the following four Items:

1) load-displacement curve

2) distribution of displacement

3) distribution of bending moment

4) distribution of ground reaction.

The displacement level to be covered by the

simulation was until the pile body begins to

yield. In this paper, the simulation Is carried out until the displacement becomes around 50% of

the pile diameter.

Figure 5.11. shows a selection of a p and a r which are in conformity with the experimental

values concerning the load-displacement curve of

the footing part of the 9 piles group. It is proved that a simulation can be conducted with

quite high precision If appropriate values can be

set up. Figures 5.12. and 5.13. show respectively

the distribution of bending moment and the

ground reaction In the case of a combination of

a p and a k that brought about the best

conformity In the case of the 9 piles group in Figure 5.11. These values also prove that the

experimental values and calculated values are in favorable conformity with each other. In addition, the same level of conformity was obtained

for the distribution of pile displacement.

Values of a p and a k were also selected by

simulation for other cases In order to get such

conformity, the result of which is shown in Table

291
5.2. In the case of series piles, the front piles

showed a value of « p that was on the same level

as the case with a single pile, and the values of rear piles decreased gradually. In comparison

with the single pile, the values of « p and a k of

parallel piles decreased as the number of piles

In the parallel direction Increased. In the case

of the vertical and batter pile group using 9 piles, which Is a combination of series and

parallel piles, the values of a p and a k showed a decrease more than with the cases where series

piles and parallel plies were used separately.

These phenomena are perhaps attributable to the

decrease of the coefficient of ground reaction

due to the Increase of equivalent load width and

the decrease of passive resistance due to the

mutual overlapping of the area of resistance of

each pile.

5.3. Influence of The Non-LInearlty of Flexural

Rigidity of The Pile Body

Since the piles used for the loading tests which were dealt with by the analyses In 5.1. and

5.2. were commonly steel pipe plies, and since the

range of displacement covered by the analyses

was until the pile body begins to yield, the pile

body Itself Is treated as an elastic body. On the

contrary, with respect to the range exceeding

the yield point, the Influence of non-llnearlty of

the flexural rigidity of the pile body becomes

remarkable In the cases of both cast-ln-place

piles and steel pipe piles.

In general, the relationship between the

bending moment, M, and the curvature, * , when

the bending moment acts upon a concrete pile Is recognized at the beglnlng of cracks of concrete

(C), at the beglnlng of the yield of tensile

reinforcement (Y) and at the ultimate phase of

section (U). This Influence was determined to be

taken Into consideration In analysis of loading

tests, and Is modellzed by a tri-llnear curve

which goes through the points obtained for the

circular reinforcing concrete section which

receives the axial force, N, and the bending

moment, M (Figure 5.14.). For an example, a case analysis'"' concerning

loading tests of cast-In-place piles Is Intro- duced.

Figure 5.15 shows the load-displacement curves measured during the loading test and

obtained by analysis. At the analysis, the

characteristics of ground reaction are modellzed

as a bl-llnear model. At small displacement, the

calculated values In which the pile body Itself Is treated as an elastic body. Is coincident with

experimental values. However, at large

displacement these values are different. On the

other hand, the calculated values In which the

non-llnearlty of the flexural rigidity of pile

body Is taken Into consideration, Is coincident

with experimental values at large displacement.

As so far discussed. It was verified that the

behavior of pile foundations during large-scale

deformation could be simulated If the Influence

of ground resistance, the non-llnearlty of the

rigidity of the pile body and the effects of a pile group are taken Into consideration. With

respect to the analytical model used for design

calculations of general pile foundations, further

studies will be conducted along with the

displacement level that must be taken Into consideration In the design.

6. CONCLUSION

An attempt was made to Introduce the limit states design method to the design of pile

foundations. In order to grasp the behavior of

piles and to establish an analytical model,

analyses were conducted on horizontal loading

tests. The results of this study can be summarized as follows:

1) Based on the Idea that the elastic limit of pile

foundation is the limit where the reversibility

Is guaranteed, the point where the residual

displacement does not make a sudden increase

is determined to be the elastic limit In the

design. This elastic limit is roughly In conformity with the yield point on the

load-displacement curve.

2) The coefficient of fluctuation of the elastic

limit described above exceeds 100%, and

scatters a great deal. The elastic limit Is 3.5% of the pile diameter In the case of the

entire average of all piles, 4% In the case of

steel pipe piles, and 1% In the case of

cast-ln-place piles (all these figures are based on the logarithmic normal distribution).

In the case of cast-ln-place piles, there Is a possibility that the occurrence of cracks to the pile body might give influence to the

determination of the yield point.

3) A calculation formula of the coefficient of

horizontal ground reaction used for stability

calculations within the elastic limit was proposed. Giving consideration to the influence

of the deformation mode In the direction of

the depth of plies, the equivalent load width

of the foundation was proposed to be •/ D//3 . The coefficient of horizontal ground reaction

prescribed In the 1980 edition of the

"Specifications" was under the condition that

the standard displacement was 10mm; however,

the coefficient of horizontal ground reaction

In this paper was determined to be calculated

when it was IX of the pile diameter following

the Idea described In 2). It was confirmed

that the estimation precision of the coefficient of horizontal ground reaction

obtained by these methods was on the same

level as the 1980 edition of the

"Specifications."

4) It was confirmed that the behavior of pile

foundations during large-scale deformation

292
could be simulated, if the non-linearity of

ground resistance, influence of the pile group, and the non-linearity of the flexural

rigidity of the pile body are taiten into

consideration.

While this study was conducted, data of

horizontal loading tests of piles were offered by

many institutions. A part of this study has been

conducted as Joint research with the Japanese

Association for Steel Pipe Piles. The authors

would like to express their gratitude to many

researchers who cooperated with this study.

REFEI^ENCES

1) The American Association of State Highway and

Transportation Officials: Standard Specifications for Highway Bridges. 1983.

2) Ministry of Transportation and Communications,

Ontario: Ontario Highway Bridge Design Code

and Commentary, 1983.

3) Canadian Geotechnlcal Society: Canadian Foundation Engineering Manual (2nd Edition). 1985.

4) Japan Road Association: Specifications for

Highway Bridges, Part IV, Specifications for

Substructures, 1990. (in Japanese)

5) Technology Research Center for National Land

Development: Study on the Limit States Design

Method for Highway Bridge Foundation Structures (Part 3). March, 1991. (In Japanese)

6) 0]<ahara, Takagi. Nakatanl. Taguchl and

Sakamoto. Study on the Characteristics of

Horizontal Resistance of Plies Based on Loading Test Data: Technical Memorandum of

P.W.R.I., No. 2721. January, 1989. (In Japanese)

7) Ogasawara, Hanko. Gose and Kawaguchi: Study

on the Yield Strength of Pile Foundations.

Journal of Structural Engineering, Vol. 37A. March, 1991. (in Japanese)

8) Det Norske Varltas : Rule for the Design,

Construction and Inspection of Offshore

Structure, 1977.

9) Okahara, Kimura, Asama, Koyama : Behavior of

Laterally Loaded Pile Under Large Deflection,

8th U.S. -Japan Bridge Engineering Workshop,

1922.

10) Kimura, Nakatanl, Hamada, ito: Horizontal

Loading Test for Cast-ln-place pile In the

Field, Proc. of The 46th Annual Conf. of The

Japan Society of Civil Engineers, 1992.

11) Okahara, Takagi, Nakatani and Kimura: Study

on the Bearing Capacity of A Single Pile and

the Design Method of Cylindrical foundations.

Technical Memorandum of P.W.R.I., No. 2919. January, 1991. (in Japanese)

293
load. P

UltliuK load. Pu

ricid load. Py

or

Figure I. 1. Limit states for stability

calculations of foundations

Figure 3.2. A case in which the point where

the residual displacement of the ground increases suddenly is defined as the elastic limit

Table 2.1 Data obtained from horizontal loading Icsls

Type of Pile Number of date

Steel pipe pile 120

Cast in pile 29

I'C pile 31

I'llC pile 24

Total 204

(note) Steel pipe piles. PC piles and PlIC piles ere placed by the driven pile •ethod.

Displacement (to)

Figure 3. 1. Load - head displacement

(horizontal loading tests)

Legend: O - Beasured values — leibull distribution ctrve

field load; Py = r»a«(l-e I)

Horizontal displaceecnt

of the pile head, S Figure 3.3. A case in which the yield load on the Weibull curve is defined

as the elastic 1 imi

t

(a) steel pipe pile

Sry/D(%)

(>>) Cast-in Place pile

Sy: yield djsplaceoent defined

by the leibull distribution curve, ^s(aa)

Sry: rield displaceient obtained fro» the point there

the residual displaceoent increases suddenly (la)

D: Pile diajBeter (ea)

Figure 3.4. Comparison of yield strain deformation obtained from formula of the Weibull distribution curve and the point where

the residual displacement increases suddenly

294
0

0

1

0

i

'

B 0

0

0

°

O-dJi 0

0 °

0

40 4-

Figure 3.5. Correlation between Sy and E5o

Nuibcr ot dBla - 204

For Sy: lean value = 20.7m

Standard deviation 24. Obb"'

^1

\

jn n 0 10 20 30 to 50 60 70 BO 90 100

S, (nl

Figure 3.7. Probability density distribution of Sy

1 —

o o

B

r

o

+

i o

1 B

e

° ? * o

° o 8 9* Oo

S o °e °

' + + a o

o o

O O O oo oo

o o o

0 o oo

1

o

Figure 3. 6. Correlation between Sy/D and E5o

)0 1 0 20 3 0 to 5 0 6 0 70 80 9 0 10 0 110

S,/D (I)

Figure 3.8. Probability density distribution of Sy/D

- steel pipe piles (Probability not to eicced 15. 9«)

(b)

Figure 3.9. Probability density distribution

of Sy and Sy/D. expressed

by logarithmic normal distribution

295
Table 3. 1 Statistical value of S,

110

p!ica'«lle

II II 14 104

titcal () J.5-4I.S I.I -49.1 4.0 -44.1 1.1-124.1 •IB Tilt* («} :i. 1 10.5 14.1 D.l 10.1 SttnUti 4c*iaUoa (*•> 11.

1

10. 1 11.9 1. 9 14.0

C»ertlcl«at «f varlatloB (X) III It 71 tl III

Sr

llon-eice<4 yrobtlllllr

9 («) 4.0 1.0 1.1 4. e 1.0

10 <«) 9.1 1.7 4 1 9.t 4.1

I9.« (t) 7. J 1.4 9.1 I.I 9.4

30.0 (1) tl.O 9.1 l.t 1. I 14

Table 4.1. Sprlnj conalant In tht vertical tfirecUoo to the ailt of plle« (H > J)

Klgldly conntct«d pll* httft43« flln9a-conn«ctf4 plla h*«d«

h • 0 h - 0 h 0 0 h • 0

iicie' 4CIS'

jcib' leiB'

11 Bhl ' » 1 11 • Bhl' •O.S

X], rj ItIB" 0 0

M

4CIt ll*Bhl'<0.9

1 • Bh 11 < Shi ' • 1 ICIB 0 0

« 1 Characterljlic value of pllet • " 7 ieT '* '

X 1 >> '-J-

(•!

* ' Coefficient of horlzonlal jroum) reaction (tf/»2) 0 • Pile dieaeter (a) " ' Fleiural rinlilHy of piles (tf/a2)

h • Length in the atlal direction above the dcalgq (round of piles (a)

Table 3.2 Statistical value of S, /D

Rvaber of ••tie

Standtrd deflation

CocfMcleol of t»rlatlon(t>

Table 4.2 Relationships between Eo and a p of elastic foundation

Coefficient of varlillon of the |ro«nl t, (l|f/ct') •

Cocrtlclcnl of variation of the fromni aeatDred Inildo korlni porta 1.1

Coefficient of variation of gained kr aneoaflocd or triailtl ceaprcailea tott 0.1

Coefficient of variation of tatlaated frai l.'SM of R rilia of aiandard peaelratle* toal 0.}

t Abscissa

a Rotarr dlsplsccscnt

displscMcnt S , Vertical dieplaceaent

Figure 4.1. Analytical model in the stability calculation

of pile foundations

Table 4 3. E, and a

Modulus of detonnation Bo dtg/em')

o

Ordinary

stata

During

earthquake Modulus of dafonution obtained fron the repetition curve of 30on diameter plate loading test I 2

Modulus of deformation measured in a bored

hole 4 8

Modulus of deformation obtained from the unconfined coiif>re8sion teat or trlaxial compression

test of speciments

4 8

Modulus of deformation estimated by E| - 381) from N value of standard penetration test

I 2

(note) In esse of a tindstore. the values of ordinary occasions ihouU be oscd.

296
10.000 IS.OOO 20,000 :s.ooo

k value by the 1980 edition of the

'SoecificaliOTffi for RUhtBy BrldKCG* (tf/B3) Figure 4.2. Reverse operation k valueCdisplacement: lOmm) and estimated k value in the 1980 edition of the "Specifications"

For h value ratio; lean value = 1. 548 Stamlard deviation ' 1.480 Coerricient of variation • 961

-(.I -II -M -II -I.I II II M M 1.1 l.l

Figure 4.5. Distribution of Log {k value ratio : reverse operation k value (displacement: IX of pile

diameter) / k value in proposed formula)

5,000 10,000 15,000 JO.OOO 25,000

» value by the proposed fonula (tf/«3)

Figure 4.3. Reverse operation k value (displacement:

IX of pile diameter) and estimated k value

by the proposed formula

Initial gradient

0 k DisplacciKMit of pile

Figure 5. I. Ground resistance characteristics in DNV

0.100

0.075

0.1

0.025

0.000

DisplaccDcnt at pile S Figure 4.4, Distribution of Log (k value ratio : reverse operation k value (displacement: lOmra)) VigMve 5.2. Bilinear-type ground resistance characteristics

/ k value in the 1980 edition of "Specifications"

297
Table 5. 1. Host adequate parameters

Test

Pile a p ak Ground

D X t(lIDl) Um)

S I (t> 19lx 5.3 4.5 2 0 15 Sand S 2 0 3l9x 6.9 15. 0 6. 0 6.0 Sand

S 3 * 610x 9.5 21. 0 3.5 15 Sond

S 4 (4 610x12.7 36.0 4. 0 15 Sand

S 5 * 813x15.0 17. 0 4.5 15 Sond C 1 19lx 5.3 5.0 0.5 1.0 Soft Clay C 2 <t> 813x14 0 30.7 2.5 15 Stiff Clay C 3 <t> 1000x27.2 20.0 1.0 2.5 Stiff Clay C 4 «M500xl3. 0 33. 1 1.0 1.5 Soft Clay C 5 «S 2000x22. 0 51.5 2.0 6.0 Soft Clay

a K

Figure 5.3. Contour map of errors (S3)

lolcd

0 JO <o so 10 100 0 I SPL ACEMENTl MM )

Figure 5. 4. Loud-displacement Curve (S3)

\ Usdinl direction

Sinsltplle 2 parallel pilM 2 «fri« pllet

3 parallel piles J scries pll"

Figure 5. 6.

9 froup pile*

MOMENTl T «M

1

10 10 so «0 50 60

e )

Calucu Isted

Figure 5.7. Test method for group piles (p|R|)

Figure 5.5. Distribution of bending moment of pile (S3)

298
Fillfnc

concrete £

Fillln< concrete

D 'D

Direction of the horlzonlil loadlnt

Steel pipe pile: D 31S.S

n » 9 pile*

Figure 5.8. Horizontal loading tests of piles (MEPC)

Sinile pile

2 parallel

2 seri es plies

3 series pilej

5 group piles

3 parallel piles

0 20 <0 10 BO too 170 110 110 .110 200 Displacement 6 (mm)

Figure 5. 9. Load-displacement curve

TOO ISO Displacement 5 (moi)

Figure 5. II. Load-displacement curve of the 9 piles

Bending Hornient HCtfu) Bending Horuent l!(tfiij)

-4. 0 -2. 0 0. 0 2. 0 0 -2. 0 0. 0 2. 0

.ExperiBental

Calculated

Rear Pile pront Pile

I'iRurc 5. 12. Distribution of bending moment of

the 9 piles

Ground Reaction pCtfia/iii') Ground Reaction pCtfu/m')

-1. 0 0. 0 I. 0 2. 0 0. 0 1. 0 2. 0

PIRI (Front Pile)

PIRI (Rear Pile)

Rear Pile Front Pile

I II II II II II II II Clsplaceient of Uie pile head (S) / pile dla»eter (0) I IM («) Figure 5. 13. Distribution of ground reaction of Figure 5.10. Ratio of load share to the central pile (9 piles used)

^ ^.^^^

299
Table 5.2. Comparison between experimental

values and analytical values

CASE a a a.

Front oilcB •iddle piles Hear piles

Single pile 1. 5 (3.0) 3.0 -

2 series

piles

1. 5 (3.0)

3. 0 (0.8)

1. 0

3 series

piles 1.0

(3.0) 3.0(1.3)

1.0

(0.8)

0. 5

2 parallel

piles 1.0

(2.5)

2.

5

3 parallel

piles 0.75

(Center 2. J) (Side 3. 2) 2.5

9 iroop

pi Ics 0.35

(Center 2. 5) (Side 2.4) 2.5

( Centerl. 6) (Side 1.3) 1.0

(center J 3) (Side 1. 2) 1.0

Figures in parenthesis are eiperlaental values

Curvature <f> Figure 5. 14. M-0 of Cast-in-place Pile

so

^ 60

a.

20

0

Line .r

y

Experiuc ital i-Linear

/ J

i

i f

h

/

0 50 100 150

Displaceient <5 [mm] Figure 5. 15. Load-displacement curve

of Cast-in-place Pile

300
Seismic Effects on Port and Harbour Structures

by

Hajime Tsuchida' and Setsuo Noda^

ABSTRACT

In Japan, many port facilities were damaged by the earthquakes. Efforts has been being continued for designing and constructing facilities of larger seismic

resistance with less construction cost. Design

standards for the earthquake resistant design were

proposed and used widely. After earthquakes of large

scale, performance of port facilities in a zone of sever shaking were investigated in details. Results of those

investigations were reflected to revision of the design

standards.

This report presents Japanese experience on

earthquake damages, current practice in seismic

design and countermeasures against earthquake

disaster in connection with port facilities. KEYWORDS: Earthquake damage, earthquake

resistance design, liquefaction of soil. Port and

Harbour structures,.

1. DAMAGE EXPERIENCE

1.1 Earthquakes which caused seismic

damage

Many of the port facilities in Japan were damaged by

earthquakes. The Kanto earthquake in 1923 is well known as one of the earthquakes which caused

serious damage to many kinds of structures. Port

facilities in Yokohama and Yokosuka were damaged

by that earthquake, too. However, none of

seismographs recorded peak value of the ground

motions in the areas where damage to port facilities took place.

In 1960's, a number of deployed accelerographs

increased considerably and since then, more reliable information on ground motions became available. Major earthquakes which caused damage to port

facilities since that period are given in Table 1.

The peak acceleration shown in the table were all recorded in port area except Niigata. However, the instrument which recorded the Niigata earthquake was

located not much distant from the Niigata Port. In the 1973 Nemuro-Hanto-oki earthquake, the most

serious damage took place in Hanasaki Port. Unfortunately, no accelerograph was installed at the

port. Although the peak ground acceleration in Kushiro Port is given in the Table 1, the damage in Kushiro Port by this earthquake was not serious.

In the 1987 Chiba-ken-Toho-oki earthquake, there was no damage to port facilities; but liquefaction was

observed at many locations in artificially filled land.

Field reconnaissances in detail on seismic

performance of port facilities (damaged and

undamaged) were carried out after the major

earthquakes and reported (Bureau for Ports and

Harbours, Ministry of Transport, et al., 1964, 1965,

1968. and 1973; Tsuchida. et al., 1979; Tsuchida, et

al., 1985; lai, et al., 1988).

1.2 Damage examples

1) Liquefaction of ground

During the Niigata earthquake in 1964, seismic

damage to large variety of structures took place due

to liquefaction of sandy grounds. By this earthquake

people started to recognize the importance of

liquefaction and very intensive research activities were undertaken by many researchers. Liquefactions were observed in almost all major earthquakes which

hit coastal areas in Japan.

2) Gravity Tyf>e Quaywall

By the 1983 Nippon-kai-Chubu earthquake, a quaywall using concrete caissons in Akita Port was

seriously damaged. Fig. 1 shows the typical cross

section and displacements of the quaywall (broken

line : after earthquake, solid line : original). The

damage was sliding of the caissons towards the sea

side more than 1 meter and settlement of backfill was

about 1.5 meter in maximum. It was evident that liquefaction had taken place behind the quaywall.

1) President, Coastal Development Institute of

Technology, 3-16, Hayabusa-cho, Chiyoda-ku,

Tokyo, 102 Japan

2) Director, Structural Engineering Division, port

and Harbour Research Institute, Ministry of

Transport, 3-1-1, Nagase, Yokosuka, 239 Japan

301
3) Sheet pile bulkhead

Fig. 2 also shows a cross section of the steel sheet

pile bulkhead (Ohama No. 2 Wharf) in Akita Port. The sheet piles were broken due to large pressure caused by the liquefied backfill. Adjacent to this bulkhead, there located another sheet pile bulkhead

(Ohama No. 1 Wharf) of very similar construction.

There was no liquefaction at this site and the

structure survived without any damage.

4) Trestle type pier

Since 1960's many trestle type piers with steel pipe

piles were constructed. Some of them were subjected

to strong ground motions. However, none of them was damaged by the earthquakes. Fig. 3 is a cross

section of the tresde type pier in Muroran Port. This

pier experienced ground motions of peak acceleration 209 Gal, during the 1968 Tokachi-oki earthquake.

The structure was designed for the seismic coefficient 0.10.

5) Cellular bulkhead

The cellular bulkhead in Muroran Port was subjected

to the ground motions of 209 Gal, but no damage

was seen. Fig. 4 is showing a cross section of the bulkhead.

6) Breakwater

No damage due to the earthquakes listed in Table 1 was seen on the breakwaters. During the Kanto

earthquake in 1923, a breakwater in Yokosuka Port

settled more than 5 meters. Fig. 5 is a cross section of the breakwater.

7) Warehouse

Many warehouses were damaged by earthquakes in past. The warehouse in Gaiko District of Akita Port was seriously damaged by the 1983 Nippon-kaiChubu earthquake. The floor slab settled

significantly and the foundation piles were broken.

2. OBSERVATION OF STRONGMOTION EARTHQUAKES

The authors and their colleagues have observed the strong-motion accelerations during earthquakes for 29

years in Japan. As of December 1990, 4802 strong

motion earthquake records have been obtained by the observation network of 81 strong-motion

accelerographs deployed throughout Japanese port areas (Kurata and lai, 1991). Fig. 7 shows one of

records recovered from this network during the major

earthquake. The records were the ground accelerations

at Hachinohe Port during the 1968 Tokachi-oki

earthquake which caused serious damage to many

kinds of engineered structures.

To study seismic wave propagations and strains of

302

ground due to seismic waves, the observation of

ground motions at individual points have been found

to be insufficient. An array observations have been

considered essential. Many research groups started observation of this type. PHRI deployed

seismometers three dimensionally in the field of the Tokyo International Airport (Haneda). Fig. 8 is showing the deployment of seismometers of the PHRI array in Haneda.

3. CURRENT PRACTICE IN

EARTHQUAKE RESISTANT DESIGN

3.1 Procedure

Structures in ports and harbours in Japan are designed

in accordance with the Technical Standard of Port and

Harbour Facilities and its Commentary (Japan Port and Harbour Associadon, 1989). Chronologically,

the procedure for earthquake resistant design was

compiled as early as in 1950, revised, supplemented

or legislated in 1959, 1967, 1971 1975 and 1979. The most recent revision in 1989 aims at assuring,

in practice, the greater safety against liquefaction; i.e. the lesson learned from the case history in 1983

Nippon-kai-Chubu Earthquake of magnitude 7.7.

According to the Technical Standard, the steps to be

taken for earthquake resistant design are as follows:

(a) take into account the effects of seismic

inertia force upon the safety and stability of

structures or the effects of seismic

displacement upon lifeline facilines and

(b) assess the liquefaction potential of soil due to earthquakes and, if necessary, take measures against liquefaction. Current procedures taken for these steps will be

briefly reviewed in the following sections.

3.2 Stability against seismic inertia force

Seismic inertia force considerably affects the seismic

stability of port facilities such as quaywalls, piers

and breakwaters. Those effects are, in the roudne

design practice, treated by a method well known as seismic coefficient method; the seismic inertia force

is idealized into a static horizontal load of which

magnitude is specified by the seismic coefficient.

The seismic coefficient is determined by taking

1) regional seismicity,

2) condition of foundation soil and

3) importance of structures

into consideration. The regional seismicity in Japan

is studied by Kawasumi (1951) with case history data of 343 earthquakes since the year of 599 A.D.. The

results obtained by Kawasumi are critically reviewed

and idealized, for use in design procedure, into
seismic zoning with three regions, of which regional

seismic coefficients are specified as 0.15, 0.10 and

0.05.

Because the results given by Kawasumi are of the

alluvial ground, a factor should be multiplied to the regional seismic coefficients for taking variety in

soil conditions into account. The factors are given

for three kinds of subsoil conditions, specified as

0.8, 1.0 and 1.2.

In addition to the subsoil condition, importance of

structures should be considered for determining

seismic coefficient. The factors to be multiplied to the seismic coefficients are given for four levels of

importance, specified as 1.5, 1.2, 1.0 and 0.5.

Seismic inertia force thus specified will, when

loaded, produce increase in the stresses acting in material of structures. Such stresses will be

compared with the allowable limits, which are

specified as 1.5 times high as those specified for the

static stability analysis.

Thus, basically all kinds of structures are designed by

seismic coefficient method for the routine practice.

Among the various port structures, piers with piles

(i.e. trestle type piers) show characterisdc dynamic

response during earthquakes. For the piers of this type, the validity of the design will be checked by

the analysis in which dynamic response of the pier has been taken into consideration (Yamamoto et al, 1970).

4. VALIDATION WITH CASE HISTORY

AND STRONG MOTION DATA

4.1 Effect of vertical component on stability

Every fives years or so often the structures designed

by the simple procedure presented in the previous secuon have been exposed to great earthquakes in Japan. With the strong-motion observadon network

deployed by the Port and Harbour Research Institute since 1963, validity of the simple procedure has been

critically evaluated by the research groups of the same research insdtute.

During the earthquake, structures receive vertical seismic inerua force as well as horizontal one. It is the practice in design, however, that the seismic

inertia force is represented by nothing but a

horizontal load. Uwabe et al.(1976), using verdcal

components of 574 strong-motion accelerograms,

concluded that ratios of vertical vs. horizontal

components in the peak values are less than 1/2 as shown in Fig. 8 and the vertical component has

indeed little effect on the stability of gravity type

quaywalls.

4.2 Relation between max. acceleration and

seismic coefficient

During the earthquake, the peak value of the seismic

inertia force is applied instantaneously. The seismic

load considered in the design procedure is, however, a continuously applied load. If duration of intense

seismic loading is limited to a very short period,

nothing but a insignificant deformation will be

induced. Besides, if slight deformation is allowed for dense saturated sand (i.e. non-liquefiable sand),

dilatant behavior under undrained condidon will instantaneously increase the strength of soil. Thus,

the instantaneously applied peak value of the seismic

inerda force can be higher than the seismic load to be

considered in the seismic coefficient method. Noda

and Uwabe (1975), using case history data of 129

gravity type quaywalls during 12 earthquakes, studied the relation between the maximum ground

acceleration recorded by SMAC-B2 type

accelerograph and the seismic coefficient for gravity

type quaywall. The results are obtained as shown in Fig. 9 with the reladon to be used for the conversion

from maximum acceleraUon to seismic coefficient as follows.

kh = a/g (a < 0.2g) (1) kh = l/3(a/g)i/3 (a > 0.2g)

in which

kf,: seismic coefficient a : maximum acceleration (Gal) g : acceleranon of gravity (Gal)

Kitajima and Uwabe (1979) conducted similar study

for anchored steel sheet pile quaywalls and, using 1 10

case history data, concluded that the relation for the gravity type quaywall is also applicable for the steel sheet pile quaywalls.

As shown in Fig. 10, the reladon in Eq.(l) contains

conservadve errors. If an average relation is sought

for represendng the general correlation between these

quantities, it will be given as follows.

kh = 0.5 a/g (2)

Obviously the relation by Eq.(2) will not be used in pracdce for determining the seismic coefficient from

maximum acceleration. This relation will, however,

be discussed again in the following section in the context of maximum acceleration to be used for assessing liquefacdon potennal.

4.3 Stability against liquefaction
Soil liquefaction, if occurs, considerably affects the

stability of structures. A design criteria to assess

liquefaction potential for a sandy ground in coastal areas was originally proposed based on field experiences and laboratory tests (Tsuchida, 1970).

Liquefaction potential, in the current practice, is assessed by the following procedure which is a

revision of the original one (lai et al, 1989).

1) Classify the soils by comparing the grain size accumulation curves with the ranges shown in Fig.lO;

Soils within the range (A) : very easily

liquefiable soil. Soils within the range (BO or (Be) : easily

liquefiable soil but less liquefiable than the

soils within the range (A).

Consider that no liquefaction will take place for the

soils which do not belong to either the ranges (A),

(Bf) or (Be).

2) Obtain the equivalent N-values from SPT Nvalues with corrections for effective confining

pressures to 0.66 kgf/cm2 and with corrections for fines content. Conduct the earthquake response

analysis of ground by equivalent linear method and

obtain the equivalent acceleration from the maximum

shear stress ratio (the ratio of maximum shear stress over vertical effective stress) by multiplying the

factor of 0.7g as a conversion factor.

3) Use the design charts shown in Fig. 1 1 in accordance with the ranges for grain size accumulation curves. Assess the liquefaction

potential in accordance with the regions to which the equivalent N-value and the equivalent acceleration

belongs;

Region I : Consider that liquefaction will take place.

Region II : Conduct a further study such as laboratory test of soils if desirable.

Otherwise, consider that liquefaction will take place.

Region III : Conduct a further study such as laboratory tests of soils if high reliability is required in the stability of the soil deposit.

Otherwise, consider that liquefaction will not take place.

Region IV : Consider that liquefaction will not take place.

Present criterion has evolved from the studies since 1964 Niigata Earthquake (Tsuchida and Hayashi,

1971). Though the evolution process is quite

independent from the studies by Tokimatsu and

Yoshimi (1983) and Seed et al.(1985), all of these

criteria show remarkably good agreement as shown in Fig. 12 (lai et al, 1989).

In assessing liquefaction potential by the present

procedure, maximum accelerations should be

specified for the input motions to be given at the base rock for earthquake response analysis. Based on

the recent study on seismic zoning (Kitazawa et al, 1984), maximum accelerations as an rock outcrop

motions ( i.e. double amplitudes of incident waves)

are specified for five regions as 300, 250, 200, 150

and 100 Gals.

It is important to assure that the level of safety

against liquefaction should be well balanced with the

level of safety against seismic inertia force. As partly

suggested in the previous section, Eq.(2) shows such

balance are indeed assured in the present design

procedure.

Based on the assessment of liquefaction potential,

adequate measures will, if necessary, be taken against

liquefaction. Relevant studies can be referred to on

designing the area of ground compaction (lai et al, 1987) and on designing spacing of gravel drains (lai et al, 1988).

ACKNOWLEDGEMENTS

The authors express sincere appreciation to Dr. Takamasa Inatomi, Chief of Structural Dynamics

Laboratory; Dr. Tatsuo Uwabe, Chief of Earthquake

Disaster Prevention Laboratory, Dr. Susumu lai. Chief of Geotechnical Earthquake Engineering

Laboratory, and Mr. Yasuo Matsunaga, Member of

Geotechnical Earthquake Engineering Laboratory of

Port and Harbour Research Institute for their considerable support to prepare this report. REFERENCES

Bureau for Ports and Harbours, Port and

Harbour Research Institute, and The First District Port Construction Bureau of Ministry of Transport

(1964) : "Damage to Harbour Structures by the Niigata Earthquake (Part 1)." Bureau for Ports and Harbours, Port and

Harbour Research Institute, and The First District Port Construction Bureau of Ministry of Transport

(1965) : "Damage to Harbour Structures by the Niigata Earthquake (Part 2)." Bureau for Ports and Harbours, Port and

Harbour Research Institute, and The Second District Port Construction Bureau of Ministry of Transport,

and Port and Harbour Division of Hokkaido

Development Bureau of Hokkaido Development

Agency (1968) : "Damage to Harbour Structures by

the 1968 Tokachi-Oki Earthquake."
Bureau for Ports and Harbours, and Port and

Harbour Research Institute of Ministry of Transport

and Port and Harbour Division of Hokkaido

Development Bureau of Hokkaido Development

Agency (1973) : "The Repon on the Damage to Port Structures and the Investigation of the Tsunami

Caused by the Nemuro Hanto Oki Earthquake on

June 1973."

lai, S., Koizumi, K. and Kurata, E. (1987) : "Basic Consideration for Designing the Area of the Ground Compaction as a Remedial Measure against

Liquefaction," Technical Note of Port and Harbour

Research Institute, No.590, 66p.

lai.S., Koizumi, K., Noda, S. and Tsuchida,

H.(1988) : "Large Scale Model Tests and Analysis

of Gravel Drains," Proceedings of Ninth World

Conference on Earthquake Engineering, Tokyo,

Vol.III, pp.261-266.

lai, S., Urakami, T., Mutoh, Y. and

Kikuchi, M.(1988) : "Liquefaction at Coastal Area

and Performance of a Quaywall during 1987 Off East Chiba Prefecture Earthquake," Technical Note of

Port and Harbour Research Institute, No. 616.

lai, S., Tsuchida, H. and Koizumi, K.

(1989) : "A Liquefaction Criterion Based on Field

Performances around Seismograph Stations," Soils and Foundations, Vol.29, No. 2, pp. 52-68. Japan Port and Harbour Association (1980)

: "Technical Standards for Port and Harbour Facilities in Japan (1989 Edition).

Japan Port and Harbour Association (1989)

: "Technical Standards for Port and Harbour Facilities

in Japan (1989 Edition)."

Kawasumi, H. (1951) : "Measures of

Earthquake Danger and Expectancy of Maximum

Intensity throughout Japan as inferred from the Seismic Activity in Historical Times," Bulletin of

Earthquake Research Institute Vol.29, Tokyo

University.

Kitajima, S. and Uwabe, T. (1979) : "Analysis on Seismic Damage in Anchored Sheetpiling Bulkheads," Report of Port and Harbour

Research Institute, Vol.18, No.l, pp.67- 127. Kitazawa, S., Uwabe T. and Higaki N.

(1984) : "Expected Values of Maximum Base Rock

Accelerations along Coasts of Japan," Technical

Note of Port and Harbour Research Institute,

No.486, 137p.

Kurata, E. and lai, S. (1991) : "Annual

Report on Strong-Motion Earthquake Records in Japanese Ports (1990)," Technical Note of the Port and Harbour Research Institute, No. 705. Noda, S., Uwabe, T. and Chiba, T. (1975) : "Relation between Seismic Coefficient and Ground

Acceleration for Gravity Quaywall," Report of Port and Harbour Research Institute, Vol.14, No.4, pp.67-

111. Seed, H.B., Tokimatsu, K., Harder, L.F.

and Chung, R.M. (1985) : "Influence of SPT

Procedures in Soil Liquefaction Resistance

Evaluations," Journal of Geotechnical Engineering

Division, ASCE, Vol.1 11, No.l2, pp.1425-1445.

Tokimatsu, K. and Yoshimi Y. (1983) : "Empirical Correlation of Soil Liquefaction Based on SPT N-value and Fines content," Soils and

Foundations, Vol.23, No.4, pp. 56-74. Tsuchida, H. (1970) : "Design Cntena for assessing Liquefaction Potential of Sandy Ground and

Countermeasures," Proceedings of Annual

Conference for Research Presentation of Port and

Harbour Research Institute, pp.(3)l-33.

Tsuchida, H. and Hayashi, S. (1971) : "Estimation of Liquefaction Potential of Sandy

Soils," Third Joint Meeting of U.S. and Japan Panel on Wind and Seismic Effects, UJ^^R, pp. 1-16. Tsuchida, H., Inatomi, T., Noda, S., Yagiu,

T., Tabata, T., Tokunaga, S. and Hirano, T. (1979)

"The Damage to Port Structures by the 1978 Miyagiken-oki Earthquake," Technical Note of Port and

Harbour Research Institute, No. 325, 50p.

Tsuchida, H., Noda, S., Inatomi, T., Uwabe, T., lai, S., Ohneda, H. and Toyama, S. (1985) : Damage to Port Structures by the 1983

Nippon-kai-Chubu Earthquake," Technical Note of

Port and Harbour Research Institute, No. 511, 447p.

Uwabe, T., Noda, S. and Kurata, E. (1976)

"Characteristics of Vertical Components of StrongMotion Accelerograms and Effects of Vertical Ground

Motion on Stability of Gravity Type QuayWalls,"

Report of Port and Harbour Research Institute,

Vol.15, No.2, pp.239-317.

Yamamoto, R., Hayashi, S., Tsuchida, H., Yamashita, I. and Ogura, K. (1970) : "Evaluation of

Seismic Stability of Trestle Type Pier with Vertical

Steel Piles," Report of Port and Harbour Research

Institute, Vol.9, No.l, pp.179-228.

305
Table 1 Recent earthquakes which caused damage to port faciHties

Date Name Mag. Highest recorded

peak acc. in Gal (Location)

1964/06/16 Niigata earthquake 7.5 159 (Niigata)

1968/05/16 1968 Tokachi-oki 7.8 233 (Hachinohe)

1973/06/17 1973 Nemuro-Hannto-oki 7.4 164(Kushiro)

1978/06/12 1978 Miyagi-ken-oki 7.4 280 (Shiogama)

1982/03/21 1982 Urakawa-oki 7.1 253 (Hiroo)

1983/05/25 1983 Nihonkai-Chubu 7.7 205 (Akita)

1984/08/07 1984 Hyuuganada 7.1 268 (Hososhima)

1987/12/17 1987 Chiba-ken-Toho-oki 6.7 171 (Chiba)

Koorlno post (701)

JO Concrete cavenent

• t7 • ±S '

Call son ',

Fllllni

1

Backfl 1 1 sand

H

{iJ.Ox IS. I X U. i)

1

Dssign depth [—

|

V -13.50 \ Plinning d«pth A -u.oo J

13.00

*-

, .10.40 .I *3.oo 'l i j <.Tnrtr:]^M.30

|

Fig. 1 Cross section of gravity type quaywall

(Gaiko Wharf) in Akita Port

steel pipe pile « 0. SO icjn

1-2 9. 00 C.TO.C 3,00

Fig. 3 Trestle type pier (West No. 2 Pier,

East Side) in Muroran Port

n.jo 2.00

o.so g_

^ H.t.l 'I.SS ^ l.T.L ±0.0

Fig. 2 Cross section of damage to steel sheet pile

bulkhead (Ohama No. 2) in Akita Port

-9.10

»UO0

Fl I I I ng jand

Steel sheet pi le T.S.P.F 1 = 10. !0

Fig. 4 Cellular bulkhead (Nittsu WharO

in Muroran Port

306
Seal*

Fig. 5 Breakwater in Yokosuka damaged by Kanto earthquake (1923)

200. r

0.

-200.

200.

0.

-200.

200.

0.

-200.

(GAU) S-252

(GAL) S-2S2 WEST

(GAL) .S-252 POWN

10. IS. 20. TIME (SECONDS)

25. 30.

Fig. 6 Accelerations of 1968 Tokachi-oki earthquake at Hachinohe Port

2550m

8 .'CDO dZ) f '•) O C D

(

) C DOC

New A-Runway

(a) Plan

(,0.8 Tunnel

No, 5 Tunnel

(b) Profile ' Seismometer

Fig. 7 Deployment of seismometers of array observation system

of PHRI(Haneda Airport)

307
Q Ground AccelerJClon

• Obierved Acceleration

0 iOO 200 • 300 100 500 Ground Acceleration-gals

Fig. 9 Relation between seismic coefficient and ground acceleration (after Noda et al, 1975)

' ' Sond I Grovel Cloy | Sill | Send | Grovel Porlicle Size'(mm) Porlicle Size'(mm)

Fig. 10 Ranges of grain size accumulation curves for liquefiable soils (after lai et al, 1989)

308
(a) For soils in the range (A) (b) For soils in the ranges (Bf) or (Be)

Fig. 11 Design chart for assessing liquefaction potential (after lai et al., 1989)

High " Domoge

Poteni iol Inter- No Signif i

;mediote ;cont Domoge

M = 7.5

'-( To n b i

Tokimotsu ond Yoshimi

(1983)

Seed et cl (1985)'*''

Present Study

)t; Corrected for the effective verticpl stress of t kgf /cmM 98 kN / m^)

i"*; Averoge Japanese SPT energy

ratio (73%) is used for comporison

0 5 10 15 20

Norrriolised SPTN-volue, N

25 30

i

Fig. 12 Comparison with other criteria (after lai et al, 1989)

309

Implementation of Executive Order 12699, "Seismic Safety of Federal and

Federally Assisted or Regulated New Building Construction "

by

Richard N. Wright*

ABSTRACT

The Interagency Committee on Seismic Safety

in Construction (ICSSC) assists federal

agencies involved in construction to develop

and incorporate earthquake hazards reduction measures in their ongoing programs. ICSSC

proposed an executive order for seismic safety

in construction that became the basis for

Executive Order 12699, "Seismic Safety of

Federal and Federally Assisted or Regulated

New Building Construction," dated January 5,

1990. ICSSC is working with federal agencies,

state and local governments and private sector organizations to implement the Executive

Order.

KEYWORDS: building codes; building

standards; earthquake hazard reduction;

existing buildings; retrofitting buildings; seismic

safety

1. THE EARTHQUAKE PROBLEM

Losses of property approaching $10 billion,

losses of 62 lives and severe disruption of

human activities from the October 17, 1989,

Loma Prieta, California, earthquake reminded

the United States of the severe threats posed

by great earthquakes. In perspective, losses were modest because the earthquake was far from the San Francisco Bay area where the

major damages occurred. Unless enhanced

efforts are made to reduce earthquake

hazards, a large earthquake close to a major

metropolitan area can kill tens of thousands of

people, cause tens of billions of dollars of

direct property losses, and, through consequent

losses, severely disrupt economic activity for

the whole nation.

It is important to note that earthquake hazards

are not only a California problem. Forty-six of

our fifty states are vulnerable to strong

earthquakes, Figure 1. In the eastern half of

the country, the area that will be affected by

an earthquake could be 10 times as large as for a similar earthquake in California.

Records from historic earthquakes have

illustrated this pattern. Figure 2, which is believed to be due to differences in rock and

overlaying soil. One or more earthquakes,

equal to or larger than that that struck

California in 1989, are expected to affect U.S.

metropolitan areas in the next ten years.

However, while earthquakes are an inevitable

hazard, they are not inevitable disasters.

Properly designed and constructed buildings

and lifelines (public works and utilities structures) can resist economically the effects of strong earthquake ground shaking.

2. POLICY REQUIREMENTS

The National Earthquake Hazard Reduction

Program (NEHRP) was established by the

President in 1978 in accord with the

Earthquake Hazards Reduction Act of 1977

(P.L. 95-124) to reduce loss of life, injuries,

and economic costs resulting from the occurrence of earthquakes in the United

''National Institute of Standards and

Technology, Gaithersburg, MD 20899

311
States. The Interagency Committee on

Seismic Safety in Construction was established

in 1978 to assist the federal agencies involved

in constructions to develop and incorporate

earthquake hazard reduction measures in their

programs. Twenty-nine agencies participate in ICSSC.

On January 5, 1990, the President issued

Executive Order 12699, "Seismic Safety of

Federal and Federally Assisted or Regulated

New Building Construction." The Executive

Order is based on a proposal by the

Interagency Committee on Seismic Safety in Construction.

The Executive Order responds to the policy

requirements created by the Earthquake

Hazards Reduction Act of 1977. The Act

requires "the development and promulgation

of specifications, building standards, design

criteria, and construction practices to achieve

appropriate seismic resistance." It calls for

"the examination of alternative provisions and

requirements for reducing earthquake hazards

through federal and federally financed

construction, loans, loan guarantees and

licenses." The Act requires attention not only

to federal facilities, but also to federally

assisted or regulated construction. Each

agency is responsible for its own actions.

Neither the Federal Emergency Management

Agency (FEMA) nor ICSSC are given

authorities over the agencies' programs.

3. PROVISIONS OF THE EXECUTIVE

ORDER

The requirements for the earthquake safety of

federal buildings have these purposes: to reduce risks to the lives of occupants; to reduce secondary risks of failures; to improve

the capability for important federal buildings

to be functional during or after the emergency;

and to reduce losses of public buildings. Any

new federal buildings entering the detailed

design stage after January 5, 1990, shall be

designed and constructed in accord with

appropriate seismic standards.

For federally leased, assisted or regulated

buildings, the purpose of the Executive Order

is to reduce risk to lives of occupants of

buildings leased for federal uses or purchased

or constructed with federal assistance; to reduce risk to the lives of persons potentially

affected by earthquake failures of federal

assisted or regulated buildings, and to protect

public investments. For space constructed and

leased for federal occupancy, the Executive

Order applies to agreements executed since

January 5, 1990.

Federal domestic assistance programs affect most new residential construction because

states, municipalities and developers want to be eligible for VA or FHA mortgages. The

Executive Order requires agencies, within 3

years, to plan and initiate measures to assure appropriate consideration of seismic safety in domestic assistance programs.

The concurrent requirements are extremely

important to the private sector and to the

federal agencies. Agencies and ICSSC are not expected to develop federal standards for new

building construction. Administration policy in 0MB Circular A-119 for the use of nationally

recognized standards to the extent possible in federal programs is reflected in the Executive

Order. Agencies would use nationally

recognized private sector standards unless none are available that meet special agency

requirements. The local building code may be

used if the agency determines it provides

adequate seismic safety. ICSSC is working

collectively to provide guidance to the agencies so that each agency doesn't itself have to carry out an assessment of the more than 15,000

local building codes in the country.

Each agency is responsible for issuing its regulations or procedures, planning for

implementation through its own budget

process and regularly reviewing its regulations

and procedures. FEMA is to request from

312
each agency annually the status of its work and

to report on execution of the order annually to Congress.

4. IMPLEMENTATION OF THE

EXECUTIVE ORDER

ICSSC is now working on implementation of

the Executive Order. Guidelines [1] have

been developed for implementation of the

Executive Order. ICSSC also is working with

the private sector in implementation of the

Executive Order. The National Earthquake

Hazards Reduction Program has developed, in cooperation with the Building Seismic Safety

Council and the Applied Technology Council,

"Recommended Provisions for Development of

Seismic Regulations for New Buildings."

These are being used in the national standards

and model codes. The American Society of

Civil Engineers is considering the NEHRP

Recommended Provisions in revision of its widely used national Standard A7, "Minimum

Design Loads for Buildings and Other

Structures."

A study with the Council of American Building

Officials (CABO) [2] compared model code

provisions to NEHRP Recommended

Provisions. Based on its review of the CABO

assessment, the ICSSC recommended as substantially equivalent to the NEHRP

Recommended Provisions:

- 1991 ICBO Uniform Building Code

- 1992 Supplement to the BOCA National

Building Code

- 1992 Amendments to the SBCC Standard

Building Code

State or local ordinances adopting and

enforcing these model codes are deemed

adequate for implementing the Executive

Order.

These nationally recognized private sector standards will become important resources for new federal buildings. ICSSC also is working

with the private sector in preparing guidance

documents for the federal agencies to use to assess the adequacy of the local building codes

in dealing with the seismic safety of federally

assisted or regulated facilities. A study with

the National Conference of States on Building

Codes and Standards [3] assessed the status in the United States of the adoption and

enforcement of adequate seismic provisions.

Considering the strong efforts of the private

sector and state and local government to implement seismic design and construction

standards for new buildings, and the federal

agencies' implementation of the Executive

Order, most U.S. new building construction

should be appropriately seismically resistant by

the mid-1990s.

5. REFERENCES

1. Todd, D. and Bieniawski, A., editors.

Guidelines and Procedures for

Implementation of the Executive Order on

Seismic Safety of New Building

Construction, NISTIR, National Institute of

Standards and Technology, Gaithersburg,

MD, May 1992.

2. Council of American Building Officials,

Assessment of the Seismic Provisions of

Model Building Codes, NISTGCR 91598,

National Institute of Standards and

Technology, Gaithersburg, MD, publication

pending.

3. National Conference of States on Building

Codes and Standards, Seismic Provisions of

State and Local Building Codes and Their

Enforcement, NISTGCR 91599, National

Institute of Standards and Technology,

Gaithersburg, MD, publication pending.

313
314
315

U.S. Army Corps of Engineers Earthquake Engineering Research Program

by

W. D. Roper*, M. E. Hynes**, and R. F. Davidson*

ABSTRACT

To ensure the rapid recovery of Corps

infrastructure and provide effective, rapid

response to earthquake-induced flooding,

landslides, and other life threatening

emergencies, the Corps has recently organized

a focused research and development program,

called the Quake Program, to address the

primary sources of seismic damage, namely

ground shaking and ground failure, and their

impact on Corps facilities. Quake Program

efforts are prioritized according to the needs

of our field engineers who are responsible for

the design, construction, operation and

maintenance of the Corps' flood control and

navigation facilities. The objectives of the

program are: (1) to develop practical, more

effective seismic design procedures where they

are now missing, (2) to reduce the uncertainty

in our seismic safety assessments and avoid

costly defensive design, (3) to reduce the

risk of catastrophic failure of Corps

infrastructure, primarily dams, and (4) to save money on new construction and remediation

efforts. The Quake studies incorporate the

research results of universities and other

government agencies in this country and abroad

to develop seismic design tools tailored to Corps needs. This paper describes some of the

problems we have encountered in our projects

which motivated the establishment of the Quake

program at this time, the scope of the studies

planned for the next few years, and the

coordination of these activities with others

involved in earthquake-related research.

KEYWORDS: buildings; dams; earthquakes;

embankments; foundations; ground motions;

liquefaction; locks; outlet works; seismology;

walls; lifelines.

1 . INTRODUCTION

As part of our Civil Works mission, the Corps

of Engineers designs, constructs, operates and

maintains approximately 400 flood control dams

and reservoirs, and about 150 navigation locks

and reservoirs. We have built thousands of

miles of levees, and do most of the dredging

needed to keep our nation's ports open. In addition, the Corps has been assigned certain

emergency responsibilities under the U. S. Federal Response Plan that would be activated

after a major earthquake. It is in connection

with our Civil Works mission that the Corps is

currently organizing an earthquake engineering

research program, called the Quake Program.

This paper describes a number of problems we have encountered in our large construction and

seismic safety projects which motivated the

development of this new research program, the

organization and scope of the studies planned

for the next few years to address the

perceived gaps in our knowledge, and the

coordination of these studies with the

activities of universities and other

government agencies in this country and

abroad.

2. THREATENED FACILITIES AND RESEARCH NEEDS A sudden great earthquake poses the danger of

catastrophic failures in major works of the

Corps of Engineers including dams, levees,

river control structures, navigation channels,

harbors, airfields, and hazardous and toxic

waste sites, as well as essential military

facilities. At least 39 U.S. states are subject to strong earthquakes and more than 20 million people are at risk. In the central

United States, where we have little modem

experience with earthquakes, at least 20 locks

and dams, 51 pumping plants, 19 reservoirs,

and 23,000 miles of levees could be impacted

by a major seismic event in the New Madrid

area.

In spite of major advances over the last 25 years, serious gaps in our knowledge base

exist in the areas of earthquake hazard

predictions; site characterization for

seismically sensitive parameters; constitutive

behavior, failure mechanisms and material

properties of rock, soils, concrete and

composite (reinforced or remediated) materials

under seismic loads, and the stress and

deformation response of sites and facilities

to seismic loading. Cost-efficient

remediation and defensive design techniques

are needed in addition to careful calibration

of fast advancing numerical methods to actual

field performance. Part of such calibration

includes rapid deployment to sites of

earthquake damage to collect and monitor

transient data. In addition, some of these

same monitoring techniques can be developed

for use in immediate urban search and rescue efforts in damaged areas.

It is our experience that every time the Corps

is faced with a serious earthquake engineering

safety study it becomes a research project

because the state of the art in earthquake

engineering falls far short of our needs.

Thus, earthquake safety studies tend to take a long time since we must perform the research

necessary to develop design solutions. These

research studies require considerable

experimentation in the field, in the

laboratory, and on the computer. Each step in

this solution process involves a great deal of

uncertainty. Once a logical line of reasoning

to solve the problem has been developed the

results may be tentative because new knowledge

provided when earthquakes occur or with a major state-of-the-art advance may necessitate

revisions.

This uncertainty combined with disastrous

consequences in the event of a failure of one

of our critical structures requires diligent

* Headquarters, U. S. Army Corps of Engineers,

Washington, D. C. 20314-1000

•k A U. S. Army Engineer Waterways Experiment

Station, Vicksburg, Mississippi 39180-6199

317
attention to continuously improve our seismic

design capabilities and to confirm that our existing designs conform to the highest

standards of safety. At this time, we design

our structures defensively, to account for the

shortcomings in the state of the art to analyze and design efficiently for earthquake

problems. Our research is aimed at improving

our ability to design effectively, to reduce

uncertainty in our end results and reduce the

cost of our construction.

3. PROBLEMS ENCOUNTERED IN OUR PROJECTS

The motivating force in all Corps Civil Works

research is the need expressed by our field

engineering offices for better design

procedures and tools. Engineering problems

have been encountered in a number of large

Corps projects that have highlighted the need

for research in earthquake engineering. These

problems involve our inability to accurately

predict earthquake ground motion levels, our

incomplete understanding of materials in situ,

material behavior and mechanisms of failure

during and after earthquake loading, the need

for validation and simplification of our

analytical approaches, and the need for more

efficient, economical methods of remediation.

Three projects are described next to illustrate the knowledge gaps we have been

faced with in our efforts to achieve

efficient, well engineered designs for seismic

loading of these key structures. Two projects

are existing dams which have been found to need remediation to withstand their design

earthquakes. The other project is a new navigation structure in the design stage.

3.1 Folsom Dam and Reservoir Project

The Folsom Dam and Reservoir Project is

located on the American River about 20 miles

northeast of the city of Sacramento,

California (see location map in Figure 1) . The reservoir is retained by about 4.5 miles

of earth and rockfill embankment dams and a

14 00-ft long concrete gravity section which

includes the spillway. At maximum pool, the

reservoir retains one million acre-ft of

water. The design earthquake for this project

is a magnitude 6.5 event occurring 15 km away. Catastrophic failure of this dam is estimated

to result in 6,000 deaths, put downtown

Sacramento under six ft of water, and cause

six billion dollars' worth of property

damages. Clearly this is a critical project.

The embankment shells are rolled-fill

constructed from gravelly dredge tailings. A

portion of one auxiliary embankment dam.

Mormon Island Auxiliary Dam, has upstream and

downstream shells founded on these loose,

gravelly dredge tailings. Figure 2 shows a plan and Figure 3 shows cross sections of this

dam. This is the only section of the Folsom

Project found to be unsafe. We are now in the

process of designing and constructing a remedial measure at this location. At the time this safety evaluation was started

in 1978, there were no established procedures

to measure the properties of these gravelly

materials, in situ or in the laboratory, and

to assess their earthquake behavior.

Consequently, we had to perform the research

necessary and incorporate the work of others

to develop testing and analysis procedures

adapted to gravelly soils for which there is

little or no earthquake experience — largescale in situ density tests, geophysical tests

in steel-cased holes, interpretation of

geophysical tests in loose gravels and

rockfill, laboratory compaction and density

tests for gravels, interpretation of Becker Rammer penetration tests and equivalent SPT

values, cyclic loading behavior in large-scale

laboratory tests and corrections for membrane

compliance for extremely loose and moderately

dense gravels, and estimation of residual

strength for gravels)

.

For the concrete gravity section, there were

also problems, such as realistic ways of

determining the adequacy of the complex,

three-dimensional connection between the

embankment wing dams and the partially

embedded concrete gravity dam (see typical

section in Figure 4) subjected to seismic

loads, procedures for assigning values of

reservoir bottom absorption for hydrodynamic

response analyses of the concrete dam, and

methods of accurately determining the

potential for sliding deformation in the dam

or in fracture planes in the rock foundation.

Through almost ten years of testing, analysis,

and conservative interpretation, we were able

to acquire enough information to conclude that

only the 800-ft long section of Mormon Island

Auxiliary Dam with embankment shells on loose

gravels required remedial action. We are now

faced with the poor understanding the

profession has today concerning the actual

behavior of liquefying soils and the design,

construction, verification and seismic

behavior of remediated (composite) zones. A

possible remediation plan under consideration

involves a system of drains (shown in Figure

5) to control pore pressures.

We now find that we do not really understand

the effective stress states and stress-strain

behavior of soils as they are deforming while

excess pore pressures increase during seismic

loading, and after liquefaction as they

develop residual strength. We find we do not

fully understand the seismic behavior of

improved soils with zones of higher density

from dynamic compaction, or with stone columns

with higher permeability, or reinforced with

piles, and the pore-pressure, stiffness and

strength interaction of these relatively loose

and dense zones under dynamic loads. We cannot accurately predict the change in

density to be expected with distance from a remedial action. We cannot easily verify that

we have indeed achieved the desired results,

since the remedial action typically radically

increases horizontal stresses which are difficult to measure and for which none of our

seismic safety evaluation procedures have been

developed and calibrated. For example, SPT

blowcounts corrected to one tsf is a widely

adopted practice, however, the correction

factor Cj^ was developed for horizontal stress

levels typically found in natural deposits,

not the high values that result in remediated

zones

.

This wide range of uncertainty in the design

and effectiveness of remedial actions causes

us once again to adopt conservative

interpretations of our analytical and test

results. In this project, our efficiency as engineers is hampered because the design tools were not readily available but had to be

developed, and the end results still have

considerable uncertainty requiring costly

defensive design.

318
3.2 Sardis Dam and Reservoir Project

Sardis Dam is located in northwestern

Mississippi, 70 miles from the southernmost

extension of the New Madrid fault zone in the

central United States (see location map in

Figure 1, project plan in Figure 6, and

typical sections in Figure 7). Sardis is a hydraulic fill dam, and study of its seismic

safety began right after the near failure of

the hydraulic fill Lower San Fernando Dam, Los

Angeles, California, in the 1971 San Fernando

earthquake. More than once in the process of our seismic safety evaluation we were about to conclude that the dam was safe, only to have

new developments in the state of the art of

earthquake engineering force us to reconsider.

One of the first issues was the level of

shaking to be expected at the site — early

estimates were that it was too small to be of concern. In 1980, Dr. Ellis Krinitzsky of the

Waterways Experiment Station at Vicksburg,

Mississippi, conducted a survey of 18 individuals and organizations, all of whom

were considered to be experts at determining

ground motions. All were given the same "hypothetical" sites, one of which was, for

practical purposes Sardis Dam. The peak

ground accelerations that were recommended

ranged from 0.03g to 0.5g — from

imperceptible to very strong. We eventually

reached the conclusion that the expected

ground motions were sufficiently large to require us to assess the liquefaction

potential of the dam and foundation.

After extensive investigations of the sandy

portions of the dam and foundation, we had

concluded that the dam was safe, when Dr. H.

B. Seed (Seed et al. 1983) published data from

China indicating that some clays were susceptible to liquefaction — up to that time

clays had generally been assumed to be nonliquefiable. Weak, sensitive clays meeting

the Chinese criteria were in fact found in the

foundation of Sardis Dam, and in 1988 an upstream berm was completed in the area believed to be most likely to fail. It was recognized at that time that more work was needed to evaluate the safety of the rest of

the dam.

The site characterization difficulties for

this site and several other complexly

stratified alluvial dam sites in the central

United States challenged our in situ

investigation techniques and consequently we

improved our ability to perform and interpret

cone penetrometer tests. We pursued

laboratory investigations to better understand

the behavior of these fine-grained soils under

cyclic loading and to better define the

criteria indicated by the data from China.

These studies helped define additional

portions of the dam and foundation that

required remedial action. We devoted more

than a decade to development and validation of

a sophisticated effective stress nonlinear

large strain finite element technique to estimate the deformed shape that would result

from an earthquake and to develop design

parameters for the remediated zones. The

deformed shape of the embankment after

earthquake shaking using this technique is

shown in Figure 8

In January 1992, nearly 20 years after the San

Fernando earthquake that started this

reevaluation, the conclusion of the Sardis

studies is that additional remedial work is

needed at a cost of several million dollars.

Design of the proposed solution — driving

several thousand large concrete piles (see

Figure 9) — will be a challenge to the state

of the art. Because of the current lack of

design tools, we are again forced into a certain amount of concurrent research and

design, even though that is not usually a very

efficient way to do either.

3.3 Olmsted Locks and Dam Project

Another challenge we now face is to design and

construct a new double lock and dam on the

Ohio River at Olmsted (see location map in

Figure 1) . A portion of the dam lies down

flat on the river bed so that barge traffic

only needs to use the lock in periods of low

water. This river carries some of the

heaviest barge traffic in the world. Loss of use of this important shipping route would

have a serious impact on the economy of our nation as evidenced by the economic damage

caused by the severe drought which made

shipping impossible on portions of the

Mississippi River in 1991. While we believe

that the potential for loss of life caused by

earthquake- induced failure of this structure

is small, the economic losses could be

substantial, and failure could shut down

navigation on the Ohio River for an extended

period.

This important navigation link is located on the northernmost extension of the New Madrid

fault zone. Since v;e have little modern

experience with significant earthquakes in the

central United States, the earthquake ground

motion levels appropriate for design remain

controversial, as they were for Sardis Dam. Since the seismic design level this project is largely an economic decision, the many issues

that enter into a detailed risk analysis (such

as ground motion levels, uncertainty in

engineering design aspects, and quantification

of consequences) remain controversial because

of differences in opinion as to how to define

them, and because the construction costs associated with these aspects are substantial.

The design of this massive, complex, rigid

structure remains a challenge. The realistic

analysis of stiff, rigid, embedded, submerged

walls subjected to dynamic loads is a distinct

challenge to our dynamic soil-water-structure

interaction analysis capability. We have much

to learn yet before we have accurate,

efficient, practical tools for dynamic design

of its pile foundations. Our methods for

determining hydrodynamic effects need

improvement and validation. To add further

geotechnical interest, there is an old

landslide on the landside of the locks. The

challenge here is to estimate the material

strengths in this old landslide and estimate

the extent of deformations in an earthquake.

Like Sardis and Folsom Dams, the Olmsted

Project involves concurrent research and

design.

On projects such as Olmsted, we would like to be able to make intelligent decisions

concerning the tradeoffs between the cost of

providing additional seismic resistance and

the expected damage to be avoided — direct

and indirect, tangible and intangible. At the

present time, we are not able to define these

aspects with sufficient accuracy to be useful.

Improvements will require, for example, better

estimates of expected ground motions,

including recurrence rates, and better methods

319
of predicting structural damage for a given

level of shaking. It has also become clear as

we progress in this project and others that

the tools for estimating vertical ground

motion, empirically and analytically, have not been developed and validated.

The Folsom Dam and Sardis Dam seismic safety

evaluations and the Olmsted Locks and Dam design phase are three examples of our field

experiences in the shortcomings of the state

of the art in earthquake engineering. These

and other projects like them lead us to conclude that once we have done the best we

can to evaluate the seismic safety of our

critical facilities within the current

limitation of the state of the art, the

remaining uncertainties are large,

uncomfortably large, considering the costs and

risks involved. These issues have thus

motivated us to organize the Quake Program.

Our planning process and the current scope of

studies in the program are described in more

detail in the next section.

4. FOCUS AND SCOPE OF QUAKE PROGRAM

4 . 1 Program Focus and Planning Process

Since the motivation for Corps-funded research

related to Civil Works is primarily driven by

needs in the field, it is the nature of Corps

Civil Works research programs that they are highly problem-oriented, and generally

classify as applied research. These programs

are also directly focused on the Corps Civil

Works facilities. Thus, the niche that these

programs fill, the Quake Program in

particular, is a link between the more basic

research pursued at universities and the

rather unique Corps field engineering needs

for practical design of critical flood control

and navigational facilities.

The Corps is actually responsible for the

enormous capital investment that the United

States has made in our system of dams, locks

and related facilities. As public servants,

it is our mission to protect public safety and

to spend public dollars wisely. This serious

responsibility sharpens our perspective of

research needs for the Corps. Consequently,

our selection of research topics is driven by

creative ideas that help us solve a real field

problem, rather than creative ideas that would

be simply intellectually interesting to study.

The process by which we formulate our study

topics and prioritize them is a combination of

(1) needs identified by senior engineers in

Corps field offices, (2) knowledge gaps

identified by research elements within the

Corps, and (3) Corps policy guidance provided

by program monitors from Corps Headquarters.

We shape our program based on suggestions

provided by a Field Review Group composed of

the senior field engineers, who review and

prioritize all proposed topics. The Field

Review Group provides guidance for the

direction of various studies to better meet

field needs, and suggests new topics for

consideration.

The Corps research elements manage and execute

the program, suggest topics to the Field

review group for their consideration, provide

guidance on technical feasibility, and

distinguish between problems actually

requiring research and technology transfer (if

the needed tools are actually available but

have not yet made their way into practice)

Technical Monitors from Corps Headquarters

maintain an oversight role throughout the

process to provide technical and policy

guidance for the program. The proposed

program incorporating the field comments and

priorities is approved, guided and reviewed by

Corps senior policy officials to assure

coherent, sensible, productive, cooperative,

mission-oriented expenditure of our limited

research funds.

4.2 Structure and Scope of Proposed Studies

The program structure follows the solution

procedure for an earthquake engineering

problem: 1) What are the ground motions? 2) How will they be delivered to the

site/facility? 3) Analyze the facility for stress and deformation using appropriate

boundary conditions, properties, material

behavior and numerical and physical models, 4) Estimate actual performance, and 5) Revise

design or implement remedial action if needed.

The program has three primary focus areas (1) earthquake ground motions, (2) geotechnical

earthquake engineering, (3) structural

earthquake engineering, and one secondary

focus area in emergency response. Other Corps

civil works related disciplines, such as hydraulics, electrical and mechanical, have

indicated that they had no pressing earthquake

engineering problems requiring research, and

thus are not included in the program at this

time. As recommended by our Field Review Group, the

initial efforts are directed toward those

areas where greatest loss of life and property

would result from an earthquake in conjunction

with those structures most likely to fail from

earthquakes. That area is unquestionably

reservoir dams. Within the arena of reservoir

dams, the Field Review Group recommended

prioritization of research efforts dealing

with liquefaction, followed by nonlinear

behavior of concrete dams. Thirdly, they

recommended that the next area of emphasis

should be navigation locks and dams and other

hydraulic structures. In addition, the Field

Review Group recommended that additional

strong motion instruments be placed in

representative structures at sites where

earthquake events are likely to occur, to provide additional data for validation of

modeling and analysis techniques.

Evaluating the adequate seismic performance of our facilities usually consists of estimating

the deformations that we expect and deciding

whether the facilities can safely withstand

these deformations. There are presently

serious limitations in our ability to estimate

deformations, and improvements will require

(1) better estimates of expected ground

motions, (2) improved analytic tools for

calculating deformations including better

constitutive models, (3) improved site and

laboratory investigation methods that provide

the information needed to use those analytic

tools, and (4) validation of items 2 and 3 with full scale field data, particularly from

earthquakes and possibly centrifuge testing.

4.2.1 Earthquake Ground Motions

The greatest uncertainty for all of our

projects, be they earth dams, concrete dams,

locks, levees, harbors, etc., is by far the

level of earthquake ground motion that we should design for. Determining the incoming

320
ground motions and other possible seismic

hazards is common to all design efforts. It

is in this step that we define the "loads" we

will design for. We find that with each new earthquake we learn a great deal. In this

topic area we have planned several work units

to improve our procedures for selecting design

ground motions for engineering analyses. This

effort includes: 1) reconnaissance of

earthquake stricken locations world-wide to collect full-scale performance data, 2) continued development of ground motion and

response spectra design charts and data bases,

3) development of empirical and analytical

procedures to estimate vertical motions in

time and frequency domains, 4) investigation

of the effects of fault mechanics on vertical

motions, 5) two- and three-dimensional

boundary effects on ground motions, traveling

waves and spacial variation of motions, and 6) instrumentation of a site and structure to capture detailed full scale response data for

calibration of analytical procedures.

The continued effort to improve our ability to determine design ground motions is essential

since this aspect of earthquake problems

typically is a prime controlling factor in the

overall cost of a project. Overestimation of

ground motions leads to costly overdesign, and

underestimation may lead to loss of life and

costly, time consuming remediation.

4.2.2 Geotechnical Earthquake Engineering

This portion of the Quake Program is focused

on liquefaction and its effects on our

facilities. There are four main topic areas

addressed:

* Site characterization and material behavior,

to include development of innovative methods

such as geophysical techniques for in situ

identification of potentially liquefiable

soils; rapid, less expensive assessment of

dynamic material properties in situ; and

investigation of the behavior of soils as they

approach liquefaction, liquefy, and after

liquefaction.

* Seismically-induced deformations to include

a series of work units to develop improved

methods for estimating deformations, to include the effects of three-dimensional

boundary conditions and geometry, and to examine the potential behavior under strong

aftershocks.

* Remedial measures and ground improvement to include a series of studies on the fundamental

behavior of remediated soils which are now

composites; development of procedures for

realistically analyzing these composite

materials; investigation of the effects of

lateral stresses and migration of excess pore

pressures into improved ground.

* Dynamic soil-structure-water interaction

problems to include design of deep and shallow

foundations, walls, tunnels and shafts, and

embedded and submerged structures, in

cooperation with the structural counterparts.

The best type of data for improving our design

procedures is full scale performance data.

The above topics are planned to be integrated

with data from reconnaissance of earthquake

damaged sites and quantitative information

from an instrumented structure mentioned in

the previous section.

4.2.3 Structural Earthquake Engineering

The structural earthquake engineering studies

focus on concrete dams and the outlet works

essential for controlled release of the

reservoir. In the course of working on these

types of projects it has become clear that the

output from finite element analyses is not easily translated to the kind of information

needed by structural designers; so, another

aspect of the proposed studies is to develop

such basic tools. The areas for research

planned at this time are:

* Seismic structural response of concrete dams

to include sliding structural stability,

nonlinear material behavior and failure

mechanisms, nonlinear behavior of rollercompacted concrete, dynamic uplift pressures,

seismically-induced displacements, design

earthquake level selection, nonlinear behavior

of arch dams, field and model verification

tests, reconnaissance of earthquake stricken

sites to collect full scale performance data,

and instrumentation of a site to collect

detailed, quantitative full scale performance

data for calibration of analytical procedures.

* Seismic response of outlet works to include

overturning stability, nonlinear material

behavior, equipment-structure interaction,

bridge-tower interaction, and tower-abutment

interaction.

* Development of basic engineering tools for

extraction of design force information from

response spectra analyses, and pre- and postprocessors for finite element output.

* Additional soil-structure-water interaction

studies for piles, walls, concrete channels,

tunnels and shafts, in cooperation with the

geotechnical counterparts.

4.2.4 Emergency Response

Another important mission of the Corps is

emergency response in the aftermath of a natural disaster. We have certain

responsibilities under the Federal Response

Plan that would be activated after a major

earthquake, such as assisting with removing

debris, and restoring transportation routes

and utilities. We are also often asked for

assistance by others. For example, after the Loma Prieta earthquake, the Corps had more

than three hundred engineers in the field

supporting the U. S. Federal Emergency

Management Agency.

Our plans for this portion of the Quake

program are still in the formative stage. At

this time we envision (I) periodic review of

progress in the general field of earthquake

engineering to select information that would

impact our emergency response mission, (2) coordination of Quake research and emergency

response research, and (3) action to assist

transfer of technology to Corps emergency

response offices.

5. COORDINATION OF QUAKE PROGRAM WITH OTHERS

For many years, the Corps has exercised well

established, existing forums for keeping

abreast of current research in earthquake

engineering, primarily through national and

international, governmental and private

professional activities. We typically work

with leaders in the field in our earthquake

engineering safety evaluation problems, and

321
perform research in our Corps Laboratories.

We learn much from participation,

communication and coordination by our

headquarters, field and research laboratory

staff with:

a. Activities of national and international

professional societies such as American

Society of Civil Engineers, Earthquake

Engineering Research Institute, Seismological

Society of America, U. S. Committee on Large

Dams, and the International Committee on Large

Dams.

b. Activities of national organizations such

as the National Research Council Committee on Earthquake Engineering and Transportation

Research Board, and the Electric Power

Research Institute.

c. Activities of federal agencies, such as the four primary agencies involved in the

National Hazards Earthquake Reduction Program,

namely the Federal Emergency Management

Agency, the U. S. Geological Survey, the

National Science Foundation (NSF) which

includes the National Center for Earthquake

Engineering Research, and the National

Institute for Standards and Technology, as

well as the many contributing agencies.

d. Activities of university organizations

such as the consortium of California

Universities for Research in Earthquake

Engineering, and the Earthquake Engineering

Research Center at the University of

California, Berkeley.

e. Activities of international government

groups such as the Joint U.S. -Japan Panel on Wind and Seismic Effects.

f. Activities of private companies such as Woodward Clyde and Associates, Dames and Moore

and Associates, H. J. Deggenkolb and

Associates, and others involved in state-ofthe-art earthquake engineering projects.

Through our Field Review Group meetings, close

communication with federal agencies, and work

with other experts, we develop the perspective

needed to plan our research program in

earthquake engineering to meet our Civil Works

mission.

6. CONCLUSIONS

In the development of a new federally funded

research program, it is essential to carefully

identify research needs and expected products,

coordinate and communicate with other federal

agencies, plan and execute the proposed

program, and transfer results to the federal

field engineers.

We proposed the Quake Program in response to

an outcry from our district and division

offices who repeatedly encounter the

shortcomings in the state-of-the-art in

earthquake engineering in their seismic safety

evaluations of existing and design of new

Civil Works projects, primarily flood control

and navigation facilities. The focus of our proposed Quake Program is applied research —

to take fundamental research contributions

such as scientific agency and university

research, and develop practical design tools

tailored to Corps needs for use in Corps

offices. At this time, the focus of the

program is on the seismic safety of reservoir

dams — the improvements needed in ground

motion selection and geotechnical and

structural engineering to reduce uncertainties

and develop more efficient, effective

procedures for assessing dam safety under

seismic loading and designing remedial actions

if required.

7 . REFERENCES

Finn, W. D. L. , Ledbetter, R. H. , Fleming, R.

L. , Jr., Templeton, A. E., Forrest, T. W. , Stacy, S. T. 1991. "Dam on Liquefiable

Foundation: Safety Assessment and Remediation," 17th Congress on Large Dams, Vienna,

Austria, June 1991.

Ledbetter, R. H. , Finn, W. D. L. , Nickell, J. S., Wahl, R. E., Hynes, M. E. 1991. "Liquefaction Induced Behavior and Remediation for Mormon Island Auxiliary Dam," International

Workshop on Remedial Treatment of Liquefiable

Soils, UJNR, Tsukuba Science City, Japan.

Seed, H. B. , Idriss, I. M. , Arango, I. 1983.

"Evaluation of Liquefaction Potential Using

Field Performance Data," Journal of Geotechnical Engineering, American Society of Civil

Engineers, Vol. 109, No. 3, pp. 458-482.

322
Figure 1. Locations of three Corps projects

Figure 2. Plan of Mormon Island Auxiliary Dam, Folsom Project,

showing locations of field investigations

323
EL.37e

EL. 312

I \ 1 I I I 1 500 700 900 1100 1300 1420

DISTANCE, FT.

Figure 3. Idealized section of Mormon Island Auxiliary Dam,

Folsora Project, showing estimated shear wave velocities in fps

Figure 4. Typical section of Wing Dam, Folsom Project, at

interface with Concrete Gravity Dam

324
325
I

SARDIS DAM

LITTLE TALLAHATCHIE RIVER, MISSISSIPPI PLAN- MAIN DAM & ABUTMENT DIKES

Figure 6. Plan of Sardis Dam, Mississippi

TYPICAL SECTION OF DAM

Y///////A CLAY OR COMPACTED FILL

\-}V:^;;^;;.y;A HYDRAULICALLY PLACED SAND

millllll SILTS [^^^3' ALLUVIAL DEPOSIT

Figure 7. Idealized section of Sardis Dam, Mississippi

326
Figure 8. Initial and post-liquefaction estimated shapes for

Sardis Dam (without remediation, after Finn et al. 1991)

/\

f1 p

Figure 9. Conceptual plan for remediation of Sardis Dam by

strengthening and stiffening upstream foundation materials

(after Finn et al. 1991)

327

A Progress Report on the Worldwide Earthquake Risk Management

(WWERM) Program

by

S. T. Algermissen', Walter W. Hays', and Paul R. Krumpe^

ABSTRACT

Considerable progress has been made in the Worldwide Earthquake Risk Management (WWERM)

Program since its initiation in late 1989 as a cooperative program of the Agency for International

Development (AID), Office of U.S. Foreign Disaster Assistance (OFDA), and the U.S. Geological

Survey. Probabilistic peak acceleration and peak Modified MercalU intensity (MMI) maps have been

prepared for Chile and for Sulawesi province in Indonesia. Earthquake risk (loss) studies for dwellings

in Gorontalo, North Sulawesi, have been completed and risk studies for dwellings in selected areas of

central Chile are underway. A special study of the effect of site response on earthquake ground motion

estimation in central Chile has also been completed and indicates that site response may modify the

ground shaking by as much as plus or minus two units of MMI. A program for the development of

national probabilistic ground motion maps for the Philippines is now underway and pilot studies of

earthquake ground motion and risk are being planned for Morocco.

KEYWORDS; Earthquake; hazard; international; risk

1. INTRODUCTION

The WWERM Program was developed as an outgrowth of an Executive Briefing on "Strategic Planning

to Reduce Economic Impacts of Earthquake Hazards throughout the World" held at the National

Academy of Sciences, Washington, D.C., on March 8 and 9, 1988 (Williams and Hays, 1988). This

Executive Briefing established the need for a more quantitative evaluation of the potential for earthquake

disasters and for the assessment of the economic impact and casualty loss potential of such disasters. In addition, the need to assess the long-range economic losses associated with destructive earthquakes was also outlined.

Algermissen and others (1991) reported on the WWERM Program from its inception through mid- 1991.

Here we report on progress in the WWERM Program from mid- 1991 through May 1992.

2. OBJECTIVES OF THE PROGRAM

The initial objectives of the program were to develop an action plan and conduct a pilot program in each

of three geographic regions to demonstrate the utility, importance, feasibility, and need for uniform

'U.S. Geological Survey

^Department of Energy

329
global earthquake hazard mapping and risk assessment/management as a contribution to the International

Decade for Natural Disaster Reduction (IDNDR).

The initial three geographic areas for pilot studies were Morocco, Peru/Chile, and Indonesia. Since the

initiation of the project in 1990, the scope of the work has been greatly expanded through the funding

of additional program in Indonesia and the Philippines by the AID Mission in each country. The

supplemental funding from the AID Office in Indonesia has permitted the following additional work:

(1) a comprehensive assessment of earthquake ground motion in North Sulawesi Province; and (2) the

calculation of estimates of expected future losses to masonry dwellings in Gorontalo, North Sulawesi.

Funding from the AID Mission in the Philippines has allowed us to undertake a comprehensive program

of earthquake ground-motion assessment for the entire country together with pilot studies of future

expected earthquake losses. As in Indonesia, OFDA has provided supplemental funding that has made

it possible to purchase and transfer a Macintosh computer and associated equipment to the Philippine

Institute of Volcanology and Seismology (PHIVOLCS) for training and technology transfer for seismic

hazard mapping and risk assessment.

As work progresses, we are continuing to expand the technical horizons of WWERM with the hope that the project will eventually become global in scope.

3. PROGRESS

3.1 Chile

Because of funding and technical results available from an earlier OFDA-USGS Project "Seismic Zoning

and Implementation of Improved Earthquake Resistant Design in Chile" which in many ways

complemented the WWERM Program, it has been possible to develop a considerably broader program

in Chile with results and products having widespread application. These products include: (1)

probabilistic acceleration and seismic intensity maps of the entire country for an exposure time (period

of time of interest) of 50 years and 10 percent chance that the mapped values will be exceeded; (2)

comprehensive evaluation of the "site effect" in central Chile, that is, the contribution of the surface

geology to the modification of ground shaking during earthquakes; and (3) collection of dala for loss assessment.

A probabihstic ground acceleration map for a time period of interest of 50 years is shown in Figure 1. A similar map of seismic intensity for central Chile is shown in Figure 2. These maps are prepared for "average" site conditions in Chile. The primary use of the acceleration maps is to provide the ground

motion basis for the earthquake resistant design of structures either for a specific building or as part of

the seismic provisions in a building code. The primary use of the intensity maps is to present the

expected levels of damage (in descriptive form) for disaster preparedness, mitigation, and land use

planning. Figure 3 shows the magnitude of the variafion in ground shaking (in units of intensity)

associated solely with variations in surface geology. Figure 4 shows how these variations in shaking can be incorporated into probabilistic seismic intensity maps to provide a more realistic estimate of the

maximum ground shaking levels that may occur in an area in a specific time period of interest. The

level of refinement of estimated ground mofion shown in Figure 4 is generally not available, even for most areas of the United States, and represents a state-of-the-art product. The importance of surface

geology in controlling ground shaking during earthquakes was first appreciated iii Chile after the M5=7.9,

central Chile earthquake of 1985. This is, however, the first study to describe this effect in a

330
quantitative manner. We anticipate that the intensity maps in this study will be of considerable use in disaster preparedness, mitigation, and land use planning in central Chile.

During October 1991, USGS personnel participated in an international seminar and a separate workshop,

both related to the WWERM Project. The first meeting was organized by UNDRO, MAPFRE (a

Spanish reinsurance company, ONEMI (the national emergency office in Chile), the Universidad de

Chile, the Caja Reaseguradora de Chile (a Chilean insurance company), and the USGS. This "Seminario

Intemacional" held in Santiago from October 21-24, 1991, was entitled "Disastres Sismicos Urbanos,

Codigo de Construccion Sismoresistente y Seguros" (Urban Seismic Disaster, Earthquake Resistant

Construction Codes and Insurance).

A workshop entitled "Applicacion de Mapas Probabilisticos de Movimientos Fuertes para la Reduccion

del Riesgo Seismico en Chile" (Application of Probabilistic Maps of Strong Motion to the Reduction

of Seismic Risk in Chile) was organized by the USGS and the Universidad de Chile during the week

following the international seminar. This highly technical workshop was designed to make the results of the two OFDA-USGS projects in Chile available to scientists, engineers, and disaster management

specialists in Chile. In particular, scientists and engineers from the major Chilean universities and from

major consulting firms involved in the development and revision of the earthquake resistant design

provisions of the Chilean National Building Code, together with disaster preparedness and mitigation

specialists from ONEMI (the national emergency office in Chile), were invited to attend and participate

in the workshop. The papers presented at the USGS-Universidad de Chile workshop are currently being

edited. These papers will be published in a special issue of "Revista Geofisica", a professional journal

published by the Pan American Institute of Geography and History with wide circulation in Latin

America.

3.2 Indonesia

The earthquake hazard and risk study of North Sulawesi, Indonesia, was completed in August 1991, in advance of a workshop held in Bandung in September 1991. Products of the Pilot Study include: (1)

synthesis of the geologic structures related to earthquake occurrence in the area; (2) expected peak

horizontal ground acceleration on both hard rock and soft soil that have a 90 percent probability of not being exceeded in 50 years; (3) development of inventory and earthquake vulnerability relationships for dwellings; and (4) estimation of both future average annual loss and catastrophe potential for the

dwelling stock in Gorontalo, North Sulawesi. Papers on all aspects of the Pilot Study were presented

at the workshop in Bandung. A technical paper "Probabilistic Seismic Hazard Estimates in North

Sulawesi Province, Indonesia" describing the results of the Pilot Study will be presented at the

SECOND-ASIA CONFERENCE on Engineering for Mitigating Natural Hazards Damage that is scheduled for Yogyakarta, June 22-26, 1992, (Thenhaus and others, 1992).

3.3 The Philippines

The WWERM Program will begin in the Philippines in June 1992. The intention is to: (1) prepare

national probabilistic hazard maps for all of the country; (2) to undertake pilot studies of future

earthquake losses in selected areas of the country; and (3) to conduct training and technology transfer

in earthquake hazard and risk assessment in conjunction with the Philippine Institute of Volcanology and

Seismology (PHTVOLCS).

331
3.4 Morocco

The Morocco phase of WWERM has been much delayed, first because of the Persian Gulf War, and

second, because of the accelerated program of hazard and risk assessment that was undertaken in Indonesia. The work in Morocco will include a demonstration project of earthquake hazard (ground

motion) and risk (loss to dwellings) in selected areas. The work will be done cooperatively with the Center for Technical and Scientific Research (CTSR) in Rabat, Morocco. Work in Morocco will begin

in August 1992.

4. SUMMARY

Major studies of earthquake ground motion estimation have been completed in Chile and investigations

of risk to dwellings in Central Chile are nearing completion. Similar studies have been completed in Sulawesi, Indonesia. Very successful workshops highlighting the technical results have been held in Chile and Indonesia. A major study of earthquake ground motion and pilot studies of seismic risk to dwellings will begin in the Philippines in June 1992 followed by pUot studies of earthquake ground

motion and risk in Morocco.

5. REFERENCES

Williams, M.E. and Hays, W.W., 1988, Strategic Planning to Reduce Economic Impacts of Earthquake

Hazards throughout the World, Proceedings of the Executive Briefing, National Academy of

Sciences, Washington, D.C., March 8-9, 1988, U.S. Geological Suivey, Open-File Report 88-

361, 149 pps and Appendices

P.C. Thenhaus, I. Effendi and E. Kertapati, Probabilistic Seismic Hazard Estimates in North Sulawesi

Province, Indonesia, 1992, to be presented at the SECOND-ASIA CONFERENCE on

Engineering for Mitigating Natural Hazards Damage

S.T. Algermissen, Paul R. Krumpe, and Walter W. Hays, 1991, The worldwide earthquake risk

management program. Proceedings, 23rd Joint Meeting of the U.S-Japan on Wind and Seismic

Effects, Tsukuba, Japan, May 12-17, 1991.

S.T. Algermissen and Paul C. Thenhaus, 1992, The Worldwide Earthquake Risk Management Program:

A Program for Global Earthquake Hazard Assessment, presented at the 1992 Joint Spring

Meeting, May 12-16, 1992, of the American Geophysical Union in Montreal, Canada

332
Figure 1. Probabilistic ground motion acceleration map of Chile for a 50 year time pe.'iod

of interest The ground motions mapped have oniy an estimated 10% chance of living exceeded

in 50 years. Maps of this type are widely used as design maps in the earthquake resistant

design provisions of building codes.

333
50 Year

Intensity

221

2£

35=

Figure 2. Probabilistic ground intensity map of Chile for a 50 year time

period of interest. The ground motions mapped have only an estimated

10% chance of being exceeded in 50 years. Maps of this type are often

combined with site specific shaking maps (such as Figure 3) to produce

probabilistic ground shaking maps that include the site specific response

associated with the near surface materials beneath a site.

334
7rw 70°W

Figure 3. Site specific response factors aSj=ociaied with the near surface materials in

central Chile, This map is coaibLQed with the regional ground motions shown in Figure 2

to produce the map in Figure 4.

335
Figure 4. Probabilistic ground motion intensity map of central Qiile for a 50 year time period

of interest. This map combines the mapped values of Figure 2 wilh those of Figure 3 to produce

a probabilistic ground motion map that includes the effect of near surface materials. These maps

are valuable in disaster preparedness, mitigation, land use plarming and eaiiliquake risk (loss)

studies because they represent an estimate of the maximuni damage (and consequently loss) that

is expected to occur in the period of time considered in the mapping.

336
Development of Advanced Reinforced Concrete Buildings Using High-Strength

Concrete and Reinforcement Researches on Structural Performance

Conducted in 1991

by

Tatsuo Murota', Hisahiro Hiraishi^ Takashi Kaminosono', Masaomi Teshigawara\

Hitoshi Shiohara^, and Hideo Fujitani*

ABSTRACT

A five-year National Project has been promoted by

the Ministry of Construction of Japan since 1988 to develop such structures as super high-rise rein- forced concrete buildings using high strength and

high-quality concrete and reinforcing steel. The

project is simply referred as to "New RC". This pa- per describes the outlines of some research subjects which were conducted in 1991, related to

structural performance.

Damage due to Bending and Straightening of High Strength Reinforcing Bars.

Strain Aging Properly of High Strength Reinforcing Bar Performance of H^echanical Connector for Splicing Reinforcing Bars Performance of Member using Reinforcing Bars with High Yield Ratio and splicing

Threaded Couplers

Framework of Draft Code for High Strength

Reinforcing Bars

1. INTRODUCTION

Reinforced concrete (hereafter referred as to "RC") has widely been used for medium rise buildings

because of the low cost, excellent durability and

easy maintenance, so on. However, it Is Impossible to realize the super high-rise buildings and

buildings with long spans which are currently re- quired from the social point of view. If the material strength remains within ordinary ones. The Project alms to produce high strength concrete

of the specified strength of from 300 kgf/cm2 to 1,200 kgf/cm2 ( 30 to 120 MPa ) and high strength

and high quality reinforcing steel bars of the yield strength of 4,000 kgf/cm2 to 12,000 kgf/cm2 ( 400

to 1,200 MPa ), and to develop new fields of RC

buildings by utilizing them. (Fig. 1)

2. RESEARCHES ON REINFORCING STEEL

BARS AND REINFORCED CONCRETE

ELEMENTS

2.1 Research Hems

2.1.2 Fundamental Mechanical Properties of Reinforced Concrete using High Strength

Materials

a. Mechanical Properties of Confined High

Strength Concrete

b. Local Buckling of Longitudinal Reinforcing Bars In Flexure Hinge Region

c. Constitutive Model for High Strength Concrete and Reliability of Non-Linear Finite Element Analysis

d. Strength of Plate Anchorage of Beam Longitudinal bar In Exterior Beam-Column-Joint

e. Development of Standard Bond Test Proce- dure for High Strength Concrete

L Development of Design Equation for Bondsplitting Strength in Flexure Member

2.2 Results

2.2.1 Development of High Strength & High

2.1.1 Development of High Strength & High

quality Reinforcing Steel

a. Development of Technology for Producing

High Strength & High Quality Reinforcing Bars, and their Fundamental Mechanical

Properties.

b. Weldablllty of High Strength Reinforcing Bar

by Gas Pressure Welding Method

c. Weldablllty of High Strength Reinforcing Bar

by Butt-Welding by Special Process

1 Director. Structural Engineering Department, Building Research

institute (BRI) 2 Head. Structural Division, Structural Engineering Department,

(BRi) 3 Head, DInamIc Division, Structural Engineering Department,

(BRi) 4 Senior Research Associate, Large Scale Structure Testing Divi- sion, Production Department (BRi) 5 Senior Research Associate, 2-nd Earthquaite Engineering Divi- sion. liSEE (BR)

6 Research Associate, Stuructural Division, Structural Engineering Department (BRi)

337
Quality Reinforcing Steel

a. Development of Technology for producing

High Strength & High Quality Reinforcing Bars, and their Fundamental Mechanical

Properties

Sample of four different grades of high

strength deformed reinforcing bars had been

produced. The specified yield strength of the grade USD685 are 685 MPa and designed as high strength mild steel with clear

yield plateau. The specified yield strength of the grade USD980 is 980 MPa. It is designed

as high-strength steel without yield plateau. Both of these two grades are intended to be

used for longitudinal reinforcing bars. The

grades USD785 and USD1275 are intended

to be used exclusively for transverse reinforc- ing bar including hoops, stirrups with specified yield strength of 785 MPa and 1275 MPa

respectively. Basic mechanical properties In- cluding (1) chemical ingredients, (2) yield

strength and its scatterings, (3) fracture

strain, (4) ratio of yield strength to tensile

strength, (5) yield plateau length (6) minimum

allowable bend radius were investigated.

b. Weldability of High Strength Reinforcing Bars

by Gas Pressure Welding Method

In order to asses the possibility of splicing by gas pressure welding method, tensile tests and bending tests were carried out with re- spect to grade USD685 D25 reinforcing bars. Gas pressure welding splice could transfer

full tensile strength of the reinforcing bars. It was impossible to bend 180 degrees, be- cause they broke during bending due to

brittle fracture affected by welding heat.

c. Weldability of High Strength Reinforcing Bars

by Butt-Welding by Special Process

High strength welding rod with strength of 1,000 MPa was developed for splicing by

butt-welding method. In order to asses the possibility of butt-welding of the grade

USD685 D25 reinforcing bars, tensile tests were carried out. Butt welding splice could

transfer yield strength but it could not transfer full tensile strength of the reinforcing bars due to brittle fracture of the part affected by

welding heat.

d. Damage due to bending and straightening of High Strength Reinforcing Bar

In order to asses the damage caused during

bending and straightening, tests of 90 degree

bending followed by straightening were car- ried out with respect to D25 with grade

USD685 and USD980. inside radius of bending was 3.6 db, 4.8 db and 8.0 db; db

is nominal diameter of reinforcing bar. Rein- forcing bars with normal deformation of grade USD685 and USD980 showed no damage by appearance after 90 degree bend

as far as inside radius is equal or larger than

4.8 db. Reinforcing bar with threaded defor- mation of grade USD685 could be bend if inside radius is equal or larger than 4.8 db,

while USD980 needed minimum inside radi- us of 8.0 db. To get no damage during

straightening, minimum inside radius of 8.0 db were needed.

e. Strain Aging Property of High Strength Reinforcing Bar

In order to asses the strain aging property

of high strength reinforcing bar, strain aging

tests of grade USD685 reinforcing bars were

carried out. The reinforcing bars were

stretched until the tensile strain attained to 50 % of fracture strain, then strain aging were enhanced by high temperature condition (100 "0 *

1 hour ). The graded USD685

reinforcing bars produced by "in-line quenched and tempered" process showed

no strain aging.

f. Performance of Mechanical Connector for Splicing Reinforcing Bars

High strength threaded sleeves coupler was

developed. The mechanical connector was

intended to be used for splicing threaded de- formed reinforcing bars. In order to asses

the reliability of the splicing system, tensile tests of spliced reinforcing bars were carried

out. The developed splicing system could

transfer full strength of the USD685 reinforc- ing bar.

g. Performance of Members using reinforcing bars with high yield ratio and Splicing

Threaded Couplers

The effect of high-yield ratio and splicing

couplers on the strain concentration at flex- ure critical section are investigated by tests

of half-scale three cantilever beam speci- mens with tensile reinforcement ratio of 0.5 %. Specimens were subjected to statically cyclic load to failure. Figure 2.1 compares

the force-deflection relations of specimens

using longitudinal reinforcements with yield

ratio of 90 % and 75 %. Fracture of longitu338
dinal reinforcements was observed during

the test of specimen using the bars whose

yield ratio was 90 %. The necessity to limit maximum yield ratio for high strength steel

in the code were recommended.

h. Framework of Draft Code for High Strength

Reinforcing Bars Framework of draft code for Reinforcing Bars were proposed.

2.2.2 Fundamental Mechanical Properties of Re- inforced Concrete using High Strengtti

Material -

a. Mechanical Properties of Confined High

Strength Concrete

Analytical models which predict compressive

stress-strain relation for confined concrete

are surveyed form literature. From the point of view of (1) simplicity, (2) correlation. (3) universal applicability from normal strength concrete to high strength concrete, the effec- tiveness of existing models are evaluated

using test results of uniaxial compressive

tests obtained In this research project. Figure 2.2 shows the comparison of the prediction of ultimate strength by typical existing model and the model proposed. The model

proposed showed good correlation for square section column, while the classical model of Richart showed good correlation for circular column.

b. Local Buckling of Longitudinal Reinforcing Bars In Flexural Hinge Region

Twenty four stubbed column specimens using high strength longitudinal bars were subjected to monotonicaily increasing compressive force, in order to investigate the effect of hoop spacing on the behavior of local

plastic buckling of longitudinal reinforcement. Figure 2.3 (a) shows the typical geometry

and the dimensions of the specimens. Two

slits were made between loading steel plate stabs and longitudinal reinforcement so that the compressive force shoukJ be transmitted by steel bars only. All of the specimen

showed the local plastic buckling after compressive yield of longitudinal reinforcement. As Increasing the spacing of the hoops, ulti- mate strength decreased and strength de- graded suddenly after ultimate strength reached as shown In Figure 2.3 (b).

c. Constitutive h^odels for High Strength Concrete and Reliability of Non-Linear Finite Element Analysis Several existing non-linear finite element

analysis codes are selected for analytical

platforms and were Implemented with nonlinear constitutive model considering the mechanical properties of high strength

material. Figure 2.4 shows the typical model

parameters essential for modeling of high

strength concrete. Reliability of the pro- grams have been examined by comparison

between analytical results and tests results of benchmark specimens selected from

member tests carried out In the New RC

project.

d. Strength of Plate Anchorage of Beam Longitudinal bars in Exterior Beam-Column-Joint

Pull-out tests of 56 end plate anchorages

were carried out, which simulate the beam

longitudinal rebars developed In exterior beam-coioumn-joint to Investigate the force

transfer mechanism and to develop design

method of the end plate. Figure 2.5 explains

the concept of the tests and the test

variables.

e. Development of Standard Bond Test Proce- dure for High Strength Concrete The effect of the shape of deformation on the surface of reinforcing bars on tensile splitting bond strength developed in high strength concrete with compressive strength of approximately 100 lutPa were Investigated by 9

simple bond test specimens.

f. Development of Design Equation for bondsplitting Strength in Flexure Member

Bond-splitting strength tests of 26 high

strength concrete beams were carried out to derive empirical design equation for bond

strength which takes into accounts the ef- fects of (1) concrete compressive strength

(2) bond length (3) amount of transverse re- inforcement, (4) number of legs of stirrups or hoops. Design equation was proposed

based on the tests.

3. RESEARCHES ON STRUCTURAL PERFORMANCE OF MEMBERS

3.1 Research Items

339
3.1.1 Flexural Behavior of Columns and Beams

a. Bond splitting failure of columns and beams

after flexural yielding

b. Deformation capacity of columns under

high-axial force after flexural yielding

c. Minimum reinforcement ratio of beams

d. Numerical analysis using micro models

3.1.2 Shear Behavior of Columns and Beams

a. Shear behavior of columns and beams

b. Bond splitting failure of columns

c. Shear behavior of beams with small

openings

d. Numerical analysis using micro models

3.1.3 Shear Behavior of Beam-Column Joints

a. Shear behavior of interior and exterior beam-column Joints

b. Anchorage of exterior beam-column joints

c. Shear behavior of Interior and exterior beam-column joints under bi-directional

loading

d. Numerical analysis using micro models

3.1.4 Shear and Flexural Behavior of Shear

Walls

a. Shear capacity of shear walls

b. Flexural deformation capacity of shear walls

c. Deformation capacity of shear walls under

bi-directional loading

d. Development of analytical models and parametric studies

3.2 Results

3.2.1 Flexural Behavior of Columns and Beams

a. Experimental study on columns were per- formed to investigate the relationship be- tween axial stress and deformation capacity.

Figure 3.1 shows the results.

In Figure 3.1 , ti means axial stress level (r\ =

axial force divided by compressive strength

of concrete and cross sectional area of column).

The specimens S6. S7 and S10 which are subjected to high axial stress {r\ = 0.5 ) have

smaller deformation capacity than the others

(11 = 0.35. 0.15).

Vertical Cracks occurred along the height in the specimen S7 with shear span ratio of 1 .5

( 71 = 0.5 ) ( Fig. 3.2 ).

b. In case of the beams with double layer longi- tudinal reinforcement, bond splitting failure of the second layer longitudinal reinforcement occurred early compared with that of the first layer longitudinal reinforcement.

3.2.2 Shear Behavior of Columns and Beams

a. Shear Strength of columns obtained by the

test is higher than that calculated by A-meth- od of AIJ (1j, but lower than that by B-method

of AIJ [1].

b. Shear strength of the shear failure type of beams can be evaluated by A-method of AIJ

(11-

c. Two specimens with an opening were tested. Figure 3.3 shows the crack pattern of them. The specimen L1 has reinforcement with long

anchorage for the opening, and the speci- mens S1 has reinforcement with short an- chorage for the opening.

3.2.3 Shear Behavior of Beam-Column Joints

a. In the interior beam-column joints of New

RC, coefficient for effective compressive

strength concrete v decreases rapidly and

the joint will fail early compared with the joint by normal strength RC.

b. In the exterior beam-column joints of New

RC, anchorage strength of beam reinforce- ment with double layer arrangement is small- er than that of beam reinforcement with sin- gle layer arrangement.

3.2.4 Shear and Flexural Behavior of Shear

Walls

Five flexural walls were tested to investigate the ductility under bi-directional lateral

loading. Variables in this test are (1) loading

hysteresis, (2) ratio of axial stress. (3) concrete strength, and (4) confinement of col- umns and wall panel.

Figure 3.4 shows the loading system and

Fig. 3.5 shows the loading path. Figure 3.6 shows the horizontal force and

displacement relationships. Deformation

ductility of the shear wall under bi-directional cyclic loading was influenced by the axial

strain of columns and wall panel.

4. DESIGN GUIDELINE

340
Design guideline committee Is In activity to mal<e a

design guideline and design criteria for New RC

buildings, and to show an example designed

buildings. This section shows an outline of a design guideline and related research Items to It.

4.1 Scope

This design guideline Is adapted to a seismic design for an building which height is not more

than 200 meters, and consists of moment re- sisting frame with/without multi-story structural

wall.

Eccentricity of stiffness and mass in the build- ing should be avoided.

Two hundred meters in height is correspond to 60

storied building for residence or 40 storied buildings

for office. Up to 200 meters in height, lateral design load is, In general, determined by an earth- quake load, not by an wide pressure. Following four types of buildings are temporarily designed and

discussed to make a design guideline.

(1) Sixty stories apartment type buildings : Stmctural systems are a moment resisting frame and moment resisting frame and multl story

stnjctural wall.

(2) Forty stories office type buildings : Structural systems are a double tube and tube core

wall

(3) Twenty five stories office type buildings : Structural systems is a moment resisting frame

(4) Fifteen stories office type buildings : Structural systems are a moment resisting frame and moment resisting frame and multl story

structural wall

4.2 KAaterlals

Materials in Zone I are specified in this design

guideline.

Normal strength materials are allowable to be used. Combination of strength of concrete and steel will be provided in Structural Requirement.,

4.3 Design Principles

4.3.1 Design for gravity load

Allowable stress design is adopted.

Allowable compressive stress of concrete is one third of specified concrete strength (Fc). allowable tensile stress of steel is 196 - 215

MPa, and allowable compressive stress of

steel is two thirds of its yield strength.

if necessary, cracking width should be

checked.

Young modulus ratio of steel to concrete, and creep

coefficient should be determined.

4.3.2 Design for seismic load

Against large level earthquake load , any members should not reach their yield strengt^. Against extra large level earthquake load , building drift s^hould be within design drift (1/80

at each story ). 4 Against assuring design drift all member

should be safe.

in order to satisfy these criteria, following design

procedure are proposed.

(1) Standard design base shear coefficient (Co): By static analysis, yield of members shoukJ

not be allowed at Co = 0.2. And Co of 0.25 is given to a building at assuring design drift. Design stress for member is derived based

on the stress at assuring design drift considering the effect of higher mode, fluctuation

of strength, importance factor for members,

and accidental torsion. By dynamic analysis,

above design criteria is also checked.

(2) Design direction:

in general, a building should be designed to have same safety margin in any direction, in this design guideline, two principle direction and one weak direction of building are

discussed. One weak direction may be 45

degree direction. Analytical method in 45

degree direction is now considering and un- der discussion, it is possible to be proposed

as a simple method that flexural and shear

Note:

1 earthquake molion with 25 kine of its maximum veloctly 2 earthquake motion with 50 kIne of Its maximum velocity and

artllk;lal earthquake motion proposed In New RC project (New RC wave). 3 This story limit is from maintaining of structural performance

4 ConsMerIng fluctuation of earthquake characterlsttes, analytteal model, some safety factor Is Introduced. Assuring design drIH Is determined as a factored design drift. This factor has not been

determined yet.

341
and axial stress in 45 degree direction Is 1 .4 times and 2 times stress obtained from analysis in principle directions, respectively.

(3) Estimation of axial load:

Effect of vertical component as well as tw^o directional components of earlfiquake on axial load Is now discussing. But two directional effect wilt be counted by design in 45

degree direction.

(4) Analytical model and method:

Previous dynamic analysis is necessary to determine the adequate design drift, dimensions of members, and lateral distribution for static design etc.. Simple and convenient dynamic analysis model and method is developed.

Dynamic analysis should be done by more

precise model more than a plane frame

model.

Static design should be done by the method

of non-linear incremental analysis using more precise model more than a plane frame model. Three dimensional analysis is recommended.

(5) Multi story stmctural wall

The effect of multi story structural wall on the

vibration mode during earthqual^e motion is now discussing. And three dimensional ef- fect of connecting beam, slab due to the

stmctural wail would be tal<e into the design.

(6) Assuring the deformation capacity of member : Considering a fluctuation of earthqual<e characteristics, analytical model, and gravity

effect, deformation capacity of member

should be assured more than a member deformation at an assuring design drift. Defor- mation capacity of a member is determined by the minimum estimated value.

(7) Torsion : Torsion due to the progress of plasticity of members should be taken into the design.

4.3.3 Design for wind pressure

If the lateral load by the wind pressure is smaller than that by the seismic load at Co = 0.2. design for wind pressure would be omitted.

Design for wind pressure is not debating in Design

Committee at all.

4.3.4 Design for other loads

Load by temperature, shrinkage, unequal set- tlement, accident etc. should be considered.

Design for the loads by temperature, shrinkage, un- equal settlement, accident etc. is not debating In Design Committee at all.

4.4 Design of member

4.4.1 Flexural strength

Flexural strength should be calculated based

on the characteristics of materials. Simple

equation will be provided based on

experiments.

Equation relating to experiments with mean value and deviation will be proposed from structural per- formance committee.

4.4.2 Another strength

Strength of shear, bond, beam-column-joint,

and structural wall will be provided.

Equation relating to experiments with mean value and deviation will be proposed from structural per- formance committee.

4.4.3 Deformation capacity

Members should have more deformation capacity than that at an assuring design drift.

In design, flexural yield and failure is supposed.

This failure model change to compressive, bond, or shear failure at its final limit. Deformation capacity

of a member could be estimated at the point where

failure mode changes.

Based on these deformation capacity, assuring design drift is able to be determined. Both assuring

drift determined by maintaining a building function and by deformation capacity of members should be

adopted. Balanced steel ratio at ultimate strength should be

accounted.

4.5 Earthquake motion for design

As well as recorded earthquake motions, artifi- cial earthquake wave should be used for a

342
dynamic design.

Artificial eanliqual<e wave would be proposed. As

well as recorded motion, this wave should be used

for dynamic design. Mailing method of New RC

wave will be explained. New RC wave is proposed

as the wave on technical soil. Wave at the base of a building should be translated from the wave on

the technical soil considering the interaction be- tween soil and building.

4.6 Foundation

omitted.

4.7 Structural requirements

Expected Items for structural requirements are

listed. Items that can be determined by design will be omitted. For rest of them structural require- ments are determined.

4.7.1 Column and beam

(1) Section (2) Flexural steel (3) Shear reinforce- ment (4) Openings in beam

4.7.2 Wall

omitted.

5. OTHER RESEARCH ITEMS

5.1 Development of a super high rise flat slab type building

5.1.1 Target buildings

Target buildings are shown in Figure 5.1,5.2 and

their stmctural outlines are tabulated in Table 5.1

5.1.2 Literature research On next two items, literature research have been

going.

(a) Comparison research on the strength of the

joint between experimental results and

estimation.

(b) Strength and deformation capacity of a wall.

5.1.3 Experimental work on the behavior of the

|oln of wall or column and slab. On next four experimental series have been

planned.

(a) Inner column and slab joint; 4 specimens

(b) Side column and slab ; 8 specimens

(c) Corner column and slab ; 4 specimens

(d) Wall and slab; 7 specimens

5.2 Feasibility Study on New RC Mega

Structures

Trial Designs of Mega-Structures are carried out by

using New RC materials. ^Strength of concrete is from 600 to 1.200 kgf/cm (60 to 120 MPa ) and

strengt|i of reinforcing steel bar is 7,000 to 1 2,000

l<gf/cm ( 700 to 1200 MPa). Six trial design models are listed below and Figures 5.3-5.8 show them. The structural design criteria which such megastmctures have to satisfy will be defined.

(1) Frame structure of 200 meters height

(Fig. 5.3)

(2) Frame structure of 300 meters height

(Fig. 5.4)

(3) Frame structure of 300 meters height

(Fig. 5.5)

(4) Frame sfmcture with K-brace of 200 meters

height (Fig. 5.6)

(5) Frame structure with one way brace of 200

meters height (Fig. 5.7)

(6) Frame structure with brace of 300 meters

height (Fig. 5.8)

REFERENCES

[1] Architectural Institute of Japan : 'Design Guideline for Earthquake Resistant Reinforced Concrete Building Based on Ultimate Strength

Concept," May 1988 (in Japanese)

343
12000

8000

4000

I 300 600 900 1200 STRENGTH OF CONCRETE

Fc (kg/cni)

ZONE ; TYPE OF RC BUILDINGS & MATERIALS

A: LOW-RISE BUILDINGS IN COMMON

B: HIGH-RISE BUILDINGS CONSTRUCTED IN LAST DECADE

I : HIGH-STRENGTH CONCRETE & REINFORCEMENT

D-l: ULTRA HIGH -STRENGTH CONCRETE & HIGH -STRENGTH REINFORCEMENT n-2: HIGH -STRENGTH CONCRETE & ULTRA HIGH -STRENGTH REINFORCEMENT

ni: ULTRA HIGH -STRENGTH CONCRETE & REINFORCEMENT

Fig.l Strength of materials and fields of research and development

344
Stress-Strain Relationship of Steel - CR9 0.:R9 0

C R 7 5 :

I

2 4 S Tensile Strain , %

10

Fig. 2.1 Test of Cantilever Beams using High-yield Ratio Longitudinal Reinforcing Bars

345
0.0 0.5 1.0 1.5 2.0 2.5 3.0

Fig. 2.2 Model Predicting Ultimate Strength Inclease of Confined Concrete

346
Geometry of Typical Specimen

Fig. 2.3 Effect of Hoops Spacing for Preventing from Locai Bucl<iing of Longitudinai Bar

347
Compressive Strain , mm/mm

A A Stevens

9 U U U U lOJ

Tensile Principle Strain / Compressive Principle Strain

Fig. 2.4 Example of Constitutive Model for Concrete for Non-Linear FEM Analysis

348
Anchorage Device

Steel Plate and Nut

Nut only

Longitudinal Reinforcement . D19 (USD685) or y

Traneverg© Reinforcement Ratio (Pw, ^|

0 % 0.2 % or 0.4 %

Fig. 2.5 Tensile Strength Test of Plate Anchorage Developed In Exterior Beam Column Joint

349
story Drift Angle { xO.Ol rad.)

FiprS.l Horizontal Force and Story Drift Angle Relationship

S6 l/200rad S7 1/lOOrad S8 l/50rad S9 1/lOOrad SIO 1/lOOrad

Flg'3.2 Crack pattern of Specimens ( at Maximum Horizontal Force )

350
1

1

II N

^^^^^ \

"1

1

(a) LI — ^-^ r^i -r-i r-^

1 t ( V

(b) S 1

Fig 3.3 Crack pattern of Specimens ( at Maximum Horizontal Force )

Loading Frame

Reaction Wall

Actuator

( for Horizontal Direction )

Specimen Actuator

( for Vertical Direction )

Fig 3.4 Loading System

f ®

(10) (0

® 1

(Dd)

I

(D®

Fig 3.5 Loading Path

351
R y

-1/50 -1/100 1/100 1/50

'5

I

1 1 1 1 1 1 1

-20 -15 -10 -5 0 5 10 15 20

» 6 y (mm)

(cj Bl-Dlrectlonal Loading ( Y-Direction )

3,6 Horizontal Force and Displacement Relationship

352
Table 5.1 outline of structure

(1 ) Core wall type (2) Wall type

Pu rpose Apar t men t Resor t Hotel

Stories 50 F . B - K P - 2 40 F Height G L + 151. 5m G L + 123m

Story 3.0m 1 F : 4. 5m 3. Om 1 F : 6. Om

Height

Area 41. 8 X 4 1. 8m (typical floor) 1. 090 nrf 16. 2 X 16.2m (core arer)

Concrete Fc= 98(For slab Fc = 56 7 )MPa Fc= 98MPa

Steel a y=980MPa o y=980MPa

o y=295MPa(for slab)

Section H'all thickness 70 cm Wall thickness 40cm

Boundary beam 70 x 90cm Boundary beam 50 x 300cm

Column 90 X gOctn ( in ter lor)

80 X 80cni (exterior)

Slab 20or25cni Slab 20cm

353
RESIDENCE

JLL

41 •tor

2or

ENTRANCE HALL

Fig. 5. 1 (b) Section of core wall type Fig. 5. 2 (b) Section of wall type

354
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357
IIIIII1

.1IiIi
Dynamic Testing of Reinforced Concrete Frames

by

Pamalee Brady' and Steven C. Sweeny^

ABSTRACT

This paper discusses ongoing research to study

the behavior of concrete frame structural systems typical of U.S. military construction,

and ways of improving their resistance to seismic loading. One test program focuses on

reinforced concrete frames with masonry infill. Frame stiffness and its interaction with the

infill is investigated. In-plane and out-of-plane

behavior is being smdied, including tests on

plain infill and infiU repaired both before and

after damage to the wall has occurred. The

second test program focuses on evaluating the

capacity of lightly reinforced concrete frames

and the feasibihty of retrofitting them with

viscoelsatic dampers to improve their seismic

performance. Both beam/slab/column and

slab/colimin details typical of existing three

story concrete frame barracks on the west coast

are being tested with and without the dampers

to study building response. KEYWORDS: concrete frames, dynamic

testing, non-ductiie, repair.

1. INTRODUCTION

Most permanent buildings on Army

installations have been constructed, in accordance with seismic criteria applicable

during their design, to provide some protection

from seismic hazards. However, concern for

the seismic safety of these buildings arises from experiences in recent earthquakes, where

structures such as hospitals, communications

centers, emergency facilities, and other

essential and high-loss-potential facilities have

been damaged by ground shaking or other

seismic hazards. Since the early 1970s the

Army has maintained seismic design criteria that are current with national codes and

standards. After the 1971 San Fernando

Earthquake the Army realized there may be

hazards associated with their existing building

inventory. Research was initiated to develop

methods for evaluating these existing

structures. In 1988 technical manual, TM 5- 809-10-2, "Seismic Upgrading of Existing

Buildings," was published to standardize the

procedure for determining building

vulnerability analytically.

The large inventory of built structures means that reducing seismic hazards in existing

buildings is a complicated and costly issue that must be carefully addressed. The Army alone

currently operates and maintains about 164

installations and major activities in the United

States. More than 127,000 buildings are located on these installations. A phased plan

involving inventory screening, hazard indexing,

Research Structural Engineer, U.S. Army Construction Engineering Research Laboratory, P.O. Box 9005, Champaign, IL, 61826-

9005. ^Research Structural Engineer, U.S. Army Construction Engineering Research Laboratory, P.O. Box 9005, Champaign, IL, 61826-

9005.

359
preliminary seismic safety evaluations, detailed

seismic safety evaluations and development of

seismic hazard mitigation plans has been

initiated by each of the military services to reduce military seismic hazards. These

activities have been accelerated in response to the National Earthquake Hazard Reduction

Program Reauthorization Act of 1990.

The U.S. Army Construction

Engineering Research Laboratory (USACERL)

has performed preliminary seismic evaluations

of buildings at several Army and Air Force

installations in Seismic Zones 4 and 3. The

objective of these evaluations is to determine

whether a building provides an acceptable

degree of safety, or whether a more extensive

detailed investigation is required. At three Air

Force bases in Seismic Zone 4, 56 buildings were identified as significant seismic hazards

requiring detailed analysis and possible

upgrading. Projected cost to bring these

buildings up to current seismic standards is $118 M.

Among the vulnerable buildings, we

identified several standard barrack designs used

by both the Army and Air Force during

extensive construction in the 1950s, 1960s and

1970s. Approximately 80 percent of the

vulnerable facilities evaluated were of two

common structural systems. These systems are

1) concrete frames without infill and 2) concrete frames with full infill, infill with

openings, or partial height infill walls. The

systems have performed pworly in earthquakes

and the American Concrete Institute code

requirements changed significantly in 1971 to address the identified problems. Details of the

reinforced concrete frames constructed prior to these code changes are the primary reasons for

poor behavior of the structural systems and are defined below:

1. Lap splices in columns occur at the base of the columns which for lateral loading is the region of maximum

moment. These splices were also not

well confined by tie or spiral

reinforcing.

2. Positive beam moment reinforcing was

discontinuous at the colimin.

3. Little or no reinforcement was

provided in the beam-column joints

4. Construction joints are located above

and below the beam column joint.

Additionally, infill walls alter the behavior of

the concrete frames. Infills are much stiffer than the surroimding frame and tend to share or resist the lateral loads in the system.

Research has shown that partial height walls

produce a short column effect in the frames

and therefore larger shears in the columns.

Because of the prevalence of

these standard designs, the cost benefit of a study to accurately assess the capacity of the

existing systems with the goal of developing

upgrading solutions to improve their behavior

is extremely high. Only recently has research

addressed assessing the strength and ductility

of existing buildings and dynamic tests have

not been performed on these structural system

types.

USACERL has undertaken two related research programs to investigate the seismic

hazard associated with buildings having these

structural systems. The programs evaluate the

seismic resistance of the buildings' critical elements to determine the extent of any

deficiencies so that practical strengthening

concepts can be developed and their

effectiveness assessed. One program is directed toward conceptually modelling a prototype structure of concrete frames with

masonry infill walls and the other uses scale

models of concrete frame prototype structures.

The prototype structures upon which

the research is based are three story barrack

buildings. There are four prevalent building

configurations with similar structural systems;

in plan they represent H, rolling pin, hammer

head, and T shapes (Figure 1). A typical

360
building of the H configuration consists of

three separate structures each with overall

dimensions of 3566.2 cm by 1178.7 cm (117

ft-0 in. by 38 ft-8 in), (Figure 2). Each

structure is six bays in the longitudinal

direction and two bays in the transverse

direction. Bay width is 586.7 cm (19 ft - 3

in.) and story height is 304.8 cm (10 ft). Exterior transverse frames contain full height

infill walls. Interior longitudinal frames have no infill and exterior longitudinal frames have

partial height infill walls. The balance of this paper discusses the d>'namic testing programs

to evaluate these reinforced concrete frame

systems.

2. Concrete Frames with Masonry Infill

2.1. Objective.

The objective of this project is to conceptually model a transverse frame with

masonry infill to study dynamic behavior,

failure mechanisms, and the effects of repair

and rehabilitation. Two small scale concrete frame models are designed to be used in a multi-year research effort. A strong frame,

with larger beam and columns and ductile

detailing, and a weak frame with smaller

elements and non-ductile detailing will be

tested. The test program for each frame is being conducted in three stages. The first stage is determination of the behavior of the

bare frame, including the effects of initial cracking and column post-stressing on the

behavior of the model. The second phase is determination of the effect of the infiU on

frame behavior and failure mechanisms of the

infill both in-plane and out-of-plane. The fmal

phase is determining the effect of repair and

retrofit of the infill on system behavior.

2.2 Test Specimen.

The strong frame is one-half scale of a frame designed in accordance with Chapter 21

of ACI 318-89 (Figure 3). The columns are 15.2 X 15.2 cm (6 in x 6 in), and 88.9 cm (35

in) from base to the center of the beam-column

joint. Longitudinal reinforcement consisted of

6 - #3 grade 60 deformed bars. Stirrups are formed from 6 gauge basic wire. A 1.6 cm

(5/8 in) brass tube runs down the center of the

column to accommodate the post-stressing

cable. The beams are 12.7 cm wide by 15.2 cm deep (5x6 in), 1 37.2 cm (54 in) center to center of the beam-column joints. Each beam

has 3 - #3 grade 60 deformed bars top and

bottom and 6 gauge basic wire stirrups. In

addition, the beam has an unsymmetric top

flange 5.1 cm (2 in) thick extending 24.1 cm

(9.5 in) from the edge of the beam. Both

beams and columns extend 15.2 cm (6 in)

beyond the joint to provide for continuity of

the reinforcement. Concrete is a 2.7 kN/cm'^

(40(X) psi) mix with a maximum aggregate size

of .95 cm (.375 in). The weak frame has the same clear dimensions for infill as the strong

frame with smaller beam and column

dimensions and non-ductile reinforcing. The

columns will be 12.7 x 12.7 cm (5 x 5 in) with

4 - #3 bars and 6 gauge wire stirrups. Final

design of this frame has not been completed.

To facilitate dynamic testing on the shaketable two identical frames in parallel are tested (Figure 4). The two frames are coupled

with a reinforced concrete slab. The slab also

provides for attachment of 25.4 kN (5.7 kips)

additional mass during the in plane tests. Diagonal tension cross bracing is provided at the ends between the two frames to prevent

torsion during in plane and to provide all lateral support out of plane. Columns are poststressed during testing to simulate dead load

stresses.

Instrumentation consists of an array of

strain gauges, displacement transducers, and

accelerometers. One frame is heavily

instrumented and the second frame lightly

instrumented to verily similar response

between the two frames. Strain gauges are mounted on longitudinal bars at the base of the

columns and in the beam near the beamcolumn joints. Two gauges, at 2.5 and 10.2

361
cm (1 and 4 in) from the base or joint, are mounted at each location. At each strain gauge

pair displacement transducers monitor cracking

and curvature. Global displacement

transducers are attached from a fixed reference

frame to the model at the base, mid-height of

the column, mid-depth of the beam, and at

. mid-width of the slab. Displacement

transducers within the shake table actuators are used to measure table displacements.

Accelerations are measured at the base, midheight of each column, mid-span of the beam,

center of the slab, the top of the mass plates

and at the shake table. Accelerations will also

be measured at various locations on the infill when tested.

2.3 Test Procedure.

Dynamic testing was conducted on the

USACERL Biaxial Shock Test Machine

(BSTM). Three tests are used to measure the

behavior of the model under dynamic loading;

a) a random frequency sine wave, b) free vibration, and c) a time scaled El Centro

record. The frame is initially subjected to a low level random sine wave motion to determine the natural frequency and to approximate the dynamic amplification of the

base motions. The random input is increased

until a significant response is measured. This

input level is used for all subsequent random

vibration tests. Next, free vibration of the

frame is measured. The model is excited with

a sine wave at a frequency slightly higher than

the natural frequency. Input motions are increased again until significant response is measured. The table motion is then stopped

and the structure allowed to oscillate freely.

The final test is an El Centro record with a time scale factor of 1.414. For the bare frame

only, each of the three tests are conducted at four levels of post-stress applied to the

columns. The frames are tested at poststressing forces of 13.3, 26.7, 40.0, and 53.4

kN (3, 6, 9, and 12 kips). For frames with

infill, only the 53.4 kN post-stressing will be

used.

The frame or infill are damaged by

subjecting the system to increasingly stronger

earthquake ground motions until significant

cracking develops. It is the intent to destroy

the infill, not the frame, and to change the

response by cracking the frame or infill without yielding the reinforcement. In order to achieve the necessary forces to crack the system, the time scaled El Centro record had to be filtered to remove the large displacements

induced by low frequency motion. A high

pass filter with a cutoff frequency of 2 Hz was

used to produce the necessary seismic record.

This input motion was used at increasing levels

imtil significant damage was observed.

Random vibrations are input after each test to measure change in model response. After

cracking the bare frame, the model is subjected

to the same three tests at each of the four

levels of column stress as before cracking to evaluate the effects of cracking on bare frame

response.

2.4 Initial Observations.

The Pre-cracking response behavior was as expected for a linear elastic system.

The natural frequency for all levels of poststress remained at approximately 11.6 Hz. The

structure showed very good single degree of

freedom response. Damping was approximated

by both log decrement method ftt)m the free vibration curves and the half-power method

from the random vibration curves with good

agreement. Damping for the uncracked model

for all levels of post-stressing averaged of

2.2%. Response to the El Centro record was

also consistent for all levels of post-stressing.

The structure was tested with the

filtered El Centro record at four levels. The

first two tests gave a peak acceleration

response of .43g and 1.08g resf)ectively with

little indication of damage. The third test showed a peak acceleration of 1.62g and a peak displacement of 3.0 mm (0.12 in). Random vibration and ft-ee vibration tests indicated a significant shift in the natural

362
frequency. The fourth test was performed at an input level 25% higher than test 3. Peak

acceleration was measured at 1 .72g with a peak

displacement of 4.8 mm (0.19 in). Again,

random and free vibration tests indicated a

significant change in natural frequency as shown by the transfer funcition magnitudes in Figure 5, and the model was considered

sufficiently damaged.

Initial cracking on the structure had an

effect on both stiffness and damping.

Damping calculations by both log decrement

and half-power method show that the damping

coefficient change from 2.2% to 4.3%.

Working backwards through the equations of

motion for a damped system using the damping

and natural frequencies measured in the tests show that the stiffness of a single frame

changed from 82.0 to 44.8 kN/cm (46.8 to 25.6

kips/in).

After cracking, the model was again

subjected to the random vibration, free vibration, and unfiltered El Centro tests at the same input level run on the pre-cracked model

at the four levels of post-stressing. The postcracking behavior was much different then the pre-cracked behavior. A slight increase in natural frequency with increasing post-stress was observed. Again, both the log decrement

and half-power methods for computing

damping give good correlation. While the values for damping of the cracked frame vary, there is no correlation with post-stress values.

The average value for the coefficient of

damping in the cracked frame is 4.3%.

Calculation of frame stifftiess gives more

consistent results. The drop in stiffness after cracking is significant, and both pre and postcracked results indicate a slight trend towards

increased stiffness with increased poststressing, which was expected.

Most important for structural engineers

is how the shifts in stiffness and damping

affect the behavior of the structure during an

earthquake. Typically, increased damping is

considered beneficial for energy dissipation. In

this case however, the effect of the decreased

natural frequency caused an undesirable shift towards the predominant frequencies of the El

Centro N-S seismic record. The result was a

significant increase in peak acceleration and

displacement of the cracked model.

3. Lightly Reinforced Concrete Frames

3.1 Objective.

The objective of this research program

is to evaluate the effects of the vulnerable

details identified above on overall behavior of

nonductile concrete frames by experimentally

assessing the capacity of scale models drawn

from two prototype structural systems. With

this knowledge we may better assess the complete capacity of the frame systems.

Additionally we wiU test the effectiveness of

viscoelastic (VE) dampers to improve the

behavior of concrete frames under earthquake

loading. The program is being worked

cooperatively with Professors Wood and

Foutch at the University of Illinois. It will be

conducted in three phases. In the first phase,

currently ongoing, tests of scale model frame

subassemblages will be conducted to study

their behavior and assess their capacities. A

test of a scaled three story nonductile structure will be conducted in phase two. The third phase of the project will be devoted to technology transfer. Conventional

strengthening schemes will be compared with

upgrading using VE damper technology and

the cost-effectiveness of different solutions

evaluated. The following sections of this paper

will focus on the first phase of the program,

assessing the capacity of the existing frames

through testing of frame subassemblages.

3.2 Test Specimens.

A total of four specimens representing

the critical second floor joint of two lightly

363
reinforced concrete frame systems will be

constructed. Two 6/lOth-scale models of a

typical beam/slab/column joint have been

tested on the USACERL Biaxial Shock Test Machine. Two 1/2-scale flat plate/column

joint specimens will be tested this simimer on

the shaketable. Geometric similitude between

the prototype and test specimens was

maintained. Subassemblage members were

reinforced to produce the same

demand/capacity ratios as those of the

prototype elements.

The beam/slab/column test specimen measures 243.8 cm by 284.5 cm (8 ft-0 in. by

9 ft-4 in.) in plan and is 182.9 cm (6 ft) tall. The column is 18.4 cm (7.25 in.) square. The

slab is 7.6 cm (3 in.) thick and the beam

extends 16.5 cm (6.5 in.) below the slab. (Figure 6) Reinforcement details of the subassemblage follow ACI 318-51 code

provisions. The longitudinal column

reinforcement of the model consists of 4 - #5

bars; the bars are spliced 20 d^^ at the top of

the slab. Transverse reinforcement is provided

by #2 ties spaced at 17.8 cm (7 in.) o.c, the

minimum column dimension, along the height

of the column. The ties have 90 degree hooks.

Longitudinal beam reinforcement is 4 - #3 bars

top and bottom. Bottom bars were

discontinuous and extend 8.9 cm (3-1/2 in.) into the column. This extension provides

similitude with the positive moment capacity of

the prototype structure based on available bond

strength. Where bars were continuous in the prototype structure, such as beam centerlines

and midheights of colimins, bars were hooked

in the test specimen to ensure development.

The transverse reinforcement in the beams was

provided by #2 bar open stirrups spaced 7.6

cm (3 in.) o.c. arranged typical of gravity load

conditions; they did not extend the full length

of the beam.

The flat plate/column specimens have

nominal dimensions of 304.8 cm by 304.8 cm

(10 ft-0 in. by 10 ft-0 in.) and are 182.9 cm (6

ft) high. The column is 18.4 cm (7.25 in.)

square and the slab is 9.2 cm (3-5/8 in.) thick. (Figure 7) The reinforcement consists of two

mats of #2 perpendicular bars at the top and

bottom of the slab. Standard spacing within

the column strip was 7.6 cm (3 in.) o.c. and

12.7 cm (5 in.) o.c. in the middle strip.

Column reinforcement was the same as that of

the beam/slab/colimin model.

To maintain gravity stresses and

produce the appropriate reactive mass in the

models their mass was augmented by steel weights. A total of 26.6 kN (6 kips) was

added to the beam/slab/column model, 13.3 kN

(3 kips) at the top of the column and 13.3 kN

(3 kips) on the slab. Inflection points of the

columns and beams were assumed to occur at their mid-height and mid length respectively.

The test fixture provides for rotation at these

points. (Figure 8)

The joint region of each model is instrumented

to measure curvature of the beam/slab,

curvature of the colimin above and below the

beam/slab and shear in the joint and column.

Global displacements are measured from a fixed reference frame at the top of the column,

slab and base of the column. Accelerations are measured at the top of the column, slab and

base in the direction of load and at the slab in

the transverse direction.

3.3 Test Procedure.

White noise and free vibration tests are conducted on each model to determine the

characteristic frequency and damping of the

model at various points in the test program.

The El Centro 1940 N-S record, time scaled by

a factor of 2.0, is used to excite the model at increasingly greater magnitudes to produce

damage in the model. The specimens are

tested to failure. The effects of structure geometry and configuration, yield and collapse

capacities and energy dissipation characteristics are evaluated. A viscoelastic damper is incorporated into the second model of each

type and the influence of the increased viscous

364
damping on the elastic and inelastic dynamic lightly reinforced concrete frames to improve

response of the structure investigated. their seismic behavior.

Parametric studies will assess the differences in performance of the models with and without

dampers.

3.4 Initial Observations.

Tests of the beam/slab/column

specimens with and without dampers have been

conducted. The failure sequence of the lightly

reinforced joint was as expected; first, the

discontinuous positive reinforcement slipped

from the joint. This was associated with a period shift in the model. As the tests continued a hinge formed at the base of the

upper colunm, indicating slip in the column lap

splice, and the joint region of the beam/column

deteriorated increasingly. The tests were

stopped to prohibit failure of the now unstable

upper colimin. Damage to the joint also made

the structure extremely flexible. Capacity of

the joint was generally as predicted.

A preliminary review of the data from

the tests with the viscoelastic dampers shows

reduced accelerations and displacements of the

structure as compared with the specimen

without dampers for the same input excitation.

After testing with the dampers, they were removed from this model and it was tested at the peak excitation of the first model. The

failure mode was similar but more dramatic.

Further study of the data from both

tests is required to evaluate the response of the

subassemblages, the improved response due to dampers and the effect of the test sequence on model response. Similar tests will be

conducted this summer on slab/column

subassemblages with and without dampers.

More detailed analyses will be compared with

experimental results. The work summarized

above constitutes the first phase in a three phase program. Results of this work will be

used to develop a better imderstanding of the

behavior of the structures and to develop

simplified design approaches for upgrading

365
•T' ROLLING PIN

•H' HAMMER HEAD

Figure 1. Standard building configurations,

rr

- 39'e'

PARTIAL PLAN

39B1

BUILCING PLAN

.— it >9f —

SCCTIDN Q-Q 1SCCTIDN b-b

Figure 2. Typical "H" building dimensions.

366
Figure 4. Assembled model with masses and bracing

367
V

9

25

20

S 15 g

t>- 10

^ 5

Test 2 (uncracked)

Test 4 (cracked

Test 3

) fi

5 10 Frequency [hz]15

Figure 5. Change in the response due to cracking.

r-<r t

#ys. 4'-r • <ac. BOTTOM

#ys. y-3/r <o.c. top

*'OjC. bottcn

-I I- 7 V<

Figure 6. Beam/Slab/Column Subassemblage.

368
Figure 8. Test Configuration.

369

li

Shaking Table Collapse Tests of RC Columns

by

Chikahiro Minowa^ Nobuyuki Ogawa^ Tadashi Mikoshiba\ and Norio Oyagi*

SUMMARY

Precise control of shaking table should be desirable

for aseismic tests of structures. Large-scale shaking

table of National Research Institute for Earth Science

and Disaster Prevention (NIED) has employed the

digital control system Ho, In order to verify this digital control characteristics, the collapse tests of reinforced concrete columns were conducted. The reinforced concrete column damage pattern

differences between static cyclic loading tests and

dynamic shaking table tests were observed. In this study, the spacing of shear reinforcement bars in columns were variable. Four types of spacings of shear rein- forcement bar spacing are used. As results of the tests,

it was verified tnat the maximum horizontal strengths of columns, the deformation capacities, and the damage

patterns were changed with the shear reinforcement

oar spacings, and also with testing method differences between statics and dynamics. Furthermore, P- 5 effects were discussed. KEY WORD: Shaking Table, Digital Control, Analog

Control, Reinforced Concrete, Dynamic Collapse

Tests, Static Cyclic Loading Tests, P- S Effects

1 1NTRODUCTION

Shaking Tables were used for the tests of base

isolation methods, vibration control methods, liquefactions, and verifications of aseismic design codes, and

so on. Some of these tests shall require the precise

reproduction of earthquake records. Preparatory

excitations are permitted for elastic region tests of

structures. However, preparatory excitations can not be used for liquefaction tests and structure damage

tests. Recently, NEID employed the distal control system using DSP board in order to miprove the response properties of the large-scale shaking table. The collapse tests of reinforced concrete columns

(model scale: 1/3) were carried out so as to verify this digital control method.

Many studies concerning about the ultimate hori- zontal strengths of coliunns were already conducted. However, most of these studies were carried out by the

static cyclic loadings of one column, or psuedo-dvnamic

loadings. In this investigation, realistic dynamic loading

which would simulate earthquake responses of structures, were done by the use of a shaking table, and the dynamic damage characteristics of columns were

observed. A small test structure of one story and one

bay had four columns. These columns were made by

reinforced concretes. For kind of spacings of shear reinforcement bars were used. In addition to the shaking table test, static cyclic loading test had been

done on behalf of the comparisons. The time variations of damage patterns, the ultimate strength and the story deformations were observed.

2 TEST METHODS

Four kinds of reinforced concrete columns were

tested on the shaking table. The over view of shaking

table tests is shown in Photo 1. Analog control methods

were used two times, and also digital control methods

were used two times. Additionally, steel column tests were conducted. The new digital control system consists of a personal computer and a digital signal processing

(DSP) board. The control methods are selected on a key board. The outline of this control method is shown

in Fig. 1. In the case of an analog control, the H2

controller shall be bypassed, and this system is used as a wave form generator. In this test series, the shaking

table were drive by the strong earthquake motion

record of 1968 Tokachioki, Hachmohe E-W, which was mode in the use of FFT band pass filter 0.4Hz-50Hz.

The real time scale was adopted. Fig. 2 shows the main

records of shaking table responses in the analog and

digital controls. This digital control methods would give the improvement of gain characteristics and robustness

to the sheiking table. The control frequency increased from lOHz to 13Hz with this digital control method.

The digital control responses are appeared to be

inferior to analog control responses in time domains. It will be necessary to consider the nonlinearity of servo

valve in the H2 controller. One story X one bay test structure was assembled on the shaking table. Four columns had the section of 13cm X 13cm, and the height of 85cm. The diameters

of four main steel bars were 16mm. The columns were

fixed in steel roof girder and steel base frames with

bolts. The roof of test structure was made by steel frame. On the roof, three concrete mass weight were fixed. Total weight of roof was 27.8ton. The spans of this test structure were 2m in loading direction, and 2m in perpendicular direction. In order to mitigate the shock

loads of falling roofs in the shaking table during damage

excitations, the roof support frame was installed inside of the test structure.

1 Former Director, Disaster Prevention Division,National Research Institute for Earth Science and

Disaster Prevention, Science and Technology Agency

2 Head, Earthquake Engineering Laboratory, ditto 3 Senior Researcher, ditto 4 Research Member, Earthquake Engineering Laboratory, ditto

371
Therefore, the roof of 27.8ton fell about 15cm in damage excitation. The four types of columns with the shear reinforcement bar spacing of 50mm, 105mm,

170mm, and 285mm were tested in the shaking table. The outline of the test structure are shown in Fig. 3. The details of colunms are shown in Fig. 4, and the

physical properties of columns are concrete compressive strength 230kg/cm2 - 300kg/cm% and steel bar

tensile strength 3600kg/cm2.

In a static cyclic loading test, the pull type oil jacks

which were connected in the roof frame, were used. Two jacks were set in right side, the others were set in

left side. Cyclic loading level were 1/1000, 17200, 1/100,

1/50, 1/20 in story deformation angles. After cyclic test, loads increased gradually until column collapses. The

one dimensional shaking table of NEED was used in this

test. The dimension of shaking table is 12m by 12m. The

table weight is ISOton. The maximum power is 360tons. The maxunum velocity and displacement are 75cm/s

and 24cm. The acceleration levels of these tests were

0.5G and the velocity were 70cm/s approximately.

CoUapse responses of test columns were observed by

video records, and measured by various sensors.

3 STATIC TEST RESULTS

Two kinds of columns were tested. In the static cyclic

loading test of column @285ram, the shear cracks appeared in the region between about 10cm and 40cm

from two column ends at the cyclic loading of defor- mation angle 1/100. At the cycuc loading of 1/50, the shear cracks increased diagonally, and, the main steel bars of two colunms in XI frame were buckled at the middle height, with the separation of concrete and main

steel bars, at last the root fell down. In the static cyclic

loading test of column @50mm, the bending cracks concentrated in the region from two colunm ends till 10cm or 20cm positions at cyclic loading of 1/50 and

1/25. At cyclic loading of 1/5, the compressive crushes

of two column ends began to grow immediately. At last, the concrete of two column ends were crushed com-

[)letely, and the main steel bars were buckled. The

oad-Displacement hysteresis loops of these two static cyclic loading tests are shown in Fig. 5 and 6. The crack

sicetches of tnese loadings are shown in Photo 2.

In the shaking table test of @285mm, initial shear

cracks grew immediately, and three columns were cut near column tops, the other one was cut near a column

bottom. The main steel bars buckled.

In the test of @170min, the initial crack types were same to @285 test case. However, the columns col- lapsed in next opposite acceleration with the separation

of concretes and steel bars, and with the crush of core

concretes.

In the test of @105nmi, the shear bending cracks were appeared near two column ends. These cracks increased and collapsed with the separation of concretes, and with buddings of main steel bars. In the test of @50mm, the initial shear bending

cracks appeared in two column ends. These cracks grew

gradually, and the compressive concretes crushed with main steel bar buckling. The crush re^on of columns were limited in column ends.

5 P-S EFFECTS

The Acceleration - story displacement hysteresis

loops of Fig. 7 shows the effects of roof weights: that is P- 6 effects. The reinforced concrete columns were

collapsed without hysteresis loops which would be

enough to analysis. Therefore, the data which were

recorded in preparatory tests using steel columns(Photo 7), were analyzed. The steel columns

have the H shape section of 100mm x 100mm x 6mm x 8mm, and the height of Im. The frame conditions were same to the reinforced concrete test cases. A simple model with rigid girders, shown in Fig. 8,

is considered. The equations of this model are expressed as follows.

m ( V + ) - m ( u + + B ) + v - 0

V

{ V + v„ )

: + m ( tt + tt + g ) " h + u

/ ;—;•

» ^( h+ „

« h + tt

4 DYNAMIC TEST RESULTS

Four kinds of columns were tested in the shaking

table. The natural frequencies of the elastic regions of each test structure was 4.5Hz approximately. The

damping ratios in elastic regions was estimated 5% - 10% by AR model analysis. Distinct differences between four test structures were not found. The plastic

region dominant frequencies of each test structure was

about 3Hz at the time just before collapses. Four test structures collapsed at the begiiming of main wave.

Acceleration-story displacement hysteresis loops of

four shaking table tests are shown in Fig. 7. Damaged

columns are shown in Photo 3,4,5 and 6.

The kij v term of first equation will be replaced by

horizontal restoring force characteristics of columns

f(v). The outline of f(v) is indicated in Fig. 9. The

responses were calculated with these models. Parameter values were assumed as XI = 1cm, X2 = 1.3cm,

BKl = 9ton/cm, BK2 = 7ton/cm, BIG = 0.2ton/cm,

BK4 = 5ton/cm, h = 80cm, m = 0.027ton/gal,

kg = 750ton/cm. Damping ratios were assumes as 0.04

for horizontal responses, and 0.01 for vertical responses. Fig. 10 is the simulation responses with the consideration of P-3 effects. Fig. 11 didn't consider the P- effects. Up-down components of input were thought

372
as noises which were generated by shaking table mechanisms. The comparison of two cases indicates the needs of P- d analysis for reinforced concrete tests.

6 CONCLUSION

The dynamic maximum horizontal strengths and the dynamic deformation capacities were almost agreed

with static ones. However, the damage patterns were

changed with the shear reinforcement bar spacing, and

also with test method differences between statics and

dynamics.

P-S effects are important for collapse properties of columns.

The shaking table digital control methods improves

the shaking table gain characteristics, slightly.

ACKNOWLEDGMENT

This experiment was started with the suggestion of Professor Heki Shibata, University of Tokyo. These

tests were carried out with cooperations of Tokyu

Construction Co.,LTD.The authors wish to express tne appreciations to them.

REFERENCES

Maekawa, A. (1991), Application of H2 Control to Electric-Hydraulic Shakmg Table, JSME D&D Conferences Kobe (In Japanes)

Minowa, C, (1985), One Story Frame Collapse Tests by Using the Two Dimensional Shaking Table (Horizontal and Up-Dovra), Report of NRCDP No.34

373
Analog Control SysUm

Hj Digital

Controlltr

Analog

Controller

Tabit

Fig. 1 Outline of Shaking Table Control

Fig. 2 Shaking Table Responses

374
500 4000 500

WEIOFTT 27ton

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STEEL PLATE GIRDER

J LJL

' " fesa^

SAFETY FRAME

TEST

COLUMN

SHAKING TABLE

m

1500 1

2000 1500

Fig. 3 Outline of Test Structure

375
oCJ

I I L

-40.00 0.0 40.00

displacemenl (mm)

Fig. 5 6285mm Static Loops

o

O

o

o

«*-'

c?o

Ooo

* 1 / fff

1 -200.00 0.0 200.00

displacement (mm)

Fig. 6 650mm Static Loops

376
relative disp.

accel

.

relative dlsp.

accel

.

900gal

265. 6iTwn

» t I I

0 SOmin

606.2gal

265.9nin

t I I I I

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relative disp. relative disp.

accel

.

accel

.

1024.1 gal

81^0""" 8 28 5inin

Fig. 7 Dynamic Loops

377
t: u

7777

E: YOUNG MODULUS

I :GEOMETRICAL MOMENT

OF INTERIA OF COLUMN

m:MASS OF TEST STRUCTURE

A: SECTION AREA OF COLUMN

Q: SHEAR RESISTANCE OF

COLUMN

N: AXIAL RESISTANCE OF

COLUMN

Fy: HORIZONTAL EXCITATION

FORCE

F..: UP-DOWN EXCITATION

FORCE

k^=2. J2EI/h' , ;t^=2-EA/h

Fig. 8 Analysis Model

RESTORING FORCE

STORY DISPLACEMENT

Fig. 9 Restoring Force Model

378
Fig. 10 With P - (5

I

Fig. 11 Without P - S

379
Photo 1 Over view of Test

2 m5m (Static) • photo 3 @285nmi (Dynainic)

380


A Hollow Clay Tile Wall Seismic Performance Program Overview

by

J. E. Beavers, W. D. Jones, and W. C. T. Stoddart^

ABSTRACT: An overview of a multiyear

hollow clay tile wall (HCTW) program being

conducted by Martin Marietta Energy Systems,

Inc., at the Oak Ridge Y-12 Plant,' for the

U.S. Department of Energy is presented. The

purpose of the HCTW program is to determine the load capacity of unreinforced infilled HCTW buildings when subjected to earthquakes. Progress to date tends to indicate that

extensive retrofit of such structures may not be

warranted in low-to-moderate seismic zones.

1. INTRODUCTION

Many structures in the United States, especially the central and eastern United States, are constructed of unreinforced masonry. One

type of unreinforced masonry, hollow clay tile (HCT), was used extensively as an infill during

building construction throughout the United

States, especially during the early to middle

1900s. As a result, the U.S. Department of

Energy (DOE) has a significant inventory of

buildings with infill walls constructed of

unreinforced HCT. These buildings do not

meet today's seismic standards for new

construction.

in the spring of 1990 and includes in situ,

laboratory, full-scale building, and shake table

tests; nondestructive evaluation (NDE) development; and analytical studies. The program is focused on the goals of providing information

and guidance in the use of this information for

assessing the behavior and, if required, identifying cost-effective methodology for retrofitting these buildings.

Nineteen types of in situ and laboratory tests are being performed, ten of which are both in situ and laboratory. The total number of tests to be performed is approximately 110 in situ

and 190 laboratory. The NDE program consists of developing techniques to determine

HCTW construction quality and infill boundary

conditions. The analytical portion of the

project consists of research on existing analysis

methods and on the development of analytical

evaluation techniques for buildings and

components.

Building analysis consists of modeling and

analyzing entire structures. Potential analysis

techniques are either equivalent static analyses

Bennett and Flanagan (1992) conducted an

extensive literature review that revealed no

relevant research on the performance of

hollow clay tile walls (HCTWs) under seismic

loading. To determine the seismic performance

of HCT buildings, the Center for Natural

Phenomena Engineering (CNPE) of Martin

Marietta Energy Systems, Inc. (MMES), is conducting an HCTW testing and analytical

program for DOE. The program was initiated

^ Presented at the Tenth World Conference on Earthquake Engineering, July 19-25, 1992,

in Madrid, Spain.

^ Center for Natural Phenomena Engineering Martin Marietta Energy Systems, Inc., Oak Ridge, Tennessee, 37831.

' Managed by Martin Marietta Energy

Systems, Inc., for the U.S. Department

of Energy under contract number

DE-AC05-84OR21400.

383
(such as the UBC code Rw factors) or dynamic methods, such as time history analysis.

Modeling of the HCT infills may need to be

simplified, using, for example, equivalent

struts. Component analysis involves developing an appropriate constitutive model for the HCT

and an appropriate idealization. All components that are tested, such as prisms, in-plane

infills, and out-of-plane infills (air bag tests), will be analyzed, and parametric studies will be

performed. It is expected that this approach

will lead to the development of simpler ideali- zations that can be used in the analysis of

entire building structures.

Following the Loma Prieta earthquake of

October 18, 1989, a significant number of

buildings constructed of HCT in the San

Francisco area were found to be damaged, in- cluding the city hall of Oakland, California. In

addition, those that were not damaged or experienced little damage are being examined

for the need for retrofit. The results of this HCTW program will provide significant insight

into the performance of such structures and

provide a methodology for evaluating HCT

buildings that better represents their

performance. In fact, in low-to-moderate

seismic zones, the results may indicate that,

although many such structures were not designed for seismic loads, major retrofit may not

be required, as might be indicated from using

conventional analysis techniques in the

evaluation process.

2. SPECinC GOALS AND BUILDING

CODES

As noted above, the general goal of the

HCTW program is to develop recommendations for assessing the behavior of buildings

constructed of infilled, unreinforced HCTWs.

The program focuses on HCTW buildings

located in zones of low-to-moderate seismicity,

although the results are directly applicable to high seismicity zones. If infilled, unreinforced

HCTW building construction can be shown to

have inherent capacity above that normally

assumed by design codes, retrofit may not be

required in low-to-moderate seismicity zones if a risk-based approach is adopted. A risk-based

approach is based on the principle that, while

the building may not meet today's codes and

standards requirements, the expected seismic

performance is reasonably understood and has

sufficient margins of safety. To help accomplish these goals, testing and improved

understanding are required.

A second principle that has an impact on the

issue of whether or not to retrofit in low-tomoderate seismicity zones is the cost increment for retrofitting. In many cases, the cost

for retrofitting a building to a higher seismic

capacity rather than to a lower seismic capacity

is not much more than the cost for retrofitting to the lower seismic capacity; hence, the incremental cost may be very low. For example, the

cost for installing a l-in.-diameter (25.4-mm)

bolt required for the retrofit of a building in a moderate or high seismic zone is basically the same cost as that for installing a 0.5-in.- diameter (12.7-mm) bolt in the same building

for retrofit in a low seismic zone. Thus, the

most significant cost impact for retrofitting

buildings in low-to-moderate seismic zones is not the seismic capacity level of retrofit, but

the decision, or requirement, to retrofit.

Building codes in the United States generally

specify that the capacity of unreinforced

masonry, due to its brittle nature, be limited to a stress less than the stress that causes

cracking. However, the overall seismic capacity

of an infilled masonry building panel can be an

order of magnitude higher than its load

capacity when the first crack occurs or as indicated by baseline design codes. This leads

to a key proposition in the risk-based approach

to a "no retrofit philosophy" for existing

construction in low-to-moderate seismic zones. This proposition is that, even though HCTWs

may crack when earthquakes occur, overall

stability (no collapse) will be maintained. Thus,

significant cracking of infilled walls is being

384
described by the authors as acceptable damage

for buildings in low-to-moderate seismic zones. Therefore, the principal basis of the research

and development program on HCTWs

described in this paper is to investigate and

develop understanding and computational

methods for determining post-cracking

behavior, including ultimate load capacity, of

infilled, unreinforced HCTW building construction.

3. HCTW PROGRAM DESCRIPTION

The HCTW program at MMES is divided into

four elements of work in a classical engineering work breakdown structure (WBS),

shown in Figure 1.

WBS 1.2 is HCTW testing which represents

the part of the program that includes static, in

situ, and laboratory testing of HCTWs and wall

components. WBS 1.3 is analytical modeling

and analysis, which consists of research of

methodologies, pre- and post-test analyses, and

the development of evaluation methodologies

that incorporate the results of all research

efforts. WBS 1.4 is the seismic shake table

facility, which is designed to show that the

evaluation methodologies selected, based on

static testing, are conservative when tested

dynamically. Thus the evaluation methodology

selected will provide a better, but still conservative, seismic performance estimate of

HCTW buildings. The final element of the

HCTW program is WBS 1.8, NDE research

and development. For proper evaluation of an

existing HCTW building, it is imperative that

the existing condition of the HCTWs and the

boundary conditions between the HCTWs and

the structural framing be determined. Thus, a

significant amount of effort in the HCTW

program is the establishment of NDE

techniques that can correlate controlled

laboratory-constructed specimens to in situ wall construction and that can be relied on to properly define key parameter inputs to the

evaluation methodologies.

4. DESCRIPTION OF HCTWS

The specific infilled, unreinforced HCTWs

studied in this program are constructed of

multi- or single-wythe HCT, usually 8 in. (203.2 mm) or 12 in. (304.8 mm) in width,

using running bond, with all cores laid horizontally. Double-wythe walls in this program are constructed of 4-in. (101.6-mm) and 8-in. (203.2-mm) HCTs staggered to give over- lapping head joints, as shown in Figure 2. The

4-in. (101.6-mm) and 8-in. (203.2-mm) tiles are

also staggered from course to course. In this type of construction, no continuous collar joint

exists; however, in each row, the 4-in. (101.6-mm) and 8-in. (203.2-mm) tiles are separated by a 0.5 in. (12.7 mm) to 1 in. (25.4

mm) gap. In forming the full wythe bed joint,

some mortar has fallen into this gap. There is no vertical or horizontal reinforcement in the

walls. The HCTWs are placed within the steel

and concrete framing with varying offsets to the framing centerline depending on the particular building. In some cases, HCT pilasters,

integrated with the continuous HCTW, are

actually framed around a column as the

HCTWs transverse the length of a building.

5. PREVIOUS HCTW PROGRAM

ACTIVITIES

Aspects of the HCTW program have been

going on since 1989; however, the program

was not fully developed, as shown in Figure 1, until mid-1990. Beavers, Bennett, and

Flanagan (1991) discussed an earlier version of

the program and referred to some test results,

while Fricke and Jones (1990, 1991) described

the results of some of the earlier tests of

element WBS 1.2, defined as a subelement

WBS 1.2.1, push-test. Nine push-tests to measure the shear strength of the mortar bed

joint were reported. The average shear

strength was found to be 99 psi (683.1 kPa),

with a standard deviation of 32.5 psi (224.3

kPa). Butala, Jones, and Beavers (1991) have

described some of the work in WBS 1.3.2,

385
pre- and post-test analyses in preparation of

testing activities of WBS 1.2.14, wall-column

bending. By examining analytically the potential implications of story drift on the out-ofplane performance of such walls, they found

that building drift in the out-of-plane direction

imposed more bending stress on a wall panel

than did inertial forces on the wall panel itself. This stress caused horizontal cracking to occur

at a Y level less than 5% (5%g). They also

reported that the cracking would occur along

several mortar joints, such that the steel

framing would act independently at a fre- quency of 0.75 Hz; therefore, the inertial

forces acting on the wall would be small, and

it would remain stable and standing. The

question posed here is, How much in-plane

strength remains, given the out-of-plane

horizontal cracking? This concern will be

further addressed by testing in WBS 1.2.14.

Chua (1991), using existing analysis methodologies, conducted a parametric study of

boundary conditions, Poisson's ratio, and other

parameters to determine a best approach to estimate ultimate out-of-plane behavior of

infilled HCTWs. As noted by Beavers,

Bennett, and Flanagan, Chua found that for

static out-of-plane loads, the modulus of

rupture and the modulus of elasticity had the

most significant effect on first panel cracking,

while Poisson's ratio and shear moduli had

little effect.

6. RECENT HCTW PROGRAM RESULTS

Under the HCTW testing element WBS 1.2, one major subelement is the air bag test, identified as WBS 1.2.7. An air bag test is the

application of out-of-plane loading to an

infilled wall using a pressurized air bag (Dawe

and Seah 1989). The first air bag test (Fricke,

Huff, and Jones 1992) was conducted in

October 1991. The test was conducted on a

8-in. (203.2-mm) single-wythe HCTW 28 ft (8.53 m) wide by 12 ft (3.66 m) tall infilled within a steel frame of W14X142 columns, a

W30X108 overhead beam, and a concrete

floor slab. This test was conducted on the first floor of a five-story building. A schematic of

the test setup is shown in Figure 3, with the

actual wall and corresponding reaction frame

shown in Figure 4.

For design purposes, conventional methods

using code allowables and a static analysis

would generally be based on one-way action in

the short (vertical) direction. This procedure

would predict a uniform loading limit

equivalent to an inertial load of approximately

O.lg. However, as discussed by Fricke, Huff,

and Jones, the test results revealed an

equivalent inertial load capacity of around 3^,

thirty times their results obtained from the

conventional approach. These researchers also

looked at other alternative conventional

evaluation methods and compared the results to their test results. In all cases, the test results revealed a minimum in situ out-of-plane

seismic capacity of 13 times the least conservative conventional method. The basic reason

for the underestimated capacity resulting from

conventional methods is the lack of consideration of two-way arching action that actually occurs in infilled walls. The load deflection curve for the central point of the wall is shown

in Figure 5.

More recently, Flanagan, Bennett, and Barclay

(1992) conducted three laboratory in-plane

tests, on full-scale infilled HCTWs. These tests were part of a sixteen-test program under

subelement WBS 1.2.9, designed to study

failure mode, shear wall performance, load

transfer mechanism, energy absorption, and

hysteretic behavior of HCTW infilled frames

having various frame to infill stiffness ratios. The walls were 8 ft (2.4 m) in height and

width, constructed with single-wythe 8-in. (203.2-mm) HCTs, placed between the flanges

and against the column webs, as is much of the

construction of DOE facilities. For each test, the overhead beam size was held constant at W12X35 while the column sizes for each test varied (W10X12, W10X30, and W10X45).

386
For determining the in-plane capacity, conservative conventional analyses usually result in defining that capacity as being the load at which the first shear crack occurs in the

infilled wall. The results of the three tests when applied to a first floor interior wall of a

typical three-story DOE building showed that

the equivalent in-plane seismic capacity was

approximately 0.3^ at first panel cracking.

Applying the ultimate capacities of the tests to

this typical building indicated an equivalent

in-plane seismic capacity of approximately 0.8g.

A typical failure mode at the completion of

each test is shown in Figure 6, while Figure 7

shows a typical hysteretic curve of the in-plane

load versus deflection.

Again, the analysis of these tests showed that

conventional analysis techniques are inadequate to estimate the ultimate performance of

unreinforced HCTW structures of the type

described herein.

7. OTHER TEST RESULTS

As shown in Figure 1, unit block tests (WBS

1.2.3), bond wrench tests (WBS 1.2.11), and

prism compression tests (WBS 1.2.4) are also

being conducted in support of the overall

program. Fifteen unit block tests (ten in compression parallel to cores and five splitting

tensile strength perpendicular to cores) have

been conducted with an average net compression stress of 5539 psi (38,194 kPa) and a standard deviation of 634 psi (4,391 kPa). The

average splitting tensile strength was 351 psi

(2420 kPa) with a standard deviation of

106 psi (731 kPa). The results of prism compression tests showed an ultimate capacity of

228 psi (1573 kpa) when loaded perpendic-ular

to the cores.

Preliminary bond wrench tests on the air-bagdamaged wall have shown an average ultimate

tensile capacity of the bed joint mortar to be

16 psi (110 kPa). As the HCTW testing

program continues, additional tests and

corresponding results will be reported.

8. SUMMARY AND CONCLUSION

The HCTW program is a multiyear testing and

analytical effort to address the seismic

performance of steel and concrete frame

structures with infilled HCTWs. The first year

of an approximate three-year program has

been completed. Results to date support the

assertion that infilled HCTW structures can have significantly more inherent strength than

conventional analysis and code allowable

stresses would imply, and must be treated as structural systems.

Some progress has been made in developing

alternative analytical methods, but more test data are needed to support current results. The program is being documented, with technology transfer as one of its goals. As further

testing is conducted and analytical techniques

verified and/or developed, results will be

published. Based on the preliminary results of

the HCTW program, it is believed that many

unreinforced HCTW buildings in low and

moderate seismic zones may not require

retrofit, when evaluated using a risk-based

approach. Further testing and analytical

development are expected to confirm this assertion.

9. REFERENCES

Beavers, J.E., R.M. Bennett, & R.D. Flanagan.

1991. Research on unreinforced hollow clay

tile walls and developments of computational

mechanics. U.S. /China Joint Workshop on

Computational Mechanics, September 24-28,

Beijing, China.

387
Bennett, R.M. & R.D. Flanagan. 1992.

Structural Testing of Clay Tile Walls: Testing

and Evaluation Program, Y/EN-4325, Center

for Natural Phenomena Engineering, Martin

Marietta Energy Systems, Inc., Oak Ridge,

Tennessee, USA

Butala, M.B., W.D. Jones, & J.E. Beavers.

1991. Out-of-plane behavior of hollow clay tile walls infilled between steel frames. Proc. of the

Third Department of Energy Natural

Phenomena Hazards Mitigation Conference,

October 15-18, St. Louis, Missouri, USA.

Chua, L.S. 1991. Out-of-plane behavior of

unreinforced hollow clay tile masonry infilled wall panels. Thesis presented for the Master of

Science degree. University of Tennessee,

Knoxville, Tennessee, USA.

Dawe, J.L. & C.K. Seah. 1989. Behavior of

masonry infilled steel frames. Canadian

Journal of Civil Engineering, 16: 865-76.

Flanagan, R.D. & R.M. Bennett. 1991.

Analytical modeling of masonry infilled steel frames. Proc. of the Third Department of

Natural Phenomena Hazard Mitigation

Conference, October 15-18, St. Lx)uis, Missouri,

USA

Hanagan, R.D., R.M. Bennett & G. Barclay.

1992. Experimental Testing of Hollow Clay

Tile Infilled Frames. Proc. of the Sixth

Canadian Masonry Symposium, University of

Saskatchewan, July 15-17, Saskatoon,

Saskatchewan, Canada.

Fricke, K.E. & W.D. Jones. 1990. A test program to determine the structural properties

of unreinforced hollow clay tile masonry walls

of the DOE Oak Ridge plants. Proc. of the

Fifth North American Masonry Conference,

University of Illinois, Champaign, Illinois,

USA: 747-758.

Fricke, K.E. & W. D. Jones. 1991. A test program to determine the structural properties

of unreinforced hollow clay tile masonry walls

of the DOE Oak Ridge plants. Masonry

Society Journal 9: 56-63.

Fricke, K.E., T.E. Huff & W.D. Jones. 1992. In situ lateral load testing of an unreinforced

masonry hollow clay tile wall. Proc. of the Sixth

Canadian Masonry Symposium, July 15-17,

University of Saskatchewan, Saskatoon,

Saskatchewan, Canada.

388
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389
Figure 2. Cut-section of an infill HCTW.
Figure 4. Air bag test HCTW with loading frame.

Figure 6. In-plane laboratory

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391
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392
The Valuation and Designation Systems for Structural Calculation Programs

in Building Engineering

by

Keiichi Ohtani'

SUMMARY

In the recent several decades, the

computer uses are popularized rapidly

to all area of the building world which

are involved research, development,

design, construction and others.

Furthermore, the diffusion of the

computer uses will be spread more and more in future. In the background of

these situation, the valuation

committee for the computer programs was established in the Building Center of

Japan in 1974, for the purpose to

examine the programs and ensure these

quality and reliability. This

committee deals with the continuous

treatment programs which have general

purpose and are used in the structural

calculation reports for the approval of

requisition. The administration step of

the designation by the Minister of

Construction for the evaluated programs

was started in 1977. This valuation

and designation systems are unique

without parallel in the rest of the

world. By the establishment of this

systems, the program maker, user and

administrative official are recipient

of favours themselves. This report

describes the present situation of the

above mentioned systems.

KEY WORDS ; Valuation of Program,

Designation of Program, Structural

Calculation Program, System of the

Approval of Requisition, Building

Standard Law.

1. INTRODUCTION

The computer uses for the all

field of the building world, especially

the field of structural calculation and

sun-shadow calculation, gained

experience about 30 years. Hereafter,

the computer uses will be increased

more and more for all sorts of

business

.

In the background of these

situation, the valuation committee for

the computer program of the structural

calculation was established in the

Building Center of Japan in 1974, for

the purpose to maintain the program

quality for the structural calculation

reports by using the computer and to

attempt the convenience of the approval

of requisition by the administrative

office. Furthermore, the

administrative measure of the

designation system by the Minister of

Construction to the program, which has

finished the valuation, was started in

1977.

These valuation and designation

systems are epoch-making one and

unprecedented one for the foreign

countries, then other countries are paid attention to these systems. This

report describes the present situation

and the contents of the valuation and

designation systems for structural

calculation program.

2

.

THE SYSTEM OF THE APPROVAL OF

REQUISITION

At the case of design and

construction of Building in Japan, we must be according to the rule of "the

Building Standard Law". It is the

obligation that the designer reports

the design specification (drawings,

structural calculation report and etc.)

to the administrative office for the

approval of requisition before the

beginning to construct a building. The

administrative office is checked the

received design specification by the

rules of "the Building Standard Law",

and given the permission of

construction to the designer when he

judged that the design proper.

However, other system, which is

recognized one by one by the Minister

of Construction, is set up for the

building which is exceed higher than 60 m, and the building which is used the

special technique.

It is a characteristic of building

administration in Japan that the design

specification is recognized by the

method of approval of requisition.

3. CIRCUMSTANCES OF THE PROGRAM

VALUATION AND DESIGNATION SYSTEMS

T) Director , Disaster Prevention

Research Division, National Research

Institute for Earth Science and

Disaster Prevention, Science and

Technology Agency, Tennodai 3-1,

Tsukuba-shi, Ibaraki-ken 305 JAPAN

393
Above-mentioned, the rule of "the

Building Standard Law" is decided on the approval of requisition for the

designed building. In recent year, the case of structural calculation report

for approval made by the using of

computer are increased more and more.

The building director (administrative

official), who is participated in

requisition work, is received as the

black box as the contents of program,

and is judged for the property of

structural calculation report.

Furthermore, by the data communication

in use of computers and telephone lines

and the popularization of personal

computer, the computer use by the third

person, who has no judging ability, is

increased more and more. It is a social pressing need to establish the

appropriate plan for the computer use

of structural technique confusion and

maintain the quality of program.

In the background of these

situation, Ministry of Construction was established the investigation committee

of the program valuation works in the

Building Center of Japan in 1971. By

the report from this committee (June,

1972), the valuation committee for the

computer programs was established in

the same center, for the purpose to

examine the programs and to ensure

these quality and to directly reflect

in the administrative management of the

valuation results. The activity of the

committee was started from January,

1974 . Further, Ministry of Construction

was established the designation system

by the Minister of Construction for the

program which is finished the valuation

in 197-7. This designation system is

enhanced the effect of the valuation

system and is reflect to the treatment

for the administrative works. And, we were waiting to start this system from

beginning the valuation activity.

If the owner who has a valuated

program apply for the Minister of

Construction by the definite treatment,

the Minister will be given the

designation to the program.

Accordingly, the designated program has

tow characteristics; One is the confirm

of the applied area and calculation

method through the valuation, other is

the acceptation of logical accuracy by

the Minister. By this measure, the

position of the valuation system is

defined to the treatment on the

building administration. The

standpoint of applicant (program

owner), administrative official,

committee member and user are cleared

for each other. Together with

immeasurable effects for the the

maintain of building safety and for the

qualitative improvement of the program.

4

.

EFFECTS AND CONTENTS OF THE

VALUATION AND DESIGNATION~SYSTEMS

The considerable effects of these

systems are as follows;

1) The program maker need not exolain

the reliance of the program to the

administrative official or user individually.

2) The user use to free from care at

the first utilization of the

valuated program.

3) The administrative official need

not check to reliance of the

program, check only the objective

building properties.

The proposed reasons by every

company is mentioned these effects at

the application of the valuation. Now,

this valuation system is fixed to work

for the public good, and the

administrative office estimate the

extremely effective system to this

valuation

.

To reply the expected effects, the

following subjects are examined by the

committee

.

1) Errors affecting the results of

structural calculation in the

program itself.

2) Misapplication of the program.

(protection and caution)

3) Misusages of calculation results.

(protection and caution)

In the evaluating of the

committee, pertinence of the program is

investigated by the explanation of its

contents, the test running which is

done repeatedly and the other tests.

This committee deals with the

continuous treatment programs (packaged

programs) which have general purpose

and are used in the structural

calculation reports. It is a big

problem to deal with the copyright or know-how of the program at the

valuating work, the members of the

committee are consisted from

university, national research institute

and administrative office employs.

5. THE STATE OF THE SYSTEMS

At the early stage, the object of

valuation was only limited to the

program for wide use computer. The

programs for public and private usage were deals with this valuation

committee. And also, the committee

valuated with appropriateness the

394
-jrogram employment by the different

organ with the program maker. The

Building Standard Law Enforcement

Order, especially the part of

structural calculation, was drastic

changed in July, 1981. At this moment,

22 programs were valuated (some

programs were designated) already. But

the according of change of the Building

Standard Law Enforcement Order, above

mentioned 22 programs were lost effects

of the valuation and designation.

After the effectuation of the

Building Standard Law Enforcement

Order, the new programs (include the

above 22 programs with revision) were

the use of wide use

The usage of personal

which have a good

were progressed rapidly

more and more. By the considering this

condition, the valuation committee,

under the auspices of the Ministry of

Construction, determined to deal with

the programs for personal computer in

March, 1982, and was started to this

activity in July of same year.

valuated to

computer

computer

,

performance

,

system for change of program.

2) To make the active plan for the more efficiently valuation.

We want to hope that this

valuation and designation systems will

be developed more and more by the

exertion and cooperation of the persons

concerned

Moreover, beside this valuation

system for the building, the other

organizations were established the

similar systems, such as the aseismic

design program in high-pressured gas

facilities and the analytical program

in surveying.

At the present, the 250 matters,

in which this number is including with

the minor modification or the addendum

of function of program, were completed

to the valuation work. (Table 1 and

Figs. 1, 2) In total, the 101

different programs, include the ones

for public use, private use and

personal computer, were completed to

the valuation, and the 85 organizations

had the valuated program for the

employment of the structural

calculation work. But nowadays, the

designation system to the program for

the personal computer and the part of

program of the calculation of retained

horizontal strength were not

established. Therefore, the 68

organizations (47 programs) were received the designation from the

Minister of Construction.

6. CONCLUSION

The valuation and designation

systems for the structural calculation

programs in building engineering are fixed to the building world in the

background of the popularization of

computer uses and the administrative

mechanism of the approval of

requisition system for building. The

number of the valuated program amount

to 101 cases and the designated program

amount to 68 cases now. This valuation and designation

systems have some problems as follows;

1) To fully equip the corresponding

395
Table 1. THE STATE OF VALUATION AND DESIGNATION SYSTEMS

(February, 1991)

Valuation Designation

Public use 17

166

10

Private use 55 37

Personal Coiputer 29

Prosrai Eiployient 85 21

Modification or Addendum 64

Total 2 5 0 6 8

Fig. 1

co

I

81 82 83 W 85 86 87 88 89 90 91*

Year *tnmFeb.9i

^pOiOcuse ^private use @personal comp. mU emplcyment modflcallon or addendum

THE YEAR TREND FOR THE NUMBER OF VALUATED PROGRAMS

250

200

ISO

100

SO / ^^-^ •^y^r«oa**^

rrlv«t« tt««

IMbllc OS*

Fig. 2

ei 62 63 64 8S 8« 87 86 89 90 91*

YEAK *uotil r«b. 1991 THE YEAR TREND FOR THE ACCUMULATED NUMBER

OF VALUATED PROGRAMS

396
Highlights of the 13 March 1992 Erzincan (Turkey) Earthquake

by

Mehmet Celebi*

ABSTRACT

The March 13, 1992 Ms = 6.8 Erzincan

earthquake in Turkey is highhghted here.

The epicenter of this earthquake was lo- cated 7.7 km from the eastern end of the

North AnatoHan fault. The strong motions

recorded in Erzincan had peak ground ac- celerations of approximately 0.5 g, accompanied by a pulse of 2 seconds. The duration of the earthquake was 7 seconds. This

earthquake caused collapse of about 150

buildings—mainly to 4-5-story reinforced,

concrete-framed buildings with infill walls.

This damage, which is discussed, can be

attributed to non-compliance with seismic

codes.

KEY WORDS: earthquake; North Anatolian fault; magnitude; acceleration, code,

spectra.

1. BACKGROUND INFORMATION

At the eastern end of the North Anatolian fault in Turkey, a magnitude 6.8 earthquake occurred on March 13, 1992 at 19:19

local time (17:19 UTC). National Earthquake Information Center (NEIC) reported

the depth at 28 km and the epicentral co- ordinates as 39.706°N and 39.570°E. The '

epicenter is 7.7 km SE of the center of the

City of Erzincan, the coordinates of which,

from the Gazetteer (1984), are 39.75°N and

39.50°E. The City of Erzincan has a population of 91,000 plus approximately 30,000

military personnel. Official count of lost

lives was 800. However, unofficially, as manj

as 3,000 persons may have lost their lives

during the earthquake.

The City of Erzincan is located on the edge

of an alluvium basin lying on an alluvium

fan at an elevation of 1,150 m. It is bordered

in the north and south by high and steep

mountain ranges, some reaching altitudes

of 3,800 m. The geology of the basin can be best described as 100 m of interbedded

sand and gravel overlying lacustrine sediments. The depth to bedrock in the basin

is estimated to be between 1,200-1,500 meters. At Erzincan, the depth to bedrock is estimated to be 300-500 meters. Borehole

logs are available only for the top 200-225

meters.

The North Anatolian fault, well known for

its similarity to the San Andreas fault in

California, is a right-lateral strike- slip fault.

The recurrence intervals of large earthquakes are also similar to the San Andreas

(California) recurrence intervals. The last

large earthquake of 1939 leveled the City of

Erzincan of that time, killing 39,000 people.

The 1939 earthquake had a magnitude 8.0

and epicentral coordinates of 39. 7° N and

39.7°E (Ambraseys, 1970). Since 1939, there

have been 60 magnitude 6 and larger earthquakes along the North Anatolian fault

*U.S. Geological Survey MS/977, 345 Middlefield Road

Menlo Park, CA 94025-3591 USA

397
(Ambraseys, 1970). Ambraseys states that

historically, there are quiet periods of seismicity of approximately 150 years. A general map showing the North Anatolian fault,

the location of the City of Erzincan is provided in Figure 1. The unique topography

of the epicentral area and the basin of Erzincan is shown in Figure 2.

As a consequence of the 1939 earthquake,

the City of Erzincan was relocated to the

north of its original location to its current

location (towards the hills) with the idea

that the new buildings would be on firmer

ground and shallower depth of alluvium.

Furthermore, in the new city, the number

of stories was limited to 3—a restriction,

which, in the 1970s and 80s was slowly forgotten and abandoned. At the time of the

13 March 1992 earthquake, the tallest building in the city was 7 stories. Only 2

engineered buildings (the Railroad Station

Building and the Military Headquarters

Building) from the pre-1939 era survived

the 1939 earthquake and also the recent

earthquake.

It was reported that several buildings that

collapsed during the 13 March 1992 earthquake were repaired after sustaining severe damage during a smaller (M = 5.6) 1983

earthquake.

TYPICAL CONSTRUCTION

Typical construction in the City of Erzin- can is of reinforced concrete frame with infill walls, generally of hollow-tile masonry

construction. The infill walls vary in height,

sometimes up to window levels and sometimes as piers extending between floor to

ceilings. The roofs are generally of woodtruss with clay tiles. Gable end walls, generally not braced, are also constructed with

hollow tile masonry. Shear wall construction is seldom used. There is some adobe

(mostly single-story) and stone-masonry

construction.

The majority of the inventory of buildings

in Erzincan is 4-5 stories. A significant percentage of the inventory is single-story

houses, although there are many 2-3-story

buildings also. Very few of the buildings in

the City of Erzincan are 6-7-story buildings.

There are also some typical industrial buildings (steel frame of sugar refinery, sawtooth, roof-type reinforced concrete cotton

textile mill and a new plant of precast gable)

in the Erzincan basin.

DAMAGE

The City of Erzincan had most of the engineered structures in the earthquake stricken

Erzincan Basin and vicinity and it suffered

the most damage. Other population centers that experienced damage were smaller

towns and villages where the majority of

buildings were not engineered. In a small

scale, the basin and the damage patterns

in Erzincan due to the earthquake resemble the Mexico City (Michoacoan) earthquake of 1985. Figure 3 shows a typically

pancaked four-story builchng and Figure 4

shows a typical reinforced concrete frame

building in the state of collapse, its infill walls spalled out and plastic hinges developed at the column ends due to insufficient

and/or incorrectly implemented shear re- inforcement. On the other hand, a singlestory, partially timber-reinforced adobe

398
building located in downtown Erzincan (Fig- ure 5) survived the earthquake without any

damage.

Typical samples of damage are illustrated

in Figures 6 to 10. The general view of Unsan Flour Mill, approximately 10 km east

of Erzincan, is seen in Figure 6. The details

of the short column clearly indicate insufficient and inappropriately placed shear re- inforcement (Figure 7). Similarly, the fourwinged Military Hospital, damaged extensively beyond repair, is seen in Figure 8. The hospital lost one of its wings, lost one

story in one of the two central wings, and

sustained heavy damage to the remaining

two wings. This was one of the three hospitals in Erzincan. All three hospitals were

out of service as a result of the severe damages during the earthquake. The L-shaped

wing of the seven-story Social Security Hospital totally collapsed and the State Hospital was severely damaged.

Some of the very few steel structures in

Erzincan, at the Sugar Refinery, were heavily damaged. In Figure 9, the main refinery

building is seen. The infill walls of the re- finery sustained heavy damage. The stack

of the calcination unit of the refinery collapsed (Figure 10) damaging part of the

main refinery building in the process.

Approximately 150 buildings either totally

collapsed or were damaged beyond repair.

STRONG-MOTIONS RECORDED

The Earthquake Research Center (ERC) of

the General Directorate of Disaster Affairs

of the Ministry of Public Works and Settlement (Republic of Turkey) operates and

maintains the strong ground motion network of Turkey. The network, initiated in

1973, consists of 69 SMA-1 type accelorographs and 36 Wilmot seismoscopes. In the

earthquake stricken area, there are three

tri-axial strong-motion accelerographs. All

three stations recorded the 13 March 1992

main shock. The relative locations of these

three stations (Erzincan, Refahiye and Tercan) are seen in Figure 1. The analog films were recovered immediately after the earthquake and processed to be digitized at ERC

headquarters in Ankara. These three tri- axial records (as digitized by ERC) are seen

in Figure 11.

The record from Refahiye has peak accelerations of 0.07 g (NS), 0.07 g (EW) and 0.04

g

(vertical). The records from Tercan have

clear peculiarities (no clear P and S waves,

etc.) that are not yet clarified; therefore, no

discussion of this record will be made. Both

records possibly need careful re-digitization

and processing.

The record from within the City of Erzin- can has peak accelerations of 0.39 g (NS),

0.49 g (EW) and 0.24 g

(vertical). The

record is significant because (a) it is one of

the few near-field recorded strong-motions

for a damaging earthquake, and (b) it shows

a prominent pulse of approximately 2 sec- onds duration. Using the ERC digitized uncorrected accelerograms, baseline correction

and filtering (high-pass Butterworth filter at 0.125 Hz and order 4) was done at the

United States Geological Survey (USGS)

offices at Menlo Park, Cahfornia to obtain

the velocity and displacement time-histories seen in Figure 12. From the acceleration

time-history, the duration of the earthquake

is estimated as approximately 7 seconds calculated on the basis of the 5-95% rule.

399
applied to the integrated square of the

recorded acceleration time-history (Figure

13).

The acceleration response spectra of the

(NS, EW and vertical components) of the

earthquake for 5% damping are shown in

Figure 14. The EW spectrum clearly identifies

the predominant peaks at approximately

0.2, 0.3, 0.65 and 2 seconds. Of these, the

2-second period is the source frequency and

the others are attributable to the site fre- quencies, possibly incorporating basin re- flection effects. The record shows significant

energy at approximately 0.2-0.35 seconds

(clearly within the range of the frequencies

of the 4-5-story reinforced concrete framed

buildings with infill walls).

In Figure 15, the time-histories and the cor- responding Fourier amplitude spectra of the

dominant direction (and its orthogonal component) of the Erzincan ground accelerations are shown. The dominant direction of

128° is calculated using the recorded NS

and EW acceleration records and minimizing their cross-variance. This is ac- complished by the relationship:

0 = 0.5 tan-^ {2an/{(7l - a^)}

where and (7| are the variances of the

recorded horizontal orthogonal motions (NS

and EW directions) and a12 is their cross- variance. When the NS and EW records are rotated by 0, the orthogonal components

in the dominant direction of the recorded

acceleration time-histories are determined.

The significance of the 128° is that the strike

of the North Anatolian fault near the Erzincan region is also approximately the same

angle. This may possibly be a strong rea- son as to why the shaking in Erzincan was

apparently stronger in the direction of the

fault and possibly adversely affected the fate

of some of the buildings.

CODES

At the time of the 13 March 1992 Erzin- can earthquake, the seismic resistant design

code in effect in Turkey was the nationwide

code enacted in 1975 entitled "Specifications for Structures to be Built in Disaster Areas." The complete code is included

in the previous as well as the latest issue

of the "Earthquake Resistant Regulations:

A World List," a publication of the Inter- national Association for Earthquake Engineering (1988). The basic design philosophy

of this code is elastic. This particular code

recommended a seismic coefficient of 0.1 for

the highly seismic zone. The 1975 code was

advanced for its time as it included a spectral coefficient that incorporated the period

of vibration of the underlying soil layers.

Such a factor entered into the UBC code in

1976.

Since the damages to engineered structures

in Erzincan are very much related to the

implementation of the code provisions, significant highlights of the 1975 code are summarized below. Also, the new draft code

(not yet enacted) will be highlighted to provide a comparison with the 1975 code.

1975 Code

Method of analyses for aseismic design of

buildings is based on the equivalent lat- eral static forces applied at each floor level

(as in most other codes including the Uniform Building Code). The code stipulates

that, in the absence of rigorous dynamic

analyses, the equivalent lateral static force

400
concept be accepted as the minimum re- quirement for all reinforced concrete or

steel framed structures that are not higher

than 75 meters above the foundation level

and that have regular load bearing systems.

The provisions are also applicable for masonry buildings, chimneys, towers and elevated tanks.

The lateral load is calculated by:

F = CW

where

C = seismic coefficient = CqKSI

and

Co = seismic zone coefficient,

K = structural type coefficient,

S = spectral coefficient,

/ = building importance coefficient.

The structural type coefficient and the

building importance coefficient are similar

to those in the various versions of the UBC.

For one- and two-story buildings, 5 = 1,

K = 1 and for all masonry buildings S = 1.

Some of the significant and unique aspects

of the provisions of the 1975 code are:

a) Seismic Zone Coefficients are based

on the Seismic Zoning Map. The map

divides Turkey into four seismic zones and has been issued in 1972 also by the

Ministry of Public Works and Resettlement following a ruling by the Council

of Ministers. The map (shown in Figure

16) clearly follows the North Anatolian

Fault (Figure 1) as the major Seismic

Zone 1 region including the City of Erzincan. The largest coefficient adopted

for Zone 1 is 0.1.

b) Spectral Coefficient (with a maximum value of 1.0) is based on the formula:

S = l/[O.S + T-To]

where

T = natural period of the structure.

To = predominant period of the

underlying soil.

Compared with codes of other countries

in 1975, the Turkish Code was one of

the first to include the spectral coefficient into the code. UBC included spectral coefficient in 1976.

Reinforced Concrete structures are con- structed as "ductile systems." There is no

ductility reduction factor (as reflected by

the low seismic zone coefficient of 0.1 for

Zone 1. Detailing specifics for reinforced concrete structures recommend confinement re- gions for both the uppermost and lowermost sections of columns and for beamcolumn joints. One significant deficiency is the lack of provision or recommendation

for using deformed reinforcing bars. Again,

based on observations of damaged or collapsed buildings, except for a very few cases

(less than 10), only undeformed bars were

used.

New Draft Code

At the time of the earthquake, a draft of a new version of the code was being circulated for final review. The proposed draft

brings significant changes and approaches

to the seismic design provisions, particularly in the calculation of the seismic coefficient (C) in the equivalent lateral static

force formula F = CW:

401
C = CoIS/R . The coefficient Co, in the draft code, is proposed to be 0.4 for the seismic Zone 1. The

spectral coefficient is revised as:

5 = l/[1.0 + r-To]^/^

The coefficient R is now called the structural type factor and incorporates both the

K factor of the 1975 Code and the Reduction Factor of the UBC 1991. For example R is 3.3 for reinforced concrete-framed

structures and 5.0 for reinforced concreteframed structures with "enhanced ductility." Assuming all other coefficients as 1.0,

then, C = Co (new)/i? effectively is 0.08-

0.12. Comparison with C = Co(1975) =

0.1, indicates that for the structures with

the ductility provisions similar to that of

the 1975 Code, the coefficient C is increased

by 20%. For "enhanced ductility" C is decreased by 20%.

CONCLUSIONS

One of the important factors in the fate of

the collapsed buildings during the 13 March

1992 Erzincan earthquake was non-comphance and lack of implementation of the seismic design code rather than significant deficiency in the design code. The 1975 Code

provided sufficient and pragmatic detailing

provisions for reinforced concrete-framed

buildings; if followed, it would probably

have prevented the collapse of many of the

lost buildings. The damages suffered during this earthquake should be a reminder of

the vulnerability of many underreinforced

concrete buildings in the mid- and eastern

United States.

The repair and retrofitting methods applied

in Erzincan and vicinity should be closely

monitored to derive lessons about their ef- fectiveness during future earthquakes. During this recent earthquake, such an opportunity was lost because some of the buildings that performed well, that were hghtly

or heavily damaged or that totally collapsed

during the 1992 event had been repaired

following the 1983 earthquake. Due to lack

of documentation of the various retrofit and

repair techniques, many of the lessons that

could have been learned were simply missed.

It is imperative that dissemination and outreach of new techniques and modern design details of new construction as well as

retrofitting of existing structures must be

conveyed to the practicing engineers and

contractors.

REFERENCES

Ambraseys, N. N., 1970, Some Characteristic features of the Anatolian Fault Zone,

Techtonophysics, v. 9, pp. 143-165.

Anonymous, 1984, Gazetteer of Turkey,

v. 1, 2nd ed., Defense Mapping Agency,

Washington, D.C., (stock no: GAZGNTURKEYVl), Sept. 1984.

402
26* 28* 30* 32' 34' 36* 38^^ 40^ 4r 44"

Figure 1. General map of Turkey, the North Anatolian fault and the location of strongmotion stations (at Erzincan, Tercan and Refahiye) that recorded the earthquake.

Figure 2. General topographical map of the Erzincan basin (contour intervals 500 ft.)

[Source: Defense Mapping Agency, Aerospace Center, St. Louis, Mo., 1989]

403
Figure 3. Typically pancaked building in Erzincan.

Figure 4. Typical damaged building exhibiting collapse mechanism.

404

i
ill
Figure 7. An external column of Unsan flour mill exhibiting short column and deficient

detailing lack of shear reinforcement.

Figure 8. The Military Hospital—lost one wing, lost a story of one of the two middle wings

and sustained heavy damage to the remaining two wings.

406
Figure 9. General view of the main sugar refinery buildings.
«G ERZINCAN (TURKEY) EARTHQUAKE OF MARCH 13, 1992 - MAIN SHOCK

S 1000 2

i -500

n-1000.

NORTH-SOUTH : inax-0.39g

VERTICAL : nja3t«=0.24g EAST-WEST : niax-0.49g

STATION: ERZINCAN

16 18

200 5 100 6 0 ^

-100

-200.

2 4 6 8 10 12 14 TIME (SEC)

ERZINCAN (TURKEY) EARTHQUAKE OF MARCH 13, 1992 - MAIN SHOCK

20

TION: REFAHI NOkTH-SOjUTH : niax=0.07g

i

1

: inax=0.07g

2 4 6 8 10 12 14 16 18 TIME (SEC)

ERZINCAN (TURKEY) EARTHQUAKE OF MARCH 13, 1992 - MAIN SHOCK

20

STATION: TERCAN

0 2 4 6 8 10 12 14 16 18 20 TIME (SEC)

Figure 11. Plots of uncorrected strong-motion acceleration records (courtesy of General

Directorate for Disaster Affairs of the Ministry of Public Works and Settlement, Republic

of Turkey).

ERZINCAN (TURKEY) EARTHQUAKE OF MARCH 13, 1992 - MAIN SHOCK

1000

500

0

-500

•1000.

NORTH-SOUTH : max>.0.39g

, - , VERTICAL : max-=0.24g

EAST-WEST : max-0.49g

0

STATION; ERZINCAN

8 10 12 TIME (SEC)

14 16 18 20

200

100

0

-100

-200

ERZINCAN (TURKEY) EARTHQUAKE OF MARCH 13, 1992 - MAIN SHOCK

NORTH-SOUTH : max- 105 cm/s VERTICAL : max- 24 cm/s ^ EAST-WEST : max- 80 cm/s ^ STATION: ERZINCAN

0 8 10 12 TIME (SEC)

14 16 18 20

100 ERZINCAN (TURKEY) EARTHQUAKE OF MARCH 13, 1992 - MAIN SHOCK

50

0

-50 —

VERTICAL : max= 13.4 cm

5

EAST-WEST : max- 31.1 cm

STATION: ERZINCAN

2 4 6 8 10 12 14 16 18 20 TIME (SEC)

Figure 12. Acceleration, velocity and displacement time-histories of the Erzincan record.

408
ERZINCAN (TURKEY) EARTHQUAKE OF 13 MARCH, 1992

li < < > . , ,

,

TIME (SEC)

Figure 13. Plot of normalized summed squared (NS and EW) acceleration records.

ERZINCAN (TURKEY) EARTHQUAKE OF MARCH 13, 1992 - MAIN SHOCK

1600 I

1 1 1 '—'—'—'—T-T 1 ' ' '—'—'

r-r~i r

1 1 1 1—i—i—rPERIOD (SECONDS)

Figure 14. Acceleration response spectra (5% damping) of the Erzincan record.

409
Figure 15. Horizontal components of the Erzincan acceration time-histories rotated to the

dominant direction and corresponding ampHtude spectra.

Figure 16. Earthquake zoning map of Turkey.

410
Beach Profile Response to Storm Surge and Waves

by

Nicholas C. Kraus^ Magnus Larson^, Bruce A. Ebersole' , and Jane McKee Smith'*

ABSTRACT

The Corps of Engineers interest in storminduced beach profile change stems from its mission in engineering shore protection

projects. Beach fill is a popular method of

shore protection, and a major design consideration is the degree of protection the fill provides. Mathematical modeling provides a quantitative tool to evaluate the response of

natural beaches and beach fills to storms, and

the SBEACH model was developed to fill this engineering need within the Corps. The

paper summarizes development and application of the model, and ongoing research

aimed at improving it. The SBEACH model

simulates many of the important processes

that characterize beach profile change during

storms, and it has proven to be reasonably

accurate as a quantitative predictor.

However, applications have demonstrated the

need for improving the model, and research

efforts are under way to extend model

capabilities. The recently completed

SUPERTANK data collection project involved use of an extensive and diverse array of

instruments to study cross-shore sand trans- port processes and beach profile change at prototype scale. Analyses of the data

collected during SUPERDUCK will improve

understanding of the complex storm erosion

process and guide development of future

model improvements.

KEYWORDS: beach; beach fill; erosion;

model; storm.

1. INTRODUCTION

Beach nourishment is becoming the preferred

method of shore protection because: (a) it is

often the least expensive alternative, (b) use of sand fill provides flexibility, particularly

when "hard" engineering structures such as groins, revetments, or seawalls are already in place, (c) some States are now establishing

policies against the use of hard shore protection structures, and (d) it is a natural method

of protection, providing an aesthetically

pleasing environment and recreational benefits. However, recreational benefits can no

longer be used to justify Federal participation

in a beach nourishment project. Benefits

derived from recreational use can be considered, but projects must be justified on the

basis of storm-protection benefits, such as prevention of beach erosion and shoreline

recession, and reduction in flooding and wave

damage.

Reliable engineering methods are needed to aid in the design of protective beach berms

and dunes that can withstand the impacts of

storms. Mathematical modeling of beach profile and fill change provides a general and

quantitative approach for representing the

simultaneous influence of hydrodynamic, sediment and geometric factors important in the

storm-induced erosion process. Kraus (1989)

reviews use of beach change models in the

coastal design and engineering process.

^Senior Scientist, ^Supervisory, Research

Hydraulic Engineer, ''Research Hydraulic

Engineer, Coastal Engineering Research

Center, US Army Engineer Waterways Experiment Station, 3909 Halls Ferry Road,

Vicksburg, MS 39180-6199.

^ Assistant Professor, Department of Water

Resources Engineering, Lund Institute of

Technology, University of Lund, Box 118,

Lund, Sweden S-221 00.

411
Successful numerical simulations are limited

only by the development of the underlying

theory and availability of data to refine and

validate the model. This paper reviews sand

transport processes and beach profile change

associated with storms, development and

application of a mathematical model to simulate storm-induced beach profile change,

and ongoing research aimed at improving

both understanding of the governing processes

and numerical model accuracy.

2. BEACH CHANGE DURING STORMS

Strong onshore-directed winds that occur

during a storm result in the generation of waves that eventually break on the beach.

Storm waves usually remove sand from the

beach face and transport it both offshore and

alongshore. Frequently, a linear bar is creat- ed in response to the tendency for offshore

transport (linear in the alongshore direction).

During the peak of the storm, when breaking wave heights are greatest, most beach erosion occurs and the linear bar is most pronounced.

Figures la-lc show the evolution of a section

of beach at the Coastal Engineering Research

Center's (CERC) Field Research Facility

(FRF) in Duck, NC, during the period 9-17

September 1985, in response to the passage of

a northeaster storm (maximum significant wave height of 2.1 m) (from Howd and

Birkemeier, 1987). Figure la shows the beach

prior to passage of the storm, and Figure lb

shows the beach shape at the time of peak

storm intensity. Note the presence and uniformity of the linear bar. During the storm

sand can be transported in the alongshore

direction at potentially high rates; however,

beach profile evolution appears to be dominated by cross-shore processes. Figure Ic shows the same stretch of beach during the

latter stages of the storm, as its intensity wanes. The beach begins to assume a very

three-dimensional shape. However, the

important point is that during the time of

greatest beach erosion potential, beach

change seems to be dominated by cross-shore

processes and transport gradients.

Onshore winds and breaking waves associated

with storms also increase the mean level of

the water surface near the shore above what

would normally be expected due to tidal action. Erosion caused by high-energy breaking waves is exacerbated by abnormally high

water levels, which expose the previously

subaerial portion of the beach berm and dune

system to direct wave attack. As a result,

considerable quantities of sand can be

removed from the pre-storm berm and dune

system and deposited offshore. If water levels remain well below the elevation of the dune

crest, scarping of the dune can occur without

significant overtopping of the dune. Under

these conditions, if wave action persists long

enough, severe erosion can occur and impact

coastal, commercial, and residential structures

(see Figure 2). In areas where no dunes exist, or when rising water levels approach or exceed the elevation of the dune crest, landward overwash of dune sand can occur in addition to offshore transport of foreshore

sand. Because dune crest elevation varies

considerably along the beach, overwash

usually begins at low points in the dune

system and overwash fans develop. As the

water level rises to the predominant dune

crest elevations, large-scale overwash can

occur.

As storm wave energy subsides, wave conditions change from very steep wind waves to swell waves with low steepness (ratio of inci- dent wave height to wave length). Swell waves tend to rebuild the beach, and the

beach recovery process begins quickly following passage of a storm. Beach recovery is most noticeable in the form of an emergent

foreshore berm. The newly formed berm is comprised of sand that was removed from the

pre-storm foreshore, transported offshore

under the action of storm waves, and then

transported onshore under the action of swell

412
waves. Examples of this process are shown

below for two New Jersey beaches.

The aforementioned cross-shore transport

processes associated with storms occur over a short period of time (on the order of hours

and days), but are of major engineering concern because of the potentially catastro- phic nature of foreshore and dune erosion

and inundation. Planning and design of

shore-protection projects must consider

potential damage produced by storms. Major

factors that control cross-shore sediment

transport and storm-induced beach profile

change are: (a) beach profile shape prior to the storm, (b) grain size characteristics of the

beach sand, (c) wave and water level climate

during the storm, (d) presence of hard shoreprotection structures, and (e) longshore nonuniformity in beach conditions.

3. STORM-INDUCED EROSION

MODELING

All models of storm-induced beach profile

change developed for engineering use have

relied on the assumption that the beach profile evolves primarily in response to cross- shore processes. Also, these models have

been based on some assumption about the

shape of the profile as it evolves in response

to varying wave conditions and water level. Early models were effectively time independent in predicting beach profile change from

a pre-storm condition to an eroded condition

following the storm (Edelman, 1972;

Vallianos, 1974; Swart, 1976). Vellinga (1983,

1986) developed a simple mathematical model

for estimating severe erosion of protective

dunes impacted by storms that might be

expected along the Dutch coast. Predictions

from these models are independent of the

temporal evolution of the storm driving forces

(waves and water level); however, erosion is sensitive to the shape and duration of the

storm hydrograph, particularly phasing

between the storm-induced water level rise and the normal astronomical high tide. Wave

properties also change during the storm, as mentioned previously. More reliable esti- mates of erosion will result if the timedependent nature of the forcing functions is considered.

Realistic time-dependent simulation of beach

and dune erosion was first accomplished by

Kriebel (1982, 1986) and Kriebel and Dean

(1985), who extended the static equilibrium

beach profile concept of Bruun (1954) and

Dean (1977). Kriebel and Dean (1985)

demonstrated that their model produced

reasonable results through comparisons with

limited field data on dune erosion. Principal

limitations of the model are: (a) weak dependence of predictions on incident wave characteristics (height and period); (b) somewhat

unrealistic profile shape, in the sense of not being able to predict bar formation; and (c) limited capability to simulate accretion and

berm buildup on the foreshore. Bars are a

natural defense mechanism on beaches, and

serve as a reservoir for sand that is trans- ported offshore during the storm. It is important to reproduce this type of geomorphic

feature.

Larson and Kraus (1989) and Larson et al. (1990) developed a mathematical model

called "SBEACH," an acronym for StormInduced BEAch Profile CHange model, that overcomes limitations of previous models. A

short description of the model follows.

The net direction of sand transport at any

time during a storm is defined according to an

empirical relationship between the deepwater wave steepness and dimensionless fall speed

of the sand particles. These parameters best

predicted transport direction observed in large-scale laboratory experiments and in the

field, and frequently emerged in empirical

correlations of dynamic and morphologic

properties of the profile. Figure 3 shows the

separation of erosional and accretionary events observed in large-scale wave tank

experiments and 99 events in the field, along

413
with the empirical relationship used to differentiate between the two conditions.

The magnitude of the transport rate is calculated in four distinct zones across the beach

profile, as depicted in Figure 4, analogous to zones of different wave properties identified

in hydrodynamic studies of the surf zone (e.g.,

Svendsen et al., 1978; Basco, 1985). Empirical

predictive relationships for the transport rate

in each zone were developed using sequential

measurements of profile change made in the

laboratory during large-scale testing conducted by the U.S. Army Corps of Engineers

(Saville, 1957; Kraus and Larson 1988) and

the Central Research Institute of Electric

Power Industry (CRIEPI) in Japan (Kajima

et al., 1982). Sand transport rates in the

broken wave zone are calculated based on the

difference between the amount of local wave

energy dissipation and the equilibrium energy

dissipation level at which sand movement

would cease (the equilibrium beach profile). In other words, the transport rate is proportional to the degree to which the beach profile

is out of equilibrium. Available data support use of a linearly increasing transport rate in the swash zone, from zero at the limit of wave

runup to the value calculated at the inner

limit of the broken wave zone. In the breaker

transition zone (between the breaker and

plunge points), the rate increases exponentially from the value calculated at the outer

limit of the broken wave zone, the plunge

point. In the pre-breaking zone, the transport

rate decreases exponentially from its maximum value at the break point. Empirical

decay coefficients that depend on wave height

and grain size were determined based on

analysis of the large-scale laboratory data.

Expressions for predicting the position of the

break point, the position of the plunge point

relative to the break point, and the limit of wave runup are defined using formulations

found in the technical literature, and are given in Larson and Kraus (1989). The

numerical wave model of Dally et al. (1985)

together with the cross-shore momentum

equation are used to calculate the wave height

needed to define the break point, energy

dissipation, set up, and other wave-related

quantities at discretized points along the

profile. ;

An explicit finite-difference solution scheme

is used to solve the mass conservation equation for sand using the calculated transport

rates. The seaward limit of the model domain

is set at the location where the calculated

transport rate in the pre-breaking zone is negligible during the course of a storm. The

shoreward limit is defined to be the limit of wave runup for the case of a sandy beach, or

the transport rate is defined to be zero if a

seawall or revetment is present. Avalanching

is included to limit the steepness of the beach

slope at the landward boundary. Typical time

and space steps used in model simulations of

storms are 5-20 min and 1-5 m, respectively.

The water level and offshore wave conditions

are input at each time step, and wave height

across the shore and profile change are then

calculated.

4. MODEL APPLICATION

The empirically based SBEACH model should

be calibrated prior to its use in investigating

beach fill response to storms, because main

model parameters were established for a limited range of wave heights and periods, and

grain sizes. Calibration requires high-quality

data, including immediately pre- and poststorm beach profile data, grain size data, and

the time history of wave and water level conditions during the storm.

Here, an example is given of model calibration using a data set from the Point Pleasant

and Manasquan Beaches located on the south

and north sides, respectively, of the

Manasquan River jetties on the coast of New

Jersey. Point Pleasant Beach is a relatively

wide sandy beach with no active coastal struc- tures except a 678-m long jetty to the north;

414
Manasquan Beach is a relatively narrow sandy

beach with short groins spaced at approximately 200- to 300-m intervals. Details on the

site are given in Larson et al. (1990).

The beach profile was surveyed to wading

depth along eight transects spaced at approximate 200-300-m intervals at Point Pleasant

and along nine transects midway between

groins at Manasquan on March 27-28, 1984.

A strong extratropical storm (northeaster)

arrived on March 29, halting fathometer

surveys that would extend the profile

measurements to deeper water. The storm

caused significant erosion; and, after the

storm passed, profile surveys were repeated

along the transects. The post-storm surveys

taken on April 2-3, 1984, include 3-4 days of

post-storm swell waves with low steepness.

Water level changes were recorded hourly by

a tide gage located inside the inlet, and wave

height and period were calculated from

measurements taken by a wave buoy located

directly offshore from the inlet in 15 m of

water. Water level and wave height and

period between the times of profile surveys

are shown in Figures 5 and 6. A mean initial pre-storm profile was developed using an earlier measured subaqueous

profile joined to the measured subaerial

profile from 27-28 March. The mean subaqueous profile along the two beaches differed

substantially, with the water depth along Point

Pleasant profiles increasing monotonically

with distance offshore, whereas the

Manasquan Beach profiles had a subaqueous

terrace located near the ends of the groins. A

representative grain size of 0.5 mm was esti- mated from previously collected data. These

beach and storm data were used as model

input, and beach profile change for the storm was simulated. Calibration was determined

based on a least-squares minimization

between measured and simulated beach

profiles.

Typical results showing measured and simulated profile changes at Point Pleasant and

Manasquan are shown in Figures 7 and 8. Because of the small heights and long periods

of the waves during the latter stages of the

storm (after March 30), partial beach

recovery undoubtedly occurred prior to the

post-storm survey. This is evidenced by the

presence of the foreshore berm in each figure.

At elevations above the berm, predicted volumetric erosion is reasonable; and neglecting

sand volume in the emergent post-storm

berm, total foreshore erosion volume seems

to be predicted very well. Bar position is also

well-predicted at Point Pleasant but less so at Manasquan, perhaps because of the longshore

influence of groins at Manasquan. Predictions of beach recovery following the storm

are qualitatively correct, but quantitatively

inadequate. Additional discussion of model

results for Manasquan and Point Pleasant are discussed in Larson and Kraus (1991).

5. ONGOING RESEARCH

Results from this and many other applications

of the SBEACH model have identified areas where further research is needed to improve

the model. Better understanding of the rela- tionship between the driving hydrodynamic

forces and the entrainment and transport of

sediment is needed to reduce the empiricism

of the model. Another area of research

involves determining the optimal way to include the effects of random wave breaking.

The large-scale laboratory tests that produced

data used to develop the model only involved

generation of monochromatic waves. Monochromatic waves are a simplified representation of waves that impact most beaches.

Other areas of much-needed research include

study of the dynamics of wave impacts on

berms and dunes, dune overwash, swash

processes, and the post-storm beach recovery

process.

Recently, a major laboratory project called "SUPERTANK" supported by several work

415
units in the U.S. Army Corps of Engineers

Coastal and Dredging Research Programs was completed (Kraus et al., 1992).

SUPERTANK focussed on cross-shore hydrodynamic and sand transport processes, and was designed to address the critical research

needs outlined above. The 8-week project was conducted in the large wave tank facility

of Oregon State University in Corvallis, OR

(Figure 9). SUPERTANK drew participation

of researchers from 20 Federal agencies,

universities, and private companies in the

United States, Denmark, Japan, The

Netherlands, and Sweden.

SUPERTANK is believed to be the most

densely and comprehensively instrumented

nearshore processes data collection project ever conducted in the laboratory or the field. At the peak of data collection activities, the wave tank was instrumented with 16 resis- tance wave gages, ten capacitance wave gages,

18 two-component electromagnetic current meters, 34 optical backscatterence sensors (OBS), ten pore-pressure gages, three

acoustic sediment concentration profilers, one

acoustic-Doppler current profiler, one fourring acoustic benthic stress gage, one laser Doppler velocimeter, five video cameras, and

two underwater video cameras. Figures 10a

and 10b show components of the instrument

array (wave gages, current meters, and OBS

sensors). These instruments provide detailed

measurements of the nearshore hydrodynamics (waves and currents) and sediment

concentration needed to quantify sand

transport. Numerous fluorescent sand tracer studies were also conducted for the same

purpose. During the experiments approximately 350 beach profile surveys were made

to record the response of the beach profile to wave action. Surveys were made with an

auto-tracking infra-red survey system call a geodimeter, which targeted a prism attached

to a survey rod mounted on a moveable

carriage. The survey system allowed accurate

detailed surveys to be made in minimal time.

Sixty-six different wave conditions were run

for a total of 129 hours of wave action; 70%

of the wave conditions involved random

waves.

Test series focussed on the following: (a) beach erosion and accretion due random and

monochromatic waves, (b) erosion in front of

a seawall, (c) response of offshore subaqueous

berms, and (d) dune and berm flooding and

erosion. Figures 1 1 a and b show the condition of a dune before and during tests to investigate the dynamics of dune overtopping

and overwash. Data return from the instrument array was remarkably high (99%).

Preliminary results support many of the

design procedures being used to investigate

beach profile and fill response.

6. CONCLUSIONS

Cross-shore sand transport processes during

storms are complex, particularly foreshore

sand transport and the dynamics of dune

erosion and overwash. The SBEACH model

has been developed to quantify storm-induced

beach profile change, and it successfully

simulates many of the major morphologic

responses observed during storms, such as foreshore erosion, dune and berm scarping,

bar development, and post-storm recovery

(with less accuracy). Because of the limited

range of wave and beach conditions used to develop the model, calibration of SBEACH is recommended prior to application for project

design. However, because of the rapid re- sponse of the beach to storm events and short

time available to mobilize for data collection,

it is usually difficult to obtain accurate data

sets that are suitable for calibration purposes.

Data from Duck, NC, and Point Pleasant and

Manasquan, NJ, illustrate how quickly the

beach responds. There are several areas in which the SBEACH model can be improved.

For example, better descriptions of the effects of random waves on beach erosion and accretion and their relationship to sand transport

are needed, as are improved descriptors of

the dynamics of dune and berm inundation,

416
and subsequent erosion and overwash.

Existing technology (laboratory facilities, computers, and instrumentation) allow complex

and comprehensive investigations to be undertaken that will improve our understanding of

the physics of sand transport and beach

profile change. The SUPERTANK data

collection project will provide valuable

information for improving the SBEACH

model and our understanding of storminduced erosion.

7. ACKNOWLEDGEMENTS

We would like to express our appreciation to Mr. Jeffrey Gebert, Oceanographer, U.S.

Army Engineer District, Philadelphia, for

providing the beach erosion data set at Point

Pleasant and Manasquan, NJ, to Mr. William

A. Birkemeier for providing the figures

showing storm-induced beach change at the

CERC FRF, and to all those who made the

SUPERTANK project such a great success.

The contributions of N.C. Kraus, J.M. Smith,

and B.A. Ebersole were made as part of the

activities of work units in the Coastal and

Dredging Research Programs. The contribution of M. Larson was also supported by the

Coastal Research Program and funded by the

U.S. Army Engineer Waterways Experiment

Station's CERC through the U.S. Army

Research, Development, and Standardization

Group - UK. Permission was granted by the

Chief of Engineers to publish this information.

8. REFERENCES

L Basco, D.R. (1985). "A Qualitative

Description of Wave Breaking," J. Waterway, Port, Coastal and Ocean Engineering,

111(2), pp. 171-188.

2. Bruun, P. (1954). "Coast Erosion and the

Development of Beach Profiles," Technical

Memorandum No. 44, Beach Erosion

Board, U.S. Army Corps of Engineers.

3. Dally, W.R., Dean, R.G., and Dalrymple,

R.A. (1985). "Wave Height Variation

Across Beaches of Arbitrary Profile," J. Geophys. Res.,9Q{C6), pp. 11917-11927.

4. Dean, R.G. (1977). "Equilibrium Beach

Profiles: U.S. Atlantic and Gulf Coasts,"

Dept. Civil Engrg., Ocean Engrg. Rep. No.

12, U. of Delaware, Newark, DE.

5. Edelman, T. (1972). "Dune Erosion

During Storm Conditions," Proc. 13th

Coastal Engrg. Conf., ASCE, pp. 1305-1311.

6. Howd, P.A.,and Birkemeier, W.A. (1987).

"Sorm-Induced Morphologic Changes

During DUCK85," Proc. Coastal Sediments

'91, ASCE, pp. 834-847.

7. Kajima, R., Shimizu, T., Maruyama, K.,

and Saito, S. (1982). "Experiments on

Beach Profile Change With a Large Wave

Flume," Proc. 18th Coastal Engrg. Conf.,

ASCE, pp. 1385-1404.

8. Kraus, N.C. (1989). "Beach Change

Modeling and the Coastal Planning

Process," Proc. Coastal Tjone '89, ASCE,

pp. 553-567.

9. Kraus, N.C, and Larson, M. (1988).

"Beach Profile Change Measured in the

Tank for Large Waves, 1956-1957, and

1962," Tech. Rep. CERC-88-6, Coastal

Engrg. Res. Center, U.S. Army Engr.

Waterways Experiment Station, Vicksburg,

MS.

10. Kraus, N.C, Smith, J.M.,and Sollitt, CK.

(1992). "SUPERTANK Laboratory Data

Collection Project," Proc. 23rd Coastal

Engrg. Conf., ASCE, in prep.

11. Kriebel, D.L. (1982). "Beach and Dune

Response to Hurricanes," Unpublished

M.S. Thesis, U. of Delaware, Newark,

DE.

417
12. Kriebel, D.L. (1986). "Verification Study

of a Dune Erosion Model," Shore and

Beach, 54(3), pp. 13-21.

13. Kriebel, D.L., and Dean, R.G. (1985).

"Numerical Simulation of TimeDependent Beach and Dune Erosion,"

Coastal Engrg.,9, pp. 221-245.

14. Larson, M., and Kraus, N.C. (1989).

"SBEACH: Numerical Model for

Simulating Storm-Induced Beach Change;

Report 1, Empirical Foundation and

Model Development," Tech. Rep. CERC89-9, Coastal Engrg. Res. Center, U.S.

Army Engr. Waterways Experiment

Station, Vicksburg, MS.

15. Larson, M., and Kraus, N.C. (1991).

"Mathematical Modeling of the Fate of

Beach Fill," Coastal Engineering, 16: 83-

114.

20. Vallianos, L. (1974). "Beach Fill Planning - Brunswick County, North

Carolina," Proc. 14th Coastal Engrg.

Conf., ASCE, pp. 1350-1369.

21. Vellinga, P. (1983). "Predictive

Computational Model for Beach and

Dune Erosion During Storm Surges,"

Proc. Coastal Structures '83, ASCE, pp. 806-819.

22. Vellinga, P. (1986). "Beach and Dune

Erosion During Storm Surges," Delft

Hydraulics Comm. No. 372, Delft

Hydraulics Lab.

16. Larson, M.,and Kraus, N.C, and Byrnes,

M. R. (1990). "SBEACH: Numerical

Model for Simulating Storm Induced

Beach Change; Report 2, Numerical

Formulation and Model Tests," Tech.

Rep., Coastal Engrg. Res. Center, U.S.

Army Engr. Waterways Experiment

Station, Vicksburg, MS, in press.

17. Saville, T. (1957). "Scale Effects in Two

Dimensional Beach Studies," Trans. 7th General Meeting lAHR, 1, pp. A3- 1 -A3- 10.

18. Svendsen, I.A., Madsen, P. A., and Buhr

Hansen, J. (1978). "Wave

Characteristics in the Surf Zone," Proc.

14th Coastal Engrg. Conf. , ASCE, pp. 520-

539.

19. Swart, D.H. (1976). "Predictive

Equations Regarding Coastal Transports,"

Proc. 15th Coastal Engrg. Conf., ASCE,

pp. 1113-1132.

418
419
LWT Data

0.1000 --

0.0100 -

0.0010

0.0001

CE/CRIEPI • CE Erosion, 0,22 mm • CE Eroston, 0.40 mm O CE Accretion, 0.22 mm O CE Accretion. 0.40 mm

-•-CRiEPI Erosion. 0.47 mm

CRIEPI Erosion, 0.27 mm

-B-CHIEPI Accretion. 0.47 mm

CRIEPI Accretion. 0.27 mm

20.0 H,/wT

(a) monochromatic laboratory waves

Field Data, Mean Wave Height

0.1000

0.0100

IX

0.0010

0.0001

[7

- - <> ...j. i c

....

>QQ..

ot>

1 t P^SsH^ l>i 111

ACCRETION

J EROSION

i ^

O

...].

....

1

1 — 1—hH,/L. = 0.00070 (Ho/wT)^ HH \ 1 1—1 Mill 1 0.1 1.0 10.0 20.0 IVwT

(b) field conditions

Figure 3. Criterion for predicting beach erosion and accretion

420
Swash

Zone

Plunge

Point

Break

Point

Breaker Transition Broken Wave Zone , Zone , Prebreaking Zone

Figure 4. Regions of cross-shore sand transport

Water Level (m)

Manasquan, NJ

1984

1 1 1 1

1 1 1 1 26 27 28 29 30 31 1 2 3

March / April

Figure 5. Water level measured at Manasquan Inlet during the March 1984 storm

421
14

Wave Height and Period

Manasquan, NJ

1984

29 30 31

March / April

Figure 6. Wave height and period measured at Manasquan Inlet during the March 1984 storm

Elevation (m)

\

Point Pleasant Beach, NJ

1984

Initial 27 Mar

Post-Storm 2 Apr

Calc. Erosion 30 Mar

Calc. Recovery 2 Apr

1

1 1 1 ^

50 100 150

Distance Offshore (m)

200 250

Figure 7. Storm-induced profile change at Point Pleasant Beach, New Jersey

422
Elevation (m)

Manasquan Beach, NJ

1984

Initial 26 Mar

Post-Storm 3 Apr

- Calc. Erosion 30 Mar

Calc. Recovery 3 Apr

1 1 1 1

-2

50 100 150

Distance Offshore (m)

200 250

Figure 8. Storm-induced profile change at Manasquan Beach, New Jersey

Figure 9. Large-scale wave tank at Oregon State University

423
424

I

I

i

I
Developing a Seismic Safety Program

by

Donald G. Eagling, P.E.*

ABSTRACT

Although the field of seismic safety is in the

domain of the engineering profession, funding to carry out a seismic safety program is normally

derived from the budgets that managers (or

owners) need to carry out the missions of their

organizations. Consequently, it is generally the

manager rather than the engineer who must make

most of the key decisions which are needed to carry out an effective seismic safety program. It is also the manager who must resolve issues of

risk management which accompany seismic

hazards and involve hfe safety, damage control,

potential evacuation costs, potential liabilities and

cost-benefit considerations. It is therefore very

important that facility managers have a broad

understanding of the comprehensive nature of an

effective earthquake program as well as an awareness of the socio-political, legal and

economic risks that often impede the progress and

success of such programs.

This paper provides practical guidance for

managers who must carry out a seismic safety

program including advice for selection of an

earthquake safety consultant, the main elements

of a balanced program, and risk management

techniques for avoiding common pitfalls and

achieving expeditious results.

KEYWORDS: risk management; seismic safety.

1. INTRODUCTION

The technical aspects of earthquake safety are normally within the domain and control of the

engineering profession and academia. On the

other hand, Uie cost of earthquake safety is a drain on the very budgets that managers (or

owners) of facihties need to carry out the

missions of their organizations. This is true whether the enterprise is funded privately or by

government sources. Consequently, the

responsible manager must be heavily involved in the planning, scheduling and funding of an earthquake safety program, if it is to be

successful. Thus, it is very important that

managers receive practical professional advice so they can carry out a cost effective program and

properly manage the associated risks. In the end,

the responsibility for decision must lie with the

manager, but good communication and mutual

trust with a practical earthquake engineering

consultant is essential to provide the manager

with an extension of expertise in this specialized

field. On the other hand, the engineering

profession has a public obligation to see that

earthquake safety is achieved at minimum cost and not lost in the process of lengthy and costly

investigation and voluminous reporting.

This paper provides advice for the owner or manager who must carry out an earthquake safety

program from the perspective of an engineer who

has had such an experience as a facility manager.

The scope, depth and focus required to carry out

a satisfactory program will vary considerably

with the age of a facility, the risk involved and the

quality of design which was applied during its construction history. For a new and growing

facility the focus will be on design and

construction. For an older facility the need to evaluate existing conditions and prioritize projects

for abatement of seismic hazards will receive

most attention. For the majority of sites,

however, a balanced program will be most

effective in preventing further development of

new hazards while reducing the backlog of old

ones.

Those structural engineers who are experienced in earthquake engineering and have reviewed a number of facilities, both in government and

Retired Plant Manager, University of

California, Lawrence Berkeley Laboratory,

Berkeley, California 94720

427
private enterprise, have found a wide variety of

serious seismic deficiencies that the owner or manager were unaware existed. This is not unusual even in areas of the country where

seismic design provisions have been part of the

building code for many years.

East of California, few conventional buildings in

the United States have been designed for

earthquakes, even where there has been a history

of earthquakes of sufficient intensity to damage

buildings. At those locations where the potential

for seismic destruction exists along with a legacy

of hazardous buildings and contents, the prospect

of carrying out a comprehensive earthquake

safety program is indeed challenging.

The owner or manager who is responsible for a

facility is usually unfamiliar with earthquake

engineering and may tend to look for answers in

techniques that are more sophisticated than are required to solve the actual problems in

earthquake safety. The approach to solutions of

these problems can become so academic,

ponderous, and expensive that abatement of the

seismic hazards simply doesn't get done in a timely manner.

In recent years, the state-of-the-art in seismology,

geo-technical theory, and the dynamic analysis of

structures has progressed tremendously. Spurred on by the need to resolve questions in seismic

safety for nuclear power plants, the field has

become very compartmentalized and specialized.

The great strides made in these specialties have

contributed significantly to the field of earthquake

engineering and public safety. Unfortunately, it is easy for the responsible manager or plant

engineer to "fall into a crack" between these

experts who quite naturally tend to resolve

seismic questions based on their own specialties.

Often more time and money are expended

analyzing the problems in earthquake safety than

would be needed to design practical solutions to these same problems. To gain a better

perspective it is important to understand that most

problems found in existing construction are a

result of not implementing what has been known

and observed about earthquake-resistant

construction for a long time. Structural engineers

who have observed and studied earthquakedamaged buildings are able to diagnose hazardous

deficiencies in existing buildings rather easily and

efficiently, often without complex calculafions.

They know which building types are hazardous.

They also know standard methods for

rehabilitating them.

2 . SELECTING AN EARTHQUAKE SAFETY

CONSULTANT

The most important thing a manager can do to

initiate an effecfive and economical earthquake

safety program is to hire an experienced

earthquake engineer as an earthquake safety

consultant who is strong on design and tends to keep analysis straight-forward and simple.

Occasionally, there is good reason to apply

structural dynamics to provide a better

understanding of a complex problem, but not

often. The earthquake engineer, working closely

with the manager, should advise on selecfion of

other consultants, define their scope of work,

coordinate their work with the overall program

and ensure that the owner is not vicumized by

unnecessary or impractical studies.

Selecfing the right structural engineering firm to counsel a practical way through the complexifies

and pitfalls which can befall earthquake safety

programs is the most important decision a manager can make.

This is also true when selecting geo-technical

consultants. The level of sophisficafion in state- of-the-art techniques for predicfing the intensity

of ground shaking is intimidating. There is a strong tendency for both consultants and clients

to believe the predictions to be more accurate than

history shows they are. This tendency may lead

participants to spend more money and time than

the exercise is worth. The illusion of security

thus developed is apt to be in direct proportion to the degree of sophistication applied.

Even when structural dynamics is to be

employed, the selection of an effective groundmotion input can be a relatively simple matter. There is littie to be gained by exhaustive site studies because history shows tiiat the prediction

of ground motion is indeed an inaccurate science.

The inaccuracies of input often can be

accommodated in good structural design.

Potential earthquake engineering and geotechnical consultants should be asked to explain

428
in simple terms how they expect to approach the project, what techniques will be applied and what

they expect to find. They should provide

examples of previous work, names of clients, and

cost history. The owner should be sure that the

lead consultant selected has strong design

experience and has made field investigations of

earthquake damage.

3. THE BALANCED PROGRAM

An effective earthquake safety program is analogous to an effective lateral-force-resisting

system; it should have no weak links. Several

basic precautions should be taken in establishing a program; of primary importance are the

following five:

First, make certain that plaimed new buildings are not being inadequately designed while the process

of earthquake safety review of existing buildings

is underway. This is a profound admonition, but

the possibility of its happening is real. It can be

avoided by using plan-check or third-party

review, prior to construction, to ensure that the

calculations and designs of new structures and

rehabilitation projects are done properly to resist earthquakes. It is embarrassing for a manager to find that a newly designed and constructed

building is worse than an old one, but it happens.

Design criteria should be clearly defined and

simple to use. Standards and model codes are readily available and should be used. Complex

approaches or criteria should not be applied

unless the need is clearly established. Criteria

should be practical. At most sites, many building

modifications, experimental setups, equipment

installations, cabinetry, nonstructural components

and other projects are designed regularly by

architects, mechanical engineers, designers, and

other nonstructural engineers who do not have

sufficient background in seismic analysis and

design. Nevertheless, if the criteria are simple

and straight-forward these minor subprojects will

be designed and built with adequate earthquake

resistance. The more significant structural

problems should be handled by registered

structural engineers. Also, research or production facilities using hazardous materials

should be carefully reviewed for seismic safety

by a structural engineer. A professional

engineer's stamp and signature as well as a third- party review should be required.

Second, the site should be reviewed for geologic

and seismic hazards. Potential conditions that are inherently hazardous in ground shaking should be

identified. These may include the following:

1 . Unstable slopes and existing landslides;

2. Areas of low-density saturated granular soils subject to densification and subsidence;

3 . Areas of low-density saturated granular soils subject to liquefaction;

4 . Areas where sensitive clays may be subject to strength loss under heavy ground shaking;

5. Areas where flooding would occur due to

failure of an up-slope levee or dam.

Active faults should be identified and a geologic

map of the site developed.

The site review need not be rigorous in detail

unless the potential hazards pose a high risk for an existing building or lifeline. If a new building or improvement is planned, the specific siting can be examined in more detail. The main point is to recognize potential hazards and take them into account. For example, it would be folly to permit

the typical one-third increase in allowable soil bearing capacity for seismic loading in an area of

sensitive clays subject to strength loss under

ground shaking. The initial review should be

quite broad and superficial in character. It is important that it be carried out by a geologist or

soils engineer who understands the nature of soil dynamics, preferably one who has experience

with earthquakes. Generally, except for fault rupture, each of the potential hazards that may

exist can be mitigated through standard

stabilization practices, or by simply avoiding

them in the case of new construction. Sometimes

fault movement can be accommodated, or the

effect of fault movement mitigated, if it is known

where surface mptures are likely to occur.

Third, survey and evaluate all existing buildings

and stmctures to determine their earthquake safety

ratings. A structural engineer experienced in

earthquake damage investigation should do the job. The assessment should be kept simple. A

basic requirement is to find out what has to be

done to insure that each building has a predicable

lateral-force-resisting system. The job of

rehabilitation should be started one step at a time,

reducing liability on a priority basis.

429
Given a limited budget it is important to detemiine

which buildings will have the greatest benefit for

the money spent for improvements in life-safety

and property damage. Of course, practical risk

management must also address those sociopolitical issues which encompass and plague

earthquake safety programs and pose questions of

public and personal liability. Practical risk

management is discussed in the next section of

this paper.

Fourth, make an earthquake-hazards survey of

the contents each building including operations,

equipment, hazardous materials storage and

nonstructural building elements. Obvious

hazards, such as loose-item overhead storage,

should be corrected by building managers or

operations supervisors. Most operational hazards

are obvious to one simply observing the scene and imagining an earthquake taking place.

Tipping hazards, such as storage cabinets, tall files, library shelving, and similar furnishings,

should be braced or anchored. Tie-downs shoiJd

be installed on plant equipment such as transformers, emergency generators, tanks,

elevator drives, fans, motors, and similar units. Apply a simple and judgmental priority system to use limited resources economically.

Fifth, develop an emergency plan to recover from

a destructive earthquake. Apply the scenario

technique to develop a probable model for the

aftermath of an intense earthquake. Use those

department heads who will have to handle the

recovery to lead the planning effort. Reduce

obstacles to recovery by eliminating obvious

hazards and ensuring that the supplies and

equipment that will be needed in an emergency

will in fact be available. Lifelines, such as water

supply lines, power systems, storm and sanitary

sewers, transportation and communications

systems also should be surveyed with

earthquakes in mind. The consequences of

possible facility losses can be mitigated by careful

emergency planning, and the potential for loss of

a given facility reduced by "hardening" the

lifelines that would likely be in jeopardy during an earthquake.

Self-help planning, preparation, and training

should be key elements in any emergency

response plan for earthquake safety. Make sure that building managers and operations

supervisors understand this fact and let them take

part in the preparation oflocal emergency plans.

In the process of administering an earthquake

safety program, a manager will encounter

differences of opinion among the engineers and

consultants involved. These differences reflect the perspectives of experts and tend to give the

managers some insight into the practical state-ofthe-art. It also reminds us that there often is more

than one answer to a given problem. When a problem is particularly difficult and costs, risks, or liabilities are high, it is unquestionably

worthwhile to get more than one opinion.

The question arises, how does a manager resolve

technical differences of opinion between two

consultants on subject-matter about which the

manager feels inadequate.

The best answer lies in the manager's usual role, that of managing the multi-disciplinary functions

of a technical complex such as a major research

and development laboratory or a sophisticated

production facility. The development of good

communication and mutual trust with a practical

earthquake engineering consultant will provide

the manager with an extension of expertise in this specialized field much as it does in any other

specialized field under his or her management. In the end the responsibility must lie with the

manager and it is important to realize ahead of

time that technical differences of opinion are apt

to arise about earthquake safety management. A

good earthquake engineering consultant will be

interested in the goal of practical earthquake

safety rather than earthquake engineering for its own sake. This extension of expertise through

selection of wise counsel is a challenge that most

managers face in other facets of their

responsibilities.

When questions regarding technical differences of

opinion or criteria persist, it is important that they

are resolved by some due process within a

technical framework that will stand the test of

future technical and legal review. The designer

and the manager should be reasonably protected

by the due process involved assuming each

fulfills his professional responsibilities

satisfactorily.

4. RISK MANAGEMENT

Risk management from the manager's point of

view is an important part of a comprehensive

430
earthquake safety program. It is, of course,

inherent in the establishment of lateral force

criteria for the design of buildings and equipment

including research facilities and experimental set- ups. Earthquake codes establish minimum

requirements for life safety and essentially

provide protection against collapse. Damage

control is not a prime objective. Many code

provisions, such as the prescribed limitations on

"drift" (deflection based upon prescribed lateral forces), have the effect of reducing damage, but one must realize that the real earthquake will cause deflections much greater than "code

deflections." When damage control is an

important consideration, the design must accoimt

for these larger deflections. This protection is not

inherent in ttie Code. One approach is to analyze

a structure with the objective of predicting or

estimating the location and extent of damage that will probably result from a major earthquake. In

this way, additional attention to design detail can

be applied specifically in the area of concern to "buy insurance" against damage for little extra

cost, an example of good risk management.

Liability is certainly a legitimate concern in risk

management. Often, however, this concern is translated into a legalistic solution, rather than a

practical solution which actually mitigates the

seismic hazard. For example, the Code is generally not retroactive so it is not legally

incumbent upon the "responsible official" to upgrade an existing building to current standards.

Nevertheless, some earlier editions of the Code

made it "legally" possible to design and construct

a hazardous building. A non-ductile reinforced

concrete frame building is a good example. A

decision not to review an existing building

because it was once designed to Code is a

legalistic solution to avoid Hability, but it does

nothing to mitigate any seismic hazard that may

exist. This legalistic position under the protective

umbrella of the Code is becoming more difficult to assume because the engineering profession is now much more aware of hazardous buildings

which have been "built to Code." One cannot be

certain to avoid liability by remaining ignorant of

the hazard. The legal issue may well be whether

or not such a building is commonly known to be

hazardous by the profession.

On the other hand, the risk of liability should be

managed carefully when structural hazards are revealed as a result of seismic safety surveys and

reports. For example, if a building is reported to be a collapse hazard, the official responsible for

the safety of the occupants should take steps to mitigate the hazard. It is important to actively

seek funds to abate the hazard and to inform the

occupants that the building is deficient. While

these steps will not guarantee immunity, failure to take them certainly increases liability.

The problems of funding rehabilitation work are unusually difficult and the solutions time

consuming. The period during which the hazard

may continue to exist increases the risk of

liability.

A public agency cannot legally go out of

business; therefore, it cannot spend its available

funds so heavily for rehabilitafion that it cannot

fulfill its prescribed mission. This fact weighs

against liability, but does not provide immunity.

In the event that a damaging earthquake results in

litigation, the pertinent issue is: what funds were

available to the responsible official and what were

they used for? With this in mind, one approach

to managing the risk of liabiUty is to take the basic

risk-reducing steps which can be identified

immediately (assuming the hazard cannot be

easily abated in a short item), then follow due

process to find a permanent solution to the

problem. Examples of emergency steps are: adding temporary supports, reinforcing structural joints, installing epoxy grouting, removing

potential hazards, or even reducing the occupancy

loading. Normally, some emergency funding can

be found for such purposes while the more timeconsuming tasks of evolving a permanent

solution and developing adequate funding take

place. The important point is that responsible

action (within constraints) must take place if hability is to be minimized.

It is likely that a seismic safety survey will turn up a number of structurally deficient buildings

and facilities. This has been the common result where such reviews have been carried out. For

years, many buildings were designed with nonductile, reinforced-concrete frames that were then

permitted by Code, but are now known to have

poor seismic resistance. Many older buildings

have no formal or predictable lateral force

resisting system. Sometimes building alterations

have reduced or destroyed the resistance

incorporated in the original design. The point is, a seismic safety survey will likely present the responsible official with a multiplicity of hazards

and risks to manage.

431
It is important to mitigate risks on the basis of

priority, but it is even more important not to get

bogged down in a complex series of studies or a methodology which slows the process of

abatement. A simplistic priority system based

upon some due process and responsible

professional judgment is sufficient. As with the

Richter scale for measuring the magnitude of

earthquakes, it is not as important that the result is accurate as it is that relative size (or priority) is easily and quickly established.

The same selection principle should be applied

when seismic safety surveys are initially carried

out. That is, the priority system for the sequence

in which buildings and facilities are surveyed

should be simplistic and direct. Obvious

problems, possible collapse hazards and high risk

facilities such as those with toxic dispersible

contents, should be reviewed early. Emergency

recovery facilities, high occupancy buildings and

lifeline facilities should also be early on the list.

It is important that the survey should not be held

back by an academic approach to the multiplicity

of potential hazards and the complexity of the

problem. The recommendation is to keep the approach simplistic, rely on good professional

judgment and move forward.

One stumbling block to seismic safety is the ever

present concern for accurately estimating the intensity of the potential earthquake that a given

site might experience. For various reasons, poliUcal, academic and psychological, the

immediate, basic need to find out whether a brace

is indeed missing unfortunately too often

becomes secondary to guessing how big the

earthquake is going to be. Commonly, the

problems uncovered in a seismic survey have less to do with lateral force criteria to be applied than

with other design deficiencies such as missing

links, brittle members or connections, lack of

continuity or just poor construction. When a building or facility is found deficient, the size of

potential seismic input is only one of the considerations which may be brought to bear on

the corrective measure. Usually there will be

ample time to develop detailed lateral force criteria after the real problem is revealed.

The recommendation is: don't delay the seismic

survey in order to study the potential seismicity of

the site. Experience shows that this approach is not good risk management.

The design criteria for new buildings are rather well established in the latest Codes, but this is not

true for rehabilitation work. Here good risk

management requires more careful consideration.

The lateral force provisions in the Code provide

good guidelines for rehabilitation design, but

often, lateral force resistance is only part of the

problem. As discussed, brittleness, lack of

continuity or redundancy, deflections, poor

detaihng, poor workmanship, and many other

possible deficiencies may exist. From the

standpoint of risk management, it is even more

important than in the design of new facilities that the designer of rehabilitation work give particular

attention to a complete seismic diagnosis and

criteria development.

Once it is determined that a building has a serious

structural deficiency that must be corrected,

another kind of problem is often present. The

building may have other Code deficiencies by

current standards that are not central to the main

hazard. That is, the main structural deficiency

may be a collapse hazard, but the other Code

deficiencies may not present life safety hazards.

The question may tiien arise: will the responsible

official (or the engineer who designs tiie rehabilitafion work) be put in a position of

liability if the design does not correct all the

deficiencies by current Code standards? Often it is not economically feasible or even good risk

management to correct everything. For example,

funding may be used to correct two collapse

hazards rather than spend it all on one building to bring everything up to Code.

The recommendation is to achieve the most life safety for the funding available, but mitigate

possible liabilities for the design professional by

careful due process. For example: a "criteria

board" can be established consisting of

professionally knowledgeable members such as

the rehabilitation designer, the plan checker

(preferably an independent consultant), and the

"building official" (the in-house person

responsiblefor design and construction).

The authority to set seismic criteria should then be

officially delegated to this "expert" group. Court

judgments in the United States have held that the

responsible authority will be immune from

liability for acting in a discretionary manner if this person has the authority and knowledge to do so. The authority must be properly delegated and

432
"knowledge" implies professional judgment in the

practice involved. The rehabilitation design

engineer will be reasonably protected by the due

process involved assuming he fulfills his design

responsibilities satisfactorily.

High on the list of effective risk management

techniques related to seismic safety is the so- called "third party plan check" that has been

mentioned earlier in tiiis paper. This independent

plan check, together with proper field inspection

of construction (whether new or rehabilitation

work), is highly recommended as one of the best

ways of assuring seismic safety in structural

design and construction.

In "earthquake country," engineers generally

believe the buildings they design will be subjected

to a damaging earthquake during the lifetime of

the structure. They don't consider the hazard to be one that in aU likelihood will not occur. They

are also aware that the earthquake is the real master inspector. If there is a gap in the lateral- force-resisting system, the consequences will be more serious than statistical. This point of view

has a very positive effect upon risk management

as applied to design.

In areas where earthquakes are very rare events, it is difficult for engineers to take them seriously,

particularly in the design of conventional

structures. It is a fact that few buildings in the

midwestem and eastern United States have been

designed for earthquake safety, even in those

areas where it is well known that damaging

earthquakes have occurred in the historic past.

Many buildings are constructed of unreinforced

unit masonry, one of the building types

particularly susceptible to earthquake damage and

collapse. A great deal of progress could be made

by simply avoiding the use of unreinforced unit

masonry in new construction.

In "earthquake country," the choice of criteria for

seismic design can be a relatively simple matter, if one believes the great earthquake is imminent. In

this case, the design earthquake can be taken

relatively close to the maximum credible

earthquake. This might correspond to 0.2g base

shear using the static equivalent design approach

or to 0.8g using ground spectra acceleration for an inelastic dynamic analysis. This design

approach is based on what is known as the

"minimax decision," since it minimizes the

maximum losses in the future.

In areas of the country where the potential for a

great earthquake exists, but the probability is extremely low, the choice of criteria can be very

difficult. The maximum credible earthquake is not a practical choice for the design of most

conventional structures in such areas. A more

fundamental consideration relates to the decision

whether or not to design for seismic forces at all. However, considerable seismic resistance can be

achieved for very little extra cost by simply

applying the principles ofequivalent static lateral force design and making sure that the system is continuous and ductile. The insurance available

in such a minimal approach is a true bargain in

risk management. The lateral-force factor to be

utiHzed is of secondary importance. However, if Building Code values are used, one has some

assurance that this choice is properly coordinated

with other design provisions of the Code.

5. CONCLUDING COMMENTS

This paper is based upon experience with the

earthquake safety program at the Lawrence

Berkeley Laboratory (LBL) where 34 buildings were strengthened over a 20 year period

following the San Fernando Earthquake in 1971.

As a consequence of that experience, LBL

pubhshed a "Seismic Safety Guide" in 1983 for

the U.S. Department of Energy (DOE). It was

written for those managers of DOE facilities (and

engineers) who had little background in

earthquake engineering. Its guidelines include

hazard identification and evaluation, site planning, the evaluation and rehabilitation of

existing facilities, the design of new facilities,

lifelines, operational safety, emergency planning

and the management of risks and liabilities.

Since 1983 great progress has taken place in

earthquake engineering. Also, a number of

damaging earthquakes have taken place where

structures were designed to resist them. As a

result a wealth of new information has become

available. A new edition of the "Seismic Safety

Guide" will be published in 1993 to incorporate

and interpret this new material so that the

managers of DOE facilities can achieve a comfortable level of understanding for the

decisions they must make relative to seismic risk

management.

433
The format for the "Guide" will follow the

original edition. Each chapter will be written by

an experienced professional for an audience

composed of managers or engineers with little background in earthquake engineering. Comment

and advice from the facility manager's perspective

will be provided in the foreword to each author's

section. For those interested in a much more

detailed treatment of this paper's topic, the new

"Seismic Safety Guide" will be published in the

Spring of 1993.

6. REFERENCES

1. Eagling, D.G., et al, (1983). "Seismic

Safety Guide." LBL-9143, Lawrence

Berkeley Laboratory, University of

California, Berkeley, CA.

2. Eagling, D.G., (1992). "Seismic Safety

Guide, Synoptic Description of the 1993

Edition." Lawrence Berkeley Laboratory,

University of California, Berkeley, CA.

434
Desk-Top Model Buildings for Dynamic Earthquake Response Demonstrations

by

, A. Gerald Brady*

ABSTRACT

Models of bmldings that illustrate dynamic

resonance behavior when excited by hand

are designed and built. Two types of buildings are considered, one with columns

stronger than floors, the other with columns

weaker than floors. Combinations and variations of these two types are possible. Floor masses and column stiffnesses are chosen

in order that the frequency of the second

mode is approximately five cycles per sec- ond, so that first and second modes can be

excited manually. The models are expected

to be resonated by hand by schoolchildren or persons imfamihar with the dynamic res- onant response of tall buildings, to gain an

imderstanding of structural behavior during earthqucLkes. Among other things, this

experience will develop a level of confidence

in the builder and experimenter should they

be in a high-rise building during an earthquake, sensing both these resonances and

other violent shaking.

KEY WORDS: building models; construction; dynamic; earthquake; resonance; vibration.

1. INTRODUCTION

The physical characteristics and construction of desk-top model structures that demonstrate earthqu2ike shaking effects have

probably been described in many research

thesis reports. The models here, though,

are primarily hands-on displays where the

participant provides the shaking input by

hand in the form of a shding to-and-fro motion, of an easily held block of wood act- ing as the fotmdation for the structture. The

models axe suitable for schoolteachers in re- gions of high seismicity to acquaint children

with the kind of motion they might experi- ence if they are in a tall building during an

earthquake, rather than a low-rise school

building, and to give children a perspective of how the high-rise buildings they are

familicir vnth in their cities shake during

earthquakes in ways that are understood,

and have therefore been swrcovmted for, by

the structural engineering profession. The

models axe also suitable for showing this

behavior in a simple fashion to people outside the seismic engineering profession.

Almost every university or institution with

a strong ezirthquake engineering component

will have had, at some time, a motor-driven

exhibit shaking a platform in the northsouth and east-west directions simultaneously while a structure bolted to the plat-

*U.S. Geological Survey MS/977, 345 Middlefield Road

Menlo Park, CA 94025-3591 USA

435
form illustrates its seismic response. A more

current form for this display is a computer

program where earthquake magnitude and

epicentral distance can be entered, together

with structure height and construction material. Detailed structural behavior can be

shown on a color screen, and catastrophic

collapse provides audience excitement. The

models described in this paper, however,

axe not comphcated, and their hands-on feature provide a different experience. It can

actually be diflScvdt to find the perfect fre- quency and thus the perfect timing to excite, slowly and steadily, a flexible or resilient structure into one of its vibrational resonances. Once learned, the abihty to excite these model resonances remains, as does

the ability to resonate other structures Uke

fishing rods, handrails, and small flagpoles,

for example.

The models show how a bmlding resonates

in a particular direction when the ground

shakes in a steady vibrational to-and-fro

motion in the same direction. The frequencies at which this happens have been calculated first, and the models are designed

and constructed so that the frequencies are within the range that the operator can readily provide, namely from 72 to 5 cycles per

second (cps). This range includes the reso- nant frequencies of some typical tall buildings (Housner and Brady, 1963). The models show how quickly a resonating mode

can be excited and how little input amplitudes are needed to maintain excessive,

highly visible, structural amplitudes.

There are two basic models described here,

representing the following buildings:

A. Strong column-weak beam (bending

beam) buildings: Buildings whose columns

are feir stronger and stifFer than the floors

and floor beams so that the cantilever bending characteristics of the colmnns are tmaffected by the presence of the floors.

B. Weak colimMi-strong beam (shear beam)

buildings: Buildings whose floors and floor

beams, being {ai stiffer than the colvunns,

remain flat and horizontal during vibrations.

The colimms are forced into slight 5-bends

between floors.

There are many versions and combinations

of these two that permit modeUng:

1. Buildings that are different in height.

2. Buildings which throughout their height,

behave as though both types A and B above

have an effect.

3. Buildings whose lower stories have such

massive colurons that type A is in effect

there, but whose upper story columns have

less strength and type B is in effect.

4. Buildings whose stiffness is increased by

the addition of shear walls, sections of exterior wall reaching the entire height, and

whose wall direction is the same as the direction of shaking.

5. Buildings that are not symmetric in plan

with respect to the shaking direction, and

which therefore twist slightly, along a vertical axis, when shaken.

Placing two structural models on the same

base can provide an illustration of the

pounding of adjacent structures not vibrating in imison.

Although a fundamental structural reso- nance might have a low resonant frequency, a higher resonance (or overtone or harmonic

436
in musical terms), can be excited at close

to six times (for type A) or three times

(for type B) the fundamental frequency,

depending on the structural type, and

the models axe planned to be able to show

this.

Construction of the models is based on

wood blocks of suitable size to represent

horizontal floors, and strips of thin sheet

alimainum to represent vertical columns

whose flexibihty or stiffness can be adjusted

by specifying different widths and thick- nesses. They have various low amoimts of

damping, depending on the joint details,

but damping is assumed insignificant in the

analysis. There is no effort to incorporate

specific energy dissipation capability in the

models, although some exists in all moving joints. The resvdt is that resonant amplitudes of motion are large, particularly

if, by mistake, input amplitudes are maintained at high levels once resonances are achieved. This can lead to damage in the

form of weakened screwed joints, permanently bent columns or column fatigue fail- ure imless reasonable care is taken.

2. SPECIFICATIONS

Because these models axe hand-held and

excited into resonance by shaking of the

base with a steady vibrational to-and-fro

motion, the highest expected resonsmt fre- quency is Umited by the highest frequency

able to be provided by the user, about 5 cps

for most people. The base must be small

enough to be grasped between thumb and

fingers because the highest frequency res- onances require the user to overcome the

inertia of the model which builds up in proportion to the square of the frequency, and

to overcome the tendency to overturn.

For ease of transportation, the height of the

tallest models is hmited to 2 ft.

Figure 1 illustrates the two basic models

and construction details for the joints. Fig- ure 1(A) represents two vertical cantilevers

rigidly attached to the base and able to deform by bending under the action of horizontal inertia forces when the base is shaken.

Horizontal deflections are representative of

the columns of tall buildings whose columns

at all elevations are much stronger or stiffer than the floors and floor beams at the same

elevations. The floors, where they meet

these colxunns, bend readily, and their con- tribution to the bending resistance of the

joint is insignificant. The column bends,

throughout its height, as though the floors

have no effect. In the type A model, the

floor bending effectively takes place in the

joint itself, whereas in real structures it is distributed over the total extent of the floor.

Fig\ire 1(B) illustrates the model of a building with stiff, unbending floors. During vi- brations, the floors remain horizontal and

the joints do not rotate, forcing the columns

in the vicinity of the joints to be vertical. In

between floors, the columns bend into very

slight 5-bends.

The sketches adjacent to the model of Fig- ure 1(A) show two deformed shapes of the

basic 2ft-high vertical cantilever model as it vibrates in its first and second modes. The

base is shown having zero motion. During

an earthquake, or, in this case, during hand

excitation, the base is moved at the same

frequency that the model exhibits. With

models having insignificant damping, the

base motion required at resonance has sur- prisingly low amplitude. The most effective

technique, once the resonant frequency is

437
established, is to vibrate the base shghtly

ahead of the structure. The second mode

of model A has a theoretical frequency 6.27

times faster than the fundamental mode. A

model with half the height, also sketched,

has a fundamental frequency four times the

fundamental mode. Calculations for frequencies, given specific member sizes, may be

found in the Appendix.

Adjacent to the model of Figure 1(B) Eire the deformed shapes of the rigid floor/

flexible column model as it vibrates in its fundamental and second modes. The sec- ond mode has a frequency approximately

three times the fundamental. A model of

half the height has a fundamenteJ frequency

two times that of the fundamental mode.

Calculations for frequencies may be found

in the Appendix.

3. CONSTRUCTION

The construction of these models has been

made simple by the selection of hardware

that is readily obtainable. The colimms are cut from aluminum sheet in thicknesses of

0.031" (or %") and 0.010", obtainable from

a metal dealer and cut on a shear at the

dealer or local sheet-metal shop. The floors are from 1" x 2" pine moulding, smoothed

four sides (finished section %"x 1^2")- The

bases for those models with no aluminum

attached to the longer dimension may be

wider, for stability against overturning, for

example 1" x 3". Models with vertical shear

beams for stiffening require bases the same

size as the floors above. Short lengths {%")

of plastic or rubber hose, inside diameter

Ys", are required for the hinges of model

type A. They hold the floors in position

between the columns, threaded on the #8

roundhead wood screws. Washers imder the

heads of the #8 sheetmetal screws in model

type B are built up to an outside diameter of ^2" to ensure proper bending of the

columns just above and below these connections.

The base of every model is 4" long, presenting a convenient size to grasp. Floors

having hinges are ^4" shorter for each end

that is hinged, so that the type A model

has 3^2" floors. The floors of the Type B

model are 4" long. Coliman, floor and hinge

detail for these and other models Usted in

the Introduction can be seen in Figure 1.

4. CONCLUSIONS

These models axe simple to make and are challenging to excite into resonance. They

fulfill the desired goal of providing a handson experience for the public in tall building dynamics. Seeing the response of the

models to "ground" motion may help allay

the fears of persons experiencing an earthquake.

5. APPENDIX

5.1 Buildings with strong columns and

weaker floors

Figure 2(a) shows the x — y coordinate sys- tem for a cantilever beam, placed vertically,

with constant physical characteristics: moment of inertia /, mass density p, cross- sectional Eirea A, and Young's modulus E.

Neglecting shear distortions and rotary inertia (see, for example, Hurty and Rubinstein, 1964), the deflection y{x,t) satisfies

^

+ ^.^ = 0 (1)

Figure 2(b) shows two such cantilevers

representing columns, tied together through

floor beams with connections allowing free

438
rotation of the beams. Representing the

floor mass as m/, total floor mass A//, and

assimiing the floor mass is uniformly spread

over the height /, equation (1) becomes

where

+

2pA + Mf/l

2EI = 0 (2)

Looking for a natxireJ mode solution

y(x,f) = y{x)smujt (3)

where w is a modal circular frequency, leads

to

where

13' =

2pA-\-Mf/l

2EI

Looking for a solution of the form

(4)

(5)

(6)

where A is complex, and using the boundary conditions

y(0) = y'(0) = y"{l) = y"'{l) = 0 (7)

yields

cos^/cosh^/= -1 (8)

Solving numerically, for example by trial

znd error with a calculator,

/5/ = 1 • 875 , 4 • 694 ,

. . . (9)

Substituting in equation (5),

2EI

2pA + Mf/l^

.A^„,n = l,2,3...

(10)

^^2

iVi = 3-516, iV2 = 22-035, -^- = 6-27 (11) Ni

Selecting / = 24", h = 1.5", d = (1/32)",

rrif = 0.1 Ib./g, h = 4"; and using E =

10 X 10^ psi, p = 0.0975 lb. /g/in\ g =

386.4 in/s^ / = bd^/12, A = bd, Mf/l =

rrif/h, yields

/i = 0.9 cps , /2 = 5.7 cps (12)

and the frequency of a building of half this

height is

fl/2 = 3 • 6 cps (13)

5.2 Buildings with strong floors and weaker

columns

Figure 3(a) and (b) show two structures of

height / under the action of a horizontal

shearing force F, which displaces the top

of the structures a distance yt. Figure 3(a)

is a vertical shear beam with cross-section area A, shear modulus G, and mass density

p. Figure 3(b) represents an iV-story building whose floors are strong and remain horizontal, forcing the weaker columns to be

vertical in the vicinity of the beam-column

joints. The two columns have Young's modulus E, and each has moment of inertia

I. The known wave equation governing the

dynamic displacements of (a) can be trans- ferred to the structure (b) provided their

stiff"nesses under the action of the force F,

and densities, are equivalent.

For the vertical shear beam (see, for example, Hurty and Rubinstein, 1964),

F ^yt

(14)

439
Co.) (b) (c) (d)

Cq) (^) (d)

Figure 1. (A) Model type A with wood floors pinned at each end; (B) Type B with floors

rigidly connected to weaJcer columns—(a) general view, (b) first mode shape, (c) second

mode shape, (d) first mode shape of half- height model. Details: 1, 0.031" x X^j^' aluminum

column; 2, 1" x 2" x 3y2"-long wood floor; 3, ^^S-lYi" roundhead wood screws, two per

hinge, %" apart; 4, ^%4" hole allowing adequate clearance; 5, ^f^' length of plastic tube,

Vs" I.D.; 6, 0.010" X 1" aluminum column; 7, 1" x 2" x 4"-long wood floor; 8, #8-iy4"

rotmdhead sheetmetal screws, two per joint, ^f^" apart; 9, ^764" hole; 10, #8 and Yi"

washers.

440
For the iV-story btiilding,

24EI yt F = N

Equating stiffnesses,

GA =

2AEII

(15)

(16)

Considering free vibrations of the vertical

sheax beam,

dx^ G dt^

(17)

Looking for a natural mode solution of the

form

leads to

where

y{x,t) = y{x)smu)t

y"(x) + Py{x) = 0

l2 P 2

Looking for a solution of the form

y = A cos fcz + B sin kx

with boundary conditions

y(0) = y'{l) = 0

leads to

(18)

(19)

(20)

(21)

(22)

TT

ifc/ = (2n-l)-, n = l, 2, .. (23)

Substituting equation (23) into (20) yields

where

= 1 , iVa = 3 , N2/Ni = 3 (25)

Transfer the stiffness from Figure 3(a) to

3(b) using equation (16), and transfer the

density by equating the total mass (ignoring the mass of the thin aluminum columns

compared with the floors):

pAl = Mf = Nmf = ^ - nif (26)

Selecting / = 24", 6 = 1", d = 0.01", m/ =

0.1 Ib./g, h - 4"; and using E = \Q x

10^ psi, g = 386.4 inM / = hS l\2\ and

substituting equations (16) and (26) into

(24), yields

/i = 1 • 4 cps , /2 = 4 • 2 cps (27)

The frequency of a building of half this height

IS

fi/2 = 2 • 8 cps (28)

Figure 4 reproduces photos of some of the

models and their mode shapes.

6. REFERENCES

1. Housner, G. W., and Brady, A. G., 1963,

Natural periods of vibration of buildings: JouT. Eng. Mech. Div., ASCE, v. 89, EM4, p. 31-65.

2. Hurty, W. C, and Rubinstein, M. F.,

1964, Dynamics of Structures, Prentice

Hall.

441
r

7^

-/77>

•J/

(b)

Figure 2. (a) Cantilevered bending beam placed vertically, (b) schematic for analysis of

type A model.

Figure 3. Two structures resisting horizontal forces by shearing action, (a) Cantilevered

sheeir beam placed vertically; (b) schematic for Einalysis of type B model; (c) interstory

detail.

442
(Tc) Cd)

Figure 4. (a) Type A and B models; (b) floor and column detail; (c) first mode of type B

model, frequency 1.9 cps; (d) second mode of type B model, frequency 5.6 cps.

443

Manuscripts Authored for

Panel Meeting but Not

Presented Orally

Determination of Windborne Missile Design Parameters

by

James R. McDonald*

ABSTRACT

Tornadoes and other high winds pick up and

transport objects and debris, which can cause

extensive damage to structures. This paper

addresses the problem of windborne missile

impact on facilities which, if damaged, could

result in injury to people or damage to the

environment. The problem has two parts; (1)

what missiles should be included in the criteria,

and (2) what is the impact resistance of different

wall and roof panels to the postulated missiles.

This paper briefly addresses both issues. The

document. Guidelines for Design and

Evaluation of Department of Energy Facilities

Subjected to Natural Hazard Phenomena,

(UCRL 15910) specified a missile criteria for

moderate and high hazard facilities. The

selection of the missiles to be considered and

their parameters by the Natural Hazards

Phenomena Committee are described in the

paper. A series of impact tests were then

conducted at Texas Tech University to identify a

set of wall and roof panels that could satisfy the

UCRL 15910 missile impact criteria. The

deterministic criteria is selected to satisfy stated probabilistic performance goals of the DOE

facilities. KEYWORDS: debris; impact; missiles;

tornadoes; wind; wind resistant design

I. INTRODUCnON

Model building codes (e.g. SBCCI, 1991) and

national wind load standards (e.g., ASCE, 1988)

do not require building designs to consider

loads produced by tornadoes. The probability

of a building at a specific geographic location

experiencing a tornado with particular wind

speed is considered too small for ordinary

buildings. However, there are circumstances

where design for tornadoes is necessary or

desirable. Facilities that have high dollar value

materials or perform critical functions, such as a

fur storage warehouse or a corporate computer

data center, may need to be designed to be

tornado resistant. Facilities that store or process

toxic or radioactive materials may also require

consideration of tornado effects. Nuclear

reactors have traditionally been designed to be

tornado resistant. Critical equipment and

facilities needed to conduct a safe shut down of

the reactor are housed in tornado-hardened

buildings or enclosures.

A tornado damages buildings or facilities by

three principal mechanisms: (I) wind, (2)

atmospheric pressure change (APC), and (3)

missile impact. This paper addresses the third

damage mechanism: tornado missile impact.

High winds and tornadoes pick up and transport

various objects and debris. Under the right set of circumstances, these objects may be

accelerated to high speeds, carried to great

heights, and travel long distances. Common

terms for debris transported by the winds are wind-generated or tornado-generated missiles.

Tornado winds are more efficient in transporting

missiles than "straight" winds in hurricanes and

thunderstorms.

When a tornado missile strikes the wall or roof

of a building, two independent responses are

possible: penetration or structural response. A

penetration impact is caused by a long,

prismatically- shaped object such as a timber

plank, pipe, or pole. When the missile .strikes the

wall, one of four possible events takes place:

(1) the object strikes the wall at some oblique

angle and bounces harmlessly off the wall

(oblique strike)

(2) the missile strikes essentially on end, and

penetrates some distance into the wall

(penetration)

(3) the missile strikes the wall with sufficient

force to scab material from back face of

wall; it may penetrate but does not

perforate the wall (back face scabbing)

(4) the missile passes through the wall and exits

at some slower speed (perforation with

residual velocity).

Professor of Civil Engineering, Texas Tech

University, Box 41023, Lubbock, TX 79409.

447
Structural respouse occurs when a large, heavy

object strikes a structural component and

produces excessive plastic deformations in the

material. An example is an automobile striking a reinforced concrete wall panel with sufficient

force to produce yielding of the reinforcing

steel and crushing of the concrete. If a heavy

object strikes a column with sufficient force, the

column could collapse and initiate a progressive

failure as other columns in the structural system

collapse one after the other. Structural response

impact is not addressed in this paper. Many buildings and facilities operated by the

U.S. Department of Energy (DOE) house toxic or nuclear materials that present potential danger

to plant personnel or the environment should a high wind or tornado damage the facility.

Recognizing the potential problem, DOE

through Lawrence Livermore National

Laboratories, supported the development of

Design and Evaluation Guidelines for DOE

Facilities Subjected to Natural Phenomena

Hazards. A committee consisting of wind,

earthquake and flood experts produced the

guidelines, which are referred to as UCRL

15910. The approach adopted by the Natural

Hazards Phenomena (NPH) Committee is to use

a deterministic design and evaluation criteria to meet probabilistic performance goals. Hill et al. (1991) describes the treatment of wind and AFC

loads on DOE facilities. The objective of the

study described herein is to establish the basis

for the missile criteria and impact-resistant

requirements for design or evaluation of IX)E

facilities.

Four facility-use categories are defined in the current version of UCRL 15910 (DOE, 1990):

(1) general use, (2) important or low hazard, (3)

moderate hazard, and (4) high hazard facilities. Only the latter two categories require

consideration of windbome missiles. A facility

will be subject to either high wind or tornado

design criteria, depending on its geographic

location. Both criteria require consideration of

windborne missiles. Tornado missile criteria is more conservative than straight wind.

The remainder of this paper is divided into three

sections that present (1) basis for missile

trajectory parameters, (2) impact resistance of

wall and roof panels, and (3) conclusions and

recommendations.

2. BASIS FOR MISSILE PARAMETERS

The principal questions to be answered about

tornado missiles are (1) what types are

transported by the winds, (2) how fast do they

fly, (3) how high, and (4) how far.

2.1 Windborne Missiles

I'ield damage experience is the best means for

identifying the types of missiles that can be

picked up and transported by high winds or tornadoes. For purposes of discussion, objects can be classified as lightweight, medium, and

heavy missiles (McDonald and Kiesling, 1988).

Lightweight missiles include roof gravel, pieces

of gypsum board, fiberglass insulation, sheet

metal panels, and small tree branches. These

missiles primarily break glass or damage

building finishes. Medium weight missiles are, by far, the most

common. The category includes pieces of

timber plank from damaged or destroyed wood

buildings, electrical conduit, plastic pipe, small

communication antennas, fence posts, small steel

joists, purlins, or girts. These missiles will be

picked up by thunderstorm or hurricane winds,

but are more likely to achieve the highest

trajectory speeds when picked up by tornado

winds. They can achieve speeds sufficient to

perforate ordinary stud wall construction, walls

with brick veneer, unreinforced and ungrouted

concrete masonry, and concrete wall panels less than four inches thick.

Heavy missiles include a variety of objects such

as large diameter pipes, utility poles,

automobiles, trash dumpsters, storage tanks, and

railroad cars. Except for the pipes and poles,

these objects tend to roll and tumble on the

ground rather than become airborne. The poles

and pipes can penetrate or perforate all walls

with equivalent resistance less than or equal to a 12 in. thick reinforced concrete wall. Large,

heavy objects cause structural response rather

than penetration failure.

Field investigations of many windstorms have

led to the identification and classification of

windborne missiles. Lightweight and medium

weight missiles are common in relatively weak

tornadoes, hurricanes and straight winds.

Transportation of heavy missiles is an indication

of high winds in an intense tornado.

Because of the adverse consequences of core damage to nuclear power reactors, the USNRC

design criteria requires plants to be designed for a very conservative list of heavy weight missiles

(USNRC, 1981). As a result of the USNRC

448
missile criteria, facilities that house equipment or systems required for safe shutdown of power

reactors are required to have 24-in. thick concrete walls and 18-in. roof slabs. This very

conservative criteria controls design and

contributes to high construction costs.

Probabilistic risk assessments (PRAs) of heavy

tornado missile events suggest that the

probability of reactor core damage from missile

impact is of the order of I x 10-8.

The probabilistic performance goals established

for DOE facilities in UCRL 15910 can be

achieved with much less conservative windborne

missile criteria. Because PRAs are time

consuming and expensive, a deterministic

approach was adopted, which leads to designs

that meet the performance goals. The approach

is based on post-storm damage data and

judgments of experts in wind damage

assessments.

2.2. Missile Flight Characteristics ,

Flight characteristics are the parameters that determine how fast, how high, and how far missiles travel in a windstorm. The important

parameters are weight, surface area, drag and lift coefficients, release mechanism and height of

missile at release. The weight, area and

aerodynamic coefficients determine if a missile,

when released, will be transported by the wind or merely fall to the ground under the influence of

gravity. The height above ground at release

determines how long a missile will remain

airborne before striking the ground. Another

factor is whether the missile is restrained or

unrestrained before being picked up by the

wind. If it is unrestrained, it may be simply

pushed along the ground by the wind. If it is restrained and the winds suddenly overcome the

restraint, the missile is more likely to fly higher

and faster than if it were not restrained.

In the case of a tornado, the location within the

tornado wind field where the missile is released

determines the trajectory it will follow. Missiles

that release on the right hand side of the tornado

path centerline (looking in direction of tornado

travel) will generally travel in the direction of

tornado translation with the trajectory

converging toward the centerline. Objects

picked up on the left side of the path centerline

are carried by the tangential winds and may

initially move in the opposite direction to tornado tranlation and travel around the back

side of the tornado vortex.

In field investigations, the initial location ofthe

missile and its final impact location usually can

be determined. What happens in between can

only be determined if movies or videos of the

storm were available. Such films are very rare. Thus, the only feasible approach to understanding missile trajectories is by computer

simulation.

2.3 Trajectory Simulation

Two approaches to trajectory simulation have

been used in the past: deterministic and

probabilistic.

2.3.1 Deterministic

Missile aerodynamics are modeled using tlight

parameters, injection mechanism, wind field

model and dynamic equations of motion

(McDonald, 1975). Flight parameters include

drag and lift coefficients, area, and weight of

missile. Drag and lift coefficients are obtained

from wind tunnel tests. The flow of wind in the

tornado vortex is modeled as realistically as

possible. Photogrammetric analyses of the flow

observed in movies or videos of tornado funnels

are used to identify the wind flow patterns as well as theoretical considerations. Several

tornado wind field models have been published

in technical literature. The DBT-T? model by

Fujita (1978) is based on photogrammetric

analysis and extensive field observations of

damage patterns by Dr. Fujita. Other models

include lloecker (1960) and Simiu and Cordes

(1976).

Injection of the missile into the wind field can be

as simple as releasing the missile at some

preselected point within the three-dimensional

tornado wind field where it will achieve

maximum speed, height and/or distance traveled. Another approach is to systematically place the

missiles at regular intervals across the tornado

path and release them at some predetermined

wind speed. Trajectories are then calculated for

each missile and the worst conditions are

tabulated.

Missiles can be modeled as a three-degree-offreedom point model or a six-degree-offreedom rigid model. Calculations are simpler

for the point model. The approach was used to successfully predict the impact locations of the

steel beams that were picked up by the Bossier

City Tornado (McDonald, 1981).

The six-degree-of-freedom model requires a set of drag and lift coefficients for each orientation

449
of the missile in space relative to the wind

direction vector. Such measurements have been

made in the wind tunnel for a few objects. The

computational effort for the six-degrce-offreedom model is many times larger than for the three-degree-of-freedom model, with no solid

evidence that it is more accurate or realistic (Redman et al., 1976). Neither model considers

the effects of small scale turbulence in the

tornado wind field.

2.3.2 Probabilistic Simulation

The probabilistic approach to trajectory

simulation is essentially a Monte Carlo

simulation. Tornado wind field characteristics

(maximum wind speed, path width and

direction) are randomly selected as are certain

missile parameters at each time step of the

trajectory calculations. The missiles are randomly selected according to some weighted

frequency scheme. By simulating many years

of tornado missile experience, statistics on the

types of missiles and their trajectory

characteristics can be obtained. See for example

Twisdale and Dunn (1978). The site specific

approach, which attempts to model the real situation as accurately as possible, must be

applied at each facility being considered for

tornado effects. To carefully model a site by

this method requires extensive effort and

considerable cost.

2.4 VCRL 15910 Missile Criteria

Since UCRL 15910 is based on a deterministic

approach to satisfy probabilistic performance

goals, the deterministic approach to tornado

missile criteria was adopted by the NHP

Committee.

Since general use and low hazard facilities are designed by criteria in model codes or national

standards, which have no provisions for

winborne missiles, UCRL 15910 also does not specify a missile criteria for these facilities. Moderate and high hazard facilities located in areas where tornadoes are not a serious threat

have a minimum windbome missile criteria. The

minimum missile criteria accounts for objects or

debris that could be picked up by straight winds,

hurricanes or weak tornadoes.

Moderate and high hazard facilities located in tornado prone areas must satisfy tornado missile

criteria. Because the tornado missile trajectory

calculations require specialized software and

involve considerable time and effort, the NPH

Committee established missile criteria that was

independent of tornado design wind speed. This was possible because the range of tornado

design wind speeds for the 26 DOE sites was

relatively narrow. Thus a single criteria for all moderate and high hazard facilities could be

established.

Two medium and one heavy weight missiles are considered in the design and evaluation of DOE

facilities. A timber missile is typical of debris

found in the destruction of office trailers,

storage sheds, residences and other light framed

timber structures. Hundreds of these missiles can be generated in the destruction of a

residential neighborhood. A steel pipe missile

represents a class of debris that includes

electrical conduit, liquid or gas piping, fence

posts and small columns. The pipe missile is less frequently available for transport and does not

fly as high or as fast as the timber missile, but is more destructive upon impact. Tornadoes can

roll and tumble automobiles, pickup trucks,

small vans, forklifts and storage tanks of

comparable size and weights. These missiles do

not become airborne, but can damage wall

panels, frames or columns by excessive

structural response.

The three missiles produce varying degrees of

damage. A specific type of construction is required to stop each missile. The timber missile

is capable of breaking glass and perforating

curtain walls or unreinforced masonry walls.

Reinforced concrete or masonry walls are required to stop the pipe missile. Timber and

pipe missiles can perforate weak exterior walls

and emerge with sufficient speed to perforate

interior walls and damage HVAC ducts, filter enclosures or control equipment. The impact of

a rolling or tumbling automobile produces

failure by excess structural response. Bearing

walls, rigid frames and exterior columns are

particularly susceptible to heavy missile impacts.

Specific missile criteria in UCRL 15910 are summarized in Table 1. Critical missile

parameters include (1) material, (2) weight, (3)

impact speed and (4) maximum height achieved.

Impact speeds are expressed in terms of

horizontal and vertical components. They

should not be combined vectorially, because the

maximum horizontal speed does not occur at the same point in the trajectory as the maximum

vertical speed. The maximum horizontal speed

generally occurs at maximum height of

trajectory above ground, while the maximum

vertical speed occurs at impact with the ground.

450
In general, the maximum horizontal speed is used in the design of walls or other vertical

barriers. The maximum vertical speed is used to design roof decks and other horizontal barriers.

Both maximum horizontal and vertical speeds

should be considered (separately) on sloping

surfaces.

The designer or evaluator using UCRL 15910

criteria is not required to determine missile

impact speeds. He designs new barriers or

assesses existing barriers based on results of

impact tests. The impact test results for the

timber plank and steel pipe missiles are described in the next section. Prom the test

results, specific wall or roof barriers are known

to meet the UCRL 15910 missile criteria.

3. IMPACT RESISTANCE OF WALL

AND ROOF PANELS

3. 1 Background

Penetration, perforation and scabbing equations

used for tornado missile impact design in the

1960s were derived from military applications of

ballistic missile impacts. The masses and shapes

of the military projectiles were quite different

from those generated by tornado winds.

Gwaltney (1968) presented a comprehensive

review of missile generation and protection in nuclear power plants. A number of penetration

formulas are referenced in the Gwaltney paper:

(1) Modified Petry formula

(2) Ballistic Research Laboratory (BRL)

formula

(3) National Defense Research Committee

(NDRC) equation

(4) Army Corps of Engineers formula

Because none of the ballistic formulas were

directly applicable to tornado missile impact

design, several agencies and companies initiated impact test programs. Most of the missiles and

barriers tested in the 1970s related to the U.S.

Nuclear Regulatory Commission (USNRC)

criteria for design of commercial nuclear plants. Test results were reported for steel rods (rebar),

steel pipes, timber planks, utility poles, and

automobiles. Based on test results, a modified

version of the NDRC equation was proposed by

Kennedy (1975). Rotz (1975) developed an

empirical equation based on a series of impact

tests sponsored by Bechtel Power Corporation

(Velasso, 1975). The NDRC and Rotz equations

are applicable to steel pipe missiles for

impacting on reinforced concrete panels. The

test results were for impact speeds higher and

panels much thicker than those required by

UCRL 15910.

3.2 Missile Impact Tests

The Tornado Missile Impact Facility (TMIF) at Texas Tech University is capable of testing the

resistance of various wall and roof panels to the

impact of medium weight missiles. The tests include both the 2x4 timber plank (15 lbs) and

the 3-in. diameter steel pipe (75 lbs). A

compressed air cannon is capable of propelling

the 2x4 plank to 150 mph and the pipe to 75

mph. Figure 1 shows the general arrangement

of the facility. Figure 2 is a photo of the cannon. A rigid reaction frame supports the test panels. The speed of the missile just prior to impact is measured by an optical timing gate.

3.2.1 2x4 Timber Plank Missiles

Three series of tests were conducted with the 2x4

plank missile. The first series of tests determined the impact resistance of conventional

residential wall construction as defined in Architectural Graphics Standards (Packard, ed.,

1981). Wall panels 4 ft wide by 8 ft high were

constructed with 2x4 Douglas fir studs at 16 in. on center. The inside wall material was 12 in. gypsum board. The exterior cladding is described in Table 2. As shown in Table 2, the

2x4 plank pertorates the stud walls at speeds of

about 55 mph. Missile trajectory simulations of

the 2x4 plank suggest that if the missile is transported by the winds, it will most likely

achieve speeds greater than 55 mph. Thus,

ordinary stud wall construction will be

perforated by the 2x4 timber plank missile.

Stud walls clad with insulation board and brick veneer are capable of resisting the 2x4 plank

missile at speeds less than 120 mph (see Table

2)-

The second series of tests involved impact of the

timber plank missile on concrete masonry walls

(McDonald and Bailey, 1985). Eight-in. and 12-

in. concrete masonry unit (CMU) walls with

various combinations of reinforcement and

grout in the vertical cells were tested. Unreinforced and ungrouted walls offer little resistance to perforation. If the missile strikes the face shell of an 8-in. wall, it perforates at impact speeds as low as 60 mph. The 12-in.

CMU wall offers slightly more resistance

451
because the face shell is thicker. Reinforcing

each vertical cell with a 1/2-in. diameter (#4)

rebar and filling the cell with a high slump

concrete grout produces a wall that is capable of

stopping the 2x4 plank at any impact speed. At

impact speeds greater than 100 mph, the nndssile will splinter. Horizontal trussed joist

reinforcement at 16 in. on center helps prevent

cracking of the masonry panel. Table 3

summarizes the results of the second series of

tests on CMU walls.

The third series of tests involved impact of the

2.\4 timber plank on unreinforced and

reinforced concrete wall and roof panels. The

2x4 timber plank will not perforate a 4-in.

unreinforced concrete panel (4000 psi nominal

compressive strength), but it will badly crack it. Adding 1/2-in. diameter rebar at 12-in. on

center each way on the inside face of the panel

will prevent the cracking. Impacts of the plank

at speeds greater than 100 mph on slabs thicker

than 4 in. will cause the missile to splinter.

3.2.2 Three-In. Diameter Steel Pipe Missiles

A series of impact tests were conducted on

reinforced concrete and masonry panels that were designed to satisfy the UCRL 15910 missile

impact criteria. Six concrete walls with 6-in., 8-

in., 9-in., or 10-in. thicknesses were tested at impact speeds of 50 and 75 mph. All concrete panels performed better than the Modified

NDRC equation predicted.

Eight-in. and 12-in. CMU walls that were

grouted and reinforced with 1/2 in. diameter

rebar were tested. The 3-in. diameter pipe

perforated the 8-in. CMU wall at an impact

speed of 75 mph. Back-face scabbing occurred

at an impact speed of 50 mph. Pieces of

material weighing approximately 7.5 lbs came

off the back face of the panel. The threshold

perforation speed of the 12-in. CMU wall was

found to be approximately 60 mph.

Two other masonry panels were tested with the

pipe missile. One was a 9 1/2-in. cavity wall

consisting of two wyths of clay brick masonry

and 2500 psi concrete in the cavity. Number 3

rebar was placed vertically at 8-in. on center. Horizontal joint reinforcement was placed at 16-

in. on center. The panel was impacted by the

steel pipe at 50 mph. Cracks appeared on both

the front and back face, but no perforation on

scabbing occurred. This panel was not tested at 75 mph. The second masonry panel was

constructed with 8-in. reinforced and grouted

CMUs and a 4-in. clay brick veneer. This panel was impacted by the pipe at 50 mph. The

impact speed appeared to be near the threshold

of back-face scabbing. Table 4 summarizes the

resuhs of these tests.

4. CONCLUSIONS AND

RECOMMENDATIONS

Several series of impact tests using the 2x4

timber plank and the 3-in. diameter steel pipe

have been conducted at the Texas Tech

University TMIF. Results of these tests form the

basis for the following recommended wall and

roof panels for missile protection of moderate

and high hazard facilities as defined in UCRL

15910. (See Table 2 for missile criteria.)

4.1 Moderate Hazard - Straight Wind

The 2x4 timber plank impact at 50 mph requires

a minimum 8-in. CMU wall with one #3 rebar

grouted in each vertical cell and trussed

horizontal joint reinforcement at 16 in. on

center.

4.2 M(xlerate Hazard - Tornado

The timber plank impact at 100 mph requires a minimum 8-in. CMU wall with one #3 rebar

grouted in each vertical cell and trussed

horizontal joint reinforcement at 16 in. on

center; 6-in. concrete wall with #3 rebar at 12 in. on center each way 1 1/2-in. from the inside

face. The 3-in. pipe impact at 50 mph requires a minimum 12-in. CMU wall with #4 rebar

grouted in each vertical cell and trussed

horizontal joint reinforcement at 16 in. on

center; 8-in. concrete wall with #4 rebar at 12-in. on center, each way 1 1/2-in. from each face.

4.3 High Hazard - Straight Winds

High hazard facilities for straight winds have the same criteria as moderate hazard, except the wall

panels are susceptible to impact up to 50 ft above ground level.

4.4 High Hazard - Tornado

The 2x4 timber plank impacting at 150 mph

requires a minimum of 8-in. CMU wall with one #4 rebar grouted in each vertical cell and trussed

horizontal joint reinforcement at 16-in. on

center; 6-in. concrete wall with #4 rebar at 12-in. on center, each way, 1 1/2-in. from each face. The 3-in. diameter pipe impacting at 75 mph

requires a minimum 10-in. concrete wall

452
reinforced with #4 rebar at 12-in. on center,

each way, 1 1/2-in. from each face.

4.5 Recommendations

The wall panels described above will satisfy the

criteria of UCRL. 15910. The criteria will meet

the performance goals for the category of

facility considered. Additional tests are needed

to determine the resistance of doors, grills and

other wall cladding systems. The impact

resistance of steel plate panels is untested.

Additional tests are needed to validate the BRL

equation for steel plates.

Considerable additional study is also needed to establish and verify procedures for calculating

structural responses to heavy weight missiles.

5. ACKNOWLEDGEMENTS

Financial support for the Texas Tech University

missile impact tests from the National Science

Foundation and Lawrence Livermore National

Laboratory are gratefully acknowledged.

6. REF^ERENCES

1. ASCE, 1988: Minimum Design Loads for

Buildings and Other Structures, American

Society of Civil Engineers, New York, NY.

2. DOE, 1990: Design and Evaluation

Guidelines for DOE Facilities Subjected to Natural Phenomena Hazards, UCRL 15910,

U.S. Department of Energy, Washington,

D.C.

3. Gwaltney, R.C., 1968: Missile Generation

and Protection in Light-Water-Cooled

Reactor Plants, ORNL-NSIC-22, Oak Ridge

National Laboratory, Oak Ridge, TN.

4. Fujita, T.T., 1978: Workbook of Tornadoes

and High Winds for Engineering

Applications, SMRP Research Paper No.

165, Department of Geophysical Sciences,

the University of Chicago, Chicago, IL.

5. Hill, J R., McDonald, J R. and Miuray, R.C.,

1991: Probabilistic Performance Goals for

Wind and Tornado Design, 23rd Joint Panel

Meeting on Wind and Seismic Effects.

6. Hoecker, W.H., Jr., 1960: "Wind Speed and

Air Flow Patterns in the Dallas Tornado of

April 2, 1957," Monthly Weather Review,

Vol. 88, pp. 167-180.

7. Kennedy, R.P., 1975: A Review of

Procedures for Analysis and Design of

Concrete Structures to Resist Missile Impact

Effects, Holmes & Narver, Inc., Anaheim,

CA.

8. McDonald, JR., 1975: Flight

Characteristics of Tornado-Generated

Missiles, Institute for Disaster Research,

Texas Tech University, Lubbock, TX.

9. McDonald, JR., 1981: "Incredible

Tornado Generated Missiles," Proceedings,

Fourth U.S. National Conference on Wind

Engineering Research, Vol. 1, Ifniversity of

Washington, Seattle, WA, pp. 29-36.

10. McDonald, J R., 1990: "Impact Resistance

of Common Building Materials to Tornado

Missiles," Journal of Wind Engineering and

Industrial Aerodynamics, Vol. 36, Filsevier Science Publishers B.V., Amsterdam, pp. 717-724.

11. McDonald, J R. and Bailey, J R., 1985:

"Impact Resistance of Masonry Walls to Tornado-Generated Missiles," Proceedings,

Third North American Masonry

Conference, The University of Texas at Arlington (June 3-5, 1985), pp. 84-1

through 84-1 1.

12. McDonald, J R. and Kiesling. E.W., 1988:

"Impact Resistance of Wood and Wood

Products to Simulated Tornado Missiles,"

Proceedings, International Conference on Timber Engineering, Seattle, WA

(September 19-22, 1988).

13. Redman, G.H. and Associates, 1976: Wind

Field and Trajectory Models for TornadoPropelled Objects, EPRI-308, Electric Power

Research Institute, Palo Alto, CA.

14. SBCCI, 1991: Standard Building Code,

Southern Building Code Congress

International, Inc., Birmingham, AL.

15. Rotz, J.V., 1975: Summary of Missile Test

Results on Concrete Panels, Proceedings,

ASCE Specialty Conference on Structural

Design of Nuclear Plant Facilities, Vol. 1-A,

New Orleans, LA.

16. Simiu, E. and Cordes, M., 1976: TornadoBorne Missile Speeds, NBSIR 76-1050,

National Bureau of Standards, Washington,

DC.

453
17. Twisdale, L.A. and Dunn, W.L., 1981:

Tornado Missile Simulation and Design

Methodology, Vol. I: Simulation,

Methodology, Design Applications and

TORMIS Computer Code, EPRl NP-2005,

Eilectric Power Research Institute, Palo Alto,

CA.

18. Twisdale, L.A., Dunn, VV.L. and Chin, J., 1978: Tornado Missile Risk Analysis, EPRl

NP-768, Final Report, Electric Power

Research Institute, Palo Alto, CA.

19. USNRC, 1981: Standard Review Plan

3.5.1.4, U.S. Nuclear Regulatory

Commission, NlIREG-08000. Washington,

DC.

20. Vassallo, F.A., 1975: Missile Impact Testing

oj Reinforced Concrete Panels, Report No.

HC-5609-D-1, Calspan Corporation,

Buffalo, NY.
Table 1. Windborne Missile Criteria in UCRL 15910

_M.iderate,ilazani .tiigkHazaK

Straight winds 2x4 timber plank, 15 lb @ 50 mph (horiz.); max. 2x4 timber plank, 15 lb @ 50 mph (horiz.); max.

height 30 ft. height 50 ft.

Tornadoes 2x4 timber plank, 15 lb @ 100 mph (horiz.); max.

height 150 ft; 70 mph (vert.)

3 in. dia. steel pipe, 75 lb @ 50 mph (horiz ); max.

height 75 ft; 35 mph (vert.)

2x4 timber plank, 15 lb @ 150 mph (horiz ); max.

height 200 ft; 100 mph

(vert.)

3 in. dia. steel pijie, 75 lb @ 75 mph (horiz.); max.

height 100 ft; 50 mph

(vert.)

3000 lb automobile @

25 mph; rolls and tumbles

Table 2. Results of 2x4 Timber Plank Impact Tests on

Typical Stud Wall Panels (McDonald, 1990)

Perforation

PaneLDe&criptiQn Speadv-triph

Masonite Siding 54

Insulation Board and Masonite Siding 54

Plywood (1/2 in.) and Masonite Siding 52

Plywood (1/2 in.) 52

Plywood (3/4 in.) 53

Stucco 53

Lapboard Siding 53

Insulation Board and Lapboard Siding 52

Insulation Board and Brick Veneer 120

455
Table 3. Estimated Speed for Perforation of CMU Walls

by 2x4 Timber Plank Missile (McDonald, 1990)

Perforation

Wall Construction Speed, mph

8-in CMU Wall

Grout and re bar in every cell >130i

Grout only in every cell >1302

Intermittent vertical reinforcement 653

Horizontal Joint Reinforcement Only 65^

IJnreinforced 60

1 2-in. CMU Wall

Groin and rebar in every cell >130

Grout only in every cell >130

Intermittent vertical reinforcement 70^

Horizontal Joint Reinforcement Only 703

Unreinforced 65

^Missile will splinter.

^Missile will not perforate, but wall may crack.

3Face shell of ungrouted cell will be perforated.

Table 4. Results of 3-in. Diameter Steel Pipe Impact Tests

Panel Description

Impact

speed, mph Damage

6-in. reinforced concrete panels

8- in. reinforced concrete panels

9-in. reinforced concrete panel

9-in. reinforced concrete panel

10- in. reinforced concrete panel

8-in. CMU panel with vertical

cells reinforced and grouted

12-in. CMU panel with vertical

cells reinforced and grouted

35 Small cracks, no perforation or scabbing

50 Small cracks, no perforation or scabbing

50 No cracks, perforation or scabbing

75 Radial cracks on back face, 1 1/2-in.

penetration, impending scabbing

75 No significant cracks, perforation or scabbing

50 Missile perforated panel; 7.5 lb pieces came

off back face

60 Cracks on front and back face; impact speed

near threshold for scabbing

456
457

Advances in Wood Engineering and Construction Research

by

Erwin L. Schaffer'

ABSTRACT

The Forest Products Laboratory has served

the public for nearly 80 years and will

continue to serve by providing information

on wood-based material properties and

advice on how to apply this information to structural products. The audience is broad

and our national and international

cooperators are many. We welcome your

requests for assistance and stand ready to provide help through our many media. The

examples of research we've collaborated in over the last 10 to 15 years illustrate that we

care about the efficient use of our timber

resource and the world's resource, while

promoting the durability, economy, and

safety of wood products in construction.

KEYWORDS: construction; engineering;

design stress; fire endurance; wood

assembly.

1. INTRODUCTION

The Forest Service of the U.S. Department

of Agriculture supports the largest, most

comprehensive forestry research program m

the world. The research for utilization of

timber—from conversion to lumber and

other forest products to recommendations

for efficient use in construction—has been

focused at the Forest Products Laboratory

(FPL) in Madison, Wisconsin, since its establishment in 1910.

Because timber testing was recognized early as both an industrial engineering problem

and an important aspect of forest

management, the brand-new FPL found its fund of knowledge in engineering already

well established. Researchers concentrated

on the fundamental problems of determining

the mechanical properties of wood and wood

products. They evaluated the effects of fire,

drying, and preservatives on mechanical

properties, and prepared lumber grading

rules. Efforts in characterizing the strength

properties of structural timbers led to one of

the earliest ASTM standards for grading

structural timbers. Broadening of the species

base in wood utilization by FPL also had a strong and favorable impact on national

forest management.

Researchers in the 1920s developed the

equipment for testing large structural

members, expanding FPL's ability to

develop design data for wooden columns

used in buildings, bridges, and other such

structures.

Large-scale construction research was

successful in the design and economics of

the laminated arch in the 1930s. Exhaustive

screening of appropriate fasteners and joint

configurations provided the basis for modem

industry standards. The broad public need

for information on the engineering

properties of wood and FPL's wealth of

knowledge complemented each other in the

publication of the first edition of the Wood

Handbook in 1935.

^Assistant Director, Wood Products

Research, USDA Forest Service, Forest

Products Laboratory, Madison, Wisconsin

53705-2398. The Forest Products

Laboratory is maintained in cooperation

with the University of Wisconsin. This

article was written and prepared by U.S.

Government employees on official time, and

it is therefore in the public domain and not

subject to copyright.

459
Expanded fire-testing facilities and the

development of the "stressed-skin" or sandwich panel principle, prefabricated wall

units, and improved moisture barriers were

also part of FPL's contribution in the 1930s.

When World War II broke out, FPL

engineers sought improved materials and

construction for aircraft—first assessing

alternate solid-wood species, then

experimenting with the Lab's "sandwich"

construction. Laminated product strength

testing helped fill the need for ship timbers

and other military structures.

After the war, renewed research by FPL in

housing showed that lighter framing could conserve wood and still meet performance

standards. Those principles were supported

by a definitive book on construction

techniques for the homebuilder. Engineers at FPL developed a series of plans for more

affordable housing and studied earthquake

and tornado damage to better understand a

building's response to natural forces.

As homeowners of the 1970s increased their awareness of ecological concerns and the

need for energy conservation, FPL

responded with guidelines for thermal

insulation, moisture and noise control, and

renovation and remodeling techniques. The

concerns of economy, strength, and speed in

new construction were met in the TrussFramed System, a prefabricated system

suitable for framing houses and small

commercial buildings.

In the 1980s, basic work in determining

strength properties of members and systems

continued, including the assessment of

lumber grades and the effects of

environmental conditions, loads,

constructions, fasteners, and treatments.

Building components are increasingly

modelled and analyzed to help assess and

meet performance standards. A structural

test frame subjects an entire house to realistic loads that help measure the

performance of materials and construction.

Computer-aided modelling of structural

behavior, moisture movement,and

performance of wood members under fire

exposure are opening new opportunities for

rational analyses of entire building systems.

2. MEETING USER'S NEEDS

The Forest Products Laboratory has

evidenced the ability to meet the needs of a variety of interests. The most popular view

of FPL is the traditional one for any R + D

laboratory—a producer and disseminator of

research results. The FPL is this but much

more! We serve a multitude of "users" via

many media. Publications—from highly

technical reports, to summary compendiums

that integrate results in major areas of

research, to popular articles for the

journalistic press—are a key medium, but

other less obvious media are sometimes

tools as well (see Appendix).

Whom do we view as "users" of our

research, and how do we serve them? Let's deal with answers to these simple questions

to illustrate how we relate to our "audience"

and might better serve them. Our real

challenge is to "package" and "facilitate" the use of our research.

We've attempted to group our national and

international research-users as follows:

User group Specific users

General public Consumers

Industry Companies,

associations

Legislative bodies Federal, State Regulatory groups Code-making agencies

Federal and State agencies

Standards' writers ASTM, ANSI,

ISO, DIN

Professional/Technical ASCE. CIB, RILEM

groups

Educators Universities Foresters Federal, State, Private

Research peers

460
As we develop our research planning and

conduct, we identify whom we expect to "serve" by conducting the work and through

what media we expect to transfer the results to the user. Technology transfer, therefore, is taken seriously as a responsibility of

researchers. Of the users cited, providing

research results to research peers solely to build a knowledge base is currently

considered the weakest use of our resources.

Based upon this variety of users, what are

the typical media employed for technology

transfer in addition to publications? Media

are as many and varied as user groups:

Presentations: Meetings of common

interest groups, which provide unique

opportunities to pass the word.

Advice: Inquiries or informal meetings

regarding a common concern that draw

information from completed research.

Committee Participation: Incorporation of

research results into appropriate standards,

guides, specifications, design criteria, etc.

Cooperation: Cooperation with users who

would like to adopt results but require some

assistance to do so.

Translation: Interpretation or modification

of research results to allow ready adoption

and use by educators, consumers, legislators,

code-making agencies, and regulators.

Review: Examination of proposed

guidelines prepared by others for

consistency with completed research.

Extrapolation/Interpolation: Evaluation

and guidance on extending research results

to similar applications.

Reports: An opportunity for users to

interact with researchers and benefit from all stages of research—from planning to

performance through review of completed

work.

All of these media reflect the variety of roles

that FPL researchers satisfy appropriate to the need.

The FPL is a problem-solving R + D

laboratory. The above media and users

reflect the intense effort to generate useful

information. We see ourselves as advisors

only. We have no regulator or standardswriting authority, but we will provide

assistance to those who have such authority.

A typical example is our advisory fimction

to the National Forest Products Association

on design procedures, criteria, and

procedures for setting allowable properties

for structural lumber and fastenings that lead

to the National Design Specification.

3. CODE IMPACTING RESEARCH

The Forest Products Laboratory, as a member of the wood resource community,

has a long history of providing guidance,

solving problems, and generating new

concepts and products that have had an

impact on codes and standards. Some wood

engineering research and technology transfer

activity over the last 10 years should be of

special interest to this Pacific Rim Group.

Following are examples of activities that

have resulted in useful products, which may

not be familiar to you. We hope that briefly

describing the research and results will be

informative as well as useful.

3.1 Truss-Framed System

The Truss-Framed System (TFS) is an

innovative new building system for

residential and light commercial

construction. The system offers substantial

savings in both materials costs and

construction time, with improvement in

structural durability and strength. The idea

for TFS was conceived at FPL in the late

461
1970s to meet the need for less costly, high

quality home construction.

The system's key structural component is a unitized floor-roof truss, which is joined

together by regular wall studs and spaced at 24-in. (610-mm) intervals. The system is a marriage of factory and site-built

construction. Trusses are fabricated in a

plant under quality conditions and delivered

prebuilt to the construction site, where they

are erected by a small crew or with light

mechanical equipment such as a crane or

forklift (Fig. 1). The entire unitized frame is constructed from nominal 2- by 4-in. (38- by

89-mm) dimension lumber rather than the

more expensive nominal 2- by 8-in. (38- by

191 -mm) or 2- by 10-in. (38- by 184-mm)

lumber used in conventionally designed

homes (Fig. 2). Because the trusses can be

designed to span the width of most homes,

supports are not needed in the basement and

load-bearing walls are not necessary on the

first floor. Once connected with sheathing,

the structurally engineered trusses create a strong and durable wood frame building

(Fig. 3).

Truss framing was developed after field

observations of building damage from high

winds or earthquakes showed that roof-towall or wall-to-floor joints often failed

before the structural members. These

observations were confirmed by FPL full- scale house tests. Since the unitized frame

assures structural continuity between the

critical joints, TFS buildings are often

sturdier than those constructed through

conventional means.

The innovative system also requires about

30 percent less structural framing lumber

than the conventional stick-built home.

Because less on-site effort is needed to

frame a structure, the average size home can

usually be erected in a few hours. As a

result, builders have reported savings of

about 10 percent of the overall costs by

using the TFS building technique.

Since January 1982, U.S. builders have used

TFS to construct over 2,000 homes,

apartments, and office buildings in 31 states. The TFS has been assigned a public patent

and is available to anyone who wishes to use

it. A TFS Construction Manual has been

published by the National Association of

Home Builders Research Foundation (1983)

in cooperation with the Forest Products

Laboratory. The manual includes sections on

the design, detailing, and construction of

TFS buildmgs.

3.2 Calculated Fire Endurance

To improve flexibility in structural wood

design, efforts continue to develop models

and design procedures for determining the

fire endurance ratings of wood members and

assemblies. Traditionally, the fire endurance

rating has been obtained by testing the

assembly in an ASTM E 119 (ASTM, 1983)

test furnace. These ratings are published in

listings such as the Underwriters

Laboratories' Fire Resistance Directory

(1986), Gypsum Association's Fire

Resistance Design Manual (1984), and those

in the model building codes. These listed

ratings are limited to the actual assembly

tested. As a result, designers are limited in

the type of fire rated assembly that they can

use. Modifications such as adding insulation,

changing member size, changing or adding

interior finish, or changing the spacing

between members are generally not

permitted. Extrapolation or interpolation of

ratings to or for similar assemblies is often

questionable. In recent years, two fire

endurance design procedures for wood that

allow greater flexibility have gained U.S.

building code acceptance. Two codeapproved calculation procedures have been

advanced by the National Forest Products

Association to predict fire endurance. These

are (1) additive calculation procedure for

walls and floors, and (2) procedure for

design of 1-h fire-resistive exposed wood

members (nominal >6 in. (140 mm). In

462
addition, other procedures have been

proposed or are being developed.

The fire endurance of a wood member or assembly depends on the performance of its protective membranes (if any), the charring

of the structural wood element, and the

structural capacity of the remaining

uncharred portions of the structural wood

elements. Models have also been proposed

for fire-exposed light-frame floor

assemblies, fire-exposed glued-laminated

beams, and fire-resistive coatings on wood.

The first two models emphasize

compatibility with reliability-based design

for fire exposure performance.

3.2.1 Fire-Exposed Light-Frame

Floor Assemblies

Woeste and Schaffer (1981) developed a model for time-to-failure of fire-exposed

unprotected wood joist floor assemblies.

Replicate ASTM E 119 (ASTM, 1983) tests were run to validate appropriate values for

the critical parameters. The model has been

extended to floor-truss assemblies. Figures 4

and 5 show how the truss assembly is designed and how time to failure of

unprotected truss assemblies theoretically

varies with applied load. This model is used

as part of a second-moment reliability

analysis of the floor assemblies. In the

future, reliability analysis will hopefully

provide systematic and realistic evaluations

of building fire safety.

3.2.2. Fire-Exposed Glued-Laminated Beam

A second approach to evaluating the fire

endurance of a wood member is to assume

the uncharred region consists of layers

(Fig. 6). The compressive and tensile

strengths and modulus of elasticity of each

layer are assumed to be fractions of the

room temperature values. Using one 1.5-in.

(38-mm) heated layer with reduced

properties, Schaffer et al. (1986) analyzed a beam using transformed section analysis.

This result was then used to calculate an

equivalent zero-strength layer (6) The 5 was

estimated to be 0.3 in. (7.6 mm) thick. This

zero-strength layer (5) was added to the char

depth (PO to obtain the total zero-strength

layer. The rest of the member was then

evaluated using room temperature property

values.

Bender et al. (1985) developed a reliabilitybased model to predict the strength of gluedlaminated beams under normal temperature

conditions. Using the approach of a + b

zero-strength layer, they extended the model

to include fire endurance analysis. The char

rate (P) and thickness of the zero-strength

layer (6) are assumed to remain constant.

The glued-laminated model uses

transformed section analysis to determine

the stresses within the laminates. In the fire

endurance model, the char depth is increased

at each time increment until the calculated

stresses within the laminates exceed the

corresponding tensile strength values. The

critical moment permitted by lateral

torsional buckling is also calculated. The

results compare favorably with tests and

allow both a mean time to failure and

variability to be calculated. The method

allows more flexibility in design for fire

performance through the ability to control

lumber quality to enhance fire endurance in

fabrication of glued-laminated beams, and

also provides a basis for reliability-based

strength fire design.

3.2.3 Fire-Resistive Coatings

At this time, the fire endurance of wood

members needs to be improved. The steel

industry improves the fire endurance of steel

members by covering them with fireresistive materials or coatings. Fire-resistive

coatings are not available for use on wood.

Coatings available for wood are only

designed to reduce flame spread. We

evaluated the ability of fire-retardant

coatings and fire-resistive coatings to

improve the fire resistance of plywood

463
(Figs.7,8). Based on these small-scale tests (White, 1984), we developed empirical

equations that can be used to determine the

coating thickness needed to increase the

time before a given char depth is obtained in

the standard fire test. These equations can be

combined with fire endurance models for a fire-exposed wood member or assembly to obtain the fire endurance rating of a fireresistive coated wood member or assembly

(White, 1986). Validation of these equations

is underway.

3.3 Gypsum Wallboard Protected

Wood Assemblies

The rate of heat release (HRR) is

increasingly used to assess the overall

contribution to fire growth of both

individual materials and assemblies.

Gypsum panel (0.6 in. (16 mm) thick)

protected wall assemblies containing either

wood members treated with fire retardants

or steel members were exposed to fire using

ASTM Method E 119 (1983). The purposes

of the test were (1) to obtain additional

information on the performance of fireretardant-treated, wood-based structural

assemblies exposed to fixe and (2) to extend

methods of measuring the rate of heat

release in individual construction materials

to full-scale structural assemblies.

The results of this study provide information

about the fire performance of fire-retardanttreated wood assemblies in terms of HRR.

At least 23 min elapsed from the start of the

ASTM E 119 test to the time when wall

assemblies containing treated wood began to contribute heat (Fig. 9). This result

illustrates the effectiveness of gypsum

wallboard and fire-retardant-treated studs in

restricting contribution of heat from the

assembly to fully developed fires within

compartments. Two factors contribute to the

substantial time delay in HRR: the thermal

barrier effect of the gypsum wallboard and

the fire-retardant treatment of the wood

members. Heat contributions were in the

range of up to 80 to 100 Btu/ft^/mm (15,128

to 18,910 W/m2) near the end of the test. Average HRR values for the assemblies

containing treated wood were 20 to

70 Btu/ft2/min (3,782 to 13,237 W/m^) over

the active period of heat release.

Results of this study demonstrate that

current methods used to measure HRR in

individual construction materials can be

extended to full-scale wall assemblies, and

that protection of wood stud assemblies with

gypsum wallboard can inhibit stud

contribution to fully developed fires for

20 mm or more.

3.4 Roof, Floor, and Wall

Composite Action

Although the structural performance of a

full-scale, light-frame house has not been

well-understood, the low incidence of

structural failures strongly indicates that the

system has been at least adequate and even

overbuilt. We have been developing

methods to predict the structural

characteristics of floor, wall, and roof

systems individually. We also endeavor to describe how these components interact

when assembled into a structure.

3.4.1 Floors

As a result of extensive research during the

past decade, the structural performance of

light-frame floors under uniform and

concentrated loads is well understood. A

comprehensive computer-based analysis

program developed by researchers at Colorado State University (CSU) provides a model of the structural interaction of the

floor joist, sheathing, and connectors. In a cooperative research project, CSU and FPL

used this model to predict the performance

characteristic of floors (McCutcheon et al., 1981; Vanderbilt et al., 1983). Research at FPL also showed that a different type of

model can closely predict the deflection

performance of floors and this model can be

464
easily adapted to microcomputers

(McCutcheon, 1984). The modeled

influence of 0.75-in. (19-mm) floor

sheathing on assembly deflection as compared to 2 by 8 (38 by 191 mm) joists

alone is illustrated in Fig. 10. The joists have

high variability in their stiffnesses. As the

sheathing bending modulus increases, the

variation among joist deflections decreases.

Note that much of the variability is

eliminated by using a sheathing with a modulus of elasticity of 1 x 10^ Ib/in^

(6.89 kPa), which is a realistic value for

common floor sheathing materials. Further

increases in sheathing stiffness have only a small effect. This example assumes no

interaction between the joists and sheathing

in the direction of the joist span. Additional

work has been aimed at predicting the

ultimate strength of floors (Wheat and

Moody, 1984).

Most floor analysis has been aimed at predicting performance under stationary or

static loads. However, some research

indicates that the dynamic or impact

performance of floors may often be the

critical factor in consumer acceptance.

Although some work in this area will soon be completed in Canada, the dynamic

performance of floors needs to be better

understood—particularly as new products

capable of spans beyond the traditional 14-

to 16-ft (4.3- to 4.9-m) lengths become more

widely available.

3.4.2 Walls

Walls perform several distinct structural

functions that require different types of

analysis. In a racking or shear mode, walls serve to keep the structure square and

prevent cascading failure of other walls

("dominoing"). Analytical methods to

predict both the strength and stiffness

performance of walls under shear loading

take into consideration the nailing between

sheathing and studs that provide shear

rigidity (McCutcheon, 1985; Tuomi and

McCutcheon, 1977). Recently, the

significant contribution of gypsum board or "drywall" to the overall shear rigidity was

quantified (Wolfe, 1983; Patton-Mallory et

al., 1984). Racking loads of 0.5-in.

(12.7-mm) gypsum-wallboard single-faced

stud walls consistently are 1,100 lb (454 kg)

at 0.5-in. (12.7-mm) deflection. Predicting

and validating the combined effects of walls

sheathed on both sides with dissimilar

materials are proving useful to designers and

code-making agencies (Patton-Mallory and

McCutcheon, 1987).

One function of walls is to transmit roof

and/or second-story loads to the foundation

through axial loads in the studs. A second

function is to resist the bending loads acting

perpendicular to the surface. These two

functions are satisfied by their combination

in an analysis procedure developed by

Oregon State University (OSU), which

requires as input the properties of interior

sheathings, exterior sheathings, studs, and

fasteners to predict the strength and stiffness

performance of light-frame wall systems.

Cooperative work between OSU and FPL

has yielded estimates of the effects of the

major variables on wall performance

(Polensekand Gromala, 1984).

Investigations of wall behavior, similar to research on floors, have primarily

considered only static loads. However,

dynamic loads produced by earthquake or wind gusts control wall performance in

many parts of the United States (Soltis et al., 1981). Thus, new work is underway in the area of dynamic performance of walls under

shear loading (e.g., Polensek, 1988).

3.4.3 Roof

The excellent performance of the residential

pitched truss roof system confirms that this

system adequately resists structural loads.

These metal-plate-connected trusses are the

most highly engineered component of the

wood framing. However, design does not

465
take into account all the factors that

influence overall structural performance of

roof systems. An extensive research

program at FPL has been underway to better

characterize the performance of light-frame

roof systems. The objective of the research

is to develop a model, similar to the models

used in floor and wall research, to predict

the load distribution and deflection

performance of roof systems (Moody, 1984).

Research on truss modeling has been

updated with a modern analysis method for

evaluating truss designs (Suddarth and

Wolfe, 1984). Work is continuing to

improve the methods of predicting the

performance of trusses with metal plate

connectors. Work reported by Wolfe et al. (1988) has substantially contributed to our understanding of load distribution and

composite action in plywood roof sheathed

metal-plate truss roof assemblies. The

deflection prediction of load-carrying

capacity and effect of repetitive members

(trusses) on 24-in. (610-mm) centers were

evaluated from five full-scale light-frame

roof assemblies characteristics and

compared to model predictions.

The actual deflection of the assembly is compared with that of loaded pitched trusses (3:12) alone in Fig. 11. This dramatically

illustrates the benefit to stiffness of the

assembly provided by the composite

plywood sheathing-truss assembly compared

to the trusses alone.

The distribution of load between adjacent

trusses with load applied to one truss in an assembly (truss load influence matrix) is shown in Fig. 12. The National Design

Specification currently sets allowed

repetitive member increases for lumber joist

design stress at 15 percent. This figure

indicates that for load distribution from a

stiff truss (truss 5) to a less stiff truss (truss

6), the repetitive load-sharmg factor is 1 1/45

or 24 percent—substantially greater than the

15 percent now allowed for floor or roof

joists! More detailed analyses and reports

are in preparation. We are counting on these

results to be used in the development of

methodology for reliability-based design of

truss roof assemblies.

3.4.4 Full Structure

To better understand the overall structural

performance of the entire light-frame

structure, investigators at FPL evaluated the

structural performance of a conventional

full-scale house during progressive stages of

construction (Tuomi and McCutcheon,

1974). Wind loads and snow loads were

simulated in a large structural test frame

(Fig. 13). In wind-load simulation, plywood

sheathing on the walls provided more than

adequate racking resistance, probably far m

excess of normal requirements. In fact, we

were unable to really test the walls because

the failures in tests first occurred in the

connection between the loaded wall and the

floor, and later at the connection to the

foundation.

Commercially fabricated metal-plateconnected trusses on 24-in. (610-mm)

centers were used for the roof system. Tests

showed no evidence of distress in the roof

system up to 2.5 times the design load. At

this point, one member of one roof truss

failed, but no collapse occurred. The trusses

failed at more than three times the design

load; had this been a snow load, the roof

would have collapsed.

Of special interest is an assessment of three

roof-to-wall plate connections in providing

uplift resistance under extreme winds

(Conner et al., 1987). The three nailed or nailed/lag-screw connections are shown in

Fig. 14. The uplift withdrawal capacities of

these connections were compared with

probability distributions for occurrence

likelihood of extreme winds. Figure 15

shows that 90 percent of severest wind

speeds (tornadoes) fall below 157 mph, and

a Type-H connection can be expected to

466
withstand from 110 to 180 mph. Type M

appears sufficient for most windspeeds from

70 to 120 mph. Less effective is Type L.

Thus, a reliable design for fasteners to

withstand expected winds requires a full

reliability-based analysis of fastener

variability or a function of wind load. Also

necessary is an assessment of duration of

peak wind speeds to compensate for

applying the duration of load factor to the

connections.

3.5 Improved Lumber Properties

and Design Procedures

Information is available for selecting

construction lumber for various applications

based on species and characteristics of

wood. However, the assignment of strength

properties covers a range of values, so

selection is based on the lower end of the

range. This often results in over-design for

much of the material. Research has been

conducted to determine the actual strength

and stiffness of material as used in the field

and how the duration of time loads applied

affect the material.

3.5.1 Field Testing

For years, tests of small, clear wood

specimens formed the basis for developing

engineering stresses for lumber. However,

evaluations in the 1970s indicated that this

method is inadequate when lumber is used in

highly engineered applications. Thus, a very

extensive field testing program, called the

"In-Grade Testing Program," was conducted

in cooperation with representatives of the

lumber industry during much of the 1980s

(Green, 1983a,b). The program was

completed in 1988. The major result of this

extensive testing program is currently

leading to new standards for assessing

structural lumber properties. This will

eventually lead to increased quality control

and more efficient utilization of lumber

when applied during the 1990s.

3.5.2 Duration of Load Adjustment Factor

Historically, a duration of load (DOL)

adjustment factor has been used in design to account for a strength reduction of wood

members after an elapsed time under load

(Gerhards, 1977). The DOL adjustment

factor (Fig. 16) is one of the primary factors

used to determme allowable design stresses from clear wood strengths, and it has been

obtained from bending tests on small clear

specimens of only one species. The DOL

adjustment factor has been extrapolated to

all stress conditions, such as bending,

compression, and tension, as well as to all grades, sizes, and species of structural

lumber containing a wide range of strengthreducing characteristics. This method of

extrapolation is being reassessed. The FPL is conducting an extensive research program in

cooperation with the Forintek Corporation in

Canada to establish DOL relations for

lumber that include the effects of stress condition, species, grade, and size.

Results of the DOL study and of the In- Grade Testing Program will allow

comprehensive redefinition of the

engineering properties of lumber. The

objectives of these study areas are to promote structural designs that use wood

efficiently and to form a data-base for future

reliability-based design codes.

3.6 Unified Behavior Model for Fasteners

Characterizing the strength of nailed and

bolted joints, as previously noted, was

among the early investigations conducted at FPL. T.R.C. Wilson initiated nailed joint

behavior and design work in 1917. George

W. Trayer was responsible for the early

study on steel bolts in 1927. These and

subsequent studies up to 1980, both at FPL

and universities, were the basis for much of our understanding of mechanical joint

behavior and for current design procedures

and properties in the National Design

Specifications. During the last 50 years, a

467
multitude of studies conducted globally have

investigated one or more of the properties

that affect connection behavior. Direct

comparisons have been difficult because

studies have usually differed in measuring

more than one connection property or have

been unique to the country conducting the

tests. In addition, recent failures of bolted

joints have raised doubts about our basic

understanding of their behavior and the

design criteria used.

A general "yield" theory for estimating the

ultimate lateral load of nailed or bolted joint

was developed by Johansen (1941) and

confirmed by experiments in Europe (Aune

and Patton-Mallory, 1986). The yield theory assumes the bearing capacity of a bolted or nailed connection is attained when either

(1) the compressive strength of the wood

beneath the bolt or nail shank is exceeded or

(2) one or more yielding plastic "hinges"

develop in the shank. A typical joint yield

load curve as a function of length of bolt in

the center member in a three-member joint is illustrated in Fig. 17. This curve is substantially more conservative for large

bolts but more liberal for small bolts, as seen

in Fig. 18.

The codes of several European countries

(CIB 1983) have adopted allowable loadcarrying capacities for bolted and nailed

connections based upon this model. A series

of studies have been conducted, and more

are well underway, to validate yield model

theory to design of joints typical in North

America (e.g., Aune and Patton-Mallory,

1986; Soltis et al., 1986). Results to date

continue to support the flexibility of the

method and eventual endorsement of the

procedure for assigning bolted and nailed

joint properties and design criteria in the

National Design Specification. We are grateful to European researchers for

publicizing the technique and showing its attractive practical attributes.

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D.R., eds. Proceedings, Recent advances in

spruce-fir utilization technology. Soc. Am.

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conference "From Stump Thru Mill—Recent

Advances in Spruce-Fir Utilization

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International Conference of Building

Officials, Whittier, CA.

Johansen, K.W. (1941). "Forsog med

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bygningstekvik.

Lie, T.T. (1977). "A method for assessing

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and columns." Can. J. Civ. Eng. 4: 161-169.

McCutcheon, W.J. (1984). "Deflection of

uniformly loaded floors: A beam-spring

analog." Res. Pap. FPL 449, U.S.

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McCutcheon, W.J. (1985). "Racking

deformation in wood shear walls." J. Struct. Eng. 111(2): 257-269.

McCutcheon, W.J., Vanderbilt, M.D.,

Goodman, J.R., and Criswell, M.E. (1981).

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variability on floor stiffness." Res. Pap. FPL

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Madison, WI.

Moody, R.C. (1984). "Improved analysis

procedures for roof systems." Wood Truss

Council of America. Woodwords. 2(8). 2 p.

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(1985). "Sheathing properties and

component performance in light-frame

structures." In: Structural Wood

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pp. 93-99.

Moody, R.C, and Sherwood, G.E. (1986).

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167-173.

NAHBF (1983). "Truss-framed

construction—A manual of basic practice."

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Research Foundation, Rockville, MD.

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of one-hour fire-resistive exposed wood

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(1987). "Predicting racking performance of

walls sheathed on both sides." Forest Prod.

J. 37(9): 27-32.

Patton-Mallory, M., Gutkowski, R., and

Soltis, L. (1984). "Racking performance of

light-framed walls sheathed on two sides."

Res. Pap. FPL 448, U.S. Department of

Agriculture, Forest Service, Forest Products

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Polensek, A. (1988). "Effects of testing

variables on damping and stiffness of nailed

wood-to-sheathing joints." J. Testing Eval,

Sept., American Society for Testing and

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Polensek, A., and Gromala, D.S. (1984).

"Probability distributions for wood walls." J. Struct. Eng. 11(3): 619-636.

River, B.H., and Gillespie, R.H. (1980).

"Adhesives in building construction—past,

present, and future." In: Proceedings.

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VA, pp. 31-42.

Schaefer, E.M., and Vanderbilt, M.D.

(1983). "Comprehensive analysis

methodology for wood floors." J. Struct.

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Schaffer, E.L., and Woeste, F.E. (1981).

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unprotected floor truss." In: Proceedings,

Metal Plate Wood Truss Conference, Forest

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Schaffer, E.L., Marx, CM., Bender, D.A.,

and Woeste, F.E. (1986). "Strength

validation and fire endurance of gluedlaminated timber beams." Res. Pap. FPL

467, U.S. Department of Agriculture, Forest

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Madison, WI.

Senft, J.F., Bendtsen, B.A., and Galligan,

W.L. (1985). "Weak wood—fast grown tree make problem lumber." J. Forestry, August,

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high thermal performance walls —hot,

humid summer climate." Res. Pap. FPL 455,

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(1981). "Seismic performance of low-rise

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Soltis, L., Hubbard, F.K., and Wilkinson,

T.L. (1986). "Bearing strength of bolted

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construction." NAHB Research Foundation,

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"Controlling moisture in houses." Solar Age,

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Thompson, E.G., Vanderbilt, M.D., and

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of wood under bolts." Tech. Bull. 332, U.S.

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Tuomi, R.L., and McCutcheon, W.J. (1977).

"Predicting racking strength of light frame

walls." ASCE J. Struct. Div. 104(ST7):

1131-1140.

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S.M. (1988). "Performance tests of lightframe roof assemblies." In: Proceeding,

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HGURE CAPTIONS

Figure 1. Construction of the truss-framed

system (TFS) enables a small crew of

laborers to frame an average house in only a few hours.

Figure 2. Unitized truss frames are

fabricated in a plant to insure maximum

quality control and material efficiency.

Figure 3. A simple TFS home after

completion

Figure 4. Floor-truss design subjected to test conditions of ASTM E 119. Upper chord

was loaded with tanks simulating a uniform

load of 55.1 lb/ft2 (269 kg/m^) that resulted

in a combmed live and dead load of 60 Ib/ft^

(293 kg/m2). 1 ft = 0.3048 m. 1 in. = 25.4 mm.

Figure 5. Relationship of applied load to

time-to-failure for floor-truss assembly of

Fig. 4.

Figure 6. Illustration of charring and

effective residual sections for wood beams

and columns.

Figure 7. Coated plywood sheathed stud

wall section for fire-resistance test.

Figure 8. Example of intumescent coating

for fire-resistance enhancement before and

after fire exposure.

Figure 9. Heat release rate (HRR) levels

from fire-retardant-treated wood-gypsum

wallboard assemblies. 1 BTU/(ft2 • min) = 189 W/m2.

Figure 10. Sheathing acts as a beam

spanning across floor joists and reduces the

variability in floor deflection caused by

variation in the stiffness of the floor joists.

The floor deflection under uniform load is shown for four different sheathing

stiffnesses. Note that the realistic sheathing

stiffness of 1 x 10^ Ib/in^ (6.89 kPa) greatly

reduces the variability.

Figure 11. Assembly influence on truss deflection for Group I 3:12 assembly.

Trusses individually loaded to their design

load value outside the assembly show a

larger average deflection as well as greater

variability in deflection than when tested as part of an assembly loaded to the same

design load value.

Figure 12. Truss load influence matrix. As

each truss was loaded in the assembly, loads were distributed to other truss reactions as

indicated. The number in each cell is the

percentage of the total applied load

measured at the reactions of the truss noted on the truss reaction axis. The height of the

graphic in each cell is proportional to the

load carried by that truss when only the

"loaded truss" is loaded directly.

472
Figure 13. Structural loading frame and test house with partially complete roof.

Figure 14. Joint configurations: Connection

details for low, medium, and high strength

joints.

Figure 15. Velocity pressure as a function of

wind speed with three joint capacities

superimposed. 1 mph = 0.447 m/s; 1 psf

(lb/ft2) = 0.042 kg/m2.

Figure 16. The "Madison" duration of load

adjustment factor for design stresses.

Figure 17. Symbolic normalized joint yield

load compared to 1/d ratio for three-member

joint with side members one-half thickness

of center member and equal bolt-bearing

strength.

Figure 18. Comparison of yield theory and

Trayer's (1932) empirical results for (a)

parallel-to-grain and (b) perpendicular-tograin loading. Also included are Trayer's

experimental results for Douglas-fir main

member with steel side plates ().

APPENDIX

Information About Publications Available

from the USDA Forest Products

Laboratory

The Forest Products Laboratory publishes a great number of reports and articles on

various aspects of wood utilization and

research. To facilitate acquisition, our

reports are classified into publication lists. Periodic releases on new publications are provided during the year. Subject areas that

may be of special interest are as follows:

Biodegradation and preservation

of wood

Economics

Energy from wood

Finishing (exterior only)

Fire performance of wood

Particleboard and panel products

Veneer and plywood

Architects, builders, engineers,

and lumber retailers

Detailed instructions for ordering

publications in these areas are given in the

publication list for each area. If you wish to be placed on our mailing list for one or more

of the publication lists, please call or write

Current Information Office

Forest Products Laboratory

USDA Forest Service

One Gifford Pinchot Drive

Madison, WI 53705-2398

(608) 264-5637

473
Figure 1. Construction of the truss-framed system (TFS) enables a small crew of laborers to frame an average house in only a few hours.

474
Figure 2. Unitized truss frames are fabricated in a plant to insure maximum quality control and

material efficiency.

475
Figure 3. A simple TFS home after completion

476
17'-5" Overall length of floor truss

3/4" Plywood

On

'»

Figure 4. Floor-truss design subjected to test conditions of ASTM E 119. Upper chord was

loaded with tanks simulating a uniform load of 55.1 Ib/ft^ (269 kg/m^) that resulted in a combined live and dead load of 60 Ib/ft^ (293 kg/m^). 1 ft = 0.3048 m. 1 in. = 25.4 mm.

477
0 4 8 12 16 20

TIME TO FAILURE (MIN)

M 148 530

Figure 5. Relationship of applied load to time-to-failure for floor-truss assembly of Fig. 4.

478
Figure 6. Illustration of charring and

effective residual sections for wood beams

and columns.

Figure 7. Coated plywood sheathed stud

wall section for fire-resistance test.

479
Figure 8. Example of intumescent coating for fire-resistance enhancement before and after fire

exposure.

480
Time (min)

ML85 5585

Figure 9. Heat release rate (HRR) levels from fire-retardant-treated wood-gypsum wallboard

assemblies. 1 BTU/(ft2 • min) = 189 W/m2.

481
Joist number

Figure 10. Sheathing acts as a beam spanning across floor joists and reduces the variabihty in

floor deflection caused by variation in the stiffness of the floor joists. The floor deflection under

uniform load is shown for four different sheathing stiffnesses. Note that the realistic sheathing

stiffness of 1 x 10^ Ib/in^ (6.89 kPa) greatly reduces the variability.

482
ASSEMBLY INFLUENCE ON TRUSS DEFLECTION

3:12 ASSEMBLY—NO GABLE END

0.9 T

0.1 H 1 1 1 1 1 1 1

1 2 3 4 5 6 7 8 9

TRUSS

Figure 11. Assembly influence on truss deflection for Group I 3:12 assembly. Trusses

individually loaded to their design load value outside the assembly show a larger average

deflection as well as greater variability in deflection than when tested as part of an assembly

loaded to the same design load value.

483
Figure 12. Truss load influence matrix. As each truss was loaded in the assembly, loads were

distributed to other truss reactions as indicated. The number in each cell is the percentage of the

total applied load measured at the reactions of the truss noted on the truss reaction axis. The

height of the graphic in each cell is proportional to the load carried by that truss when only the

"

484
Figure 13. Structural loading frame and test house with partially complete roof.

485
Rafter

3-l6d box nails

Double 2x4

top plate

Conventional nailed

Rafter

2- l6d serrated nails

3- l6d box nails

Double 2x4

top plate

Extra nailed

8in - lag bolt

Rafter

I6d box nails

Double 2x4

top plate

8-inch lag bolt

ML85 5296

Figure 14. Joint configurations: Connection details for low, medium, and high strength joints.

486
Figure 15. Velocity pressure as a function of wind speed with three joint capacities

superimposed. 1 mph = 0.447 m/s; 1 psf (Ib/ft^) = 0.042 kg

487
Figure 16. The "Madison" duration of load adjustment factor for design stresses.

488
Mode I

failure

Mode 2

failure

Mode 3

failure

Figure 17. Symbolic normalized joint yield load compared to 1/d ratio for three-member joint

with side members one-half thickness of center member and equal bolt-bearing strength.

489
80i

Figure 18. Comparison of yield theory and Trayer's (1932) empirical results for (a) parallel-tograin and (b) perpendicular-to-grain loading. Also included are Trayer's experimental results for

Douglas-fir main member with steel side plates ( ).

490
The Impacts of Siting Lifelines in Close Proximity Upon Their Earthquake Vulnerability

by

William Bivens^ and Phillip Lowe''

ABSTRACT

The Federal Emergency Management Agency

(FEMA) has initiated a program to examine the

potential increases in lifeline earthquake vulnerability

due to their close placement or collocation. To date, the study has focused on developing a practical

engineering approach that builds upon available data and analysis techniques for evaluating individual

lifelines or lifeline components. The methods have

been applied in California at the Cajon Pass, which

is located north and east of Los Angeles. The site lifelines analyzed included: highways, railroads, and

their bridges; natural gas bulk transmission pipelines;

refined petroleum product transmission pipelines;

high voltage electric power transmission lines; and

fiber optic telephone communication lines. The San

Andreas fault crosses at approximately the midpoint

of the Cajon Pass canyon. During the study,

improved methods to analyze the vulnerability of

buried pipelines and bridges were developed. The

study found that a fragmented regulatory control system, environmental concerns about disturbing the surface conditions, and the economics of right-of-way

acquisition and lifeline construction led to over 100

separate lifeline collocations within the study area. Depending upon the lifeline, collocation impacts were

estimated to increase the delay to restore lifeline temporary service from 20 to 60% compared to the

restoration time if no collocation were assumed to

exist. FEMA now intends to extend the study to examine other locations in the U.S. and to refine the analysis methods.

KEYWORDS: collocation, earthquake, damage,

impacts, lifelines

1. INTRODUCTION

In the U.S., lifelines are presently being sited in

"utility or transportation corridors" to reduce their right-of-way environmental, aesthetic, and cost impacts on the communities that rely upon them.

The individual lifelines are usually designed,

constructed, and modified throughout their service

life. This results in different standards and siting

criteria being applied to segments of the same

lifeline, and also to different standards or siting

criteria being applied to the separate lifelines systems

within a single corridor. Presently, the siting review

usually does not consider the impact of proximity or

collocation of the lifelines on their individual risk or

vulnerability to natural or manmade hazards or

disasters.

There have been cases when some lifeline collocations have increased the levels of damage

experienced during an accident or an earthquake.

For example, water line ruptures during earthquakes

have led to washouts which have caused foundation

damage to nearby facilities. A transportation lifeline failure has led to the subsequent failure of a collocated fuel pipeline, and the resulting fire caused

considerable property damage and loss of life. Loss

of electric power has restricted, and sometimes

failed, the ability to provide water and sewer services or emergency fire fighting capabilities.

In response to these types of situations, the Federal

Emergency Management Agency (FEMA) is examining the use of such corridors, and FEMA has

initiated a program to study the impact of siting

multiple lifeline systems in confined and at-risk areas.

The overall FEMA project goals are to develop

managerial tools that can be used to increase the understanding of the lifeline systems' vulnerabilities and to help identify potential mitigation approaches

that could be used to reduce those vulnerabilities. Another program goal is to identify methods to enhance the transfer of the resulting information to

lifeline system providers, designers, builders,

managers, operators, users, and regulators.

In the initial project of this program', a screening tool was developed to identify the critical lifeline

'Federal Emergency Management Agency,

Washington, DC 20472

''INTECH Inc, 11316 Rouen Dr.,

Potomac, MC 20854

491
collocation locations and the conditions that makes

them critical. The probable time required to restore

the lifeline service after an earthquake was used to measure the impact of collocation compared to an assumed no collocation situation.

The analysis methods were based on using existing data bases and lifeline vulnerability to earthquake

evaluation methods so that a practical, engineering

approach could be developed. During the study,

improved methods to analyze buried lifelines and to account for soil liquefaction, as well as improved

methods to analyze potential bridge vulnerability to an earthquake event were developed. The resulting

analysis tools have been successfully applied to the

lifelines sited at the Cajon Pass, California. These

lifelines included: highways, railroads, and their bridges; natural gas bulk transmission pipelines;

refined petroleum product transmission pipelines;

high voltage electric power transmission lines; and

fiber optic telephone communication lines.

2. CONCLUSIONS

A very useful screening tool has been developed.

The tool can be used to identify the critical lifeline collocation locations and the conditions that make

them critical. It can be used to identify areas of

technical uncertainty and poor siting practices, and its use can identify important research and development

opportunities.

At the Cajon Pass it was found that:

• ground movement due to landslides and

liquefaction are more important than ground

shaking for estimating the vulnerability of most

lifelmes. Bridges are the exception. For them,

ground shaking can be a critical part of the estimation of their vulnerability;

• fiiel pipeline failures had the greatest impact on

other nearby lifelines during the immediate

recovery period after an earthquake;

• current siting practices for fiber optic cables indicates that more severe telephone

communication failures than have been

experienced in past earthquakes can be anticipated

in fiiture earthquakes when fiber optic systems

provide the basic telephone service (compared to the current hard wire systems);

• lifeline siting practices do not fully consider the

potential impact that other nearby lifelines could

have on the lifeline being evaluated;

• commimication, electric power, and fuel pipeline

lifelines can generally be analyzed as a set of

discrete collocation points. The restoration of

service at one location is generally not impacted

by the restoration work at other collocation points,

because the repair equipment and personnel can

access the work site from a number of alternative paths. Thus, the critical restoration of service

time is the longest time estimated for any one

collocation point in those lifelines;

• transportation lifeline restoration of service is highly dependent on sequentially repairing the

lifeline damage along the lifeline itself, as heavy

equipment access to the next damage location often requires use of the transportation system

itself;

• some collocation can produce benefits, as the nearby parallel location of a similar lifeline can provide the detour or access route to the damaged

sections of the other lifeline. However,

intersecting lifelines generally result in the failure of one lifeline increasing the risk of failure of the

lifelines it crosses; and

• additional applications of the analysis methods are needed to assure that the results obtained are not

in some way unique to the Cajon Pass and also to see if additional improvements in the analysis

methods can be developed.

3. THE ANALYSIS METHODS

The analysis used a four Step approach: 1, data

acquisition; 2, calculation of individual lifeline vulnerability; 3, calculation of collocation damage;

and 4, interpretation of the collocation impacts on the individual lifeline systems.

./ 3.1 Data Acquisition

In this step all the required information on the lifeline routes, methods of construction, materials of

construction, location of the collocation points, collection of the topographic and geologic data, identification of probable landslide and liquefaction

locations, performing an earthquake analysis to determine the location of the regions of constant Modified Mercalli Indices (MMI), and the plotting of

the data with respect to the lifeline locations are performed. In the first application, the data was

492
input into a PC computer and analyzed and plotted

using commercially available computer aided design

(CAD) programs. The MMI isobars were developed

using the methods of Evemden^. This method was

used because it has been verified by comparison with

historical earthquakes, it incorporates the local sediment conditions into the model and the sediment

conditions are generally available in the U.S.

Geological Survey geologic data base, it is easy to use, and it is readily available to other researchers. When this was performed for the Cajon Pass, 103

separate collocation locations that required further analysis were identified in the 220 square kilometer

study area (approximately 90% of the collocation

locations were found in a 65 square kilometer region

along the foot of the Cajon canyon). A number of

the lifelines were located in the expected fault zone of

the San Andreas fault.

3.2. Calculation of Individual Lifeline Vulnerability

Damage matrices^ were used to estimate the earthquake damage to each individual lifeline at each

collocation point. The analysis was performed as if the lifeline were an isolated lifeline at that location. The damage matrices related the percent damage and

the probability of that damage to the MMI index.

The analysis also accounted for earthquake-induced

landslides, and to a lessor extent earthquake-induced

soil liquefaction. In the method applied at the Cajon

Pass, improved damage matrices were developed for determining liquefaction impacts on buried lifelines'*. In addition, an improved screening method to estimate the shaking damage to bridges was

developed. It was based on adapting bridge retrofit requirements analysis methods developed by the California Department of Transportation*.

The important items that are calculated for each

lifeline at each collocation point are the estimated

damage state (or percent damage) and the estimated

time to restore the lifeline to the needed service level. Both of these parameters can be calculated using the matrices and tables of Rojahn-*.

3.3. Calculation of Collocation Damage

This analysis step builds upon the results obtained

from the two previous Steps. A lifeline zone of

physical influence was defined to describe the physical region in which one lifeline could impact

other nearby lifelines. Then, based on the individual

lifeline damages calculated in Step 3.2, and the zone

of influence, a collocation damage scenario was

defined. The damage matrices of step S.2 and the

restoration of lifeline service tables of Step 3.2 were

used to calculate the percent of damage and lifeline restoration of service time for each lifeline for the assumed collocation damage scenario.

One of the easier direct impacts to hypothesize is that the collocation conditions will lead to an increase in the damage state of one or both of the collocated

lifelines (if there are more than two collocated

lifelines this applies to all of them). It is easy to understand the damage state, as it relates to a physical condition. Because the individual lifeline damage states assuming no collocation are known,

those values can be used to help understand how the

lifeline could impact another nearby lifeline. If, for example, light damage of a pipeline had been

calculated, it would be expected to cause no direct change in the damage state of a nearby bridge.

However, if the bridge had been estimated to collapse, it would be reasonable to estimate that within the bridge's zone of influence it would lead to failure of the pipeline (this example also illustrates that the impacts are not necessarily reciprocal).

As another example of how collocation impacts on damage state can be estimated, consider the condition

of a pipeline and a fiber optic conduit hung from a bridge. The earthquake vibration may not be enough

to cause serious damage to the bridge or to the pipeline or conduit if they were not collocated with

each other. However, the vibrations may cause the anchors holding the heavy pipeline to the bridge to

fail. As the pipeline sags (but does not fail) it could

fall onto the lower conduit, causing it to fail. The

collocation damage state hypothesis would then be: no impact on the bridge; a small increase in damage

state of the pipeline to account for the work required

to rehang the pipeline; and catastrophic failure of the

fiber optic conduit.

Special attention should be given to the collocation of

fuel carrying lifelines with other lifelines that have

the ability to provide an ignition source. The

resulting fire and/or explosion could lead to significant collocation damage. Similarly, broken

pipelines which eject fluids could lead to foundation

erosion problems that would result in increased

damage to nearby lifelines.

493
3.4. Interpreting the Collocation Impacts on

the Individual Lifeline Systems

It was established that the most realistic measure of

the impact of collocation was the "most probable

incremental change in the restoration of service time"

for each lifeline. This is defined as the product of

the probability of the collocation damage occurring

times the incremental increase in restoration of

service time. The incremental increase in restoration of service time is the restoration time calculated for the collocation damage state minus the restoration time calculated for the lifeline at the same location but assuming no collocation existed.

There are several ways to estimate the change in the probability that damage will occur, none are exact and there are no statistics available from the literature on earthquakes. However, there are some insights available to guide the analysts.

If the probabilities for damage to two lifelines, assuining no collocation conditions, are PI and P2,

they represent an upper bound on the probability that a collocation damage would occur. For example, if the probability that lifeline 1 would fail is PI, and it

is known that if lifeline 1 fails it will cause, with a 100% probability, damage to lifeline 2, then the probability that lifeline 2 receives collocation damage

is also PI (e.g., PI x 100%). Similarly, the upper

bound on the probability that lifeline 2 has damaged

lifeline 1 is P2.

As a practical matter, the collocation damage likely will be less, since there is seldom a 100% chance that the collocation damage scenario will occur. A useful measure of the probability that the collocation event has occurred is the product of the two probabilities

that the single independent events that were used to develop the collocation scenario have occurred (the

independent events are the estimate of the damage

state of each lifeline assuming there was no

collocation). In the present case, that is found by

multiplying PI x P2. The product can be interpreted as follows. It represents the increase in probability

that the two independent lifeline damages will occur

during the same initiating event. If both events must

occur before the collocation damage scenario can take

place, then it is a measure of the probability of the collocation damage scenario.

The actual probability that the collocation event will occur should be a number between the numerical

limits of PI and (PI x P2) for having lifeline 1 cause

additional damage to lifeline 2, and P2 and (P2 x PI)

for having lifeline 2 cause additional damage to

lifeline 1. It is recommended that for calculational purposes, the product PI x P2 be used to characterize

the hypothesized collocation damage scenario.

4. REFERENCES

1. Lowe, P., et. al., "Collocation Impacts on the Vulnerability of Lifelines During Earthquakes

With Application to the Cajon Pass, California,"

FEMA 226, February, 1992.

2. Evemden, J., et. al., "Interpretation of Seismic

Intensity Data", Bulletin of the Seismological

Society of America, V 63, 1973.

Evemden, J., et. al., "Seismic Intensities of

Earthquakes of Conterminous United States - Their Predictions and Interpretations", U.S.

Geological Survey Professional Paper 1223, 1983.

3. Rojahn, C. and Sharpe, R., "Earthquake Damage

Evaluation Data for California", ATC-13, Applied

Technology Council, 1985.

4. See reference 1.

Lowe, P., et. al., "A New Method to Relate Soil Liquefaction to Pipelines With Applications to Collocated Lifelines", Fourth US/Japan Workshop

on Earthquake Resistant Design of Lifeline

Facilities and Countermeasures Against Soil Liquefaction", Takai University, Honolulu, May

1992, also to be published as workshop

proceedings by the National Center for Earthquake

Engineering Research, NCEER.

5. Maroney, B. and Gates, J., "Seismic Risk

Identification and Prioritization in the CALTRANS Seismic Retrofit Program",

California Department of Transportation,

September 1990.

"CALTRANS Seismic Risk Algorithm for Bridge

Structures", SASA Division of Structures,

California Department of Transportation, June 30, 1990.

494
Nonlinear Research Needs for Concrete Gravity Dams

by

Robert L. Hall* and Wayne G. Johnson*

ABSTRACT

The Corps of Engineers' Engineer

Technical Letter (ETL) 1110-2-303

outlines a sequence of analyses of

concrete gravity dams subjected to

the maximum credible earthquake

(MCE). The first step is a twodimensional linear-elastic response

spectrum or time-history analysis

with five percent damping. This

analysis may include the effects of

hydrodynamic loads , foundation

flexibility, and absorption of the

reservoir bottom (Chopra, 1978).

For this step, five percent of

critical viscous damping is assumed

unless tensile stresses exceed

fifteen percent of f'^.. If the

tensile stresses exceed 0.15 X f'^,

adjustments are made in the damping

to account for some cracking of the

concrete. These different levels of

damping combined with different

concrete strengths are assumed to

produce a conservative and

appropriate procedure.

Nonlinear dynamic calculations were

performed to evaluate guidelines of

the ETL. The analyses revealed that

existing tools generally produce

reasonable results; however, the

studies demonstrated the need for

further development of nonlinear

analysis tools. Limitations

presently exist in the modeling and

assumptions of material properties

(Fenves, 1987), damping assumptions,

and effects of water cavitation and

intrusion into cracks (Dowling,

1987). To ensure the safety of

concrete dams subjected to strong

ground motions , further research is

needed on nonlinear analysis methods

and the corresponding parameters

that govern these complex geometric

and materially nonlinear models.

1. INTRODUCTION

The Corps presently supports the

program SDAM (Cole and Cheek, 1988)

and an acceleration- time-history

program EAGD-84 (Fenves and Chopra,

1984) for the analyses of concrete

gravity dams. These codes are supported through the numerical

maintenance modeling program. Both

codes assume linear-elastic material

response and can be used to perform

the seismic analyses according to

procedures outlined in ETL 1110-2-

303.

Mlakar (1986) evaluated the

presently accepted procedure by

performing a nonlinear analysis of

three different dams of different

heights subjected to two different

earthquake records . The earthquake

ground motions and the reservoirstructure dynamic interaction were modeled with the ADINA 84 finiteelement code. For the cases

investigated, the simpler analytic

models seem to be conservative for

evaluation of concrete gravity dams.

* US Army Engineer Waterways

Experiment Station, Vicksburg, MS

39180-6199

495
However, the elastic procedures

incorrectly located the region of

maximum cracking (Mlakar, 1986).

Fenves (1987) evaluated this ETL bycalculating the nonlinear response

of the Pine Flat Dam subjected to a scaled horizontal Taft ground motion

for five different load cases. These five load cases considered a full and empty reservoir and peak

acceleration varying from 0.18 g to

0.45 g. The results demonstrated

that cracks can form but remain

stable under low amplitude ground

motion; however, under large

amplitude ground motion, Fenves

found that the cracks propagate

across the entire cross - section of

the studied monolith.

2. DETAILED STUDIES

Mlakar' s research was funded by the

U.S. Corps of Engineers' Structural

Research Program in order to

evaluate the current ETL 1110-2-303.

This evaluation was done by

performing nonlinear analyses of

three nonoverflow- gravity dam cross

sections which were selected to

characterize the Corp's population

of concrete gravity dams. The N65W

and vertical components of the 1966

Parkfield, California earthquake

recorded at Temblor No. 2 Station

(CIT File Nos. B037-1 and B037-3) were used for each analysis. The

records from this earthquake were chosen because the peak ground

acceleration and frequency content

are representative of strong ground

motions . The nonlinear analyses

were performed using the general

purpose finite-element code ADINA.

The constitutive behavior of the

concrete was described by Bath and

Ramaswamy (1979). The hydrodynamic

loading of the reservoir was modeled

by adding concentrated nodal masses

on the upstream face corresponding

to the distribution described by

Chopra (1978). All foundations were assumed rigid.

The shortest section analyzed

represented the Richard B. Russell

Dam on the Savannah River. This

structure is 185 ft high and has a modulus of elasticity of three

million psi. The finite-element

grid contains 65 elements and 247

nodes providing 594 degrees of

freedom. The two-dimensional cross

section experienced no nonlinear

behavior when subjected to the

Parkfield ground motion. The ground

motions were then tripled in order

to determine the nonlinear response

of this gravity dam. Figure 1 shows

the cracked regions of the dam due

to tripled Parkfield accelerations.

The second gravity dam modeled has a height of 300 ft, but it is not

representative of any particular dam

and is labeled "Standard Dam." The

finite -element model for the dam

consists of 594 degrees of freedom.

The modulus of elasticity for the mass concrete was assumed to be

three million psi. The ADINA

analysis showed two cracked zones on

the downstream surface at the

elevation of slope change and at a slightly lower elevation (Figure 2)

.

These cracks progress to the

upstream surface. An examination of

the ETL procedure for this situation

reveals that no indication of

cracking above the base of the dam

would be realized.

The third dam analyzed was the

Dworshack Dam on the Clearwater

River in Idaho . This gravity dam is

638 ft high and it has an estimated

modulus of elasticity of five

million psi. This structure was chosen to represent the tallest dam

owned by the Corps of Engineers.

The finite -element model of this dam

has 80 elements and 363 nodes with

726 degrees of freedom. The

496
nonlinear analysis indicates a cracked zone which transect the cross section of the elevation of

change in downstream slope.

Cracking also initiates at the

upstream edge of the base and soon stabilizes (Figure 3) . At later

times the cracking propagates from

the downstream face at the elevation

of the slope change. The ETL

procedure indicates that cracking is

expected through the base of the

structure.

Fenves performed a nonlinear seismic

analysis of the tallest (400 ft)

nonoverflow monolith of the Pine

Flat Dam. The finite -element model

of the structure consists of 162

nodes comprising 136 quadrilateral

elements with a total of 315 degrees

of freedom. The finite element

model of the impounded reservoir

water extends upstream 1,200 ft from

the dam and consists of 224

constant-pressure elements. By

modeling the reservoir, the analysis

accounts for the interaction between

the reservoir and dam. Stiffness

proportional damping was used to

provide five percent critical

damping of the fundamental frequency

and a modulus of elasticity of 3.25

million psi was used. The S69E

component of the 1952 Taft ground

motion was taken as a horizontal

component acting in the upstreamdownstream direction. The peak

ground acceleration from these

ground motions was 0.18 g.

When the reservoir was empty, the

Taft ground motion had to be scaled

up by a factor of 2.5 to an acceleration of 0.45 g in order to

initiate cracking. However, when

the reservoir was assumed full (381

ft) , the nonlinear analysis

indicates cracking from unsealed

Taft ground motions . The cracking

occurs at the head of the dam where

a stress concentration exists

because of the assumed rigid

foundation. When ground motions

were scaled to 0.27 g, cracks are shown to propagate further along the

base. When ground motions were

- scaled to 0.36 g, cracking also occurs on the downstream slope where

the change in the slope occurs. The

extensive cracking led to a numerically unstable solution which

indicates that stability analyses

are needed. Figure 4 gives a summary of these results.

The ETL procedure was followed using

the computer program EAGD-84. Five

load cases were investigated for the

Pine Flat Dam. Table 1 summarizes

the five cases and results. For cases 1 , 3 , and 4 the maximum

tensile stresses were less than 638

psi (.15 f'c), so no cracking of the

monolith was presumed. Cases 2 and

5 were performed again assuming 7-

percent viscous damping which

equates to 14-percent hysteretic

damping (Table 2) . Case 2 resulted

in maximum tensile stress of 563 psi

at the downstream face which exceeds

0.10 X f'c and, according to the

ETL, a crack should be assumed at

two locations and sliding stability

analyses performed for the portion

of the dam above this plane. For

Case 5, the 0.10 X f'^ criteria was exceeded on the upstream and

downstream faces as well as at the

heel. The ETL criteria again

require stability analysis to be

performed.

These nonlinear analyses clearly

indicate the importance of the

nonlinear response of the Pine Flat

Dam to these ground motions. The

ETL criteria appeared to be

reasonably accurate for these five

analyses. However, the procedure

needs further investigation because

there is no theoretical basis for

assuming that tensile cracking

results in increased energy

497
dissipation which is implied by

greater damping ratios.

3. NONLINEAR RESEARCH NEEDS

The Panel on Earthquake Engineering

for Concrete Dams Committee on Earthquake Engineering, Division of

Hazard Mitigation, through the

National Research Council has

recently completed a publication

entitled, "Earthquake Engineering

for Concrete Dams: Design,

Performance, and Research Needs"

(NRC, 1990). This publication

presents details for needed research

in the nonlinear seismic response of

concrete dams. The following are

the items which should be addressed

in future research programs:

3.1 Material Testing of Mass

Concrete

Further testing of mass concrete

under dynamic loads is needed. It

is needed to determine tensile

cracking of the mass concrete under

multiaxial stress states which

represent the in-situ strain paths

in a concrete dam during a seismic

event. These tests must quantify

the effects of strain rates,

concrete mixtures, and aggregate

size. Concrete samples should be a mixture of cores from actual dams

and carefully prepared laboratory

samples

.

3.2 Development of Materials Models

for Concrete

Once the test data are available,

realistic numerical models for

tensile cracking under dynamic loads can be developed. These models must

allow for multiaxial stress states,

strain rate effects, shear transfer

by aggregate interlock and criteria

for tensile cracking and propagation

of cracks. The studies must include

smeared- crack approach, fracture

mechanics principles, crackingconsistent damping, and the

discrete -crack approach.

3.3 Modeling of Other Nonlinear

Mechanisms

Since the limiting tensile strength

of the concrete is across lift

surfaces, it is important to develop

construction joint models. These

models will redistribute the forces across a joint as the two surfaces

open and close. The degradation of

the concrete across the joints must

be modeled due to the damaging

number of loading cycles from an earthquake

.

3.4 Numerical Procedures for

Computing Nonlinear Response

The numerical material and joint

models can then be incorporated into

a finite -element program with time

integration for the equations of

motion or other discretization

methods for solving dynamic

nonlinear equations . The procedures

must also include interaction with

the impounded water, flexible

foundation rock, and the reservoir

bottom absorption. These procedures

will need to be refined to take

advantage of vector and parallel

processes in the latest computers.

3.5 Parameters and Detailed

Response Studies

With the development of accurate

numerical procedures, parametric

studies can be performed to

determine the sensitivity of the

nonlinear response to the parameter

describing the nonlinear models.

These studies would identify the

significance of tensile cracking and

joint opening with respect to either

failure or dynamic response.

Finally the following factors should

be determined: effects of ground

498
motion characteristics, water

compressibility, foundation

flexibility, reservoir bottom

absorption, and modeling issues.

3.6 Dynamic Testing of Dam Models

Testing of dam models is essential

for verifying nonlinear numerical

procedures. These tests may require

association with international

augmentation testing and use of

facilities such as a large

earthquake simulator recently

installed at the research laboratory

of the Ministry for Water

Conservancy and Hydroelectric Power

in Beijing, China.

3.7 Identification of Design

Criteria

Accurate design criteria based on numerical and experimental studies

could then be developed. Nonlinear

seismic analysis may not become a standard practice in design but will

certainly define the tensile

strength of the structure and the

post-cracking stability.

3.8 Investigation of EarthquakeResistant Design Measures

These nonlinear capabilities will

allow for the investigation of

innovative measures for increasing

the seismic safety of concrete

gravity dams . These tools could be

used to study effects in the

geometry of dams, jointing schemes,

and joint materials to dissipate

energy

.

4. CONCLUSION

Nonlinear capabilities are important

in determining the seismic stability

of concrete gravity dams. The

capability to develop a complete

program as described above is beyond

single capability of the Corps of

Engineers. However, the Corps can strategically utilize the research

being performed through the National

Science Foundation and other

organizations to answer these

complex issues and develop the

necessary tools and criteria.

Nonlinear analysis will provide the

necessary insight for the

development of reliable cost saving

earthquake resistant designs,

5 . ACKNOWLEDGMENT

Acknowledgment is made to Drs . Paul

Mlakar and Gregory L. Fenves who

conducted the analyses contained in

this paper. The paper was prepared

under the general supervision of

Messrs. Bryant Mather, Chief,

Structures Laboratory; and James T.

Ballard, Assistant Chief, Structures

Laboratory; and under the direct

supervision of Dr. Jimmy P. Balsara,

Chief, Structural Mechanics

Division.

COL Larry B. Fulton, EN, was the

Commander and Director of the U.S.

Army Engineer Waterways Experiment

Station during the preparation of

this paper. Dr. Robert W. Whalin

was Technical Director.

6 . REFERENCES

1. Bathe, K. J. 1979. "On ThreeDimensional Nonlinear Analysis

of Concrete Structures,"

Journal of Nuclear Engineering

and Design . Vol 52, pp 385-409.

2. Chopra, A. K. 1978.

"Earthquake Resistant Design of

Concrete Gravity Dams," Journal

of the Structural Division . American Society of Civil

Engineers, Vol 104, No. ST6 , pp

953-971.

3. Cole, R. A. and Cheek, J. B.

1985. "Seismic Analysis of

499
Gravity Dams , " Technical Report SL- Engineer Technical Letter No

86-44, US Army Engineer Waterways 1110-2-303, Washington, DC

Experiment Station, Vicksburg, MS.

4. Dowling, M. J. 1987.

"Nonlinear Seismic Analysis of

Arch Dams," Report No. EERL 87-

03, Pasadena, CA.

5. Fenves, G. 1987. "Earthquake

Induced Cracking in Concrete

Gravity Dams," American Society

of Civil Engineers.

6. Fenves, G. and Chopra, A. K.

1984. "EAGD-84, A Computer

Program for Earthquake Analysis

of Concrete Gravity Dams,"

Report No. UCB/EERC- 84/11

,

Earthquake Engineering Research

Center, University of

California, Berkeley, CA.

7. Fenves, G. and Chopra, A. K.

1984. "Earthquake Analysis and

Response of Concrete Gravity

Dams," Report No. UCB/EERC

84/10, Earthquake Engineering

Research Center, College of

Engineering, University of

California, Berkeley, CA.

8. Mlakar, P. F. 1986.

"Nonlinear Response of Concrete

Gravity Dams to Strong

Earthquake -Induced Ground

Motion," JAYCOR Report No.

J650-86-002/1335, for the US

Army Engineer Waterways

Experiment Station, under

Contract No. DACW39-85-M-4964.

9. National Research Council.

1990. Earthquake Engineering

for Concrete Dams: Design.

Performance . and Research

Needs . National Academy Press,

Washington, DC.

10 . US Army Corps of Engineers

,

"Earthquake Analysis and Design

of Concrete Gravity Dams,"

500
Table 1. Maximum tensile stress in Pine Flat Dam due to scaled horizontal

Taft ground motion from first linear dynamic analysis.

Max. Stress (psi)

Case Water (g) /if Up Down Heel

(1) (2) (3) (4) (5) (6) (7)

1 Empty 0.18 0.10 130 256 4

2 Empty 0.45 0.10 466 666 271

3 Full 0.18 0.10 217 250 361

4 Full 0.27 0.10 415 363 539

5 Full 0.36 0.10 510 580 718

Table 2. Maximiom tensile stresses in Pine Flat Dam due to scaled horizontal

Taft ground motion from second linear dynamic analysis.

Max

.

Stress (psi)

Case Water (g) Up Down Heel

(1) (2) (3) (4) (5) (6) (7)

1 Empty 0.18

2 Empty 0.45 0.14 383 563 198

3 Full 0.18

4 Full 0.27

5 Full 0.36 0.14 459 479 642

501
(a) t = 3.84 sec. (b) t = 3.92 sec.
(c) t = 4.48 sec. (d) t = 4.56 sec. Figure 2. Cracked zones of Standard Dam.

503
(a) t = 4.04 sec. (b) t = 4.16 sec.
(a) Empty reservoir, ag = 0.45 g (b) Full reservoir, a^ = 0.18 g

(c) Full reservoir, ag = 0.27 g (b) Full reservoir, a = 0.36 g

Figure 4. Tensile cracks in Pine Flat Dam due to Taft ground motion scaled to

peak ground acceleration ag.

505

Seismic Risk Assessment Methodology for Water Systems

by

Craig E. Taylor* and Stuart D. Werner**

ABSTRACT

In this paper, we discuss some of the

methodological highlights of the five-year

voluntary project resulting in the American

Society of Civil Engineers (ASCE) monograph

entitled Seismic Loss Estimates for a Hypothetical Water System . Highlights of this project include application of the monograph in progress to regional loss studies and discussions

of valuation methods, seismic attenuation

models, analysis of liquefaction-induced ground

failure potential, analytic models for estimating

losses from given input earthquake intensities,

and probabilistic modeling of earthquake activity

and of risks. Based on the results of this project, we make recommendations on future

steps to be undertaken in research, the

development of a manual, and the integration of

earthquake risk methods into system planning.

KEYWORDS:

1. INTRODUCTION

In 1986, the Seismic Risk and the Water and

Sewage Committees of the Technical Council on

Lifeline Earthquake Engineering (TCLEE) of

ASCE undertook a project to demonstrate how

direct earthquake loss estimates can be evaluated

for a water system. Two years after project

inception, a paper on the project progress was

presented at the Ninth World Conference on

Earthquake Engineering in Tokyo (See Taylor,

et al., 1988). Four years after project inception,

a paper was delivered on project progress at the

annual ASCE convention (see Taylor et al., 1990) . Since the state-of-the-art still requires

much improvement, this paper explained why

the monograph was not designed to be a manual:

After five years of effort involving about forty

participating engineers, the monograph entitled

Seismic Loss Estimates for a Hypothetical Water

System was completed (see C. Taylor, ed.,

1991) .

To develop primary loss estimates, the following

steps were undertaken:

• Exposure Definition . The characterization

of a hypothetical demonstration water

system, including component definition,

spatial definition and replacement cost estimation.

• Seismic Hazard Identification and

Definition . The characterization of seismic

hazards including earthquake sources,

regional attenuation patterns, local strong

ground motion amplification factors, and

local susceptibilities to permanent ground

failures.

• Component Loss Algorithm Development . The development of component loss algorithms (or, for other types of studies,

component reliability or fragility models),

and

• Seismic Risk/Loss Calculation . The

calculation of expected direct system losses

(or, for other types of studies, the

calculation of system reliability and/or

secondary losses) and of various

breakdowns on these losses (e.g., by

'Dames & Moore, 911 Wilshire Boulevard, Suite 700, Los Angeles, California 90017 USA

"Dames & Moore, 2101 Webster Street, Suite 300, Oakland, California 94612-3011

507
earthquake source, by component type, by

locations in high liquefaction susceptibility

zones).

The project implementation began with the

assignment of specific project steps to lead

engineers, with a group identified for assistance

and review. Accomplishment of these tasks was

given very wide latitude, with an emphasis on

statements of potential difficulties and how

potentially innovative approaches can be used to overcome these difficulties. Project coordination

made it possible for project steps to be taken so

that they fit into an overall framework~a

demonstration earthquake loss analysis and

resulting monograph being the ultimate goals.

In the remainder of this paper, we first outline some of the highlights of the project and then

discuss some future steps that might be taken to improve seismic risk applications to water

systems.

2. SELECTED PROJECT HIGHLIGHTS

The highlights of the project result from special

efforts made by voluntary participants in for

what was many a new context: an earthquake

risk analysis. Highlights include:

• The use of the draft monograph in progress

as material for regional earthquake loss studies

• The revelation that uniform valuation

methods would greatly assist in water

system planning as well as in exposure

definition for earthquake risk analysis

• The clarification of selection criteria for attenuation functions and of the need for

improved modeling of the probability of

liquefaction-induced ground failure

• Attempts to overcome the use of subjective

estimates of losses through analytic

modeling of filtration plants and pipelines,

respectively, and

• Use of probabilistic modeling of

earthquakes so that potential losses of

earthquake scenario selections are overcome.

After reviewing these highlights, we make

recommendations on future steps that might be

taken.

The aforementioned project highlights will be

discussed in turn. Those highlights discussed at length previously will be discussed here only

briefly.

2.1 The Benefit to Required Earthquake Loss

Studies

When the TCLEE project began, many of the

participants were not experienced in all steps

needed to perform an earthquake loss analysis

for a water system. However, as the project

evolved, many participants not only fulfilled their tasks successfully but also became better acquainted with the overall steps needed to undertake a seismic risk analysis. At least three

regional earthquake loss studies resulted from

the experience of project participants

demonstrate the diverse ways how the

collaborative TCLEE project has increased the awareness of practitioners of the benefits of the

state-of-the-practice methods illustrated in the

monograph. These include earthquake loss evaluations of the Seattle water system

(Ballantyne et al., 1990), the water system and

other utilities in the City of Everett, Washington

(Ballantyne et al., 1991), and of the water

system in the City of Portland (see Wang et al., 1991).

2.2 The Role of Valuation (Cost Estimation

Procedures)

Although technological development and

research attention has rightly focused on ways to

facilitate inventory development for lifeline networks, the greater emphasis in this project

turned out to be the relative absence of uniform

valuation methods for water system components.

The experience of project participants in water

system engineering and planning was

508
demonstrated in their sensitivity to the

uncertainties involved in estimating replacement

and repair costs. Often overlooked in scientific and engineering risk studies, these uncertainties can easily yield 30 percent variations in some

earthquake loss estimates-at least for specific

components. (A more complete discussion of

this issue is provided in Taylor et al., 1990.)

2.3 The Role of Ground Attenuation and

Liquefaction Probability Models in Seismic

Risk Analysis

Although covered elsewhere in the growing

earthquake literature, the monograph contains

general criteria for the selection of suitable

functions for modeling how earthquake ground

motions attenuate with distance from the

earthquake source. For the specialist as well as the interested engineer, these criteria are expounded in terms of:

• types of variables, including the dependent

strong-motion parameter to be used in conjunction with component loss models

and the independent parameters such as earthquake magnitude selected, distance

from source to site, and local site effects.

• range of variables for application and

empirical evaluation (e.g., accuracy of

distance estimates, lower and upper

earthquakes magnitudes considered, and

selection of valid strong-motion records)

• functional form, or form of the equation

applied and statistically evaluated.

The models used to assess liquefaction-induced

ground failure hazards are based on an

extrapolation from standard in-situ methods

rather than on general liquefaction probability

matrixes. Adaptation of in-situ methods,

whether to critical components directly or to microzones through extrapolation from existing

boring log and geological data, potentially

permits this element of seismic risk analyses to be improved through research designed to examine ftirther these standard methods. As one

example, the current problem of how earthquake

magnitude (or, somewhat more precisely, the

number of strong ground motion cycles) affect liquefaction probabilities is a problem for seismic risk analysis as well as a typical research

problem for standard methods.

2.4 The Role of Component-Loss Models in Seismic Risk Analyses

The latitude that was provided task leaders

produced noteworthy results with respect to the

development of component loss models. These

models represent loss (as a percent of

replacement value) as a function of some

measure of earthquake intensity (e.g.. Modified

Mercalli intensity, peak ground acceleration,

peak ground velocity, etc.). Response to this task typically required clarification of the

component-loss model to be developed and, as expected, a lively discussion of what notion of

intensity was suitable.

Resulting component loss models were

developed from (a) readily-available existing

models i.e., the Applied Technology Council

(ATC-13) models), (b) empirical data and/or

engineering judgment and/or (c) analytic

methods. In many cases, as with water filtration plants, past earthquakes provide only scant evidence of potential earthquake damage.

Furthermore, this evidence is only rarely

documented in terms of dollar loss relative to replacement value, and it is often applicable to

filtration plants whose seismic design provisions were very different from those of the plants

analyzed in the TCLEE project. In other cases,

as with pipelines, investigators have made

significant strides in compiling damage data in earthquakes.

Given the dearth of empirical damage or loss

data, the uniqueness of some structures, the need

to explain damage through analysis, and the

subjectivity of readily-available models, task

leaders in two topic areas developed analysis

methods in order to arrive at component loss models. Those two topic areas were filtration plants and pipelines, respectively. For this

509
presentation, we shall discuss very briefly the

application of analytic methods to filtration plants.

Analytic methods consist of the use of dynamic

response calculations, actual earthquake data,

and/or experimental data to establish algorithms

that interrelate dynamic response with potential

earthquake losses. A wide range of dynamic

response descriptors and input motion intensity

descriptors can be used in these methods as long

as (a) the input motion intensity descriptors can be correlated with the potential earthquakes

simulated for the water system; and (b) the

dynamic response descriptors can be used to establish meaningful relationships between

dynamic response and estimated losses.

The TCLEE Monograph contains an example of

the application of an approximate analytic

method to estimate a loss model for a chemical

feed building. This method consists of (a) the

development of design spectra for the building,

based on empirical compilations of ground

response data for different geologic conditions;

(b) estimation of a force-deformation relationship

for the building that incorporates the building's

yield limit and its ultimate capacity; (c) use of

these results to obtain approximate dynamic

characteristics of the building in terms of its natural period and damping ratios at yield and at ultimate; and (d) estimation of the potential

damage to the building as a function of the

ground acceleration, by using the design spectra

from Step (a), the approximate dynamic

characteristics of the building from Step (c), and

certain simplifying assumptions that relate the

degree of damage to the amount of inelastic deformation estimated for the building at each

ground acceleration level.

2.5 Probabilistic Procedures for System-wide

Loss Calculation

A major emphasis of the monograph was the use of probabilistic rather than deterministic

procedures for calculating potential earthquakeinduced losses in geographically dispersed

networks. At a minimum, probabilistic

modeling incorporates loss estimates from an

entire suite of representative earthquakes. In

contrast, deterministic modeling considers only

a small sample of earthquake scenarios.

Whether or not probabilities are somehow

assigned to these scenarios, enormous biases

enter into the selection. For those who are wellversed in earthquake risk analysis, it becomes

possible to select the earthquake scenarios that provide the desired loss result, whether high or low. In the monograph, we avoid these

potential biases through the use of probabilistic

methods initially developed as recently as the

mid-1980's.

3. RECOMMENDATIONS FOR FURTHER

RESEARCH AND APPLICATION

The state-of-the-art in assessing the earthquake

risks to water systems can still benefit from

improvement in a number of areas, as implied

for instance by the discussion of highlights.

That discussion yields the following research

recommendations:

• the need for continuing review and

improvement of valuation procedures,

especially as the ability to transfer data

improves

• the ongoing need for improvements in all geoscience models, including those of

earthquake activity sources and rates, attenuation models, and models of the

response of sites owing to local soil conditions; the need for improved models to estimate liquefaction probabilities was

apparent in this project

• the strong need to replace subjectively

derived component loss models with models

based on experience and/or analysis.

• the need for a continuing re-examination of

how parameter variability affects loss estimates (this includes continuing progress

in clarifying the variability of individual

parameters).

510
• compilation of data to assist in the future

development of models for calculating

secondary losses.

Already earthquake loss and risk estimation

procedures have been used in:

• regional earthquake loss estimation studies

including water systems

• studies to develop guidelines for state regulatory activities with respect to culinary

water facilities, and

• studies designed to assist in developing loss prevention programs for specific systems or subsystems.

Although these earthquake risk studies are

clearly providing important insights for

planning, design, and retrofit of water systems,

there is a need to further develop and refine the current risk modeling and analysis procedure.

The monograph itself was designed first for the

audience of participants themselves and second

for an audience of other potential investigators

interested in current methods and difficulties in earthquake loss estimation. Within the limited

scope of this paper, only suggestions can be

made of how to further this technology transfer. A manual could be a next step, especially if some of the above research issues that outline

the step-by-step seismic risk analysis process are addressed. Assignment of a working group to define further possible ways to effectively apply

seismic risk analysis to water systems could be

another step. Along these lines, earthquake risk

methods can be incorporated into current National Institute of Standards (NIST)/Federal

Emergency Management Agency (FEMA)

efforts to determine how to develop and

implement earthquake risk reduction strategies

for lifeline networks.

4. REFERENCES

1. Applied Technology Council, ATC-13,

1985. Earthquake Damage Evaluation Data

for California . Redwood City, California.

2. Ballantyne, Donald B., E. Berg, J. Kennedy, R. Renau, D. Wu, C. Taylor,

C. Crouse, R. Eguchi, and C. Tillman,

1990, "Earthquake Loss Estimation

Modeling of the Seattle Water System,

"Federal Way, WA: Kennedy/Jenks/Chilton

for the United States Geological Survey,

October.

3. Ballantyne, Donald B., William F. Heubach, C.B. Crouse, Ronald T. Eguchi,

Craig E. Taylor, Craig W. Tillman, Stuart D. Werner, Dennis K. Ostrom, and Felix

Wong, 1991, "Earthquake Loss Estimation

for the City of Everett, Washington

Lifelines," Federal Way, WA: Kennedy/

Jenks/Chilton for the United States Geological Survey, May.

4. Taylor, Craig E. (1991) editor. Seismic

Loss Estimates For A Hypothetical Water

System New York City: The American

Society of Civil Engineers, Technical

Council on Lifeline Earthquake Engineering

Monograph No. 2, August.

5. Taylor, C.E., W. Elliott, H. Cornell, and

D. Ballantyne (1990), "The Demonstration

Project on Seismic Loss Estimation for Water Systems," pp. 1-13 in Recent

Lifeline Seismic Risk Studies , ed. Anne S. Kiremidjian, New York City: The

American Council of Civil Engineers,

Technical Council on Lifeline Earthquake

Engineering Monograph No. 1, November.

6. Taylor, C.E., R.T. Eguchi, L.R. Wang and

J. Isenberg (1988), "Illustrative Methods for

Deriving Earthquake Losses Expected to a Water System," Ninth World Conference on

Earthquake Engineering, Tokyo.

511
7. Wang, Leon R.L., Joyce C.C. Wang, Isao

Ishibashi, Donald B. Ballantyne, and

William M. Elliott, 1990, "Development of

Inventory and Seismic Loss Estimation

Model for Portland, Oregon Water and

Sewer Systems," Norfolk, VA: Old

Dominion University, Department of Civil

Engineering, for die U.S. Geological

Survey, April.

512
Bridge Scour/An Element of Bridge Management

by

J, Sterling Jones' and Stanley R. Davis^

ABSTRACT

Several catastrophic bridge failures in the

United States were attributed to bridge scour. These events spurred the engineering

community to take a hard look at methodologies for evaluating bridges for scour. The Federal Highway Administration (FHWA)

took the lead in assembling the technology and

State Highway Agencies have taken various

approaches for incorporating bridge scour as a major element in existing bridge management

programs. A program that is being used by

the state of Maryland is described in this paper

as an example of how the bridge scour assessments are being handled and as a basis

for projecting the magnitude of the program

nationwide.

KEYWORDS: bridges; bridge management;

bridge scour; hydraulics; waterways.

1. INTRODUCTION

The Federal Highway Administration (FHWA)

and the State Highway Agencies in the U.S.

are undertaking a major venture to assess the

nation's approximately 481,000 bridges over waterways for scour. This venture was prompted in part by two catastrophic failures

which resulted in several fatalities and caught

the attention of safety officials, news media

and legislators. One was the Schoharie Creek

bridge in N.Y. which collapsed in April 1987

and caused 10 fatalities. The other was the

Hatchie River bridge in Tennessee which

collapsed in April 1989 and caused 8 fatalities.

These events were milestones in this nation's

concept of bridge scour as a significant

component of the overall bridge management

program. They spurred the engineering

community to take a hard look at existing

methods for evaluating new and existing

bridges to resist damage from scour. At that

time, there was no comprehensive guidance or

direction generally acceptable to the

engineering community on how to address

bridge scour in design and inspection.

To respond to this lack of guidance, FHWA

headed up a national cooperative effort among

engineers from The National Transportation

Safety Board, State Highway Agencies, U.S.

Geological Survey, consultant firms and

universities to assemble available information

and publish what later became Hydraulic

Engineering Circulars 18 "Evaluating Scour at Bridges" (Richardson, et. al. 1991) and 20,

"Stream Stability at Highway Structures"

(LaGasse, et. al. 1991). FHWA and State

Highway Agencies updated key items on the

National Bridge Inventory to include more

information on bridge hydraulics and scour. FHWA developed a demonstration project on

underwater inspection and training courses on

' Hydraulics Research Engineer, Federal

Highway Administration, Turner Fairbank

Highway Research Center, McLean, VA

22101-2296

^ Consultant/Hydraulics Engineer, Maryland

Department of Transportation, State

Highway Administration, Office of Bridge

Development, Baltimore, MD 21203-0717;

Previously Chief of the FHWA Hydraulics

Branch, Washington, D.C.

513
stream stability and bridge scour. FHWA then

set timetables for the State Highway Agencies

to assess the susceptibility of bridges to scour and to evaluate and take appropriate action

for the bridges that are scour susceptible.

After some discussion with State Highway

Agencies, FHWA set two target dates for

completion of a bridge scour assessments.

April 1992 was set as a target date for

completion of a screening to determine which

bridges are scour susceptible. January 1997 was set as the target date for completing the

more detailed evaluations of the scour

susceptible bridges to determine which ones

are critical and require remedial action.

State Highway Agencies faced a major bridge

management problem in implementing this program. First, scour information was not well

documented in the past. Second, there were

limited resources both in personnel and in funds to do the scour evaluations. Third,

there was a perceived liability and

accountability issue that requires a process that

is consistent and supportable based on prudent

and reasonable judgment. Bridge scour by its nature involves some risk, but State Highway

Agencies have to balance the risk against

available resources to determine which bridges

are evaluated first and what detail of

evaluation is appropriate.

Each State Highway Agency has developed

some type of management approach that is adaptable to the information available to make

prudent and reasonable judgment. This paper

describes an approach that was developed by

the State of Maryland and relates it to a national perspective.

2. CHARACTERIZATION OF BRIDGES

Bridges in the United States are listed in a National Bridge Inventory (NBI) which is updated by State Highway Agencies and

maintained by the FHWA Elias (1992)

studied data in the NBI to establish a profile

of bridges in the inventory. Based on his

analysis, which reflects a December 1990

inventory, there are 577,000 bridges with

approximately 481,000 (or 83%) over

waterways. Of the 481,000 bridges over

waterways, approximately 90,000 (19%) are

actually culverts which are classified as small

bridges because they span more than 20 feet.

The remaining 391,000 bridges have the

following characteristics: • Approximately 50% of the bridges in the U.S. have been built within the last 35 years • Approximately 7% are more than 75

years old • Approximately 40% of the bridges are simple span construction • Approximately 40% of the bridges

have Average Daily Traffic (ADT) less than 100 vehicles per day • Approximately 28% have ADT greater

than 1,000 vehicles per day.

Based on the nearly complete national

screening results, approximately 25% of the

bridges are being categorized as scour

susceptible and will require more detailed

evaluation to determine which ones are scour

critical. Based on a relatively small number of

the evaluations that have been completed,

approximately 11% of the scour susceptible

bridges are being categorized as scour critical.

The State of Maryland has 1354 Federal-aid

bridges over waterways in The National Bridge

Inventory. Four hundred forty of these (32%)

are culverts. Two hundred fifty six (19%) of

the Maryland bridges are in tidal waterways

and will have deferred evaluations for three

reasons: (1) there is a general lack of

technology to evaluate scour in tidal areas; (2)

tidal estuaries tend to dampen the velocity

currents for the inland bridges, and (3) the

really critical bridges on the coastal waterways

are under constant surveillance anyway.

Maryland considers that all of the 658

remaining bridges over waterways are scour

514
susceptible and require a rating. These bridges

will be subjected to varying degrees of

evaluations depending on the severity of the scour ratings.

3. DEVELOPMENT OF A BRIDGE

SCOUR MANAGEMENT PROGRAM

The Maryland Highway Administration

recognized the need to draw resources from

several different offices and from consultants.

A key decision in the use of consultants was

that the state must retain the responsibility for

assessing the risks involved in the bridge scour management program, and for deciding on the

actions necessary to maintain the integrity and

safety of the bridge structures.

One of the first things done was to form an

interdisplinary team to provide guidance and

direction to the team, and to promote

coordination and cooperation among the many

offices and individuals involved with the

general bridge program. Responsibilities of

various offices involved in the bridge scour management program were assigned as listed below:

3.1 Bridge Hydraulics Unit

• Technical expertise, guidance and

direction • Overall plan coordination • Establish agenda for

interdisciplinary team

• Prepare bridge scour evaluation

reports and review scour reports

prepared by others • Prepare status reports, • Coordinate program with FHWA,

USGS, etc. • Maintain oversight of research

efforts • Coordinate with other state DOTS,

University of Maryland, etc. • Conduct screening and

prioritization studies of bridges for

vulnerability to scour damage

> • Rate Bridges for Item 113 of The

National Bridge Inventory • Coordinate Training Needs;

Conduct In-House Training as Necessary to Meet Program

Objectives

3.2 Office of Bridge Development

• Overall management decisions on

program approaches and directions,

coordination with EHWA and

counties, etc. • Bridge Design Division lead office

for designing bridges over

waterways; makes decisions on

designing foundations to resist scour using guidance and

recommendations from Bridge

Hydraulics Unit

3.3 Bridge Inspection and Remedial

Engineering

• Decisions on what actions to take

and relative priorities on bridge

needs

. • Bridge Inspection Program

• Emergency concerns, closing of

bridges, etc. • Administering Action Plan for scour

critical bridges • Maintaining records on the

National Bridge Inventory

3.4 Office of Materials and Research

• Soil borings, stream bed and bank

samples, etc. • Evaluation of foundations in rock;

scourability of rock, etc. • Other guidance on GeoTech

matters

The Maryland State Highway Administation

then set up two agreements to establish a data

base for the program. First, and very

important, was an agreement with the U.S.

515
Geological Survey (USGS) to have two threeman teams conduct a preliminary field survey

of non-tidal bridges over waterways that were

in the NBI. These crews assigned a USGS

rating factor to each bridge based on the

following:

• Stability of channel bed and banks • Approach and "get away" flow

conditions to bridge • Degree of observable scour at bridge • Susceptibility to damage from scour based on conditions observed at the

site.

The USGS teams also rated NBI items 61

(channel), 71 (waterway adequacy) and

special state item 321 (scour) for each bridge

they surveyed. The fact that there were only

two teams gave a degree of uniformity to these

ratings that gives them a high weight in a ranking system.

Second, was a contract with a consultant to develop a spread sheet that could be readily

managed by state engineers. The spread sheet

included the following factors for each bridge:

• Traffic (ADT)

• Superstructure (continuous or simple spans) • Foundation (spread footing, piles,

etc.) • Waterway area (Item 71) • Channel Stability (Item 61) • Observable Bridge Scour (Special

State Item 321) • USGS Rating Factor (from field survey) • (SHA) Engineer Rating Factor

(Based on engineer's evaluation of

all available field and office data)

Numbers are assigned to each of these factors

and the spread sheet priorities are updated at least monthly as additional information

becomes available.

4. SCREENING AND SCOUR

ASSESSMENT

The Maryland State Highway Administration

uses several levels of screening and

assessments to determine which bridges

require a full scour evaluation. Advance

screening eliminated culverts and tidal bridges

for reasons cited previously. The preliminary

screening process essentially assigns numbers

to the spread sheet factors and ranks bridges

from worst to best. The office scour assessment involves some preliminary scour computations, review of foundation

information, and special circumstances for the

bridge. At this stage some bridges can be

categorized as scour critical and others can be

categorized as low risk and can be eliminated

from further assessment. The office

assessment is a relatively low cost effort that

requires approximately one person-day per

bridge. The spread sheet rating factors,

especially the (SHA) engineer rating factor, are revised after this assessment. The field and office scour assessment involves site visits by State Highway engineers, more detailed

data gathering, and more precise

computations. Again, the spread sheet rating

factors are revised after this assessment. Some

bridges then will require a full scour evaluation

which typically involves taking soil borings,

rerunning water surface profile models to get

flow and velocity distribution, estimating

expected scour depths, and determining

structural integrity of bridges with partially

exposed foundations. Table 1 shows the

expected number of bridges and the relative costs for each process.

5. FACTORS USED TO DEVELOP THE

SCOUR PRIORITY NUMBER

A central focus of Maryland's Bridge Scour

Program is the spread sheet ranking system

which is continuously updated after each

process level and whenever new data or better

technology becomes available. This system

keeps the highest risk bridges at the center of

516
attention.

The factors used to develop a ranking are of

interest and deserve some discussion. Table 2

summarized the spread sheet factors with

guidelines used by Maryland for assigning

numerical values that calculate a scour priority

number for each bridge. Using Maryland's

guidelines, the lower the scour priority

number, the greater the likelihood that the

bridge will be damaged by scour. APT reflects primarily on the value of lost time, and traffic running costs if an important

bridge fails. It also is a secondary indicator of

the probability of fatalities if a bridge does fail, although there are not enough data to establish any correlation between ADT and

the probability of fatalities.

Superstructure type is important because

continuous spans have more redundancy and

less chance of catastrophic failure with little or no warning. This is also an indicator for the

probability of fatalities.

Foundation type is definitely one of the most

important factors. Historically, most of the scour related failures have been bridges on

spread footings.

Waterway adequacy reflects the probability

that pressure flow scour is likely to occur. It also may be an indicator of contraction scour.

It is an available piece of data that has merit

in a ranking system.

Channel stability reflects the tendency for

degradation and channel migration. This

factor should have a high weight because

instable channels generally have more local scour and react to small channel changes.

Bridge Inspection Observations are important

because they highlight the bridge inspectors

identification of potential scour problem.

The USGS rating factor is especially important

in Maryland because of the uniformity of that

factor since the same survey crews inspected

and rated all of the bridges. The same survey crews also rated the NBI items.

The SHA Engineer rating factor takes into

account the accuracy of records and

conservatism of engineers at various periods of

time. This factor is based on an overview of

all of the information in the system, and is the

most dynamic factor as the screening and

assessment progresses.

The "screamer" factor allows for any bridge to

rise to top of the list at any stage of the

assessment if anyone reports evidence that a bridge is in danger.

The scour priority number is calculated by

multiplying the numerical values assigned to each of these factors. The ratings can be

criticized because there is no provision for

weighing factors according to importance and

reliability of information. It would appear that

the wide range of values for some of the less important factors, such as waterway adequacy,

would make them have more weight than they

are intended to have and lead to erroneous

ranking. This was an initial concern with the use of the system, but subsequent testing has

satisfied Maryland State Highway

Administration engineers that the system,

while not perfect, does a reasonable job of

prioritization. This success is due to several

moderating effects, including the dynamic

nature of the program, the flexibility to revise

numerical values as more information becomes

available, and the central tendency of the NBI

rating numbers.

Other State Highway Agencies have developed

similar bridge scour management programs.

There has been at least one attempt by Elias

(1992) to consolidate the best features of the

programs into one that can be promoted

nationally, but that procedure is still in draft

stage and has not been tested by any state.

517
7. SUMMARY

The Federal Highway Administration took the

lead in assembling available information,

developing training courses, and providing

demonstration projects to deal with an

accelerated program to assess the nation's

bridges for scour susceptibility. The State

Highway Agencies have had to integrate this program into their overall bridge management

program with limited resources.

Maryland is a relatively small state with fewer

bridges than the average state, but it is a sampling that reflects the magnitude of the

overall scour program in the United States. If the average $4,000 per bridge cost for

managing the bridge scour program is extrapolated to the total 391,000 bridges

(excluding culverts) over waterways, the total cost of this program nationally will exceed $1.5

billion over the next decade.

Maryland has taken a somewhat pragmatic

approach in developing its bridge scour program, placing emphasis on engineering

judgment to limit the number of bridges

requiring full scour evaluation. This approach,

coupled with an active bridge inspection

program, works for that state. The following

considerations were essential to the success of

their program:

• Importance of the Bridge

Inspection Program in identifying

and prioritizing potential scour problems • Use of the Scour Priority Number

Spreadsheet as a management tool

for evaluating existing bridges

8. REFERENCES:

1. Richardson, E.V., Harrison, L.J., and

Davis, S.R. (1991). FHWA Hydraulic

Engineering Circular No. 18, "Evaluating

Scour at Bridges," Publication No.

FHWA-IP-90-017, Washington, D.C.

2. LaGasse, P.F. et. al. (1991). FHWA

Hydraulic Engineering Circular No. 20,

"Stream Stability at Higway Structures,"

Publication No. FHWA-IP-90-14,

Washington, D.C.

3. Elias, Victor (1992). "Strategies for

Managing Unknown Bridge Foundations,"

Unpublished Draft Publication No.

FHWA-RD-92-030, FHWA R&D,

McLean, Va. 22101.

• Support of top management

® Formation of an interdisciplinary

team

• Knowledge of and sensitivity to concerns about scour by the

engineers in the bridge

management program

• Development of a plan of action to

identify the tasks to be addressed • Assistance of the FHWA and the

USGS in training efforts and in

developing the program

• Written directives establishing

policies and rsponsibility for the scour program

518
Table 1. Maryland Bridge Scour Program

Process Expected No. of

Bridges

Estimated Cost per

Bridge

Expected Cost

Preliminary Screening 658 $ 515 $ 340,000

Office Scour Assessment 658 $ 350 $ 230,000

Field & Office Scour

Assessment 219

(Approx. 33%)

$ 1600 $ 350,000

Full Scour Evaluation 110

(Approx. 17%)

$ 15,000 $ 1,650,000

Grand Total $ 2,570,000

Average Cost per Bridge $ 4,000

Table 2. Maryland Factors used to Rate Bridges for Scour Susceptibility

Factor Criteria and Numerical Indicators

ADT > 30,000 veh/day = 1 5 - 30,000 veh/day = 2

< 5,000 veh/day = 3

Superstructure simple span = 1 continuous = 2

Foundation Spread footing, erod mat'l. = 1 Unknown = 2

Short Pile = 4

Long Pile > 20' =6

Found on Solid Rock = 6

Waterway Adequacy

NBI Item 71

1 to 9

from Inventory Data

Channel Stability

NBI Item 61

1 to 9

from Inventory Data

Bridge Inspection Notes Regarding Scour

Special Maryland NBI Item 321

1 to 9

from Inventory Data

USGS Rating Factor severe = 1 moderate = 2

mild = 3

SHA Engineer Rating Factor Preliminary = 3

Revised after Office Assessment

severe - - 1 moderate = 2

mild = 3

"Screamer" Factor Overrides all others and places a bridge at the top of the list.

519
I
U.S. Coordinated Program for Masonry Building Research - Seventh Year Status

by '

,

,

James L. Noland'

ABSTRACT

The U.S. program, which consists of twentyeight specific research tasks, is coordinated with

a parallel program in Japan. Both are conducted

under the auspices of the UJNR Panel on Wind

and Seismic Effects. The primary purpose of the

U.S. program is to support the development of a

limit state design procedure for masonry as well

as provide experimental data, analytical

methods, etc. Several technical accomplishments have been realized in both modeling

and experimental areas. The technology transfer

plan includes dissemination of research reports,

topical reports and a summary report on the

specific research areas as well as by pre- sentations and papers. Design and criteria recommendations will be an extremely important mode of technology transfer and are expected to support limit state design standard

development. Current status of the program is presented. Completion is expected by July

1993.

KEYWORDS: masonry; progress; strucmres; technology transfer, UJNR

1. INTRODUCTION

The U.S. Coordinated Progr;un for Masonry

Building research is an integrated set of 28

specific research tasks being conducted by the

U.S. Technical Coordinating Committee for

Masonry Research (TCCMARAJS). The U.S.

program is coordinated with a parallel program

in Japan to exchange information and concepts

for the mutual benefit of both countries. Both

programs are conducted under the auspices of

the United States - Japan Natural Resources

Panel on Wind and Seismic Effects.

2. PROGRAM OBJECTIVES

Primary program objectives are:

1) Development of design and criteria recommendations for limit state design

of reinforced masonry buildings and

components.

2) Development of a consistent ex- perimental database on the behavior of

masonry materials, components and

systems.

3) Development of analytical non-linear

models for research and design office use for detailed analysis, system analysis, and dynamic loads determination.

4) Improved material and subassemblage

experimental procedures for obtaining

masonry properties.

5) Improved masonry fabrication procedures and standards.

6) Developing an increased awareness among engineers, architects, code

bodies and the public of the capabilities

of reinforced masonry in all seismic

zones.

7) Interfacing with standards development

groups to support development of a consensus limit state standard for masonry.

lAtkinson-Noland & Associates, Inc. 2619 Spruce Street Boulder, CO 80302

521
3. APPROACH

With NSF support, the Technical Coordinating

Committee for Masonry ResearchAJ.S.

(TCCMAR/U.S.) was formed in February 1984.

TCCMAR/U.S. was and is comprised of re- searchers from academic and industrial organizations who have strong backgroimds in

research into the properties and characteristics of

reinforced masonry materials, structural components and systems, analytical techniques,

structural dynamics, building codes, and earth- quake engineering. TCCMAR researchers are

listed in Table 1 . TCCMAR participants defined

the research program conducted, the research

and will analyze and interpret the results.

The U.S. research plan is a phased step-by-step

program of separate, but coordinated research

tasks. Emphasis is being placed upon intra-task

information exchange and consultations so the

experimental and analytical efforts are mutually

supportive. The research tasks comprising the

coordinated program are listed in Table 2.

The program includes both experimental and analytical research tasks. The experimental and

analytical efforts are conducted in parallel with

each becoming more complex and sophisticated

as knowledge and experience accumulate. The

U.S. plan is thus a "building block" approach.

The program began in October 1985 and will be

completed by July 1993.

4. CURRENT STATUS

The program is approximately 90% complete.

Remaining work consists of completing experimental efforts on reinforcing bar bond and

splicing, experimental and analytical studies of

the behavior of a five story research building,

development of limit state design and criteria

recommendations, and final technology documentation. Research task reports which have

been published are listed in Table 3.

Construction of the five story research building

has begun. It is expected that it wiU be completed, instrumented, and checked by Jime 1992.

Testing should be done through the summer.

Funding for the full scale research has been pro- vided by the National Science Foundation, the

Department of Energy, and the U.S. masonry

industry (organizations such as the Council for Masonry Research, the Masonry Institute of

America, the California-Nevada Concrete

Masonry Association, and thirty regional

masonry associations).

Work on Technology Documentation and

Design and Criteria Recommendations will begin formally with a meeting on 10 April 1992.

Plans and assignments will be made at that time.

TCCMAR Category 2 researchers have done

preliminary work to support design and criteria recommendations; this, and a draft of a limitstate design standard previously prepared, will

be the starting point.

The Design and Criteria Recommendations

project will basically consist of refining the draft limit state design standard, validation of the

design standard by trial designs of buildings of

different types in different seismic zones, and re- finement of the standard as required.

Technology Documentation will involve the

preparation of a series of monographs on key

topics to clarify and summarize research

findings. The dociunents will be published by

the Federal Emergency Management Agency.

Funding for Technology Documentation and

Design and Criteria Recommendations has been

provided by the National Science Foundation,

the U.S. Army Corps of Engineers, and the

Federal Emergency Management Agency.

522
TABLE 1 TCCMAR RESEARCHERS

Daniel Abrams Samy Adham

University of Illinois Agbabian Associates

Urbana, IL El Segundo, CA

Richard Atkinson Russell Brown

Atkinson-Noland & Associates Qemson University

Boulder, CO Qemson, SC

Robert Englekirk Robert Ewing

Englekirk,Hart & Sabol Ewing & Associates

Los Angeles, CA Rancho Palos Verdes, CA

Ahmad Hamid Mike Mammons

Drexel University US Army Corps of Engmeers

Philadelphia, PA Vicksburg, MS

Gary Hart Gilbert Hegemier

Englekirk,Hart & Sabol/UCLA Umv. of CA-San Diego

Los Angeles, CA La JoUa, CA

John Kariotis Richard Klingner

Kariotis & Associates Univ. of Texas-Austin

South Pasadena, CA Austin, TX

Ronald Mayes James L. Noland

Computech Engin. Services Atkinson-Noland & Assoc.

Berkley, CA Boulder, CO

Max Porter M.J.N. Priestley

Iowa State University Univ. of CA-San Diego

Ames, lA La JoUa, CA

Freider Seible P.B. Shing

Univ. of CA-San Diego University of Colorado

La JoUa, CA Boulder, CO

Leonard Tulin

University of Colorado

Boulder, CO

523
TABLE 2

RESEARCH TASKS

SUBJECT RESEARCHER

Grouted Hollow-Unit Clay & Concrete

Masonry in Compression

Grouted Hollow-Unit Concrete Masonry

Under Combined Compression & Flexure

Grouted Hollow-Unit Clay Masonry Under

Combined Compression & Flexure

Masonry Test Methods for Limit State Design^

Nonlinear Force-Displacement Models of

Reinforced Masonry Components & Systems

Nonlinear Finite Element Models for

Analysis of Reinforced Masonry

Nonlinear Lumped Parameter Models for

Dynamic Response Analysis of Reinforced

Masonry Systems

Nonlinear Models of Floor Diaphragms

Nonlinear Models of Reinforced Masonry

Walls Acting Out-of-Plane

Experimental Study of One-Story

Reinforced Masonry Shear Walls

Experimental Study of Two-Story

Reinforced Masonry Shear Walls

Experimental Study of Three-Story

Reinforced Masonry Shear Walls and

Development of Automated Feed-Back Test

Method for Stiff Structures

Experimental Study, Static & Dynamic,

of Out-of-Plane Behavior of Reinforced

Concrete Masonry Walls

R. Atkinson

A. Hamid

R. Brown

R. Atkinson

R. Englekirk

R. Ewing

J. Kariotis

M. Porter

R. Mayes

B. Shing

R. Klingner

G. Hegemier

F. Seible

A. Hamid

S. Adham

524
SUBJECT

Experimental Study, Static & Dynamic,

of Out-of-Plane Behavior of Reinforced

Qay Masonry Walls

Experimental Study, Static & Dynamic,

of Flanged Reinforced Masonry Walls

Analytical Study Based on Prior Tests,

of Wall-Floor Intersections^

Experimental Study (Preliminary) of

Reinforcing Bar Bond & Splices in

Grouted Masonry

Experimental Study (continuation) of

Reinforcing Bar Bond and Splices in

Grouted Masonry^

Experimental Study, Static & Dynamic

Behavior of Scale Model Three-Story

Reinforced Masonry Buildings

Development of Reliability-Based

Limit-State Design Methodology for

Masonry Buildings

Development of Reliability Factors for

Masonry Materials, Components, & Systems

Design of Five-Story Reinforced Masonry

Research Building

Design and Construction of Large

Specimen Research Building^

Experimental Study of Nonlinear

Behavior of a Five-Story Building

Under Cyclic Lateral Loads^

Design & Criteria Recommendation for

Limit State Design and Documentation and

Technology Docimientation^

RESEARCHER

R. Mayes

N. Priestley

G. Hegemier

L. Tulin

R. Atkinson

M. Hammons

D. Abrams

G. Hart

G. Hart

J. Kariotis

G. Hegemier

F. Seible

G. Hegemier

F. Seible

J. Noland

Overall Program Coordination^ J. Noland

^Research Tasks to be completed

525
TASK REPORTS^

Task No. Author/s - Title

1.1-1: Atkinson and Kingsley, Comparison of the Behavior of Clav & Concrete Masonry in Compression. September 1985. 151 pgs.

1.2(a)-l: Hamid A.A., Assis, G.F., Harris, H.G., Material Models for Grouted Block Masonry. August 1988. 67 pgs.

1.2(a)-2: Assis, G.F., Hamid A.A., Harris, H.G., Material Models for Grouted Block Masonry. August 1989. 134 pgs.

1.2(b)-l: Young, J.M., Brown, R.H., Compressive Stress Distribution of Grouted Hollow Clay

Masonry Under Strain Gradient. May 1988. 170 pgs.

1.3-1: Atkinson, R.H., An Assessment of Current Material Test Standards for Masonry Limit

States Design Methods. June 1991. 38 pgs.

2.1-1: Hart, G. and Basharkhah, M., Slender Wall Structural Engineering Analysis Computer

Program (Shwall, Version 1.01), September 1987. 68 pgs.

2.1-2: Hart, G. and Basharkhah, M., Shear Wall Structural Engineering Analysis Computer

Program (Shwall, Version 1.01). September 1987. 75 pgs.

2.1-3: Nakaki, D. & Hart, G., Uplifting Response of Structures Subjected to Earthquake

Motions. August 1987. 200 pgs.

2.1-4: Hart, G., Sajjad, N., and Basharkhah, Inelastic Column Analysis Computer Program

(INCAP, Version 1.01), March 1988.

2.1-5: Hong, W.K., Hart, G.C., Englekirk, R.E..Force-Deflection Evaluation and Models for

University of Colorado Flexural Walls. December 1989.

2.1-6: Hart, G.C. Jaw, J.W., Low, Y.K., SCM Model for University of Colorado Flexural WaUs. December 1989. 31 pgs.

2.1-7 Hart, G., Sajjad, N., Basharkhah, M., Inelastic Masonry Flexural Shear Wall Analysis

Computer Program . February, 1990. 41 pgs.

2.1-8 Hart, G., Englekirk, R., Srinivasan, M., Huang, S.C., Drag, D.J,., Seismic Performance

Study. DPC Gymnasium. Elastic Time History Analysis Using SAP90. February, 1992.

2.1-9 Hart, G., Englekirk, R., Srinivasan, M., Huang, S.C., Drag, D.J., Seismic Performance

Study. TMS Shopping Center. Elastic Time History Analysis Using SAP90. February,

1992.

2.1-10 Hart, G., Englekirk, R., Jaw, J.W., Srinivasan, M., Huang, S.C., Drag, D.J., Seismic

Performance Study. RCJ Hotel February. 1992.

2.1-11 Hart, G., Englekirk, R., Jaw, J.W., Huang, S.C., Drag, D.J., Seismic Performance Study.

2-Story Masonry WaU-Frame Building. February, 1992.

2.1-12 Hart, G., Englekirk, R., Jaw, J.W., Huang, S.C., Drag, D.J., Seismic Performance Smdv.

2-Story Masonry Wall-Frame Building Designed by Tentative Limit States Design

Standard . February, 1992.

3Task Reports are available through the Earthquake Engineering Research Library, Berkley, CA, and The

Earthquake Engineering National Center, SUNY, Buffalo, NY.

526
2.2-1: Ewing, R., El-Mustapha, A., Kariotis, J., FEM/I - A Finite Element Computer Program

for the Nonlinear Static Analysis of Reinforced Masonry Building Components, December 1987 (Revised June 1990).

2.2-2: Ewing, R.D., Parametric Studies on Reinforced Masonry Shear Walls Using FEM/I, A

Nonlinear Finite Element Analysis Program , March 1992.

2.2-3: Ewing, R.D., Finite Element Analysis of Reinforced Masonry Building Components

Designed bv a Tentative Masonry Limit States Design Standard , March 1992.

2.3- 1: Ewing, R.; Kariotis, J.; El-Mustapha, A., LPM/I, A Computer Program for the Nonlinear,

Dynamic Analysis of Lumped Parameter Models. August, 1987. 200 pgs.

2.3-2: Kariotis, J., El-Mustapha, A., Ewing, R., Influence of Foundation Model on the Uplifting

of Structures. July 1988. 50 pgs.

2.3-3: Kariotis, J., Rahman, M., El-Mustapha, A., Investigation of Current Seismic Design Provisions for Reinforced Masonry Shear Walls, January 1990. 48 pgs.

2.3-4: Kariotis, J., Rahman, A., Waqfi, O., Ewing, R., Version 1.03 LPM/I - A Computer

Program fo the Nonlinear. Dynamic Analysis of Lumped Parameter Models, February

1992.

2.3-5: Kariotis, J., Waqfi, O., Ewing, R., A Computer Program Using Beam Elements for the

Nonlinear. Dynamic Analysis of Lumped Parameter Models. February 1992.

2.3-7: Kariotis, J., Waqfi, O., Recommended Procedure for Calculation of the Balanced

Reinforcement Rafio. February 1992.

2.4(b)-l: Button, M.R., Mayes, R.L., Out-of-Plane Seismic Response of Reinforced Masonry

Walls : Correlation of FuU-Scale Test and Analytical Model Results, March 1991. 65

pgs.

3.1(a)-l: Scrivener. J.. Summary of Findings of Cyclic Tests on Masonry Piers. June 1986. 7 pgs.

3.1(a)-2: Shing, P.B., Noland, J., Spaeh, H., Klamerus, E., SchuUer, M., Response of Single-Story

Reinforced Masonry Shear WaUs to In-Plane Lateral Loads. January 1991. 136 pgs.

3.1(b)-l: Seible, F. and LaRovere, H., Summary of Pseudo Dynamic Testing. February 1987. 46

pgs.

3.1(c)-l: Merryman, K., Leiva, G., Antrobus, N., Klingner, R., In-Plane Seismic Resistance of

Two-Story Concrete Masonry Coupled Shear Walls. September 1989. 176 pgs.

3.1(c)-2: Leiva, G., Klingner, R., In-Plane Seismic Resistance of Two-Story Concrete Masonry

Shear Walls with Openings. August 1991.

3.2(a): Hamid, A., Abboud, B., Farah, M., Hatem, K., Harris, H., Response of Reinforced Block

Masonry WaUs to Out-of-Plane Static Loads. September 1989. 120 pgs.

3.2(b)-l: Agbabian, M., Adham, S., Masri, S., Avanessian, V. Traina, Out-of-Plane Dynamic

Testing of Concrete Masonry WaUs. Volumes 1 and 2, July 1989. 220 pgs.

4.1- 1: Limin, H., Priestley, N., Seismic Behavior of Hanged Masonry Shear Walls. May 1988.

119 pgs.

4.2- 1: Hegemier, G. Murakami, H., On the Behavior of Floor-to-Wall Intersections in Concrete

Masonry Construction: Part I: Experimental

4.2-2: Hegemier. G.. Murakami. H.. On the Behavior of Floor-to-Wall Intersections in Concrete

Masonry Construction: Part II: Theoretical

5.1-1: Porter, M., Sabri, A., Plank Diaphragm Characteristics. July 1990. 226 pgs.

527
5.2-1 Porter, M., Yeomans, F., Johns, A., Assembly of Existing Diaphragm Data. July 1990.

142 pgs.

6.2-1: Scrivener, J., Bond of Reinforcement in Grouted Hollow-Unit Masonry: A State-of-the- Art. June 1986. 53 pgs.

6.2-2: Soric. Z. and Tulin. L.. Bond Splices in Reinforced Masonry. August 1987. 296 pgs.

7.1-1 Paulson, T., Abrams, D., Measured Inelastic Response of Reinforced Masonry Building

Structures to Earthquake Motions. October, 1990. 294 pgs.

8.1-1: Hart. G.. A Limit State Design Method for Reinforced Masonry. Jime 1988.

8.1- 2: Hart, G., Expected Value Design in the Context of a Limit Sate Design Methodology. February 1990.

8.2- 1: Hart, G., Zorapapel, G.T., Reliability of Concrete Masonry Wall Structures. December

1991.

9.1-1: Kariotis, J.C., Johnson, A.W., Design of Reinforced Masonry Research Building. September, 1987. 42 pgs.

9.1-2: Kariotis, J.C., Waqfi, O.M., Trial Designs Made in Accordance with Tentative Limit

States Design Standards for Reinforced Masonry Buildings. February 1992.

9.2- 1: Seible. F.. Report on Large Structures Testing Facilities in Japan. September 1985. 120

pgs9.2.2: Seible, F., Design and Construction of the Charles Lee Powell Structural Systems

Laboratory. November 1986. 65 pgs.

9.2.3: Seible, F., The Japanese Five-Story FuU Scale Reinforced Masonry Building Test. January 1988. 100 pgs.

11.1-1: TCCMAR, Summary Report: U.S. Coordinated Program for Masonry Building

Research. September 1985 to August 1986. 190 pgs.

11.1-2: Stams Report: U.S. Coordinated Program for Masonry Building Research. November

1988. 170 pgs.

528
Appendix

Task Committees A-K Reports
I

I

i
Appendix: Task Committees A-K Reports

(A) STRONG-MOTION INSTRUMENT ARRAYS AND DATA

Date : May 19, 1992

Place: National Institute of Standards and Technology

Gaithersburg, MD

Attendees: U.S. Side - A. G. Brady (Chairman)

H. Meyers

P. E. Harben

(USGS)

(NOAA)

(DOE)

Japan Side S. Noda (Chairman)

K. Ohtani

K. Shioji

(PHRI)

(NIED)

(PWRI)

MISSION STATEMENT -

1. Objective :

The objective of the Task Committee is to coordinate the provision of strong-motion

earthquake data to researchers and to the practicing engineering community. The

Task Committee focuses on instrumentation; recording, processing, and analyzing

strong-motion data; research to ensure high-quality data dissemination; and analysis

of ground motion and structural motion in dynamic response to earthquakes.

2. Scope of Work

The T/C coordinates strong-motion research, processing strong-motion data, and

disseminating information on the recorded behavior of structures. In addition to

regular exchange of data and publications, the T/C's technical approach includes:

Planning and conducting T/C workshops and meetings, held generally in conjunction with UJNR Joint Meetings.

Conducting related technical sessions at professional society meetings.

Creating procedures for the dissemination of significant strong-motion

digital data with regard for the rights and expectations of (a) the owner,

(b) the users of data, and (c) the earthquake engineering community.

Creating an up-to-date information exchange on computer processing

procedures and related research so all users of processed data are aware

of valid ranges and practical limitations.

531
* Coordinating research progranns for appropriate instrumentation and

dynamic analysis of structures with base isolation or other responsereducing systems, under earthquake excitation.

* Coordinating research programs using both permanent and temporary

arrays and data derived to study site response.

3. Accomplishments

* Catalogs of strong-motion earthquake records observed both in the

United States and Japan are being exchanged. U.S. data are published

in the "Seismic Engineering Program Report", and the Japanese data are published in "Strong-Motion Records in Japan". Digital data is exchanged on tape, floppies, and CDROMs. Many other publications

also were exchanged.

* Preliminary Reports of the Southern and Northern California earthquakes,

from both the U.S. Geological Survey and California Division of Mines

and Geology, were exchanged.

* Prompt report of the Tokyo Bay Earthquake, February 2, 1 992, from the

Strong-Motion Earthquake Observation Council, NIED, was exchanged.

* From the USGS, a CDROM containing "Digitized Strong-Motion

Accelerograms of North and Central American Earthquakes 1 933-1 986",

May, 1992, by Linda Seekins was exchanged, together with a software

package for PC processing and a User Manual: "Basic Strong-Motion

Accelerogram Processing Software";Version 1 .0, May 1 992, by April M.

Converse.

* From the National Geophysical Data Center, NCAA, disks containing the

"Strong-Motion Data Catalog", and software (all records in the global

public domain through 1992) were exchanged. The User Manual is "An

Earthquake Strong-Motion Data Catalog for PC's, SMCAT" Version 2.0,

July 1990, by Lee Wesley Row III).

4. Future Plans

* Coordinate the analysis of records from arrays: downhole 3-D arrays

(geophysics and earthquake engineering) and structural arrays (structural

engineering).

* The Science and Technology Agency, and the National Science

Foundation will be asked to sponsor a joint workshop in Hawaii, in the

fall of 1993, on Digital Recording Technologies, for approximately 20

532
invited participants, including Japanese and US researchers. The

workshop will be for 2 days, followed by a technical site visit.

5. Infornnation Exchange

* Performed within the domestic community through technical meetings,

conferences, seminars, and workshops.

* Performed through participation in post-earthquake field investigations

with concentration in areas where strong-motion stations have recorded

data.

* Performed with the counterpart side through T/C workshops; visitor

exchanges, seminars, lectures; and participation in annual Joint Panel

meetings.

* The most crucial information exchange is rapid description of recovered

strong motion records (and subsequently, digital strong-motion data on

tape or disk) from significant earthquakes.

6. Impact

* Most managers who process strong-motion recordings use techniques

based, or closely associated with, the work of this T/C and its members.

Foreign networks fitting this category include:

AUSTRALIA - Australian Seismological Center, Bureau of Mineral

Resources, PO Box 378, Canberra ACT 2601.

CANADA - Pacific Geoscience Center, PO Box 6000, Sidney, BC. V8L

4B2.

ITALY - The Italian National Commission for Research and Development

on Alternative Energy Sources (ENEA), Rome.

MEXICO - Japan-Mexico Earthquake and Disaster Prevention Center,

Mexico City.

PERU - Japan-Peru Earthquake and Disaster Prevention Center, Lima.

TAIWAN - SMART Array

TURKEY - University of Bogazici, 80815 Bebek, Istanbul.

533
YUGOSLAVIA - Institute of Earthquake Engineering and Engineering

Seismology, Skopje.

Use of seismological and strong-motion data, in convenient form, is facilitated in seismically active countries having no such data of their

own.

Dissemination of digital data from significant earthquakes, taking into

account the conditions of the 3rd paragraph in the Scope of Work,

increases the research data base; application of this data in design

practice aids in reduction of earthquake hazards.

Barriers

Lack of adequate funding for T/C joint endeavors.

Establishing joint research efforts and maintaining close interaction

between participants are difficult unless visiting research appointments

are established. (Funding and staff availability are often barriers).

534
Report of Task Committee on

(B) LARGE-SCALE TESTING PROGRAM

Date: May 20, 1992

Place: National Institute of Standards and Technology

Gaithersburg, MD

Attendees: U.S. Side - H. S. Lew (Chairman)

W. E. Roper

(NIST)

(COE)

Japan Side - K. Ohtani (Chairman)

H. Yamanouchi

Y. Yamaguchi

(NIED)

(BRI)

(PWRI)

MISSION STATEMENT

1 . Objective :

The objective of the Task Committee is to develop performance data of full-scale

dynamic properties of structures to better resist wind and seismic loads through both

laboratory testing of prototype structures and field testing of structures in situ.

Improved full-scale test data will validate the results of small-scale model tests and

substantiate the results of computer analyses of structural behavior.

2. Scope of Work

The Task Committee develops its research agenda in coordination with other

appropriate T/Cs, such as T/C's "C" and "D", for full-scale evaluation of buildings, and

other structures except bridges.

Plans and conducts workshops and joint meetings to identify research

topics and develops joint research programs.

Coordinates research projects carried out by various laboratories in the

U.S. and Japan. Facilitates publication of research results and

implementation of findings in codes and standards.

Facilitates exchange of research personnel, technical information and

available testing facilities.

535
3. Accomplishments

* The second U.S. Technical Coordinating Committee on Precast Seismic

Structural Systems Program (PRESSS) was held in June 1991 at San

Diego, California.

* The Japan PRESSS Research program is continuing at several institutions

including BRI, University of Tokyo, Kyoto University, and Yokohama

National University. The Japan TCC Meetings were held during the past

fiscal year.

* The second U.S. -Japan Joint Technical Coordinating Committee on

PRESSS was held in October 1991 at Tsukuba, Japan.

* The Hybrid structural system is being considered as the fifth topic for the

joint research program following PRESSS. In preparation for the joint

research, the Japanese Planning Group on hybrid structures met five

times since September 1991.

* The first U.S. planning meeting on hybrid structures was held in San

Francisco, February 1 992. Several Japanese researchers participated in the meeting.

4. Future Plans

* The National Science Foundation (NSF) has funded the third phase(5-

story structural test) of the U.S. -side research of the U.S. -Japan

Coordinated Program for Masonry Building Research. The testing of a five-story full-scale masonry structure will be carried out at the

University of California, San Diego during 1992.

* The third Joint Technical Coordinating Committee (JTCC) on PRESSS

will be held in the U.S. in October 1992. The U.S. TCC on PRESSS will

meet prior to the third JTCC.

* In support of the International Decade for Natural Disaster Reduction

(IDNDR) program, techniques will be explored to disseminate findings of

Joint Research Projects carried out under the auspices of T/C "B" to

countries with high seismic risks.

* A joint U.S. -Japan workshop on hybrid structural systems is planned to

be held in September 1992 in the San Francisco bay area.

* Information exchange of following topics between appropriate U.S. and

Japan Organizations will be continued;

536
a. Application of high strength construction materials, such as concrete and steel, to structures in high seismic zones,

b. Testings of large-scale structures otherthan buildings and bridges.

* Because the dynamic response of large hydraulic structures is unique, it

is recommended that a study be conducted on the feasibility of creating

a new task committee on seismic behavior of these structures such as dams, navigation locks and pumping stations (see Appendix). A proposal

for creating a new Task Committee will be made to the full panel at the

25th Joint Meeting.

5. Information Exchange

* Technical reports and research documentation were exchanged with

participating organizations in both countries.

6. Impact

* The results of the past joint research projects on reinforced concrete,

steel and masonry structures have been published in both U.S. and

Japanese technical journals. A number of key findings have been

incorporated into U.S. and Japanese building codes and design

guidelines.

* As a result of joint research projects, the state of practice and

understanding in earthquake engineering has been improved in both

countries. The results from the joint program above further stimulated

research in both countries.

* Facilitate joint research projects utilizing available testing facilities in both

countries.

7. Barriers

* There are no major technical barriers for the attainment of the Task

Committee mission. However, different building practices and needs

require careful planning for effective joint research projects.

* Securing funding to support exchange of research personnel is difficult

in both countries. Funding agencies of both countries should be

informed about the technical benefits resulting from the joint panel

activities and encouraged to participate in researcher exchange and joint

research programs.

537
APPENDIX

The seismic safety issue of large hydraulic structures is important since the risk of a major disaster due to a failure of such structures increases each year with the growth

of population downstream.

Experimental and observed measures need to be compared to analytical procedures.

For example, the analyses for linear and nonlinear response need to be verified with

experimental and observed data. An exchange of details concerning analysis, design,

and construction practices can highlight the success and needs for further research.

538
Report of Task Committee on

(C) REPAIR AND RETROFIT OF EXISTING STRUCTURES

Date : May 19, 1992

Place : National Institute of Standards and Technology

Gaithersburg, MD

Attendees : U.S. Side: Ken Chong (T/C "C", Chair) (NSF)

Bob Fuller ( T/C "D", Chair) (HUD)

Jack Scaizi (T/C "D", Incoming Chair) (NSF)

Howard Nickerson (T/C "C") (Naval Facilities Engineering

Command)

KharaitiLal Abrol (T/C "C") (Department of Veterans Affairs)

Larry Hultengren (DOS)

John R. Hayes, Jr. (USACERL)

Chris Stoddart (Martin Marietta/DOE) (Observer)

Charles E. Anderson (Bureau of Reclamation)

Nicholas J. Carino (Visitor, NIST)

Ugo Morelli (FEMA)

Japan Side: Hajime Asakura (Temporary member) (ACTEC)

Hiroyuki Yamanouchi (Temporary Chairman "C&D") (BRI)

Yoshikazu Kitagawa (BRI)

MISSION STATEMENT

1 . Objective :

The objectives of the Task Committee are to encourage research and provide

technology transfer activities in materials, components, and whole buildings and

structural systems related to the repair and retrofit of existing buildings and other

structures. The scope will encompass studies related to materials of construction,

structural components as well as prototype structural systems.

2. Scope-of-Work

The T/C work includes information exchanges, planning, and hosting workshops and

seminars. Workshops are generally held in conjunction with the annual Panel's

meeting. They provide T/C members with an opportunity to review related research

and to direct plans for the future, share and implement results. The T/C conducts

related technical sessions at professional society's meetings. The T/C studies new

materials and methods to accomplish repair and retrofit operations on buildings; e.g.,

fiberglass reinforced plastics, high performance concrete, post-installed anchor bolts,

carbon-fibre wire, structural adhesives, and exploits use of automation and robotics

539
for repair and retrofit. Work focuses on coordinating research projects in U.S. and

Japan to minimize duplication and to maximize benefits.

3. Accomplishments

*Status reports were delivered on recent related NSF initiatives:

Quantitative Nondestructive Evaluation for Large Structural Systems . This

initiative was issued in June 1990. Plans are underway for the second

round in 1992. Interdisciplinary research teams are encouraged. Four

projects were selected for support in the first round. Details of these

projects and a description of the initiative are included in the proceedings

of the 23rd UJNR Joint Meeting in Tsukuba, Japan.

Repair and Rehabilitation Research for The Seismic Resistance of

Structures . In early 1990, NSF announced a new 5-year research

program that focuses on a coordinated research effort in the seismic

repair and rehabilitation of existing structures. During the first year of

the research program, FY 1990, NSF funded a total of nine projects.

In addition, cooperation and coordination efforts were initiated with other on-going

seismic repair and rehabilitation programs, conducted by the National Center for

Earthquake Engineering Research (NCEER), the Earthquake Engineering Research

Institute (EERI) through its Existing Building Committee, the Federal Emergency

Management Agency (FEMA), the Building and Fire Research Laboratory (NIST), The

California Seismic Safety Commission (CSSC), and other organizations. The first

coordination meeting of FY 1991 was held in Salt Lake City on February 13, 1991.

* 3 day ASCE seminars on Rehabilitation, Planning and Structural Repair of

Concrete and Masonry were held in, Oakland, CA (3/90); Boston (4/90); and

Washington D.C. (5/90).

* A workshop on Nondestructive Evaluation, (NDE) of Civil Structures and

Materials was held in October 1990, in Boulder, Colorado. Participants came

from many disciplines of engineering in academics and industries. A hardcover

proceedings (406 pages) is available from the University of Colorado,

Department of Civil Engineering.

* A follow-up workshop on NDE was held in May 1992, in Boulder, Colorado.

* Proceeding of the Workshop on Evaluation, Repair, and Retrofit of Structures , (NISTIR 4515), held at NIST in May 1990, was published by NIST and

distributed at the 23rd UJNR Joint Meeting in Tsukuba, Japan. Copies of the

proceedings were also distributed to authors, workshop participants, and other

researchers and practitioners in the U.S. Twenty-two papers were presented

540
at the NSF sponsored workshop, of which 17 were on repair and retrofit of

structures. Delegates totaled 24, 12 were from the U.S. and 12 from Japan.

In the U.S., FEMA published ATC 21, A Handbook for Seismic Evaluation of

Existing Buildings , and ATC 22 Techniques for Seismicallv Rehabilitating

Existing Buildings . The U.S. Department of Energy and U.S. Postal Service are involved in extensive repair and retrofit projects.

A report on "Retrofitting Buildings for Seismic Safety" was published by the

National Academy of Press in 2/91.

In Japan, the Guidelines for Evaluating and Strengthening Existing Buildings

have been revised, and the Guidelines for Evaluation of Earthquake Damage and

Repair Methods of Buildings was published. Both guidelines were available in August 1990, and are being implemented.

Several on-going repair/retrofit projects at the US Army Construction

Engineering Research Laboratory (CERL) include:

o Repair/Retrofit of Non-Ductile Reinforced Concrete Frames. Shake table

tests of beam-column subassemblages and scaled frames, both as-built

and as upgraded using viscoelastic dampers in frame bracing.

o Biaxial Response of Reinforced Concrete Frames, shake table tests of in- plane vs out-of-plane response of reinforced concrete frames with

masonry in-fills. Future research to retest damaged walls after postearthquake repair.

0 Strengthening of Precast Concrete Panel Connection Details Planned

FY93 new start project.

Future Plans

A second round of related NSF initiatives are planned for FY 92:

Quantitative Nondestructive Evaluation for Constructed Facilities.

Repair and Rehabilitation Research for the Seismic Resistance of

Structures.

NSF/SHRP Initiative on Structural Control

Advanced Polymeric Materials (NSF/EPRI)

The feasibility of establishing a Civil Infrastructure Systems Research Program

is being studied at NSF.

541
Supports NAE/NRC on a study of infrastructure research for 21st century.

New initiative on innovative steel research for construction, NSF with AISI and

AISC.

Repair and retrofit of existing structures are being studied in association with

a possible federal initiative on adaptation due to the global changes (MARS)

program.

New initiatives are being developed by NSF to focus on central and eastern

U.S. existing structures/buildings.

In support of IDNDR activities, T/C "C" plans to participate in task committee

activities related to repair and retrofit of damages due to various disasters.

The T/C "C" will consider joint support of a proposed U.S. -Japan workshop on

"Smart and High-performance Material/Systems", by T/C "G". The workshop

will possibly be conducted in 1993 in conjunction with the 25th Joint UJNR

panel meeting in Tsukuba, Japan.

A possible session on wind damages will be planned in the future in the U.S.

The role of testing in support of evaluation of existing structures,

characterization of the current conditions and qualification of retrofit

approaches should be carefully examined and might be a possible topic for a future workshop.

In Japan, BRI is considering to propose a new comprehensive research program

on new structural design methodology based on the desired structural

performance.

In Japan, state-of-the-art technology for diagnosis, repair and retrofitting of

damaged structures will be introduced and discussed at the IDNDR International

Symposium on Earthquake Disaster Reduction Technology in commemoration

of the 30th Anniversary of USEE.

Information Exchange

Published workshop reports and individual research papers; they were

exchanged with the U.S. and JAPAN-side members.

Numerous papers presented at the 24th UJNR Joint Panel Meeting are of

interest to T/C "C".

542
Impact

Contributed to development of U.S. and Japan design criteria for repairing and

strengthening reinforced concrete, steel, and other buildings.

T/C recommendations were incorporated into building codes and professional

practices, as illustrated by rehabilitation programs used by the City of Los

Angeles to upgrade existing masonry buildings to resist earthquake forces.

Influenced changes to U.S. and Japan design practices to upgrade existing

structures.

Barriers

Greater progress needs to be made to implement available research results.

Additional funding sources need to be identified to support technical activities

to support committee missions and to support participant attendance at Joint

Panel Meeting workshops.

543
Report of Task Committee on

(D) EVALUATION OF STRUCTURAL PERFORMANCE

Date : May 19, 1992 (Joint Meeting with T/C "C")

Place : National Institute of Standards and Technology

Gaithersburg, MD

Attendees : U.S. Side: Ken Chong (T/C "C", Chair) (NSF)

Bob Fuller ( T/C "D", Chair) (HUD)

Jack Scaizi (T/C "D", Incoming Chair) (NSF)

Howard Nickerson (T/C "C") (Naval Facilities Engineering

Command)

Kharaiti Lai Abrol (T/C "C") (Department of Veterans Affairs)

Larry Hultengren (DOS)

John R. Hayes, Jr. (USACERL)

Chris Stoddart (Martin Marietta/DOE) (Observer)

Charles E. Anderson (Bureau of Reclamation)

Nicholas J. Carino (Visitor, NIST)

Ugo Morelli (FEMA)

Japan Side: Hajime Asakura (Temporary member) (ACTEC)

Hiroyuki Yamanouchi (Temporary Chairman "C&D")(BRI)

Yoshikazu Kitagawa (BRI)

MISSION STATEMENT

1 . Objective :

The objectives of this Task Committee are to develop disaster mitigation policies and

programs to improve the capacity of existing structures to resist wind and seismic

forces. To provide adequate performance evaluation, each country will coordinate

development of condition assessment, screening, and structural analysis

methodologies. Structures will be analyzed and instrumented, and then evaluated

after disasters.

2. Scope of Work

* Develop an inventory of "benchmark" structures that have been analyzed

and instrumented to measure wind and seismic responses.

* Develop a uniform system for screening and analyzing wind and seismic

resistance capacity of structures in each country.

544
* Develop sensor technology, instrumentation, and "expert" systems to

provide condition assessment of existing structures.

* Evaluate masonry and precast concrete buildings, and seismic base

isolation systems; include results in each country's catalog of

"benchmark" structures for post-disaster performance evaluation.

* Conduct workshops at future UJNR Joint Panel Meetings, cooperatively

with T/C "C".

Accomplishments

* TCCMAR Reports by Englekirk and Hart were published in Feb. 1992.

Copies were furnished to the Japanese side.

* Further status reports on recent U.S. legislation were presented at the

24th Joint Panel Meeting: 1990 National Affordable Housing Act . Sec.

947 on Seismic Safety Standards. National Earthouake Hazards

Reduction Program Reauthorization Act: Public Law 101-614. Nov. 16.

1 990 . Sec. 8 on Seismic Standards.

* Information was exchanged on the activities of each country in the

International Decade on Natural Disaster Reduction (IDNDR) , pertaining

to winds and earthquakes. A report entitled: Hazard-Proofing the Nation.

Strateav to Reduce Natural Disasters was developed by the

Subcommittee on Natural Disaster Reduction, a part of the U.S. Federal

Coordinating Council for Science and Technology, and printed in June

1992. Included in the research and implementation strategies are risk

assessment and mitigation programs. Copies will be mailed to the

Japanese members.

* Status reports of several FEMA - ATC projects were discussed, including

those on seismic risk assessment of existing structures. FEMA is planning a Loss Assessment Research Project to be conducted over the

next several years.

Future Plans

* Develop a reporting system and data collection form for building

inventory and performance standard, and compile a computer data file for exchange between countries. Report at each Joint Panel Meeting.

* Continue to encourage participation by private industry, consulting

engineering firms, universities, funding agencies and State and local

government agencies involved in instrumentation, evaluation, and

condition assessment of existing buildings.

545
* Monitor development of sensor technology and instrunnentation of

structures for condition assessment, and to measure long-term aging

effects on building components.

* Participate in coordination of Joint Panel activities in the International

Decade for Natural Disaster Reduction (IDNDR), pertaining to structural

performance evaluation.

* A workshop on Structural Performance Evaluation was discussed to be

held in Japan 1993. The workshop will place emphasis on NonDestructive Evaluation (NDE).

5. Information Exchange

* Technical documents exchanged, and pertinent Japanese papers

delivered at the 24th Joint Panel Meeting.

* Japanese side will send to the U.S. T/C Chairman a copy of the "Design

Guidelines for Medium Rise Reinforced Masonry Buildings" developed by

TCCMAR-Japan. U.S. Side will explore feasibility of translating this

manuscript into English.

6. Impact

* Enables the development of revised standards and building code

provisions to prevent destruction from high winds and earthquakes.

* Provides engineers with the capability to conduct condition assessment

of existing structures for the purpose of developing cost effective repair,

retrofit, strengthening or demolition programs.

* Will increase the confidence level of the public, regulators, policy makers

and engineers, to be assured that the majority of major structures will

perform adequately to resist high wind and earthquake forces.

7. Barriers

* Insufficient resources to document load-carrying capacity of existing

buildings subjected to probable extreme winds and earthquakes.

* Difficulty in developing probability risk maps for tornadoes, typhoons,

and hurricanes.

* Limited resources for instrumenting buildings, for conducting postdisaster evaluations, and for participating in foreign travel for

coordination of activities.

546
Report of Task Committee on

(E) GROUND MOTION AND SEISMIC DESIGN FORCE

Date : May 19, 1992

Place: National Institute of Standards and Technology

Gaithersburg, MD

Attendees: U.S. Side - Walter Hays, (Acting Chairman)

Jim Hill Jeff Kimball

Mike Mahoney

Mehmet Celebi

(USGS)

(DOE)

(DOE)

(FEMA)

(USGS)

Japan Side Kazuhiko Kawashima (Chairman)

Keiichi Ohtani

(PWRI)

(NIED)

MISSION STATEMENT

1 . Objective :

The objectives of this Task Committee are to develop techniques and exchange

information and experiences on: 1 ) engineering characterization of ground motion, 2)

mapping techniques for displaying ground motion data, 3) microzonation, 4) siting and

design applications, and 5) earthquake hazard risk assessment.

2. Scope of Work

Plan and conduct cooperative research, T/C workshops and meetings, These

meetings and workshops are generally held in conjunction with the annual Panel

Meeting on Wind and Seismic Effects. Activities include:

- Conduct cooperative research programs

- Exchange research personnel

- Exchange information on a timely basis

- Cooperate in contributing to the goals of the IDNDR

3. Accomplishments

The First U.S. -Japan Workshop on Hazard/Risk Assessment and Design Earthquake

Loading was sponsored by the Task Committee E, and was held on May 28-29, 1 991

547
at the Public Works Research Institute, Tsukuba, Japan. Four presentations from U.S.

-side and ten presentations fronn Japan-side were made. During the technical site

visits, tours to the sites of large bridge constructions were made and technical

discussions on seismic design loading were conducted.

4. Future Plan

Plan the 2nd Workshop on "Ground Motion and Seismic Design Force" with emphasis

on subduction zone earthquakes. This Workshop will be held in Japan in May 1993.

Three themes will be developed: DData analysis and modeling, 2) Engineering

characterization of ground motion and ground strain, and 3) Siting and design

applications. Topics will include: attenuation of ground motion with magnitude and

distance; linear, nonlinear, and energy response spectra for rock and soil; prediction

of strong motion; probabilistic representation; uncertainty; types and importance of

structures; soil structure interaction; scenario earthquakes; loss estimation; and zoning

maps for siting, design and construction.

5. Information Exchange

Technical reports and research documentation were exchanged with participating

organizations in both countries.

6. Impact

Task committee efforts influence the national seismic codes concerning input seismic

forces.

7. Barriers

Lack of funding for travel to conduct workshops and related meetings has been the

principal barrier in the past. To increase benefits to funding agencies of each country,

a study tour is being planned.

548
Report of Task Committee on

(F) DISASTER PREVENTION METHODS FOR LIFELINE SYSTEMS

Date : May 20, 1992

Place : National Institute of Standards and Technology

Gaithersburg, Maryland

Attendees : U.S. Side - R. Dikkers (Chairman) (NIST)

(FHWA)

(FHWA)

(NIST)

(DOE)

H. Bosh

J. Cooper

R. Chung

J. Hill P. Harben (LLNL-DOE)

Japan Side - K. Kawashima (Acting Chairman)

0. Nakano

H. Asakura

(PWRI)

(HDB)

(ACTC)

MISSION STATEMENT

1 . Objective ;

The objectives of Task Committee "F" are to improve: 1) the behavior of lifeline

systems during earthquakes and 2) engineering and other seismic countermeasures

such as damage estimation techniques and inspection procedures. These technologies

provide users with improvements to existing standards of practice for safety and

serviceability of lifeline systems. Lifeline systems such as gas, oil, water, and sewage

pipelines; and power, communication, and transportation systems are crucial to the

survival and health of a city or community. Earthquakes affecting such systems cause

severe social and economic disruptions to communities and cause human suffering to

the residents.

2. Scope of Work

* Plan and conduct T/C workshops and meetings. The purpose is to exchange

the state-of-the-art knowledge and practice, and to identify key cooperative

opportunities for exchange and studies which effectively implement research

results.

* Conduct related technical sessions at professional society meetings.

* Develop research programs on selected lifeline systems (similar in concept to

the Large-Scale Testing Program). For example, through joint demonstration,

produce criteria for emergency design and loss evaluation and on-line

549
monitoring systenns for lifeline operation.

* Conduct special activities such as translating, printing, and distributing MOC

PWRI's publication Manual for Repair Methods for Civil Engineering Structures

Damaged bv Earthguakes .

* Develop performance standards and a manual on repair, restoration, and retrofit

of lifelines.

3. Accomplishments

* Conducted the 4th Joint Workshop on Earthquake Disaster Prevention for

Lifeline Systems on August 19-21, 1991, in Los Angeles, California.

* Contributed technical papers on lifeline systems at the 24th joint panel meeting.

* Participated in promoting the goals of the IDNDR.

* Developed a seismic design manual for underground structures (Japan Side).

4. Future Plan

* Conduct the 5th Joint Workshop on Earthquake Disaster Prevention for Lifeline

Systems on October 26 - 27, 1 992, in Tsukuba, Japan. A technical study tour

will follow the technical sessions.

* Encourage current efforts in both countries to establish seismic design

guidelines and standards for lifeline systems. Existing UJNR channels should

be fully utilized to facilitate the exchange of relevant information concerning

standards development. Possible collaboration of joint development lifeline

systems standards and guidelines should be pursued.

* Continue observation of instrumented buried pipes in Parkfield, California, and

Chiba, Japan, and exchange data.

5. Information Exchange

* Domestic community: through technical meetings, conference sessions,

seminars, and workshops.

* Between U.S. and Japan: through T/C Workshops and annual Joint Panel

Meetings.

* Visits by researchers and participation in technical meetings sponsored by

organizations such as NCEER, EERI and ASCE.

550
* Short-term (3 to 12 months) researcher exchanges to participate in on-going

lifeline research.

6. Impact -

* Contributed to the increase of field engineers' technical knowledge on repair

methods, through seminars with lectures by T/C members using the MOC

Manual n Japan: more than two thousand participants attended seminars.

* Contributed to the development and assessment of design, construction, and

operating standards for lifeline earthquake engineering (TCLEE/ASCE) via T/C

members' participation in committee work.

7. Barriers

* Lack of sufficient funds to conduct annual Task Committee Workshops and to

publish proceedings and reports.

551
Report of Task Committee on

(G) PASSIVE, ACTIVE, AND HYBRID CONTROL SYSTEMS

Date : May 20, 1992

Place : National Institute of Standards and Technology

Gaithersburg, MD

Attendees : U.S. Side - S. C. Liu (Chairman) (NSF)

Jack Hayes (CERL)

Chris Stoddart (Martin Marietta/DOE)

Richard Sause (Lehigh Univ.)

Japan Side - Y. Kitagawa (Chairman) (BRI)

M. Fujiwara (PWRI)

H. Yamanouchi (BRI)

MISSION STATEMENT

1 . Objective :

The objectives of the Task Committee are to: (1 ) develop research plans in passive,

active, and hybrid control of structures and equipment, (2) implement such control

techniques as computerized motion reduction or modification devices/systems in

actual design and construction of structural engineering facilities under seismic or wind environments, (3) promote U.S. -Japan cooperation in structural control research

(4) bring together governmental, academic, and industrial participants in joint pursuit

of such efforts, and (5) contribute to IDNDR by organizing joint research and other

technical activities in structural control on the basis of international cooperation.

2. Scope of Work

The UJNR T/C "G" works closely with other organizations to provide the leadership

in control research efforts by facilitating the exchange of technical data and

information through the established UJNR mechanisms. Work includes:

i) Providing technical assistance, consultation and coordination to UJNR

affiliated research agencies in the initiation, development, and execution

of their programs in structural control research.

ii) Promoting joint government-university-industry collaborative efforts to

facilitate technology transfer and practical implementation.

552
iii) Sponsoring and conducting multi-disciplinary workshops and meetings

to identify key areas of research and opportunities for cooperation, and

to exchange new knowledge and experience in practice.

iv) Developing promotional and demonstrative activities to stimulate public

awareness and interest in this emerging field of research.

v) Establishing performance standards, design specifications, guidelinesand

code recommendations for application in new construction as well as

retrofit/rehabilitation of existing structures with control systems.

vi) Initiating research in new intelligent materials, high-tech sensors, optimal

control systems design, and laboratory and field experiments of

prototype and full-scale structures.

3. Accomplishments

i) The U.S. Panel on Structural Control Research worked closely with its counterpart Japanese Panel in exchanging technical information and

developing joint research plans.

ii) In cooperation with T/C "J", a "U.S. -Japan Workshop on Earthquake

Protective System (Passive Control System)" was held at NCEER,

Buffalo, New York, on September 4 and 5, 1991. A study tour to

inspect base-isolated bridges across the U.S. was conducted by

participants prior to and following the Workshop. I. G. Buckle of NCEER

and UJNR Panel members S. C. Liu, M. Shinozuka, K. Kawashima, and

61 researchers from the U.S., Japan, and Taiwan participated in the

technical activities. Proceedings of the Workshop is being published by

NCEER.

iii) The "Japan National Symposium/Workshop on Active Structural

Response Control" was held on March 23-25, 1991, in Tokyo, Japan.

The Symposium was organized by Japan National Working Group on

Active Structural Response Control, Earthquake Engineering Research

Liaison Committee, Science Council of Japan (SCJ). UJNR Panel

members S. C. Liu, M. Shinozuka, Y. Kitagawa, and K. Kawashima

participated in the Symposium's technical activities. The workshop

identified three major areas of research in structural control. They are control theory, devices and systems, and structural design. Preliminary

Proceedings of the Symposium was published by SCJ.

iv) The "International Workshop on Recent Development in Base Isolation

Techniques for Building" organized by the Architectural Institute of Japan

(AIJ) was held in April 1992, in Tokyo, Japan. UJNR panel members Y.

Kitagawa and K. Kawashima participated in the Workshop's technical

553
activities. The Workshop identified four specific topics for future efforts:

design concept and code requirements, estimation of seismic input and

structural response, behavior of isolators and dampers, and practical

applications and field observations. Proceedings of the International

Workshop was published by AIJ.

v) A trilateral U.S. -Italy-Japan Workshop/Seminar on Intelligent Systems

was held in Perugia, Italy, June 27-29, 1991. Investigators from the

U.S., Italy, Japan, and other Western European countries gave

presentations of papers and participated in group discussions. UJNR

members S. C. Liu, H. J. Lagorio, Ken Chong, William Anderson, etc.,

contributed papers and H. J. Lagorio participated in the meeting.

vi) Under the NSF initiative on "Structural Control Research for Safety,

Performance, and Hazards Mitigation," sixty seven (67) proposals were

submitted by the U.S. research community. Approximately ten (10)

proposals are being selected for funding during the first year (FY 92) of

the 5-year program.

Future Plans

i) In cooperation with T/C "J", a "Second U.S. -Japan Workshop on Passive

and Active Control of Bridges Against Earthquakes" is scheduled for

December 1992 at PWRI, Tsukuba. A study tour to inspect baseisolated bridges across Japan is being planned for the participants prior

to and following the Workshop (Workshop Co-Chairmen: Ian

Buckle/NCEERand K. Kawashima/PWRI).

11) In cooperation with T/C "J", support of development and implementation

of innovative base-isolation technology for bridge structure will continue.

ill) Specific Task Committee work plans are being developed for coming

year (1992-1993) and for the next five years.

iv) Research cooperation between the U.S. and Japan will be pursued using

such channels as the UJNR and other inter-governmental mechanisms as

the U.S. and Japan Panel on Structural Control. Specific plans to

promote "The U.S. -Japan Cooperation Research Project" on active and

hybrid control systems are being studied by NSF, BRI, and PWRI.

v) Joint research projects on hybrid control of bridges are on-going by

PWRI, NIST, FHWA, NCEER, and University of California, Irvine.

vi) The Construction Engineering Research Laboratory (CERL) will initiative

work in July 1992 on active control of building structures using shape

memory alloy. UJNR member Jack Hayes is project coordinator of this

554
research.

vii) A joint U.S. -Japan Workshop on "Smart and High-Perfornnance Materials

Systems" will be planned in the near future. If possible this workshop

shall be in conjunction with the 25th Joint UJNR panel meeting in

Tsukuba, Japan. T/C "C" will be invited to join and help sponsor this

Workshop.

viii) Joint research projects on active and hybrid control of buildings are being

developed by BRI, NCEER and University of California, Irvine.

ix) At BRI, research programs on structural control of buildings have been

developed and implemented under the direction of UJNR member Y.

Kitagawa. Shaking-table tests of scaled control models will be carried

out in January 1993. Collaborative research program involving possible

organizations and private companies are being developed.

5. Information Exchanges

Technical materials on structural response and control systems were exchanged.

6. Impact

The T/C "G" has stimulated research communities from both sides to initiate multidisciplinary studies and cooperative investigation in structural control.

7. Barriers

No technical barriers are identified at the present time. However, lack of sufficient

financial backing from both sides may present difficulties in achieving the objectives

of T/C "G".

555
Report of Task Committee on

(H) SOIL BEHAVIOR AND STABILITY DURING EARTHQUAKES

Date : May 19, 1992

Place: National Institute of Standards and Technology

Gaithersburg, MD

Attendees U.S. Side - A. G. Franklin (Chairman)

M. G. Katona

(WES)

(AFCEL)

Japan Side - K. Yokoyama (Temp. Chairman)

Y. Yamaguchi

J. Koseki

(PWRI)

(PWRI)

(PWRI)

MISSION STATEMENT

1 . Objective :

Government agencies with responsibility for public works have the need to assure

their seismic safety and to provide economical protection against earthquake hazards.

The objective of this Task Committee is to assist in meeting those needs by enhancing

the available technology for predicting the dynamic behavior of soils, analyzing

dynamic soil-structure interaction, and modifying the earthquake behavior of

foundations and earth structures in order to assure their safe performance during

earthquakes.

2. Scope of Work

Present technical papers at annual joint panel meetings on technological

developments and on state-of-the-art and practice related to soil behavior

and stability during earthquakes.

Exchange information and technical data relating to field performance,

research, and methods of practice.

Plan and conduct T/C workshops in coordination with proposed or on- going cooperative research programs.

The Task Committee plans and conducts programs of cooperative

research, exchange of researchers between U.S. and Japanese research

institutions, publication of research results and recommended practice.

The benefits realized from available research funds can be amplified by

556
bringing greater resources to bear on problems of nnutual interest, and

by coordinating research tasl<s to make the best use of the respective

strengths of the two sides' research capabilities.

Exchange visiting researchers using mechanisms provided by formal and

informal cooperative research programs. Such exchanges promote

technical information exchanges, familiarization with methods of practice

on the respective sides, and mutual understanding and cooperative

relationships between researchers on the two sides.

Accomplishments

U.S. researchers presented 5 papers and Japanese researchers presented

10 papers, on soil behavior and stability during earthquakes, at the 23rd

and 24th Joint Panel Meetings. 5/91, 5/92

The International Workshop on Remedial Treatment of Potentially

Liquefiable Soils was conducted at PWRI in Tsukuba from 22 to 25

January 1991. Nineteen papers were presented. 1/91

Mr. Ken-ichi Tokida, the T/C (H) Japan Chairman, visited St. Louis and

met Mr. R. H. Ledbetter of the U.S. Army Engineer Waterways

Experiment Station and discussed the Proceedings of the International

Workshop held in January and plans for the next Workshop. 3/91

Proceedings of the International Workshop on Remedial Treatment of

Potentially Liquefiable Soils was published and distributed to the

members of Task Committee (H) and participants of the Workshop.

3/91, 5/91

The English language report on the expert team of Japan Disaster Relief

Team (JDR) on the earthquake in the Philippines of July 16, 1990, was

provided to the U.S. side. 5/91

T/C (H) has relinquished its responsibility for consideration of marine

structures subject to earthquakes and the measurement of sea floor

earthquake motions to the new Task Committee (K) on Wind and

Earthquake Engineering for Offshore and Coastal Structures, recognized

at the 23rd Joint Panel Meeting. 5/91

Mr. Jun-ichi Koseki of the PWRI is a one-year Guest Researcher at

Massachusetts Institute of Technology, working on Liquefaction Risk

Analysis. 8/91

Dr. Richard H. Ledbetter, of the U.S. Army Engineer Waterways

Experiment Station, visited Japan in September 1991. Information

557
exchange and fundamental discussion to start the U.S. -Japan joint

research through T/C "H", on remedial treatment of liquefiable soils were

conducted, and fundamental subjects were agreed upon. 9/91

* Mr. Yoshikazu Yamaguchi of the PWRI is a one-year Guest Researcher

at the U.S. Army Engineer Waterways Experiment Station where he is studying water seepage in dam foundations. 12/91

Future Plans :

* The Waterways Experiment Station has initiated a new research project

"Repair, Evaluation, Maintenance, and Rehabilitation Research (REMR)"

(for the U.S. fiscal-years 1992 - 1996). 10/91

* The Public Works Research Institute and the Building Research Institute

have initiated a new research and development project "Research and

Development on Earthquake Disaster Prevention in Large Urban Area"

(for the Japan-fiscal years 1992 - 1996). 4/92

* The National Research Institute for Earth Science and Disaster

Prevention, the Building Research Institute, and Wayne State University

have initiated the cooperative research "Research on Earthquake Pile

Resistance in Liquefied Soils". 4/92

* Identification of research tasks on preventive measures against soil

liquefaction that are of common interest, in accordance with the results

of the planning meeting held at Tsukuba in May 1 989 and at Yokohama

in September 1991 . 9/92

T/C (H) proposes to hold the next U.S. -Japan Workshop on Remedial

Treatment of Potentially Liquefiable Soils in Japan, in October 1993.

10/93

T/C (H) at the 22nd Joint Panel Meeting, discussed the need for

technological innovation and research to assist in saving lives following

natural disasters caused by earthquakes. The principal areas of need

were identified as improved technology for locating survivors and the

development of new equipment and systems for safe excavation of

debris to remove survivors from collapsed structures. T/C (H) recognizes

this subject is outside their area of responsibility but feels it is important

and should be discussed by the full Joint Panel.

558
The Task Committee suggests the following options for implementation:

(1) Explore a new Task Committee which will address this

topic area and provide the focal point to share experience

and research in this area.

(2) Incorporate this topic into another existing T/C.

T/C (H) will continue to explore methods whereby this topic will receive

appropriate attention.

5. Information Exchanoe :

* T/C workshops, exchange of visitors, participation in annual joint panel

meetings, and presenting technical papers.

* Exchange of researchers between U.S. and Japanese government

laboratories.

* Visits by researchers and participation in technical activities such as post-earthquake field investigation.

* Fostering professional community contacts in the U.S. and Japan,

publication of papers in journals, publication of research reports by the

respective agencies, participation in professional society meetings and

conferences, and workshops.

6. Impact

* Through the exchange of researchers and the diffusion of earthquake

related technical data, experience, and information on methods of

practice, the state of technology in geotechnical earthquake engineering

has been raised on both sides.

* Through the cooperative research program on in-situ testing of soils

which was carried out after the 1983 Nihonkai-Chubu Earthquake,

improved and more accurate methods of using in-situ tests for evaluating

the liquefaction potential of soils were achieved: these have been in general use in the United States since 1985. The Loma Prieta

Earthquake stimulated renewed awareness of the need for further

research in soil liquefaction and amplification effects of soft soils.

* Through the workshops on remedial treatment of potentially liquefiable

soils held in 1 988 and 1 991 , beneficial information on recent technology ~ were exchanged and contributed to progress future research of both

U.S. -and Japan-sides.

559
Barriers

Insufficient U.S. and Japan Government funds allocated to adequately

address earthquake engineering research.

Department of Army administrative limitation on foreign travel.

560
Report of Task Committee on

(I) STORM SURGE AND TSUNAMI

Date : May 20, 1992

Place: National Institute of Standards and Technology

Gaithersburg, MD

Attendees : U.S. Side - H. Meyers (Chairman)

W. E. Roper

(NOAA)

(COE)

Japan Side - K. Yokoyama (Acting Chairman)

K. Shioji

(PWRI)

(PWRI)

MISSION STATEMENT

1 . Objective :

The objective of this Task Committee is to mitigate damages from storm surges and

tsunamis through shared technologies, research, information sharing, and cooperative

work. Storm surge and tsunami are hazards capable of inflicting damage of disastrous

proportions. Storm surges are associated with hurricanes and typhoons where high

winds and the mounding of water under the low barometric pressure of the storm's

eye can cause extensive flooding and severe wave action. These often can be tracked

for days in advance of a landfall but the location of the landfall center is difficult to

predict. Tsunamis are predominately caused by underwater earthquakes and, to a lesser extent volcanic and land slide activity. Depending on the distance from the

source, tsunamis arrive within minutes to up to a day after generation.

2. Scope of Work:

Exchange results of research on storm surge and tsunami occurrence,

generation, propagation, and coastal effects. This includes observations

on historical, current, and theoretical tsunamis. Of particular interest is the effort by U.S. and Japan to acquire deep ocean tsunami

measurements.

Exchange results and status of anti-storm surge and tsunami activities

including analysis of the problem, planning, warning, and engineering

approaches.

561
* Exchange information on planned and on-going projects relating to storm

surge and tsunamis.

* Exchange information on development of technologies such as computer

programs to predict travel times, land-fall locations, run-up heights, and

wave characteristics and analysis; improved instrumentation, and use of

satellite communications for detection and warning.

* Facilitate dissemination through exchange of literature, technical reports

at joint meetings, special workshops, joint projects, and direct interaction

among participants.

3. Accomplishments

* Copies of preliminary marigrams from the April 25, 1992 Northern

California earthquake and tsunami were distributed to the Japanese side.

* The U.S. side has compiled detailed information on all tsunamis affecting

the U.S. and its possessions. The publication, "United States Tsunamis

1690-1989" has been printed and given to the Japanese side.

* A Second Tsunami Workshop was held on November 5-7, 1990 in Honolulu, Hawaii. Nine members participated from the Japanese side

and about 20 from the U.S. side. Site visits were made to Hilo as part

of the Workshop. Proceedings were published and were sent to all members.

4. Future Plans

* Develop exchanges of data and information on activities related to

tsunami measurements including deep ocean measurements. Deep

ocean bottom gages will be recovered in June 1992 from the ocean

bottom stations which are expected to have recorded the small tsunamis

for the April 25, 1 992, Northern California earthquake and tsunami. This

event was recorded by Hawaiian and U.S. West Coast shore based

marigraphs. The seafloor topography of the source area will be re- surveyed. These will be shared with the Japan side.

* Develop activities and information exchanges to support the International

Decade of Natural Disaster Reduction.

* Explore and promote joint undertakings such as tsunami gauge testing,

scientist exchange, satellite communication for tsunami warning, and

assistance in contacts within the U.S. and Japan.

562
* Increase the participation of storm surge experts in the Panel and T/C

meetings.

* The U.S. and Japanese sides agree that there is a need to hold

worl<shops at about 2 or 3 year intervals to exchange the latest research

and technology relating to various aspects of storm surge and tsunami

hazards. A third workshop including both storm surge and tsunamis is planned for August, 1 9-20, 1 993 in Tskuba, Japan, near the time of the

International Union of Geodesy and Geophysics (lUGG) Tsunami

Commission Symposium. The Workshop will include about 10 academic

researchers and some additional Government Task Committee members

from the US side and similar numbers from Japan. There will be about

an even split between the storm surge and tsunami problems.

* 1 996 will be the 1 00th anniversary of the great Sanriku Earthquake and

tsunami in Japan and the 50th anniversary of the April 1 , 1 946 Aleutian

earthquake and tsunami. These events led to the creation of warning

systems and much of the current activity in this field. The Task

Committee is beginning to plan for a Joint Commemorative Symposium

or other observance.

* The U.S. side is completing a detailed study of tsunamis affecting the US

West Coast including copies of all available marigrams. This will be

published in the fall of 1992 and shared with the Japan side. Also, at

the request of UNESCO-ITSU, the U.S. side will publish a catalog of

tsunamis affecting the West Coast of Mexico compiled by Sanchez and

Farreras, the Mexican authors.

5. Information Exchange

* Continue exchange of publications, bibliographic data, technical reports,

and personal contacts.

6. Impact

* Increased cooperation among scientists and engineers working in these

fields.

* Hasten implementation of new modeling and design techniques

developed by the U.S. and Japan.

7. Barriers

* Insufficient funding from government and private sector organizations to ~~ participate in T/C workshops and the Panel's annual joint meetings.

563
The tsunami problem is less severe in the U.S. than in Japan which leads

to a smaller community of researchers. There are few U.S. researchers

focusing on tsunamis at any specific location; thus highlighting the

importance of T/C periodic meetings and workshops.

There are different approaches and degree of significance in dealing with

problems in the U.S. and Japan. For example, in the U.S. much of the

hazard is from remote sources which leads the government to emphasize

developments of warning and evacuation schemes. In Japan the

engineering approach is needed because there are more locally generated

tsunamis which often do not provide sufficient times for adequate

warning and evacuation. This barrier is partially being addressed by

Inviting U.S. engineers and non-government specialists to participate

during the annual meetings.

The storm surge research community is less well organized and lacking

the specialty societies, journals, and international organizations to define

and stimulate work in the field. The Task Committee (I) can help fill this

gap.

564
Report of Task Committee on

(J) WIND AND EARTHQUAKE ENGINEERING FOR TRANSPORTATION SYSTEMS

Date : May 19, 1992

Place : National Institute of Standards and Technology

Gaithersburg, MD

Attendees : U.S. Side - J. D. Cooper (Chairman) (FHWA)

C. Barrientos (NOAA)

R. Dikkers (NIST)

J. Scaizi (NSF)

H. S. Lew (NIST)

Japan side - M. Fujiwara (Chairman) (PWRI)

0. Nakano (Hokaido Development Bureau)

MISSION STATEMENT

1 . Objective :

The objectives of this Task Committee are to plan, promote, and foster research on

the behavior of highway bridges when subjected to wind and seismic forces and to

disseminate research results and provide specifications and guidelines based on the

Task Committee's findings. Surface transportation systems play a vital role in the

movement of goods and people. Highway bridges are especially influenced by the

forces of wind and earthquakes because of their open exposure to those forces.

2. Scope of Work:

The scope of work is applicable mostly to highway bridges without limitation on their

size and function including: existing bridge and new bridge designs; whole system of

bridges; and/or single components of a bridge.

The mission is performed through:

* Conducting workshops, exchanging researchers, developing methods for

design evaluations and test procedures, developing inspection

techniques, rehabilitation and maintenance specifications and policies,

and performing cooperative research programs and other relevant

cooperative administrative activities.

565
3. Accomplishments (past 2 years)

* The Japanese side hosted the first U.S. -Japan Workshop on Seismic

Retrofit for Bridges under the auspices of T/C "J" December 17 to 22,

1 990 to discuss and review practical seismic retrofit applications in both

countries. Nineteen technical presentations were made by 5 U.S.

researchers and 30 Japanese researchers. Technical topics include the

history of developing retrofit techniques for bridges; damage assessment

from the Loma Prieta earthquake; discussion on condition assessment

and prioritization of bridges for seismic retrofit; seismic strengthening

methods; and research in progress. Site visits were made to view

retrofit technology applied to bridges on Tokyo metropolitan

expressways. Workshop Proceedings were disseminated in both

countries.

* In cooperation with T/C "G", the U.S. side hosted an Ad Hoc Workshop

on Seismic Response Control of Bridge Structures in September 1991 at

the National Center for Earthquake Engineering Research to continue

discussion and review practical seismic active, passive and hybrid control

applications in both countries. Participants toured seismically isolated

bridges in the U.S.

* Conducted coordinated research studies on seismic performance of

bridge piers and columns. Details of this work involved determining the

performance of reinforced concrete piers and columns subjected to

dynamic cyclic loading, performing model tests on the failure of

reinforced concrete piers, testing full and large-scale concrete columns

and for the behavior of concrete filled steel tubes.

* Held the eighth bridge workshop in Chicago, Illinois in May 1992 which

was followed by site visits to major bridges along the Mississippi River

at Burlington, Iowa; Quincy, Illinois; Alton, Illinois; and St. Louis,

Missouri. Main themes of the workshop included: (1) Long-Span

Bridges; (2) Wind Effects and Codes; (3) Seismic Issues; (4) Design; (5)

Bearings; (6) Foundations; and (7) Evaluation, Rating and Management.

Held the seventh bridge workshop in Tsukuba, Japan in May 1991,

which was followed by site visits of Higashi Kobe Bridge, Kansai

International Airport, and Akashi Kaikyo Bridge.

* Both sides presented technical papers at the 23rd and 24th UJNR Joint

Meetings and participated in technical study tours to buildings, bridges,

factories, construction sites, and instrumented structures.

4. Future Plans

* Conduct the 9th bridge workshop just prior to the 25th Joint Panel

566
meeting in Tsul<uba, Japan.

* Plan to host a 2nd U.S. -Japan Worl<shop on Seismic Retrofit of Bridges

in the U.S. The Worl<shop will focus on practical seismic retrofit

measures.

* Continue to investigate and exchange technical information on improved

seismic retrofit and strengthening procedures for highway bridges based

on experimental, analytical, and field studies. This exchange should

include information on maintenance of existing bridges.

* Continue experimental research study on the seismic performance of

bridge piers and columns, and support emphasis on base isolation and

hybrid control of bridges in cooperation with T/C "G". The 2nd

Workshop on seismic control is planned for December 1992 in Tsukuba.

* Develop a coordinated research study on seismic, aeroelastic, and

aerodynamic response of cable-supported bridges with emphasis on

cable inspection, vibration, control and corrosion protection.

* Continue a coordinated research study to compare the seismic design

criteria for bridges in Japan and the U.S. and discuss the method and

analysis procedures for bridge column design. Information on application

of limit state design method continues to be exchanged.

* Encourage a coordinated research study on seismic response and control,

system identification techniques, and nondestructive evaluation of bridge

structures.

* Encourage a coordinated research study on use and performance of

structural materials including new materials.

* Encourage a coordinated study on performance of jointless bridges.

Information Exchange

* Exchanged technical reports, research program documentation,

construction logs, design plans, and assorted photographs and videos,

including technical data appropriate for investigating the Philippine

earthquake of 1 990.

* FHWA hosted PWRI visitors Messrs. K. Kohno, H. Kanzaki, A. Fujimoto,

and J. Inoue. Mr. Inoue studied for five months at the Turner-Fairbank

Highway Research Center and visited State highway agencies in Virginia,

Maryland, Pennsylvania and California to study bridge maintenance and

management. Mr. Inoue prepared a report entitled "Comparison of

567
Highway Bridge Maintenance Between tlie U.S. and Japan".

* Mr. Mohsen Sultan, California Department of Transportation conducted

a study on the comparison of Seismic Design for U.S. and Japanese

bridges during a one year exchange with PWRI in Tsukuba.

6. Impact

* The free exchange of literature, instrumentation technology, and

earthquake response data has led to the advancement in design

technology, and development of preliminary retrofit guidelines, and to

the calibration and verification of specifications.

* Greater uniformity was achieved in wind engineering test procedures and

modeling resulting in more efficient solutions to the aerodynamic bridge

problem.

* The workshop and study tours in the two countries have facilitated

technology transfer. Examples include retrofit technology, cable

protection and vibration suppression of cable-stayed bridges.

* A set of priorities for needed research from a more international

viewpoint has reduced the conduct of duplicative'research efforts.

7. Barriers . .

There are no major technical barriers to achieve the Task Committee mission.

However, financial barriers, especially funding needed to participate in workshops and

panel meetings and for conducting actual bridge tests and large-scale model tests,

continue to limit the effectiveness of the work in this Task Committee's work.

568
Report of Task Committee on

(K) WIND AND EARTHQUAKE ENGINEERING FOR OFFSHORE AND COASTAL

FACILITIES

Date : May 20, 1992

Place: National Institute of Standards and Technology

Gaithersburg, MD

Attendees: U.S. Side C. Smith (Chairman)

G. Brady

W. Roper

M. Miller

(MMS)

(USGS)

(COE)

(WES)

Japan Side - S. Noda (Temporary Chairman) (PHRI)

K. Yokoyama (PWRI)

J. Koseki (PWRI)

MISSION STATEMENT

1 . Objective :

The objective of this Task Committee is to develop technical insights necessary to

mitigate damage to offshore and coastal facilities due to extreme wind and seismic

effects. The Task Committee will plan, promote, and develop research initiatives to

meet this objective, and will disseminate the results of their research for incorporation

into future specification or design guidelines.

Criteria for the design of offshore and coastal facilities may differ greatly from their

onshore counterparts. These differences can arise due to their unique design or mass

distribution, to the fluid/structure or wind/structure interaction, to the placement of

foundation elements in or on soft, fully saturated soils that can be subject to large

hydrodynamic pressures, and to the lack of specific environmental data or the

engineering experience that has been developed for most onshore sites.

2. Scope of Work

The Task Committee's work spans a broad spectrum of scientific and engineering

disciplines to include a variety of geoscience, structural, mechanical, and marine

topics. Structures of interest include offshore oil and gas facilities, aids to navigation,

breakwaters, bulkheads, and other coastal facilities. The Task Committee will provide

569
the forum for the communication between the above disciplines as they impact the

stated objective. Items of specific interest include the following:

* Sponsoring and conducting workshops and meetings to identify key

areas of research, opportunities for cooperation, and the exchange of

knowledge.

* Predicting strong ground motions for offshore and coastal sites including

assessing the effects of basin geometry, linear, and nonlinear local

geological effects using actual seafloor response measurements.

* Determining the dynamic response and the interaction of

structure/foundation/soil systems to seabed motions and/or extreme

wind forces.

* Assessing the dynamic response and behavior of various operational

facilities mounted on offshore and coastal structures.

* Developing assessment methodologies for seismicity and other

characteristics of potential seismic sources (e.g. faults) for offshore and

coastal sites in regards to how these conditions relate to structural

design criteria.

* Promoting the implementation of new research results into current design

and construction processes.

* Developing research efforts to include laboratory and field programs to

obtain data on the response of offshore and coastal facilities to extreme

wind and seismic forces.

* Creating performance standards, design specifications, guidelines, and

code recommendations for application to new construction as well as remedial action for existing facilities.

3. Accomplishments '

* Established T/C scope of work and T/C operational plan.

4. Future Plans

* Hold a U.S. /Japan Workshop on Wind and Earthquake Engineering for

Offshore and Coastal Facilities the week before the 25th Joint UJNR

Meeting at the Port and Harbour Research Institute in Yokosuka, Japan.

* Develop workshop agenda and finalize list of speakers from both the

U.S. and Japan Side.

570
Coordinate, where possible, on-going research on wind and earthquake

engineering for offshore and coastal facilities of interest to the members

of the Task Committee.

5. Information Exchange

* With Japan and U.S. members through T/C meetings and annual joint

meetings.

* With the Engineering and Scientific community through meetings,

conference sessions, seminars, and workshops.

6. Impact

* Increased cooperation among scientists and engineers concerned with

the seismic design of offshore and coastal facilities.

* Increased awareness of problems and risks associated with seismic

events affecting offshore and coastal facilities.

7. Barriers

* Insufficient funding to sponsor a strong research initiative and to support

travel and the exchange of personnel.

Not enough awareness, especially on the U.S. -side of the importance of

investigating the risks associated with seismic concerns for offshore and

coastal facilities.

571
I

I

I
NIST-114A U.S. DEPARTMENT OF COMMERCE

(REV. 3-90) NATIONAL INSTITUTE OF STANDARDS AND TECHNOLOGY

BIBLIOGRAPHIC DATA SHEET

1. PUBUCATKMI OA REPORT MUMB£R

NIST/SP-843

2. PERFOBinilQ OAOANIZATtON REPORT MUUBER

PUBUCAT10M DATE September 1992

TITLE AND SUBTITLJE Wind and Seismic Effects

Proceedings of the 24th Joint Meeting of the U.S. -Japan Cooperative Program in Natural

Resources Panel on Wind and Seismic Effects

5. AUTHOR(S)

Noel J. Raufaste

6. PERFGRMINQ OROANIZATION (IF JOINT OR OTHER THAN NIST. SEE INSTRUCTIONS)

U.S. DEPARTMENT OF COMMERCE

NATIONAL INSTITUTE OF STANDARDS AND TECHNOLOGY (BFRL)

QAtTHERSBURQ, MD 20899

7. CONTRACT/GRANT NUUBS<

a. TYPE OF RS>ORT AND PeRK>0 COVBtED

Final

9. SPONSORINQ ORGANIZATION NAME AND COMPLETE ADDRESS (STREET, CrTY, STATE, ZIP) Same as //6 above

10. SUPPLEMENTARY NOTES

11. ABSTRACT (A 200-WORD OR LESS FACTUAL SUMMARY OF MOST SIGNIFICANT IHFORMATWK. OOCU**eMT IHCUJC€S A SlGKlfTCAKT SJBLIOGPJl^-Y OR

LITERATURE SURVEY, MENTION IT HERE.)

This publication is the proceedings of the 24th Joint Meeting of the U.S. -Japan Panel on Wind and Seismic

Effects. The meeting was held at the National Institute of Standards and Technologj', Gaiihersburg. Mary land,

during May 19-22, 1992. The proceedings include the program, list of members, panel resolutions, task

committee reports, and 45 technical papers.

The papers were presented under five themes: (I)-Wind Engineering. (II) -Storm Surge and Tsunamis,

(III) -Joint Cooperative Research Program, (IV) -Earthquake Engineering, (V) - Summaries of Task

Committee Workshop Reports (oral presentations only).

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disaster reduction; earthquakes; geotechnical engineering; ground failures; lifelines;

liquefaction; masonry; repair and retrofit; risk assessment; seismic; soils; standards;

storm surge; structural engineering; tsunamis; and wind loads

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