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Wind and

Seismic

Effects

NBSIR 88-3703

Proceedings of

the Nineteenth

Joint Meeting
T

ff he National Bureau of Standards' was established by an act of Congress on March 3, 1901. The M Bureau’s overall goal is to strengthen and advance the nation’s science and technology and facilitate their effective application for public benefit. To this end, the Bureau conducts research and provides: (1) a

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The Bureau’s technical work is performed by the National Measurement Laboratory, the National

Engineering Laboratory, the Institute for Computer Sciences and Technology, and the Institute for Materials

Science and Engineering

.

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: Some divisions within the center are located at Boulder, CO 80303. ’Located at Boulder, CO, with some elements at Gaithersburg, MD.
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Gaithersburg, Maryland 20899

Wind and

Seismic

Effects

PROCEEDINGS OF

THE 19TH JOINT

MEETING OF

THE U.S.-JAPAN

COOPERATIVE PROGRAM

IN NATURAL RESOURCES

PANEL ON WIND AND

SEISMIC EFFECTS

Issued January 1988

Noel J. Raufaste,

EDITOR

Center for Building Technology

National Engineering Laboratory

National Bureau of Standards

Gaithersburg, MD 20899

U.S.»DEPARTMENT OF COMMERCE,

C. William Verity, Secretary

National Bureau of Standards,

Ernest Ambler, Director
.

.
PREFACE

The U.S. -Japan Cooperative Program in Natural Resources (UJNR) was established

in 1964 for the exchange of technical information and experience mutually

beneficial to the economics and welfare of both countries. The Panel on Wind

and Seismic Effects held its first joint meeting in 1969 in Tokyo, Japan. The

joint meeting has been held annually ever since, alternating in Japan and the

United States. The Panel is composed of 15 U.S. and 6 Japanese agencies,

co-chaired by the National Bureau of Standards and the Public Works Research

Institute. Results from this Panel's efforts have impacted building and

structure code changes in both countries.

The Nineteenth Joint Meeting was held at the Public Works Research Institute,

Tsukuba, Japan, May 12-15, 1987. The Panel featured five themes; 30 technical

papers were presented out of the 38 authored manuscripts. Ten Panel

task committees promote joint cooperation; they held their meetings

during this period.

These proceedings include the program of the Nineteenth Joint Meeting, the

Panel resolutions, all technical papers, and the task committee reports.

Partial support to maintain the US Panel was provided by the National

Science Foundation, Department of State, Bureau of Reclamation, Department

of Navy, Department of Housing and Urban Development, Federal Emergency

Management Administration, Nuclear Regulatory Commission, US Geological

Survey, and the National Bureau of Standards.

Noel J. Raufaste, Secretary

US Side, Panel on Wind and Seismic Effects

m
ABSTRACT

The Nineteenth Joint Meeting of the U.S. -Japan Panel on Wind and Seismic

Effects was held at the Public Works Research Institute, Tsukuba, Japan,

from May 12 - 15 , 1987 . This publication, the proceedings of

the Joint Meeting, includes the program, list of members, Panel resolutions,

Panel charter, task committee reports, and technical papers.

Papers were presented to the Panel under five themes. These were?

Theme I - WIND ENGINEERING

Theme IT - EARTHQUAKE ENGINEERING

Theme III - STORM SURGE AND TSUNAMI

Theme IV - U.S. -JAPAN COOPERATIVE RESEARCH PROGRAM

Theme V - RESULTS OF RECENT TASK COMMITTEE WORKSHOPS

Subjects covered in the papers presented include?

(1) characteristics of strong winds;

(2) wind loads on structures and design criteria;

(3) earthquake ground motions;

(4) soil liquefaction studies and methods to improve liquefaction

resistance;

(5) seismic loads on structures and design criteria;

(6) stress analyses of pipelines during earthquakes;

(7) full-scale seismic experiments;

(8) earthquake hazard reduction program;

(9) use of the microcomputer for earthquake studies;

(10) quantitative evaluation of damage caused by winds and earthquakes;

(11) tsunami research projects;

(12) coordinated masonry building research;

(13) bridge design to resist natural hazards.

KEYWORDS: Accelerograph; bridges; codes; concrete; design criteria;

disaster; earthquakes; earthquake hazards; earthworks;

geotechnical engineering; ground failures; liquefaction;

masonry; pipeline; repair and retrofit; seismicity; standards;

storm surge; structural engineering; tsunami; and wind loads.
CONTENTS

PAGE

PREFACE iii ABSTRACT .. iv CONTENTS . . . .

.

. v PROGRAM FOR 19th JOINT UJNR MEETING ix

LIST OF MEMBERS xv

LIST OF TASK COMMITTEE MEMBERS xxvi

LIST OF U.S. PARTICIPANTS xxviii

FORMAL RESOLUTIONS xxx CHARTER xxxiii THEMES AND TECHNICAL PAPERS 1

THEME Is Wind Engineering

Mean Wind Force Measurements on the

Deer Isle-Sedgwick Suspension Bridge 3

Harold R. Bosch

Field Experiments for Wind Pressure 12

Kishor C. Mehta

Aerodynamic Characteristics of Continuous Box Girder

Bridges Relevant to their Vibrations in Wind 16

N. Narita, K. Yokoyama, H. Sato, and Y. Nakagami

On Interpreting Model and Prototype

Responses of Suspended-Span Bridges to Wind 23

Robert JrL Scantan

THEME II; Earthquake Engineering

The United States Geological Surrey

Accelerograph Network in the U.S.-Operations.

Record Processing and Research ............ ............... ............ 33

A. Gerald Brady

Estimation of Source Time Function Based on Strong

Ground Motions and Its Application for Engineering .............................. 37

Yoshikazu Kitagawa, Takahito Inoue, and Tatsuo Murota
CONTENTS

PAGE

Strong-Motion Instrumentation Program

Corps of Engineers .................................. 44

F. K. Chang, R. F. Ballard, Jr., and A. G. Franklin

Comparisons of Dynamic Shear Strength

Between Fine Sands and Gravels 53

Norihisa Matsumoto, Nario Yasuda, Masahiko Ohkubo, and Shoji Arakawa

Preliminary Study of Shear Design for RC Piers. ................................. 62

Hiroshi Shinohara, Michio Okahara, Shouichi Nakatani,

Keiji Taguchi, and Atsushi Funakoshi

Some Recent Studies on Lifeline Earthquakie Engineering

in the United States: Experimental and Field Study ............................... 74

T. D. O’Rourke and J. Isenberg

Some Recent Studies on Lifeline Earthquake Engineering

in the United States: Analysis and Design 90

Masanobu Shiinozuka and Leon R. L. Wang

A Review of Base Isolation for Upgrading Earthquake

Performance of Building Structures 108

Nishikant R. Vaidya and Robert J. Smith

Consideration of Earthquakes in Highway Earthworks 122

Yasuyuki Koga, Akiomi Shimazu, and Eiichi Taniguchi

Damage, Strengthening, and Retrofit of Bridges. 132

James H. Gates

Large Scale Model Tests and Analysis of Gravel Drains 137

Susumu Iai, Katsuhiko Koizumi, Setsuo Noda, and Hajime Tsuchida

Experimental Study on Degree of Fixation at Pile Head Joint and Failure

Strength of Prestressed High Strength Concrete Piles 151

Yoshihiro Sugimura and Tsutomu Hirade

Schemes for Seismic Upgrading of a Mqjor Hospital 166

Joseph V. Tyrrell

vi
CONTENTS

PAGE

THEME III: Storm Surge and Tsunami

Experimental Investigation on the Damage of Offshore

Seawall due to the 1983 Nlhonkal-Chubu Tsunami 177

Katsutoshi Tanimoto and Hiroichi Tsuruya

A Summary of the Tsunami History of the United States 187

Herbert Meyers, P. A. Lockridge, and J. F. Lander

An Estimation of Tsunami Generated by Submarine Landslide 200

Sin-Iti Iwasaki and Hiroshi Takahashi

A Computer Program for Producing Tsunami

Travel-Time Chart 209

Masami Okada and Masaji Ichikawa

Numerical Simulation and Experiment on Tsunami Run-Up 217

Takaaki Uda, Atsushi Omata, and Yasuhisa Yokoyama

THEME IV: U.S.-Japan Cooperative Research Program

Effect of Number of Loading Cycles and Loading Velocity of

Reinforced Concrete Bridge Piers. 225

Toshio Iwasaki, Kazuhiko Kawashima, Kinji Hasegawa,

Tatsuhiko Koyama, and Takeshi Yoshida

Study of In Situ Testing for Evaluation of Liquefaction

Resistance and Occurrence 239

Jeffrey A. Farrar

U.S.-Japan Coordinated Earthquake Research Program on Masonry Buildings, Seismic Behavior of Three-Story

Full Scale Block Planar Frame Under Cyclic Lateral Loading 281

Shin Okamoto, Yutaka Yamazaki, Takashi Kaminosono,

Masaomi Teshigawara, and Hisahiro Hiraishi

U.S. Coordinated Program for Masonry Building

Research- 1986/ 87 292

James L. Noland

vii
CONTENTS

PAGE

MANUSCRIPTS AUTHORED FOR PANEL MEETING

BUT NOT PRESENTED ORALLY

Seismic Investigation and Improvement of

Electrical Utility Systems 299

Joseph V. Tyrrell

Wind Tunnel Test on Behavior of Rooftiles

under Strong Wind 305

Hisashi Okada and Tasuo Murota

Seismic Microzonation Using Urban Ground Infonnatlon .......................... 317

Shinkichi Kishi, Toshikazu Morohoshi, Keiichi Qtani,

and Hiroshi Takahashi

Simplified Evaluation Procedure of Seismic Stability

of Embankments on Liquefiable Ground ......................................... 327

Yasuyuki Koga and Osamu Matsuo

Strong Motion Observation System of PWRT .................................... 339

Toshio Iwasaki, Yasushi Sasaki, and Tetsuro Kuwabara

Evaluation of Economical Losses Induced by the

Nihonkai-Chubu Earthquake of 1983 352

Toshio Iwasaki, Kazuhiko Kawashima,

Naomi Obinata, and Takashi Kanoh

Analysis of Strong Motion Earthquake Records on January 9, 1987 ...................................................... 364

Nobuyoshi Yamaguchi, Sadaiku Hattori, Shinsuke Nakata,

Hatsukazu Mizuno, and Masanori liba

Appendix; Ten Task Committee Reports. 375

Til]
PROGRAM FOR 19th JOINT UJNR MEETING

Opening Session

Tuesday - May 12, 1987(10:30 - 12:00)

Conference Hall (8th Floor), Public Works Research Institute

Call to order by Dr. Toshio Iwasaki , Secretary-General, Japan Panel,

Director of Planning and Research Administration Department, Public

Works Research Institute

Remarks by Mr. Toshio Hirose, Engineer General, Ministry of

Construction

Remarks by Dr. W. McPherson, Science Officer, Office of Science

and Technology, Embassy of the United States of America

Remarks by Mr. Masayasu Miyabayashi , Director, International Affairs

Division, Science and Technology Promotion Bureau, Science and

Technology Agency

Remarks by Mr. Shunichiro Kamijo, Chairman, Japan Panel, DirectorGeneral, Public Works Research Institute, Ministry of Construction

Remarks by Dr. R. N. Wright, Chairman, U.S. Panel, Director, Center

for Building Technology, NEL, National Bureau of Standards, Department

of Commerce

Introduction of Japan Panel Members by Japanese Chairman and U.S. Panel

Members by U.S. Chairman

Election of Conference Chairman

Adoption of Agenda

Memorial Addresses on the late Dr. T. Okubo by Dr. H. S. Lew and Dr. N. Narita

Adjourn

Commemorating Planting

Group Photograph

1

X
Session Program of the 19th Joint Meeting

Tuesday-May 12

Opening Session 10:30-12:00

Lunch 12:00-13:00

Speaker

International Conference Room

Technical

Session

Session 13:00-17:00

1 (Theme II EARTHQUAKE ENGINEERING)

13:00-13:20

13:20-13:40

13:40-14:00

14:00-14:20

Chairman:S. Kamijo

The United States Geological Survey

Accelerograph Network in the U.S. -- Operations, Record Processing and Research

A. G. Brady*

Estimation on Source Time Function based on Strong Motions and Its Application for

Engineering

Y. Kitagawa*, T. Inoue , T. Murota

Strong-Motion Instrumentation Program Corps of

Engineers

F. K. Chang, R. F. Ballard, A. G. Franklin*

Discussion

Session 2 (Theme II EARTHQUAKE ENGINEERING)

Chairman: S. Kamijo

14:20-14:40 Dynamic Behavior of Rock-fill Dam

K. Ohtani*, K. Baba, T. Nakai

14:40-15:00 Comparisons of Dynamic Shear Strength Between

Fine Sands and Gravels

• No Matsumoto*, No Yasuda, M. Okubo , So Arakawa

15:00-15:20 Discussion

15:20-15:40 Break

Session 3 ( Theme II EARTHQUAKE ENGINEERING)

15:40-16:00

16:00-16:20

16:20-16:40

16:40-17:00

Chairman:S. Kamijo

Preliminary Study on Shear Design for RC Piers

H. Shinohara, M. Okahara* , S. Nakatani

,

K. Taguchi, A. Funakoshi

Some- Recent Studies on Lifeline Earthquake

Engineering in the United States:

Experimental and Field Study

To O'Rourke, Jo Isenberg, (L. R. L. Wang*)

Some Recent Studies on Lifeline Earthquake

Engineering in the United States: Analysis and

Design

M. Shinozuka, L. R. L. Wang*

Discussion

Reception 18:00-

X
Wednesday-May 13

Technical

Session

Session 9:00-12:00

4 (Theme II EARTHQUAKE ENGINEERING)

9:00- 9:20

9:20- 9:40

9:40-10:00

10 : 00-10:20

10:20-10:40

10:40-11:00

Chairman :R. N. Wright

A Review of Base Isolation for Upgrading

Earthquake Performance or Building Structures

N. R. Vaidya, R. J. Smith*

Consideration of Earthquakes in Highway

Earthworks

Y. Koga* , A. Shimazu, E. Taniguchi

Damage, Strengthening, and Retrofit of Bridges

J. H. Gates*

Large Scale Model Tests and Analysis of Gravel

Drains

S. Iai*, K. Koizumi, S. Noda, H. Tsuchida

Discussion

Break

Session 5 (Theme IV U.S.-JAPAN COOPERATIVE RESEARCH PROGRAM)

Chairman:R. N. Wright

11:00-11:20 Effect of Number of Loading Cycles and Loading

Velocity of Reinforced Concrete Bridge Piers

T. Iwasaki , K. Kawashima*, K. Hasegawa,

T. Koyama, T. Yoshida

11:20-11:40 Study of In Situ Testing for Evaluation of

Liquefaction Resistance and Occurrence

J. A. Farrar, (R. Ledzian*)

11:40-12:00 Discussion

Lunch 12:00-13:00

Technical Session

Session 6 (Theme IV U

13:00-13:20

13:20-13:40

13:40-14:00

14:00-14:10

13:00-14:00

S. -JAPAN COOPERATIVE RESEARCH PROGRAM)

Chairman:R. N. Wright

U.S. -Japan Coordinated Earthquake Research

Program on Masonry Buildings — Seismic

Behavior of Three-Story Full Scale Brick

Planar Frame under Cyclic Lateral Loading —

S. Okamoto, Y. Yamazaki*, T. Kaminosono,

M. Teshigawara, H. Hiraishi

U.S. Coordinated Progam for Masonry Building

Research 1986/87

J. L. Noland, (H. S. Lew*)

Discussion

Break

Task Committee Meetings 14:10-17:00

B : Large-Scale Testing Programs

E:Natural Hazard Assessment and Mitigation through Land

Use Programs

H:Soil Behavior and Stability during Earthquakes

I :Storm Surge and Tsunamis

J :Wind and Earthquake Engineering for Transportation Systems

Reception 13:00-

XI
Thursday-May 14

Technical Session

Session 7 (Theme

9:00-12:00

I WIND ENGINEERING)

9:00- 9:20

9:20- 9:40

9:40-10:00

10 : 00-10:20

10:20-10:40

10:40-11:00

Chairman:S. Kamijo

Mean Wind Force Measurements on the Deer IsleSedgwick Suspension Bridge

H. R. Bosch, (C. F. Galambos*)

Field Experiments for Wind Pressure

K. C. Mehta, (Co F. Galambos*)

Aerodynamic Chracteristics of Continuous Box

Girder Bridges Relevant to Their Vibrations

in Wind

N. Narita, K. Yokoyama*, S. Sato, Y. Nakagami

On Interpreting Model and Prototype Responses

of Suspended-Span Bridges to Wind

R. Ho Scanlan, (K. Yokoyama*)

Discussion

Break

Session 8 (Theme II EARTHQUAKE ENGINEERING)

Chairman :Ro N. Wright

11:00-11:20 Experimental Study on Degree of Fixation at

Pile Head Joint and Failure Strength of

Prestressed High Strength Concrete Piles

Y. Sugimura*, T. Hirade

11:20-11:40 Schemes for Seismic Upgrading of a Major

Hospital

J. Vo Tyrrell*

11:40-12:00 Discussion

Lunch 12:00-13:00

Session 9(Reports of Workshops) 13:00-14:20

13:00-13:30 Task Committee (C)

H. S. Lew*

13:30-14:00 Task Committee (J)

He Sinohara*

14:00-14:20 Discussion

14:20-14:30 Break

Task Committee Meetings 14:30-17:00

A: Strong-Motion Instrumentation Arrays and Data

C:Repair and Retrofit of Existing Structures

D: Evaluation of Performance of Structures

F:Disaster Prevention Methods for Lifeline Systems

G:Wind Characteristics and Structural Response

i

i

XU
Friday-May 15

Technical Session

Session 10 (Theme III

9:00- 9:20

9:20- 9:40

9:40-10:00

10 : 00-10:20

10:20-10:40

9:00-12:00

STORM SURGE AND TSUNAMI)

Chairman :R. N. Wright

Experimental Investigation on the Damage of

Offshore Seawall due to the 1983 Nihonkaichubu Earthquake Tsunami

K. Tanimoto* , H. Tsuruya

A Summary of the Tsunami History in the United

States

H« Meyers*, P. A. Lockridge, J. F. Lander

An Estimation of Tsunami Generated by

Submarine Landslide

S. Iwasaki*, H. Takahashi

Discussion

Break

Session 11 (Theme III

10:40-11:00

11 : 00-11:20

11:20-11:40

11:40-12:00

STORM SURGE AND TSUNAMI)

Chairman:R. N. Wright

A Computer Program for Producing Tsunami Travel

Time Chart

M. Okada* , M.' Ichikawa

Numerical Simulation and Experiments on Tsunami run-up

T. Uda*, A. Omata, Y. Yokoyama

Discussion

Break

Lunch 12:00-13:00

Reports of Task Committees 13:00-15:00

Chairman:S. Kamijo

A: Strong-Motion Instrumentation Arrays and Data

B : Large-Scale Testing Programs

C: Repair and Retrofit of Existing Structures

Devaluation of Performance of Structures

E: Natural Hazard Assessment and Mitigation through Land

Use Programs

F:Disaster Prevention Methods for Lifeline Systems

G:Wind Characteristics and Structural Response

H:Soil Behavior and Stability during Earthquakes

I: Storm Surge and Tsunamis

J:Wind and Earthquake Engineering for Transportation Systems

15:00-15:20 Break

Adoption of Charter 15:20-15:40

Chairman:S„ Kamijo

Adoption of Final Resolution 15:40-16:10

Chairman:S. Kamijo

xrn

16:10 Break
Closing Session

Friday-May 15, 1987(16:30-16:50)

International Conference Room

Opening Remarks by Dr. T. Iwasaki , Secretary-General , Japan Panel,

Director, Planning and Research Administration Department, Public Works

Research Institute

Remarks by Dr. R. N. Wright, Chairman, U.S. Panel, Director, Center for

Buildinig Technology, NEL, National Bureau of Standards

Remarks by Mr. S. Kamijo, Chairman, Japan Panel, Director-General,

Public Works Research Institute

Closure by Dr. T. Iwasaki

xiv
LIST OF MEMBERS

U.S. PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1987

Dr. Richard N. Wright, Chairman

Director, Center for Building

Technology

National Bureau of Standards

Gaithersburg, MD 20899

(301) 975-5901

Dr. S.T. Algermissen

Office of Earthquake Studies

Branch of Earthquake Tactonics

USGS

Denver Federal Center

Stop 978, Box 25046

Denver, CO 80225

(303) 236-1611

Dr. Eddie N. Bernard

Director, Pacific Marine

Environmental Laboratory

7600 San Point Way, NE

Seattle, WA 98115

(206) 526-6239 FTS 392-6800

Dr. A. Gerald Brady

Physical Scientist

Office of Earthquake Studies

U.S. Geological Survey

345 Middlefield Road

Menlo Park, CA 94025

FTS 467-2881

(415) 323-8111 X2881

Mr. Lee Butler

Waterways Experiment Station

Office WESCV-2

P.0. Box 631

Vicksburg, MS 39108

(601) 634-2405

Dr. A.G. Franklin

Chief, Earthquake Engineering

and Geophysics Division

Geotechnical Laboratory

U.S. Army Engineer Waterways

Experiment Station

Vicksburg, MS 39180

(601) 634-2658

Mr. Noel J. Raufaste, Secretariat

Structures Division

Center for Building Technology

National Bureau of Standards

Gaithersburg, MD 20899

(301) 975-5905

Dr. Celso S. Barrientos

Supervisory Physical Scientist

NOAA/NESDIS

1825 Connecticut Ave., N.W.

Universal Bldg., Rm. 518, E/A132

Washington, DC 20009

(202)673-5400

Dr. Roger D. Borcherdt

Chief, Branch of Engineering

Seismology and Geology

Office of Earthquake Studies, USGS

345 Middlefield Road

Menlo Park, CA 94025

(415) 323-8111 X2755

Dr. John J. Burns, Jr.

Head, Structures Section

Office of Research

U.S. Nuclear Regulatory

Commission

Washington, DC 20555

(301) 443-7860

Dr. Charles G. Culver

Chief, Structures Division

National Bureau of Standards

Gaithersburg, MD 20899

(301) 975-6048

Mr. G. Robert Fuller

Chief, Standards Branch

Manufactured Housing Construction

Standards Division, Room 9156

Dept, of Housing and Urban

Development

Washington, DC 20410-8000

(202) 755-6920

XV
Mrc Charles F. Galambos

Chief, Structures Division

Office of Engineering and

Highway Operations R & D

Federal Highway Administration

Department of Transportation

6300 Georgetown Pike

McLean, AV 22101

(703) 285-2087

Dr. Michael P. Gaus

Section Head of Critical

Engineering

Division of Emerging and

Critical Engineering Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

(202) 357-9500

Dr. Walter W. Hays

Deputy for Research Application

U.S. Geological Survey

Office of Earthquake Survey

905 National Center

Reston, VA 22043

(703) 648-6711

Mr. Richard W. Kramer

Civil Engineer

Technical Review Staff, D-3210

Bureau of Reclamation

P.0. Box 25007

Denver, CO 80225

FTS 776-8539 or (303) 236-8539

Mr. Paul K. Krumpe

Office of Foreign Disaster

Assistance

Agency for International

Development

Department of State

Washington, DC 20523

(202) 647-9758

Mr. James H. Gates

Office of Structures Design

Department of Transportation

1120 N Street

Sacramento, CA 95807

(916) 445-1439

Mr. Peter E. Gurvin

A/FBO, SA-6

Department of State

Washington, DC 20520

(202) 235-3689

Dr. William B. Joyner

Geophysicist

Office of Earthquake Studies

Branch of Ground Motion and

Faulting, USGS

345 Middlefield Road

Menlo Park, CA 94025

(415) 323-8111

Mr. Richard Krimm

Office of Natural and

Technological Hazards

Federal Emergency Management Agency

Washington, DC 20472

(202) 646-2871

Dr. H.S. Lew

Leader, Construction Engineering

Group

Center for Building Technology

National Bureau of Standards

Gaithersburg, MD 20899

(301) 975-6061

Dr. Shih-Chi Liu

Director, Earthquake Hazards

Mitigation Section

1800 G Street, N.W. , Rm. 1128

Washington, DC 20550

(202) 357-9780

Mr. Robert MacDonald

Geologic Service Branch

Code D 1630 Bureau of Reclamation

Building 67, Denver Federal Center

Denver, CO 80225

FTS 776-4195 of (303) 236-4195

XVI
Dr. Richard D. McConnell

Office of Construction

Veterans Administration

811 Vermont Avenue, NW

Washington, DC 20420

(202) 233-3103

Dr. Richard D. Marshall

Structural Engineering Group

Structures Division

Center for Building Technology

National Bureau of Standards

Gaithersburg, MD 20899

(301) 975-6071

Dr. Francis G. McLean

Chief, Division of Research and

Laboratory Services, D-1500

Bureau of Reclamation

P.O. Box 25007

Denver, CO 80225

FTS 776-5981 or (303) 234-3303

Mr. Herbert Meyers

Chief, Earth Geophysics

N0AA

325 Broadway

Boulder, CO 80303

(303) 497-6215 or FTS 320-6215

Mr. Joseph T. Schaefer

Chief, Techniques Development

Unit

National Severe Storm Forecast

Center

National Weather Service, NOAA

601 E. 12th Street

Room 1836

Kansas City, MO 64106

(816) 374-5672

Dr. Lawrence A. Soltis

Supervisory Research Engineer

Engineering Design Criteria

Forest Products Laboratory

Forest Service, Dept, of

Agriculture

Madison, WX 53705

(608) 364-5910 or FTS 364-5910

Alternates

Dr. Clifford Astill

Division of Emergency and

Critical Engineering

National Science Foundation

1800 G Street, NW, Room 1130

Washington, DC 20550

Mr. John F. Meehan

Principal Structural Engineer

Research Director

Structural Safety Section

Office of the State Architect

Department of General Services

Sacramento, CA 95805

(916) 445-8730

Dr. John B. Scalzi

Program Director, Systems

Engineering for Large Structures

National Science Foundation

1800 G Street, NW, Room 1130

Washington, DC 20550

(202) 357-7710

Mr. Robert J. Smith

Department of the Army

Office of the Chief of . Engineers

DAEN-ECE-D

Washington, DC 20314-1000

(202) 272-0220

Mr. Joseph Tyrrell

Director, Civil Structural Division

Naval Facilities Engineering Comman

200 Stovall Street

Alexandria, VA 22332

(703) 325-0047

Mr. Michael Changery

National Climatic Center

Federal Building

Asheville, NC 28801

(704) 259-0765 or FTS 672-0765

xvii
Dr. A.J. Eggenberger

Division of Engineering and

Critical Engineering

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

Mr. James Houston

WESCV-Z

P.0. Box 631

Waterways Experiment Station

Vicksburg, MS 39180-0631

Mr. Robert R. Ledzian

Senior Staff Assistant for

Research, W0-220

Bureau of Reclamation

Department of the Interior

18th at C Streets, N.W.

Washington, DC 20240

(202) 343-6703

Ms. Janina Mirski

Chief, Structures Division

Veterans Administration

811 Vermont Avenue, N.W.

Washington, DC 20420

(202) 389-2394

Mr. Lucian G. Guthrie

Structural Engineer

Office of Chief of Engineers

HQUSACE (DAEN-ECE-D)

Washington, DC 20314-1000

(202) 272-8673

Mr. Roger Kenneally

Structures Section

Office of Research

U.S. Nuclear Regulatory Commission

Washington, DC 20555

(301) 443-7860

Mr. T.K. Lew

Naval Civil Engineering Laboratory

L51

Port Huenene, CA 93043

(805) 982-5785

Mr. Ronald J. Morony

Program Manager

Buildinng Technology Research Staff

Department of Housing and Urban

Development

Washington, DC 20410

(202) 755-0640

xviii

/
JAPAN PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1987

Mr. Shunichiro Kamijo, Chairman

Director-General

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Toshio Iwasaki, Secretary-General

Director

Planning and Research Administration Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi,

Tsukuba-gun , Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tokunosuke Fujitani

Senior Research Staff

Physical Meteorology Research Division

Meteorogical Research Institute

Meteorological Agency

1-1 Nagamine, Yatabe-machi

Tsukuba-gun, Ibaraki-ken 305

Tel. 0298-51-7111

Mr. Minoru Fujiwara

Head, Bridge Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Sadaiku Hattori

Special Senior Scientist

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

XIX
Dr. Masaya Hirosawa

Director

International Institute of

Seismology and Earthquake Engineering Department

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Yuji Ishiyama

Head, Planning and Investigation Division

Research Planning and Information Department

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-rnachi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Mamoru Katsumata

Head, Seismological and

Volcanology Division

Meteorological Research Institute

Meteorological Agency

1-1, Nagamine, Yatabe-machi , Tsukuba-gun

Ibaraki-ken 305

tel. 0298-51-7111

Dr. Kazuhiko Kawashima

Head, Earthquake Engineering Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Yasuyuki Koga

Head, Soil Dynamics Division

Construction Method and Equipment Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

XX
Mr. Kiyoshi Kurashige

Head, Typhoon Research Division

Meteorological Research Institute

Meteorological Agency

1-1, Nagamine, Yatabe-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-51-7111

Dr. Norihisa Matsumoto

Head, Fill Type Dam Division

Dam Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tatsuo Murota

Director

Structural Engineering Department

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

Mr. Masatoshi Nagaoka

Head, Second Geographic Division

Geographical Survey Institute

Ministry of Construction

Kitazato 1-banchi, Yatabe-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-1111

Dr. Nobuyuki Narita

Deputy Director-General

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi,

Tsukuba-gun, Ibaraki-ken 305

Tel. 0298-64-2211

xxi
Dr. Setsuo Noda

Chief, Earthquake Resistant Structures Laboratory

Structure Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi , Kanagawa-ken 239

Tel. 0468-41-5410

Mr. Keiichi Ohtani

Head, Earthquake Engineering Laboratory

Second Research Division

National Research Center for Disaster Prevention

Science and Technology Agency

3-1, Tennodai , Sakura-mura, Niihari-gun

Ibaraki-ken 305

Tel. 0298-51-1611

Mr. Michio Okahara

Head, Foundation Engineering Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Shin Okamoto

Director

Research Planning and Information Department

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi , Tsukuba-gun

Xbraki-ken 305

Tel. 0298-64-2151

* Dr. Yasushi Sasaki

Head, Ground Vibration Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

xxii
Mr. Hiroshi Shinohara

Director

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Yoshihiro Sugimura

Head, Foundation Engineering Division

Structure Engineering Department

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Hiroshi Takahashi

Director-General

National Research Center for Disaster Prevention

Science and Technology Agency

3-1, Tennodai , Sakura-mura, Niihari-gun

Ibaraki-ken 305

Tel. 0298-51-1611

Dr. Katsutoshi Tanimoto

Chief, Breakwaters Laboratory

Hydraulic Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi , Kanagawa-ken 239

Tel. 0468-41-5410

Dr. Hajime Tsuchida

Deputy Director-General

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi, Kanagawa-ken 239

Tel. 0468-41-5410

xxm
Dr. Takaaki Uda

Head, Coastal Engineering Division

River Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Hiroyuki Yamanouchi

Head, Structural Dynamics Division

Structural Engineering Department

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2151

Mr. Koichi Yokoyama

Head, Structure Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Tel. 0298-64-2211

•a* Secretary
HONORARY MEMBERS

Dr. Masami Fukuoka

Professor, Science University of Tokyo

Dr. Shiro Ibukiyama

Vice President, Ohbayashi Road Construction Co. , Ltd.

Mr. Mitsuru Nagao

President

New Structural Engineering Ltd.

Mr. Kenji Kawakami

Managing Director

Kubota Inc.

Dr. Kaoru Ichihara

President

Central Consultant Inc.

Dr. Kazuto Nakazawa

President

Civil-Works Research Laboratory

Mr. Yoshijiro Sakagami

Director, Vice-President

Tokukura Construction Co., Ltd.

Dr. Ryuichi Iida

Director

Japan Dam Engineering Center

Dr. Masateru Tominaga

Counselor

Public Works Research Center

XXV
LIST OF TASK COMMITTEE MEMBERS

Task Committee

A. Strong-Motion Instrumentation

Arrays and Data

B. Large-Scale Testing Programs

C. Repair and Retrofit of

Existing Structures

D. Evaluation of Performance

of Structures

E. Natural Hazard Assessment

and Mitigation Through

Land Use Programs

F. Disaster Prevention Methods

for Lifeline Systems

G. Wind Characteristics and

Structural Response

H. Soil Behavior and Stability

During Earthquakes

U.S. Side

* A.G. Brady

WoB. Joyner

AoF. Franklin

FoG. McLean

H. Meyers

* HoS. Lew

JoB» Scalzi

R e J« Smith

* J o B c Scalzi

CoFo Galambos

P.E. Gurvin

H.S. Lew

R»D. McConnell

R.J. Morony

* GeRo Fuller

RcDo McConnell

J cP a Meehan

JcB, Scalzi

R.J. Smith

* SoT. Algerraissen

AoG. Brady

GoR. Fuller

R.D. Marshall

* S.C. Liu

J.B. Scalzi

* R.Do Marshall

C.S. Barrientos

C.F. Galambos

M.P. Gaus

* AoG. Franklin

FoG. McLean

R.J. Smith

Japanese Side

* Hajime Tsuchida

Sadaiku Hattori

Masaji Ichikawa

Yuji Ishiyama

Keiichi Ohtani

Yasushi Sasaki

Hiroyuki Yamanouchi

* Keiichi Ohtani

Yasuyuki Koga

Setsuo Noda

Shin Okamoto

Koichi Yokoyama

Hiroyuki Yamanouchi

* Masaya Hirosawa

Toshio Iwasaki

Shin Okamoto"

Hiroshi Shinohara

* Shin Okamoto

Yuji Ishiyama

Michio Okahara

Hiroshi Shinohara

Hiroyuki Yamanouchi

* Kazuhiko Kawashima

Sadaiku Hattori

Toshio Iwasaki

Masatoshi Nagaoka

Yoshihiro Sugimura

* Toshio Iwasaki

Kazuhiko Kawashima

Tatsuo Murota

Setsuo Noda

Keiichi Ohtani

Yasushi Sasaki

* Keizo Masamura

Tatsuo Hanafusa

Tatsuo Murota

Nobuyuki Narita

Koichi Yokoyama

* Yasushi Sasaki

Yasuyuki Koga

Norihisa Matsumoto

Michio Okahara

Yoshihiro Sugimura

Hajime Tsuchida

XXVI
I. Storm Surge and Tsunamis

J. Wind and Earthquake Engineering

for Transportation Systems

*C.S. Barrientos *Takaaki Uda

E.N. Bernard Masaji Ichikawa

L. Butler Keizo Masamura

M.Pe Gaus Hiroshi Takahashi

H. Meyers Katsutoshi Tanimoto

*C.F. Galambos *Hiroshi Shinohara

C.G. Culver Minoru Fujiwara

A.G. Franklin Toshio Iwasaki

J.B. Scalzi Kazuhiko Kawashima

Nobuyuki Narita

Michio Okahara

Yasushi Sasaki

Koichi Yokoyama

*Chairman

XXYU
LIST OF U.S. PARTICIPANTS

Dr. Richard N. Wright, Chairman

Director, Center for Building Technology

National Bureau of Standards

Gaithersburg, MD 20899

Mr. Noel J. Raufaste, Secretariat

Structures Division

Center for Building Technology

National Bureau of Standards

Gaithersburg, MD 20899

Dr. A. Gerald Brady

Physical Scientist

Office of Earthquake Studies

U.S. Geological Survey

345 Middlefield Road

Menlo Park, CA 94025

Dr. A. G. Franklin

Chief, Earthquake Engineering & Geophysics Division

CEWES-GH Geotechnical Laboratory

U.S. Army Engineer Waterways Experiment Station

Vicksburg, MS 39180

Mr. Charles F. Galambos

Chief, Structures Division

Office of Engineering & Highway Operations R & D

Federal Highway Administration

Department of Transportation

6300 Georgetown Pike

McLean, VA 22101

Mr. James H. Gates

Office of Structures Design

Department of Transportation

1120 N Street

Sacramento, CA 95807

Mr. Robert R. Ledzian

Senior Staff Assistant for Research, WO-220

Bureau of Reclamation

Department of the Interior

18 & C Streets, N.W.

Washington, DC 20240

Dr. H.S. Lew

Leader, Construction Engineering Group

Center for Building Technology

National Bureau of Standards

Gaithersburg, MD 20899

xxvm
Mr. Herbert Meyers

Chief, Earth Geophysics

National Oceanic and Atmospheric Administration

325 Broadway

'Boulder, CO 80 3 03

Prof. R. H. Scanlan

Department of Civil Engineering

Johns Hopkins University

Baltimore, MD 21218

Mr. Robert J. Smith

Department of the Army

Office of the Chief of Engineers

HQUSACE ( CEEC-ED

)

Washington, DC 20314-1000

Mr. Joseph Tyrrell

Deputy Chief Engineer

Naval Facilities Engineering Command

200 Stovall Street

Alexandria, VA 22332

Prof. L.R.L. Wang

Department of Civil Engineering

Old Dominion University

Norfolk, VA 23508

xxi

x
FORMAL RESOLUTIONS

RESOLUTIONS OF THE NINETEENTH JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS (UJNR)

Public Works Research Institute

Tsukuba Science City, Japan

May 12-15, 1987

The following resolutions are hereby adopted:

1. The Nineteenth Joint Panel Meeting provided an opportunity to exchange

valuable technical information which was beneficial to both countries.

In view of the importance of cooperative programs on the subject of wind

and seismic effects, the continuation of Joint Panel Meetings is

considered essential.

2. The following activities have been carried out since the Eighteenth

Joint Meeting:

a. Guest researchers from both countries performed joint research that

advanced the state of wind and earthquake engineering.

b. Technical documents, strong motion records including those from the

Mexico Earthquake of 1985, research reports, earthquake damage

records and workshop proceedings were exchanged.

c. Workshops and Planning Conferences were held:

(a) Planning Conference on Disaster Prevention Methods for Lifeline

Systems — Task Committee (F ) , August, 1986, Charleston, South Carolina

(b) Third Workshop on Repair and Retrofit —- Task Committee(C) , May, 1987, Tsukuba

(c) Planning Conference on Evaluation* of Seismic Resistance of

Existing Buildings

Task Committee(D) , May, 1987, Tsukuba

(d) Third Bridge Workshop —- Task Committee ( J ) , May, 1987, Tsukuba

The findings promoted improved knowledge for wind and seismic

disaster mitigation.

d. The Coordinated Research Program on Large-Scale Testing of Masonry

Structures is currently underway. Research is coordinated by the

Joint Technical Coordinating Committee on Masonry Research (JTCCMAR).

*

3. At the Eighteenth Joint Meeting, a charter was drafted for this Panel on Wind and Seismic Effects as suggested at the 11th UJNR Plenary

Conference, December 1985, Tokyo. The charter was approved at the

Nineteenth Joint Meeting.

XXX
4 o The Panel recognizes the importance of the following items:

a. Translate and disseminate the Japanese MOC "Manual of Repair Methods

for Structures Damaged by Earthquakes" to U.S. engineering

profession

.

bo Exchange information on application of base isolation and active

damping systems.

c. Collect strong motion data on performance of buried pipeline systems.

d. Obtain experimental verification of the effectiveness of retrofitting

and strengthening methods for structures and soils.

5. The Panel endorses the following workshops proposed by the Task

Committees

:

a. Workshop on Strong-Motion Earthquake Observation, Task Committee (A),

to be held in the San Francisco area, August 1987.

b. Fifth Workshop on Repair and Retrofit, Task Committee (C), to be held

in conjunction with the 20th UJNR Joint Meeting.

c. Workshop on Wind Characteristics and Structural Response, Task

Committee (G)

,

to be held in the coming year.

d. Workshiop on Remedial Measure against Liquefaction, Task Committee

(H), to be held in conjunction with the 20th UNJR Joint Meeting.

e. Second Workshop on Storm Surge and Tsunami, Task Committee (I), to be

held in conjunction with the 20th UJNR Joint Meeting.

f. Fourth Bridge Workshop, Task Committee (J), to be held in conjunction

with the 20th UJNR Joint Meeting.

Scheduling for workshops shall be performed by the chairmen of each Task

Committee with concurrence of the Joint Panel chairmen. Results of each

workshop shall be presented at subsequent Joint Panel Meeting.

6. The Panel recognizes the importance of continued exchange of personnel,

technical information, research and recorded data, and methods to

mitigate losses from earthquakes and strong winds. The Panel also

recognizes the importance of using available large-scale testing

facilities in both countries. Thus, these activities should be

strengthened and expanded. To facilitate such exchanges, the Panel will

continue to encourage them and provide official Panel sponsorship.

7. The Panel recommends that for future Joint Panel Meetings, priority

attention be given to the following items:

a. Task Committee Chairmen should give attention to identifying joint

research needs, priorities, development plans, and recommendations

for future research programs. Each Task Committee's Chairmen are encouraged to examine their Committee's objectives and to modify the

objectives, as appropriate, to meet the need for improved technology.

b. Task Committee Chairmen should exchange summaries of their Task

Committees activities, current and proposed, to the other Task

Committees for improved understanding of Panel activities. Task

Committee Chairmen should furnish copies of Task Committee

correspondence to the Secretariates of both sides.

XXXI
8= Recognizing that wind and seismic disaster mitigation plans benefit

disaster-prone countries worldwide, that many developing countries

suffer from natural disasters such as extreme winds, storm surges,

earthquakes and tsunamis, and that both the U.S. and Japan have been

involved in international technical cooperative programs, the Panel will

continue efforts to develop and coordinate projects which provide aid

and training to developing countries and maintain the exchange of

technical information.

9. The Twentieth Joint Meeting of the UJNR Panel on Wind and Seismic

Effects will be held at the Washington D.C. (USA), May 1988. Specific

dates, program, and itinerary will be proposed by the U.S. Panel

with concurrence by the Japan Panel.

xxxii
CHARTER OF THE UJNR PANEL ON

WIND AND SEISMIC EFFECTS

OBJECTIVES

1) Encourage, develop, and implement the exchange of

wind and seismic technology between appropriate

United States and Japanese organizations to share

scientific and technological knowledge.

2) Develop strong technical links of scientific and

engineering researchers between the two countries

and encourage exchanges of guest researchers.

3) Conduct joint research in areas of winds and seismic

technology including exchange of available research

equipment and facilities in both countries.

Publish findings from joint research efforts.

4) Conduct cooperative programs to improve engineering

design and construction practices and other wind

and earthquake hazard mitigation practices.

Publish results from cooperative programs.

CURRENT TOPICS AND SUBJECT AREAS OF INTEREST

1) Strong Motion Instrumentation Arrays and Data

2) Large Scale Testing Program

3 ) Repair and Retrofit of Existing Structures

4) Evaluation of Performance of Structures

5) Natural Hazard Assessment and Mitigation Through

Land Use Programs

6) Disaster Prevention Methods for Lifeline Systems

7) Wind Characteristics and Structural Response

8) Soil Behavior and Stability During Earthquakes

9) Storm Surge and Tsunamis

10) Wind and Earthquake Engineering for Transportation

Systems

xxxiii
COOPERATIVE ACTIVITIES

1) Conduct annual joint panel meetings alternating

locations between the United States and Japan.

2) Publish proceedings of annual meetings and of task

committee events.

3 ) Exchange data and information between both countries

4) Exchange guest scientists and engineers.

5) Develop cooperative research programs on mitigating

the effects of wind and seismic forces on structures

Concerning these programs, exchange available

research equipment and facilities in both countries,

if necessary.

6) Conduct task committee meetings and workshops in

areas identified in "Current Topics and Subject

Areas of Interest" to facilitate exchange of

technical information.

7) Establish and maintain effective communications

between scientists, engineers, and administrators

of the two countries.

PANEL MEMBERSHIP

1) Members of the panel are personnel of government

agencies designated by the agencies.

2) Other experts may be selected, as temporary members,

from appropriate disciplines representing industry,

academia, and research organizations.

CHARTER MODIFICATIONS

This Charter may be revised by the concurrence of the US

and the Japanese sides.

xxxiv
Theme I

Wind Engineering

Mean Wind Force Measurements on the

Deer Isle-Sedgwick Suspension Bridge

By Harold R. Bosch^

ABSTRACT

This paper presents the results of a wind tunnel investigation of the Deer

Isle - Sedgwick Bridge. The structure

studied is a 45 year old suspension

bridge located in the State of Maine on

the New England coast. Static force

measurements were performed on a 1/25

scale section model of this bridge by

the Structures Division of the Federal

Highway Administration (FHWA). All

tests were conducted in the G. S.

Vincent Aerodynamics Laboratory at the

Turner-Fairbank Highway Research Center

(TFHRC). A low-speed wind tunnel with a

6 ft by 6 ft ( 1 . 8m by 1.8m) cross

section was employed to provide wind

velocities up to 40 fps (12.2m/s).

KEYWORDS: Aerodynamics; Drag; Lift;

Moment; Stability; Suspension bridge

1. INTRODUCTION

The wind tunnel research reported herein

is only part of a larger research

program being conducted jointly by FHWA

and the Maine Department of

Transportation, see reference 1. Work

on this project is divided into 4 basic

areas

:

o Field measurement of wind

characteristics at the bridge site

and structural response to wind as

well as traffic loading,

o Static and dynamic analysis of the

structure at each of its various

stages of development,

o Wind tunnel testing of section

models representing the existing as

well as proposed new shapes,

o Development and verification of

remedial or rehabilitation measures.

Wind tunnel tests were conducted on a 1/25 scale section model of the Deer

Isle-Sedgwick Bridge. The large scaling

ratio used permits accurate reproduction

of details from the existing bridge

cross section. Since the structure has

a shape similar to that of the original

Tacoma Narrows Bridge and has exhibited

significant wind induced response

throughout its many years of service,

several modifications to the basic

bridge section were also evaluated.

Section model studies were performed in

both smooth flow and simulated

large-scale turbulent flow conditions,

with wind speeds ranging from 0 to 40

fps (0 to 12.2m/s). RMS response

amplitudes were measured for the

vertical and torsional

degrees-of -freedom to identify regions

of aerodynamic sensitivity. Flutter

derivatives were computed using damping

and frequency measurements obtained from

free oscillations of the section model

under a variety of wind conditions.

Mean lift, drag, and pitching moment

forces were obtained for the basic

section model and the "best" alternate

at several different wind speeds with

wind angles of attack ranging from -10

to +10 degrees.

The experimental program outlined above

is, in fact, quite comprehensive and

research is still underway. The limited

space available here does not permit

discussion of all the work which has

been completed or is underway.

Therefore, this paper will concentrate

on the investigation and evaluation of

mean wind forces. First, a description

of the Deer Isle-Sedgwick Bridge will be

provided. Next, the test setup,

instrumentation, and procedures will be

discussed. Test results, in the form of

force coefficients, will be presented

for the existing shape and Alternate A.

Some comparisons will be made between

this bridge and the original Tacoma

Narrows Bridge. Finally, conclusions

will be drawn from the results

presented.

2. DEER ISLE-SEDGWICK BRIDGE

The bridge is a girder-stiffened

suspension bridge similar in cross

section to the original Tacoma Narrows

Bridge. As illustrated in figure 1, the

structure consists of a main span 1080

ft (329.2m) in length, side spans of 484

ft (147.5m) each, and approach spans of

130 ft (39.6m) each, giving a total

length of 2308 ft (703.5m). The

structure is symmetrical with two 213 ft

(64,9m) high towers. The roadway

represents a vertical curve with

tangents of 6.5% grade at each end to

provide vertical navigational clearance

of 85 ft (25.9m) at midspan. The 2-lane

roadway itself consists of a 4.5 in

(11.4cm) reinforced concrete slab having

a width of ’20 ft (6.1m). The stiffening

girders are 6.5 ft (2.0m) deep and

spaced 23.5 ft (7.2m) apart. Spacing of

'Federal Highway Administration, HNR-10

6300 Georgetown Pike, McLean, VA 22101

3
the suspender cables is 28 ft (8.5m).

Deck and tower details are presented in

figure 2.

3. TEST SETUP

The section model was suspended in front

of the wind tunnel nozzle by means of

six force transducers, as illustrated in

figure 3. Four of these transducers

were attached to the laboratory test

frame above the test area to measure

model lift and moment forces resulting

from the wind flow. The two remaining

sensors were mounted on the wind tunnel

pilasters, located upstream of the test

area, for measuring model drag forces.

The force transducers consist of stiff

cantilever beams with a full strain gage

bridge located near each fixed support.

This configuration was chosen for its

high strain output and small

displacement characteristics. Lift and

moment sensors were preloaded by the

weight of the model itself. To prevent

longitudinal or lateral movement of the

model during tests, horizontal

restraining wires were attached to the

end bracket crossbars at the center of

gravity of the section. Preloading of

the drag sensors was accomplished by

attaching restraining wires between the

crossbars and the laboratory wall

downstream of the wind tunnel nozzle.

All wires were tensioned by 2 lb (0.9kg)

weights and wire lengths were long to

ensure that vertical components do not

influence the force measurements.

4. TEST INSTRUMENTATION

Signals from the transducers were routed

through shielded cables to an array for

instrumentation, as illustrated in

figure 4. A set of signal conditioners

and amplifiers was used to energize the

strain bridge circuitry as well as

amplify the incoming analog signals. In

addition, the "built-in” low pass

filters were used to remove any

electronic noise present. Further

"cleaning" of the signals was

accomplished by passing them through a set of anti-alias low pass filters to

eliminate extraneous fluctuations. All

analog inputs were monitored through a

select switch connected to a digital

voltmeter

.

To monitor the true wind speed of the

wind tunnel, a pitot tube was extended

into the flow at the nozzle. This

sensor was connected via tubing to a high precision electronic manometer.

The manometer provides a digital display

as well as an analog output which is

proportional to the differential

pressure measured.

All conditioned, analog inputs were

connected to a high speed, automated

data acquisition system (DAS). This DAS

was employed to efficiently sample,

digitize, and store the experimental

measurements onto digital magnetic tape

for later reduction and analysis.

5.

TEST PROCEDURE

Before starting any measurements, the

sensitivity of the force-balance system

was adjusted to provide the maximum

allowable signals for the anticipated

range of applied forces. The system was

then calibrated by applying known

weights to simulate either drag, lift,

or pitching moment forces and recording

the analog outputs.

Static force measurements on the section

models were taken only under laminar

flow conditions. The tests were

conducted at wind angles of attack

varying from +10 to -10 degrees. For

each 1 degree position within this

range, the model angle was fixed and

force measurements were taken at a

series of steady wind velocities ranging

from 4 to 26 fps (1.2 to 7.9m/s). At

each new velocity, the wind tunnel was

allowed to stabilize before measurements

were taken. With wind angle and

velocity established, the seven data

channels were sampled at a rate of 20

cps for a period of 40 to 60 seconds.

The DAS was started and stopped manually

and the digitized data was recorded

automatically in a multiple record file

on tape. Periodically, air temperature

and barometer readings were taken for

later use in correcting the wind tunnel

velocity measurements. During testing,

the force-balance system was checked for

temperature drift and adjusted as

necessary.

Force measurements were conducted on the

existing cross section and Alternate A

only, illustrated in figure 5. Raw data

obtained for these two cases was

reduced, analyzed, and displayed using

the computer system illustrated in

figure 6. For each event or tape file,

a 20 second window of data was selected

for analysis and the mean values

calculated for each channel. These

"channel means" were then used together

with the system calibration factors to

determine the drag, lift, and pitching

moment forces per foot of section. Drag

and lift forces were translated to chord

and normal forces using the following:

DCO S '-*: ~ Lsin-=*: (1)

Lcos'-*: + Dsinc*: (2)

4
The orientation of these forces is

indicated in figure 7. All forces were

then reduced to dimensionless

coefficients defined as:

Cd = 2D / (9 V£ B) (3)

Cl = 2L / (9 V-B) (4)

c , , = 2M / (9 V-B- ) (5)

Cx = 2X / (9 V-B) (6)

Cz = 22 / (9 V-B) (7)

6. FORCE RESULTS

6 . 1 Existing Cross Section

Force coefficients for the existing

configuration are displayed graphically

in figures 8 through 12. These figures

represent a consolidation of all test

results obtained within the velocity

range 10.8 to 18.6 fps (3.3 to 5.7m/s).

The forces measured exhibit only slight

variation throughout the velocity range

investigated which supports the view

that violation of Reynolds number

scaling here will have little effect on

test results. The drag coefficients

vary from 0.36 to 0.48 over a 20 degree

range of wind angle. For a horizontal

wind with a mean velocity of 100 mph

(161km/h) and 50 mph (80km/h), the

respective drag forces are calculated to

be 240.4 and 60.1 Ib/ft (357.8 and

89,4kg/m) with corresponding pressures

of 37.0 and 9.2 psf (1772 and 440pa).

In comparison, the AASHTO Bridge

Specifications, reference 2, recommends

that pressures of 50 and 12.5 psf (2394

and 599pa), be used at the respective

velocities and a 0 degree horizontal

angle. These figures represent an

average drag coefficient of 1.955 for

common girder shapes. This coefficient

is based upon span depth (d) rather than

deck width (B) as demonstrated below:

P = D/d = (1/2)9 Vs (d/d) Cd

P = (. 5) ( .002378) (146. 667- ) ( 1 ) ( 1 . 955

)

P = 50 psf ( 2394pa ) (8)

The "equivalent" Deer Isle drag

coefficient of 1.446 is somewhat less

than the conservative AASHTO value. The

minimum and maximum coefficients give

pressures of 33.3 and 44.4 psf (1594 and

2126pa ) , respectively, at the higher

velocity. Although the high velocity

and high wind angle yield pressures

approaching the AASHTO value, it should

be noted that this condition is not

likely to occur. In point of fact, wind

angle tends to decrease as wind speed

increases. Figure 13 presents a wind

angle envelope developed from data

obtained at the Severn Bridge, see

reference 3. Such information is

frequently used as a guide in the

selection of realistic design

conditions

.

The lift curve for this cross section

develops a negative slope in the middle

region, -5 to +5 degrees, and flattening

of slope near the ends. These are

undesireable aerodynamic characteristics

since the negative slope indicates a tendency for galloping in vertical modes

and flattening means stalling of the

deck section at moderate angles of

attack. To determine the critical

velocity for onset of galloping, the

following relationship may be used:

2m*: to+ (1/2)9 VB ((dCi/d«)+ Cd ) = 0

from figure 9, ( dC l /d •=: ) = -2.2037/rad

from figure 8, Cd = 0 . 40 at =-: = 0°

2(74. 5) ( .015) ( . 213) (2tt ) +

( . 5) ( .002378) (Vcr) (23. 5) (-1. 8037) = 0

Vcr = 40.47 mph (65.12km/h) (9)

It should be noted that this computation

is for the fundamental vertical mode of

the structure with an assumed damping

ratio of 1.5 percent. This is

particularly important since the

critical velocity is directly

proportional to the assumed damping and

frequency

.

The moment curve exhibits a negative

slope over nearly the entire range of

angles. This characteristic points to

the potential for autorotation in

torsion and is thus undesireable.

6 . 2 Alternate A

Force coefficients for this

configuration are also presented in

figures 8 through 12, representing a consolidation of all test results for

wind velocities in the range 5.6 to 25.8

fps (1.7 to 7 . 9m/s )

.

It should first be noted that since this

model configuration is inherently more

stable, it was possible to obtain static

measurements over a much wider velocity

range than previously. Although there

is slightly more variation in the forces

with velocity, the total is still

considered to be relatively small and,

continues to substantiate the view on

Reynolds number scaling. In this case,

drag coefficients vary between 0.20 and

0.48, with a value of about 0.27 at 0

degrees. This results in a pressure of

25.0 psf (1195pa), which is less than

5
half the AASHTO value of 50 psf (2394pa)

for a velocity of 100 mph (161km/h).

Both the lift and moment curves for this

alternate indicate that the aerodynamic

characteristics of the bridge can be

greatly enhanced by this modification.

The slope of the lift curve is positive

over the entire range of wind angles.

In addition, the moment curve is mostly

positive in slope except for large

positive angles.

7. DEER ISLE AND TACOMA NARROWS

COMPARISON

Since the dramatic collapse of the

Tacoma Narrows Bridge in 1940, much

effort has been expended to study the

aerodynamic stability of long span

bridge superstructures. This event is

perhaps the best known and most

thoroughly documented example of a bridge deck failure due to aeroelastic

instability. Wind tunnel studies have

been conducted by many organisations

including the University of Washington,

California Institute of Technology,

FHWA, and others resulting in a wealth

of experimental information. The cross

section of the original Tacoma Narrows

Bridge, with stiffening, plate girders 8

ft (2.44m) deep and spaced 39 ft

(11.89m) apart, is quite similar to the

Deer Isle configuration. The structural

similarities together with the

availability of extensive aerodynamic

data make it possible to compare results

obtained for each bridge.

The University of Washington conducted

static force measurements on a 1/20

scale section model of the original

Tacoma Narrows Bridge, see reference 4.

Drag coefficients for this bridge model,

designated Aa, have been used here to

compute drag pressures for vertical wind

angles ranging from -10 to +10 degrees.

These pressures are plotted along with

the Deer Isle values in figure 14.

Review of the results indicates that

drag forces are generally equivalent for

the two bridges with only slight

variation at large positive angles of

attack. This small difference is likely

due to the smaller depth-to-width ratio

of 0.20, unvented curbs, and slightly

higher deck position of the Tacoma

Narrows section.

Force measurements were also performed

on a second Tacoma section model,

designated B, which was to the same

length scale as the first but with a larger aspect ratio. This model was

designed as a "universal model" on which

a variety of truss, deck, and plate

girder arrangements could be studied.

With a depth-to-width ratio of 0.29,

version "Bd" of this bridge model more

closely approximates the Deer Isle

proportions. Therefore, lift

coefficients for Model Bd have been used

to determine lift pressures for

comparison with Deer Isle values.

Resulting lift curves for the two

bridges are presented in figure 15.

Both sets of data clearly demonstrate

similar undesireable aerodynamic trends,

i.e., flattening of slope at moderate

angles and negative slope at small

angles

A review of pitching moment data

obtained for each structure also leads

to a like conclusion. Both bridge

section models exhibit a negative slope

in the moment curve over the range of

wind angles tested. As noted earlier,

this is an undesireable characteristic.

8. CONCLUSIONS

Based upon the results of these force

measurements, the following conclusions

may be drawn:

o Force coefficients obtained for both

the Existing Shape and Alternate A

show little variation with change in

wind velocity. This supports the

view that violation of Reynolds

number criterion in this case will

have minimal impact on test results.

o Drag pressures computed for the

Existing Shape as well as Alternate

A are somewhat less than the

conservative AASHTO values. In

fact, measurements on the alternate,

which is a relatively streamlined

cross section, indicate drag forces

roughly half those recommended for

design

o The lift coefficient curves obtained

for the Existing Shape exhibit a negative slope. This trend

generally indicates an undesireable

tendency for galloping response in

the vertical mode. In addition,

there is a flattening of the slope

at moderate vertical wind angles

which implies potential for stalling

of the deck section.

o In a similar fashion, the curves for

pitching moment coefficient

demonstrate a consistently negative

slope. Here, the slope indicates a tendency for autorotation in the

torsional mode.

o Force measurements on asymmetric

Alternate A, on the other hand,

clearly demonstrate that aerodynamic

characteristics of the basic cross

section can be greatly improved

6
through simple modification of the

shape. In this case, drag is

reduced significantly, while lift

and moment curves present strongly

positive slopes.

o Comparison of the Deer Isle static

force results with those obtained by

others for the original Tacoma

Narrows Bridge illustrates a number

of important similarities.

9. REFERENCES

(1)

Bosch, H. R. , "A Wind Tunnel Study

of the Deer Isle - Sedgwick Bridge

(Phase I)," Report No.

FHWA/RD- 87/027, Federal Highway

Administration, Washington, D. C.

,

June 1984.

(2) “Standard Specifications for

Highway Bridges," AASHTO,

Thirteenth Edition, 1983.

(3) Roberts, Sir Gilbert, "Severn

Bridge - Design and Contract

Arrangements," Proceedings of the

Institute of Civil Engineers,

Volume 41, pp. 1-58, September

1958. . (4) Farquharson, F. B. , et. al .

,

"Aerodynamic Stability of

Suspension Bridges with Special

Reference to the Tacoma Narrows

Bridge," The University of

Washington, Engineering Experiment

Station, Bulletin Number 116, Part

I, June 1949.

DEER ISLE - SEDGWICK BRIDGE

Little Dear Isle

Sedgwick

(Elevation)

130 ft 484 ft 1080 ft 484 ft 130 ft (39.6m) (147.5m) (329.2m) (147.5m) * (39.6m)

Irp ri rjL_ re

f icr u VT

(Plan)

24 ft (7.3m)
Fig. 2. Bridge deck and tower details

Fig. 3 . Force transducer setup

8
Fig. 4. Instrumentation system. Fig. 6. Data processing system.

9
Fig 0 7. Orientation of forces.

1.0

0. 8

0. 6

0. 4

0. 2

CD 0.0

- 0. 2

-0. 4

- 0. 6

“0. 8

• - 1.0

* + * . * ** * 1

+ + +

+ + + *

* *

* *******

+ Existing

* Alternate A

II —* *

ALPHA (DEG)

Fig. 8. Drag force coefficient.

1.0

0. 8

0.6

0. 4

0.2

CM 0.0

- 0.2

-0. 4

-0.6

- 0. 8

- 1.0(VSGD(D^C\JSC\J^(DODSC\J

^

I I I I •—0 «—

ALPHA (DEG)

Fig. 10. Moment force coefficient.

* • *

' * ®

i **»»** » * * *

+ i + ^ + i

+ Existing

* Alternate A

1.0

0.8

0.6

0. 4

0.2

CX 0.0

- 0.2

-0. 4

- 0. 6

- 0. 8

- 1.0

c

ALPHA (DEG)

Fig. 11. Chord force coefficient.

+ + + + ^ • $ + +

* *

i

. * +

t1 + ^

* „ . * * * *»» #

***

+ Existing

* Alternate A

1.

0.

0.

0.

0.

CL 0.

- 0.

- 0.

- 0.

- 0.

- 1 .

«- + + + + +

* +

*

>

******

*****

* +

* + +, 4, * +

* Existing

* Alternate A

CM S

I I

QD CO

I I (MSCM"<fCOCDS(M

ALPHA (DEG)

1.0

0.6

0. 4

0.2

CZ 0.0

- 0. 2

-0. 4

- 0.6

- 1 . 0

i

+ *

*******

*

*

•

*

+ + + •*• +

*

+ Existing

* Alternate A

CM Kl

I I

CD

I COI CM S CM CO CD

I ALPHA (DEG)

IS CM

Fig. 9. Lift force coefficient. Fig. 12. Normal force coefficient.

10
Fig. 13. Wind angle limits Severn Bridge

.

-4

DGR ISLE - SEDGWICK BRIDGE

0 TACOMA NARROWS BRIDGE OCOEL-BeD

* 0

CM ** (Q OS <33 CM

ALPHA (DEGREES) Fig. 15. Lift pressure - Deer Isle and Tacoma

11

Field Experiments for Wind Pressure

BY

KISHOR C. MEHTA

1

ABSTRACT

A project sponsored by the National Science

Foundation is underway to collect wind and

building surface pressure data in the field. A

30 ft x 45 ft x 12 ft building will be con- structed on a concrete pad in flat, open terrain

in Lubbock, Texas. The building structure is

designed to support on four wheels located in corners of the building so that the building can

be rotated on the concrete pad. A 100 ft tall

meteorological tower will be instrumented at

four levels to measure wind speed and direction.

Building surface and internal pressures will be

monitored using differential pressure transducers. A computer will control the data acquisition by triggering collection and termination

of record. Extra precautions are taken in

design of the experiment to collect reliable

data in a wind regime that exceeds 20 mph. The

paper describes planning of the field site, the

instrumentation and the data acquisition system.

KEY WORDS: Building; Field Data; Pressures;

Wind

1. INTRODUCTION

Wind-induced damage to buildings and other

structures causes significant economic loss in

addition to injuries and, at times, fatalities.

Continuing urban sprawl increases the population

of man-made structures, increasing the potential

future cost resulting from wind-induced damage.

It is recognized that the bulk of the damage is

caused to low-rise ordinary buildings. Enhanced

understanding of wind pressures on building

surfaces can lead to improved construction of

low-rise buildings and has potential of mitigating future damage. Pressure coefficient

values that are available in the literature are,

to a large extent, determined from wind tunnel

tests. A number of full-scale field tests in

natural wind environments have been conducted to

substantiate wind tunnel test results. Additional field test data can advance understanding

of wind pressures on building surfaces and can

pave the way for improvement in wind tunnel

technology.

A project is underway at Texas Tech University

to acquire wind and building surface pressure

data in the field. Specific objectives of the

National Science Foundation sponsored project

are: (1) to acquire a reliable data base for

external wind pressures on building surfaces and

internal pressures in the building, and (2) to assess the- effects of wind direction fluctuations on wind pressures.

The paper describes planning of the field site,

the 30 ft x 45 ft x 12 ft building, the meteorological and pressure instrumentation, and the

data acquisition system. The experiment will be

implemented within the next six months.

2. FIELD TEST FACILITY

The test site is in open terrain on land owned

by Texas Tech University in Lubbock, Texas. An

aerial view of the site is shown in Figure 1. The site is within a 10-minute drive from the

main campus where the research team is headquartered. Lubbock is located on the high

plains of Texas. There are no hills or valleys

within a 20-mile radius of the city. The

terrain is extremely flat, with ground elevation

changing approximately 5 ft in a 1000 ft distance. Most of the land surrounding Lubbock is

used for agricultural purposes with principal

crops of cotton and sorghum. This flat and open

terrain eliminates possible anomalies in wind

characteristics caused by terrain.

The Lubbock area is an appropriate site for wind

pressure experiments in the field because it

experiences strong winds many times in a typical

year. Table 1 summarizes by month the number of

hours of high winds recorded at the National

Weather Service station in Lubbock in 1981. The

National Weather Service station is within seven miles of the proposed site. Because of the flat

and open terrain in the Lubbock vicinity, the

winds at the NWS station and the site are of the same intensity. In the wind data shown in Table

2, the wind speed was at least 20 mph for 864

hours, mostly in the spring and fall seasons.

The early spring months hold significant promise

of frequent strong winds. The open terrain of

the proposed site makes it possible to conduct

experiments in a relatively simple terrain

exposure.

The field test facility is illustrated in Figure

2. The rectangular building with dimensions 30

ft x 45 ft x 12 ft will be placed on a concrete

pad. The unique feature of the field facility

is placing the building in a circular track so that it can be rotated. The internal structural

system of the building is designed to be supported on four wheels, with each wheel located

near each corner of the building. A circular

railroad track will be embedded in the concrete

pad such that the top of the track will be flush

Professor of Civil Engineering, Institute for

Disaster Research, Texas Tech University,

Lubbock, Texas 79409.

12
with the concrete surface,, Mechanical screw

jacks will raise the building and support it on the four wheels. The building will be rotated

manually to the desired direction and lowered to

let it rest on the concrete. Once a directional

position is obtained, the building will be

anchored using four anchor bolts. The building

will be constructed without architectural

features so that the external surfaces will be

as smooth and as simple as possible.

Pressures on the building surfaces will be

measured at a number of locations using pressure

taps. A pressure transducer inside the building

will monitor internal pressures. Locations of

pressure taps will be selected for each specific

data recording event.

The 100-ft tall meteorological tower will be

located approximately 150 ft from the building.

The tower will be a three-legged truss tower.

Wind speeds will be measured at four levels:

100 ft, 60 ft, 33 ft and 12 ft. Wind direction

will be measured at two levels: 100 ft and 12

ft. In addition, temperature, barometric

pressure and relative humidity will be measured

to assess stability of the atmosphere during the

data recording period. The speed and direction

measuring instruments will be placed on 6 ft

long booms. This distance will assure that

interference of the tower structure on winds

being recorded will be eliminated except for

wind direction that comes from the direction of

the tower. The wires from the instruments will

be brought down to the ground along the tower

and transferred to inside the building from

underground.

3. INSTRUMENTATION

Two distinct types of instruments are utilized

in the project, meteorological and pressure

measurement instruments. Horizontal wind speed

and wind direction will be measured using a Gill

three-cup anemometer and microvane #12002

supplied by Ralph M. Young & Co. One set of

these instruments has been acquired and is being

tested in the laboratory. One of the tests in

the laboratory is the assessment of accuracy

with a 250-ft of cable. If the tests of ac- curacy, linearity and stability are found to be

satisfactory, additional instruments will be

purchased. No decision has been reached for

instruments to measure temperature, barometric

pressure and relative humidity.

Pressure taps will be instrumented with differential pressure transducers. One of the

important characteristics of the differential

pressure transducer needed for this project is

the low range of transducer. The full range

needed is ±0.5 psi; this low range will provide

the desired accuracy of 0.005 psi. Two com- mercial companies are identified that can supply

differential transducers with this low range.

Validyne Engineering Company provides diaphragm

type pressure transducer with ±0.5 psi full

scale range (model DP45). This is a variable

reluctance diaphragm type of transducer. It is

rugged with steel housing and can be used for

gas or liquid. The range of the transducer can

be changed by simply replacing the diaphragm;

thus, the transducer is flexible. The disadvantage of this transducer is the cost. It is

in the range of $400-$500 per unit.

Omega Engineering, Inc. can provide a 170 series

differential pressure transducer with range of

±0.5 psi. The output for this transducer is

35mV ± 2mV. It uses a silicon sensor chip for

diaphragm material. Some of its characteristic

specifications are: linearity of ±1.5% of full

scale output; compensated temperature range of

32 to 122°F; thermal zero effect of ±3 mV; and

stability over one year of ±1.5% of full scale

output. Suggested cost for this unit is $60 per

transducer.

One transducer of each type is purchased. They

are being tested in the laboratory for accuracy,

stability and suitability for the project. In addition to the instruments mentioned above,

calibration equipment for wind speed and differential pressure is being reviewed for acquisition.

A critical item in differential pressure

measurement is the reference pressure. Two

methods of obtaining reference pressure are being checked: (1) pitot tube with wind vane,

and (2) opening below ground surface. Both

methods will be implemented in the field. The

one that is found to be more stable will be used

for reference pressure in differential pressure

measurements.

4. Data Acquisition System

Data acquisition will be controlled by a dedicated IBM XT computer which will be housed at

the test facility. It is equipped with hard

disk, monitor and graphic capabilities. It will

be programmed to automatically trigger data

collection and termination. In addition, it

will be programmed to dump data directly into a Bernoulli box equipped with a 20M byte removable

floppy cartridge. This system will be capable

of collecting and storing 6 hours of data for 36

channels at the rate of 10Hz for each channel.

The data will be collected in digital form using

a Metrabyte Dash-8 analog to digital converter.

It is planned to collect a 15-minute continuous

record, once the collection is triggered by the

computer. The computer will monitor wind for

one minute and trigger a new data collection if

the wind speed is above threshold value. The

threshold wind speed will be 20 mph.

The data acquisition system is designed to

obtain basic statistical properties such as mean

and standard deviation immediately after collection of data. In addition, it will provide

time-history plot to spot anomalies in any of

the data collection channels. The floppy

13
cartridge is removable; it will be brought to

the Texas Tech campus, where the data will be

put on magnetic tape through the Texas Tech

computer. Additional analysis of the da.ta will

be performed on the main-frame computer on the

Texas Tech University campus.

5. Conclusi on

This project, sponsored by the National Science

Foundation, should provide reliable field data

for wind pressure on surfaces of a 30 ft x 45 ft

x 12 ft building. Use and analysis of data will

depend on the reliability of the data. It is

anticipated that wind direction fluctuation and

non-stationary winds have potential of providing

new knowledge in wind pressures that is not

available through wind tunnel testing. Design

of the experiment that permits rotation of the

building should permit collection of wind

pressure data for specific angle of attack. The

field site and instruments should be in place

within the next six months.

TABLE 1

NATIONAL WEATHER SERVICE HOURLY WIND SPEED

DATA FOR LUBBOCK, TEXAS IN 1981

NUMBER OF HOURS BY WIND SPEED RANGE AND MONTH

Month

Wind Speed Range (mph)

20-30 30-40 40-50 Total

s

January 9 0 0 9

February 27 0 0 27

March 113 25 1 139

April 135 4 5 144

May 153 11 0 164

June 128 3 0 131

July 24 0 0 24

August 2 0 0 2 September 5 0 0 5

October 59 6 0 65

November 82 10 0 92

December 56 6 0 62

TOTALS 793 65 6 864

FIGURE 1. AERIAL PHOTOGRAPH OF THE FIELD SITE
100 FT HIGH

MET TOWER

FIGURE 2. FIELD TEST FACILITY
Aerodynamic Characteristics of Continuous Box Girder

Bridges Relevant to their Vibrations in Wind

N. Narita,

1 K. Yokoyama, 2 H. Sato 3 and Y. Nakagami4

ABSTRACT

The wind tunnel experiments concerning the wind-induced

vibrations of continuous box girder bridges are presented in this paper. The vibrations were measured both in smooth and

turbulent flow. Their characteristics and the effects of turbulence were investigated and explained in view-point of the aero- dynamic damping and the external lift force.

KEY WORDS : Aerodynamic Damping, External Lift Force, Continuous Box Girder Bridge, Wind-Induced Vibration.

NOTATION

B : width of the bridge deck

CL : lift coefficient D : depth of the bridge deck

f : frequency

fi : natural frequency of i th mode

fr : reduced frequency (= fB/U)

h : displacement of vertical bending vibration HLh . frequency response function between lift force and

heaving motion Im : imaginary part

Ji : joint acceptance function for i th mode

1

: bridge length

m : mass per unit length

Mi : generalized mass of i th mode (= m(z) 4> i 2 (z)dz)

qi : i th generalized coordinate for h

5 Power Spectral Density Function

U : mean wind speed Ur: reduced wind speed (= U/(fB) = 1/fr) u , w : fluctuating wind speed of longitudinal and vertical component

x : axis along the main flow

y : horizontal axis perpendicular to x

z : axis along the bridge axis

6 , 6 1 : logarithmic structural damping (of i th mode)

5a,6ai: logarithmic aerodynamic damping (of i th mode)

p : air density, a : reduced mass (=m/(pB 2 )),

<t> i i th mode shape

1. INTRODUCTION

Since the collapse ot the Tacoma Narrows Bridge, a great deal of

attention has been paid to the wind effects on bridges, resulting

in a great number ot studies on the subject. These studies have enabled the construction of long-span bridges to take place.

However, there is much that remains uncertain concerning the mechanism of the aerodynamic forces and wind-induced vibra- tions ot bridges. This is due to the complexity of the nature of

flow around a structure, as well as the complicated configuration of the structure.

In general, structures are apt to vibrate when the product of

their natural frequency and their representative leneth of the section is small compared to the wind speed at the site, and

when their mass and dampmg are small as well. Therefore,

research on wind effects is indispensable for construction of

long-span bridges, such as suspension bridges and cable-stayed

bridges, whose natural frequencies are very small. Since to in- crease mass is not economical as well as unfavourable due to the reduction of natural frequency, and to increase dampmg

is not practical, the most effective way to reduce wind-induced

vibrations is to select the bridge deck section so that it has favourable aerodynamic characteristics as well as sufficient

rigidity. Such sections are a truss deck with small solidity ratio and a streamlined box deck which has a iarge ratio of deck width

(B) to depth (D) (say, B/D>5). These bridge deck sections have been applied to most of the long-span bridges.

However, the recent trend of increasing the span length of con- tinuous box-girder bridges presents a new problem. The in- creased span length decreases the natural frequency of vertical bending mode to values as low as those of cable-stayed girder

bridges. To bear the static load by its deck alone, the depth of

the bridge section must be larger than that of a suspension or cable-stayed bridge, resulting in a small B/D, which is aerodynamically unfavourable. In Fig. 1 plotted are the changes of the

lift slope (dCF/da = dCL/da + CD) with B/D [ 1 ]

. The lift slope, dCF/da, can be considered as a representative value for the aerodynamic admittance and the aerodynamic derivative

in the low frequency range (or high wind speed range). The

aerodynamic damping would be negative for a bridge deck with

small B/D ratio. It may be correctly anticipated that the windinduced vibrations of such a bridge would be violent.

In the Public Works Research Institute, Ministry of Construction, Japan, several wind tunnel expenments have been co- ducted concerning the wind-induced vibration of continuous

box girder bridges. Among these studies, the followings are described in this paper.

1) the full model test for the proposed bridge design,

2) the sectional model test for the measurement of aero1 Dt. of Eng., Deputy Director General, Public Works Research

Institute. Ministry of Construction, Japan.

2 Head, Structure Division, Structure and Bridge Department,

P.W.R.I.

3 Senior Research Engineer, Structure Division. P.W.R.I.

4 Engineer, Local Road Division, Road Bureau, Ministry of Construction. (Formerly, Research Engineer, Structure Division,

P.W.R.I.)

16
mm

|

' '

'
dynamic damping of box gilder bridges, and

3) the single span elastic model test for the measurement of

external lift force.

2. FULL MODEL TEST FOR THE PROPOSED BRIDGE

DESIGN

The proposed bridge was a four-spanned continuous box girder

bridge (150 + 190 + 190 + 150m). B/D changes from 2.2 to 4 1 along th bridge axis. The general view of the bridge is shown in Fig. 2.

The 1/120 full elastic model was constructed in the boundary

layer wind tunnel at PWRI, whose test section is 6m wide,

3m high and 27m long. The stiffness of the model was simulated

by two steel spines, on which correctly scaled deck segments were mounted. The reduced generalized mass for the first mode

(Ml//o <t>j J dz)/(p B

J )) was 32. The logarithmic damping

in still air was 0.02 for the first mode, and somewhat smaller

for the higher modes. The natural frequencies of the model

and the prototype are shown in Table 1.

The vertical vibrational displacement at the midpoint of the center span was measured both in smooth and turbulent flow. The turbulence was generated by spires. The turbulence intensities Iu and Iw were about 12% and 8%, and the integral scales Lxu and Lxw were about 0.8m and 0.3m, respectively. The

measured vibrauonai displacements are shown m Fig. 3.

From the test results, it was found that the proposed bridge

had strong possibility of limited amplitude vibrations of 1st mode, 2nd mode, 3rd mode and 5th mode in the low wind

speeds, and divergent amplitude vibrations in higher wind speed

in smooth flow. On the other hand, the magnitude of limited amplitude vibrations decreased remarkably in the turbulent flow,

and the divergent amplitude vibration observed in smooth flow turned to be the less divergent but more random vibration in the turbulent flow.

3. MEASUREMENT OF AERODYNAMIC DAMPING

In general, wind-induced vibrations are caused either by self- excited forces (negative aerodynamic damping) or external

forces (approaching turbulence, vortex excitation). When the summation of structural damping and aerodynamic damping

is 0 or negative, the aerodynamic instability will take place.

The aerodynamic damping of the box girder bridges was measured in the Low-Speed Wind Tunnel-B of PWRI, whose test section is lm wide, 2m high and 3m long. The turbulence was generated by the coarse grid whose mesh size is 0.25m

and whose bar size is 0.05m. The intensities of the turbulence

lu and Iw are 6.2% and 5.0%, respectively, and the integral

scales Lxu, Lxw, Lyu and Lyw are 0.09m, 0.04m, 0.04m and

0.04m, respectively. The measurement was made by the dynamic balance developed at PWRI [2], The sectional models

were shaken in heaving mode with an amplitude of B/100.

The cross sections of the tested bridge deck models are shown

in Fig. 4. Model A corresponds to the cross section at 1/6 point of the center span of the bridge mentioned in Chapter

2. Model B is a rectangular prism whose B/D = 2. Model C is the typical streamlined box girder for cable stayed or suspension

bridges.

The self-excited lift force was transformed into mass-damping

parameter as follows.

a 6 a

= -Im (HLh) / ( (8 zr fr

2

) (0.5 p U

2

1 ) ) (1

)

In this transformation, the effect of aerodynamic stiffness was

neglected, and the uniform mode shape was assumed.

a 5 a of Model A in smooth flow is shown in Fig. 5.1. Posi

tive aerodynamic damping appears in high reduced frequencies

(fr > 0.6 ), and then negative damping appears in the lower fre- quencies (0.4 < fr < 0.5). The negative damping corresponds

to the limited amplitude vibration observed in the proposed

bridge design test. Then the sharp peak of positive damping

appears at fr = 0.23. The negative damping follows this positive

peak. The critical reduced frequency at which the dampmg

turns to negative is about 0.2, which corresponds to the onset of the divergent vibration observed in the proposed bridge design

test.

a 6 a measured in the turbulent flow is also shown in the Figure.

The pattern is similar to that for smooth flow, however, the

positive damping peak shifts to the lower reduced frequency

and the peak becomes broader in the turbulent flow. It results the decrease of the critical reduced frequency or the increase of the critical reduced wind speed where the aerodynamic

damping turns to negative. The aerodynamic damping for the thinner box girder models, whose B/D ratios (3.4 and 4.0)

correspond to the cross section at 1/3 and 1/2 point of the center span of the continuous box girder bridge mentioned in Chapter 2, was also measured. Their characteristics were similar to those of Model A. These results suggest that the less diver- gent but more randam vibration observed in the turbulent flow was caused by the external lift force but not by the negative

aerodynamic damping.

In Fig. 5.2. shown is a 5 a of the rectangular prism (Model B). The characteristics of aerodynamic damping of the prism seem

similar to those of the box girder bridge (Model A).

These characteristics of a 5 a of the box girder bridge are quite different from those of the streamlined bridge decx

(Model C). .As is shown in Fig. 5.3, the aerodynamic damping

of the steamlined bridge deck is positive both in smooth and

turbulent flow. The effect of turbulence on the aerodynamic

damping seems negligible for this kind of bridge deck.

4.

MEASUREMENT OF THE EXTERNAL LIFT FORCES

Since the aerodynamic damping of the box girder bridge bec- comes positive m the turbulent flow, the random vibration can

be explained by the external lift forces induced by the approaching turbulence or by the vorticies behind the deck. The

external lilt forces were estimated from the Power Spectral

Density Function (PSDF) of the wind-induced vibration of the

single span elastic model.

17
The experiments were made in the Low Speed Wind Tunnel-A

in PWRI, whose test section is 2.5m wide, 4m high and 10m

long. The three kinds of turbulence were generated by spires

and floor roughness. The characteristics of the turbulent flow

are shown in Table 2.

The span length, deck width and depth were 1.65m, 0.1m and

0.05m, respectively. The cross section and mass were uniform

along the span. The reduced mass and logarithmic structural damping were 33 and 0.01, respectively. The elasticity of the model was provided by the simply supported alminum spine,

and the fundamental mode shape of the model was half-sine. The cross sectional shape was similar to the box girder bridge

in Chapter 2, except that B/D (= 2.0) is a little smaller.

The wind-induced vibrations were measured both in smooth

and turbulent flow. The characteristics of the vibrations (Fig. 6) were similar to those described in Chaper2, however, the

effect of turbulence intensity on the rms values of the response was not simple. While the turbulence intensity increases in the order of turbulent flow 1, 2, 3, the rms value around the reduced wind speed 6 increases in the order of 2, 3, 1. The

aerodynamic damping estimated from the auto-correlation function of the vibrational displacement suggested that the

relatively large responses in the turbulent flow 1 were caused by the small aerodynamic damping.

The PSDF of the reduced generalized force for the first mode

was estimated as follows.

fSCLCL (0 I Ji(0 1 2 /1

2 = fSqiqi (0 (Mi (2 rr fi)2

) 2

/ (0.5 p U

2 B1 ) 2

/ 1 Hsi (0 1 2 (2)

I Hsi(f) | 2 = ((l-(f/fi) 2

) 2 + (

(i + 6 ai) (f/fi) /

7T ) 2 )" (3)

The results are shown in Fig. 7. Since the PSDF of the external

force has sharp peak in smooth flow, it seems that the primary

cause for this force is vortex-excitation. The peak reduced

frequency becomes smaller in turbulent flow than in smooth

flow. It seems that the turbulence broadens and lowers the peaks of the PSDF when the intensity is not so high (smooth

flow, turbulent flow 1 and 2), and that the turbulence increases the power of the external force in case of high intensity (tur- bulent flow 3).

5. DISCUSSION

The magnitude of the randam vibration of the box girder bridge

in turbulent flow is affected by both the external lift force and the aerodynamic damping. In the experiment described

in Chapter 3, the lift force on the models at rest was measured

in the turbulent flow as well as the aerodynamic damping.

The PSDF of the lift force is shown in Fig. 5.1 together with

the aerodynamic damping. From the figure, it can be found

that the large power of the lift force is associated with the

large positive damping.

It may be interesting to note the close correlation between

the PSDF of the external lift force and the aerodynamic damping plotted against reduced frequency for the box girder bridges.

First of all, the effects of turbulence are similar, namely, turbulence decreases the peak reduced frequency of both the lift force and positive damping, and turbulence broadens their peaks. Secondly, the peak reduced frequencies are almost

identical as is shown in Fig. 5.1. These correlations were also observed in case of the rectangular prism whose B/D - 2 . As

is shown in Fig. 5.2, the peak reduced frequency of the positive damping in smooth flow coindides with the reduced fre- quency of vortex shedding [3]. The frequency of vortex shed- ding is identical with the peak frequency of the PSDF of the

external lift force. In the turbulent flow, both of these peaks

shift to the lower reduced frequency, and they are almost

identical as is shown in the same figure. These findings suggest

that the vorticies behind the body are the primary factor for both the external lift force and the large positive dampmg in case of the box girder bridge deck.

The effects of turbulence on the wind-induced vibrations of the continuous box girder bridges are shown schematically in Fig. 8. The Power Spectral Density Function of the external lift force has a sharp peak in case of smooth flow. The turbulence shifts the peak to lower reduced frequency and broadens the peak.

The aerodynamic damping has a sharp positive peak in case of smooth flow. The negative damping follows the positive damping at the lower reduced frequency, which corresponds to the onset of galloping. The turbulence shifts the positive damping

peak to lower reduced frequency and broadens the peak, which

remarkably decreases the reduced frequency for the onset of

negative damping. The broad-banded lift force in the turbulent flow is associated with the broad-banded positive damping.

This results the less divergent but more random vibration in the turbulent flow.

6. CONCLUDING REMARKS

The wind-induced vertical bending vibrations of continuous

box girder bridges have been investigated through the wind

tunnel tests. The main findings are as follows.

a. The limited amplitude vibrations in the low wind speed

and the divergent amplitude vibrations (galloping) in the high wind speed were observed in case of smooth flow. They were caused by negative dampmg.

b. The areodynamic damping has a sharp positive peak in case of smooth flow. The negative damping follows the

positive damping at the lower reduced frequency, which

corresponds to the onset of galloping.

c. The turbulence shifts the positive damping peak to lower reduced frequency and broadens the peak, which remarkably decreases the reduced frequency for the onset of

negative damping.

d. The broad-banded positive damping in the turbulent flow is associated with the broad-banded lift force, which

changes the galloping into the less divergent but more random vibration in the turbulent flow.

e. The positive aerodynamic damping cf the continuous

box girder bridges was found to be closely correlated with

the external lift force primarily due to the vortex-excitation.

18
ACKNOWLEDGEMENTS

The authors would like to thank Mr. K. Kanzaki and Mr. M. Fukuda of the Structure Division on PWRI, who prepared for the experiments and processed the experimental data.

REFERENCES

1) T. Mizota and A. Okajima : Experimental Studies of Time Mean Flows around Rectangular Prisms, Proc. of Japan Society of Civil Engineers, Vol. 312, 1981 (in Japanese).

2) T. Okubo, N. Narita and K. Yokoyama : Some Approaches for Improving Wind Stability of Cable-Stayed Girder

Bridges, Proc. 4th Int. Conf. on Wind Effects on Bldgs,

and Sts., Heathrow, 1975.

3) K. Washizu, A. Ohya, Y. Otsuki and K. Fuji! : Aeroelastic

Instability of Rectangular Cylinders in a Her.ving Mode,

J. of Sound and Vibration, 59(2), 1978.

Table 1. Natural frequencies

1st mode 2nd mode 3rd mode 5th mode

mode 5.6 (Hz) 7.3 9.9 11.2 prototype 0.40 0.57 0.81 0.99 The 4th mode was horizontal bending.

Table 2 Characteristics of the turbulent flow

Turbulent

flow 1 Turbulentflow 2 Turbulent

fow 3 Power Exponent 0.12 0.16 0.30

iu 7.7% 11.4% 20.3% Iw 5.3% 7.4% 13.0%

Lxu 0.34m 0.35m 0.36m

Lxw 0.15m 0.22m 0.1 3m

19
dCF / da

WIDTH / DEPTH

Fig. 1 dCF/da for rectangular prisms (1)

h/B h/B

Fig. 3 The vertical displacement at mid-point of the center span

20
r-CD

ii Q

CD

O

Q

O

) o

o

OJ

B* - . . —

O'OCH

o

o

CD

CDO

O

oa

cc

a

Oe

o

CJi)

t/S o0

so

rO

9'6

—If*" OJ

I I II

z(ia

2 rWg')/nsj

i£2 o g o o o

<

VO
fSCLCL(f) | J 1(f)

2 / I 2

r. m . s

0.05-

0.04 -

0.03 •

002 -

0.01

0

-0

of h / B

o smooth flow

? turbulent flow 1 o

turbulent flow 2 o

A turbulent flow 3

° A A

12 3 4

A

0

^

5 6 7 8 Ur

Fig. 6 r.m.s. of wind-induced vibration

fB/U

Fig. 7 Generalized external lift force

EXTERNAL

LIFT

r 0RCE

AERODYNAMIC

DAMPING

//IND - INDUCED

VIBRATION

*

EFFECT OF

TURBULENCE

1 decrease the frequency

of vortex - shedding

2. broaden the spectral

peak

3 increase the Ur for positive damping peak

4 broaden the positive

damping peak

5. increase the Ur for the onset of negative damping

6. increase the Ur for the onset of galloping

7 less divergent but more random v i bration

Fig. 8 Schematic Explanation of the wind-induced vibration of the continuous box girder bridges

22
On Interpreting Model and Prototype

Responses of Suspended-Span Bridges to Wind

by

Robert H. Scanlan^

ABSTRACT

Aeroelastic models of suspended-span bridges have taken a vari- ety of forms over the last half-century. A prominent engineering

tendency over much of that time has been the direct identifica- tion of (scaled) model response under wind with expected prototype response. The present paper points out the fact that such

direct identification may likely be in error. Since most model action is inevitably at variance with prototype action, interpretation of experimental model response is required. Such interpretation

is facilitated via a viewpoint that refers to theory. An analytical

formulation is employed in which, notably, self-excited aerodynamic effects are recognized through mechanisms like flutter derivatives. Thereafter, by tracing expected specific aerodynamic

and structural-dynamic contributions to both prototype and

model response, the relation of the latter to the former can be

assessed. This process sharpens the value of aeroelastic model

tests of bridges.

KEYWORDS

Aerodynamic; Aeroelastic; Bridges; Structural-dynamic;

Suspended-span bridges; Wind engineering.

1. INTRODUCTION

In problems of experimental simulation it is invariably of great

assistance to establish scaling ratios between the prototype that

is to be understood and the physical model that is intended to represent it. When the modeling may be considered as “complete” in all respects, an application of the principles of dimensional analysis alone can often identify those parameter groups

that govern the action of the prototype. However, when the

physical modeling is “incomplete” in some respects it may be- come necessary to explore the system under study more thor- oughly by first establishing a mathematical model for it and then

noting the roles of various nondimensional quantities in the action of the prototype. Thus, even if experiments fall short of

complete simulation, the several physical mechanisms involved

in them can be studied via partial models for their contributions

to the action of the prototype system as a whole.

In the wind tunnel testing of suspended-span bridges, a useful point of view is that no reduced-scale model perfectly duplicates

prototype action in all respects. Thus, invoking a mathematical

model of the system can help assess the accuracy of inferences

drawn from experiment about the prototype. This point of view

even enhances the critical interpretation of models initially in- tended to be “complete.” It serves to identify participating

mechanisms within the total system.

When constructing a mathematical model of bridge dynamic action under wind it is convenient conceptually to separate off the

purely structural part from the purely aerodynamic part. The

former is usually adequately described by a set of the eigenmodes and corresponding eigenfrequencies of the prototype,

which merely reflect in alternative form its mass and structural

stiffness distributions. For the aerodynamic part, some ingenuity

has been called for, and this area still comprises a focus for continued effort

To date, however, the measured steady-state force and moment

coefficients of bridge deck sections, their corresponding oscilla- tory-state flutter derivatives, and the concomitant aerodynamic

indicia! and admittance functions, have fairly successfully been

employed in studying the mechanisms of prototype response to wind. In fact, use of these concepts, usually derived from intentionally abridged, or partial, experiments, has now led to a state of the art in which a critical review of wind tunnel testing practices regarding bridge models can be undertaken. Such a review

is given in what follows.

2. THEORY OF PROTOTYPE ACTION

We consider the case of a flexible, suspended-span bridge under

high-velocity turbulent wind. The turbulence randomly excites the deck, which responds in a random fashion and also undergoes some degree of aeroelastic interaction, the net spectrum of

the attendant response exhibiting peaks at frequencies respectively in the neighborhood of the natural eigenfrequencies of the

structure. (The aeroelastic forces may affect the exact values of

these frequencies.)

We first consider the interactive forces related to the three de- grees of freedom (cf. Fig. 1): h (vertical), a (twist), and p (lat- eral) of a typical deck section. These actions take the

well-known linearized forms [l]:2

Lilt: Lae = |

p

U

2B [kh;<K)4+ KHjfK)!? + K2H;(K>] (1)

Moment: M -|pU2B

2 KAJ(K)4 + KaJ(K^ * L u u

+ K2A5(K)a] (2)

Drag: Dae - \ p02B [kpJ(K)£ + KP?(K)^ + K

2 p:(K)a] (3)

1 Department of Civil Engineering, Johns Hopkins University,

Baltimore, MD 21218.

2 It should further be noted that an alternate formulation of the

aeroelastic forces under random excitation is via the so-called indicml function approach [5]. This will not be followed here.

23
'
where

p = air density

U = mean cross-wind velocity

B = deck width

K = B<o/U= reduced frequency

Hj , and P* (i= 1, 2, 3) are well-known flutter derivatives , experimentally determined functions of K for some specified

bridge deck; and it is presumed that h, a, and p are all undergoing sinusoidal oscillation at radian frequency g>. [Figs. 2 and 3, for example, represent representative experimentally obtained

plots of Hj (K) and A2 (K) vs. 2tt/K.]

It is worth emphasizing here that the complete flutter derivatives have the form KH*, KH2

, etc. and that a pair like KHj

(K) is in fact an inseparable product that holds_for any given

value of 2Tr/Ko Thus, at some mean velocity U if the degree of

freedom h oscillates according to

h = hj sin u)^t + h

2

sin <u 2t (4)

then the corresponding self-excited lift force component is

Lae =! pU2b[k1h;(K1)=P cos Ult

hn<i)o + K2Hj(K2>-4p cos o)2tJ

(5)

where

K1 = B(o

1 /U,

K2 = Bc^/U,

o)|, cj2

being the respective radian frequencies of the two oscillation components.

If the deck response components are introduced in terms of the superposition of dimensionless modal contributions h'(x), cq(x),

p^x), then

h (x,t)-X €i(0 hi(*) B (6)

i a(*.t)-S«i(t)«i(x) (7)

i PM-2«i(t)Pi(x)B (8)

i where £j(t) is the generalized coordinate of mode L The aeroe1 astic part of the sectional lift is then

. rSKjH^WhjB

Lae (x)=ipU-B-[J =

+ J = +? j- (9)

Analogous forms hold for and Dae from (2) and (3).

Next consider the sectional buffeting terms. These will have the form

Lb (t)=|pU2BL(t) (10)

Mb (t)=ipU2B

2 M(t) (11)

Db (t)-ipU2BD(t) (12)

where L, M, D are respectively the dimensionless random lift, moment, and drag acting on the deck section. Their exact forms

will not be specified in the present paper, but may be expected

to be experimentally measured results. We then define the total actions on a deck section to be

Lift: L — Lae +Ljj (13)

Moment: M= Mae + Mjj (14)

Drag: D = Dae +Db

(15)

This results in the following expressions:

U,,)4pU2b{?[(^W\WC*j(x)B\ .

+ K?H5(Kpa

j (x)€

j ]+L(x,t)} (16)

1 , r ^ r/K

i AI(Kj )bj (x)B

M(x,t)= 2 pUV|jLv 0

K

jAj(Kj)a

j (x)BN. I + 0

j5i+ KjA3(Kpa

j (x)5

j J

+ M(x,t) r (17)

l -2 /v[yK

i pi(Kj)pj (x)B Kj p2<Kp“j«BY D^-apuMfK o

+

0 >i

•Kj^KjJajOOSjJ+DCv)}

The i th generalized coordinate ^responds according to the equation

!

i [^i

+

2

5i“i5i +“?{,] = Qi(0 (19)

where Ij is the generalized inertia of the entire bridge structure

in the i th mode, its damping ratio, and wj its circular fre- quency; the generalized force is

>i- fdcclc [L f|+Mffi +D

i|:]

dx (20)

The explicit form of Qj becomes

Qi(t>=|p02B

2 [(2*^

KjH2(Kpaj(x)B\ 1 + 0

jij + KjfHSCKpajCx^jJhjCx)

( v

r/K

jAJ(Kj)hj(x)B

^ K

j A2(Kj)aj (x)B\ +

j L\ o

+ 0

2 . 1 v r/K

j p i (Kp

pjWB

+ Kj A5(K

j)a

j (x)Cj J

ai(x) 4- % J.

J —

+ “ -=p ) €j + KjP3

(Kj)otj(x)€j

J

Pi(x)

+ L(x,t)h

i(x) + M(x,t)ai(x) + D(x,t)p

i(x) }

dx (21)

By carrying out the spanwise modal integrations that take the form

Gr s =

r Jd m5n ^ deck rm

(x) sn

(x) dx (22)

24
where r, s=h, a, or p; m, n— i or j, the following result can be

obtained for the generalized force:

2

+ A;(K

j)Ghjai+ A5(Kj)Ga . a. + P

,

1(K

j)G

pjp

.

+ P^)Ga.p.)ij + K

J

2 (H5(Kj)Gajh .

+ A5(Kj)°a

j ai + J

>3(Kj)Ga

jPi)«j]

+ J deck [L(x,t)h

i(x) + M(x,t)a

i(x) + D(x,t)Pi(x)]dx

(23)

Versions of this result appear in Ref. [2].

It has been assumed in arriving at (23) above that the aeroelastic

derivatives do not vary over the span, Le. that the deck section

geometry is uniform spanwise, and that the mean wind velocity

is also uniform over the spam When these conditions are not met

it is more appropriate to employ the more general eq. (21) for

Qi(t)<

3. REMARKS ON THE NATURE OF THE EXCITATION

Based on eq. (23) a number of observations can be made. First,

all bridge modes are excited by the turbulence; and, as a general

feature of the system, all modes are coupled together in the

equations of motion through the medium of the aeroelastic

derivatives and the modal integrals (22). The strength of this coupling will thus depend on the relative magnitudes of these

terms. The aeroelastic derivatives are functions uniquely of deck

cross-sectional geometry, while the modal integrals are conse- quences of the distributions of mass and stiffness in the bridge

structure.

Arguments of this sort (backed by numerical evaluations) can

serve to delineate those modes most likely to participate in windinduced aeroelastic response. Note that modes 1, 3, 9, 10, 12, and

13 of the example have over 93% of their inertia located in the

deck alone, whereas this is quite variable for other modes. Mode

1 dearly has the character of lateral sway; modes 3, 9, 10, 12 are mainly vertical modes; and mode 13 is the first strongly torsional mode. (This mode is particularly likely to pick up wind energy

through the agency of the A^ coefficients.) For modes like these the deck action represents most of total bridge action. They may

be expected to dominate the response.

The possibility that in certain cases the cross-modal integrals

(22) may have weak values suggests that on occasion intermodal

aeroelastic coupling may not be strong and that in such cases eq. (23) may reduce further to an uncoupled, single-mode form,

namely:

Qj BKj/ \

pi

^.^(H;(Kj)Gh. h . + A2(Ki)Ga. ai + P;(Ki)G

pip .j 5i

+ K?A|(Kj)Ga.ajSi (24)

implying the possibility of single-mode aeroelastic response.

If one wishes provisionally to consider single-mode flutter in mode i, either as an actual possibility or as a necessary bounding

case, this must imply that, in eq. (19), the mechanical damping

term on the left-hand side is to be balanced by the aerodynamic

damping terms of the right-hand side. Further, the natural fre- quency o)

j of the mode is modified to a value S3

j by the

aerodynamic stiffness (A^) term. This situation leads first to the frequency modifying relation:

l +‘fr A5(Ki)Ga.0.

(25)

and next to the criterion FC for single-mode flutter instability:

In some of the modes of a structure the deck plays a dominant

role, carrying a large percentage of the generalized inertia of the

whole system. If the deck structure is uniform over the span in these structures the modal integrals (22) may in such cases be ap- proximately proportional to the well-known conditions of or- thonormality between modes. In this situation some of the

integrals (22) may approach high (“unit”) values while others

approach zero values.

An investigation of the first 13 modes of a modern cable-stayed

bridge was made as an example, and two items of interest were

recorded for each of these modes:

1) the percentage of total bridge generalized inertia concentrated in the deck alone;

2) values of those integrals G. . (22) in which r = s and

rnrn

n — m, Le. values dependent on same-mode contributions.

Based on the convention that all generalized inertias 1^ were as- signed unit value, results for the ratio of deck contribution

(tydeck to total h

value’ “d ^

integrals G^, G^, G

pip

„

are presented in relative units in Table 1. This table serves to

identify a) those modes in which prominent deck action is to be

expected; b) the character (vertical bending, torsion, or lateral sway) of each mode. Further, it can be expected that for modes

with high deck inertia and strong, single-component modal character, the cross-modal integrals (22) will take on less significant values. Some may be negligible.

FC=^^LHi^G^+A2(Ki)G

«i«i

+Pi(Ki)0

PiPi]

S1

'

' (26)

in which it is understood that Kj=Bc3|/U.

Finally, the action of wind turbulence may be discussed. First, it has been verified [3] [4] that the presence of turbulence affects the values of the sectional aeroelastic derivatives H*, A?, P*. Hence the values of these quantities as measured under appropriate turbulent conditions should be employed in the analysis.

Next, the general action of turbulence is to force all bridge

modes into play, as is clear from eqs. (6)-(8), (19), (23). When

modes that are more highly damped are excited by the turbulence, their aeroelastic effects couple with others of the system

and extract energy from it, effectively delaying the onset of in- stability. This appears to account for the often observed extension of critical flutter velocities of full-bridge models to higher

values under turbulence.

A commonly held view is that two modes — typically a pure

bending and a pure torsion— must couple to produce flutter. This phenomenon defines, in fact, the kind of stiffness-controlled “classical” flutter observed in aircraft, where the aerodynamic

forces at high speed are large enough to alter two structural fre- quencies and modes appreciably toward coalescence into a flut- ter mode. While this type of occurrence is in principle possible

25
for bridge decks, and is covered by the theory given, by far the

greater number of actual occurrences depend upon a single tor*

sional mode becoming intrinsically unstable and “driving” the

rest of the system. The mechanism involved in this can immediately be detected by evaluation of the terms in eq. (23), most

particularly those involving the A2coefficient, which can lead

to negadve system damping.

In many cases the question of stability may be regarded as centered therefore on the values taken on by the A2

flutter derivative, which typically evolves from negative to positive with

increasing values of U/NB = 2ir/K. Modes of higher frequency,

corresponding to lower values of U/NB, correspond generally

to negative values of A^, whereas lower-frequency modes ap- propriate positive values of A^, tending toward instability. It is thus quite clear how turbulence, coupling-in many modes that are conducive to greater damping, extends the stability domain.

3.1 The Roles of Different Model Types

Common model types include narrow- and wide-span section, the single-span flexible, and the full-bridge aeroelastic. The roles,

results, and interpretations of each can be assessed in light of the

theory developed above.

The narrow-span section model is the model of choice for most

investigations. Tested under l aminar flow it yields the primary

flutter derivatives and static force coefficients that are usually

the most conservative ones. It is usually diffi cult, however, to

test the section model under conditions of properly simulated

turbulence. This problem has, however, been approached and

largely overcome in recent work [3] [4].

One of the largest open issues is the extent to which prototype

action should be inferred from model results. The rigid, springsupported model can represent only restricted freedoms; and

general prototype modal effects therefore cannot be fully duplicated by it. Under these conditions, model response to turbulence will not duplicate that of the prototype.

When the short-span section model is used to acquire section

aeroelastic derivatives experimentally it can conveniently be di- vorced from all but geometric simil arity requirements relative to the prototype. Since the aeroelastic derivatives are functions of

reduced velocity U/NB only (as with linear-theory derivatives), or of U/NB and response amplitude only (as with problems in- volving nonlinear theory), the geometric section model need

only respect these parameters, and is thus rapid, economical, and

relatively simple to test. It most often must be tested in any

event, whether or not models of other types are employed.

The section model, with its restricted, rigid modes, can be fitted directly into the general theory outlined above [2]. It can re- spond essentially in separate components of two distinct prototype modes at a time. When its parameters are employed with

the theory, the exact degree to which it meets (or fails to meet)

prototype action can be completely assessed, without ambiguity.

Its chief role remains, however, in assessing intrinsically aerodynamic and aeroelastic data.

The wide-span section model acts in every respect like the nar- row-span version but has the added feature of being wide

enough to include the integrated spanwise variations of the tur- bulence effects, both on the net flutter derivatives and on the

buffeting response. However, this model must also not be mis, taken to have the same response as the prototype. Model action

is again restricted essentially to two “pure” prototype modal

components. Entry of the modal and other data into the analytical framework given earlier again reveals clearly that model response is not directly equal to scaled prototype response, even in a single mode. This is obvious from the differences that occur in the modal integrals Grmsn

that appear in eq (22) for the model

and prototype. At the same time, the full appreciation of these

differences permits correct interpretation of model performance

under turbulence relative to expected prototype performance.

For example, analysis of model and prototype performance un- der random inputs [2] can yield the ratio of r.m.s. model re- sponse amplitude to expected prototype response in a single

mode of the latter. This ratio is not unity, even apart from

known scaling factors.

The single-span flexible model (often realized via the so-called “taut strip” technique) has identical problems to those of the narrow- and wide-span section models, in that it cannot be ex- pected to duplicate prototype modes, full-bridge inertias, or damping. For the simulation of cable-stayed bridges in particular, the fact that it develops a mode or modes of its own is actually a disadvantage, rather than an advantage, since these modes

cannot reproduce complex cable-stayed modes. In this respect

the wide-span, rigid section model is in fact superior, its “mode”

being simpler and more directly adapted to the extraction of

flutter derivatives.

The reasoning that leads to use of a single-span, flexible model

has often been faulty, to the extent that the effort has been to set up model performance that could be interpreted directly, on a one-to-one scale, as equivalent to prototype performance. Such

intended interpretation is illusory. If, in particular, the motivation for a single-span flexible model is to gather in and integrate the spanwise effects of turbulence, then interpretation beyond

the one-to-one scale is even more called for, since the model acts

differently from the prototype, as seen, again, via the theory outlined earlier.

In particular, referring to the modal integrals (22), these will be

vastly different from model to prototype. The deck generalized

inertia will not in general duplicate (to scale) full-bridge generalized inertia. Again, only a single (or perhaps two) center-span

modes are possible in the single-span flexible model. Inertias and

loadings of side spans are not represented. In the case of modern

cable-stayed bridges the shapes of the modes realizable by singlespan flexible models usually do not conform to prototype modal

forms.

The full-bridge aeroelastic model , assuming it to be aeroelastically

faithful, is the only one with a reasonable chance of directly duplicating prototype performance (to scale) on a one-to-one basis. For this it must be tested under correctly simulated earth’s boundary layer winds. Exploiters of such models have encountered the expense, care, and general difficulties involved with

them. In particular, their realizable hierarchy of structural vibration modes seldom lines up perfectly, one-for-one, with prototype modes. This line-up may be partiall y achieved through as many as 10 modes, though rarely. When it is recalled that a

wind-critical torsional mode of a modern cable-stayed bridge

may be the 13th or 17th mode of the structure, this leaves some

open problems.

Other problems of these model types are that they integrate all effects together, permitting less experimental appreciation of

specific local mechanisms that may contribute importantly to overall performance.

22 General Philosophy of Modeling

Throughout the history of bridge wind-tunnel modeling there has been a tendency on the part of investigators to view the

26
model as a direct, scaled analog of the prototype. As with any

other avenue of investigation, what is objectively discovered

may not always conform to preconceptions. In the case of

bridge modeling what has been discovered is that the essence of

what can be extracted from models is not their performance

perse but the intrinsic aerodynamic information that they engender, i.e., such data as the static and aeroelastic derivatives, indicial mid admittance functions, etc.

The desire to “command” the model into scaled dynamic equivalence to the prototype has led to extensive and needless concern with model structural dynamics, equivalent inertias,

artificial rotation points, damping, etc. While no model can duplicate prototype behavior perfectly, as has been argued, numerous attempts continue to be made to bring model performance

into such alignment. A much more fruitful approach is to concentrate on exactly what can be inferred correctly from model

tests. Certainly the static force coefficients and the aeroelastic

derivatives are prominent among such items of data. These, coupled with theory, offer a wide range of options regarding the

forecasting of prototype performance. This approach allows the

aerodynamicist to concentrate attention more upon his principal

art and less upon structural dynamics.

When costly attempts are made to encompass the entire problem— both structural and aerodynamic— via a full-bridge aeroelastic model, for example, a large set of parameters, including

many structural ones, is “invited” into the problem. Their totality may lessen the validity of conclusions drawn from experiment rather than augment them. A problematic situation that can

occur under these circumstances is that a “favorable” but unexplained outcome emerges from the testing. From an engineering

viewpoint the cause for this must ultimately be found, but a “global” test, in which too few detailed mechanisms have been

explained, may offer a false sense of security.

The analytical and directed experimental approaches outlined in

this paper have emphasized the role of theory in discerning the

mechanisms responsible for the stability of suspended-span

bridges. It should be noted that the theory may be considerably

extended [2] beyond the limited scope presented.

4. REFERENCES

1. Scanlan, R. H.: “The Action of Flexible Bridges Under

Wind. I: Flutter Theory; II: Buffeting Theory,” Jnl. of

Sound and Vibration, Vol. 60, No. 2, 1978, pp. 187-199 and

pp. 201-211.

2. Scanlan, R. H.: “Interpreting Aeroelastic Models of CableStayed Bridges” to appear in Jnl. EMD, ASCE , April 1987.

3. Scanlan, R, H. and Huston, D. R.: “Sensitivity of Bridge

Decks to Turbulent Wind” Proceedings Asia-Pacific Symposium on Wind Engineering, Roorkee, India, Dec. 1985, pp. 167-172.

4. Scanlan, R. H, and Huston, D. R.: “Changes in Bridge Deck

Flutter Derivatives Caused by Turbulence” Dynamic Response of Structures (Eds.: G. C. Hart and R. B. Nelson)

ASCE, 1986, pp. 382-389.

5. Scanlan, R. H.: “Role of Indicial Functions in Buffeting

Analysis of Bridges,” Jnl. Struct. Engrg. ASCE, Vol. 110,

No. 7, July 1984, pp. 1433-1446.

TABLE 1 Modal Geometric Integrals and Deck Inertias Mode

No.

Freq.

(Hz)

Ii (deck

only)

Gaiai °hihi GPiPi

1 0.171 0.992 0.1816 E-07 0.1414 E-ll 0.2242 E-06

2 0.188 0.781 0.3354 E-ll 0.0896 E-07 0.4358 E-10

3 0.270 0.935 0.3193 E-13 0.2105 E-06 0.2064 E-12

4 0.349 0.890 0.3614 E-12 0.2010 E-06 0.1662 E-ll

5 0.389 0.894 0.4876 E-07 0.1685 E-ll 0.1975 E-06

6 0.516 0.050 0.7846 E-07 0.3021 E-13 0.0739 E-07

7 0.522 0.062 0.9214 E-07 0.3644 E-13 0.0795 E-07

8 0.552 0.694 0.5313 E-07 0.0729 E-09 0.1317 E-06

9 0.555 0.947 0.1778 E-10 0.2133 E-06 0.4478 E-10

10 0.645 0.975 0.54S4 E-14 0.2201 E-06 0.3812 E-13

11 0.696 0.653 0.1239 E-06 0.4637 E-12 0.1130 E-06

12 0.714 0.988 0.2264 E-ll 0.2238 E-06 0.1817 E-12

13 0.721 0.955 0.3640 E-05 0.1375 E-12 0.2006 E-08

i

27
28
29
Theme II

Earthquake Engineering

The United States Geological Survey

Accelerograph Network in the U.S.-Operations,

Record Processing and Research

by

A. Gerald Brady^

ABSTRACT

The United States Geological Survey maintains

a strong-motion instrumentation permanent

network of accelerographs for recording potentially damaging strong accelerations. At the

project level, two projects are responsible

for the design, development and operation of

the network, and the processing of the

resulting ground motion data. Five projects

are closely linked to the performance of the

network and the processed data. The seven

projects, and the relationships between them,

are described in terms of their recent

ac c omp 1 i shment s

.

KEYWORDS: Accelerograph; Network; Operation;

Data processing.

1. INTRODUCTION

The Branch of Engineering Seismology and

Geology, within the Office of Earthquakes,

Volcanoes and Engineering, of the U.S. Geological Survey, may be considered to consist of

approximately 40 research projects, plus the

administrative and data handling projects es- sential for operation. All research projects

are headed in general by Ph.D. scientists.

The projects deal with the reduction of earthquake hazard from an understanding of the

causative faults, the passage of earthquakegenerated strong-motion waves to the surface,

and the effect on structures of strong ground

motion at the surface. The researchers

involved are therefore readily subdivided into

geologists, geophysicists and engineers,

although there is considerable overlap.

Many measurements of earthquake related

phenomena are taken — continuous in analog or discrete modes for monitoring purposes, and

intermittent in analog or discrete modes for

specific earthquake occurrences. The strongmotion instrumentation network, characterized

by permanently-installed instruments for

recording potentially damaging strong accelerations, is one such source of research data.

Seven projects are directly related with the

development and operation of this permanent

network, and with the strong-motion records it

produces, while several more projects use some

of the more significant record sets for re- search endeavors that do not form an essential

part of their basic project goals. This

report describes briefly the two projects that

run the network and process the data, together

with recent accomplishments. It follows with

a description of the five projects whose goals

are closely linked to the performance of the

primary two.

2. NETWORK PERFORMANCE AND RECORD PROCESSING

2 . 1 Strong-Motion Instrumentation Network

Design, Development and Operations

This network was initiated by the U.S. Coast

and Geodetic Survey in 1932 and recorded the

first accelerations close to a destructive

earthquake on March 10, 1933 (the Long Beach

earthquake). At the present time, the network

operates in 41 states and in Puerto Rico, and

consists of approximately 1000 recording units

installed at 600 ground sites, 27 buildings,

five bridges, 56 dams, and two pumping

plants. Cooperative arrangements for the

design, development and operation of many of

these installations are in effect with several

federal, state, and local agencies, and

advisory engineering committees. Cooperative

agreements with federal agencies ensure a spread of the network into some regions where

the return, in the form of significant

records, is not high. The densest parts of

the network include the metropolitan and

active fault areas in California, followed by

the seismically active areas of the remaining

western states, Alaska and Hawaii. Arrays of

closely spaced sensors, with typical spacing

of 100 m, and of less closely spaced sensors (spacing of 1 to 5 km), have been installed in

California locations.

Program goals include (1) the recording of

potentially damaging ground motions to be used

as input by the structural engineering, and

geophysical, professions in efforts to reduce

the earthquake hazard through a greater

understanding, and subsequently, more precise

building codes and regulations, and (2)

monitoring the earthquake response of civil

engineering structures with sensors placed at

critical locations. Hopefully, significant,

or even damaging, earthquake response would be

recorded

.

The following recent project activities have

been taken in their entirety from the 6-monthly technical summary reports of the National

Earthquake Hazards Reduction Program (Ref. 1).

2.1.1 New Instrumentation

Six ground motion stations were established in

1 U.S. Geological Survey, Menlo Park, CA

94025

33
the Western United States: three at rock

sites in San Francisco including the Golden

Gate Bridge abutment, one on the Calaveras

fault east of San Jose, and two in western

Nevada, at Yerington and Montgomery Pass.

Two four-instrument structural arrays were installed at the Corps of Engineers' John Day

and Dalles concrete dams located on the Columbia River between Oregon and Washington. At

each structure the instrumentation configuration consists of two accelerographs in the

lower gallery, one in the upper gallery, and

one on abutment rock.

A 21-channel instrumentation system was completed at a 33-story steel frame building

in Los Angeles. The structure is rectangular

for the first 12 stories above ground, and is

then topped by a 21-story triangular tower.

Transducers are located at the basement level,

and on the ground, 12th, 13th, and 33rd floors

in directions to measure base rocking, two

translational modes in each direction, and two

torsional modes.

A 15-channel structural monitoring system was installed for the Metropolitan Water District

of Southern California on a truss bridge that

was recently retrofitted with seismic isolation elastomeric bearings. The bridge has

three 180-foot long spans (55 m) that carry a 10-foot diameter (3 m) California Aqueduct

water pipe across the Santa Ana River near Riverside. Transducers have been located in

the middle of one truss section, at a pier

above and below the elastomeric bearing, in

the bridge abutment, and in a nearby recorder

housing resting on rock.

2.1.2 Recent Earthquake Records

Several hundred earthquake records were recovered during the past six months from

instrumentation located in California and

Alaska. The following summarizes some of the more important results.

Earthquake Date Magnitude Location

30 December 1985 5.2 Anchorage, AK

2 . 2 The National Strong-Motion Network: Data

Processing

Data processing of significant records from

the strong-motion instrumentation network provides computer-ready corrected accelerograms

to the engineering and geophysical research

community. Tapes containing this data, and

their spectra, are distributed by the National

Geophysical Data Center. Reports containing

computer plots are distributed by the U.S.

Geological Survey. Accompanying data on all

records, stations, and recorded events are available on a recently updated database operated on the USGS VAX computer at their Menlo

Park Western Region Headquarters (Ref. 2).

In its capacity as data processor for the USGS

strong-motion instrumentation network, the

project remains current with all recent processing procedures, particularly those in use

by other closely associated projects within

the USGS. Cooperative processing arrangements

have been made in the past with Italy and other countries in Europe, with Chile and other

countries in South America, and with Mexico

and other countries in Central America.

Local site geology investigations at the

strong-motion stations are shared by several

projects, but the relevant information from

bore holes, and velocity profiles, are intended to be centrally gathered in the

section of the database describing the

stations (Ref. 2).

2.2.1 Recent project activities

In cooperation with strong-motion network

personnel in the Civil Engineering Department

of the University of Chile, some of the more

distant records of the Chile earthquake

(Ms7.8, 3 March 1985) have been digitized and

processed. The important records with high

amplitudes were processed in Chile. One

significant record from Adak, Alaska during

the Alaska earthquake (Ms7.7, 7 May 1986) has

Records Peak Acceleration

11 Ground .08 g Structure .05 g

(Humana Hospital

7th floor)

8 July 1986

13 July 1986

5.9 N. Palm Springs, CA 68* Ground 00

r^. g Structure .12 g

( Skinner Dam

crest)

5.3 Offshore off 3 Ground .11 g Oceanside, CA

21 July 1986 6.0 Chalfant Valley, CA 16* Ground .36 g

*More than 100 aftershock records were obtained from permanent and temporary

acce lerograph stations following the North Palm Springs and Chalfant Valley

earthquakes

34
been processed for the U.S. Navy. Two of the

California earthquakes in July 1986 produced

significant records —North Palm Springs

(M5.9, 8 July 1986) and Chalfant Valley, near

Bishop, (M6.4, 21 July 1986). A set of 17

records from the North Palm Springs event are being processed, while the most significant

records from the Chalfant Valley event are from a downhole array ( 100 m deep) at Mammoth

Lakes

.

A user manual and a maintenance manual for the

strong-motion database have been prepared, for use by USGS database personnel, even though

the database itself is operating at present at

its basic level.

3.

PROJECTS RELYING ON NETWORK OPERATIONS AND

DATA PROCESSING

Three projects rely on the Network Operations

project for installation and maintenance of

accelerographs performing specific functions.

These stations have been incorporated into the

permanent network. The projects also rely on

the Data Processing project for processing any

records forthcoming from the installations.

3 . 1 National Strong-Motion Network:

Engineering Data Analysis

The El Centro Differential Array recorded the

1981 Westmorland earthquake (M5.6, 26 April

1981) and the Hollister Differential Array

recorded the 1986 Hollister Earthquake (M5.5,

26 January 1986). These arrays, with station

spacing of the order of 100 m, were designed

for investigations of differential displacements and their effect on structures with

large foundations or widely spaced supports.

Differential displacements and spectra, from

records of the two events, have been calculated (Ref. 3)

.

3 . 2 Experimental Investigations of

Liquefaction Potential

This project involves the design and installation of instrumental packages at a chosen

site near Parkfield, California, to monitor

pore water pressures in buried sand layers,

and strong ground motion in a downhole array,

during the predicted Parkfield earthquake,

should it occur. Four downhole accelerometers

will be at depths ranging from 3 to 30 m, in

addition to ground level accelerometers.

Their installation and maintenance have been

coordinated through the Network Operations

project and digitizing of any records,

including the pore pressure records, will be

coordinated through the Data Processing

project (Ref. 4).

3 . 3 Instrumentation of Structures

In the last several years the USGS has established a program where advisory committees,

formed in the more seismically active areas of

the country, develop lists of structures for

possible instrumentation. The recommendations

of committees in the San Francisco Bay Region,

San Bernardino County, and Los Angeles County,

in California, and in Charleston, South Carolina, have resulted in thorough instrumentation of at least one structure in each region

(Ref. 5). Further committee recommendations

are expected from Orange County, Calif.;

Seattle, Wash.; Reno, Nev. ; the New Madrid

area, Missouri; Boston, Mass.; Anchorage,

Alaska; Hawaii; and Puerto Rico.

The well-instrumented structures become part

of the permanent network, and any records from

future events will be processed accordingly.

Records are expected to provide useful

information for:

. checking the dynamic model, for the first

two modes in each direction, and for

torsion and rocking;

• investigating non-linear effects;

• correlating non-linearities and inelastic

behavior with damage;

° correlating ground level motion with

structural properties and damage;

• the improvement of seismic codes.

4. OTHER SOURCES OF ACCELEROGRAMS

4 . 1 Strong-Motion Accelerograms of the 3 March 1985 Central Chile Earthquake .

Strong-motion records from the more distant

stations during the 1985 Chile earthquake, and

records from the significant aftershocks, were digitized and processed by the Data Processing

project. The near field records were digitized in Chile. A staff member from the University of Chile visited Menlo Park to assist

in the processing. At the completion of all

processing, a data tape and report will be

prepared (Ref. 6), and the data will be

available for use both in Chile, the USGS, and

on request.

5. RECENT AND CURRENT RESEARCH PROJECTS USING

PROCESSED DATA FROM THE STRONG-MOTION INSTRU- MENTATION NETWORK .

5 . 1 Fundamental Modal Behavior of an Earthquake-excited Bridge

The 1984 Morgan Hill earthquake (M6.1, 24

April 1984) produced a set of acceleration

records at the freeway overpass in San Jose,

carrying I-280/I-680 across U.S.101, 12 km

from the epicenter. Peak accelerations

reached 0.1 £ at ground level and 0.16 _£ in

the box girder. The integrated displacements

at frequencies higher than 3 Hz permit the

identification of the fundamental modes and

frequencies. The horizontal modes, which

include dominant ground level motions up to

90% of deck motions, have frequencies of 3.8

35
and 3.9 Hz (Ref. 7).

5 . 2 Differential Ground Displacements

Included in the projects relying on Network

Operations and Data Processing, above, is the

Engineering Data Analysis project. As described there, data has already been recorded

on both the differential arrays. This differential ground motion, were it to occur under

the base of an extended structure, would

impose direct strains on the structure in

addition to those strains from the earthquake

loading. The analysis of these records shows

that when applied to a simple structure under

specific circumstances, these direct strains

can be more important than those strains

calculated assuming the entire foundation is

excited in unison (Ref. 8).

5 . 3 System Identification in Structural

Dynamics

This project develops methods for system identification for use in analyzing the earthquake

recordings of instrumented structures. It

forms the bridge between the instrumented

structures in the strong-motion instrumentation network and the processed records

obtained during earthquakes. Preliminary

testing of these research methods incorporates

the use of the processed records to identify

mode shapes and frequencies, and the comparison of these identified characteristics with

values from other independent sources.

6. CONCLUSIONS

The research projects within the USGS that are directly involved with the permanent strongmotion instrumentation network of accelerographs and the processing of records form a distinct group in the earthquake hazard

reduction program. The basic projects

(network operation and data processing) are closely linked to the others in this group

(geophysics and engineering problem solving).

Although the responsibilities for all projects

are clearly delineated, it is clear that the

performance of the basic projects is important

to the others, while, on the other hand, and

in the long term, the basic projects rely

strongly on good scientific use of their data.

7. REFERENCES

4 April 1981 Westmorland and 16 January 1986

Hollister earthquakes: USGS Open-File report

(in preparation).

4. Holzer, T. L., Bennett, M. J., Youd, T. L.,

and Chen, A.T.F. , 1986, Identification of a site to monitor liquefaction during the

predicted 1988 Parkfield earthquake, Cholame

Valley, California (abstract): Association of

Engineering Geologists Abstracts with Program,

Annual Meeting, 29th, San Francisco, p. 51.

5. Celebi, M. , and Maley, R. P., 1986.

Strong-motion instrumentation of structures in

Charleston, S. Carolina, and elsewhere:

Proceedings, U.S. National Conference on Earthquake Engineering, 3rd, Charleston, S. Carolina, II, p. 1273-1283.

6. Saragoni, R. , Fresard, M. , Brady, A. G.,

Celebi, M. (principle investigator), and Mork,

P. N., 1987. Processed Chile earthquake

records of 3 March 1985 and aftershocks: USGS

Open-file report (in preparation).

7. Brady, A. G. and. Celebi, M. , 1986.

Fundamental modal behavior of an earthquakeexcited bridge: U.S. National Conference on Earthquake Engineering, 3rd, Charleston, S. Carolina, III, 2225-2236.

8. Bycroft, G. N„, and Mork, P. N., 1987.

Differential ground motions and their

spectra: in preparation.

1. Jacobsen, M. L., and Rodriguez, T. R.

1987. "National earthquake hazards reduction

program, summaries of technical reports,"

Volume XXIII, USGS Open-File Report, in press.

2. Converse, A. M. , 1987. User's guide to

ESM: a database for strong-motion

information: USGS Open-File Report 87-160.

3. Bycroft, G. N. and Mork, P. N., 1987.

Differential displacements and spectra for

36
Estimation of Source Time Function Based on Strong

Ground Motions and Its Application for Engineering

BY

Yoshikazu Kitagawa 3

, Takahito Inoue 2

Tatsuo Murota 3

ABSTRACT

Dynamic response analysis is applied to earthquake resistant design of structures. It

is possible to analyze the dynamic phenomena

of structures for the earthquake to the most

accurate degree, because the theories and

computer programs concerning these analyses

have made great progress recently. But the

typical observed strong ground motions have

been used mostly as "earthquake input". As

the number or the nature of those waves are limited as yet, we must take care to use their

waveforms for design and a useful method for

synthesizing the ground motions has not yet

been developed.

In order to use the synthetic ground motions

calculated from the fault model as "earthquake

input", it is necessary to estimate the

characteristics of the source time function

correctly. In this paper, we propose the

estimation method of the source time function

with the observed ground motions by applying

the autoregressive(AR) model to the observed

ground motions. Further, we apply this

procedure to the synthetic waves calculated

from the fault model in order to verify this

method. As a results, the waveform obtained

by using this method corresponds to the source

time function of the seismic moment. These

results agree with the condition given for the

fault model analysis and it is pointed out

that this analytical method using the AR model

ia available to estimate "earthquake input"

for the dynamic response analysis of

structures

.

KEY WORDS: Source Time Function, Event,

Seismic Moment, Fault Model

1. INTRODUCTION

In order to estimate the seismic safety of

structures, the dynamic response analysis is

applied to earthquake resistant design of

structures, and the typical observed strong

ground motions have been used mostly as "earthquake input". But as the number or the

dynamic properties of those waves are limited

as yet, we must take care to use their

waveforms for design and a useful method of

synthesizing the ground motions has not yet

been developed. But recently many trials to have a realistic "earthquake input" have been

performed from both a theoretical and

observational aspect. They are pseudo-ground

motion, superposition of many seismic waves

from aftershocks, etc.. Especially in

connection with rather long period ground

motions, the method of synthesizing the ground

motion due to earthquake faulting is applied

for engineering. In order to use the

synthetic ground motions calculated from the

fault model as "earthquake input", it is

significant to grasp the characteristics of

the source time function correctly. However,

the ground motion is strongly affected by the

earthquake process itself, the propagating

path of the seismic wave, the topography of

the ground surface, and the properties of

ground surface layers.

In this paper, we propose the estimation

method of the source time function with the

observed ground motions by using the

autoregressive (AR) model for the purpose of

estimating the dynamic properties of the

ground surface and the seismic wave path.

Further, we apply this procedure to the

synthetic waves calculated from the fault

model in order to verify this method (1).

2. ANALYTICAL METHOD

The inversion method proposed by Kikuchi(2) is one method by which the distribution of events

(called strong motion sources) on the fault

plane can be estimated. In this method, the

parameters for the fault are determined by

matching the synthetic waveforms with the

observed ones. But, by using a method

proposed in this paper the source time

function is obtained directly from the

observed ground motion. Besides, Nair(3)

tries to estimate the source function and the

medium response function from the observed

ground motion by the AR model, Tomizawa et al. (4) synthesize three-dimensional earthquake

ground motion, and Yokota et al.(5) determine

the arrival times of P- and S-waves on the

1 Dr. of Eng., Head of Second Earthquake

Engineering Division, International Institute

of Seismology and Earthquake Engineering,

Building Research Institute, Ministry of

Construction, Japan.

2 Visiting Researcher of Second Earthquake

Engineering Division, International Institute

of Seismology and Earthquake Engineering,

Building Research Institute, Ministry of

Construction, Japan.

3 Director, Structure Department, Building

Research Institute, Ministry of Construction,

Japan.

37
prediction system of earthquakes. The outline

of this analytical method follows.

In order to estimate the dynamic properties of

the ground surface and the seismic wave path, we apply the AR model to the tail part of the

observed ground motion which is appeared by

the finite discrete time series

x (i=l , 2, . . .

.

, N) . As the tail part,

x_^ (i=L, L+l ,. . . . ,N ; 1<L<N) , is random data with

Few non-stationary properties, it is written

as

i-m + e

; i=L,L+l, . . . . ,N (1)

where a (m=l , 2 M) is the AR coefficient

and minimusing the average square error, M is

the AR order and e is error series. Then in

order to determine the suitability of a given

AR model, Akaike's final prediction error

(FPE)(6) is introduced.

By using the AR coefficient, a , obtained in

Eq.(l), the estimated value, x , from

observed ground motion is represented by the

following equation as the linear combination

with former values.

x = E a

6 1. nv § m

" i-m

i=M, M+l

( 2 )

This value shown in Eq. (2) includes the

dynamic properties of the observation site.

Further, the residual time series, e , defined

by Eq.(3) excludes the dynamic effects of the

ground surface and the seismic wave path and

includes the much data related to the source characteristics.

e . =x - x =x - £ a x , i i e i i » i m i-ra

; i=M,M+l, , . .

. ,N (3)

If the discrete time series, x^, shown in

Eq.(3) is a deterministic waveform, the error series becomes a white noise process. But as

the. observed ground motion is the

non-stationary random data, the residuals time

series shown in Eq.(3) doesn't accept the

non-stationary property sufficiently. The

non-stationary spike appears in the residuals

time series as the non-stationary property is

larger in the observed ground motion. The

time series, e,, which is excluded the

r i non-stationary property from the residuals

time series , e^, shown in Eq.(3), resembles

the stationary gauss ian process closely.

There is however the non-stationary property

of the amplitude which is appeared in Eq.(A)

e = e 4 r i i s ii=M, M+l N (A)

where e^ is the finite discrete time series

with tfie stationary gaussian process, 4> is

the weight function which gives 1 the

non-stationary property to the amplitude.

By integrating the waveform

g e^ obtained from

Eq.(A) shown in Eq.(5), it becomes possible to estimate the source time function as the

distribution of events in the fault plane.

M =a / e, dt

i •'si

; i=M,M+l , . . .

. ,N (5)

where a is the coefficient including the soil

constants at the fault.

Further, when the observed ground motions

include the multiple shock sequence, the onset

time and the distribution of events in the

fault plane are decided from the interval time

between events(7).

3. NUMERICAL ANALYSIS

3 . 1 Analytical model and its characteristics

In order to verify this analytical method,

this procedure is applied to the synthetic waves calculated from the fault model. The

synthetic waveforms are calculated by using

the estimation method developed by Sudo,

Kitagawa and Yosimura(8). The used

underground structure and fault parameters are shown in Fig.l and Table 1. The relation

between the fault and the observation

point (OP-1) is shown in Fig. 10 and the

epicentral distance is 113 km. The transfer

functions of the underground structure

obtained by Haskell's method are shown in

Fig„2. One is a transfer function between the

point source and the ground surface, the other

is between the baserock and the ground

surface. Both of the transfer functions show

a similar property and their predominant

frequency is from 2 to 5 Hz. The waveforms of

velocity calculated from the fault model with

one event, which have the seismic moment shown

in Fig. 7, are shown in Fig. 3 and their fourier

spectrum is shown in Fig. A. Besides, this

numerical analysis is concerned with the body wave mainly.

In order to estimate the dynamic property of

the observation site, it is necessary to select the appropriate AR order and time

length of the tall part of the observed ground

motion respectively. The reference range of

the most appropriate AR order proposed by

Akaike(6) is 2-3/k( k is number of data). We

investigate two cases of the AR order both

with the first local minimum of FPE and the

lowest value of FPE in that range. Coherence

between the observed waveform at point (C) and

the waveform applied by the AR model with two cases of the AR order are shown in Fig. 5. Regardless of the AR order, the observed wave below 2Hz is regressed by the AR model but the

agreement above 2Hz is not so good. Further,

coherence between the waveforms obtained from

residuals in Eq.(3) and the incident waves at point (B) are shown in Fig.6(a)(b). The

38
waveform obtained from the residuals with the

first local minimum of FPE of the AR order

corresponds with the incident wave at point (5). On the other hand, the waveform obtained from

the residuals with the lowest value of FPE of

the AR order agrees with the incident wave at point @ and the apparent arrival time P and S waves can be measured at the observation

point. Judging from Fig. 6, as the order of

the AR model is higher, the waveform including

more source characteristics is obtained from

the observed ground motions.

3 . 2 Property of distribution of events

The waveform obtained from Eq.(5) is shown in

Fig. 7 as a solid line. When we think that the

waveform with the arrival time of S-wave

corresponds to the source time function, this

waveform shows the earthquake as a single

event. As the area of the waveform resembles

the seismic moment, the source time function

in this figure has 0.66 * 10 26 dynecm as the

seismic moment. This result agrees with the

given condition for analysis shown in Fig. 7 as

a dotted line. Further, the results of this

procedure as applied to the synthetic body and

surface waves calculated from the fault model

with one event is shown in Fig. 8. Its waveform

is similar to the waveform shown in Fig. 7 and

the agreement between the seismic moment

obtained in this case and given for analysis

is better than in the body wave only.

The results in the case of multiple events are shown in Fig. 9. There are two peaks in this source time function. The second peak is

smaller than the first peak and the seismic

moment of the second event is underestimated

more than the given condition for analysis

but the total seismic moment is 0.68 * 10

dynecm. The interval time between two events

is longer than the given condition for

analysis because of the Doppler effect. The

location and the interval time between two

events obtained by using the results at three

observation points are shown in Fig. 10. These

results agree with the condition given for the

fault model analysis shown in Table 1.

4. CONCLUSION

In order to obtain the strong ground motion

estimation due to realistic earthquake

faulting, it is necessary to clarify the source characteristics. In this paper, we propose the estimation method of the source

time function with the observed ground

motions. For the purpose of excluding the

dynamic properties of the ground surface and

the seismic wave path, we apply the AR model

to the tail part of the observed ground

motions. Further we apply this procedure to the synthetic waves calculated from the fault

model in order to verify this method. As a result, the waveform obtained by using this

method corresponds to the source time function

of the seismic moment and these results agree

with the condition given for the fault model

analysis. It is pointed out that this

analytical method using the AR model is

available to estimate "earthquake input" for

the dynamic response analysis of structures.

From now, we plan to investigate in detail by

applying this method to the observed ground

motion with more complex source condition.

ACKNOWLEDGEMENT

We acknowledge Mr. Toshihide KASHIMA and Miss

Yukie TAKASHIMA of Second Earthquake

Engineering Division, International Institute

of Seismology and Earthquake Engineering,

Building Research Institute, Ministry of

Construction, Japan kindly much help.

(1) Y. Kitagawa, T.Inoue : Study on Source Time

Function of Strong Ground Motion, Proc.7th

J.E.E.S, pp. 157-162, 1986, (in Japanese).

(2) M.Kikuchi : Inversion of Complex Body

Waves, B.S.S.A, 72, 491, 1982.

(3) G.J.Nair : Estimation of Source Function

and Medium Response Function by Autoregressive

Method, Phys. Earth Planet. Inter., 32, pp. 36-44, 1983.

(4) M.Tomizawa, T.Matsumura, T.Demizu : Synthesis of Three-Dimensional Earthquake

Ground Motion by means of Time-Varying

Autoregressive Model, Journal of Structural

and Construction Engineering, No. 349,

pp. 10-21, 1985, (in Japanese).

(5) T.Yokota, S.Zhou, M.Mizoue : An Automatic

Measurement of Arraival Times of Seismic Waves

and Its Application to an On-Line Processing

System, Bull. Earthq. Res. Inst., Vol.56,

pp. 449-484, 1981, (in Japanese).

(6) H.Akaike : Fitting Autoregressive Models

for Prediction, Ann. Inst. Statist. Math.

,

Vol.21, pp. 243-247, 1969.

(7) H.Sato, T.Sato, N.Yamada : Rupture Process

of the May 26 1983 Japan Sea Earthquake as Inferred from Strong Motion Accelerograms.

Programme Abstr. Seismol. Soc. Japan, No. 2,

pp.18, 1983 (in Japanese).

(8) K.Sudo, Y. Kitagawa, N.Yoshimura : Strong

Motion Estimation due to Realistic Earthquake

Faulting, 16th UJNR, 1984.

39
40
30N3H3H00

41

fault

length

(L)

20

(km)

Fig.

5

Coherence

between

observed

waves

fault

width

(W)

10

(km)

and

wave

form

applied

by

AR

model

rupture

velocity

3.0

(km/sec)
wave

form

obtained

from

0)3

fQ>

03

(D5O

jj •H3

1-1 01 13

l-i Oa<

M-jO

Ha

a;a

w

ro iw u o

SI

3DN3H3HQD

touo

03a

-H

ai

JJ *H

1-1 0) 13 &3

i-i

a o a a 4-i < o

U-l c 0 3E

0) -a

03 c

(0 -H U E

ra E 13 a c 0 «o

4-1 03

01 i—

> n3

ro 3 2 13n H C 03 0 01

cj

a 3 03 a E 01 0 O >

-Q

a (0 *a 3

0)U 3 iJ 3 01 C

01 3 01 a -a 13

0) (0 -a o ii o 0 JQ 3 U O -a

40

CT> *Ha

3DN333H00

42
observed

wave o Cl r.3N[M) -13A oI oH O

QJ T3 > 0 0) Si

0) O' c c 0

•H

J-l r*

0) JJ

•H 5

Cfi c cn 0 0J

(-) UJ 0 >

CD 03 — S-l 5 UJ2 0) 4J 0)

VJ-I 0

nj 03 U4

U3

u

4J 3

f-H cn 3w 'O

0) CX 03

03

o/W

O'

•H

&4 I e.C<

(0>U 4J c 0) c

0)

W -H > .C 0

0) 0 a* o < E to

1*3 -J —« < H •>

U

1*3

u

Hn < 2 > (—i

OS

1*3

i-. 1*3 2 U —«

cn 2 /—s

L* < E 2 £— -al 2 1*3 in w < > —

1*3

Q

1*3 'J 2 2 1*3

H < 1*3 2

1*3 c cn r-* > O — 1*3 1*3 2 2 ca

r~lI 5<

TIME

(5EC )

Fig.

10

Location

of

events

and

Fig. 9

Results

in

case

of

two

events

observation

points
-
Strong-Motion Instrumentation Program

Corps of Engineers

by

F. K. Chang, R. F. Ballard, Jr., and A. G. Franklin 1

ABSTRACT

The Corps of Engineers Strong-Motion Instrumentation Program was established in 1970 for the

purposes of surveillance of seismic safety of

Corps structures, provision of performance data,

and provision of a data base for earthquake

research. The program currently comprises 400

accelerographs , 14 seismic alarm devices, 56

seismoscopes , and 34 peak recording accelerographs at 114 Corps projects in 34 states and

Puerto Rico. Responsibility for the program is

assigned to the Waterways Experiment Station

(WES) ; installation and maintenance operations

are shared by the WES and the U.S. Geological

Survey. Significant earthquake records have

been obtained in the New Hampshire earthquake of

18 January 1982 and the Borah Peak, Idaho, earthquake of 28 October 1983.

KEYWORDS: Earthquakes; Dams; Instrumentation

1 . HISTORICAL BACKGROUND

The Corps of Engineers (CE) Strong-Motion Instrumentation Program (SMIP) was established in 1970

with the primary purpose of surveillance of seismic safety of existing and future Corps structures, such as earth and roclc-fill dams, concrete

dams, and appurtenant structures. It was also

intended to provide a measure of project performance, to provide design and performance comparisons, and to develop a data base for performance

predictions and earthquake research. Responsibility for operation of the program was assigned

to the U.S. Army Engineer Waterways Experiment

Station (WES). Funding was, and continues to be,

provided by the individual Corps of Engineer districts on whose projects the instruments are located

.

In 1973, an agreement was made with the U. S.

Geological Survey (USGS), under which the

Seismic Engineering Branch of the USGS undertook the installation, and maintenance of all

Corps of Engineers (CE) strong-motion instruments. By 1978, the operations were overtaxing

the resources of the USGS, and the arrangement

was modified. The WES assumed responsibility for

installation and maintenance of instruments in

the eastern United States, which represents about

half of the Corps’ instruments. The USGS continues to maintain those instruments in the

western United States.

2. GENERAL CRITERIA FOR DESIGN OF INSTRUMENT

INSTALLATIONS

Selection of Dams to be Instrumented

General guidance on strong-motion instrument installations is given in the Corps of Engineers’

Engineer Manual (EM) 1110-2-1908 (Department of

the Army, 1976). In general, all dams in seismic

risk zones 3 and 4 of the Seismic Risk Maps

(Fig. 1) are required to be instrumented. In

other zones, selection of dams to be instrumented

takes into account the height, material type, and

zoning of the embankment, the type of foundation,

whether the embankment is founded on overburden,

and the depth to rock. Also, the need for collection of basic data is a consideration. Where

there is a likelihood of occurrence of earthquake

shocks of sufficient intensity to give useful

information, instrumentation is warranted. In

questionable cases, the following additional

rules are applied:

o Nature of Foundation - If the foundation

materials underlying the dam are composed of

sands or silty sands that might be subject to

liquefaction, the dam should be instrumented.

o Type of Construction - Regardless of seismic

risk zone, all hydraulic-fill dams should be

instrumented. Rolled earth-fill or earthand rock-fill dams, being less susceptible to

liquefaction, are considered for instrumentation on the basis of other influencing

factors.

o Height - Most dams that are more than 100 ft

(30 m) high should be instrumented.

o Presence of Known Capable Faults - If the

dam is located nearer than 25 miles (40 km)

to a known capable fault, it should be

instrumented.

o History of Seismic Activity at the Site - If

acceleration levels greater than 0.2 g have

been recorded in the vicinity of the dam, it

should be instrumented.

o Distance from Higher Risk Zone Boundaries - If the dam is located less than 100 miles

(loOkm) from a higher risk zone boundary, it

should be instrumented.

U.S. Army Engineer Waterways Experiment

Station, Vicksburg, Mississippi

44
Design of Instrument Arrays

A typical instrument array includes threecomponent accelerographs at three locations: a free-field location, on the surface of the alluvium a short distance downstream of the dam; on

the dam crest, near the center or at the highest

section; and on an abutment. On a high dam, additional installations may be made at intermediate locations on the downstream face. Special

site conditions may warrant additional free-field

or abutment locations. On the other hand, some dams in areas of low seismic risk may have instruments at only one or two locations.

3. INSTRUMENTS USED

Instruments for measuring strong ground motions

must be self-contained units. They must be

equipped with self-starting devices that will be

activated at a prescribed level of seismic activity, and then they must measure and record

three components of motion for several minutes.

Strong-motion instruments originally approved for use in the Corps' SMTP network included the

RFT-250 and RFT-350 Strong-Motion Accelerographs,

the SMA-1 Strong -Motion Accelerograph , the

PRA-100 Peak Recording Accelerograph, and the

Wilmot SR- 100 Seismoscope. Manufacture of the

RFT Accelerographs has since been discontinued,

and those in the network will be replaced by

SMA-1 Accelerographs

Limited demand and controversy over their usefulness have prompted the manufacturers to cease production of the Wilmot Seismoscope. There are

56 seismoscopes still in operation in the network; they continue to be maintained but will

eventually be phased out.

The RFT and SMA-1 accelerographs are analog type

recorders. Since 1985, the Corps has installed

eight DCA-333 digital accelerographs at R. D. Bailey (4 instruments), J. W. Flanagan (3) and

Bluestone (1) dams of the Huntington District,

Ohio River Division, for the purpose of evaluating their long-term reliability.

As a measure to minimize the need for on-site

inspections after earthquakes, seismic alarm

devices have been installed on several Corps

structures. The seismic alarm device provides a warning when a preset threshold acceleration has

been exceeded. The alarm package contains three

vertical accelerometers with individual threshold

levels commonly set at 0.05 g, 0.10 g, and 0.20 g.

After an event, site personnel can readily

determine if any of the three threshold levels were reached and how many times the level was exceeded. The alarm devices were designed and

fabricated by WES.

4. PRESENT STATUS OF PROGRAM

As of 31 March 1937, the Corps of Engineers had

in operation 400 accelerographs, 14 seismic alarm

devices, 56 seismoscopes, and 34 peak recording

accelerographs (PRA'S), at 119 projects located

in 34 states and the Commonwealth of Puerto Rico.

A map of instrumented projects is shown in Figure 2, and a tabulation of strong -motion instrumentation is given in Table 1

5.

SIGNIFICANT EARTHQUAKES RECORDED

New Hampshire Earthquake of 18 January 1982

The first instrumental earthquake records representative of seismic activity in the New England area were obtained as a result of the 18 January

1982 Franklin Falls, NH, 4.7 magnitude event,

which occurred at latitude 43.5 degrees N, longitude 71.6 degrees W. Thirty-six accelerograms

(components) were recorded, digitized, and analyzed (Chang, 1983). The transverse component of

the accelerograph record located on the right

abutment of Franklin Falls Dam (8 km from the

epicenter) recorded a maximum acceleration of

0.

55.g, which is the highest value of acceleration ever recorded in the eastern United States.

The integrated maximum velocity is 5.59 cm/sec,

which is very low in comparison with other earthquake records in the western United States or other earthquake regions.

Mount Borah, Idaho, Earthquake of 28 October 1983

During the M_ 7.3 Mt. Borah earthquake of

28 October 1983, which occurred at latitude

44.0 degrees N, longitude 113.9 degrees W, in

central Idaho, nine meaningful strong -motion

records (27 components) were recovered on three

dams. These were Dworshak Dam (concrete, 330 km

from epicenter, 1 record), Lucky Peak Dam (earth,

180 km, 3 records) and Ririe Dam (rock-fill,

179 km, 5 records). Minor damage to the intake

structure at Ririe Dam was reported. An analysis

of the records was reported by Chang (1985).

ACKNOWLEDGMENT

The tests described and the resulting data presented, unless otherwise noted, were obtained

from research conducted under the Strong -Motion

Instrumentation Program of the U. S. Army Corps

of Engineers by the USAE Waterways Experiment

Station. Permission was granted by the Chief of

Engineers to publish this information.

REFERENCES

1. Chang, F. K. , 1983. "Analysis of Strong -Motion

Data from the New Hampshire Earthquake of

18 January 1982," NUREG/CR-3327 ,- U. S. Nuclear

Regulatory Commission, Washington, D. C. 20555

2. Chang, F. K. , 1985. "Analysis of Strong-Motion

Data from the Mount Borah, Idaho, Earthquake of

28 October 1983." Miscellaneous Paper GL-85-12,

U.S. Army Engineer Waterways Experiment Station,

Vicxsburg, Mississippi 39180

3. Department of the Army, 1976. Engineer Manual

1110-2-1908, Instrumentation of Earth and Rock

Fill Dams , Part 2 of 2, 19 Nov 1976. Office of

the Chief of Engineers, Washington, D.C. 20314

45
46

Figure

1.

Seismic

zone

map

of

the

contiguous

United

States

and

Puerto

Rico
CORPS

OF

ENGINEERS

STRONG-MOTION

INSTRUMENTATION

PROGRAM

STATUS -

MAY

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47
CORPS

OF

ENGINEERS

STRONG-MOTION

INSTRUMENTATION

PROGRAM

STATUS

-

31

MARCH

1987

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48

ioga-Hammond

earth

&

roc
Seismic Alarm

Serviced

Coordinates

Station

H

Accelerographs

Selsmoscopes

Peak

Accel.

Devices

By

Lat.

Long

No.

Division

District

Project

Type

(m)

State

Operational

Planned

Oper.

Plan.

Oper

Plan.

Oper

.

Plan.

USGS

WES

N

W

X X X X X

<> QC (/} <> !/l M 1/1 (/)

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ld h n CO v© *y o o n n ro ^ n

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49

Howard

Hanson

rock-fill

72

WA

3

SMA-i

Libby

gravity

136

MT

4

RFT-250

Mud

Mountain

earth

130

WA

3

SMA-l

Uynoochee

concrete

55

WA

3

SMA-l
Seismic Alarm

Serviced

Coordinates

Station

H

Accelerographs

Selsmoscopes

Peak

Accel.

Devices

By

Lat.

Long.

No.

Division

District

Project

Type

(m)

State

Operational

Planned

Oper

.

Plan.

Oper.

Plan.

Oper.

Plan.

USGS

WES

N

W

< < I z z X 1/1 1/1 i/l CO

ld

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Q —I -J

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h n ^

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Kinzua

earth

&

gravity

55

PA

3

SMA-1

Michael

J.

Klrwan

earth

25

OH

1

SMA-1

Tionesta

earth

47

PA

1

SMA-1
TABLE

1

(Continued)

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51
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of5)

Comparisons of Dynamic Shear Strength

Between Fine Sands and Gravels

By

Norihisa MATSUMOTO 1

, Nario YASUDA2 Masahiko OHKUBO3

, Shoji ARAKAWA4

ABSTRACT

In this paper dynamic shear strength or liquefaction

strength of rockfill materials has been investigated by

using the large-scale cyclic triaxial testing equipment.

The size of saturated columnar specimen is 30-cm in diameter and 60-cm long. The gram size ranges from

0.42mm to 63.5mm.

Tests for twelve specimens of rockfill materials were performed by using the above mentioned equipment in order to study the effects of gradation and density of

the materials on the dynamic shear strength.

The followings results are obtained on the dynamic

strength of the coarse grained granular materials.

1) The specimen of higher relative density indicates the higher strength against liquefaction in the

case of the same grain size distribution.

2) The well graded materials can resist to the cyclic

shear stress better than the materials which have

smaller uniformity coefficient and the same

relative density.

3) The coarse grained materials are stronger against

liqueifaction than fine sands. KEYWARDS: Dynamic Strength, Coarse Grained

Granular Materials, Undrained Shear Test

1.

INTRODUCTION

A lot of research works have been accumulated on

the liquefaction or strength loss of sands during earth- quakes. However, available test data of dynamic shear

strength for coarse grained granular materials are very

limited because the particle size contained in the materials is large and hence the large-scale testing equipment is required.

The authors have developed the large-scale cyclic

triaxial testing equipment at Public Works Research

Institute of Ministry of Construction to obtain dynamic

deformation properties and dynamic shear strength for

rockfill materials.

This paper, therefore, describes the test results of

effects of grain size distribution and relative density on

the dynamic shear strength of saturated rock-fill materials under undrained test condition.

2.

GENERAU DESCRIPTION OF TEST

2.1 Uarge-scale Triaxial Test Equipment

The equipment used for this study is capable of test- ing 30-cm diameter, 60-cm long specimens. The main

features of performances are shown in Table 1. Fig. 1 illustrates the equipment. Friction between loading rod

and bearing is minimized by using a ball bearing. The

cyclic axial loads are applied by using an electro-hydraulic

servo valve. Hydraulic pressure ranges from 120kg f/cirr to 180kg f/cm 2

. The maximum cyclic frequency is 10Hz

for an axial load and 5Hz for an confining pressure. The two separated sensors shown in Fig. 1, each of

which is disk-shaped coil, are used as a displacement

transducer for measuring the small range of displacement

for the dynamic deformation test. However, the differ- ential-type displacement transducer attached to the outside of a triaxial cell is used in these liquefaction tests.

2.2 Preparation of Specimens and Test Condition

The cyclic triaxial tests (liquefaction test) in this study were performed on rock-fill materials for Sagae

Dam which is now under construction in northern part

of Japan by the Ministry of Construction. The quality of

rock materials consists of mainly dolerite containing

pegmatite partially.

Some physical properties of rock materials are shown

in Table 2. The grain size distributions of the specimens

are illustrated in Fig. 2. Two different gradation curves

are adopted and the uniformity coefficient is 12 for A-sample and 7 for B-sample. The maximum particle size of the specimen is 63.5mm. which is less than 1/5 of

the diameter of the specimen. This ratio is borrowed

from a maximum allowable ratio for the static triaxial

tests.

1) Dr. of Eng., Head, Filldams Section. Dam Department

Public Works Research Institute. Ministry of Construction.

Japan

2) Research Engineer, Filldams Section. Dam Department

Public Works Research Institute. Japan

3) Engineer. Filldams Section, Dam Department Public Works

Research Institute, Japan

4) Research Engineer, Filldams Section. Dam Department

Public Works Research Institute. Japan

53
The materials were placed in six layers, and each layer was compacted with the 33kg f weight electric vibratory rammer which has 950 compaction blows per minutes.

The test density was determined from the compaction

tests so that relative densities of the materials should be

60, 80, 97% in each gradation.

The procedures of cyclic triaxial tests are as follows

1) The confining pressure of 0.2kg f/cm2

is applied

to the specimen (30-cm in diamater and 60-cm

high) and then the vacuum pressure inside the

specimen is released. 2) Carbon dioxide (C02 ) of 0.05 kg f/cm 2 pressure

is supplied to the specimen for 24 hours in order

to fill the air void completely with C02 because

C02

is much easier to dissolve in water than air. 3) After the completion of replacement of air by

C02 , de-aired water of 70 liters is supplied to the

sample for 24 hours.

4) Saturation degree of the specimen is investigated

by Skempton’s B-value and test is carried out about the specimen which B-value is more than

0.96.

5) The burette of 1,000 cc is used for measuring

volume change of the sample during the consolidation of about 45 minutes.

The cyclic undrained test is carried out under loadcontrolled condition of 0.2Hz. The specimen is subjected

to the back pressure of 2.0kg f/cm2 and the confining

pressure of 4.0kg f/cm 2

, resulting in the effective

confining pressure of 2.0 kg f/cm 2. Shear stress ratio R1 is expressed as the followings

o c’ 2 o c' where r ; cyclic shear stress a d ; cyclic deviatoric stress (single amplitude) o c

' ; effective confining pressure The specimen of Toyoura standard sand also prepared

after the same manner of coarse grained granural

materials.

The digitized test results are recorded by a personal

computer through the low-pass filter of 1.0Hz by which

the noise is deleted from the measured data. The data

aquisition system of the experiment is indicated in Fig.3.

3. TEST RESULTS

Shear stress ratio of cyclic tests ranges from 0.3 to

0.7. This range of values is .determined from the dynamic

non-linear analysis of 100m high rockfill dam F.E.M.

model. Two types of earthquake motions are chosen as an input motion in which maximum acceleration is 250

gals and 350 gals each. The shear stress ratio of whole

elements ranges from 0.01 to 1 0 for the former motion

and from 0.1 to 0.7 for latter one. The shear stress ratio has a tendency to increase near the slope surface. Tested results are listed in Table 3. Falling head per- meability tests are performed after the saturation of

specimens and the permeability coefficients of

2~6x 10

“ 3 cm/sec are observed. Distinct relations were

not observed between the permeability and the gradation

or the relative density of a specimen.

Typical time histories of load, axial strain and pore

water pressure are shown in Fig. 4. Initial relative density of a sample is 97% and shear stress ratio is 0.35

Cyclic frequency of 0.2Hz is determined from the

capacity of cyclic triaxial equipment. Fig. 5 is a stress path by axial stress and pore water pressure.

Fig. 6 shows initial liquefaction potential defined as the number when the effective confining pressure becomes 0kg f/cm 2 for the first time The higher the

relative density becomes, the larger the number of

liquefaction is for A-samples. From this diagram coarse grained granular materials also indicate the liquefaction curves similar to the sands but are much stronger than

sands against the same shear stress ratio. The results of B-samples are also plotted in this figure. B-samples can be liquefied in early stage compared to A-samples for the same relative density at two

different levels of about 70% and 90% (see Fig 7). Disregarding the rubber membrane penetration to the void of a specimen can lead to the under-estimation of

liquefaction potentials. Therefore in the case of the sample of low uniformity coefficient like B-sample, the

tested result shows weaker liquefaction strength than

real one. Crush of the particle is another reason for low

strength of B-sample. Rockfill materials used in this study is easily crushed so B-sample, which uniformity

coefficient is smaller than that of A-sample, caused

larger volume change by cyclic loads. The test results of Toyoura standard sand are plotted

together in Fig. 7. The relative density ranges from 50%

to 80%. It is noticed that coarse grained materials is stronger against liqueifaction than fine sands . The process of pore water pressure build-up in the

specimen is illustrated in Fig. 8 The increase of pore water pressure occurs rather in early stage of liquefaction and gradually approach to zero effective confining pressure for both A-sample and B-sample. Compared

to sands, which line is also shown in Fig. 8, the pre pressure build-up of coarse grained granular materials occurs at the early stage of cycles.

The authors plan to investigate the effect of the

quality and grain size distributions of rockfill materials on liquefaction potential. And further studies are also required for the dissipation of excess pore pressures to evaluate the safety of rockfill dams subjected to earth- quake shaking.

4. CONCLUSIONS

From the previous considerations thae followings can

be concluded on the dynamic strength of the coarse grained granular materials.

1) The specimen of higher relative density indicates

the higher strength against liquefaction in the

case of the same grain size distribution.

54
2) The well graded materials can resist to the cyclic

shear stress better than the materials which have

smaller uniformity coefficient and the same

relative density.

3) The coarse grained materials are stronger against

liqueifaction than fine sands.

Note: This report is based on “N. MATSUMOTO. et. al. -DYNAMIC SHEAR STRENGTH OF COARSE

GRAINED GRANULAR MATERIALS, INTERNATIONAL SYMPOSIUM ON EARTHQUAKE

& DAMS, 20, MAY, 1987; BEIJING, CHINA”.

REFERENCE

1) N. Matsumoto, N. Yasuda, M. Ohkubo — Dynamic

Shear Modulus, Damping Ratio, and Poisson’s Ratio

of Coarse Grained Granular Materials, Vol. 28, No.

1 1 , 1986, Civil Engineering Journal, Japan

9

55
TABLE 1 Main features of performances

Specimen size <i>30cm x h60cm

Control system Electro- Hydraulic- Servo Wave form Sinusoidal, triangular, and earthquake waves

Axial loading

Maximum loading capacity compression - 20 tf Maximum loading capacity extension — 10 tf Frequency range 0.2 - 10 Hz

Confining pressure Max. 15 kgf/cm J Frequency range 0.2 - 5 Hz

Fig, 1 Cyclic triaxial testing equipment

1. Loading rod 6. Support

2. Gap sensor 7. Triaxial Cell

3. Lord cell (Extension & Compression) 8. Loading cap

4 Rubber membrane 9. Specimen

5. Porus metal disk 10 Pedestal

56
Table 2 Physical Properties of Materials

Gravel Sand

Spefific gravity Gs 2.509 2.644 Water content W 2.44 (%) Air-dry - Maximum void ratio e max 0.563 0.905 Minimum void ratio emin 0.182 0.595 Maximum grain size dmax 63.5 mm 0.42

Coefficient of uniformity Vc 12 (7) 1.36 Dry density of specimen rd 2.024 gf/cm 3 1.538

®

Fig. 2 Grain size distribution curve

1. Percent passing (%)

2. Grain size (mm)

3. A-sample (No. 1 - No. 9)

4. B-sample (No. 10 — No. 12)

5. Standard sand

57
Table 3 Tested Cases

No. of

sample T/Oc

' Dro (%) ec

e max e min

(1) (xlO"

3

)

1 0.32 60 0.338 0.552 0.248 6.04

2 0.48 60 0.340 ii II 3.10

3 0.65 60 0.345 n II 4.25 4 0.35 80 0.289 n II 4.82

5 0.51 80 0.292 it II 2.46

6 0.67 80 0.288 n II 5.86

7 0.35 97 0.243 n II 3.52

8 0.52 97 0.247 n II 5.54

9 0.68 97 0.245 n II 2.20

10 0.34 60 0.415 0.669 0.326 5.57

11 0.35 80 0.361 II It ‘ 3.35 12 0.35 97 0.318 II It 2.00

13 0.26 50 0.739 0.905 0.595 - 14 0.23 60 0.706 II II -

15 0.23 80 0.667 II II -

16 0.30 80 0.676

1

II

Fig. 3 Data acquisition system

1. Displacement transducer (Cap Sensor)

2. Load cell

3. Pore water pressure meter

4 Amplifier

5. Low-pass filter (1,0 Hz)

6. A/D converter

(16K, words of memory storage for each channel)

7. Personal computer (16 bit)

8. CRT

9 X-Y plotter

10. Floppy disk

58
2000

©

-2000

CD

t

1 *t ' i i r " r ~ r r r r » i r1 r ~t r t i i r i i ? t r —

r

t —r~ t1

i—1—J_i—

i

—J 1 1 1 A _i 1 i 1 T T' ~ T ! I'7-I T

"

T 1 1 T T T

-

' 1 T T ~r - T ~ T— T

1 © c

1 i- 1— 1 i 1 A 1 1 1 I—I— I lilt L 1 I 1 I i-- 1. A - i 1 L— J— U 1—

t —t — t i i r i r r t r \ i l > t

1 i t » i r i 1 t t t t

1 r~ ^vwwvywvmVYVYVVYYV^

100 ©

200 300

Fig. 4 Time histories of axial load, axial strain and pore water pressure

1 . Axial load ikgf)

2. Axial strain (7c)

3. Pore water pressure (kgf/cm 3

) 4. Time in second

0.75

©

-0.75-

-1.5

©

Fig. 5 Stress path of A-sample

1. Shear stress (kgt/cm 2

) 2. Effective mean principal stress (kgf/cm 2

) 3. Initial stress condition

59
©

Fig. 6 Liquefaction potentiaJ of specimens

1. Shear stress ratio

2. Number of liquefaction

3 A-Sample (Dro = 60%)

4. A-Sample (Dro = 80%)

5. A-Sample (Dro = 97%)

6. B-Sample (Dro = 60% ~97%)

7. Toyoura standard sand (Dro = 50% ~ 80%)

Fig. 7 Comparison of A-Sample and B-Sample on the same relative density

1. Relative density Dr (%)

2. Number of cycles Nl

60

3. A-Sample (No. 1. No. 4, No. 7) 4 B-Sample (No. 10 ~ No. 12)
©

Fig. 8 N/NC ~ Au/oc

1. Au/ac

' Pore water pressure 4. B-Sample

2. N/NC NC; Number of liquefaction

3. A-Sample

61
_

'

,

.

.

'

.
Preliminary Study of Shear Design for RC Piers

by

Hiroshi Shinohara 1

, Michio Okahara2

, Shouichi Nakatani3

, Keiji Taguchi4

, and Atsushi Funakoshi5

ABSTRACT

Earthquakes occur frequently in Japan, and thus

the safety for reinforced concrete (RC) columns as well as bridge piers shall be reviewed against bending

moments and also against shear force due to seismic

load. It is particularly important for RC piers to be

designed in such a manner that shear resistance of

RC piers will exceed the flexural resistance for pre- venting brittle collapse. On the other hand, in the

design of RC members for the substructure of highway bridges in Japan, no ultimate strength of

members is being considered at present.

Therefore, in the present study, the up-to-date

state of design methods against shear force to RC

members was surveyed on the basis of domestic and

foreign design guides and criteria and this design was examined based on data of previous loading tests on RC columns. This report shows the results of survey

using the data of previous loading tests, for the design methods being used in various countries as well as the results of the review of both the equation

for the shear strength and the required quantity of

lateral ties for preventing brittle collapse.

Keywords: Reinforced concrete pier, shear, shear

strength, lateral ties and loading tests.

1. INTRODUCTION

In Japan where earthquakes occur very frequently,

aseismatic design is considered to be an important

problem for not only structures for public facilities but also for various other structures. In the column

members for highway bridge piers, axial forces or a bending moment prevails at ordinary times but large

horizontal force occurs during earthquakes. The

safety of bridge piers must be examined not only

against the bending moment but also against shear

forces due to the horizontal force that occurs during

earthquake. That is, it is known that the RC member

shows an excellent deformation performance in case of flexural failure without diagonal cracks but the RC member shows brittle collapse in case of shear

failure. For the purpose of preventing this shear

failure, it is vitally important for earthquake resistant design that the shear resistance of RC piers will exceed its flexural resistance. Nevertheless, the conventional allowable stress method is still being

applied for the design of RC members of the substructure for highway bridges. According to the design rule of this conventional method for shear,

it is attempted to limit the average shear stress borne

by concrete member to less than the allowable shear

stress. Therefore, in the strict sense of the word,

the behavior of members subjected to external forces

greater than the design loads is not taken into

account in present design practice.

On the other hand, there is currently a tendency

to adopt the so-called ultimate state design method

in the actual design in this country, in which the

safety of members is secured by evaluating the variance of load and resistance in ultimate state basing upon probability theory. This tendency is partly seen in the recent revision of the Standard

Concrete Specifications (Japan Sciety of Civil Engineers). Namely, the check of safety for failure when subjected to ultimate load is desired even in the design of RC members for substructure of

highway bridges.

Therefore, in the present study, the typical design

guides and criteria applied in various countries were

surveyed and present design methods for shear force were revealed in order to examme the design

methods against shear force to column members

of RC bridge piers. In addition, the data of previous

loading tests for column members were collected and

compiled in order to classify failure modes and to review the shear strength. Based on the results, the design method against shear forces acting upon RC

bridge piers will be proposed in this report.

2. DESIGN METHODS IN VARIOUS

COUNTRIES FOR SHEAR FORCES

ACTING ON RC MEMBERS

2. 1 General

Design methods for RC members subjected to shear force were studied and compiled by surveying

several domestic and foreign design guides, criteria and standards, mainly with respect to the evaluation

methods and the structural items related to shear

strength.

Design guides, criteria and standards surveyed are:

Standard Concrete Specifications, 1986 (Japan

Society of Civil Engineers)

Highway Bridge Specifications, 1980 (Japan

Road Association)

ACI Building Code 318-83 (U.S.A.)

AASHTO 1983 (U.S.A.)

BS5400 part 4 (British)

CPI 1 0 (British

)

CEB/FIP 1978 (Europe)

DIN 1045 (West Germany)

1. Director, Structure & Bridge Department. Public Works

Research Institute, Ministry of Construction, Japan.

2. Head, Foundation Engineering Division, Structure & Bridge

Department, Public Works Research Institute. Ministry of Construction, Japan

3. Research Engineer, Foundation Engineering Division. Structure & Bridge Department, Public Works Research Institute, Ministry of Construction, Japan

4,

5. Foundation Engineering Division, Structure & Bridge Depart

ment, Public Works Research Institute. Ministry of Construction, Japan.

62
2.2 Comparison of Design Methods in

Various Countries

2 . 2.1 Provisions for Shear Force

Provisions for evaluating methods and structural items relating to the shear strength are compiled in Table 2.1. According to this table, the shear force is resisted by concrete and shear reinforcement, and the

required quantity of lateral ties is calculated based on

so-called classical truss analogy in every country.

Moreover, most provisions have adopted certain forms

of the ultimate state design method in order to assure the safety for ultimate strength of members.

2 . 2.2 Comparison of Shear Strength Evaluation

Methods

The principal factors affecting shear strength of

concrete are the main reinforcement ratio and axial compressive stress and so on. But these factors are in- corporated in design equation slightly differently

depending on country.

Characteristic points of these factors are explained

below.

(1) Effect of main reinforcement ratio In most of standards it is considered that the shear strength normally increases as the main rein- forcement ratio becomes larger because of the dowelling effect of main reinforcement. BS equation has

adopted the correlation between concrete strength

and the dowelling effect of main reinforcement, and

is different from other standards in that the shear

strength borne by the concrete is not increased even

if the concrete strength becomes larger where the

quantity of main reinforcement is small. (2) Effect of axial compressive stress Most of the standards have incorporated only a relatively small effect of axial compressive stress. Especially ACI equation hardly permits an increase in shear strength within the range (5 to 15 kg/cm

2

) of

stress used for ordinary bridge piers. Also, DIN

equation has not incorporated it by the reason that

it is of negligible degree However, CEB equation

expects an increase in the shear strength by axial compressive stress. (3) Influence of effective depth

Some standards consider a decrease of shear

strength as scale effect. CEB equation shows no

decrease where “d” is over 60 cm.

(4) Effect of lateral ties The effect of lateral ties is mostly based on the same concept taken from the classical truss analogy

in every standard. With respect to the upper limit of

the shear strength, the equations of every standard

have incorporated provisions based on the crushing

of web, but the crushing of the web hardly becomes

problematic since the cross section of the actual

bridge piers is large.

2.3 Review

Since various factors will affect the shear strength

of concrete, the design methods for shear in every

standard are slightly different but the concepts of

the shear strength of RC members are almost the same. However, these provisions m each country are basically for checking the shear in beam members

and not for considering the column members subjected to shear force. Also, though it is known that the shear strength of concrete increases as the shear span

ratio decreases, the effect of shear span ratio is currently not considered in the equation for checking

shear. It is desirable to make corrections based on

the shear span ratio because shear failure of column

members occurs frequently in the region where the shear span ratio is small.

3. STUDY OF DESIGN METHODS FOR

SHEAR FORCE TO RC PIERS

3 . 1 Principle of Study

The principle of this study is: (1) To collect data of previous loading tests on RC

column members carried out by various organization

in Japan and to compile and classify P-5 relations

and failure modes.

(2) To identify the failure modes of RC column

members. To classify the failure modes into four

kinds of shear failure type, shear flexural failure type, flexural shear failure type and flexural failure type.

(3) For specimens indicated the failure of shear

failure type and shear flexural failure type among the

failure modes, to survey their region and strength

and to review the empirical equation for shear

strength.

(4) Her specimens in dicstinc tHe feiiure of thie flexural failure type and the flexural shear failure type among the failure modes, and to survey and

review the amount of lateral ties arranged.

3.2 Study Based on Experimental Data

3.2.1 Collection of Experimental Data

Data of loading tests on RC column members

were collected under the following conditions:

(1) Cross sectional shape: Only rectangular sections were considered and circular and hollow sections were excluded.

(2) Concrete: Only the normal concrete was considered excluding the specimens of special concrete such as lightweight concrete.

(3) Axial steel material: Only the steel reinforcing

bars were considered excluding reinforcement by steel

rigid frame, steel plates and PC.

(4) Shear reinforcement: Only lateral ties were considered and spirals and diagonal reinforcing bars were

excluded.

(5) Axial compressive stress: Axial compressive stress were considered for the bridge piers excluding speci- mens with the value over 100 kg/cm 2

. (6) Specimen model: A model as shown in Fig. 3.1 was employed.

The experimental data were collected from 301

specimens tested by the organizations shown in Table

3.1.

3.2.2 Summary of Experimental Data

(1) Relation between maximum mean shear stress and the shear span ratio. For the experimental data obtained from 301

specimens, the relationship between the maximum

63
mean shear stress rmax (the value obtained by

dividing the maximum load Pmax by column width b times the effective depth d) and the shear span ratio a/d (the value obtained by dividing the column

height a by effective depth d) were classified by the

tensile reinforcement ratio Pr of column (the value

obtained by dividing the cross sectional area As of

main tensile reinforcement by column width b times

the effective depth d), and the results are shown in Fig. 3.2. From this, it is known that the maximum

mean shear stress increases as the shear span ratio a/d becomes smaller and tensile reinforcement ratio Pr becomes larger.

(2) Failure modes

Failure of RC column members due to horizontal

forces can be roughly divided into the following

forms:

1 ) T orm of shear failure before the flexural yielding of the main tensile reinforcement.

2) Form of flexural failure without the occurrence of diagonal cracks after the flexural yielding of main tensile reinforcement.

3) Intermediate form ot failure between (1) and

(2) in which the failure occurs as a result of a decrease in strength caused from the diagonal

cracks after the flexural yielding of main tensile reinforcement.

In the case of bridge piers, it is said that sufficient

flexibility can be maintained if the ratio of ultimate

displacement to yield displacement is greater than 4. By taking account of the above, attention was paid to flexural yielding load Py and horizontal displacement

4-5 y on P-5 curve based on the schematic diagrams of

failure modes as shown in Fig. 3.3, and the failure modes

were classified into four kinds. Classification and

features of these failure modes are shown in Table 3.2. The experimental data were then classified by the

failure mode, and the various indexes were plotted on

graphs as shown in Fig. 3.4 to Fig. 3.6. According to these graphs, it is well-recognized that brittle collapse

of S type and SB failure type occur considerably in the

following regions:

(1) Region with a shear span ratio of a/d <3 (refer to

Fig. 3.4).

(2) Region with a relatively large tensile reinforcement

ratio Pr (refer to Fig. 3.5).

(3) Region with no or little lateral ties (refer to Fig.

3.6).

3.2.3 Review of the Equation for Shear Strength

Except for the flexural failure type (B type and BS

type) among the experimental data, 41 specimens indi- cated brittle shear failure (S type and SB type). Then,

the experimental data from 41 specimens were reviewed

by making use of the design equation for shear strength

used mostly area in Japan. These equation are shown in Table 3.3. No safety factor for design was taken into account. Ratio between the maximum mean shear unit stress real (=Vc/bd or VWbd) calculated by the design equation and the maximum mean shear unit stress rmax (=Pmax/bd) obtained from test data was reviewed and

compiled.

Among 41 specimens indicated the shear failure type,

calculations were made using Eq. (ii) for 28 specimens

having no lateral ties, and the results are shown in Fig. 3.7. In addition, calculations were made using Eq.

(i) for 41 specimens including those reinforced with

lateral ties, and the results are shown in Fig. 3.8. From

this, it is apparent that the equation employed for calculating shear strength has underestimated the shear

strength in the region where a/d is small. This seems to have occurred because the effect of a/d was eliminated

for simplicity and the members to be considered when

introducing the design equation though the validity was

verified using even the shear span ratio (a/d) as para- meter in the empirical equation used for deriving Eq.

(ii) .In the same manner, calculations were made using

Eq. (v) for 28 specimens not reinforced with lateral ties and Eq (iv) for 41 specimens including those reinforced

with lateral ties, and the results shown in Figs. 3.9 and

3.10. Eq. (v) considers the effect of a/d, but the effect was underestimated in the region of experimental data

of 1.0<a/d<3.0 while the effect was much evaluated

in the region of a/d<1.0.

As a result of the review stated above, the effect of

shear span ratio (a/d) was taken into account by a simple

method based on the design equation set forth in the Standard Concrete Specifications, and then the proposed

equation for shear strength has been shown in Eq. (3.1)

V«= (4.0 — a/d) • Vc + Vs (3.1)

where, Vc is obtained from Eq. (ii) and Vs from Eq. (iii). Now, comparing the calculated values and experimental

values of the maximum mean shear stress, the results as shown in Figs. 3.11 and 3.12 are obtained. As shown in Fig. 3.11, the ratio between experimental value and

calculated value was 1.011 in mean value and 24.8% in coefficient of variation.

3.2.4 Review of Reinforcing Effect by Lateral

Ties

By identifying the failure modes as shown in Fig. 3.5, a considerably important tendency can be recognized

because the failure modes can be classified by the tensile reinforcement ratio and shear span ratio into the region

where brittle collapse occurs mostly and the region

where failure with a high flexibility occurs. As a sresult, the reinforcing effect of lateral ties on the plane with

tensile reinforcement ratio Pr and shear span ratio a/d was qualitatively reviewed by checking the quantity of

lateral ties which causes flexural failure with flexibility. In Fig. 3.13, the B type and BS type of failure mode

are abstracted and plotted by lateral tie ratio Pw of

specimens (=Aw/b's, where Aw: cross sectional ratio of

lateral ties, b: width of column, and S: spacing of lateral

ties). From this, it is recognized that flexural failures with flexibility can be bright about even if the lateral

ties are not placed at all. This means that it is more

rational to arrange the lateral ties on the basis the rela- tion between the tensile reinforcement ratio and the shear span ratio.

3.3 Proposed Design Method for Shear Force

to RC Piers

3.3.1 Principle of Design

(1) The shear strength of the column member is to be

expressed by the sum of shear strength to be borne by

64
concrete and shear strength to be borne by lateral ties. (2) Equaton for shear strength to be bome by lateral

ties is to be based on truss analogy.

(3) Equation for the shear strength to be bome by the concrete is to be corrected by the shear span, ratio of

smaller than 3. (4) The minimum quantity of lateral ties is to be given

by the classification region of shear span ratio and

tensile reinforcement ratio. 3.3.2

Checking Position and Others

(1) Checking the shear for RC bridge piers should be

made for the maximum shear force at the base of bridge

pier.

(2) The shear span ratio should be given by the following formulas (Fig. 3.14):

Shear span: a=Mmjjc jSmax (3.2)

Shear span ratio: a/d (3.3)

(3) Tensile reinforcement ratio Pr should be given by

the following formula:

Tensile reinforcement ratio: Pt=As/bd (3.4)

where, As: Cross sectional area of tensile reinforcement

3.3.3

Equation for Design Shear Strength

The following equation should be used for calculating

the design shear strength:

reinforced concrete design method in each country have

also been clarified. (2) It has been verified that various guides and

standards in every country was applied for the same

design method between beam members and column

members such as RC piers.

(3) As a result of collecting and compiling the results of loading tests conducted on column members in the

past, it has been found that the failure modes can be

roughly classified based on the relation between tensile reinforcement ratio and shear span ratio. It has been

clarified that the so-called shear failures indicating brittle collapse occur only when the shear span ratio is less than

3 in the case of the load size and sectional shape for normally designed bridge piers.

(4) It has been clarified that the shear strength bome

by the concrete varies with the shear span ratio even if the sectional shape is the same, ana tends to increase

as the shear span ratio becomes smaller. As a con- sequence, the previous equation for calculating shear

strength have been corrected based on the experimental

data. (5) The equation for calculating shear strength and the

mmimum quantity of lateral ties have been proposed,

and should be used when checking the shear in RC

bridge piers by ultimate state design method.

(3-5)

where, Vc is given by Eq. (3.6).

In the case of a/d^ t>: Vc =0.9Pd 0p Prr\/ Ocd bd/yb

In the case of a/d<3:

Vc=0.9 (4-a/d) Mpfalfcdbdhb

/Jrf=yToo7d£l.S

Pp = ?/ lOOpr^ i .5

(3n = l+2Mo/Mu^2.0

(3.6)

acd -ock /yc

ack : Design compressive strength of concrete

y e : Material coefficient of concrete

y

: Coefficient of member

Vi is given by Eq. (3.7)

Vi=Aw asy d/1 . 1 5s (3.7)

Aw : Cross sectional area of lateral ties asy : Yield strength of lateral ties

s : Spacing of lateral ties

3.3.4

Minimum Quantity of Lateral Ties

Quantity of lateral ties should be greater than the

values shown in Fig 3.15.

4. CONCLUSIONS

(1) As a result of surveying various domestic and

foreign guides and standards, it has been clarified that many have adopted the ultimate state design method or

load coefficient design method. Also, the main trends of

65
Table

2.1

Provisions

for

the

design

against

shear

forces

in

various

countries

66
67
Table

3.1

Experimental

data

collected

68 A
Table

3.3

Shear

strength

calculating

equation

used

in

the

study

69
Cantilever beam model

Fig. 3.1

Simple beam model

Specimen model Fig. 3.3 Classification of failure modes by P-5 curve

(kg/c

c 480

03

01E

I 100 E

x

ro

32.0 - Pmax bd

(kg/cm ‘

ot

SB type BS type

B type

1 6 2.4 3 2

Shear span

4.0 4 8 5 6

ratio a/d

Fig. 3.2 Relationship between maximum mean shear

stress and shear span ratio (by tensile reinforcement ratio)

Fig. 3.4 Relationship between maximum mean shear

stress and shear span ratio (by failure mode)

70
Tensile

reinforcement

ratio

Pi

(%)

Shear span ratio a/d

Mg. 3.5 Relationship between tensile reinforcement

ratio and shear span ratio (by failure mode)

Fig. 3.7 Comparison between experirr.en tal values anU

calculated values based on Standard Concrete

Specifications (for 28 specimens)

Failure mode

O S type A SB type + BS type x x B type

+ + + + +

4*

+ -H- +

10 3 X *

I A + -

ft +

® ttC XiT+X XXX +

07 7

X

T X „ * v x

*

0.2 -* * +

i

*

X -X

0 0 O' 0.2 0.3 0,4 0.5 0.6 0.7 0.8 0.9 1.0 Lateral tie ratio Pi (%)

Mean value= 1.761

Coefficient of variation=39 0%

0.0 0.5 1 0 1.5 2 0 2.5 3.0 3.5 4.0 Shear span ratio a/d

Fig. 3.6 Relationship between tensile refinforcement

ratio and lateral tie ratio (by failure mode)

Fig. 3.8 Comparison between experimental values anc

calculated values based on Standard Concrete

Specifications (for 41 specimens)

71
Fig. 3.9

Fig. 3.

: 20

:

' 5

Tmax

Teal

-

Mean value= 1 481

Coefficient of variation=27. 1%

0

0 O

0 ©oo ©

s

oo

o

8

O o

OGDO O

O

o o

0

1 1 1 1 1 'III 0.0 0 5 1 0 1.5 2.0 2.5 3.0 3 5 4 0 4.5 5.0 Shear span ratio a/d

oS 3 5

e

i “

1.5 Tmax

Teal

1.0

Mean value= 1 .01

1

Coefficient of variation=24 8%

oo

oo

0.0 0 5 1.0 1 5 2.0 2.5 3.0 3 5 4.0 4 5 5.0

Shear span ratio a/d

Comparison between experimental values and

calculated values on the basis of Japan National

Railways’ euqation (for 28 specimens)

Fig. 3.11 Comparison between experimental values

and calculated values based on the proposed

equation (for 28 specimens)

4 5 -I

Mean value=1 347

Coefficient of variation=31 .2% Mean value=0 979

Coefficient of variation=26 0%

10

1

1 5 -j Tmax

5 2 5

- e

E

I *

fTJ

l 5 Tmax

Teal

i 0

00

o oo . 8 e

0 0 0.5 1.0 1.5 2.0 2.5 3 0 3.5 4 0 4,5 5.0 Shear span ratio a/d

10 Comparison between experimental value and

calculated values on the basis Japan National

Railways’ equation (for 41 specimens)

Fig. 3.12 Comparison between experimental values

and calculated values based on the proposed

equation (for 41 specimens)

72
Fig. 3.13 Minimum lateral tie ratio Pw flexural failure type (B type, BS type) specimens Fig. 3.15 Minimum ratio of lateral ties, Pw (yc)

Bending Shear

moment force

diagram diagram

Shear span

Mmax

Smax

Fig. 3.14 Checking position and concept of a/d

73
Some Recent Studies on Lifeline Earthquakie Engineering

in the United States: Experimental and Field Study

BY

T. D. O’Rourke^ and J. Isenberg^

ABSTRACT

This paper focuses on research pertaining to pipeline systems. It concentrates on

experimental and field studies, and begins with

a genera] description of earthquake effects on

pipelines. Pipeline response to permanent

ground movements is discussed, with a summary

of observations during previous earthquakes, a discussion of soil-pipeline interaction models,

and a review of previous laboratory

experiments, as well as a planned field

experiment at a pipeline fault crossing. Pipeline

response to traveling ground waves is discussed

similarly in terms of previous earthquake

observations, soil-pipeline interaction models,

and a review of laboratory and field

experiments. A summary of experimental trends

is given and recommendations are made

regarding future research opportunities.

KEYWORDS: Earthquakes; Geotechnical

Engineering; Lifelines; Pipelines; Soil-Structure

Interaction

1. INTRODUCTION

Lifelines are systems which proxr ide

transportation for people, resources, and

services, as well as the transmission of electric

power, messages, and data. The American

Society of Civil Engineers (1) has identified four

principal lifeline systems: electric power and

telecommunications, gas and liquid fuel,

transportation, and water and sewage. The

great range of operational criteria as well as the many different mechanical and structural

features of lifelines add to their complexity, and

make it difficult to characterize research efforts

in a truly comprehensive way for all types of

lifelines. This paper focuses on research

pertaining to pipeline systems. A substantial

amount of the lifeline earthquake engineering

research in the United States has been devoted

to pipelines, principally as a result of the severe damage sustained by pipelines during

the 1971 San Fernando (2,3,4) and 1906 San

Francisco earthquakes (5,6,7). Research results

for pipelines are applicable to gas, liquid fuel,

water supply, and sewage systems, as well as various types of cabling and ducts in electric

and telecommunication systems.

The paper concentrates on experimental and

field studies, and begins with a general

description of earthquake effects on pipeline

systems. Pipeline response to permanent

ground movements is discussed, with a summary

of observations during previous earthquakes,

a discussion of soil-pipeline interaction models,

and a review of previous laboratory experiments

as well as a planned field experiment at a pipeline fault crossing. Pipeline response to traveling ground waves is organized in a similar

manner with attention to previous earthquake

observations, models for soil-pipeline interaction,

and laboratory and field experiments.

2. CONCEPTUAL FRAMEWORK FOR PIPELINE

' EXPERIMENTS

Laboratory and field experimentation for pipeline

systems should be integrated within a framework

of previous earthquake observations, analytical

model development, and applications in practice

through appropriate design and siting

procedures. A rational plan for research and

development includes: 1) characterization of the

loading environment, 2) assessment of the loading

effects on continuous and jointed pipelines, as

well as special valves, pumps, and pressure

control devices, 3) experiments designed and

organized in conjunction with specific analytical

models of soil-pipeline interaction, and 4) the use

of experimentally verified models for

improvements in siting, design, and risk

assessment. Each of these steps is discussed

under a subheading which follows.

2.1 Earthquake Loading Environment

Whereas aboveground piping responds to seismic

shaking in the same manner as other

aboveground structures and equipment, buried

piping responds to earthquake motions in such a way as to have nearly the same curx'atures and

strains as the ground itself. The earthquake

response of a buried pipeline depends closely on

the static and dynamic deformation, stiffness, and

strength of the surrounding soil. An important

part of characterizing the earthquake loading

environment is a careful assessment of soil

properties and deformation patterns, especially

when the potential exists for ground failures.

In response to an earthquake, pipelines may be

damaged by permanent ground deformation and

traveling ground waves. Both effects may have

important repercussions on pipeline performance,

although damage from traveling ground waves is frequently concentrated in pipelines which have

been weakened by corrosion or are constructed

of brittle materials such as cast iron

or asbestos cement. Permanent ground movements

1 Cornell University, Ithaca, NY 14853

^Weidlinger Associates, Palo Alto, CA 94304

74
result from irreversible deformations caused by

shear failure or volume change, and develop

concurrently with or subsequent to seismic

shaking. Because permanent movements

frequently exceed the peak ground

displacements from seismic waves, they should

be regarded as conditions of maximum distortion

for buried pipelines and cables.

2.2

Assessment of Loading Effects

Three modes of unacceptable performance can be

identified for buried pipelines: 1) rupture or excessive deformation of the pipe, 2) excessive

rotation at pipeline joints, and 3) pullout or excessive compressive distortion at pipeline

joints.

The performance of metallic pipelines, which

convey high pressure gas and liquid fuel,

depends almost exclusively on the first mode.

Modern high pressure facilities are constructed

with full penetration butt and longitudinal seam

welds, inspected by means of X-ray radiographs

or other nondestructive tests of the welds, and

composed of mild or high strength steels.

There is no well-documented case of a buried or aboveground petroleum or natural gas

transmission pipeline constructed according to

modern standards having been ruptured by the

effects of ground shaking. Large permanent

ground deformations associated with landslides,

surface faulting, and soil liquefaction are the

principal earthquake hazards. Continuous,

butt-welded pipelines possess good ductility,

and can be deformed well into the plastic range

without rupture. Compressive forces, however,

may lead to wrinkling and kinking of the pipe

wall so that additional deformation of the line

tends to concentrate at one location with

resulting instability. Permanent ground

movements which contribute to high compressive

strains in the pipe are therefore undesirable.

Toe areas of lateral spreads and landslides as

well as reverse faults are potential locations of

large compressive deformation. Pipeline

orientation with respect to transverse soil

movement also plays an important role in the

development of pipeline strain. At fault

crossings, a pipeline should be oriented to

respond in tension to differential movement. In

landslide and lateral spread areas, a pipeline

should be oriented parallel to natural

topographic contours to promote longitudinal

flexure and tensile elongation, thereby

minimizing compressive distortion of the line.

Jointed pipelines are subject to all three modes

of potentially unacceptable deformation. In some

cases, the effects of permanent as well as transient ground deformations may be

concentrated at pipeline joints so that their

performance under both static and dynamic

excitation needs to be understood. Unacceptable

performance may result from leakage at a joint

without a structural failure such as pullout or

material fracture. Joint response to earthquake

loads should therefore be evaluated under the

conditions of internal fluid pressure that are

likely to prevail in the field.

Jointed pipelines are used typically in water

trunk and distribution systems, sewage collection

systems, and natural gas distribution networks.

The great variety of joint types presents a challenge to the experimentalist in deciding which

joints to investigate. Many U. S. distribution

systems rely on relatively old cast iron pipelines,

which were constructed with lead-caulked or cement joints. Because of creep characteristics,

lead-caulked joints are likely to show significant

differences in behavior for static relative to

dynamic loading.

There are a host of valving, pumping, and

pressure control devices which also deserve

careful evaluation in assessing the earthquake

response of pipeline systems. Of special

importance are penetrations and connections

associated with subsurface and aboveground

structures, where locally high concentrations of

strain are possible.

2.3 Experiments Guided by Analytical Models

Experiments should be designed and conducted in

accordance with analytical models. This is

especially true for soil-pipeline interaction, and

applies to both laboratory and field experiments.

The use of analytical models as an integral part

of the experimental process carries a two-fold

advantage. It provides a rational means for

interpreting the results; data can be assessed

immediately, and potential sources of error can

be recognized as well as additional tests or test

modifications chosen for further refinement.

Coordination with specific models ensures that

the testing can be used to verify, or change

existing analytical procedures. This allows for

consistent and systematic improvements in

design. With the availability of sophisticated,

portable computers and data acquisition systems,

it is possible to conduct analytical simulations

and testing simultaneously, with real-time

comparisons and control of the testing process.

2.4 Improvements in Practice

Experimentation provides a rational basis for

selecting analytical models appropriate for

design, and of equal importance, for

understanding the limitations of these models. In

some cases, the uncertainty associated with the

performance of a pipeline component can be

quantified by probabilistic means for reliability

studies. Experiments also point directly to modifications in the materials or configuration of a given component to improve its seismic

response. The same experimental methods can be

used to substantiate the improved performance.

3. PIPELINE RESPONSE TO PERMANENT GROUND

MOVEMENT

75
3.1 Previous Observations

Large permanent ground movements associated

with soil liquefaction, landslides, and surface

faulting are the most troublesome sources of

potential damage to gas, liquid fuel, and water

transmission pipelines. Evidence from the 1906

San Francisco, 1952 Kern County, 1964 Niigata,

1964 Alaska, 1971 San Fernando, 1978 Miyagi-KenOki, 1983 Nihonkai-Chubu, and 1987 Ecuador

earthquakes shows an unmistakable correlation

between permanent ground movement and buried

pipeline damage. Some of the most graphic

illustrations of this kind are associated with the

1971 San Fernando earthquake, during which

surface faulting and liquefaction-induced lateral

spreads and slope movements caused failures in

gas transmission and distribution piping (8,9).

Table 1 summarizes the principal causes of

permanent ground movement associated with

earthquakes. Each category of displacement is described briefly, and specific types of

movement within a given category are listed.

Slope instability is a cause of potentially

damaging ground movement, with the greatest

seismic risks related to slopes showing signs of

instability under static conditions. It is

unlikely that any pipeline can survive the severe deformation imposed by deep

translational and ' rotational slope failures. If instability involves slumps and relatively shallow

slides, slope stabilization may be an effective

means of correcting difficulties. Field

observations and analytical studies suggest that

modern, welded steel pipelines can survive the

deformations imposed by relatively shallow and

limited slope displacements, provided that the

line is oriented perpendicular to the anticipated

direction of movement.

In addition to landslides, the principal

geotechnical hazards associated with

earthquakes are surface faulting and soil

liquefaction. Surface faulting is often regarded

as the most dangerous source of differential

movement, mostly because visible fault

displacements have been the subjects of some

striking and well-publicized observations.

Surface faulting, however, is a relatively rare occurrence. In contrast, soil displacements

caused by liquefaction are comparable in

magnitude and more widespread. Liquefaction can result in many different types of

deformation, including flow failures, lateral

spreads, subsidence, loss of bearing, and

buoyancy effects.

The most troublesome types of liquefactioninduced movements for a buried pipeline are

lateral spreads. Lateral spreads involve the

predominantly horizontal movement of competent

shallow soils on an underlying liquefied

deposit. Maximum displacements are often

several feet, and may occur on slopes as gentle

as 0.5 to 1.0*.

Assessing liquefaction and landslide hazards

for pipelines requires careful attention to geologic features, and may require an assessment

of structure, morphology, and topographic

characteristics. Two case histories help illustrate

the importance of permanent ground movements

on pipeline damage.

The first case involves lateral spreading near the

Juvenile Hall and Upper Van Norman Reservoir

during the 1971 San Fernando earthquake. The

information in this example is derived from work

by O’Rourke (10). Figure la shows a plan view

of the bedrock and alluvial sediments

surrounding the Upper Van Norman Reservoir.

Superimposed on this view are the lateral spread,

which occurred at the site, and contours of equal

groundwater depth at the time of the

earthquake. The soils in the area are chiefly

sands, silty sands, and silts, which were laid

down in alluvial fans from Weldon and Grapevine

Canyons to fill a northeast-trending depression

in the bedrock. The location of the lateral

spread can be seen clearly as trending northeast

of the reservoir between bedrock exposures.

The presence of a loose sandy silt and fine sand

at a depth of 6 to 9 m with the water table at 3

to 6 m were characteristics that made the site

susceptible to liquefaction. The lateral spreading

and pattern of ground displacement, however,

were controlled by the structure of the bedrock

depression and the geomorphic. features of the

alluvial fans which filled the depression.

Figure lb shows the mapped ground deformation,

with vectors of displacement reported by Youd

(11). The maximum soil movement was

approximately 1.75 m. Two high-pressure natural

gas pipelines were ruptured in the zone of

lateral spreading. Both were continuous,

girth-welded steel pipelines, with welds dating

from approximately 1926 to 1930.

Figure lc shows the boundary scarps of the

lateral spread and vector displacements

superimposed on a contour map of the site.

There is a clear relationship between the pattern

of spreading and the local topography. The

lateral spread was confined generally to lower

elevations between steeply rising ground to both

the northwest and southeast. Moreover, the

displacement vectors are nearly perpendicular to the contour lines.

Two earthquakes, which measured 6.1 and 6.9 on

the Richter Scale, occurred on March 5, 1987 in

the Napo Province of northeast Ecuador. The

earthquakes occurred after a month of heavy

rains, during which a total of 610 mm of

precipitation was measured in the area of the

earthquakes. The earthquake epicenters were

located near an active volcano, El Reventador.

Strong ground shaking triggered rock avalanches

and debris slides near the volcano, involving over 100 million m^ of soil and rock. Landslide

material was washed into the surrounding rivers,

causing them to swell with mud and debris. The

swollen rivers flooded mountain roads and

76
dwellings, leaving deposits of mud and saturated

sediments as deep as 10 m. The flooding and

debris slides were responsible for most of the

loss of life, which is estimated between

1000-2000 people.

Of major importance is the damage from

landslides to the Trans-Ecuadorian Pipeline.

The failure of this facility represents the

'largest single pipeline loss in history. The

Trans-Ecuadorian Pipeline is a 660-mm-diameter

pipeline, which carries oil 420 km from the

Ecuadorian oil fields east of the Andes

Mountains to the port of Esmeraldas on the

Pacific Ocean. The pipeline was of telescoping

design and constructed of X-60 grade steel,

ranging in wall thickness from 9.5 to 23 mm. In

all locations of damage, the pipeline was

constructed as an aboveground facility,

supported on 12-m centers by concrete pedestal

foundations or double pipe piles with cross beams.

As shown in Figure 2, approximately 10.5 km of

the pipeline have been completely destroyed

along the north bank of the Coca River east of

its confluence with the Salado River. An

additional 16 km of the pipeline, east of the

Malo River, have been damaged by local debris

slides, which have severed the pipeline in at

least eight places. Eight km of pipeline west of

the Salado River have deformed, with significant

distortion and displacement of aboveground

pipeline supports.

Loss of the Trans-Ecuadorian Pipeline has

deprived the country of 60% of its export revenue. It contributed to a 9% increase in the

world price of oil. The lost revenues and cost

of reconstruction of this facility are estimated

as Si to $1.5 billion.

3.2 Models for Soil-Pipeline Interaction

The ground movement patterns associated with

surface faulting, landslides, and lateral

spreading have many characteristics in common.

The maximum relative displacements in landslides

and lateral spreads are concentrated at the

slide margins where movements tend to replicate

strike and normal faulting and, thus, may be

treated by many of the same modeling

techniques that apply for surface faults.

O’Rourke, Grigoriu, and Khater (12) point out

that, although it is not possible to model with

accuracy the soil displacement patterns at all potentially vulnerable sites, it is possible to

characterize the worst case conditions which

may develop and use these as upper bound

estimates of deformation. In general, the most

destructive differential soil movements are those

which develop abruptly, such that a singleplane or knife-edge model of displacement can

be used as an effective and conservative basis

for analysis.

Figures 3a and b show a plan view of a pipeline

which is intersected by a right lateral strikeslip fault at an angle, 0f. The pipeline is

oriented so that fault displacement, df, will

cause tension in the buried pipeline. It is

assumed that sufficient geologic evidence is

available to orient the pipeline for tensile

deformation. Newmark and Hall (13) analyzed the

pipeline deformation as an antisymmetric pattern

of circular arcs, with each circular segment

spanning the fault centerline and the location of an anchor point. The distance between the fault

and the anchor point is known as the anchor

length, La. Anchors may be caused by bends,

tie-ins, and other features which develop

substantial resistance to axial movement.

Alternatively, the anchor point may represent an

effective anchor length, beyond which there is no axial stress imposed in the pipeline from fault

movement.

An analytical model was developed by Kennedy,

Chow, and Williamson (14) which accounts for

increased frictional resistance near the fault

crossing and for pipeline bending strains.

Figure 4 shows a plan view of pipeline

deformation which forms the basis for the

Kennedy, Chow, and Williamson model. Only

one-half of the antisymmetric pattern of

deformation is shown. As in Figure 3, the

anchor point represents either a local constraint

or the end of the effective anchor length. It is assumed that greater frictional resistance is

developed along the curved portion of the

pipeline than along adjoining sections because of

high horizontal soil pressures resulting from

large relative movements between the pipeline

and surrounding ground. Kennedy, Chow, and

Williamson recommend that the frictional

resistance per unit pipe area be taken between

2.4 and 3.3f, where f is the frictional resistance

per unit surface area of the pipe. Between the

end of the curved portion of the pipeline and

the anchor point, the frictional resistance is

assumed to be f.

Figure 5 compares the results from both the

Kennedy, Chow, and Williamson and the Newmark

and Hall models for a case analyzed and reported

by O’Rourke, Grigoriu, and Khater (12). The

results pertain to a pipeline of X-60 grade steel,

buried in granular soil with 7 = 19.7 kN/m^,

depth to center of pipe of 1.2 m, and interface

friction angle between the pipe and surrounding

soil of 27*. The fault displacement, df, at a maximum axial pipe strain of 0.03 is expressed as

a multiple of the pipe wall thickness, t. It should be pointed out that the maximum axial

pipeline strain does not depend on the pipe

diameter. The strain depends solely on the axial

elongation of the pipeline.

The Kennedy, Chow, and Williamson model can be seen in Figure 5 to give a more conservative

result, than the Newmark and Hall model, chiefly

because of the increased level of frictional

resistance along the curved portion of the

pipeline. In the figure, the plots are developed

for the recommended range of frictional

resistance of 2.4 to 3.3f. Over a wide range of

77
£f, the Kennedy, Chow, and Williamson model

predicts about half the fault movement capacity

as the Newmark and Hall model.

Detailed analytical studies of buried pipeline

response to fault displacement were performed

by Tawfik and O’Rourke (15) with the aid of a computer program developed especially for

evaluating pipeline performance for elastoplastic soil reactions and large plastic strains in

the pipeline steel. Their analytical results show

that the pipeline deformation differs

substantially from a circular arc and that the

full value of f does not occur at all locations, as assumed in the Kennedy, Chow, and Williamson

and Newmark and Hall models. The analytical

results obtained by Tawfik and O’Rourke show

strains higher than those predicted under

similar conditions by the Newmark and Hall

model, and substantially lower than those given

by the Kennedy, Chow, and Williamson model.

Given the importance of this type of soilpipeline interaction for many high pressure

transmission pipelines and the wide range in

analytical results, it is extremely important to substantiate by experiment the correct modeling

assumptions. A major difficulty in conducting

an appropriate experiment is the large scale

required to comply with the boundary conditions

that represent actual pipeline intersections with

soil ruptures. Analyses (12) have shown that

the effective anchor length of the pipeline

required to develop plastic strains in the steel

may be several thousand times the pipe wall

thickness. Experiments at these scales must be

conducted in the field at locations with very

special geologic characteristics.

3.3 Laboratory Experiments

Buried pipeline response to large permanent

ground movements depends closely on the forcedisplacement relationships which characterize

reactions between the pipeline and surrounding

soil. These relationships involve an elastoplastic response of the soil, which can be

represented as a nonlinear or bilinear function.

Analytical models for soil-pipeline interaction

have evolved from methods which were

originally developed to evaluate the lateral and

longitudinal soil reactions against deep

foundations. Experimental work in the U. S. has

focused on characterizing the force-displacement

relationships for buried piping subject to longitudinal and lateral displacement with

respect to the soil.

Full-scale tests of buried piping with diameters

ranging from 25 to 300 mm have been performed

in the laboratory (16,17,18) to evaluate the

lateral and uplift forces generated by relative

movement between pipe and soil as a function of

soil density and strength, depth of burial, pipe

diameter, and surface roughness of the pipe.

The force-displacement relationship has been

shown to be modeled by means of a hyperbolic

function, or in simplified form as a bilinear

representation. Data have been acquired which

allow the bilinear relationship to be prescribed

as a function of soil density and burial depth

(17).

Currently, experimental evidence is lacking for

oblique pipeline movement, involving components

of both vertical and horizontal displacement with

respect to the soil. In addition, data are also

absent for characterizing soil restraint for

pipelines buried in fine-grained soils such as clay and silt.

Tests have been performed to evaluate the

frictional resistance generated by axial movement

of a pipe in granular soil (19). These test

results tend to corroborate previous experimental

evidence for the frictional interface

characteristics of deep foundations and retaining

walls. Laboratory tests on the flexible joints of

ductile iron pipe have been conducted to characterize pullout and rotational capacity (20).

On the basis of these tests, the moment-rotation

relationship was shown to depend on the

properties of the waterproof gaskets provided

with the joints.

3.4 Field Experiments

To investigate the performance of pipelines under

conditions where permanent ground displacements

are critical, a field experiment has been proposed

that capitalizes on the predicted recurrence of

the 1966 Parkfield earthquake (21) on the San

Andreas Fault in Central California. According to the U. S. Geological Survey (22), there is a high

probability that a magnitude 6 or greater

earthquake will occur there between 1986 and

1993. Seven earthquakes have occurred on this

segment of the fault over the last 150 years with an average return period of 22 years. The

purpose of the experiment is to study the

validity of current models of welded and jointed

pipe segments in and adjacent to the surface zone of fault rupture. For welded steel pipe, the

critical design parameter is the effective anchor

length as defined in Section 3.2. Different

analytical approaches lead to significantly

different anchor lengths and hence, in some

cases, to significantly different designs at fault

crossings. For jointed pipelines, it is assumed

that ground movement is accommodated almost

entirely by relative displacement and rotation at

the joints. The Parkfield experiment may show

whether the component of fault displacement in

the direction of the pipe axis is equal to the

joint displacement and, if there are multiple

joints in the fault zone, whether they

accommodate the fault displacement equally.

To address these and other important issues,

both continuous and jointed pipes will be laid in

trenches across the fault zone in several

different locations along the Parkfield segment.

Active instrumentation in the form of strain

gages and recording equipment that will be

automatically triggered by the earthquake will be

relied upon. Joint deformations and rotations

78
will also be observed visually.

3.4.1 Jointed Pipeline Experiment

The experiments will investigate the relationship

between crossing angle and joint deformation

and the behavior of joints and pipeline

segments just outside the surface fault zone.

Ductile iron pipe with rubber gasket joints will

be used. Two test beds are illustrated in

Figure 6. In each bed, two pipe segments are used in tension and one is in compression. In

the narrower bed there will be one joint and no

complete segments; in the wider bed there will

be two joints and one complete segment. Of

course, the actual surface rupture zone does

not have distinct boundaries that are shown in

the idealized zone of Figure 6. To locate the

most likely region of surface rupture and to estimate the width of the rupture zone, we will

rely on U.S.G.S. measurements of fault creep

which we assume to be premonitory to surface

rupture.

According to one simple picture of pipeline

response to fault movement, relative

displacement along the axis of the pipe is

6 rel = a cos @ (1)

where a - fault offset and © = angle of pipe

axis with respect to fault strike. For A r 100

mm, the joint must accommodate about 100 mm

times cos30° “ 84 mm, which is close to the

allowable for rubber gasket, ductile iron pipe.

For 9 - 60*, only 50 mm of slip must be

accommodated, which should be possible without

leakage. What is unresolved by simple theory is how much joint rotation accompanies the fault

displacement. The amount by which rotation

impairs axial slip capacity can be tested in a laboratory. The Parkfield experiment will show

how much joint rotation, if any, actually occurs

under the conditions of the test. A second

issue is whether, if fault displacement is 100

mm, the pipe segment in compression will fail.

For the jointed pipeline segments, passive

instrumentation in the form of markers and

viewing ports will be designed to permit visual

observations. The viewing port illustrated in

Figure 7 will be fitted with a demountable

optical device to measure relative displacement

and rotation between adjacent segments. Active

instrumentation will be in the form of

transducers whose active element relies on

change of resistance. Three measurements

corresponding to two rotations and a displacement in the direction of the pipe axis

will be made, leading to a total of 27 active

channels.

The effects of fault creep (up to 10 mm/year)

will complicate interpretation of measurements,

because joint displacement during preearthquake creep may reduce capacity to withstand subsequent earthquake

displacements. Pipeline response to creep will

be monitored at regular intervals so that the

condition of the pipe joints at the onset of the

earthquake will be known.

3.4.2 Welded Steel Pipeline

The purpose of the welded steel pipeline

experiment is to study the mode of deformation

and slip between pipe and adjacent soil as a function of distance from the fault zone. As

before, we idealize the fault zone as having

distinct boundaries. Observations of pipeline

behavior will help evaluate how this common

oversimplification affects analysis and design.

The geometry of the experiment is shown in

Figure 8.

According to present estimates, the pipe will not

rupture nor will the deformations be

concentrated in a narrow zone. Hence it is necessary to measure flexural and extensional

components of strain along the length of the pipe

using strain gages and an appropriate recording

system. Two pipeline segments will be used, one

of which will be continuously welded, whereas

the other will have joints specially designed to accommodate fault displacement.

Three hypothetical variations of axial strains with

distance along the pipe are illustrated in Figure

9. Approximately a dozen measurement points on one side of the fault zone center (west side in

Figure 9) will be instrumented. These

hypothetical strain data correspond to three

different variations of shear stress distribution.

Finding out which of these, if any, is correct is a major goal of the experiment. A few

measurement points will be provided on the east

side to check on symmetry of shear distribution.

If the distribution is symmetrical it would imply

that the ground displacement is symmetrical

about the center point of the pipe, which would

be very fortunate.

The experiment provides an excellent opportunity

to test the influence of joints on pipeline

displacements and strain patterns. The use of

joints and special couplings has been

recommended as a means of accommodating the

vertical and horizontal components of permanent

ground movement. Traveling ground waves,

however, present a special difficulty for jointed

pipelines and it is not known how the repeated

displacements from dynamic excitation will be

concentrated at the joints nor how it will affect

their behavior. At the Parkfield fault crossing, an additional welded steel pipeline will be

constructed with joints at the location of

expected maximum ground displacement. These

will allow for rotation and axial extension to accommodate fault movement. Strain gages will

be placed at selected locations which coincide

with those on the continuous pipeline so that the

joint effects on moment and axial strain

distribution can be evaluated.

Ideally, the welded steel pipeline segments should

extend beyond the anchor distance calculated

79
according to current theories. It should also be

approximately one-quarter to one-half wavelength

long. According to the Newmark-Hall approach

(13), the anchor length is about 107 m. This is

longer than one-quarter wavelength and slightly

less than one-half wavelength. Accordingly, a

total length of 214 m represents a reasonable

compromise.

Uncertainties arising out of compromises on

length of pipeline can be eliminated by

introducing a fixed anchor point at the end of

the pipe where relative displacement between

soil and pipe would be known. So far no

reason has been found to require us to introduce artificial constraints, however.

3.4.3 Data Acquisition System

Components of a data acquisition system include

strain gages, underground cables, amplifiers, a seismic trigger, recording device, and storage

batteries with solar panel for recharging. Ten

hz resolution will be achieved. The trigger will

be set so as not to activate recording due to creep strains, which will be monitored by

periodic on-site visits.

4. PIPELINE RESPONSE TO TRAVELING GROUND

WAVES

4.1 Observations in Previous Earthquakes

Extensive empirical studies have been made to

measure the performance of pipelines in recent

earthquakes. Study of five United States

earthquakes (23,24,25) indicates that, where

wave effects dominate ground shaking, most

leaks in ferrous pipes occur where corrosion

has locally reduced pipe wall thicknesses.

Bends, tees, and joints are also locations of

frequent damage (26). The commonly accepted

measure of earthquake damage to underground

pipelines is the number of leaks or required

repairs per unit length of pipe. Postearthquake reconnaissance indicates that the

promptly appearing damage can be roughly

correlated with a measure of ground shaking

intensity such as peak acceleration, peak

ground strain, or other parameter which

characterizes ground motion as due to wave

propagation effects. These correlations are

useful for predicting damage in future

earthquakes and in probabilistic system models.

The most thoroughly studied U. S. earthquake is

the 1971 San Fernando event, where ground

rupture was responsible for numerous breaks in

the northern San Fernando Valley and where

wave propagation was responsible for less severe damage in the southern San Fernando

Valley. Data on repair frequency compiled by

Katayama (27), Figure 10, is correlated with

peak acceleration, which enables observations in

San Fernando to be compared with Japanese

experience. Figure 10 also shows that repair

frequency following the 1983 Coalinga, California

earthquake agrees reasonably well with the

other data (28).

Another possible choice of shaking intensity is ground strain, which can be estimated by

subtracting integrated ground motion records

that are separated by a known measure of wave

speed, such as group velocity. Recent work

applying these two methods to data from the San

Fernando earthquake is reported in (29) and is illustrated in Figure 11. Other measures of

shaking intensity include Modified Mercalli

Intensity as applied in (30).

The main obstacle to increasing the data base of

damage and repair statistics on the basis of

historical earthquakes is lack of reliable data on

intensity of local ground shaking. This rules out

otherwise useful data from the 1965 Puget Sound,

Washington and 1969 Santa Rosa, California

earthquakes (31).

It should be noted that the data discussed above

come from either Japanese or California

earthquakes. The possibility that the data are

biased due to ground shaking characteristic of

those regions or to other regional anomaly

requires further study. It should also be noted

that the data represent prompt earthquake

damage which appears within several days to several weeks following the earthquake. Detailed

study of the San Fernando and Coalinga

earthquakes suggests that, in addition to causing

prompt damage, earthquakes may accelerate aging

of pipeline systems, thus shortening their

effective life and permanently increasing

maintenance costs.

4.2 Models for Soil-Pipeline Interaction

A summary of the response of pipelines to transient wave effects is presented in (12). A

chief assumption of the methods reviewed there

is that pipe deformation is assumed to be

governed by the strains and displacements of the

surrounding soil. This assumption is consistent

with field measurements of Japanese researchers

(32,33,34,35,36). If it is also assumed that

transient ground shaking can be represented by

waves of constant shape, then motions at points

along various points of the pipeline will differ

only by a lag time on an absolute time scale.

These assumptions have led to simplified

relationships in which axial pipeline strains are

directly proportional to the ground velocity and

inversely proportional to propagation velocity.

Curvatures, usually small, can be related directly

to acceleration and inversely to the square of

propagation velocity.

Analytical methods for straight pipelines can be

grouped into two categories, depending on whether the pipelines are continuous or jointed.

The analytical methods of continuous pipelines

represent seismic input by specified distributions

of ground displacements (37,38) or shearing

stresses (39). These distributions are assumed

to be sinusoidal over the ground surface and are

applied along pipelines at locations which

80

j
maximize response. Methods which assume buried

pipelines to be rigidly bonded to the soil

(40,41,42) lead to expressions for axial and

bending strains which depend on the apparent

propagation velocity of the traveling ground

wave, c, along the pipeline. Rayleigh wave

velocity and combinations of dilatational, shear,

and Rayleigh velocities are recommended as a basis for c. The American Society of Civil

Engineers (ASCE) Committee on Seismic Analysis

recommends an apparent wave velocity between

610 and 915 m/sec (43).

The effects of relative displacement between soil

and pipeline can be considered by assuming

that slip occurs near the ends of the pipeline to mobilize a maximum frictional force per unit pipe

length, t u . Outside the regions of slippage, the

pipe and soil move together and, as a result,

the pipeline strain coincides with the maximum

soil strain, eg. The maximum slippage length is

t - e g

EA/tu (2)

in which A is the cross-sectional area of the

pipe and E is its modulus.

The maximum frictional force per unit pipe

length, t u , can be obtained from

tu = |D (1 + K0 ) y H tan6 ( 3

where y is the unit weight of soil, H is depth

of the pipe, K0 is the coefficient of earth

pressure at rest, and 6 is the interface friction

angle between the soil and pipe. It follows from

this model that the maximum strain in long

pipelines for full slip, where ea eg, is

ea = Xty/4 EA (4)

and this occurs over a length of pipeline of onequarter wavelength, X. Other methods of

analysis account for, in addition to peak ground

displacement and/or velocity, the shape of the

seismic ground waves (37,38). These methods

involve solutions of differential equations of

equilibrium for specified locations of the seismic

waves and can account for effects of slippage

between soil and pipe.

Seismic excitation has also been characterized

by distribution of shearing stress along

pipelines (39). If such a distribution varies

sinusoidally, in which tQ is the maximum

shearing stress, it is assumed that slip occurs when t( x ) exceeds a postulated critical value,

t cr , and follows the pattern in Figure 12 for

large earthquakes. The maximum stress in the

pipeline is

At0

/2rrt; no slip

ff max =

i ^

Xtcr /4t; slip over full length

Static analysis of jointed pipelines provides an upper bound on the relative displacement and

rotation in joints (44). The model in Figure 13

idealizes the soil as linearly elastic springs

and pipe segments as rigid. If joint forces are neglected and the wavelength is long compared

with pipe dimensions, the maximum relative joint

displacement and rotation are

uj = ^L/ci;

( 6 ) ej - aL/ct2

where eg and ct are longitudinal and transverse

components of the apparent wave propagation

velocity and L is the length of pipe between

joints.

A more refined model of the soil-pipe-joint

system is shown in Figure 14 (45). The advances

over the model described above are the addition

of dampers to represent radiation damping

effects of pipe-soil interaction and the addition

of springs and dashpots to represent joint

behavior. Ground shaking is applied by means

of a uniform traveling excitation that propagates

with constant amplitude and velocity. Equations

of equilibrium incorporate both damping and

inertia (38,39,46,47). Reduced models that neglect

these terms are used in applications and have

been extended to cases of deformable pipe

segments (44,48). The most important steps in

developing these models are improvements in the

joint models, because these are potential weak

links, and soil-structure interaction, because this

governs the relative displacements transmitted to the pipeline.

Available methods for bends and tees deal with

continuous pipelines. The pipelines are composed

of elastic linear materials and the soils can be

represented by rigid-plastic or elasto-plastic

models. Pipe-soil models are illustrated in

Figures 15 and 16. In these figures, S and M

are shear and moment in the pipe segments,

respectively, and a* is the relative displacement

between the soil and Element 1. Analyses of

these and similar models are in (42,43,49). These

methods are based on the assumption that the

bend angle does not change, which is unrealistic

for bends in thin-wall pipeline. Models of these

pipelines should be based on curved thin-wall

tubes (50,51), which account for bend flexibility

and result in a lower bending moment in the

pipe. It can be shown that

ks &* _

Si - + 0M;

20

M

* zffk*EI

r0

7T

(7)

in which A*, S, and M have the same meaning as above, r Q is the bend radius, and k* = 1 - 9/(10 + 12 [4tr 0/kD)]. Note that the bending

moment, M, varies rapidly with the bend radius,

rQ, particularly for small r G.

4.3 Laboratory Experiments
One aim of laboratory experiments that investigate pipeline response to traveling waves is to develop data on resistance to axial movement of

buried pipes. A suite of dynamic and static

experiments is reported in (52). In the static

tests, a pipe segment was embedded in a soil

box and subjected to monotonically increasing

axial force. Average values of interface friction

were found as functions of depth to diameter

ratio; as expected, the maximum frictional

resistance varied directly with the burial

depth. Subsequently, dynamic tests were

performed in a similar test fixture. Hysteresis

curves were developed for cyclic loading and

dependence of the energy dissipation on frequency and burial depth was measured. A

similar series of experiments is reported in

(19). In these tests, steel, vitrified clay, and

ductile iron pipe materials were subjected to

axial and transverse shear loading. In some

cases, water was added to the backfill to simulate field conditions. The tests were

carried out in a pressurized test fixture to investigate a range of backfill depths. One use

of the data developed from such tests is in

frictional models of the type described in

Section 4.2.

4.4 Field Experiments

There are relatively few field-scale experiments

conducted in the U. S. that investigate pipeline

response to traveling waves. The experiment

that is proposed for the Parkfield area,

described above, may provide some useful data,

however. This is because there are portions of

the- welded steel line that are distant from the

anticipated zone of maximum surface offset and

which may therefore primarily be subjected to shaking from traveling waves. These portions

will be instrumented with strain gages and will

be within a few hundred meters of strong

motion seismographs so that it may be possible

to correlate the ground shaking with the pipe

response.

A second type of field experiment involves

subjecting ductile iron pipe to conditions under

which corrosion will proceed at different rates.

This test series, conducted by the Ductile Iron

Pipe Research Institute, will yield valuable

information on corrosion-related damage, which

is a major contributor to damage in pipelines

subjected to wave propagation effects.

5. CONCLUDING REMARKS

Experimental work in lifeline earthquake

engineering should be performed within a conceptual framework of 1 ) load characterization

according to permanent ground deformation and

traveling ground waves, 2) assessing the effects

of the loading environment on continuous welded

pipelines, jointed pipelines, and valving and

pressure control devices, 3) experiments

performed in conjunction with specific analytical

models, and 4) the use of experimentally verified

models for improvements in design, siting, and

risk assessment. Laboratory research in the

U. S. has focused on characterizing the static

and dynamic performance of pipeline joints and

developing force-displacement relationships for

longitudinal and transverse movements of buried

piping relative to the surrounding soil.

Significant questions exist about the proper

analytical assumptions for models of soil-pipeline

interaction at locations of earthquake-induced

soil ruptures, such as surface faulting and slip

at the margins of landslides and lateral spreads.

The large-scale effects assocated with this type

of problem require field testing at a site with

special geologic and seismic characteristics. A

test program to resolve design and siting

questions about pipeline performance has been

planned at a crossing of the San Andreas Fault

in Parkfield, California.

ACKNOWLEDGMENTS

This work was partially supported under

Research Account No. RF 150-6580A, sponsored by

the National Center for Earthquake Engineering

Research. The authors also wish to acknowledge

the National Science Foundation, ASCE Technical

Council of Lifeline Earthquake Engineering, and

Earthquake Engineering Research Institute for

their support in efforts leading to research

reported in this paper.

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15049-CU-4 , Department of Civil Engineering

and Engineering Mechanics, Columbia University, New York, NY.

Loh, C. H., A. H. S. Ang, and Y. K. Wen,

"Spatial Correlation Study of Strong Motion

Array Data with Application to Lifeline

Earthquake Engineering," Structural Research

Series No. 503 , University of Illinois at Urbana-Champaign, IL, Mar. 1983.

Nelson, I. and P. Weidlinger, "Role of

Coulomb Friction in the Seismic Response

of Buried Pipes - General Introduction and

Overview," Report No. 18 , Weidlinger Associates, New York, NY, Mar. 1982.

Newmark, N. M., "Problems in Wave Propagation in Soil and Rock," Proceedings , International Symposium on Wave Propagation and

Dynamic Properties of Earth Materials, Albuquerque, NM, pp. 7-26.

Sakurai, A. and T. Takahashi, "Dynamic

Stresses of Underground Pipelines during

Earthquakes," Proceedings , 4th World Conference on Earthquake Engineering,

Santiago, Chile, 1969, pp. 150-162.

Shah, H. H. and S. L. Chu, "Seismic Analysis of Underground Structural Elements,"

Journal of the Power Division , ASCE, Vol.

100, No. P01, July 1974, pp. 53-62.

Committee on Seismic Analysis, "Seismic Response of Buried Pipes and Structural Components," ASCE, New York, NY, 1983.

Wang, L. R. L., M. J. O’Rourke, and R. P. Pikul, "Seismic Vulnerability, Behavior and

Design of Buried Pipelines," Technical

Report No. 9 , Department of Civil Engineering, Rensselaer Polytechnical Institute,

Troy, NY, Mar. 1979.

Shinozuka, M., R. Y. Tan, and T. Koike,

"Serviceability of Water Transmission

Systems under Seismic Risk," Proceedings , 2nd Specialty Conference of the Technical

Council on Lifeline Earthquake Engineering,

ASCE, New York, NY, 1981, pp. 97-110.

Nelson, I. and P. Weidlinger, "Dynamic

Seismic Analysis of Long Sequential Lifelines," presented at the Dec. 10-15, 1978

ASME Pressure Vessels and Piping Division

Winter Annual Meeting, San Francisco, CA.

84
47. Wang, L. R. L. and K. M. Cheng, "Seismic

Response Behavior of Buried Pipelines,"

presented at the Dec. 10-15, 1978 ASME

Pressure Vessels and Piping Division Winter

Annual Meeting, San Francisco, CA.

48. Wang, L. R. L., "Quasi-Static Analysis Formulation for Straight Buried Piping

Systems," Technical Memorandum , Department

of Civil Engineering, Rensselaer Polytechnical Institute, Troy, NY, July 1978.

49. Goodling, E. C., "Seismic Stresses in

Buried Elbows," presented at the Apr. 2-6,

1979 ASCE Convention and Exposition,

Boston, MA (preprint 3595).

50. Committee on Gas and Liquid Fuel Lifelines,

"Guidelines for the Seismic Design of Oil

and Gas Pipeline Systems," ASCE, New York,

NY, 1984.

51. Goodling, E. C.» "Buried Piping - An

Analysis Procedure Update," Earthquake

Behavior and Safety of Oil and Gas Storage

Facilities, Buried Pipelines and Equipment, Vol. PVP-77, ASME, New York, NY, 1983, pp.

225-237.

52. Wang, L. R. L., "Role and Development of

Soil Parameters for Seismic Responses of

Buried Lifelines," Earthquake Behavior and

Safety of Oil and Gas Storage Facilities,

Buried Pipelines and Equipment, Vol. PVP77, ASME, New York, NY, 1983, pp. 312-323.

Table 1. Principal Causes and Descriptions of Permanent Ground Movement

Cause Description

Landslides Mass downs lope movements of soil or rock, which may be triggered

under static or seismic conditions. Many displacement patterns

are possible. Principal forms of movement include 1) rock falls,

25 relatively shallow slumping and sliding of soil, and 3) relatively deep translation and rotation of soil and rock.

Liquefaction Displacement caused by transformation of saturated cohesionless

soils to liquefied state or condition of substantially reduced

shear strength. Liquefaction-induced movements include 1) lateral

spreading, 2) flow failure, 3) loss of bearing, 4) subsidence, and

5 ) buoyancy effects

Faulting Displacement of adjacent portions of the earth’s crust. Movement

concentrated in relatively narrow fault zones. Principal types of

fault movement include 1) strike, 2) reverse, and 3) normal slip.

Tectonic Uplift and Subsidence Regional changes in dimension associated with tectonic activity.

Generally distributed over large areas.

Vibration-Induced Densification Settlements and ground cracks arising from a volume loss caused

by vibration of dry or partially saturated cohesionless soil; most

prominent as a result of earthquake vibrations.

*

85
legend:

Displacement Seal*:

0 3.0 0m —° fTT* Offset Displacement

Vector Displacement — Surface Rupture

.Trrrf< Scarp-Hacfturea on Downward Side

o

3o Sand Bolls Location of Pipeline Repair

Relative Sllp.m

• 9m up, *9 m down -Vertical Displacement Qai Quaternary Alluvium QTs Saugus Formation (ROCK)

b) Lateral Spread Movements

Figure 1. Plan View of Lateral Spread near Upper Van Norman Reservoir

after 1971 San Fernando Earthquake (10)

a) Before Fault Movement

b) After Fault Movement

Plan and Cross-Sectional

Continuous Pipeline at a Views of a

Fault Crossing

Figure 2 Map Showing Area of Pipeline Damage

fran 1987 Ecuador Earthquakes

86

Figure 3
Viewing port

Figure 4. Plan View of Pipeline and Fault Movement as Basis for Kennedy, Chow, and

Williamson Model (12)

Angle of Pipeline" Fault Intersection, /3f(deg)

Figure 5. Comparison of Analytical Results for

Different Fault Crossing Models (12)

Figure 7. Passive Instrumentation

Detail of possible Joinc in continuously

welded pipline *

Figure 6. Array of Jointed Pipeline Segrients at Figure 8. Continuously Welded Pipeline with Two

Various Crossing Angles to Fault Strike; Joints, Laid Adjacent to Continuously

Two Fault Zone Widths Considered Welded Pipeline without Joint

87
Repolrs/km

Figure 9. Locations of Strain Measurements, Derived Distribution of Axial Forces, and Corresponding

Implied Distribution of Interface Shears for Continuously Welded Pipeline

8.0

4.0

1.6

.80

.40

.16

.08

.04

1 (Average -soil 7 /

:ondltlons —

Poor condi son /

tlons-*j

n

i L“

Good

7 condi

>011 cions

fwj

_L_1_

^10

1 T

i /

/ DATA FF REF. 80M

1 J

'

1 Ai

B / • MANAGl ~P * N. SAf

/ VALLE'

1 T0KAC1

/ 1968

JA 1972

I FERNANDO

(, 1971

II - OKI

rA 1964 1948

1923

3EST0S

r pipe

= ACT ION

1

1

*• / a nuga'

/ - A FUKUI

1 B TOKYO

/ U

1 W/ ASI CEMEN'

2 SOME

LIQUEf

.05 .1 .25 .5 1.0 2.5

Peak ground acceleration (g)

Coal Inga

V X - Data from O'Rourke (£- — with C - 2.1 kn/sec) O Serein date calculated by Iaenberg

I

10 . 100 . 1000 .

Pipe Strain (10 mm/am)

Figure 10. Repairs to Water Distribution Pipelines in Previous Earthquakes (27)

Figure 11. Pipeline Damage vs. Pipe Strain

88
Shearing stress

{ E g 1 5 ) -7

in sail

Figure 14 . Mechanical Model for a Pipeline with

Joints (39)

Figure 12. Stresses in Soil and Pipelines (39)

Detail - A

(a) ( b)

Figure 15. Representation of the Forces,

Manents, and Displacements at a Pipeline Bend (12)

r Initial position

l

I jr jr" srr r" ear / mq —' L L J L l 1

Figure 13. Static Conditions for Pipeline Joint

Rotation (44)

Detail -A

Figure 16. Representation of the Forces, Manents,

and Displacements at a Continuous

Pipeline Tee (12)

89
Some Recent Studies on Lifeline Earthquake Engineering

in the United States; Analysis and Design

BY

Masanobu. Shinozuka^- and Leon R. L. Wang^

ABSTRACT

This paper reports some recent studies on analysis and design of lifelines in the United

States. It consists of two parts: Part A

deals with component design and analysis while

Part B deals with systems aspects. Dr. Leon R.

L. Wang is mainly responsible for Part A, while

Dr. Masanobu Shinozuka for Part B. A companion

paper written by O'Rourke and Isenberg [A.O]

reports on some recent experimental and field

studies on lifelines in the United States.

In Part A, the recent developments of component

analysis and design of buried pipelines

subjected to ground strains or liquefaction

induced by seismic shaking (ground wave

effects) and large ground movements by several

researchers in the United States are reviewed

and discussed including work of the authors.

Part B delineates the unresolved issues

associated with systems aspects of lifeline

earthquake engineering, particularly those of

telecommunication transmission facilities

network systems, with additional related

comments on water transmission pipeline

systems. The text in Part B is primarily based

on the work performed by Dr. Shinozuka and his

associates. Nevertheless, it is hoped that the

text describes physical and topological issues

that still need to be addressed for the purpose

of seismic systems analysis of telecommunications networks.

The results of the studies performed on component design and analysis of lifeline

systems as described in Part A of this paper

obviously play an important role in determining

system survivability and serviceability as described in Part B.

In both parts, emphasis will be placed on those

subjects matters that are well suited for joint

research between the United States and Japan.

KEYWORDS: Buried Pipeline; Ground Movement;

Lifeline; Seismic Shaking; Soil

Liquefaction; Telecommunication

PART A: COMPONENT DESIGN/ANALYSIS ASPECTS

A . 1 . INTRODUCTION

Lifeline systems, which include water and sewer lifelines, oil and gas lifelines, electric

power and telecommunication, transmission

lines, and transportation lifelines etc. have

been damaged heavily by recent earthquakes

around the world such as 1971 San Fernando

Earthquake, 1976 Tangshan Earthquake, 1978

Miyaki-Ken-Oki Earthquake, and 1985 Mexico City

Earthquake and among others. The earthquake

damage to these lifelines has recently been

receiving more attention because of the impact

of these systems upon the populus during and

after a major earthquake due to the loss of

fire fighting capabilities (water lifeline),

disruption of transportation of human and goods

(transportation lifeline) and energy (oil and

gas lifeline) and possibility of disease (sewer

lifeline), etc.

At the present time, scientists, researchers

and engineers in the United States, Japan, and

other countries are now actively engaged in

research for adequate design for all lifeline

systems. In the United States, the efforts are represented by several specialty conferences on

the subject [A.l to A. 9], Due to its diversity

of the subject, this part of the paper will

deal only with the component analysis and

design of buried pipelines. The discussions of

transportation lifelines, electrical power and

communication lifelines etc. are not included.

A. 2. HISTORICAL BACKGROUND

The overview of seismic damage of buried

pipelines was reported by L. Wang and M.

O'Rourke [A. 10]. Recently, a comprehensive

review on Seismic response of buried pipeline

was published by T. O'Rourke et al [ A . 1 1 ] with

more than 180 references.

In general, there are three causes of seismic

hazards to below-ground lifelines, namely: a)

soil straining induced by seismic ground

shaking, b) ground movement/rupture and c) soil

liquefaction induced by ground shaking.

The major seismic hazards have been observed to come from large ground movement/rupture along

fault or soil liquefaction zones. However,

their effects are limited to local regions that

inherit such potential hazards.

‘Department of Civil Engineering and

Engineering Mechanics, Columbia University,

New York, NY 10027, USA

^Department of Civil Engineering, Old

Dominion University, Norfolk, VA 23508,

USA

90
The effects of wave propagation from seismic

ground shaking to the damages of buried

straight pipelines located within uniform firm

soil have been found to be relatively minor as reported by Isoyama and Katayama [A. 12], The

affected area is rather large, however.

Without ground failure the damage of buried

pipelines occurred mostly at regions where the

soil and/or geological conditions change and at

joints and junctions [A. 13]. Since seismic

shaking affects a large area, the design of

buried pipeline to seismic shaking is

unavoidable

% In this paper, the analysis and design of

buried pipeline components for all three

conditions will be reviewed and discussed.

A. 3. EFFECT OF SEISMIC GROUND WAVES

Under the seismic shaking environment, the

response behavior of buried pipelines has been

observed to be controlled by the ground

displacement/strains , predominantly in the

axial direction by Kubo et al [A. 14] and the

dynamic effects to be negligible by Sakurai and

Takahashi [A. 15]. To aid the design of buried

lifelines, simplified and quasi-static analyses

for the response of straight pipelines, bends

and junctions to seismic ground waves are discussed. The major assumptions and

limitations are that the ground has not failed

or liquefied and the traveling wave is constant

in speed, shape, and magnitude. Further

assumptions or deviation from above conditions

will be described at the appropriate

subsections dealing with specific components.

A. 3.1 Straight Pipelines

Analytical methods for straight pipelines can

be grouped into two categories, namely:

continuous and jointed pipelines. Continuous

pipelines are generally modeled as beams, which

are connected to an elastic medium, and their

responses can be determined in closed form. On

the other hand, numerical solutions are typically required for the analysis of jointed

pipelines

A. 3. 1.1 Continuous Pipelines

The simplest method to analyze continuous

pipelines is based on the assumption that

buried pipelines are perfectly bonded to the

surrounding soil [A. 15, A. 16]. In other words,

it assumes no relative motion between the pipe

and the ground. Thus, as upper bounds, one can

take the seismic ground strains as the pipe

strains and the seismic ground curvature as the

pipe curvatures. This is equivalent to

assuming that the pipe has no stiffness and

therefore, will follow the ground exactly.

For the analysis and design of continuous

pipelines, the upper bound of the axial strain

of the pipe, e p, will be the maximum ground

strain, e g, due to the earthquake:

e p

e g =-^vraax/Cp (A.l)

The upper bound for the maximum curvature of the

pipeline, X p, will be the maximum ground

curvature, x g:

X p

= Xg = amax/Cs (A. 2)

where Vmax is the maximum ground velocity and

amax the maximum ground acceleration during

a seismic event at the site; Cp and C s are

the longitudinal (compressive) and transverse

(shear) wave propagation velocities,

respectively, of the controlling environments

with respect to the pipeline. If a pipeline

with diameter D and the soil move together, the

maximum pipeline bending strain e ^ is

e b= + amax ®/2 cj (A. 3)

If a continuous piping system can meet above

sets of upper bound criteria (strain and

curvature) , the pipeline will be adequate

against earthquakes that produce ground

velocities and accelerations less than the

Vmax and amax an environment that has

ground wave speed greater than the Cp and C s used in the analysis.

For important projects such as nuclear power

plants, the effective apparent wave propagation

velocities must be investigated carefully.

However, for preliminary design, the wave propagation velocities, C s and Cp, at the

site may be determined by:

where G is the soil shear modulus and P the soil mass density.

Note that the effective wave propagation

velocity is affected by the characteristics of

soil of deeper layers, one should not use the

shear modulus of the soil just near or surrounding the pipe. Engineering judgement

must be exercised.

The American Society of Civil Engineers's

Committee on Seismic Analysis [A. 17] recommends

an apparent wave propagation velocity between

2000 and 3000 ft/sec (610 to 915 m/sec) for a conservative estimate of pipeline strain.

Measurements of apparent propagation velocities

by Japanese researchers [A. 18] indicate that the

recommended values are indeed conservative and

that a velocity of 3000 ft/sec (915 m/sec) would

be appropriate as a lower bound estimate for

most conditions.

One must note that Eq . (A.l) is based on the

assumption that pipelines are rigidly bonded to

the soil. Under actual conditions, ground wave displacements are transmitted to the pipeline by

91
means of shear.. In the limit, these

displacements are constrained by the amount of

shear deformation and resistance mobilized at

the soil-pipeline interface.

The effects of relative displacement between

soil and pipeline can be accounted for by the

concept of frictional slippage length [A. 19].

If it is assumed that the ground strain e g is

constant, slip will occur near the ends of the

pipeline to mobilize a maximum frictional force

per unit pipe length t u. This condition is

illustrated in Fig. 1. The maximum slippage

length Lf, is:

L f = 2 e g EA/t u

(A. 5)

in which A is the cross-sectional area of

pipeline and E is the modulus of the pipeline

material

.

The maximum frictional force per unit pipe

length t u can be obtained from

t = it D (1 + K 0

) y H tan 6 /2 (A. 6)

in which P is the unit weight of soil, H is the

depth of the pipe, KQ is the coefficient of

earth pressure at rest, and <5 is the

interface friction angle between the soil and

pipe.

If the pipeline is shorter than twice the

maximum slippage length, the maximum pipe

strain e p would be smaller than £ g # when

seismic waves are considered, it can easily be

shown that maximum £p occurs for harmonic

wave lengths greater than one-half the maximum

slippage length. The maximum strain in

pipelines for full slip where £ p

e g

is

therefore

£p £ x t u / 8EA (A. 7)

where i is the wave length.

Other methods of analysis account, in addition

to peak ground displacement and/or velocity,

for the shape of seismic ground waves [A. 20].

The methods involve solutions of differential

equations of equilibrium for specified

locations of the seismic waves and can account

for effects of slippage between soil and pipe

[A. 20]. It makes use of a conversion factor

3 o to estimate the pipe strain by

multiplying the free field strain £g,

The study [A. 20] showed that for most cases,

3 o is almost 1 (0.98 - 0.99) for straight

pipes.

A. 3. 1.2 Segmented Pipelines

For segmented pipelines, the maximum relative

joint displacements and maximum joint rotations

become important design parameters in addition

to the pipe strains and curvatures.

The methods for the analysis of pipelines with

joints are more general than those in the

previous subsection since continuous pipes can

be viewed as segmented pipes with infinitely

stiff joints.

The simplest method for the analysis of jointed

pipelines is the static solution [A.21]. It provides an upper bound on the relative

displacement and rotation in joints. Figure A.

2

shows the model considered in this method. Soil

behavior is idealized by linearly elastic

springs with stiffness k and the pipe segments

are assumed to be infinitely rigid. It is also

assumed that joint forces are negligible and the

lepgth of seismic waves exceeds significantly

the dimension of the pipe segments. According

to these assumptions, the maximum relative joint

displacement Up, and rotation Qp are

Up = e gL = VmaxL /Cp (A. 9)

2

® p — Xg L — amaxU /U s (A. 10)

where L is the length of the pipe segment.

If a buried segmented piping system can meet all

four sets of upper bounds (pipe strain and

curvature; joint displacement and rotation) for

a design earthquake, the pipeline will be

conservatively safe because in the real case,

the pipe strain and relative joint displacement

will jointly take-up the imposed ground strain

and both the pipe curvature and joint rotation

will jointly take-up the imposed ground

curvature. One must note that the above

conclusions would only be true if one can accurately estimate the maximum ground velocity

and acceleration in the region and the seismic wave propagation velocities at the site.

A more realistic representation of the soilpipe-joint system is shown in Fig. A. 3. The

equations of equilibrium incorporate both

inertia and damping terms [A. 22, A. 23]. Due to

the fact that the dynamic effects to the seismic

response of buried pipeline are negligible

[A. 14], the inertia and damping terms in the

dynamic equations of motion have been dropped.

As indicated in Fig. A. 3, a pipe segment has

axial stiffness (EA/L) and a node at each end.

The joints are represented by elasto-plastic

joint springs (Fig. A. 4). The resistance forces

that develop between the soil and the pipe

segments are represented by elasto-plastic soil

resistance springs. (Fig. A. 5) Note that the

soil slippage characteristics have been taken

into account in the plastic part of the soil

spring.

92
The equations of static equilibrium, obtained

from the variation of the total strain energy in

the soil-structure interaction system [A. 24,

A. 25] are as follows:

t ^system ] {X} « [Ksoll ] {Xg}

2n X 2n 2nxl 2n x 2n 2nxl

(A. 11)

where [KS y S £ em ] and L Kl s o i 1 J axe the

structural system and soil resistance matrices,

respectively, {X} is the nodal axial

displacement vector and {X^} is the ground

displacement vector which varies with time.

The effects of pipe segment length on pipe

strain and relative joint displacement for

three pipe lengths of 10 ft. (3.05 m) , 20 ft.

(6.10 m) , and 40 ft. (12.20 m) were studied

using a 20 segment pipeline with elastic joint

springs and under a uniform soil environment.

It was found that the longer the pipe segment

is the larger the pipe strain, and also the

larger the relative joint displacement will

be. Since the pipe segment length is

proportional to the delay time from one end of

the pipe to the other end, the longer delay

time directly affects the seismic response

behavior.

The solution of pipe motion {X} shown in Eq

(A. 11) depends on the inputs of the ground

motion, {X^}. Since {X^} is a function of

time, the solution of {X} is also a function of

time

Assuming that the wave form of the traveling

seismic excitation remains constant over the

entire length of the pipeline which consists of

n-segments, the inputs of the time-space varying

ground motions are:

f 0 t-n± < 0

XGi =

\ (A. 12)

*» A maxh( t_ni) c "ni > 0

where A max is maximum ground displacement

input; h(t) is the displacement time function;

n^ is the delay time of the seismic wave

traveling from the first support to the end

node of the i

c ^ pipe segment considered as:

n. = IL. /C. (A. 13)

L

j = 1 J J

and Cj is the traveling wave propagation

velocity of soil/geological environments

surrounding the pipe segment j

The solution of the system of static governing

equations requires the input of ground

displacement at an instant of time. The

resulting pipeline nodal displacements, X^s,

are used to determine average pipe

strain, and i

c ^ relative joint

displacement, Uj. as:

\

= < x2i “ X2i-1)/L. and

u i = x2i+l - x2i (A. 14)

By comparing these parameters within the

earthquake time domain, the maximum values of

average pipe strains, c p; relative joint

displacement. Up and their corresponding

occurrence time and location are determined.

Computer solutions and examples can be found in

Refs. A. 25 and A. 26. Several general

conclusions can be made through the parametric

study

The effects of geotechnical interfaces (change

of soil stiffness) and geological interfaces

(change wave propagation speeds) were studied.

The maximum pipe strain was found to occur near

the interface and always in the region with

higher soil stiffness. For traveling speed

case, the maximum strain was found to occur again near the interface, but in the region

with lower traveling speed.

The effects of a number of other parameters on the response of buried pipelines, such as pipe

size (diameter), pipe materials, various

combinations of soil and joint plasticity, wave propagation velocity, wave forms, duration and

magnitude, have been investigated and reported

[A. 25, A. 26], In general, buried pipelines with

longer segment lengths, in regions with lower wave propagation speed, or surrounded by stiff

soil (high spring constant) will produce large

pipe strains and relative joint displacements.

However, soft soil which may slip and soft

joint spring which may yield, will reduce both

pipe strains and relative joint displacements.

A. 3. 2 Bends and Junctions

The methods available for the seismic analysis

of bends and tees deal generally with

continuous pipelines. In most cases, the

seismic input is characterized by postulated

ground displacement waves [A. 19, A. 20, A. 27],

It is assumed that the pipelines are composed

of elastic linear materials and that soils can

be represented by rigd-plastic or elasto-plastic models.

A. 3 .2 . 1 Bends

Figure A. 6 shows the deformation of a bend when

the direction of seismic wave propagation

coincides with the pipeline referred to as element 1. The deformation of element l,A g , has two components as shown below

A e = SiLf/EA + t uL f 2 /2EA (A. 15)

where Lf is the effective slippage length,

which can be obtained by the following

equation:

93
4EA8

f 3k

3 eg k

2t J

-1 ) (A. 16)

in which 8 = V (k/ 4EI) , I is the moment of

inertia of the pipe, k = Dks , and ks is the

modulus of subgrade reaction. The force

at the junction can be approximated by

Sj. = e g

EA - t uL f (A. 17)

in which e gEA is the axial force induced in

element 1 when there is no slippage and the

soil strain is assumed to be £g and t uLf

is the axial force induced in the same element

over Lf when slippage occurs.

The soil displacement A

s can be approximated

conservatively by

priate model should be based on curved thin

wall tubes [A. 4, A. 28]. This type of

representation accounts for bend flexibility

and results in a substantially lower bending

moment in the pipeline. Details can be found

in Ref. [A. 4, A. 28]

.

A. 4. RESPONSE OF BURIED PIPELINE TO GROUND

MOVEMENTS

The analysis of pipeline response to ground

movements requires input of soil displacement

patterns and soil-pipeline interaction

resistant characteristics. The pipeline model

may be based on continuous beam, jointed beam,

shell, or combined beam and shell

representations. Because soil and ductile pipe

properties are governed by nonlinear

stress-strain relationships, the analysis may

require complex numerical procedures. The

following discusses several analytical studies.

A

s = z g^f or u ( */4) (A. 18)

where u is the free field displacement function

at quarter wave length.

The force S]^ has been related in Ref. [A. 20]

to the relative displacement between the soil

and element 1 as:

A. 4.1 Continuous Pipeline Response to Fault

Movements

Analytical treatments of continuous pipeline

deformation from fault movement have been

developed by several researchers [A. 29 to

A. 34] . This paper briefly discusses some of

their developments.

S X = (3k/ 4 8 ) ( A e - A

g

) (A. 19)

The force S 2 and the bending moment M can be

determined by elementary structural analysis

as

:

S 2 = Si/3 (A. 20)

M = Si/3 8~ (A. 21)

3.2.2 Tees

Tees can be analyzed by similar approaches.

Figure A. 7 shows the deformation of a tee when

the seismic wave travels along the element 1 without further explanation. The Si force

and moment are given below:

Si = ( e g

EA - t u

L f )/2 (A. 22)

in which

L = ( I 1+

4e

g k

-1) (A. 23)

f 2k t u

8

and

M = Si/2 J (A. 24)

The previous methods of analysis are based on

the assumption that the bend angle does not

change during the seismic excitation. This

assumption is not realistic for bends involving

thin wall pipelines, for which a more approA.4.1.1. Newmark and Hall Model [A. 29, A. 30]

Fig. A. 8 shows a plan view of a pipeline which

is intersected by a right lateral strike-slip

fault at an angle, B . The pipeline is

oriented so that fault displacement A

£ will

cause tension in the buried pipeline. It is

assumed that sufficient geologic evidence is

available to determine such orientation.

Newmark and Hall [A. 30] analyzed the pipeline

deformation as an antisymmetric pattern of

circular arcs, with each circular segment

spanning the fault centerline and the location

of an anchor point. The distance between the

fault and the anchor point is known as the

anchor length, La . Anchors may be caused by

bends, tie-ins, and other features which

develop substantial resistance to axial

movement. Alternatively, the anchor point may

represent an effective anchor length, beyond

which there is no axial stress imposed in the

pipeline from fault movement. The pipeline is

buried at a depth to centerline H in soil with

unit weight Y . The coefficient of at-rest

horizontal soil pressure is KQ , and <5 is

the angle of frictional resistance at the soilpipeline interface. Using a numerical

procedure, relative sensitivity of pipeline

response to key parameters were studied. The

most important parameters affecting the

pipeline's capacity for fault displacement are

the angle of crossing 8 , and the effective

anchored length La

. Increasing 8 from 50

to 80 degrees increases the capacity for fault

displacement by more than a factor of four.

94
The most favorable angle of crossing is

approximate 90°. Increasing of burial depth

from 3 to 10 ft (0.9 to 3.3 m

) will decrease

the maximum tolerable movement by a factor of

four because lower soil resistance from

shallower buried depth will allow larger fault

movement. Fault displacement capacity increases

in direct proportion to wall thickness.

A. 4. 1.2 Kennedy, Chow, and Williamson Model

[A. 31 , A.32

The analytical method developed by Kennedy,

Chow, and Williamson [A. 31, A. 32] represents an improvement in model sophistication principally

because it accounts for increased frictional

resistance near the fault crossing and for

pipeline bending strains. It is assumed that

greater frictional resistance is developed along

the curved portion of the pipeline than along

adjoining sections because of high horizontal

soil pressures resulting from large relative

movements between the pipeline and surrounding

ground

The Kennedy, Chow, and Williamson model gives a more conservative result than the Newmark and

Hall model, chiefly because of the increased

level of frictional resistance along the curved

portion of the pipeline. Because of large

deflection is involved, passive soil pressure

has been taken into account. Over a wide range

of B the Kennedy, Chow, and Williamson model

predicts about half the fault movement capacity

as the Newmark and Hall model.

Note that in Kennedy et al . and Newmark and Hall

models, only axial tensile force at the point of

inflection was used for equilibrium and only

axial strength of pipe was used for failure

criterion. No flexural resistance was considered

A. 4. 1.3 Wang and Yeh Model [A. 33]

Note that the omission of flexural rigidity of

pipe using Newmark and Hall's model [A. 29, A. 30]

and Kennedy et al's model [A. 31, A. 32] cannot

satisfy the equilibrium condition for a pipeline

across a "reverse" strike-slip fault that will

cause compression in the pipeline. Another

assumption in these papers was that the far ends

(away from the point of inflection) of the

pipeline approach tangentially to the undeformed

lines. Actually, the response behavior of

buried pipeline at far ends to fault movement

should be similar to a beam on an elastic

foundation.

To improve the existing results and to

generalize the formulations for buried pipelines

across any strike-slip faults that may cause either tension or compression failure of the

pipeline, Wang and Yeh Model [A. 33] included

the bending rigidity of pipe, a shear force at

the point of inflection of the curved pipe

crossing the fault zone and a boundary

condition related to a semi-infinite beam on an elastic foundation at some distance away from

the fault zone, as shown in Figure A. 9. The

effects of axial force and bending moment

interaction and large deformation will also be

included in the analysis. Thus, the problem is

two-fold non-linear; geometry non-linearity and

material non-linearity. Note that without

buckling consideration, the formulations

outlined below for buried pipeline across strike-slip faults, can also be applicable to reverse strike-slip with little modifications.

In the paper, the design criteria for buried

pipeline proposed were (a) the axial stress must

be smaller than the yeild strength, of the

material and (b) the resulting moment must be

smaller than the resisting moment capacity of

the pipe section. The influence of the axial

forces on the resisting bending moment of a steel pipe can be found in Ref. [A. 33].

A series of parametric analyses has been

undertaken to study the influence of the fault

movement effects on the design of buried

pipelines. The general conclusion was that the

resisting capacity of buried pipeline to fault

movement would be larger for larger crossing

angle, smaller buried depth and smaller diameter

pipe. However, the study found that the results

from earlier studies that were based only on axial stress criterion were unconservative.

A. 4. 2 Jointed Pipeline Re sponse to Fault

Movement

A jointed pipeline can fail by tension in three

ways: (a) excessive tensile and bending

deformation of the pipe, (b) excessive rotation

at a joint, or (c) pullout at a joint. A

simplified analytical method for evaluating

jointed pipeline response to abrupt soil

movement has been developed by O'Rourke and

Trautmann [A. 34]. As shown in Fig.A.10, the

maximum tolerable strike-slip fault displacement

df can be evaluated as being composed of a critical transverse component d c and a critical longitudinal component S^. The value

of Sl Is given by the displacement pullout

capacity of the joint. For many water pipeline

with rubber gasket joints, Sl is simply the

depth of spigot penetration in the bell end of

the adjoining pipe [A. 35]. Even when special

fixtures are provided to restrain pullout, the

axial resistance of the fixtures is generally

low compared to the forces developed by

permanent soil displacement. Accordingly, Sl can be taken as a conservative limit.

As seen in Fig. A. 10, the critical fault

movement is defined as:

tan Sf = d c /SL (A. 25)

in which 6 ^ defines the angle of pipelinefault intersection for which failure may occur

by either the transverse or longitudinal

components of fault movement.

95
This value represents the optimal orientation

of the pipeline because the fault movement will

mobilize the maximum capacities for pullout,

rotation, and pipe tension at this angle. For

angles larger than the optimal orientation,

failure by pipe bending or joint rotation will

occur at smaller fault displacements. For

angles less than this optimal orientation,

failure by joint pullout will occur.

The tolerable fault displacement can be plotted

as a function of 6 f. Fig. A. 11 shows df as

a function of 6 f for various types of

jointed pipeline [A. 34]. The cross-hatched

area in the figure refers to a zone in which

cast iron pipeline failure will depend on the

location of lead joints relative to the fault

centerline. Mechanical joints are shown to

fail by pullout at decreasing levels of

displacement for 6 f < 70 deg. Extra long,

restrained couplings are shown to accommodate

potentially as much as 13 in. (330 mm) of

abrupt soil movement.

A. 4. 3 Pipeline Response to Compressive Ground

Movement

In previous sections, most studies assumed that

the pipeline was oriented to develop elongation

in response to ground movements. There are, of

course, many instances in which the

orientations of potential ground ruptures

cannot be determined accurately, and the

pipeline may therefore be subjected to

compression

.

Two modes of deformation can occur when a pipeline is severely compressed. The pipeline

may buckle as a beam or it may deform by local

warping and wrinkling of its wall.

An analytical investigation of shell buckling

[A. 36] shows that buried pipelines buckle

dynamically under an axial compressive stress

and strain very similar to the static buckling

stress and strain. The same study demonstrates

that the soil stiffness and internal pipe

pressure can substantially increase the axial

buckling capacity of the pipe.

An experimental study [A. 37] indicates that,

for a ratio of pipe diameter to wall thickness

of 100, local wrinkling occurs at compressive

strains between 0.004 and 0.006. Compressive

fault displacement will cause strain to

accumulate locally at the point of initial

wrinkling, thereby leading to a severe concentration of strain at movements

considerably less than those causing large

tensile strains at fault crossings.

Analytical work has been performed on the

problem of beam buckling of pipelines [A. 38].

The analyses pertain to conditions in which the

tranverse vertical restraint on the pipeline is

assumed to be constant and the pipeline is

modeled as a linear elastic beam.

Solutions have been presented for two general

conditions, pertaining to -a pipeline resting on

an elastic and rigid foundation.

A. 5. PIPELINE RESPONSE IN A SOIL LIQUEFACTION

ENVIRONMENT

The performance of pipeline buried in a soil

liquefaction zone has not been studied until

recently. Preliminary experimental studies

have been initiated by Kuribayashi et al.

[A. 39, A. 40] and analytical approaches by Wang

and his colleagues [A. 41, A. 42]. The following

discusses briefly these analytical approaches.

A. 5.1 Continuous Pipelines

In the liquefiable zone, the seismic waves propagating faster through firm soils along the

pipe axis cause an incoherent movement at both

edges of the pipe submerged in liquefied

mediums. Therefore, a pipeline buried under

the liquefied area is subjected to seismic

incoherent motions which generate longitudinal

and lateral forces on the pipeline. Breaking

and buckling are two common pipeline failure

modes

.

The surrounding soil mediums of the pipe within

the liquefiable zone during the earthquake

losses shear strength due to seismic waves and

the stiffness of the soil foundation varies

(decreases) with time within the liquefiable

region. The pipe could be considered partially

supported until the soil is completely

liquefied. This paper proposed a simplified

beam-column (buried pipe) model on two time

verying strength elastic foundations, one for

liquefiable and the other for non-liquef iable zones as shown in Fig. A. 12. A finite

difference technique involved both temporal and

spatial variables is employed to solve the

dynamic responses of buried pipelines passing

through a soil liquifiable zone which is

subjected to an incoherent ground shaking.

For evaluating the dynamic effect of

longitudinal incoherent displacement (or strain

rate) several moving (loading) speeds applied

at the movable end of the pipe are chosen to

evaluate their influences on the lateral

deflection and axial load of the pipe. Details

can be found in Ref. [A. 41] and thus will not be

repeated

.

The results of load response histories showed

that the dynamic displacements and axial load

of the pipe increase with an increase in the

moving speed (strain rate) and thus increase

the possibility of dynamic buckling. Since the

boundary conditions and the moving speeds do

not represent exactly the field conditions, the

results can be considered only qualitatively.

More studies would be needed to clarify the

actual behavior of pipeline passing through

soil liquefaction zone.

96
A. 5. 2 Pipeline With A Manhole

An experimental research on the responses of

buried pipeline including a manhole (Fig* A.13) was carried in saturated sand by Kuribayashi et

al. [A. AO] . However, no analytical solution

has been reported for comparisons . The exact

response of a buried pipeline system including

manholes in a soil liquefaction environment

subjected to seismic shakings is complex and

the complete solution for such complex system

has not been founds

In a paper, Wang et al. [A. 42] proposed to

examine analytically both the horizontal and

vertical responses of the buried pipeline

experiment including a manhole under a soil

liquefaction from a realistic but still simple

mathematical model. For vertical responses

only the static analysis will be applied since

the buoyancy force behaves in a static manner.

However, for the horizontal responses, a non-linear dynamic analysis including time

varying soil spring and damping is proposed.

Case studies are carried out to evaluate the

response behavior of buried pipelines with and

without a manhole. The analytical results are comapred to the experimental results obtained

by Kuribayashi et al. [A. 40]

Analytical procedure involves critical

determination of various influential parameters

in the model including: (a) dynamic horizontal

water and soil pressures, (b) subgrade reaction

forces, and (c) damping characteristics at the

structure and soil interface. Unfortunately,

the exact values of these parameters are not

completely known and assumed values have to be

used in the analytical study.

The analytical responses are in general higher

than the experimental observation. The

discrepancy may be attributed to the sudden

change in the parabolic loading case, the

approximation of the Rayleigh-Ritz shape

function, the under estimation of the damping

ratios and finally the omission of the

soil-structure-f luid interaction.

A. 6 DESIGN CONSIDERATIONS

In the United States, there is no seismic code

for buried pipeline design except some guidelines [A. 3, A. 4]. Some aspects of the

lifeline design considerations have been

discussed by Wang [A. 43].

In the absence of seismic design codes for

buried pipelines, several passive design

considerations have been used by engineers to

reduce seismic damage and minimize hazardous

effects. Following are some common engineering

practices and recommendations:

1

Redundancy should be built into the

distribution system. More smaller pipes

should be used in lieu of a single larger

pipe to minimize reduction in operation due

to breakage of pipes.

2. Blow-off valves should be installed at a location where higher seismic activity is

anticipated, such as along a .fault line.

By this technique, water is led to a nearby

reservoir after a blow-off valve fails

during an earthquake.

3. Ductile pipe materials such as steel or ductile iron should be used to allow larger

pipeline deformations.

4. For segmented pipelines, flexible joints

such as rubber gasketed connections should

be used to provide for relative joint

movements. For anticipated large ground

movement, extra long restraining sleeves or 'Bellow Joints' should be used. When

feasible, shorter segments which will

experience less strain imposed by the

ground motion, should be used. Also,

relative joint displacements are less for

shorter segments.

5. If feasible, consideration should be given

to encasing the pipeline in a larger tunnel

in order to isolate the pipeline from the

seismic ground motion, or to lubricating

the pipeline in order to increase the

"slippage" between the pipe and the

surrounding soil.

A. 7. SUMMARY AND CONCLUSION

Part A has summarized the recent developments

in component analysis and design of buried

lifelines subjected to ground strains or liquefaction due to seismic shaking and ground

movements

To aid in the design of buried pipelines for

earthquake loads, various analysis procedures

and design considerations for various

environments have been reviewed and discussed.

In conclusion, it is important to note that the

behavior of buried pipelines is governed by the

relative displacements of the ground along the

route. Ductility or flexibility to allow

buried pipeline movement with the ground is the

most important factor for the seismic design of

such structures.

PART B: SYSTEMS ASPECTS

B. l . INTRODUCTION

Analytical models appropriate for the network

analysis of telecommunication transmission

facilities network systems must be developed.

Combining these models with those for "exchange

office buildings," representative models for

the total telecommunication network systems

should then be constructed. The utilities

equipment may have to be integrated as part of

97
the total network system. Methods should then

be developed to estimate the vulnerability (or

equivalently survivability) and the serviceability of such networks. Here, the term

"exchange office buldings" is used for

simplicity to represent station equipment,

switching facilities and related structures

including the building which houses them.

The seismic vulnerability indicated the

susceptibility of the network to a seismic

event (or a sequence of seismic events) and is

to be measured in terms of the probability;

either the probability of vulnerability or probability of survivability. In estimating

such a probability, not only the susceptibility

of the transmission facilities network but also

those of the exchange office buildings must be

taken into consideration. It appears

reasonable to model the total network system as

a linear graph consisting of vertices and

branches in which the vertices represent the

exchange office buildings and the branches

represent the transmission facilities,

typically above- and under-ground transmission

cables. Thus, a seismic event could result in

the physical damage of the network system

equivalent to the removal of the subset (or

possibly the entire set) of vertices and

branches from the original network. We could

postulate that the network has survived an earthquake if all possible pairs of stations

retain a minimum level of communication

capability. In fact, survivability may be

defined in a number of ways. On the one hand,

the survival of a network, for example, may

require more than one independent path between

all the pairs of vertices of the network. On

the other hand, in the case of those emergency

situations arising from destructive earthquake,

we may have to be satisfied (at least

immediately after an earthquake), with the

following forms of partial connectivity as the

definition for network survival: (1)

connectivity only among key pairs of stations

and (2) a certain percentage of vertex pairs

remaining communicable. In developing such

criteria, we must have a good understanding of

the extent to which and the way in which the

communication network system is needed

immediately after an earthquake for medical

assistance, civil order, fire fighting, and

other purposes.

The possible removal of the vertices mentioned

above results from damage to the exchange

office buildings caused by excessive structural

vibrations or structural failures of the

buildings housing them due to an earthquake.

Similar situations arise when underground water

transmission network systems are analyzed for

their seismic vulnerability in which, for

example, possible failures of pumping stations

(these are vertices in that case) need to be

taken into consideration. In dealing with

telecommunication systems, however, station

equipment, switching facilities and the

buildings housing them, which as a set are represented by a vertex in the associated

linear graph may in turn be modelled as a sub-network. The telecommunication network

system, with such sub-networks replacing the

vertices, appears to be a more consistent model

of the total system and may have to be

developed for ( 1 ). determining the degree of

survivability (or vulnerability) of an existing

or proposed system and (2) deriving the optimum

designs or improvement strategies on the basis

of maximum survivability or minimum

vulnerability

.

Turning now to the question of serviceability

immediately following an earthquake, the

quality of service is usually measured in terms

of the delay and/or blocking which are obviously dependent upon the post-earthquake

states of physical damage and hence

functionality of the total network system.

Those serviceability criteria appropriate for

post-earthquake emergency situations must be

developed and evaluated in terms of the

probability that the total network system will

be able to meet a specified service criterion

following an earthquake. Furthermore, study

must be made to establish general strategies

for restoring the level of service specified by

such a criterion if the total network system is

damaged to the extent that it cannot satisfy

the specified service. The relationship

between vulnerability and serviceability is

also an interesting subject for future study.

B .2 FAILURE PROBABILITIES OF TRANSMISSION

FACILITIES

The first step towards evaluating the

survivability and serviceability of a total

network system is to assess the probability of

failure of various specific items that are classified in the categories of station

equipment, switching facilities, building

structures and transmission facilities. The

transmission facilities, and in particular, the

transmission cables and conduits are mainly

considered here. The probabilities of failure

of those items belonging to the categories of

station equipment, switching facilities,

building structures and utilities equipment may

be evaluated within the context of traditional

earthquake engineering which usually deals with

building structures and secondary systems

attached to it.

Six principal types of transmission media are used for providing transmission channels: (1)

open-wire lines, (2) paired cables, (3) coaxial

cables, (4) radio, (5) wave guides, (6) optical

fibers. For existing urban telecommunication

network systems, only media (2) and (3) appear

to be important for seismic analyses, at least

at this time.

In most cases, underground cables are placed

inside conduits and then buried underground.

98
The conduit system provides physical protection

for the cables and also allows new cables to be

added inexpensively, although at the expense of an initial cost*

Several potential failure modes of cables

(although some of them occur simultaneously)

are: (1) rupture of wire conductors, which are

the high purity copper, (2) crack, distortion

or rupture of insulated materials, whcih are made of wood, pulp, or plastic, (3) rupture of

cable sheaths, which are plastic, steel,

aluminum, or lead, and (4) rupture of conduit

systems which in turn may lead to the cracking,

distortion, or rupture of the cables.

Failure mode (1) completely disrupts

communication while mode (2) will possibly

cause electrical interference (cross-talk)

between cable pairs. Modes (3) and (4) will

primarily cause water to enter, constituting

the principal hazard to which underground

cables are subjected since any significant

amount of water into the core of pulp-insulated

cables will render most of the circuits

inoperative. In this regard, a damage matrix

displaying physical failure (damage) modes and

associated resisting capacities in matrix form

should be developed for transmission cables

with and without conduits. This is a task

similar to the one performed by Kratky et al.-

[ B . 1 ] with respect to the pipes and joints used

for water transmission pipelines. Indeed, it

is primarily for the construction of these

damage matrices that the studies performed in

Part A are useful. However, the task of

developing such damage matrices for

transmission cables appears to be more demanding in terms of the effort to be

expended, since little experimental work has so far been systematically performed on these

cables and conduits.

After damage matrices are constructed, we must

develop damage probability matrices which will

indicate the probabilities of a certain type of

cables to sustain different levels of damage

under various earthquake intensities measured

usually in terms of the Modified Mercalli

Intensity (MMI) scale. The notion of a damage

probability matrix is initially used in [B.2]

to describe the general correlation between

building damage and earthquake intensity,

mainly on the basis of empirical data,

supplemented by a dynamic analysis. Figure B.l

illustrates the basic form in which the damage

probability matrix is used in [B.2] for a cost-benefit analysis of building structures.

A matrix of this form is constructed for each

of the alternative design strategies

representing the choice of a set of design

parameters such as the lateral force

requirement associated with a particular

seismic zone, materials to be used, structural

types to be designed, etc. Furthermore, such a matrix is to be developed on the basis of

theoretical analyses and subjective evaluations, as well as the damage statistics

collected from past earthquakes. The damage

state describes the degree of damage in a qualitative manner with a central damage ratio

indicating the (perhaps highly subjective)

average ratio of the repair cost associated

with each damage state to the cost of

building. The numbers in each column indicate

the probabilities with which different states

of damage will be inflicted upon the building

(designed under a particular strategy) by an earthquake of the particular MMI value

associated with that column. The reasons why a certain earthquake intensity will not

necessarily result in a certain state of damage

are, among others: (1) individual buildings, even if they are designed and constructed to

meet the same requirements, will have different

resistances depending on the skill and

idiosyncrasies of the designer and constructor

and (2) the individual ground motions, even if

they are categorized in the same MMI scale,

will obviously be different in detail.

For telecommunications networks involving

networks of transmission cables which extend over a wide underground area, a damage

probability matrix reflects the uncertainties

(a) in the ground conditions, (b) in the way in

which the cables are loaded under seismic

conditions, and (c) in the resisting capacity

of the cables (or in the failure matrices), in

addition to the obvious uncertainties

surrounding the details of the earthquake

ground motion classified in any particular MMI

scale. Therefore, such damage probability

matrices will generally be developed (a) under

each set of ground conditions, (b) for each

type of seismic events in which the cables are loaded differently and (c) for each component

of the transmission facilities (e.g., coaxial

cable with conduit). In previous work [B.3]

dealing with water transmission pipelines, the

following five types of seismic events were considered the major causes of the physical

damage to pipelines: (1) Ground motion

associated with the wave propagation; (2)

fault movement at the location of a pipe

crossing; (3) liquefaction; (4) landslide;

and (5) interference with other under- and

above-ground structures.

While the Japanese data illustrated in Fig. B.2

indicate that most of the severe physical

damage such as severing of cables appears to

have resulted from causes (1) and (4) above,

those area in the US for which study is

intended to be performed, are often surrounded

and penetrated by active faults and at the same time possibly include a large number of

liquefaction-sensitive locations of

significant size. In addition, in

telecommunications systems, cables placed

underground lead above-ground to central

offices, customers' telephone units, etc. much

more often than underground water transmission

pipelines. Therefore, interference of the

underground cables with other under-and

99
particularly above-ground structures appears to

be a serious problem which needs to be

investigated. The severing of the coaxial

cables near Sendai, Japan due to the relative

displacement between the bridge abutment and the

ground observed during the 1978 MiyagiKen-Oki earthquake, can be considered as typically resulting from such an interference.

The example of a damage probability matrix shown

in Fig. B.3 illustrates the possible damage that can occur to an arc welded steel pipe due to the

ground motion associated with seismic wave propagation and is used for the analysis

performed in [B.3] and [B.4], However,

this matrix can be used only for the case where

the damage at the corner of T- and L-shaped pipe

joints buried in soft soil is considered. When

the locations of possible damage are relatively

well defined, either due to the nature of the

seismic events (e.g., events (2)-(5) above) or

to the stress or strain concentration within the

pipes (e.g., at the corner of T-shaped pipe

joints as mentioned earlier), a damage

probability matrix established for that location

along the pipes (or along the cables for

telecommunication cables) can be used to

estimate the probability of different levels of

damage being sustained by the pipes at that

location. In Fig. B.3, "A" indicates soft soil

conditions and " B " represents a strain

concentration factor to multiply the free-field

ground strain to obtain the pipe strain. In

fact, an underground water transmission system

as shown in Fig. B.4 is considered in [B.3, B.4,

B.5]. The system is similar to that of the City

of Los Angeles, and the area in which it is

embedded is divided into a number of square

zones. With the aid of a geological map each

square zone is assigned a designation: A (soft

soil), B (medium soil), C (hard soil) and F

(zone where an identifiable fault exists).

Actually, an L designation is considered for zones susceptible to liquefaction, but this

designation was not assigned to any square zone

due to the difficulty of doing so merely on the

basis of the geological map used. In this

connection, damage probability matrices similar

to that given in Fig. B.3, were developed in

[B.3] for other seismic events. For example, a damage probability matrix is developed for the

segment of pipe that would be embedded in an L zone considering the following: larger scale

liquefaction resulting in a severer state of

pipe strain will be induced by a more intense

earthquake in more liquefaction susceptible

soils

.

However, in the case of a straight pipe

subjected to the ground motion associated with

seismic wave propagation, questions such as where along the pipe the damage might occur or how many points per unit length along the pipe

are to be considered as potential damage

locations, are by no means easy to answer. To

resolve these questions, in [B.3] and [B.4], a vulnerability analysis of underground water

transmission lifelines was performed under the

assumption that the potential damage locations

are located along the pipe at an interval of

one-half of the seismic wave length. When we have segmented pipes, however, the joints may

be the most vulnerable locations.

Following procedures similar to those described

above, damage probability matrices for

transmission facilities must be developed, and

under similar assumptions, they should be used

for a vulnerability analysis. In developing

the damage probability matrices, the

degradation of the structural integrity of the

cables and conduits due to corrosion and other

aging and wear-out effects must also be taken

into consideration.

B.3 ESTIMATION OF SYSTEM SURVIVABILITY

Once the criteria of system survivability (or

vulnerability) have been established, the

probability of system survivability (or

vulnerability) can be obtained by a variety of

approaches using damage probability matrices of

the system components. Therefore, the damage

probability matrices for station equipment,

switching facilities, building structures and

utilities equipment must also be developed.

In most communication network systems involving

electrical and electronic components with their

functions defined either on an 'on' or 'off'

basis, the methodology of vulnerability

analysis is relatively well established. For

example, the so-called "naive approach" [B.6]

advocates the use of straight forward Monte

Carlo simulation techniques in which the success (on) or failure (off) of each component

is simulated based on specified probabilities

of success and failure, and the resulting

network is checked for its connectivity

requirements. Repeated trials will then

estimate the survivability in terms of

empirical probability. The methods that

combine analytical and Monte Carlo techniques,

such as the functional simulation approach

[B.7] and the Moore Shannon approach [B.8] may

also be used effectively. The latter method is

typically a semi-Monte Carlo approach in the sense that the nubmer of disconnected

sub-networks with, say,, exactly k elements

within the total network, is simulated by Monte

Carlo methods and used in an analytical

expression to obtain the probability of network

connectivity

.

For a seismic analysis, however, the

telecommunications systems under consideration

are subjected to random seismic environments,

variable soil conditions, statistical component

strength variations, etc., thus resulting in

various levels of component damage. Indeed, to

make these otherwise unmanageable complications

analytically tractable, we have introduced

damage probability matrices for each

combination of significant parameters. It is

100
noted here that the recent trend of using

fragility curves with engineering confidence

statements is an interesting alternative to the use of the damage probability matrix approach.

The vulnerability of the total network system

can be estimated by an analytical method

similar to that employed in [B.4] (dealing with

water transmission lifeline systems) in which

the probability that connectivity between a specified vertex and any of the pumping

stations will remain intact was defined as the

survivability. However, in view of the more complicated connectivity requirements expected

for total telecommunications network systems

and in the face of the anticipated difficulty

of identifying the combined effects of various

levels of damage occurring in a number of

components in the system, the analytical method

developed in [B.4] will require substantial

modifications with a possible need for the

partial use of Monte Carlo techniques.

B.4 ESTIMATION OF SYSTEM SERVICEABILITY

Every large telecommunications network is based

upon the principle of commonly shared

equipment. It is this sharing which causes the

phenomenon of delay and blocking. To estimate

the serviceability of a system, therefore, we

must first investigate the average and peak

traffic loads under normal operating

conditions. Such traffic loading patterns

under normal conditions obviously do not apply

to post-earthquake emergency situations. In

emergency situations, the minimum capacity of

the communication system must be maintained for

the basic survival requirements of the

citizens. The reduced capacity must then be

increased by means of installing mobile systems

and/or by repair and restoration of the

original system. In order to gain empirical

insight into the problem of traffic loads,

however, normal traffic loading patterns must

be known

A future study would primarily consider those

post-earthquake emergency situtations in which

the components of the station equipment,

switching facilities, building structures,

utilities equipment, and transmission

facilities, are all subject to possible

physical damage and failure. Therefore, the

serviceability of the total telecommunications

network systems must be analyzed after the

vulnerability analysis establishes the lesser

state of connectivity that the network systems

have deteriorated into, resulting from the

damage sustained by these individual

components. This task appears to be at most as demanding in terms of analytical and numerical

efforts as the serviceability analysis

performed on underground water transmission

network systems in [ B . 5 ] . In [ B . 5 ] , the

vulnerability analysis was performed

analytically to produce a sample of degraded

networks. On each of these degraded networks.

a (nonlinear) flow analysis was performed in

the Monte Carlo sense to determine the

probability of serviceability defined in terms

of remaining levels of flow rate and water

pressure at specified vertices for use in

post-earthquake fire fighting. Similar

approaches appear to be applicable to a serviceability analysis involving

telecommunications network systems.

Speaking again of the water transmission

system, we might add that, given an earthquake,

the earthquake intensity at each potentially

critical location of the pipeline is estimated

and the damage probability matrix is used

accordingly. If the area over which the

lifeline system stretches is relatively small

compared with the epicentral distance, the

earthquake intensity may be considered uniform

throughout the area. In this case, the seismic

hazard curve given as a function of the

earthquake intensity will be convolved with the

probabilities of system serviceability (or

vulnerability) conditional to various

earthquake intensity values in order to derive

an estimate of the annual probability of system

serviceability (or vulnerability). However, if

the spatial variation of the earthquake intensity

has to be accounted for, the ensuing analysis

becomes more complicated. In [B.4], for

example, the empirical attenuation law for the

peak ground acceleration (in the surface layer)

proposed by Esteva [B.9] was combined with the

relationship, also empirical, between the

earthquake intensity (MMI) value and peak

ground surface acceleration recommended by

Gutenberg and Richter [B.10]. This combination

produced an expression for the MMI value at the

site as a function of the earthquake magnitude

M and the hypocentral distance associated with

the site. The relationship developed by Kanai

[B.ll] was also used to estimate the range of

the velocity spectrum corresponding to each MMI

value, as it is needed to estimate the free

field ground strain. These earthquake

intensity values can then be used together with

the damage probability matrices to produce a sample of degraded networks through the

vulnerability analysis. The estimation of the

probability of system serviceability then

follows. Such a probability of system

serviceability, however, is conditional to a particular earthquake with magnitude M

originating from a point source or line source

or even area source. Appropriate analysis must

then be performed to derive an estimate of the

annual probability of system serviceability

from the conditional probability.

Each step of the procedures described above

requires further scrutiny and study in order to

reduce the extent of inaccuracy and uncertainty

involved. More importantly, we must recognize

that these procedures incorporate rather simple

models of the bedrock, surface layer, fault

movement, etc., and therefore may not be

accurate enough to consider all the significant

101
aspects of free-field ground motion in relation

to the structural response analysis. For

example, interaction of surface and body waves

through lateral inhomogeneities such as dipping

layers on the edges of alluvial basins may have

to be studied for its effects on the free-field

ground motion on or near the ground surface.

B.5 CONCLUSION

The preceeding observations involving

telecommunications network systems and water

transmission pipeline systems identified the

need for further information on a number of

critical factors that affect the mitigation of

seismic hazards to these systems. Such

information is needed in order to develop a more reliable database on which more credible

vulnerability and serviceability analyses can

be performed. Ultimately, the result of these

analyses is expected to provide logical

guidelines for more technical measures of

seismic hazard mitigation by establishing the

design, retrofitting and repair criteria for

hardware systems, for evaluating the adequacy

of the system's redundancy, and for mapping out

strategies for emergency plans and restoration

schemes. Even more importantly, such a result

provides valuable input to the societal impact

analysis that must be performed from a more

global point of view. Indeed, the oil pipeline

failure due to large-scale landslides during

the recent Ecuador Earthquake has inflicted a devastating loss of economy and productivity

upon that country. In fact, we have been

extremely fortunate that no major earthquakes

as large in their scale as those of the New

Madrid, Missouri Earthquakes of 1811-1812 and

the Charleston, South Carolina Earthquake of

1886, have not struck the Central United

States. Energy distribution systems including

crude oil, petroleum and natural gas pipeline

systems are concentrated in this area from the

South and Southwest to densely populated areas

of the North and Northeast [B.12], Physical

and functional failures of such energy systems

due to seismic events and the resulting

disruptions in the supply of energy could

produce disasterous consequences of

unprecedented scale and magnitude. Such

failures would probably be induced by

earthquake-triggered large-scale landslides

and/or liquefaction. It is about time that we extended the type of analytical procedures

described in this paper, applied the results of

such procedures for the purpose of mitigating

disasters and at the same time continue to

identify and work on research issues the

resolution of which will greatly alleviate the

difficulties associated with mitigation

efforts

.

ACKNOWLEDGEMENT

The authors wish to acknowledge that most of

the works discussed in the paper have been

supported by the Earthquake Hazard Mitigation

Program of the National Science Foundation over

the years. The recent work discussed in Part B was partially supported by Subcontract No.

SUNYRF-NCEER-86-3043 under the auspices of the

National Center for Earthquake Engineering

Research under NSF Grant NO. ECE-867-07591 0 Dr. Shinozuka wishes to indicate that some of

his material is taken from Ref. B.13. Dr. Wang

summarizes the results from a large pool of

references. Although too many authors and

colleagues to be named herein, one should

recognize their contributions to the

advancement of lifeline component design and

analysis as indicated in the references.

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A. 28. Goodling, E.C., Jr., "Buried Piping-An

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104, No. TCI, Nov. 1978, pp. 91-107.

A. 31. Kennedy, R.P., Chow, A.W., and

Williamson, R.A., "Fault Movement Effects on Buried Oil Pipeline," Transportation

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A. 32. Kennedy, R.P., Darrow, A.D.C., and

Short, S.A., "Seismic Design of Oil Pipeline

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1979, pp. 119-134.

A. 33. Wang, L.R.L., and Yeh, Y.H., "A Refined

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Vol. 13., No. 1., Jan-Feb. 1985, pp. 75-96.

A. 34. O'Rourke, T.D., and Trautmann, C.H.,

103
"Earthquake Ground Rupture Effects on Jointed

Pipe," Proc. of the Second Specialty Conference

on the Technical Council onLifeline Earthquake

Engineering, ASCE, Aug. 1981, pp. 65-80.

A. 35. Singhal, A.C., "Nonlinear Behavior of

Ductile Iron Pipeline Joints," Journal of

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1 American Society of Civil Engineers, May

1984, pp. 29-37.

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under Seismic Loads," Paper presented at the

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No. 80-C2/PVP-76) .

A. 37. Bouwkamp, J.G., and Stephen, R.M.,

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C.H., "Buckling of Buried Pipelines Due to

Large Ground Movements," Earthquake Behavior

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Aida, M., and Yui , Y., "An Experimental

Behavior of Buried Pipes During Liquefaction of

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Publication, PVP-98-4, June 1985, pp. 34-42.

A. 40. Kuribayashi, E., Ieda, R., Kawamura,

M„, Tsuchiyama, S., and Yamanaka, S.,

"Behavior of Buried Pipelines with Manholes

During Liquefaction (in Japanese)," Proc. of

the 21st Annual Meeting of the Japanese Society

of Soil Mechanics and Foundation Engineering,

Sapporo, Japan, 1986.

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Effects of Soil Liquefaction and Ground

Displacement to Buried Pipeline," Proceedings

of the 1985 Pressure Vessels and Piping

Conference, ASME publication, PVP, 98-4, June

1985, pp. 43-52.

A. 42. Wang, L.R.L., Shim, J.S., and Ishibashi,

I., "Response of Buried Pipelines Including

Manhole in a Soil Liquefaction Environment,"

Paper to be presented at 3rd International

Conference on Soil Dynamics and Earthquake

Engineering," Princeton University, June 22-24,

1987 .

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Buried Pipelines Under Various Seismic

Environments," International Journal of

Pipelines, Vol. 5, No. 3, October 1986, pp.

277-287.

B.l. Kratky, R.J. et al., "Suggested

Experiments on Straight Pipes in Air and in

Soil," Report No. 9 under NSF Grant No.

FR-78-15049, Weidlinger Associates, January

1979.

B.2. Whitman, R.V., et al., "Methodology and

Pilot Application," Journal of the Structural

Division, ASCE, Vol. 101, No. ST5, May 1975, pp.

1067-1084.

B.3. Shinozuka, M., Takada, S., and Ishikawa,

H., "Some Aspects of Seismic Risk Analysis of

Underground Lifeline Systems," Journal of

Pressure Vessel Technology, Vol. 101, February

1979, pp. 31-43.

B-4 . Shinozuka, M., and Koike, T., "Seismic

Risk of Underground Lifeline Systems Resulting

From Fault Movement," Proceedings of the 2nd US

National Conference on Earthquake Engineering,

Stanford, California, January 1979, pp. 663-672.

B-5. Shinozuka, M., Tan, R., and Koike, T.,

"Estimation of Serviceability of Underground

Water Transmission Network Systems," Proceedings

of the ASCE Specialty Conference on Lifeline

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August 20-21, 1981, pp. 97-110.

B-6 . Frank, H., and Chou, W., "Topological

Optimization of Comptuer Networks," Proceedings

of IEEE, Vol. 60, November 1972.

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Reliability Analysis: Part I," Network, Vol. 1,

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Circuits Using Less Reliable Relays," Journal of

the Franklin Institute, Vol. 262, 1956.

B-9 . Esteva, L., "Geology and Predictability in

the Assessment of Seismic Risk," Proceedings of

the 2nd International Conference of the

Association of Engineering Geologists, Sao

Paolo, Brazil, 1974.

B-10. Gutenberg, B., and Richter, C.F.,

Seismicity of the Earth, (NY: Hafner) , 1965.

B-ll. Kanai, K., Seismological Engineering,

(Tokyo: Kyoritsu Press), 1973.

B-12. Coordinating Committee on Energy of the

Public Affairs Council, "Vulnerability of Energy

Distribution Systems to an Earthquake in the

Eastern United States-An Overview," American

Association of Engineering Societies, December

1986.

B-13. Shinozuka, M., "Scientific and Engineering Information Needs in Abatement of

Scientific Hazards to Telecommunications

Systems," BSSC Workshop on Development of and

Action Plan for Abatement of Seismic Hazards to

Lifelines, Denver, Colorado, November 5-7, 1986.

104
(a) SHORT ELEMENT

( b) LONG ELEMENT

Fig. A. 1 Strain in Continuous Pipeline [A. 11]

Fig. A. 4 Elasto-plastic Joint Spring [A. 25]

Fig. A.2 Static Joint Rotation [A. 21]

777777777777777777777777777777777777

p

30 Gl

V G2

*

2 , *3 K H—* H

Kfl

K

X

Gi-I LGi ^

N £

2i

kt-H

I WlHn

detail —

Fig. A. 3 A Buried Segmented Piping System Model

[A. 21] Fig. A. 6 Force/Displacement at a Pipeline Bend

[A. 11]

105
detail —

A

mm in.

Anglt of Pip •lint —Paul t Intersection. (Jf(deg)

Fig c A. 7 Force/Displacement at a Tee Junction

[Aoll]

Fig. A. 11 Joint Resistance to Fault Movement

Fig. A. 9 Pipeline Response to Fault Movement — Wang and Yeh's Model [A. 33]

DISPLACEMENT GAUGES

3PRIN8

HIM8E

Fig. A. 10 Segmented Pipeline Subjected to

Strike-slip Fault Movement [A. 11]

Fig. A. 13 Experimental Study of Buried Pipeline

with a Manhole in an Soil Liquefaction

Environment [A. 40]

106
DAMAGE

STATE

CENTRAL

DAMAGE

RATIO, %

M M INTENSITY

VI VII VIII DC

NONE 0

LIGHT 0.3 p

L>SI

MODERATE 5 —

HEAVY 30

TOTAL 100

COLLAPSE. 100

Fig. B.l Form of Damage Probability Matrix

Fig, B .4 Water Distribution System and

Simplified Soil Map of Los Angeles

Fig. B.2 Damage Sustained by Inter-city Coaxial

Cables Due to the 1978 Miyagi-Ken-Oki

Earthquake

VI VII VIII DC

MINOR 1 0.96 0.18 0

MODERATE 0 0.04 0.72 0.11

MAJOR 0 0 0.21 0,89

Fig. B.3 Damage Probability Matrix for Soil

Conditions A and a =3
A Review of Base Isolation for Upgrading Earthquake

Performance of Building Structures

BY

Dr. Nishikamt R. Vaidya^

Mr. Robert J. Smith^

ABSTRACT

This paper reviews base isolation for upgrading

the earthquake performance of building

structures. Both new construction and existing

structures are considered. Base isolation is a design strategy which attempts to reduce the

seismic forces imparted to the building

structure and the components housed within the

building. This strategy appears to be of

particular significance for buildings that are required to be operational through and following

a severe seismic event. But, it is not

beneficial in all cases. A qualitative criteria

for selecting buildings which may be potential

candidates are suggested. These criteria may be

used in the preliminary stages of building

design to screen the types that may benefit from

base isolation. Several specific schemes are presented and the manner in which they act to

isolate the structure are described. The more

prominent isolation schemes use reinforced

elastomer bearings which provide isolation in

the horizontal direction only. In the more developed schemes (on the basis of use, testing,

and analysis) the displacement response is

controlled by a friction interface, a lead

insert or by special compounding of the

elastomer. Design issues revolve around the

relatively large displacements and are generally

resolved by good detailing. Provided that this

is done, it is concluded that base isolation is a sound alternative to traditional methods of

aseismic design of building structures.

1. INTRODUCTION

Structural upgrading for the purpose of

mitigating effects of earthquakes can consider

several alternative solutions. Very broadly

they can be fitted into two categories —one is

to strengthen the structures and the other is to

isolate the structure from the seismic ground

motion. Generally, the alternatives are applicable not only in the design stage for new construction but also, if properly detailed, for

existing structures. Irrespective of the

1. Project Manager, Paul C. Rizzo Associates,

Pittsburgh, Pennsylvania.

2. Chief, Structures Branch, U.S. Army Corps of

Engineers, Washington, D.C.

choice, the solution is more easily adopted for new construction than for existing structures.

Among these alternatives for seismic structural

upgrading, the traditional approach has been to

add strength and stiffness to the structural

systems. Typically, additional bracing and/or

shear walls are considered to take the

horizontal forces generated from the design

seismic event and transmit these forces to the

building foundation safely and without

significant relative displacements. The seismic

forces chosen for design reasonably assure that

the building will not collapse in the event of

the expected severe earthquake but may be

subjected to permanent damage.

In structures of historical significance, the

traditional methods may not be the most

attractive if they mar the architectural

appearance of the facade which must be preserved

to elicit at least partially the worth of the

facility. For certain mission essential

buildings, the intrinsic value dictates that in

addition to protecting life, the structure and

systems housed therein suffer as little damage

as is feasible to reasonably assure the

continued operation of the facility through and

following the earthquake. Thus, importance of

the structure site conditions, seismic

potential, and other related issues may prompt

consideration of alternate upgrading.

Deviating from the traditional approach then,

very broadly two alternative approaches have

been considered by the profession to mitigate

earthquake effects.

• Reduction of input motion through

insulation of the structure, and

• Attenuation of input motion through

isolation of the structure.

Insulation takes the form of backpacking or incorporating a trench around the facility and

far enough away from it in free-field.

Conceptually, the trench intercepts incoming

seismic waves and, thus, effects reduction in

the input ground motion in the vicinity of the

building foundation.

In general, isolation takes the form of

supporting the building on a- flexible foundation

108
such that the seismic ground motions are attenuated as they are imparted to the structure

through the isolating mechanism. The general

concept is illustrated on Figure 1. Since the

isolation mechanism is located at the base of

the structure, it is referred to as base

isolation. This paper addresses the current

methods of base isolation.

In a broad sense, most base-isolation schemes

consist of a flexible link between the ground

and the supported structure with one or more energy absorbing devices acting either in

parallel or in series with the flexible link.

Depending upon the scheme, either the flexible

link or the energy absorbing device takes on the

greater significance. Thus, in the one extreme,

attenuation of seismic motion depends primarily

on capacity to absorb the seismic energy

imparted at the structure/foundation interface;

while in the other extreme, the flexible link

alone may be designed through a frequency shift

to result in a low transmissibility of ground

motion.

This paper reviews the criteria for selecting

base isolation as a design strategy. Various

base isolation schemes are briefly summarized

and their anticipated performance described.

The information about these schemes has been

collected from published literature, technical

papers, notes, etc. It is hoped that the

information presented here will aid in selecting

an appropriate design strategy and evaluating

the base isolation systems during the conceptual

design stage of a building structure.

2. BASE ISOLATION DESIGN PHILOSOPHY

A general overview of the basic principles of

base isolation can be found in texts on earthquake engineering (e.g. Newmark and

Rosenblueth, 1971). As this would signify, the

concept is not new; its history in seismic

applications, which dates back to the early

1900's, has been very aptly described by Kelly

(Kelly, 1979). The more recent interest has

probably been motivated by several factors,

among which the more significant perhaps are,

the need for a higher level of earthquake

performance of buildings, and equally important,

the development of reinforced elastomer bearings

which, due to their inherent mechanical

properties, have extended the range of

applicability of base isolation in earthquake

protection of structures. An example of this is

the recently completed Law and Justice Center

Building in San Bernardino, California. The

design of base isolation for this building is

illustrated on Figure 2 (Way, 1986).

In this strategy of aseismic design, the entire

structure is founded on several bearings which

are relatively flexible in the horizontal

direction and simultaneously capable of

supporting the vertical gravity load of the

structure. Isolation from severe ground motions

is effected by providing a low horizontal

stiffness to the bearings so as to provide a low

frequency to the predominant horizontal mode of

the structure vibration. Most earthquake ground

motion being in the 3 Hz to 10 Hz range is thus

effectively filtered and only relatively small

accelerations are imparted to the isolated

structure. The general principles are Illustrated on Figure 3. For simplicity, it is

assumed in this figure that the entire structure

displaces horizontally on the collection of

bearings in a more or less rigid fashion.

Figure 3 shows an idealized response spectrum of

earthquake type ground motion. The conventional

structure may have predominant periods in the

range where the ground motion is amplified. By

providing a flexible base, the fundamental

period is shifted from T1 to T2 which is outside

the range where the ground motion is

amplified. Very simply, the principle of base

isolation is to effect this shift of the

predominant mode of vibration. Indeed, in the

extreme case one could support the entire

structure on frictionless bearings which then

theoretically totally isolate the structure in

the horizontal direction. Obviously, the

resulting displacements would be significantly

large and severe measure to control these

displacements would discourage such an approach. Even a nominal shift in the

frequency, such as attempted in most base

isolation solutions, increases the displacement

response to the point where it becomes necessary

that some measure to control these displacements

be incorporated as an integral part of the base

isolation system.

The seismic behavior of base-isolated structures

can simply be looked at in terms of the response

of damped lower frequency single mass oscillators; and on this basis, compared to the

behavior of non-base-isolated structures. But

from the practical approach, the similarities

end there and the seismic performance of baseisolated structures could be radically different

from that of the conventional.

By decreasing the participation of the baseisolated structure in the seismic response, its

structural components can be generally designed

to remain within the elastic range, and thus

decrease the potential for its structural damage

during severe earthquakes, (damage which is

generally accepted in the conventional seismic

design philosophy). But because of this, the

isolation bearing becomes the all important

structural element sacrificing some of the

redundancies inherently present in the

conventional design. Details of the structural

design need special attention to assure proper

vertical and horizontal load distribution to the

dispersed and discreet bearing elements. Very

broadly then, the design philosophy attempts to

relieve the risk to the building structural

elements by an approximately equal sharing of

the risk by the bearings.

109
Much has been written and published about base

isolation and its strategy of seismic design. A

recent workshop (ATC, 1986) on base isolation

and passive energy dissipation brought together

the current thoughts about the use of these

techniques for aseismic design of building

structures. Some codes have been proposed for

the design of structures using base isolation

(SEAONC, 1986, SEAOSC, 1986). Other guidelines

present specific design issues that need to be

considered (Vaidya and Eggenberger, 1984) if

this strategy is chosen. There is still not a complete acceptance of the technique by the

profession although it is agreed by most that

the principle is sound and is applicable at

least to certain types of structures.

The apprehension results, perhaps justifiably

so, from the lack of extensive field data on performance of structures constructed on base

isolation, (Huang, et. al , 1986). For the most

part, current design has to rely rather heavily

on the results of analyses using mathematical

models and scale model shaker table tests which

generate results in a controlled laboratory

environment. Indeed, these do not encompass the

myriad of variable conditions that will be

encountered in the field; and the collection of

field data over several years of observed actual

performance remains the primary means to refine

the technique of base isolation.

But as much as the principal structural element

used in the base isolation system, the bearing,

is an engineered product admitting to QA/ QC with

quite a respectable history of use in other

applications, some in far more arduous

environment, the reliability of the system is

perhaps governed more significantly by

uncertainties in seismic ground motion on the one hand and the structure response on the

other. The former applies to conventional as well as isolated structures. Over the last two

decades, a wealth of information has been

gathered on the nature of ground motion and the sources of uncertainties in its prediction. As

far as the latter, more than conventional

structures base-isolated structures lend

themselves to a more representative mathematical

model for estimating seismic structural

response. Isolation is extracted by demanding

little participation from the superstructure and

the non-structural components in responding to

seismic ground motion, thereby reducing the

attendant uncertainties in the mathematical

model and rendering analytical effort more

tractable

.

In view of the above and in the spirit of

innovation, backed by a wealth of analytical and

experimental data, quite a few in the profession

appear convinced to consider, evaluate, and use

base isolation to mitigate earthquake effects

notwithstanding the paucity of specific field

performance data.

3. CRITERIA FOR SELECTING BASE ISOLATION

As in any design, diversity forces a choice.

Even within the traditional mold, the options

for the building system are many and varied.

The choi.ce of the optimum solution is generally

made in conference with the architect,

structural engineer, geotechnical engineer, and

the seismologist. As new materials present

themselves and eventually get absorbed into

construction use, the traditional mold loses

significance or at least has to expand its

definition to include concepts that these new materials have obviated.

Bearing technology and energy dissipators in the manner of new materials have brought to bear

their potential in the design of earthquake

resistance. Their successful use in other

areas, but nevertheless to mitigate vibration

problems in general, is anticipated to

increasingly encourage their use in reducing

earthquake damage. In this respect, they have

expanded the possible options available to the

engineer. Depending on one's perspective, their

use may be considered as a deviation from the

traditional or merely to precipitate an expansion of the traditional role. For the

moment, relegating the careful consideration of

details to the appropriate engineering phase,

which is imperative regardless of the option

chosen, the authors prefer to look upon the

latter as an appropriate role befitting base

isolation.

And in this subset of the overall design task,

the first choice is whether the building under

consideration should be base isolated; having

chosen base isolation, should the designer

design his own system or use one of the patented

systems? If the use of a patented system is

chosen, which system should be selected?

Finally, what structural issues specifically

related to the base isolation of choice need

special attention to reasonably guarantee

successful performance of the entire structural

system?

Obviously, the design process is dynamic, its

elements interacting with one another, extending

also to the design tasks prior to the subset

being considered here, such as establishing

structural design strategy, architectural

features, and siting. Both initial and lifecycle costs are an important consideration in

the design process also. Previous limited

studies (Vaidya and Eggenberger, 1984; Tarics,

1982)' have indicated that base isolation can be

provided at an equal or lower initial cost.

Improved risk is seen to favorably affect lifecycle costs. For these reasons, it is

anticipated that base isolation will rather

quickly emerge out of the research realm into

practice, if it has not already, and the anatomy

of the decision (Rigney, 1986) determined

predominantly by the deliberation of the

engineer. It is to this latter, that this

110
discussion is addressed and dedicated to

facilitate the decision making.

Accordingly, before moving on to the description

of the various types of base isolation systems,

the following paragraphs are devoted to

discussing a procedure that may be used to

establish if the concept should be considered in

the first place. The need for this first

important decision admits the possibility of an inappropriate use of base isolation. Indeed, it

should be obvious that base isolation is not a panacea that could provide earthquake protection

irrespective of the type and use of the building

structure. In some cases, it may be ideally

suited; in others it may be only marginally

suited; and in still others it may actually

degrade the performance of the building.

Therefore, the judicious selection of buildings

and isolation systems is a major step toward

assuring that base isolation will actually

enhance the seismic performance and, hence, may

be considered as an alternative approach for

upgrading structural design.

The following discussion is rather qualitative

in nature and should be supplemented with case- by-case quantitative evaluations in the event

that on the basis of this qualitative procedure,

base isolation cannot be clearly eliminated as a strategy. Without getting too encumbered with

specific detail at this point, but keeping in

mind that what is being attempted is isolation

by detuning, energy dissipation by some damping

mechanisms, or usually some combination of the

two, the flow chart on Figure 4 presents the

basic issues to be considered in the process.

Very broadly, the significant issues to be

considered in the decision making are:

• Seismic Zone

• Site Conditions

• Ground Motion

• Lateral Load Resisting Systems

• Physical Dimensions of Structure

« Facility Importance

• Materials of Construction

It is apparent, for example, that low seismic

zones may not justify the additional cost of

base isolation in reducing the already low

seismic risk. In any event, depending upon the

building physical dimensions, other loads, such

as wind, may govern lateral design. In such

zones, conventional techniques for upgrading

structural design to earthquake forces may be

the preferred approach. For Zone 2 and perhaps

Zone 3, reinforced elastomeric pads without

additional energy absorption, wind loads

permitting, may be a cost effective

alternative. While for regions of relatively

high seismic risk base isolation with an effective energy dissipation may be the only way

to reasonably assure continued operation of the

facility through and following a major

earthquake

Site conditions in combination with the seismic

zone determines the expected range of ground

motion characteristics. This is a step most

fraught with uncertainty. If, for example, the

likelihood of encountering predominantly low

frequency motion is high, the use of base

isolation may not be appropriate as detuning may

demand that the resulting system have a significantly long period resulting in

unrealistic demand for the control of

displacement. The use of an energy dissipating

system in this case may be more appropriate.

The physical dimensions, foundation conditions,

and the primary lateral load resisting system

may further determine the applicability of base

isolation and its type that could lead to an optimum performance. The facility importance

and the extent and type of damage that can be

tolerated will also impact on whether base

isolation should be used.

The following discussion is directed towards

screening potential candidates for base

isolation. Specific structures should not adopt

base isolation nor should they be eliminated as potential candidates solely on the basis of the

following criteria. Some quantitative

justification is necessary and this can usually

be obtained without much technical effort (e.g.

Kelly, et. al, 1986). It appears at the outset

that base isolation is suitable for the

following general situations:

• The site is located in a high seismic

zone

• The structure is not founded on soft

soil.

• The building is low to medium height.

• The building has a relatively low shape

factor (H/L 1)

• The contents of the building are sensitive to high frequency vibration.

• The lateral load resisting system results

in a rigid structure.

The flow chart shown on Figure 4 includes only

the major steps in the decision process.

Several details that must be considered in each

step are excluded to preserve simplicity and

clarity in the flow chart. Some of these

details are discussed in the following

paragraphs.

Architectural consideration would lead to the

selection of the structural system while siting

and site conditions would determine the seismic

zone and the seismic coefficient. The procedure

envisioned here first performs a preliminary

conventional design, say on the basis of UBC

( ICBO , 1984) or SEAOC (SEAOC, 1984). For this

design, evaluate the dynamic properties,

specifically the first mode frequency, structure

111
mass, and stiffness. Estimate the level of the

expected ductility demand in the structural

elements in the event of the design earthquake

and the maximum credible earthquake. If the

ductility demand is acceptably small, then no further consideration of design options is

necessary and the conventional design may be

refined and the costs evaluated.

Generally, the ductility demand is influenced by

the structural system, site seismic potential,

and the predominant mode frequency. If the

ductility demand is judged to be unacceptable,

then either revise the conventional design or consider base isolation.

If the base isolation option is chosen, first

estimate the required predominant mode frequency

to reduce the seismic coefficient

significantly. Establish the stiffness and

damping parameters of the base isolation

system. If the site soil stiffness is

relatively low, about the same magnitude as the

required base isolation stiffness, then base

isolation may not be appropriate. If the site

soils are firm, select and design the base

isolation system. Evaluate the isolated

structure dynamic parameters and compute system

response. If the response is acceptable, refine

the structure design and evaluate costs.

Generally, unacceptable response in the base

isolated system takes the form of large

displacements, and the attendant problems

associated with the interface of base-isolated

and non-base-isolated components of a building

system. If the displacements are unacceptably

high, revise the base isolation system design.

As can be gathered from the above discussion,

several issues need to be looked at

simultaneously. Quantitative values of

estimated response of various base isolated

structural systems would indeed expedite the

process. This work is currently being conducted

by the authors. Research work is also being

conducted on the influence of site conditions on base-isolated structure response (Constantinou,

1986). When these and other results become

available, it is hoped that specific design

criteria may emerge. But until then careful

evaluation on a case-by-case basis remains the

primary means to reasonably assure successful

performance of base isolation for earthquake

protection of structures.

4. BEARING SYSTEMS

From the brief discussion about the design

philosophy and what is being attempted, it

follows that the "ideal" base isolation system

will perform the following functions.

1.

Minimize lateral loads on the structure

and minimize the attendant relative

displacements

.

2. Safely support the vertical load of the

structure.

3. Provide restraint against other

environmental loads such as wind.

The primary elements of any base isolation

system are accordingly, isolation, energy

dissipation, and restraint. When mitigation of

the earthquake forces are effected predominantly

by friction based elements either restricted to

the base or dispersed through the structure, the

design is sometimes called a passive energy

dissipation system. In essence, this may be

looked upon as increasing the inherent damping

in a structure by enhancing the elements that cause damping in the first place. Thus, very

broadly, passive energy dissipation acts in

qui-te a different manner than does isolation.

Although these methods are discussed here, their

broad scope, applicability, and design

variations make it necessary to restrict the

discussion to only those types that are located

in the base of the structure and in this context

are treated much the same as base isolation in

which the inevitable means for energy

dissipation has taken on the greater

significance.

Of the three elements that constitute a base

isolation system, restraint assumes a relatively

subsidiary role; its use seen as an afterthought

to supplement the performance of the former two

elements, namely, isolation and energy

absorption. Isolation and energy absorption are then the more significant elements engineered

into the overall structural strategy, thus

deviating from the traditional approach, their

main tenet being to reduce the destructive

horizontal forces and motions imparted to the

structure. Pure isolation attempts to achieve

this by detuning, while energy absorption by

dissipating the earthquake energy at the

structure/foundation interface.

In most currently promoted bearing systems,

neither isolation nor energy absorption is

solely used at the exclusion of the other;

rather, these two elements are combined in a manner to produce synergism and effect the most

reduction of destructive horizontal motions.

Various combinations have been proposed; some have been actually implemented for aseismic

design; several have been tested in a laboratory

environment to substantiate their claim at

reducing seismic lateral loads; and most have

been subjected to mathematical analysis to

illustrate their potential. Oversimplifying

here only for the sake of clarity, it can be

stated that the different ways in which the

elements of isolation and energy absorption are combined result in different systems.

Consequently, one is not that radically

different from another for after all there are only two basic elements to be combined and if we keep this in mind and look at all systems in

112
this light it is not too difficult to understand

the expected performance of each.

The following paragraphs briefly describe the

bearing devices and their anticipated

performance in isolating the structures they

support. A sincere attempt is made here to

include all proposed, analyzed, tested and

actually used systems. It is only through

inadvertency that some may have been missed.

4. 1 Plain Elastomer Bearing

The simplest isolation device is a plain

elastomer bearing. Due to their inherent

properties, elastomers are quite flexible

relative to the traditional materials of

construction and placed under structures can isolate them from ground motions . Both shear

stiffness and compressive stiffness are inversely proportional to the thickness of an elastomer block. In addition, however; the

compressive stiffness depends upon the areas of

the sides of the elastomer block free to

bulge. The disadvantage of plain elastomer

bearings is that for the thickness of elastomer

required to obtain horizontal isolation the

vertical stiffness is relatively low and it is

difficult to sustain the gravity loads of the

structure without undergoing significant

vertical displacements. A three-story school

building in Skopje, Yugoslavia uses plain

elastomer bearings for isolation (Siegenthaler

1970).

4.2 Reinforced Elastomer Bearings

The most common base isolation bearings comprise

steel-laminated elastomers. These are also most

commonly used as bridge bearings. Neoprene and

rubber with durometer hardness in the range of

50 to 70 are the more commonly used elastomers

for base isolation bearings. Since shear

stiffness of the bearing depends upon the total

thickness of the elastomer, these bearings

retain the horizontal flexibility while their

vertical stiffness is significantly increased by

the presence of the steel plates. Additionally,

the elastomers being non-linear dissipate energy

through hysteresis. If no special compounding

is used, elastomer bearings of this type provide

an equivalent viscous damping of about five

percent. Bearings of this type have been used

to isolate the structures of a Nuclear Power

Station in Cruas , France.

4 . 3 Reinforced Elastomer Bearings High/Damping

Rubber

To reduce the displacement response, it may be

necessary to provide means for additional energy

dissipation. One way is to specially compound

the elastomer such that the bearing in deforming

in shear describes a relatively large hysteresis

loop thereby dissipating a proportionally larger

energy per cycle. High damping elastomer

bearings have been implemented under the Law and

Justice Center Building in California (Way,

1986).

4.4 Reinforced Elastomer Bearing with External

Energy Dissipation Mechanism

Here the performance of the elastomer bearing is

supplemented with a mechanical energy dissipator

which acts in parallel with the bearing.

Plastic deformation of steel beams either in

torsion or flexure, extrusion of lead, rubber

bumpers are some of the devices proposed to

provide energy dissipation. Several specific

devices based on the plastic deformation of

steel and extrusion of lead have been tested

(Skinner et. al, 1975; Skinner et. al, 1977;

Stiemer and Chow, 1984). Several of these

devices have been included in Bridges in New

Zealand (Tyler, 1984).

4.5 Reinforced Elastomer Bearings with Internal

Energy Dissipation Mechanism

In this concept, the reinforced elastomer

bearing includes a vertical hole in its center

which is filled with lead. In undergoing shear

deformations, the lead plug deforms plastically

in shear and thus results in energy absorption

in addition to that provided by the elastomer.

The William Clayton Building in Wellington, New" Zealand is isolated by bearings of this type

(Megget , 1978; Megget, 1984). Additionally,

several bridges in New Zealand also use bearings

of this type.

4.6 Reinforced Elastomer Bearings with a Frictional Interface

In this concept, a frictional interface is

included between the bearings and the structure

they isolate (Richli, et. al, 1980). A large

seismic event may induce slip of the isolated

structure on the frictional interface, thus,

dissipating additional energy in friction. In

addition to limiting the seismic forces imparted

to the structure, the frictional interface also

limits the shear deformation of the bearings.

Bearings of this type have been used to isolate

the structures of the Koeberg Nuclear Power

Station in South Africa (Vaidya and Plichon,

1986).

4.7 The Alexisismon (Ikonomou, 1979)

In this concept of base isolation, the

functions of supportive vertical load and

horizontal isolation are performed separately by

two components. A pot bearing carries the

vertical load and provides relatively small

frictional resistance in the horizontal

direction. An elastomer block supplies the

horizontal stiffness as a means to provide a restoring force against horizontal

displacements. In concept, this may be

comparable to the reinforced elastomer bearing

with a frictional interface. In action, the

113
elastomer force here works in parallel with the

horizontal frictional resistance. This system

has been used to isolate a six-story structure

in Greece (Ikonomou, 1979).

4.8 The Resilient-Friction Base Isolator

The resilient-friction base isolator is

comprised of a stack of flat steel rings which

enclose a central rubber core and if required

peripheral rubber cores within the flat rings.

Here also, the friction force acts in parallel

with the restoring force of the elastomer

core. The entire vertical load is supported by

the flat rings and the rubber cores only

distribute the sliding displacements. A

significant amount of analysis and scale model

tests (Mostaghel et. al , 1986) have illustrated

the potential of this system.

4.9 The Earthquake Barrier System (Caspe, 1986)

This concept relies primarily on sliding

friction to provide a force barrier. The

barrier physically consists of a friction

assemblage with a predetermined frictionslippage level attached to the structure at its

base. One suggested assemblage uses two

perpendicular sliding rails operating

independently. A horizontal diaphragm restrains

torsional motion of building columns which are supported by the friction assemblage. Hydraulic

dampers or neoprene shear springs may be added

in parallel with the friction assemblage to

enhance the performance of the barrier

systems. Analytical as well as experimental

work has been performed on this system (Caspe,

1984).

4.10 Steel Roller Bearings

Steel roller or ball bearings can provide the

vertical load carrying function with an independent device used to restrain or limit the

horizontal motion. A four-story reinforced

concrete frame building in Mexico City is

supported on roller bearings. Each bearing

consists of 100 one-centimeter diameter steel

balls. Horizontal restraints are provided by

steel limiting cables (Gonzalez-Flores , 1974).

In the event of severe ground motions, the

limiting cables may need an additional energy

dissipation device.

4.11 The Coil Spring Concept

Perhaps comparable in simplicity to the plain

elastomer pads, this concept uses coil springs

to mount the structure on the foundation

soils. This concept was studied to base-isolate

an existing five-story, wood-frame building at

the Naval Post Graduate School in Monterey,

California (Reed and Kircher, 1986). The

disadvantage here is that the horizontal

response is strongly coupled with the rocking

response of the isolated structure. This

concept has been further expanded and studied in

the GERB System (Huffmann, 1985).

The above list of isolation devices do not

include general geotechnical and structural

approaches such as soft soils, sleeved piles,

etc. Although in principle these can be

categorized as base isolation, the variations

here can be numerous and relatively general. Of

the devices that have been described above, the

reinforced elastomer bearing is the common denominator in the more developed schemes. In

terms of the additional damping mechanisms, the

reinforced elastomer bearings with the

frictional interface, special elastomer

compounding, and the internal lead plug appear

to be the most developed from the point of view

of analysis, confirmatory test, and practical

use for the seismic isolation of building

structures

.

5. STRUCTURAL CONSIDERATIONS

The use of reinforced elastomer bearings is

relatively common place in bridges where they

have been used primarily to accommodate thermal

movements of the bridge deck, and for vibration

and noise isolation for buildings and vibrating

equipment. Indeed the first such used dates

back about 30 years. However, seismic

application is relatively recent. Because of

the dynamic characteristics of building

structures and the ground motions from which

isolation is attempted, this application

deserves a careful consideration from the point

of view of the overall system response.

As mentioned earlier, the structural response of

base-isolated structures is significantly

different from conventionally founded

structures. Vertical load carrying members,

such as building columns and load bearing walls,

are discreet structural elements tied together

horizontally at floor elevations by the floor

diaphragms. The lateral loads induced during

earthquakes due to the structure mass are transmitted to the foundation by braced frames

or shear walls which themselves contain the

vertical load carrying elements. Generally,

these lateral load resisting elements may also

not be continuous but rather discreetly located

within the plan dimensions of the building. The

floor diaphragms are designed to appropriately

collect the distributed inertia forces and

transmit them to these discreet lateral load

resisting elements. Accordingly, a rigid

diaphragm at the base of the building is

imperative in transmitting building shears

uniformly to the bearings.

Special cases may require close attention to the

design of the first floor diaphragm. For

example, internal core walls such as those

enclosing elevator shafts may be the principal

lateral load resisting elements. In such a case, the first floor diaphragm should

distribute overturning moments in addition to

114
the horizontal base shear. In limiting cases of

high aspect ratios of such elements, the

overturning moment and the resulting tension on

at least some bearings may preclude the use of

this strategy of aseismic design.

The location and the design of stairs may also

require special attention. Stairwell walls may

be used for lateral load resistance. These

could be a part of the isolated structure but

special detail is required in the stair sections

traversing the interface between the isolated

building and the non-isolated grade outside.

Conceptually, a gap of about 15 inches between

the building and the perimeter, such as shown

for example on Figure 1, can comfortably

accommodate the relatively large displacements

of the base-isolated structure. Similarly, most

utilities and pipes, sewer lines, etc., can incorporate joints to accommodate displacements

on the order of 10 to 15 inches.

Nevertheless, most problems of details such as discussed above are just that—problems of good

detailing specific to the particular building

under consideration and the particular design.

Examined rationally, these issues can generally

be resolved to satisfaction. In the final

design stages, other issues such as the effect

on the seismic response due to eccentricity,

settlement of foundation soils, effects of

vertical and rocking input motion, etc., should

be investigated at least for confirmation of the

performance of the base isolation. In most

cases, these issues are secondary (Vaidya and

Plichon, 1986) but depending upon the structure

may be of significance.

Specific to the base isolation system, its

components namely, the bearing and the energy

dissipating device, should be carefully

reviewed. -Effects of manufacturing tolerances

of bearings should be examined. Some tests

should be performed on the finished bearings to

confirm their mechanical properties. Although

elastomers such as neoprene and natural rubber

have proven their durability in other

applications, the long-term effects of the

environment and aging remain controversial.

This is perhaps due to the applicability of

accelerated aging tests for the performance of

the bearings. In any event, this is an important issue particularly because elastomers

are known to stiffen with age and lack of

movement. Some means of inspection and testing

of in-place bearings may be desirable.

Actual experience with systems in place will

with time supply required performance data to

improve or refine the base isolation systems.

Newer applications will perhaps present other

issues that have not been thought of, but it

appears that most issues can be solved on a rational basis.

6. CONCLUDING REMARKS

This paper has reviewed base isolation as a method for upgrading the earthquake performance

of building structures. Both new construction

and existing buildings can potentially benefit

from it. However, base isolation is clearly not

applicable to all situations. The paper

discusses criteria for selecting the strategy

based upon the site seismic considerations, the

use, and the type of building structure.

Various base isolation systems currently

promoted have been briefly described. Some of

these systems are more advanced than others in

terms of analytical studies, confirmatory tests,

and actual use. Most base isolation schemes use

the common element of the reinforced elastomer

bearing pad. Additionally, the more successful

incorporate a frictional interface, a lead plug,

or specially compound the elastomer to provide a healthy hysteresis characteristic to the

bearing. This energy absorbing capability

controls the relatively large displacements that

characterize base isolation schemes.

Design considerations arising due to the

particular seismic response characteristics of

base-isolated structures are briefly

discussed. It is suggested that these are problems of good detailing and just as important

here as in the conventional design. Provided

that these are paid the attention they deserve,

base isolation is a viable alternative to

upgrading seismic performance of building

structures.

7. REFERENCES

1. Applied Technology Council, 1986,

"Proceedings of a Seminar and Workshop on Base Isolation and Passive Energy

Dissipation," ATC-17, San Francisco,

California.

2. Caspe, M.S., July 1984, "Base Isolation from

Earthquake Hazards, An Idea Whose Time Has

Come!," Proceedings of the Eighth World

Conference on Earthquake Engineering, San

Francisco, California, Vol. 5, pp. 1031—

1038.

3. Constantinou, M.C. and M.C. Kneifati, 1986,

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Structures," Proceeding of the Third U.S.

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4. Gonzalez-Flores , M. , 1964, "Sisteraa para

Eliminar los Esfuerzos Peligrosos que los

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1986, "Recorded Motion of a Base Isolated

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3. Huffmann, G.K, 1985, "Full Base Isolation

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7.

Ikonomou, A.S., August 1979, "The

Alexisisraon: An Application to a Building

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SP-70 , ACI , pp. 549-586.

9. Kelly, T.E., R.L. Mayers, and L.R. Jones,

1986, "Preliminary Design Procedures for

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10. Megget, L.M. , July 1984, "The Design and

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Frame Building in Wellington, New Zealand,"

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California, Vol. 5, pp. 935-342

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Concrete Frame Building," Bulletin of the

New Zealand National Society for Earthquake

Engineering, Vol. 11, No. 4, pp. 245-254.

12. Mostaghel, N. , M. Hejazi, and M.

Khodaverdian , 1986, "Response of Structures

Supported on Resilient-Friction Base

Isolator (R-FBI)," Proceedings of the Third

U.S. National Conference on Earthquake

Engineering, Charleston, South Carolina.

13. Newmark, N.M. and E. Rosenblueth, 1971,

"Fundamentals of Earthquake Engineering,"

Prentice-Hall, Inc., Englewood Cliffs, New

Jersey.

14. Reed, J.W. and C.A. Kircher, 1986, "BaseIsolation of a Five-Story, Wood-Frame

Building," Proceedings, ATC-17, San

Francisco, California.

19. Stiemer, S.F., F.L. Chow, July 1984, "Curved

Plated Energy Absorbers for Earthquake

Resistant Structures," Proceedings of the

Eighth World Conference on Earthquake

Engineering, San Francisco, California, Vol.

5, pp. 967-974.

20. Structural Engineers Association of

California, 1984, Seismology Committee,

"Recommended Lateral Force Requirements and

Commentary."

21. Structural Engineers Association of Northern

California (SEAONC), 1986, "Tentative

Seismic Isolation Design Requirements,"

Proceedings, ATC-17, San Francisco,

California

22. Structural Engineers Association of Southern

California (SEAOSC), 1986, "Draft Guidelines

for the Design of Buildings with Base

Isolators," Proceedings, ATC-17, San

Francisco, California.

23. Tarics, A.G. , 1982, "Cost Consideration of

Base Isolation," International Conference on National Rubber for Earthquake Protection of

Buildings and Vibration Isolation, Kuala

Lampur, Malasia.

24. Tyler, R.G., July 1984, "Developments in

Energy Absorbing Devices and the Physics and

Engineering Laboratory, DSIR, New Zealand,"

Proceedings of the Eighth World Conference

on Earthquake Engineering, San Francisco,

California, Vol. 6, pp. 969-976.

25. "Uniform Building Code," 1979 Edition,

International Conference of Building

Officials, Whittier, California.

15.

Rigney, R.B., 1.986, "System Selection from

an Owner's View or the Anatomy of a Decision," Proceedings ATC-17, San

Francisco, California.

26.

Vaidya, N.R. and C.E. Plichon, 1986, "On the

Concept of Base Isolation Design in France,"

Proceedings, ATC-17, San Francisco,

California

116
27. Vaidya, N.R. and A.J. Eggenberger, January

1984, "Feasibility Evaluation of Base

Isolation for the Aseismic Design of

Structures," Report to the National Science

Foundation, Washington D.C. by D'Appolonia

Consulting Engineers, Project No. 82-1365.

28. Way, D and Marshall Law, 1986, "Design and

Analysis of a High-Damping Rubber Isolation

System" (Case History of the Foothill

Communities Law and Justice Center),

Proceedings, ATC-17, San Francisco,

California.
trc

o

u!

s

z

o

*><

>

UJ

«JUJ

118

FIGURE

1

-

CONCEPTUAL

DESIGN

OF

BASE

ISOLATION
REF. WAY, 1986

FIGURE 2 - BASE ISOLATION DESIGN FOR THE FOOTHILLS

COMMUNITY LAW AND JUSTICE CENTER BUILDING

119
ACCELERATION 1

PERIOD

TYPICAL RANGE= 1 SEC- 2 SEC

AMPLIFIED ACCELERATION REGION (*0.1 SEC- 0.3 SEC)

IDEALIZED SEISMIC GROUND RESPONSE SPECTRUM

NOTES:

T, = PERIOD OF FIXED BASE STRUCTURE

T 2 = PERIOD OF STRUCTURE ON BASE ISOLATION

FIGURE 3 - PERIOD SHIFT DUE TO BASE ISOLATION

120
FIGURE 4 - FLOW CHART CRITERIA FOR SELECTING BASE ISOLATION

121
Consideration of Earthquakes in Highway Earthworks

BY

Yasuyuki Koga 1

)

Akiomi Shimazu 2)

Eiichi Taniguchi 3)

ABSTRACT

Earthwork sections in highways such as fills

and cuts have been sometimes severely damaged

by past earthquakes. It has been usually

considered to be fairly easy and economical

to repair seismically damaged earthwork

sections when damages occurred. However, the

required level of earthwork sections has

been recently rising because of the large

influence and importance of highway earthwork

sections in case of their failure.

This paper briefly presents a current

practice in highway earthworks against

seismic effects mainly on the basis of the

recently revised editions of Technical Guideline for Highway Earthwork (Japan Road

Association)

KEY WORDS: Highway Earthworks, Seismic

Damage, Seismic Design, Seismic Inspection

1 . INTRODUCTION

Because of the topography of the country, a large portion of highways in Japan are constructed in mountainous areas in which the

topography is steep and the geological and

soil property is complex, or in lower land areas whose ground is poor. It can also be

said that Japan is in heavily severe condition for the road construction and

maintenance because of heavy and frequent

rain, snow and earthquakes. Among the total

length of highways in Japan exceeding 1 .

million km, structural sections such as bridges and tunnels are only 1,700 km and a large portion of rest consists of earthwork

sections. The earthwork sections such as fills and cuts cannot always be assumed to have enough resistance against such natural

conditions

This paper summarizes the viewpoints on the

influence of the earthquake in the current

technical level of the highway earthworks,

which are mainly described in the recently

revised edition of Technical Guideline for

Highway Earthwork (Japan Road Association).

2. PREVIOUS EARTHQUAKE EXPERIENCE AND DESIGN

VIEWPOINTS IN HIGHWAY EARTHWORK SECTIONS

Earthwork structures are one of the earthquake-prone structures among conventional

civil engineering structures. Numerous

damages occurred in several recent earthquakes in Japan. For example, settlements

and cracks of embankments due to the

liquefaction of sandy ground occurred in a large scale in the Niigata Earthquake of 1964

and the Nihonkai-chubu Earthquake of 1983 as

well. In the Tokachi-oki Earthquake of 1968,

the slide of embankments was enlarged because

of the soft ground and also the increase of

water content of the fill by the rainfall

prior to the earthquake. The slides of the

embankment constructed on the cuts of

mountainous slopes occurred in the Oitakenchubu Earthquake of 1975 and the Miyagikenoki Earthquake of 1978. In addition to these

examples, slides and rockfalls took place in

a large scale in the cuts and natural slopes

above the road in the Izu-Oshima-Kinkai

Earthquake of 1978. These examples summarize

the feature of the damage to the road earthwork sections.

In spite of such a lot of damage examples, no explicit considerations are given in the

design of general highway earthwork structures. The reason is described as follows in

the Highway Earthquake Essentials (1).

(1) Repairing of earth structures such as cuts and fills is rather easy compared

with other structures. In the case of

fills, the strength of fills and ground

increases as time passes after construction, then the stability increases.

(2) In the fills and cuts of ordinary scale,

ordinary soil and geological properties,

the standard slope gradient is usually

applied, which considers the fill and

cut height, and soil classification.

Since the standard slope gradient is

obtained on the basis of long experiences that the earth structures endured

meteorological condition in Japan, such

slope gradients are assumed to ensure

the stability of the cuts and fills

against an external load below some magnitude

(3) Much is still unsolved about the dynamic

soil properties and earth pressures

connected with retaining walls, however,

it is known from the past experience

that retaining walls can functionally

stand without particular consideration

1) Head, Soil Dynamics Division, Construction Equipment and Method Department,

Public Works Research Institute, Ministry

of Construction, Japan

2) Head, Construction Method Division,

Construction Equipment and Method Department, Public Works Research Institute,

Japan

3) Dr. of Eng., Head, Planning Division,

Planning and Research Administration

Department (Formerly Senior Research

Engineer, Soil Dynamics Division), Public

Works Research Institute, Japan

122
of the seismic effect if the construction is secured. Therefore, no seismic

effect is taken account in the design of

retaining walls of ordinary size.

But it is desirable to conduct investigations against earthquakes in case of

the embankment on a sandy ground which

may liquefy during earthquakes, which is

high and close to an important structure, or whose damage may probably give

a large influence to road traffic and

roadside facilities.

Abovementioned is a summary of past

experience on seismic effect and

fundamental viewpoints on highway earthwork sections up to the present. On the

basis of such fundamental viewpoints, a little more concrete viewpoints and

calculation formulas for earthquakes are described in the technical guidelines of

highway earthworks recently revised.

3 . VIEWPOINTS OF EARTHQUAKES IN HIGHWAY

EARTHWORK SECTIONS IN THE NEW EDDITION

OF TECHNICAL GUIDELINE FOR HIGHWAY

EARTHWORK.

(1) General and fill slopes

Among the technical guideline for road

earthworks recently revised, seismic

designing is referred in Technical

Guideline for Highway Earthwork —

Countermeasures for Soft Ground (2),

Countermeasures for Slope Stability (3),

Retaining Wall, Culvert and Temporary

Structure (4) . Though the expressions

of the viewpoint of earthquakes in these

technical guidelines are not quite

identical, the fundamental viewpoint can

be shown as in the flow chart in Fig. 1. This is a simplified version of a flow chart for the highway embankments

shown in "Technical Guideline for Slope

Stability". Basic ideas hold in other

structures. That is, the conditions to be considered in the investigation for

the stability of embankments are following three items (3).

(i) Condition for embankment proper

(a) When the height of the fill exceeds

the standard values in Table 1.

(b) When the fill material comprises

soils with a high water content and

a low shearing strength (especially

volcanic soils with a high water

content)

(c) When the fill material comprises

soils with a high water content in

which pore pressure tends to increase easily (e.g., soils with a high clay and silt content, volcanic

soils)

(a) When the fill foundation is unstable

as poor ground and landslide

(including liquefiable sandy ground

during earthquakes)

(b) When the fill is easily affected by

spring water from foundation slope

(e.g., partial cut fill section,

widened embankment, fill on slope,

fill across valley).

(c) When the fill slope may be innundated or the slope toe may be eroded

during floods (e.g., fill in ponds

and along rivers)

(iii) Influence by the fill failure

(a) When serious damage may be given to adjacent structures in case of

failure

(b) When a long time is needed for

repairing work and the road function

may be considerably disturbed in case of failure (e.g., fills on slope in mountainous roads without

other alternative roads).

The flow in Fig. 1 can be classified

into following three cases from investigation items and methods.

(i) None of three conditions above

mentioned can be a problem. Standard

slope gradients will be used.

(ii) Though the fill and ground condition

is not good enough, the influence by th

fill failure will be relatively small.

Or though the influence by the fill

failure is large, such a possibility

is small because the fill and ground

condition is good. A static stability

investigation including a stability

analysis is conducted.

(iii) The fill and ground condition is not

good enough and the influence on the

adjacent structures and road functions

by the fill failure is quite large. A

static and seismic stability investigation including a stability analysis

is conducted.

In a preliminary survey stage the

quality of the fill material is

estimated considering Table-1 and above

case (i) . The method to determine the

strength of the fill material in the

detailed survey stage will be described

later in the Chapter 4.

(2) Soft ground (2)

Soft ground is an example of a poor

ground condition. When there exists a sandy ground which way liquefy or a clayey ground which may remarkably

deform during earthquakes in a section

(ii) Conditions for fill foundation where the influence by the fill failure

123
is quite large, the investigation for a seismic condition and a proper consideration must be given when needed.

Especially, when the liquefaction

phenomenon occurs during earthquakes in

a wide area of sandy ground supporting a fill, the bearing capacity for the fill

is remarkably reduced and consequently

the fill is severely damaged. Therefore

it is necessary to estimate the ground

condition. The methods used for the

estimation varies in the several con- struction stages from planning to designing

In the preliminary survey stage with

little detailed information of ground

soil property, the existence of the

ground prone to liquefaction is roughly

grasped and the possibility is estimated

from the simple information such as micro relief, N-value from the SPT and

gradation distribution.

And the investigation method thereafter

is to be determined. Subsequently it is

evaluated whether the fill is damaged on

the basis of soil property obtained from

detailed investigation and stability

analysis. Countermeasures must be

investigated if necessary.

A simplified chart is proposed to judge

the possibility of the liquefaction of

the ground and the loss of the fill

stability on the basis of N-value from

the SPT and gradation characteristics

without any stability calculation (Fig.

2) . This figure shows critical N-value

(Ncr) to give failure condition for each

soil type against specified seismic

coefficient. In the investigation of a specific fill, measured N-values must be (4)

compared with critical N-values in Fig.

2 considering the depth and soil kinds.

The evaluation of the ground as a whole

whether the fill is damaged or not must

be done from overall comparison of

measured N-values and critical N-values

along total depth of the ground. As one

of the methods, the stability coefficient for each depth is defined by

N/Ncr. It has been recently shown that

the average value of the stability

coefficient for total depth to be

considered has a reasonable correlation

with a damage fact. Therefore Fig. 2 can be used a rough estimation. A

detailed procedure, though not quite the same as that in the technical guideline,

is described elsewhere (5)

Clayey soils generally show less tendency

to decrease the shear strength than sandy

soils due to vibration. Loss of

stability and excessive settlement will occur against static load for a clayey

ground where some damage can possibly

occur during earthquakes. In other

words, if sufficient consideration is

given to secure static stability,

seismic stability can be generally kept

due to the margin to static stability.

However, some fills on clayey ground were known to be severely damaged where

static stability was assumed not high

enough. Therefore seismic stability

must be investigated in clayey ground

where remarkable damages may occur during earthquakes from detailed results

of soil exploration.

The detailed stability analysis method

of the fill on poor ground will be

described in Chapter 4.

The stability of retaining wall on poor

ground must be investigated as well.

(3) Retaining wall (4)

Usually earthquake load is considered

for only the retaining walls having

height over 8 m from following reasons. Essentially, very few retaining walls

have collapsed in past damage examples.

Large damages were mostly caused by the

collapse of the overall ground which

supports retaining walls and fills.

Therefore retaining walls of ordinary

size will stand against earthquakes

without particular seismic calculation

if they are carefully designed and con- structed under static condition. On the

other hand, because the retaining walls

over 8 m are considered to exceed the

above experience, a seismic calculation

is required. Moreover, seismic calculation is also required for the retaining

walls over around 5 m when the influence

of the collapse of the retaing wall can

be large.

Cut slope (3)

Cutting slope is based on past experience more than fill works. Standard

slope gradients in Table 2 are generally

used as a reference, because natural

ground generally consists of complex and

inhomogeneous soil layer and it gets

less stable due to weathering after

construction. In cutting also as in the case of fills, it is necessary not to apply the standard slope gradients in

Table 2 but to use such measures as lessing slope gradients depending on three conditions of ground condition,

cutting conditions and influence by the

failure as in the case of fills.

The investigation procedure of the

stability analysis of cutting slopes is

similar to that of Fig, 1 for fills.

That is, investigation of seismic

stability is conducted when the ground

or cutting condition is not good enough

and the influence by the failure is

large

However, the reliability of stability

124
calculation for stability investigation

of cut slopes is less than that of fill

slopes from the above background.

Therefore, in the consideration of

earthquakes in cut and natural slopes,

it is desirable to refer to the past

features of landslide disasters during

earthquakes and also to consider the

countermeasure being effective to the

landslide caused by not only earthquakes

but other effects. This is because the

countermeasure for the rainfall should

also be effective to earthquakes to some extent

Following facts are known from the past

earthquake damage to cut and natural

slopes

(i) There are more cases of shallow slope

failure and falling rocks in cut and

natural slopes than in the rainfall

cases

(ii) The damage ratio is high in the cut

and natural slopes where the longitudinal and cross sectional shape

is convex. The top of the failure

reaches near the ridge.

(iii) Very few damages occur to the slope

protection works which (a) considers

earth pressure, or (b) closely touches

to stable ground, or (c) is flexible,

such as lattice walls and cast-inplace concrete crib-works.

(iv) Among the stone and block masonry

retaining walls, dry masonry or mortar

masonry with steep gradient and thin

grout concrete are damaged.

(v) Concrete spraying, mortar spraying,

rockfall prevention fences work

efficient to some extent.

It is required to refer to the abovementioned and to estimate the seismic

damage possibility and to conduct

proper countermeasures of cut slopes.

4. SEISMIC CALCULATION METHOD FOR HIGHWAY

EARTH WORK SECTIONS

Following calculation formulas can be

used as a reference to conduct a detailed

seismic calculation on the basis of soil

exploration results.

(1) Fill slope (3)

Following equation can be used to obtain

a safety factor for seismic stability,

which is based on seismic coefficient

method assuming a circular slip surface.

- _

I [c • £+{ (W-u *b) cosa-kh »W .sinai tan® ] "s ~ — — —

E (W • sincc+kh *W • ~) ... (1)

where F s

: Factor of safety,

r : Radius of slip surface,

c : Cohesion of soil,

$ : Angle of shear resistance,

£ : Length of arc of slip

surface cut by each slice,

W : Weight of slice,

u : Pore water pressure,

b : Width of each slice,

k : Design seismic coefficient,

h ; Vertical distance between

the center of gravity of

each slice and the center of

slip circle.

(See Fig. 3)

The Safety factor must be greater than

unity in the seismic design of the fill

slope

As to the seismic stability calculation,

researches to consider dynamic soil

strength are now under way, and dynamic

strength characteristics of compacted

fill material and weathered residual

soils are not well known yet. However,

ordinary strength parameters c and 4) obtained from static tests can be used

for ordinary stability analysis, because

it is troublesome to do dynamic soil

strength tests for fill slopes of ordinary size and the dynamic strength of

well compacted fill material is assumed

nearly equal to its static strength.

Strength parameters can be expressed in

terms of effective stress or total

stress depending on whether pore

pressure against shearing is considered

or not. In case of the total stress

expression, only the pore pressure

from ground water level is considered.

In case of the effective stress

expression, the excess pore pressure

caused by earthquake vibration is also

considered. This situation is summar- rized in Table-3.

(2) Fills on soft ground (2)

The estimation of ground strength is

imoortant for the fills on soft ground.

The expression of soil strength of soft

ground varies depending on the soil

kinds such as sandy and clayey soils.

1) Sandy ground (liquefiable ground)

At present in Japan, the soil strength

of liqueficable ground is mostly

expressed in terms of effective

stress for stability analysis. The

safety factor for fill stability is

given by the following equation.

E ( c

' • £+(W-u0 °b-ue °b) • cosa • tancp '

}

Fs = — — ~~

E(W-sina+kh*W • —

where c' ,p

T

: Cohesion of soil and

angle of shear resistance

in terms of effective

125
stress

,

uo : Pore pressure corresponding to the ordinary

ground water,

ue : Pore pressure generated

by the earthquake,

and other notations are the same

as those in Eq . ( 1 ) .

The soil strength of liquefiable

ground can also be expressed in terms

of total stress, in which the safety

factor is calculated by

E(xf • O

F s = t- ............ O)

Z(W’sina+k^*W • —

)

where tf : dynamic soil strength

expressed in terms of

total stress.

Calculation by Eq. (2) needs excess

pore pressure caused by seismic

loading, which is estimated by dynamic

shear stress in the ground.

force)

Among these, seismic earth pressures are calculated by applying horizontal

seismic inertia force to the wedge in

the trial wedge method, in which an arbitary planar slip surface from the

toe of the back fill is assumed and the

maximum earth pressure obtained from the

force polygon of the wedge becomes an active pressure resultant Pa. Seismic

inertia force is horizontally applied at

the gravitational center of the retaining wall, which is given as K^/W using

design horizontal seismic coefficient Kh

and self weight of the retaining wall W. The safety factors and other indices for

the stability in seismic stability

analyses are as follows:

(i) Safety factor for sliding: F s >_ 1.2

(ii) Stability for overturning: Eccentric

distance |e[ <_ B/3 (B = width of base)

(iii) Bearing capacity of the foundation:

Fs >, 1.2

2) Clayey ground

It is considered that the decrease of

strength of clayey soils against

dynamic loading is not large and on the

contrary greater values than static

condition can be assumed because of

loading rate effect. However, static

strength is generally used for

stability analysis. That is, the

safety factor is obtained from the

equation below.

cc u

* £ F s = E(W- since+kft “H •

where cuFinal undrained shear

strength when the con- solidation has ended.

(3) Cut slope (2)

As mentioned before stability analyses

are not generally performed for the

design of cut slopes except the ones in

landslide area or old failure area. Stability analyses are conducted for the

designing of repairing works of failed

slopes. In this case, it is better to do stability analyses assuming not only

circular slip surface but other possible

shapes of slip surface.

(iv) Overall stability for the sliding

including backfill and foundation:

the sliding is investigated referring

to above Sections (1) and (2).

(5) Seismic investigation for semi-buried

road and common utility duct

Besides the conventional earth structures such as fills described up to preceding sections, the aseismity of

the semi-buried road constructed at soft

ground with high ground water level and common utility duct playing an important

role as a life line in a city must be

investigated.

(i) Semi-buried road (2)

The stability of the semi-buried road

constructed at the site of high ground

water level is low even for ordinary

buoyant force. The stability for the

buoyant force during an earthquake

must be confirmed considering the excess pore pressure caused by the

liquefaction when it can possibly

occur. The safety factor for buoyant

force is obtained by following

equation.

= Wb + w s + Qs

Us + U[) (5)

(4) Retaining wall (4)

Retaining walls must be stable in terms

of sliding, overturning, bearing

capacity of foundation and overall sliding to consist of backfill and foundation against self weight, external load,

earthpressure and seismic load (seismic

earth pressure and seismic inertia

where Wg: Self weight of semi-buried

road including pavement

and leveling concrete,

Ws: Load by overburden earth

above base,

Qs: Shearing resistance of

overburden earth above

base or frictional

resistance between side126
wall of semi-buried road

and back fills,

Us: Buoyant force by the

hydrostatic pressure at

the bottom of semi-buried

road

Up): Buoyant force by the excess pore pressure at

the bottom of semi-buried

road

The buoyant force Ud is estimated

referring to the formula for the common utility duct in the next

section.

The safety factor must be greater than

1.2 and 1.0 for static case and

seismic case, respectively.

The sidewall is designed as cantilever

type retaining wall.

where W$

Wb :

Qs

:

Qb :

Us:

U D

:

Weight of overburden

earth

Self weight of common duct

including contained materials and leveling concrete,

Shearing resistance of

overburden earth,

Frictional resistance of

sidewall of common utility

duct

Buoyant force by the hydrostatic pressure at the duct

base

Buoyant force by the excess pore-pressure at the duct

base

This method is essentially the same as that to obtain excess pore pressure in

the stability analysis for the fill.

The safety factor for seismic buoyant

force must be greater than 1.1.

(ii) Common utility duct (6) (6) Design seismic coefficient ( 2 , 3 , 4 )

There is no experience of the seismic

damage to the common utility ducts

because they are rather new facilities in Japan. However seismic

calculations for the common utility

duct are to be conducted at following

sites which need precaution from the

past experience of the seismic damage

to other underground structures:

Soft ground, sections whose ground

condition varies steeply, sites where

the structure changes or different

structures such as are connected

vertical ducts, and liquefiable

ground.

Except liquefiable ground, the

response displacement method is used

as aseismic calculations to obtain

longitudinal sectional force. Displacement amplifude of the ground and wave length of the ground motion are needed as design seismic input. The

former is calculated considering the

fundamental period of the surface

layer and regional characteristics,

and the latter considering the shear wave velocities of the surface layer

and bedrock, and fundamental period of

the surface layer.

The possibility of the liquefaction is

estimated by calculating the liquefaction strength and the liquefaction

resistance rate FL using the method of

the highway bridge specification (JRA) on the basis of N-value of the ground

and average grain-diameter Dcq. The excess pore pressure ratio Au/ov' is

obtained from Fl and Fig. 7 , then the

safety factor for floating up is

calculated by the equation below.

The design seismic coefficient for the

stability analysis and earth pressure

calculation of earth structures must be

carefully determined. It is generally

calculated by the following equation

considering the way to determine one for

other civil engineering structures.

kh = vi • v 2

* kho

where k^o : Standard design seismic

coefficient , 0 . 15 ,

: Seismic zone factor,

\>2 '• Ground condition factor.

The factors vi and \>2 are the same as those used in the Highway Bridge

Specification. The important factor V3

used in the Highway Bridge Specification

is assumed as 1.0 because the earth

structure is considered particularly

important if a seismic calculation is

needed for the structure.

Moreover it may be noticed that the

standard design seismic coefficient k^0

equals to 0.2 in highway bridges. This

is because the road bridge stands above

the ground and the response acceleration

can get larger than that of fills which

is united to the ground.

The design seismic coefficient is not

necessarily the maximum response

acceleration divided by gravitational

accelegration which irregularly

fluctuates. This is because the maximum

response acceleration occurs only once during the period of earthquake process

then if the state to exceed the resistance occurs for a very short period the

failure will not occur at once and it is

too severe to assume the maximum valve

to continue permanently as the seismic

coefficient method. This is very

important for a material as soil that

U S + uD

127
does not show a rapid decrease of

strength after its failure point. Consequently, the seismic coefficient kh

used for a stability analysis is

obtained by reducing the maximum value

of assumed seismic motion.

On the other hand, in the sandy ground

described in the Section (2)1), it is

sometimes required to estimate the excess pore pressure u e during an earthquake. It is generally estimated from

the comparison of seismic shear stress

and dynamic shear strength of the

ground. The seismic shear stress is

roughly estimated from the acceleration

at the ground surface. In order to simply express the above shear stress

and shear strength, they are generally

converted to a shear stress of uniform

amplitude and a certain number of

cycles. However, the liquefaction

strength in the Highway Bridge Specification is represented as the maximum

value of shear stress process which

changes irregularly. Therefore it is

reasonable to compare the above shear

strength with the maximum shear stress

which is estimated from the maximum

acceleration at the ground surface.

That is, the seismic coefficient to obtain the excess pore pressure should

be the maximum value of the ground

motion.

There is no established idea if it is

necessary to consider at once the design

seismic coefficient and the excess pore

pressure mentioned above. One idea

sometimes used is to consider the

seismic coefficient and the excess pore

pressure separately because the peak

time of seismic motion and excess pore

pressure do not generally coincide.

There are many questions to need future

research. They are; (1) if it is

reasonable to express the seismic

strength of the liquefiable ground

through pore pressure, (2) if it is

appropriate to consider the seismic

coefficient and excess pore pressure

separately, (3) how the reference factor

of safety should be determined when they

are considered separately, and so on.

5. OTHERS

(1) Seismic inspection

So far described is fundamentally based

on the several technical guidelines for 1 . highway earthwork. Though these guidelines explain the maintenance and

administration to some extent, they 2. essentially focus on construction works.

On the other hand, it is an important

issue how the aseismicity of existing

structures should be evaluated when we consider the aseismicity of road earth- 3. works. As regards to this issue, the

seismic inspection such road structures

as bridges, tunnels, pedestrian over- bridges and embankments were conducted

in 1986 through a notification by

Director-Generals of Road Bureau and

City Bureau, MOC. This kind of inspection has been conducted in 1971, 1976

and 1979. The inspection of this time

newly handles the fills on steeply

inclined ground and at approaching to bridges. The method of the inspection

is a basically simple one which con- siders the topography, the inclination

of the foundation, the fill height, the

ground condition, the catchment situation, the slope gradient, etc. It is

used to judge the aseismicity on the

basis of these items, and to extract the

sections to need some countermeasures.

(2) Rehabilitation of seismic damage (7)

It is very important to prepare to reduce the damage effect by coping with

the damage smoothly and rationally and

securing the emergency transportation

route. For this purpose, the technical

development project by MOC, "Development

of rehabilitation technique of seismically damaged structures" has been

carried out from 1981 to 1985. As a consequence. "Manual for rehabilitation

technique of seismically damaged structures" was published (7). In the part

III — part of Road Facilities of this

manual, much is described concerning

road fills, natural and cut slopes, common utility ducts, culverts as the

reference for rehabilitation.

6, Conclusions

The recent viewpoints of the seismic

effect on the highway earthworks were reviewed mainly on the basis of revised

technical guidelines for highway earthworks. It has been considered appropriate to cope with the seismic damage by

repairing works. But as the socially

required level for road functions has

been upgraded, the need for concrete

evaluation methods, investigation methods

and countermeasures of the necessary

sections against seismic effect has also

been increasing. Therefore techniques to cope with this demand should be much more developed

Reference

Japan Road Association: Highway Earthwork Essentials, 1983 (in Japanese).

Japan Road Association: Technical Guideline for Highway Earthwork — Countermeasures for Soft Ground, 1986 (in

Japanese)

Japan Road Association: Technical Guideline for Highway Earthwork — Counter128
measures for Slope Stability, 1986 (in

Japanese.)

4. Japan Road Association: Technical Guideline for Highway Earthwork — Retaining

Walls, Culverts, Temporary Structures,

1987 (in Japanese),

5. Koga, Y. and Matsuo, 0.: Simplified

procedure to evaluate the stability of

embankments on liquefiable ground, 19th

Joint Meeting, UJNR, 1987.

6. Japan Road Association: Technical Guideline for Designing of Common Utility

Duct, 1986 (in Japanese).

7. Iwasaki T., et al.: Outline of Manual

for Repair Methods for Civil Engineering

Structures Damaged by Earthquakes, 18th

Joint Meeting, UJNR, 1986.

Table 1 Standard Gradients of Cut Slopes

Soil classification Cutting depth Gradient

Hard rock 1:0.3 to 1:0.8

Soft rock 1:0.5 to 1:1.2

Sand Not dense, and poorly graded 1:1.5 to

Sandy soil

Dense

Less than 5m 1:0.8 to 1:1.0

5 to 10m 1:1.0 to 1:1.2

Not dense

Less than 5m 1:1.0 to 1:1.2

5 Co 10m 1:1.2 to 1:1.5

Sand soil maxed with gravel or rock masses

Dense, or well graded

Less than 10m 1:0.8 to 1:1.0

10 to 15m 1:1.0 to 1:1.2

Not dense, or poorly graded

Less than ICa u -0

10 to 15m 1:1.2 to 1:1.5

Cohesive soil 0 to 10m 1:0.8 to 1:1.2

Cohesive soil mixed with rock masses or cobblestones

Less chan 5m 1:1.0 to 1:1.2

5 to 10m 1:1.2 to 1:1.5

Table 2 Standard Gradients of Fill Slopes

Filling materials Height of fill (m) Slope ratio Remarks

Sand with good grading (SW), gravel, and sand mixed with

gravel (GM) (GC) (GW) (GP)

Less than Sm 1:1.5 to 1:1.8 To be applied to fills with sufficient bearing capacity at foundation ground, which are not affected by inundation.

Typical unified classes are shown in ( ) for reference.

5 to 15m 1:1.8 to 1:2.0

Sand with poor grading (SP) Less than 10m 1:1.8 to 1:2.0

1

Rock masses (including muck)

Less than 10m 1:1.5 to 1:1.8

10 to 20m 1:1.8 to 1:2.0

Sandy soil (SMHSC), hard clayey soils and hard clay

(hard clayey soils and clay

of alluvium, loam, etc.)

Less than 5m 1:1.5 to 1:1.8

5 to 10m 1:1.8 to 1:2.0

Soft clayey soils (VH ) Less than 5m 1:1.8 to 1:2.0

129
Table 3 Strength Parameters and Pore Pressures to be Used for Stability Calculation in Terms of Total Stress and

Effective Stress Methods

Total Stress Method Effective Stress Method

„ .

i Low Permeability

Seismic

c cu> $CU* tlQ c ’ > $

’ > Uo , ue High Permeability cd > ®d> uo

C CU> $cu

cd, 9d

c' , <h' uoue

Strength Parameters Obtained from CU Test

" " " " CD " Strength Parameters Obtained from CU Test

Pore Pressure by Normal Ground Water Level

Pore Pressure Generated by Earthquakes

Fig. 1 Flow of Stability Check for Highway Earthworks

(Case of Fill Slopes)

130
N value

0

Fig. 3 Seismic Stability Calculation Method

by Use of Circular Slip Surface

Soil is classified by the description

of soil profile or the Table below.

Name

Sandy Silt

Fine Sand

Medium Sand

Coarse Sand j

D 50 (mm)

D 5

< 0.071

0.071 4 D 50 < 0.25 ^ D 5 q 5 0.12 4 D so

0.25

0.12

ws

W \u D

Fig. 2 Critical N Value for Rough Estimation Fig. 4 Forces Acting on Semi-Buried Road

u

03

*4-4 0)o

O' J

U

cn

0300

03

0)u

c

03

tf) (V OC

UsOu

U

03 Ua

Excess Pore Pressure Ratio ^u~'

A u

Fig. 5 Relation between Fl and Lu

131
Damage, Strengthening, and Retrofit of Bridges

BY

James He Gates 1

ABSTRACT

Earthquake damage to bridges in California since

1932, including the 1971 San Fernando earthquake

is summarized. The 1986 damage to a seismically

retrofit bridge in the Palm Springs area is

described. The completion of the 16 year

California project to strengthen and retrofit

joints and bearings to increase their seismic

capacity is described and plans outlined for the

next phase of seismic strengthening and retrofit

of columns and substructures of bridges in

highly vulnerable seismic areas. Lessons

learned from recent earthquakes are summarized.

KEYWORDS: Bridges; Damage; Earthquakes; Retrofit

1. INTRODUCTION

This paper is an extension and update of the

previous papers presented at the UJNR in 1983

(5) and 1985 (6) and presents a summary report

describing the history of seismic damage to

bridges in California and recent efforts in

California to strengthening and retrofit of

highway bridges to increase their seismic

resistance.

1 . 1 Damage

About 1.5 moderate to large earthquakes have

occurred every year in California since 1932.

Except for the very damaging 1971 San Fernando

earthquake, bridge damage in California has been

very limited, especially in the time period

before 1971. Since the 1971 San Fernando

earthquake, eleven separate earthquakes have

been investigated in detail by the California

Department of Transportation (CALTRANS) Office

of Structures Design, Post Earthquake

Investigation Team (PEQIT). Eight formal

reports have been written describing these

investigations. This paper summarizes the

bridge damages observed in California since

about 1932, and discusses the implications of

the post "San Fernando investigations. The

recent damage to a retrofit bridge in the Palm

Springs area is also described.

1.2 Retrofit

The task of strengthening seismically deficient

joints and bearings is now almost complete for

the State owned bridges in California. This

9

paper will present a short summary of the

accomplishments of the current retrofit program

and a some insights into the future directions

of the seismic strengthening of bridges in

California.

2. DAMAGE

2.1 Bridge Damage from 1932 to 1971 (18)

In the period from 1932 to 1971 about 26

separate earthquakes with magnitudes of 5.5 or greater occurred in California in regions close

to bridge structures. The rate of seismic

events in this period average about 1.5

earthquakes per year. At least 1500 bridges

were excited by accelerations of 0.35g or more during this period, yet only 33 bridges

sustained damage. Of these 33 bridges only 15

bridges required repairs. Total repair costs

for this 40 year period has been estimated at

$100,000 in 1975 dollars.

Bridges affected in this time period were generally rugged, low short period structures

located in rural areas and consisted primarily

of stream crossings although in the 1950's some freeway and interchange structures were located

close to some events.

Only one bridge was closed to traffic in this

time period and this occurred during the 1940 El

Centro earthquake when a timber cap slipped off

of supporting timber piles. It is interesting

to note that a more modern bridge (New River Br.

No. 58-05), located at the same location was heavily damaged in the 1979 Imperial Valley

earthquake

Most of the damage to highways recorded in this

time period was associated with fill movements

and settlement rather than structural damage due

to dynamic motions.

2.2 Bridge Damage in the 1971 San Fernando

Earthquake (3,19)

The total damage to highway bridges in San

Fernando was about $6,500,000 (4). The major

damage was concentrated within a narrow region

close to and generally within the causitive

fault zone of the magnitude 6.6 event.

Two major collapses occurred in this event due

to the pulling apart of narrow hinge seats. The

route 210/5 Separation (53-1989L) and the

1 Structural Mechanics Engineer, California Department of Transportation, Division of Structures

Office of Structures Design, P. 0. Box 942874, Sacramento, CA 94274-0001

132
Northwest Connector OC (53-1985) were both long,

tall single column type concrete bridges with

several intermediate expansion joints.

The only other complete collapse occurred at the

Route 5 (Truck Lane)/405 Separation bridge

(53-1548). This bridge was a 335 foot long two

span post -tensioned bridge, 38 feet wide on two

monolithic columns. The under-reinforced

columns broke up so much that the bridge settled

to the ground at the bent level.

Damage at several other bridges was heavy enough

to require replacement. A total of 6 bridges were eventually completely replaced (including

the 3 collapsed bridges). Only the collapsed 2 span portion of the Northwest Connector was replaced and the portions which did not collapse

were salvaged and incorporated into the rebuilt

structure

A total of 58 bridges were damaged significantly

enough to require restoration work. Damage was highly varied, with about 14 bridges suffering

damage over $100,000, 23 bridges with damage

between $10,000 and $100,000 and about 21

bridges with minor damage (less than $10,000).

Several important lessons were learned from the

evaluation of this damage:

o Ground shaking in the vicinity of the

fault was much larger than expected,

o Expansion joint seats in continuous

bridges were too narrow and were unrestrained

o Column members lacked sufficient

confinement reinforcement.

As a result of this earthquake, the design force

levels for California bridges was increased

significantly, the reinforced concrete details were improved, seat widths increased and

retrofit of existing bridges was started (4).

2.3 Bridge Damage since 1971

Since 1971 The CALTRANS Office of Structures

Design has investigated several earthquakes.

Since 1979, formal reports of the most

significant of these events have been written:

Bridges Bridge Damage

Date Magnitude Location Investigated Low Moderate Heavy

2/21/73 5.7 Oxnard 30-40 5 0 0

8/1/75 6.1 Oroville 98 14 0 0

8/13/78 5.2 Santa Barbara 10-20 3 0 0

8/6/79 5.5 Coyote Lake (11) 15-20 4 1 0

10/15/79 6.4 Imperial Valley (14) 17 3 1 0

1/24/80 5.5 Greenville (13) 8-10 4 0 0

5/25/80 6.7 Mammoth Lakes (15) 17 0 0 0

11/8/80 7.0 Trinidad -Of fshore( 12) 23 0 0 1 5/2/83 6.5 Coalinga (10) 60 6 0 0

4/24/84 6.2 Morgan Hill (16) 39 0 1 1 7/8/86 5.9 Palm Springs (17) 46 4 3 0

It should be noted that the severity of damage

observed since the 1971 San Fernando earthquake

is very low and reminiscant of the years before

1971. In spite of this lack of damage, several

valuable lessons have been learned.

The effect of earthquakes on various bearing

systems is now better known as a result of the

investigations of the PEQIT studies. For

example the bearings at the Alamo River Bridge, were moved laterally by the Imperial Valley

quake of October 15, 1979 (14). The strong

shaking at this location lasted only four or five seconds and the total duration was about

thirteen seconds. We learned to be very

cautious about assuming that keeper plates,

bearing plate anchor bolts, keeper plate bolts

or welds, and similar details have any

significant effect in keeping a bridge

superstructure on its supports during a major

earthquake. All of the bearings at the end of a span do not always resist horizontal forces

simultaneously. Because keeper plates or other

devices are not set during construction with

exactly the same clearances, only a portion of

the bearings will initially resist a horizontal

force in one direction.

It is not uncommon for a line of bearings at one end of a span to be damaged to varying degrees

after an earthquake. For example the

Trinidad -Of f shore earthquake on November 8, 1980

(12) provided an example of variable bearing

damage at the south abutment of the Fields

Landing O.H., when some bearings obviously

carried a major portion of the lateral force and

others very little load.

Anchor bolts have been placed too close to the

edge of the bearing seats in early bridges and causes concrete to spall off and tear out when

subjected to horizontal loads. For example in

the Coyote Lake earthquake in 1979 (11) the Bent

2 of the 129/101 separation clearly showed this

type of damage.

Anchor bolts are frequently threaded below the

top surface of the pier or abutment seat. This

practice gives results in a reduced area for

shear and a greatly reduced resistance to

bending and failure generally occurs due to

notch sensitivity at the root of thread.

133
Grout pads under bearing masonry plates have

traditionally posed problems during and after

construction and have also been one of the main

trouble spots in minor quakes. Failure of a grout pad will allow the bearing assembly to move and subject to anchor bolts to combined

bending and shear.

Bearing rollers allow normal movements between

the sole plate and the superstructure plate.

Transverse sliding due to seismic loading takes

place on both contract surfaces. Transverse

sliding obviously does not start until the

horizontal force exceeds the vertical dead load

on the bearing multiplied by the coefficient of

friction. This action, once initiated, results

in repeated impacts on the end keeper plates

leading to a progressive bending failure of the

plates, as evidenced by bearing damage at the

First Street Separation during the Greenville

earthquake of January 24, 1980 (13).

We have learned that bridge bearings may not be

what they are represented to be on "As Built"

plans or maintenance records. Adjustments to

bearing details are sometimes made after the

initial construction. Sometimes this work may

not be of the best workmanship. Records of such

work may be inadequate or non-existent. A

conservative approach should always be taken

both in the original design and in the

retrofitting of bridge bearings.

Lastly, it should be noted that even though we

see many events which do not cause much damage,

or where the strongest motions are not located

near any bridges; the threat of heavy damage

still exists and a large event is sure to strike

a major complex of bridge structures in the

future

2.4 Bridge Damage in the 1986 Palm Springs

Earthquake

The Palm Springs earthquake occurred at 2:21

a.m. on July 8, 1986. The epicenter was located

approximately 12 miles northwest of Palm

Springs, CA. The magnitude was 5.9 ml

(Richter). Following the initial shock, many

aftershocks were recorded with magnitudes

ranging from 3.0 to 4.0. Rupture is assumed to

have occurred on the Banning Fault, a branch of

the San Andreas Fault.

Bridge damage was limited primarily to the areas

of Painted Hills and Whitewater, about 12 miles

northwest of Palm Springs. Some damaged houses

were found in the area. Palm Springs suffered

only minor damage, mainly shattered windows and

stock fallen from store shelves. There were

rock slides which closed some roads, moderately

damaged bridges, downed power lines, some damage

to an aqueduct siphon, and serious damage to an electrical substation resulting from.the

earthquake

The CALTRANS Structures Design PEQIT Team of

Steve Mellon and Mark Imbriani investigated

bridge damage in the area on July 9* 11, 1986.

The team inspected 46 bridge sites (45 state

bridges and 1 county bridge) in Riverside

County

Three bridges were found to have suffered

moderate damage and four bridges were slightly

damaged. The team discusses the damages to

these bridges in detail in their report (17).

Of particular interest is the damage to the

Whitewater Overcrossing (Br. No. 56-71). This

bridge is located about 8 miles northwest of

Palm Springs near the town of Whitewater. It is

about 2 miles from the fault and is estimated to

have received accelerations in the range of

0. 5g.

The superstructure of the bridge is a straight,

two span composite steel girder bridge, each

span being simply supported with a cast -in-place

concrete deck on the girders. The span lengths

are 156'*142'. The substructure consists of a single three column concrete bent. Abutments

are seat-type with pedestals, bearings and an expansion joint at the roadway level. The total

bridge is skewed about 37 degrees. The girders

rest on elastomeric bearing pads at all

supports. The bridge was strengthened for

seismic loads by the addition of steel

restrainer cables retrofit in both the

longitudinal and transverse directions.

Most of the damage to the structure appeared to

be caused by transverse movements. The

superstructure. Span 1 at Abutment 1, came to

rest 7 inches west of its original position but

remained on the pedestal.

At Abutment 1 the concrete was spalled at both corners where the deck overhang came in contact

with the abutment backwall. Also on the east

side, the overhang impacted the backwall causing

a crack in the backwall. Three transverse

restrainer cables failed at Abutment 1, along

with the original keeper angles on the abutment

pedestals. These pedestals were very lightly

reinforced and were heavily fractured.

At Bent 2, the cap was lightly cracked on the

inside face of both exterior columns.

The Bent 2 metal rail expansion joint sleeve

(east side) pulled apart and came to rest

misaligned. In order for this sleeve to come

apart, it is estimated that the differential

longitudinal movement had to be in excess of 4 inches

At Abutment 3, a single restrainer cable failed.

Slight cracking and spalling occurred at the

overhang -abutment expansion joint connection,

and wingwall/curb joint.

In assessing the damage to the Whitewater

Overcrossing the investigation team has offered

the following preliminary observations (17):

134
"The shear capacity of the keeper angle bolts of

Abutment 1 was very limited due to lack of

confinement by reinforcement steel in the

pedestal around these bolts. When the keeper

angles failed, the earthquake forces were taken

by the cable restrainers. It is estimated that

the cable resistance was adequate to withstand

the earthquake loads. However, one cable failed

at the turnbuckle when a stud stripped out of

the coupler because it was inserted only 5/8"

into the turnbuckle. It is conjecture that this

failure reduced the transverse capacity by 1/3,

therefore, overloading the remaining two cables

and leading to ultimate failure.

The large transverse offset of the

superstructure In this span, as compared to the

other spans, is probably due to the sudden

energy released when the other two cables broke.

The remainder of the retrofit details which

included the doweled *in cable stud standard

detail and the anchor drum, a bracket bolted to

the girder, behaved well, i.e., developed the

ultimate strength of the cable.

At Bent 2, the crack in the cap at the inside

face of both outside columns is due to lack of

reinforcement in this area. The cap obviously

was not designed for the earthquake moment

developed in this rigid frame during lateral

shaking

At Abutment 3, failure of the restrainer cable

was due to the lack of adequate threading of the

cable into the coupler (1/2"). As happened at

Abutment 1, the stud stripped from the

turnbuckle. It is surmised that shaking was subsiding when this connection failed and thus

did not overload the adjacent transverse cables

as was th§ case at Abutment 1."

3. RETROFIT

The California retrofit program began almost

immediately after the 1971 San Fernando

Earthquake (2,7) Projects which were under

construction were modified where possible by

contract change order to meet increased seismic

requirements. This effort was extended to

include bridges in the design phase.

The California Department of Transportation

identified about 1300 bridges (out of about

13,000) which were primarily deficient in seat

width. These unrestrained joints represent the

initial focus of the Departments retrofit

program. To date about practically all these

bridges have been retrofit at a cost of nearly

$55 million.

The average California retrofit project

consisted of the addition of steel restrainer

cables at hinge and expansion joints to prevent

spans from collapsing. Force levels were identical to those used on new structures (1).

It is expected that the potential for collapse

can be minimized even though extensive damage is

experienced.

The California design details for restrainer

units have evolved to the point where reliable

and economical systems are performing

satisfactorily under service conditions in the

field, although only one installation has been

tested by an actual earthquake (17).

Selection of structures for retrofit in

California was based on a priority system which

takes into account the expected accelerations at

the site, the estimated cost to retrofit the

structure, the cost of replacement in the event

of loss, the length and availability of detours

and the average daily traffic as well as other

factors which reflect the importance of the

bridge in the transportation system.

California developed a simplified restrainer

analysis procedure (8,9) which considers many of

the non-linear features of multi -hinged bridges,

such as expansion joint gaps and the impacting

of the superstructure against the abutment fill.

This procedure is now used to compute restrainer

forces in restrainers in both old and new

construction.

The installation of hinge restrainers added a considerable amount of seismic reliability to

the California bridge system. There remains

however, several weaknesses in various bridges

in the network, especially those located in areas of very high seismicity. The task of

strengthening columns and substructures will

occupy much of our time in the next several

years. By applying non-linear analysis

techniques it is possible to better predict

failure modes and analyze the effects on overall

stability when various components fail.

Preliminary indications are that by increasing

the amount of longitudinal restrainers at joints

and strengthening select columns, we can maintain the stability of long tall ramp

structures even though considerable damage will

occur. Another area for potential strengthening

is the addition of infill walls between very

weak two column bents.

The task of selection, design and construction

of this second phase of retrofit is only just

beginning and I hope to be able to keep you

informed of our progress on this long range

project

135
4.

REFERENCES

1. "Bridge Design Specifications Manual"

Standard Specifications for Highway Bridges

Adopted by the American Association of State

Highway Bridges, Thirteenth Edition, 1983. With

Interim Bridge Specifications through 1985,

Washington, D.C., With revisions by The

California Department of Transportation, Office

of Structures Design, Sacramento, California,

August, 1986.

2. Degenkolb, O.H., "Retrofitting Bridges to

Increase Their Seismic Resistance", Proceedings

of the First Seminar on Repair and Retrofit of

Structures. US/Japan Cooperative Earthquake

Engineering Research Program. National Science

Foundation, University of Michigan, Ann Arbor,

Michigan, 1980, Vol 1, pp 109-114.

3. "Field Investigation of Bridge Damage in the

San Fernando Earthquake." Bridge Department,

Division of Highways, California Department of

Transportation, Report of Investigation Team,

George G. Fung, Richard J. Lebeau, Eldon D. Klein, John Belvedere, and Adlai F. Goldschmidt,

Sacramento, California, 1971.

4. Gates, James H. , "California's Seismic

Design Criteria for Bridges" Journal of The

Structural Division, American Society of Civil

Engineers. December, 1976. pp 2301*2313.

5. Gates, James H. , "Seismic Resistant Bridge

Design in California", UJNR, Proceedings of the

15th Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects, Tsukuba, Japan, May,

1983, Paper No. 2-19.

6. Gates, James H. , "Earthquake Resistant Bridge

Design in California", UJNR, Proceedings of the

15th Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects, Tsukuba, Japan, May,

1985, Paper No. 2-7, NBS IR 86-3364, pg 146.,

May, 1986.

7. Mancarti, G.D., "New Concepts in Earthquake

Retrofitting of Highway Bridges". Paper

presented at Northwest Bridge Engineers

Conference, Boise, Idaho, October, 1981,

Updated Sept, 1984.

8. "Memos to Designers", 15-10, Earthquake

Design (Commentary), California Department of

Transportation, Office of Structures Design,

Sacramento, CA, July, 1984.

9. "Memos to Designers", 21*18, Restrainers at

Hinges and Bearings California Department of

Transportation, Office of Structures Design,

Sacramento, CA, December, 1981, Supplement //l,

October, 1981, Commentary, February, 1981.

10. "Seismic Report, Coalinga, California

Earthquake, May 2, 1983. Caltrans Office of

Structures Design, Post Earthquake Investigation

Team, James Munro and Eugene G. Klein Jr., 1983.

11. "Seismic Report, Coyote Lake, CA Earthquake

of August 6, 1979. Caltrans Office of

Structures Design, Post Earthquake Investigation

Team, Floyd Mellon, Paul J. Jurach, Richard

Hagar, 1979.

12. "Seismic Report, Eureka, CA

(Trinidad -Of f shore ) Earthquake of November 8,

1980. Caltrans Office of Structures Design,

Post Earthquake Investigation Team, Frank Semans

and Ray Zelinski, 1980.

13. "Seismic Report, Greenville, CA Earthquake

of January 24, 1980. Caltrans Office of

Structures Design, Post Earthquake Investigation

Team, Larry N. Storlie, Frank Semans, 1980.

14. "Seismic Report, Imperial Valley, CA

Earthquake of October 15, 1979. Caltrans Office

of Structures Design, Post Earthquake

Investigation Team, Paul J. Jurach, Larry N.

Storlie, George G. Miller and Roy Imbsen

(FHWA) , 1979.

15. "Seismic Report, Mammoth Lakes Earthquake of

May 25, 1980. Caltrans Office of Structures

Design, Post Earthquake Investigation Team,

Frank Semans, Eugene G. Klein Jr., 1980.

16. "Seismic Report, Morgan Hill, California

Earthquake, April 24, 1984. Caltrans Office of

Structures Design, Post Earthquake Investigation

Team, Richard Land and James Munro, 1984.

17. "Seismic Report, Palm Springs, California

Earthquake, July 8, 1986. Caltrans Office of

Structures Design, Post Earthquake Investigation

Team, Steve Mellon, Mark Imbriani, 1986.

18. "Survey of Past Earthquake Damage to

Structures", California Department of

Transportation, Bridge Maintenance Engineer, W.

J. Jurkovich, File 820.1, February 18, 1975.

19. "The Effect on State Highways of the San

Fernando Earthquake, February 9, 1971." State

of California, Business and Transportation

Agency, Department of Public Works, Submitted

Persuant to SB 682, September, 1971.

136
-
Large Scale Model Tests and Analysis of Gravel Drains

BI

Susumu Iai*, Katsuhiko Koizumi**, Setsuo Noda***, and Hajime Tsuchida****

ABSTRACT

Performance of gravel drains,

installed as one of remedial

measures against liquefaction, is

studied with large scale model

tests. The tests are conducted with

a container, set on the shaking

table, made of a stack of 64

aluminum rings; the rings of 200 cm

in diameter being stacked to the

total height of 200 cm. The results

of the shaking tests and the

analysis indicate a simple design

procedure for determining the

spacing between the gravel drains.

KEY WORDS; LIQUEFACTION, DRAIN

1. INR0DUCTI0N

Installing gravel drains is one of

remedial measures against

liquefaction. In general gravel

drains can be installed in various

shapes, but most of the gravel

drains are installed in a columnar

shape as shown in Fig.1. The gravel

drains are usually installed

without vibrations and noises and,

therefore, have an advantage over

compaction methods when the ground

has to be improved near the

existing structures.

Performance of gravel drains was

first studied by Seed and Booker

(1977) [ 4 ] by solving a kind of

consolidation equations. Several

studies followed them [3*6,7].

However, there have never been a large scale model test under

horizontal excitations to study the

applicability of the equation used

by Seed and Booker. Therefore,

present study is carried out.

2. LARGE SCALE MODEL TESTS

When the ground shown in Fig.1 is

shaken by an earthquake, excess pore

water pressure is generated in the

soils around gravel drains and, as a

result, pore water of the part of

the ground bounded by the hexagon in

Fig.1 will flow, as indicated by

arrows, into a gravel drain.

Therefore, it is enough to study the

axi-symmetric part of the ground,

bounded by the equivalent circle

shown in Fig.1, with one columnar

gravel drain at its center.

A model of the axi-symmetric part of

the ground is made in a container

which is set on the shaking table

and designed to enforce, in the

model ground, pure horizontal cyclic

shearing as in the field. This

container, shown in Photo 1 , is made

of a stack of 64 aluminum rings of

which diameter and thickness are 200

cm and 2 cm. Between the aluminum

rings of 3 cm in height, rollerbearings of 2 mm in diameter are

inserted to reduce the friction

between the rings. The total height

of the stack amounts to 200 cm.

Sands used for the tests are taken

from Gaiko district at Akita Port,

where liquefaction took place during

the 1983 Nihonkai-Chubu-Sarthquake.

Gravels used for the tests are

crushed stones. The grain size

accumulation curves are shown in

Fig. 2. The model ground was made by

poring the sand into the water. No

* Senior Research Engineer,

Structures Division, Port and

Harbour Research Institute

** Member, Earthquake Resistant

Structures Laboratory, Port and

Harbour Research Institute

*** Ph.D., Chief, ditto.

**** Ph.D., Deputy Director General,

Port and Harbour Research

Institute

137
tamping or disturbance was applied.

The model of the gravel drain of 60

cm in diameter was constructed with

a casing, which was leveled up from

time to time as the ground and the

drain were constructed. Instruments

are deployed in the array

configuration in the model as shown

in Fig=3 for measuring

accelerations, displacements, pore

water pressures, settlements, earth

pressures, and amount of drained

water during the shaking table

tests

3. RESULTS OF THE MODEL TESTS

More details on the performance of

gravel drain can be understood by

the time histories of the excess

pore water pressures. The results of

the tests at about 80 Gals are shown

in Figs.6 and 7, of which values are

measured by the instrument at P24 in

Fig. 2. These results indicate that

the gravel drain decreases the rate

of generation in pore water

pressures and increases the rate of

dissipation although the gravel

drain fails, at this particular

condition, in reducing the maximum

value of the excess pore water

pressures

The shaking table tests are carried

out under the conditions shown in

Table 1? R-204, R-302, and R-303

are reference tests without a

gravel drain and are to be compared

with R-502 and R-503? which are the

tests with a gravel drain. Among

these tests, R-302 and -R-502 are of

the virgin condition. R-204 is not

of the virgin condition in a strict

sense but practically can be

regarded as of the virgin

condition. R-303 and R-503 are

carried out after R-302 and R-502,

respectively. All the tests are

carried out with a sinusoidal input

motions of 2 Hz' for the duration of

10 seconds.

The shaking table tests resulted

the maximum excess pore water

pressures shown in Fig.4 for the

cases without gravel drains and

Fig.5 for the cases with the gravel

drain. Comparison of these figures

indicates maximum excess pore water

pressures in and around the gravel

drain are generally smaller than

the ones in the ground without a

gravel drain when the maximum

acceleration of the input motion is

about 50 Gals. However, when the

maximum acceleration is about 80

Gals, maximum excess pore water

pressures around the gravel drain

reach the same level as the ones

without the gravel drains.

These results qualitatively indicate

the effectiveness and the limitation

of the performance of gravel drains.

The rest of the present paper is

devoted to the qualitative

evaluation of the performance of

gravel drains.

4. BASIC EQUATIONS FOR GRAVEL DRAINS

Seed and Booker analysed the

performance of gravel drains by the

following equation [ 4 ];

in _ 2hg = jL L.— s r

9u>\ 3t ”

3t mv

" r 3r\^Yw

’ 9 w

_j_ 1 a / ky 9u \

%' 3z dz J

( 1 )

in which

r, z : the radial and verical

coordinates

t : the time coordinate

u : the excess pore water

pressure

Ug : the excess pore water

generated by the cyclic

shear under the undrained

condition

kg , kv

s the coefficients of

permeability in the

horizontal and vertical

directions

yw

s the unit weight of water

mv : the coefficient of volume

compressibility

138
Ug» in Eq.(l), is given by the

following empirical equation [4-3 >

in which

av£

: the effective vertical

stress

N : number of cycles of cyclic

shear stress

Nq : number of cycles required to

cause initial liquefaction

Cu/avi=l.O)

a j an empirically determined

constant that has a typical

value of 0.7

If we note that t = N f = elapsed

time of applicating cyclic shear

stress of which frequency is f Hz,

and tq = Nq f = duration of cyclic

shear stress required to cause

initial liquefaction, Eq.(2) can be

rewritten as follows;

u„ 2 / t

x l/2a

V —-S t cvo

= —arcsm

IT (t

/

Thus

lUg _ Gyp

8t airtq

(4)

The drainage of pore water

influences the value of u/av£ and

thus influences the rate of

generation of pore water pressure

Ug through Eq.(4).

In order to derive the Eq.(l),

Darcy's law and elas to-plas tic

nature of the skelton, made of sand

particles, are implicitly assumed.

It is evident, however, that those

assumptions are not valid in the

state of liquefaction in which the

sand particles completely lose

their mutual contacts [5]. Some

limitations sould exist in the

applicability of the Eq.(l).

Therefore, the results of the model

tests are compared with the computed

results by Eqs.(l) and (4)°

5. COMPARISON BETWEEN THE

EXPERIMENTAL AND ANALYTICAL RESULTS

Among the several constants

necessary for the computation by

Eqs.(l) and (4)> tq in Eq.(4) was

computed for each half cycle of

shaking from the time histories of

shear stress ratios and the

liquefaction resistance curves. The

time histories of shear stress

ratios are computed by integrating,

from the ground surface to the

arbitrary depths of the model, the

measured time histories of the

accelerations multiplied by the

density of the model ground. The

liquefaction resistance curves of

the model ground was obtained by

back-fitting the curve to the test

results without a gravel drain.

Coefficient of volume

compressibility was also obtained by

the test results without a gravel

drain. Coefficient of volume

compressibility, as shown in Fig.8,

increases when the excess pore water

pressure ratio u/av£> increases

beyond 0.5» Moreover, once the

excess pore water pressure ratio

u/

0

Yo becomes greater than 0.95,

it is difficult to find the one-toone correspondence between the

coefficient of volume

compressibility and the excess pore

water pressure ratio. These facts

indicate that the assumptions of the

Darcy's law and the elasto-plastic

nature of the skelton can not remain

valid in the higher range of the

excess pore water pressure ratio

because, in this range, significant

amount of sand particles lose the

mutual contacts. Consequently, the

application of Eq.(l) should be

limited to the lower range of the

excess pore water pressure ratio. In

139
particular, if the application of

the Eq.(l) is limited to the range

of u/avJj<0.5, constant coefficients

of volume compressibility and

permeability can be used®

Comparison between the experimental

and the analytical results without

a gravel drain, shown in Figs.9 and

10, confirms the appropriateness of

the constants used for the

analysis.

Comparison between the results

with the gravel drain, shown in

Figs. 11 and 12 , indicates

acceptable applicability of Eqs.(l)

and (4)* Applicability of Eqs.(l)

and (4) is confirmed also for the

settlements as shown in Fig.l 3«

Consequently, the applicability of

the Eqs.(l) and (4) with the

pressure independent constants is

confirmed under the condition of

u/

Vo< 0.5« Therefore, fundamental

characteristics of the gravel

drains will be studied by using

Eqs.(l) and (4) under this

condition hereafter.

6. PERFORMANCE OF GRAVEL DRAINS

UNDER SINUSOIDAL CYCLIC LOADS

Eqs.(l) and (4) indicate that the

pore water pressure ratio u/aV(t, depends on the following

dimensionless parameters;

a/b : a ratio characterizing the

geometric configuration in

which a ,b : radii of the

gravel drain and the circle

of the effective spacing

shown in Fig.l

td/ti t a ratio characterizing the

duration of the earthquake

shaking in relation to the

duration of shaking required

to cause initial

liquefaction,

T1 ~ ( ks ,tl)/( I“rTw a2 )

* a factor charactrizing the

duration of earthquake

shaking required to cause

initial liquefaction in

relation to the consolidation

properties of the ground,

R = (8/ir 2)- (k s /kd

)- (h/a) ?

: a ratio characterizing the

resistance of the drain in

relation to the permeability

of the sand,

in which k^ : the coefficient

of permeability of the gravel

drain and h : height of the

; . ; : gravel drain,

and

a : a parameter characterizing the

shape of the pore water

pressure generation curve in

Eq. (4)

Among these parameters, there are

two parameters which were not used

by Seed and Booker [43- One is T]_

which will later be shown a more

appropriate parameter than (=

Ti t^/t^) which was used by Seed and

Booker. The other is R which was

derived by Xoshikuni and Nakanodo

[10] and applied for the analysis of

the gravel drains by Tanaka et al

[6]. There is another dimensionless

parameter characterising the

consolidation properties of the

gravel drain itself. This, parameter

has a very minor influence upon the

value of the pore water pressure

ratio because the time scale of

consolidation of gravel drain itself

is very small in comparison with the

time scale of pore water pressure

generation t^. Therefore, this

parameter will not be considered.

Furthermore, the parameter t^/t]_

will be shown to have very minor

influence upon the maximum value of

the pore water pressure ratio for

most of the cases encountered in

practice. This fact will be shown at

first by using a typical solution of

Eqs.(l) and (4)» It is noted in

Fig.l 4 that, when the maximum value

of u/aVQ is less than about 0.5,

value of u/ avo reaches its steady

state value before t = tq (=0.5t (j).

Therefore, the maximum value of

140
u/Ov<3 is determined only by Tq and

that (u/avy max is not influenced

by the value of t^/tq if t^/tj_>1 .0.

Similar conclusions can be drawn

for various values of a/b, T]_, and

R. As shown in Fig. 15, the excess

pore water pressure ratio reaches

90% of the steady state value of

u/av& at the time tgo» which is

less than 2t^_ if (u/ o^q) max^^*^*

Consequently, the parameter t^/tq

has a very minor influence upon the

value of (u/cJvymax f°r mosi of the

cases encountered in practice

because, for most of the cases,

duration of earthquake shaking is

greater than about 2tq.

Consequently, only four

dimensionless parameters determine

the maximum value of the pore water

pressure ratio; a/b, Tq, R, and a. Therefore, the maximum excess pore

water pressure ratio averaged over

the horizontal section at the

middle level of the ground

(u/avy max can be shown in compact

charts as in Fig.l6. Conditions

used for the computation for Fig.16

are a = 0.7 without vertical

permeability in the ground. The

computation was carried out by

finite element method.

7. DESIGN OF SPACING BETWEEN THE

GRAVEL DRAINS AGAINST EARTHQUAKES

Once the simplified solutions shown

in Fig. 16 have been obtained,

spacing between the gravel drains

can be determined by the procedure

illustrated in Fig.17. As indicated

in this figure, there are two

important items, besides the

constants of the soils, to be

provided for determining the

spacing; the value of (u/ Ov

'Q) max

and the earthquake ground motions.

The value of (H/av £) max can be

determined by the stability

analysis. However, if the gravel

drains are installed in a loosely

deposited ground, detailed

stability analysis for determining

the value of (u/oV(^) max becomes

practically meaningless, because

very small errors contained in the

liquefaction resistance curve and

the shear stress time history has

very large influence on the

generaion of excess pore water

pressures in the ground. It is much

more important to give the accurate

liquefaction resistance curve and

the appropriate shear stress time

history.

The value of (u/av<y max apart, the

earthquake ground motions observed

during earthquakes are very

different from each other. Even for

the earthquakes of the same

magnitudes, there is a great

variation in effective duration of

the earthquake motions in relation

to liquefaction. This fact was

revealed by the study by Tokimatsu

and loshimi (1980) [7]. They define

a quantity "c" as a ratio of

"cumulative number of equivalent

cycles at time t

n against " equivalent number of cycles for a whole time history" and, using this

quantity, define the effective

duration of earthquake motions by

the following equation:

t e ~ 2(t c=Q # 5 - ^ 2 = 0 . 1

) (5)

in which

tQ : the effective duration of

earthquake motions

tc=0.5 : time at which c = 0.5

tc-0.1 : time at which c = 0.1

Similar study has been carried out

here by using the strong-motion

earthquake records obtained by the

network of the Port and Harbour

Research Institute [9] and redifing

the effective duration of earthquake

motions tq by the following

equation:

tq = 2.5(t c=Q # 5 - "kc=0.l) (6)

Here, the factor 2.5 (= 2/0.8) was

used instead of the factor 2 in

Eq.(5) because the factor 2 is

141
considered as a factor tempered by

the engineering judgement.

The results are plotted in Fig. 18.

These results indicate that there

is a trend of increase in the

effective duration .of the

earthquake motions in accordance

with the increase in magnitude of

the earthquake. However, the

scatter contained in the data is so

great that it is almost impossible

to use the average value of the

data if a reliable design is

required. For example, if the

target earthquake for the design of

the gravel drains is of the

magnitude of 7.5 or 8, and very

reliable design is required, then

the wave form of NS component of

the strong earthquake motion

recorded at Hachinohe Harbour

during the 1968 Tokachi-Oki

Earthquake (M = 7-9) [8], may be

considered one of appropriate

target earthquake motions, because

this wave form has rather short

effective duration for liquefaction

as shown in Fig. 18, in which this

earthquake motion is designated as

8-252 NS. Ishihara and Tasuda

called this wave form n shock type"

in the study of liquefaction under

irregular loading conditions [2].

It is also widely used for the

earthquake resistant design of the

structures in Japan.

Once the wave form of the target

earthquake is determined, the time

history of shear stress ratio is

computed by the ground response

analysis and the rest of the

procedure to obtain the value of tq

is straight forward. At first, the

cumulative damage t/t

q

= N/Nq Is

calculated from the time history of

shear stress ratio and the

liquefaction resistance curve [1],

Secondly, the value of tq is

calculated by the following

equations

tq = 2.5(t t /-t^ =0 .5 - tt/

1^ =0.1

in which tt/t^o.s :

H/t^=0.1 :

time at which

t/tq = 0.5

time at which

t/tq =0.1

Thirdly, ' influence of the shape of

the pore water pressure generation

curve is taken into account by

redefining the value of tq as

follows %

' ' '

tq(redefined) = tq(N1 /Nq)/(N0/Nq)

. ( 8 )

in which (N1 /Nq) and (NO/Nq)

are, as shown in Fig.19, cummulative

damages corresponding to (u/ ovQ m£LX

obtained from the experiment and the

assumed (i.e. Eq.(4) with a=0.7)

pore water pressure generation

curves.

The rest of the procedure shown in

Fig. 17 with Fig.l6 determines the

value of a/b.

As an example of the spacing between

the gravel drains, consider the case

where (u/av&) max = 0.2, k s = 3x1 0“ 2 cm/s, m v = 3x10~ s m

2 /kN, k^ = 10

cm/s,"a = 30 cm, and h = 10 m, and

consider that tq has been already

obtained as five seconds by the

above-mentioned procedure. Then, Tq

= 55*- and R = 2.7, and from Fig.l6,

a/b is determined as 0.35 and,

therefore, the radius of the circle

of the effective spacing is

determined as b = 85 cm.

8. CONCLUSIONS

The large scale model tests and the

analysis of gravel drains lead to

the following conclusions:

(l) The kind of consolidation

equations used by Seed and

Booker [ 4 ] has acceptable

applicability in the lower

range of the excess pore water

pressure ratio (the ratio of

excess pore water pressure

against the vertical effective

142
stress). Constants for those

equations does not change

regardless of the change in

the excess pore water pressure

if the excess pore water

pressure ratio is less than

0.5.

(2) The maximum value of pore

water pressure ratio for the

gravel drains under sinusoidal

cyclic loadings coincides with

the steady state value of the

pore water pressure ratio for

most of the cases encounted in

practice. This fact simplifies

the design procedure for

determining the spacing

between the gravel drains.

(3) Resistance of the drain should

be considered for designing

the spacing between the gravel

drains

(4) The earthquake ground motions

used for designing the spacing

between the gravel drains

sould be carefully chosen.

ACKNOWLEDGEMENT

The authors wish to express their

thanks to Mr. K. Mineta, Hokkaido

Kaihatsu Consultant, and Mr.

T. Yamamoto, Penta Ocean

Construction Co., for their works

in the experiments and the analysis

during their stay in the Port and

Harbour Research Institute as

visiting researchers. The authors

also wish to express their thanks

to Mr. M. Kobayashi, Chief of Soil

Mechanics Laboratory, Port and

Harbour Research Institute, for his

works for developing the finite

element program.

REFERENCES

1. Annaki,M. and Lee, K.L.(1977)

: "Equivalent uniform cycle

concept for soil dynamics,"

Journal of the Geotechnical

Engineering Division, ASCE

Vol.103, N0.GT6. pp. 549-564

2. Ishihara, K. and Yasuda, I.

(1975) : "Sand liquefaction in

hollow cylinder torsion under

irregular excitation," Soils

and Foundations, Vol.15, No.1

,

pp. 45-59

3. Ohara, S. and Yamamoto. T.

(1986) : "Gravel pile method

for preventing liquefaction,"

Proc. 7th Japan Earthquake

Engineering Symposium, pp.727-

732

4. Seed, H.B. and Booker, J.R.

(1977) : "Stabilization of

potentially liquefiable sand

deposits using gravel drains,"

Journal of the Geotechnical

Engineering Division, ASCE,

Vo. 103, No.GT7, pp. 757-768

5. Scott, R.F. (1986) : Solidification and consolidation of a liquefied sand column," Soils

and Foundations Vol.26, No. 4,

pp. 23-31

6 . Tanaka, Y., Kokusho, T.,

Esashi, Y., and Matsui, I.

( 1984 )

• n 0 n preventing

liquefaction of level ground

using gravel piles," Proc.

Japan Society of Civil

Engineers, Vol.352 (III-2),

pp.89-98 (in Japanese)

7. Tokimatsu, K. and Yoshimi, Y.

(1980) : "Effects of vertical

drains on the bearing capacity

of saturated sand during

earthquakes," International

Conference on Engineering for

Protection from Natural

Disasters, Bangkok, pp.643-655

8 . Tsuchida, H., Kurata, E., and

Sudo, K. (1969) : "Strongmotion earthquake records at

the 1968 Tokachi-Oki Earthquake

and its aftershocks," Technical

Note of the Port and Harbour

Research Institute , No. 80

9. Tsuchida, H. et al. (1966-1986)

: "Annual report on strongmotion earthquake records in

Japanese ports," Technical Note

of the Port and Harbour

Research Institute, No. 55-

No. 51 9

10. Yoshikuni, H. and Nakanodo,

H. (1974) : "Consolidation of

143
soils by vertical drain wells

with finite permeability/®

Soils and Foundations, Vol.14»

No«2, pp. 35-46

144
Photo 1 A stack of 64 aluminum rings on the shaking table

145
Percentage

Finer

by

Weight

Table 1 Conditions for the experiments

Case Number R-204 R-302 R-303 R-502 R-503

Relative Density (%) 29 31 31 33 35

Acceleration of

Shaking Table (Gal) 43 50 81 50 69

SS5

S8 6 O

t

AV2 TM

**i- “ A*H2 =3- . . A

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i it*— ©x—I —© Direction of Vibration *o ££

1151151151151 125 I 60 - I 35 J

Cross Section

Fig.l Schematic figure

of gravel drains

100

80

60

40

Grain Size (mm)

Fig. 2 Grain size accumulation

curves of the sands and

gravels

10 190 Unit : cm

^ Mtir (L) Icdlfrovur & (Honowull (A) Aco>t' wtr

A (r»<nu I ! (A) >or» «tr

• Prottur* t*u<**(M _ EprtSb ® Gavqb («

(S)(SS)

It (H)

Fig. 3 Location of instruments

deployed in the model

- /

I

-

- / // GRAVEL

/

I -

AKITA i 1 ' SAND | j

1 for DRAIN- -

1

-

- h

-

w> ©

(3 £ 0.05 o

95 0 Radial Distance (cm)

Fig. 4 Maximum

excess

pore water

pressures

(without a drain)

rGRAVEL

DRAIN

95 30 0 Radial Distance (cm)

Fig. 5 Maximum

excess

pore water

pressures

(with a drain)

146
Fig. 7 Time histories of the excess

pore water pressures at P24

(during the 240 seconds)

Fig. 9 Comparison between the

measured and the computed

time histories for excess

pore water pressures

(without a drain)

(gf /cm2

)

Fig. 10 Comparison between the

measured and the computed

maximum values for excess

pore water pressures

(without a drain)

147
150

o

CL

coCO

CD

CDx

LLJ

150

e

o

ai D

CO 05 £ ~

a>

<13 b:

I o o Measured P—24

— Computed P—24

a a Measured P—23

—

Computed P—23

° 0 Measured P—22

—

Computed P—22

0)

!_ (TO 5|cn ^

CD

u. \

CL CD

With Gravel ~ Dram

Time (sec)

Fig .11 Comparison between the

measured and the computed

time histories for excess

pore water pressures

(with a drain)

100

50

° ° Measured Level—

Computed Level—

A & Measured Level—

Computed Level—

Measured Level—

—— Computed Level—

With

Gravel

DrainD "g-Xo

- -""X 5

’

' C ravpl 1 Sand Drain

95

Radial Distance

30

(cm)

Fig. 12 Comparison between the

measured and the computed

maximum values for excess

pore water pressures

(with a drain)

CDi_3

co

CO

CD £L| O

0.5 03 ^ 03 o a: u

o

CL

_co I ? 0.5 t o

0) —

co 1

0

With

Drai

Gravel

n

® Measi

Comp

S Settle Measi

jred

uted

ment

jred

Surface -

ment

ained

. •

// ® \

/• \

r

at the A Settle by Dr Water

i © ©

U/<TU0

0.5

25 50 75

Time (sec)

100

Fig o 13 Comparison between the

measured and the computed

settlements and excess

0

td/tr

a/b

2

= 0.2

' T2 =50^-

1

1

1

1

\

l\

1 \

1 \

100

1 1 1 0.1

t / t i Fig. 14 Typical excess pore

water pressure change

pore water pressure ratio

(with a drain)

148
"t 90 T7

R = 0.01

(a) R = 0.01

R = 3

Drain Pile Radius Ratio a/b

(b) R = 3

Fig. 15 Duration of shaking required to

water pressures

attain the steady state pore

max

R=!

(a) R = 0.01 (d) R = 1

R = 3

(b) R 0.1 (e) R = 3

(c) R = 0.3 (f) R = 10

Fig. 16 Relation between maximum pore water pressure ratio, time factor,

well resistance, and radius ratio

149
Fig, 17 Procedure for determining

the spacing between the

gravel drains

Fig. 18 Relation between the

magnitude of the earthquake

and the effective duration

of earthquake motions

u/cr vo

Fig. 19 Schematic figure for obtaining correction factors for pore water

pressure generation curve

150
Experimental Study on Degree of Fixation at Pile Head Joint and

Failure Strength of Prestressed High Strength Concrete Piles

by

YOSHIHIRO SUGIMURA

1

, TSXJTOMU HIRADE

2

ABSTRACT

This paper discusses experimental result of

full-scale bending-shear tests for

pile-footing joint under various axial load

conditions, in order to investigate degree of

fixation at pile head joint and ultimate

strength of piles. As test specimens, 29 sets

in total of prestressed high strength concrete

(PHC) piles are used by combining three grades

of effective prestress (c =40, 80, 100kg/cm 2

) e and two kinds of diameter (35cm, 50cm) with

six types of joint method. Statically

indeterminate loading system is used, i.e„,

concentrated load is applied at intermediate

point between pile head joint and the other

pinned supporting point of pile for producing

simulated bending moment distribution of pile

in the ground by horizontal load.

Conclusions derived from experimental

results are classified into the following two

categories :

1) Degree of fixation at pile head joint a

in the range of allowable capacity design is

affected significantly by embedding length of

pile into footing and axial load. When the

ratio of embedding length to pile diameter

changes from 0.2 to 1.0, increases from 0.7

to 0.9, and from 0.4 to 0.75 in average, under

the conditions with and without axial load,

respectively.

2) Final state of PHC pile results in shear

failure type, especially in axial load

condition, so that it is important to make an effort to keep its ductility in the ultimate

capacity design.

Keywords : Bending Shear Test of PHC pile,

Fixation Degree at Pile Head, Shear Failure,

Aseismie Design of Pile

1. INTRODUCTION

In the earthquake resistant design of pile

foundation, it is very important to accumulate

information about degree of fixation, i.e.

,

degree of restraint for rotation at pile head

and ultimate strength of pile. Theoretical

solutions of horizontal resistance of pile are discussed by introducing the concept of degree

of fixation at pile head and effect of pile

length in the previous paper.

1 In succession,

this paper discusses experimental results of

full-scale bending-shear tests for joint

between prestressed high strength concrete

pile and footing under various axial load

conditions, in order to investigate degree of

fixation at pile head joint and ultimate

strength of piles. Some parts of these

experiments were already discussed in the

previous papers

2 ’ 3

1) Dr. Eng., Head, Geotechnical Engineering

Division, Structural Engineering

Department, Building Research Institute,

Ministry of Construction, Japan

2) Research Assistant, Geotechnical

Engineering Division, Structural

Engineering Department, Building Research

Institute, Ministry of Construction, Japan

151
2. PREVIOUS STUDIES ON PILE HEAD JOINT UNDER

HORIZONTAL FORCE

In the design method of horizontal

resistance of pile foundations it is an important problem of selection that the

boundary condition of rotation at pile head is

adopted as rigid joint or pin joint. Stress

becomes large but displacement is small at

pile head in the case of rigid joint. On the

other hand, displacement is large but stress

becomes small at pile head in the case of pin

joint. Therefore, by selecting either of the

two, structural designers bring the detail of

joint near to the adopted condition in

practical design.

Experimental studies on pile head joint

under horizontal force have been performed

from more than ten years ago,

4-' 6 and it has

been made known that rigidity and strength of

pile head joint becomes large for the case of

rigid joint. Since the circular notice from

the chief of Planning section, Road Bureau,

Ministry of Construction was presented in May

1972, and specified that piles and footing

should be fixed in principle, design

directional to rigid joint has been

constituting the main current in the field of

road bridge. T. Asama and Y. Adachi explains

that the reason why rigid joint is adopted is

the most common method in the field of public

works and also displacement is an important

factor for the case such as bridges, 7,3

although this design guide does not always

prohibit pin joint. Therefore, studies on structures in public works, 9 ~17 wholly in some

sense, have been carried out with the

intention of rigid joint based on this guide.

In the field of structural design of

building foundation, there are some research

works according to this guide, 18 "* 20 but

philosophy such as design directional to pin

joint or design not sharing horizontal force

to piles still remains. This situation shows

not only variation of design philosophy of

pile head joint but indef initeness of relation

between design detail and real behavior of

joint. However, taking the opportunity of

discovery of earthquake damage of PHC piles

due to the Miyagiken-oki Earthquake in 1978, a new look is taken at pile head joint as an important problem. According to the results

of survey by the authors, 21 * 22 earthquake

damage type of piles is estimated to be near

to rigid joint condition rather than pin joint

condition expected in design process, so that

condition of pile head joint is an important

parameter on earthquake damage of piles. And

also, in the supplementary design guide of

circular notice from the chief of Building

Guidance section, Housing Bureau, Ministry of

Construction presented in September 19 84,

calculation is recommended to get horizontal

resistance of pile by considering degree of

fixation at pile head. 23 Under such

background, experimental studies have been

performed actively to investigate degree of

fixation at pile head and ultimate strength of

piles in the field of building foundation, 24 ~

44 and a trial is made to examine

synthetically the past papers on degree of

fixation and strength of pile head joint. 45

There appears also in the field of public

works, an proposal to make design method more rational by introducing effects of angle of

rotation at pile head. 46

3. DEGREE OF RESTRAINT FOR ROTATION AND

DISPLACEMENT FACTOR AT PILE HEAD

The term of degree of fixation at pile head

has been often used as the concept showing the

rigidity of pile head joint. This is a parameter which shows ratio of fixation of a certain joint at pile head to the case of

rigid condition, and is divided broadly into

two categories, i.e., degree of fixation to bending moment and degree of fixation to displacement. Here, let examine, the

substance of degree of fixation at pile head,

152
by taking as an example, pile of infinite

length of which theoretical solutions were

derived in the previous paper.

1 Table 1 shows

various degrees of fixation at pile head for

the case that boundary conditions are given as bending moment and angle of rotation 0^

with known value of inverse of rotational

spring k. The numbers in each row in Table 1 show the corresponding references which make

mention of degree of fixation based on experimental results. References of the row

c) do not always use the same definition

expressed in the table, but are classified

into the same group because of similarity of

phlosophy to degree of fixation.

According to Table 1, degree of fixation to

bending moment can be determined directly

from the boundary condition at pile head and

changes from 0 to 1 when pile head joint

changes from pin condition to rigid condition,

while degree of fixation to displacement R^q

is a fuction including changing from 2 to

1. Besides, if piles are infinite length, the

value of R^

0 changes extremely. 1 Then, and

yO

are called as degree of restraint for

rotation and displacement factor in strict

sense, respectively, and distinguished each

other in this paper at need.

The reason of a large number of the

references corresponding to the rows c) and d)

in Table 1, is related to the fact that

cantilever type tests are adopted as the

simplest experimental method and investigated

test results attaching importance mainly to displacement because rigid joint has been the

main current. These are classified into the

group of degree of fixation at pile head to displacement, i.e., displacement factor. As

mentioned above, it is reasonable to distinguish degree of fixation in a narrow

sense, i.e., degree of restraint for rotation

from displacement factor.

On the other hand, there are some cases in

which degree of restraint for rotation is

determined by the definition shown in the row

b) based on bending moment from cantilever

type tests. But, in this case, it needs the

aid of coeffeicient B concerning ground

condition. 1 It is reasonable to get degree of

restraint for rotation directly from

laboratory tests on pile member not through

condition of soil deposits. Therefore, the

best way is to use the definition of the row

a) in Table 1 determined directly from bending

moment through staically indeterminate

experiment. In the following chapters, only

the difference of rotational restraint

condition at pile head is discussed by

assuming the rotation of footing or foundation

beam does not exist.

A. OUTLINE OF EXPERIMENTS

A . 1 TEST PROGRAM

Statically indeterminate loading system is

used as shown in Fig. 1. The reason is to adopt the method used continuously by the

authors, 22 which is capable of getting the

degree of fixation directly from bending

moment at pile head, and of keeping the

similarity of bending moment distribution of

pile in the ground due to horizontal force, as mentioned in chapter 3. The loading points B,

C and the supporting point A are treated as free for rotation by using Teflon lubricant

device support. Footing is fixed to the floor

of test room by using the two PC bars.

It is advisable to use alternating cyclic

loading system essentially, but one direction

cyclic loading system is adopted in these test

series, by considering the necessity to avoid

the influence of test device used in the

previous paper,

22 or difficulty of turnning

test specimen round.

Each value of loading step P at the loading

point corresponding to allowable, design crack

153
and design failure bending moment of pile, is

calculated for the case that pile head joint

is assumed to be rigid joint. And, in

principle, after repeating 3 cycles, 5 cycles,

and 1 cycle at these loading steps, load is

increased to the final failure state.

Experiments are performed in two phase test

series. In each case, preliminary load

corresponding to about one third of allowable

load is acted firstly, and after confirming

stability of test specimen, main test is

started.

4.2 TEST SPECIMENS

Outline of test specimen is shown in Fig.

2, for the case of 20Z*. Structural

properties of pile section in phase 1 test

series are shown in Table 2 and Fig. 3. And

strength of concrete of each specimen and

detail of pile head joint are shown in Table 3 and Fig. 4, respectively. Test specimens, 24

in total, are selected by combining pile

diameters of 35cm and 50cm, pile grades of A

(effective prestress a

e=40kg/cm 2

) and C (effictive prestress a =100kg/cm 2

) , and axial

e loads of 0, 30 and 60t.

The detail of joint method in phase 1 test

series are as follows. Type X is the most

common method that pile cut together with PC

bars is buried simply in footing with 10cm

depth, and plain concrete is filled inside of

pile up to length corresponding to pile

diameter. Type Y is the same method as type X

except that reinforced concrete is filled

inside of pile with a little longer length.

Type Z is the method that pile cut leaving PC

bars 50 times as long as those diameter is

buried in footing with 10cm depth, and covered

by steel pipe with concrete outside and filled

with plain concrete inside of pile to 40cm

long from pile top.

The types of test specimen and strength of

materials in phase 2 test series are shown in

Table 4 and Table 5, and structural section of

pile and detail of pile head joint are shown

in Fig. 5 and Fig. 6. In this case, piles in

35cm diameter of grade B (effective prestress

O

g=80kg/cm 2

) are used for supplement of phase

1 test series.

The piles were manufactured with separater

at the pile end and cut the outer part of PC

bars after curing, in order to estimate

relaxation of prestress due to cut off of pile

head at the job site. Change of prestress

during this process was measured by strain

gages set up on PC bars and these measured

values are listed in Table 4 as the measured

effective prestress. Axial load condition is

30t in all the cases.

Similar three types of joint method, i.e.,

X* , Y' and Z' are selected in phase 2 test

series, but by using three different embedding

length in order to investigate effects of

embedding length of pile head in footing. The

type X’ is the same method as type X in phase

1 test series, but two kinds of embedding

length are selected, i.e., 10cm and 35cm,

which are called as test specimens 10X 1 and

35X' , respectively. Type Y' is the method

pile head in footing is reinforced by rising

steel bars and spiral hoops with embedding

length of 20cm which is called as test

specimen 20Y’. In this case, reinforcing cage

is not unified with pile. Type Z' is the

similar method as type Z in phase 1 test

series, but steel pipe (inner diameter : 35. 4cm,

length :35cm, thickness :0.6cm) is glued to pile

with epoxy resin, and 8 anchor steel bars of

35cm bond length (D10-8) are welded to steel

pipe. Embedding length are selected as 10cm

and 20cm which are called as test specimens

10Z' and 20Z', respectively. In all the

specimens in phase 2 test series, plain

concrete is not filled inside of pile.

154
TEST RESULTS

The size of footing and ratio of

reinforcement is basically common in phas? 1 and phase 2 test series, except that tbe n.gh.r

steel bars are bended downward and binder to steel pipe pile for the case of Z' type, aid are bended upward around pile surface for the case of Y' type. The shortest distance from

the center of pile section to the side of

footing is about 1.7 times as long as pile

diameter in the case of 35cm pile diameter,

and 1.2 times in the case of 50cm pile

diameter.

4.3 MEASUREMENT PROGRAM

Xhe following items are measured in cnroon with phase 1 and phase 2 test series.

1) loads at loading points B and C, and

reaction force at supporting point A for

determination of bending moment distrihunnn

of pile

2) displacements at loading points B am C, at supporting point A and at pile head jent E

for getting of deflection of pile

3) angle of rotation at pile head

4) angle of rotation of footing

5) strains at pile surface for checking of

bending moment distribution of pile

In addition to the above, in phase 1 r=st series, rosette gages are used to measmre

three directional strains in pile. And a_so

,

in phase 2 test series, measurement points are

increased as many as possible, such as displacement for deflection of pile, scrams

of PC bars for getting bending moment

distribution of pile, mold gages and

strainmeters of steel bars set in footinr to

t’neck behavior of joint part, axial load

t'nange during test, and strains of. PC bams

:ixing footing to the floor of test room. As

m example, measurement location of the se

>f specimen 20Z' in phase 2 test series it hown in Fig. 7.

Figure 8 shows envelope curves of the

na-ri mi up displacement of pile at virgin loads

in phase 2 test series. The maximum

displacement of each case is shown by the

avanage of the measured values of the dial

rages on the upper surface (D3) and the lower

surface (Dll) of pile as shown in Fig. 7. Figure 9 shows an example of deflection of

tile and strain distribution of pile at each

loading step.

It is noticed from these figures that the mum displacement occurs at the location a little to supporting point A from loading

point C and the slight displacement is also

observed both at supporting point A and pile

head joint E. Initial rigidity in the range

of small load is the largest for the case of

35S’ in which embedding length is equal to IB

(B spile diameter) and becomes small as embedding length is short. The maximum

strength is the largest for the case of 35X'

,

anc becomes smaller in the order of 10Z'>10X'

>20Z'>20Y' within the range from 27. 3t to 34. 9t. And the corresponding displacement of

pile is scattering in the range from 2 to 4cm,

changing smaller in the order of 35X'>10X'>

10Z' >20Y ' >20Z '

Figure 10 shows the relation between

bending moment and displacement at loading

point C, and Fig. 11 shows the relation

between bending moment and rotation of angle

at pile head joint E. Rotation of angle at pile head is calculated from the quantity of

pullout at the upper side and push in at the

lower side of pile at the end of footing

measured by the dial gages (D15, D16) , as shown in Fig. 7. But almost all part is

consisting of pullout at the upper side and

quantity of push in at the lower side is very

small. From these figures, linear relation

between bending moment and displacement is

observed until cracks occur in pile member.

155
As for rotation of angle at pile head 0 , the

followings are recognized as characteristics

around the allowable bending moment of pile

member

1) Under the condition of axial load, 0^

becomes small*

2) As for the effect of joint method, the

tendency that 9^

becomes larger in the order

Z, Y, X is recognized from the results of

phase 1 test series. But in some cases, the

order is reversed X and Y. From the results

of phase 2 test series, it is remarkable that

6^ becomes smaller as embedding length is longer.

3) Wien the pile diameter is large, 0^

becomes larger.

4) As for the grade of pile, 0^ becomes

larger in the order C, B, A.

6. DEGREE OF FIXATION AT PILE HEAD AT THE

ALLOWABLE BENDING MOMENT OF PILE

Figure 12 shows the comparison of the

degree of fixation at pile head in each

test when the bending moment of pile reaches

its allowable value. The abscissa is the

ratio between embedding length of pile into

footing and pile diameter a,, and the part of

d

outer steel pipe pile is included in the

embedding length for the case of joint type Z. The ordinate is the degree of fixation at pile

head, which is corrected to avoid the effect

of rotation of footing 0 and displacement at

r the point E and A by the following equations

(see Fig. 13)

a

r

( 1 )

«r

' *£ * “e

• «E - W (te F + ?E

- ?A>

(2)

«Ef ’ I «*i - *» T? V <3)

in which,

: bending moment at point E determined

from the measured values of the loads

P^, ?2 anc^ the reaction force at supporting point R

A

AMj.: influence component to the bending

moment at point E determined from the

measured values of the angle of

rotation of the footing 0p, and the

displacements y and y at points E A

and A

M^: theoretical bending moment at point E corresponding to the conditions of pin

support at point A and fixed joint at

point E El : bending rigidity of pile

a,b,c,£ : length of each part of specimen

shown in Fig. 13

The following characteristics are derived

from Fig. 12.

1) Embedding length of pile into footing is one of the most important influence factor on degree of fixation at pile head a , and

becomes larger as embedding length is longer.

When the ratio of embedding length to pile

diameter a, changes from 0.2 to 1.0, a

d r changes from the range of 0. 3-0.5 to the range

of 0.7-0. 8 under the condition without axial

load, and from the range of 0.6-0. 8 to the

range of 0. 8-1.0 under the condition with

axial load. If a, is nearly equal to 1,

d

condition of pile head becomes similar to rigid joint.

2) The range of due to the condition with

or without axial load has width of about 0 . 2 , respectively, but can be divided clearly into

the two groups, i.e., the lower region without

axial load and the higher region under the

influence of axial load.

3) Degree of fixation at pile head is

influenced by the condition if anchor steel

bars are unified with pile head or not, even

in the case of equal embedding length. If

both are welded, degree of fixation becomes

large by the bonding effect of anchor steel

bars
7. FAILURE TYPE AND ULTIMATE STRENGTH OF PILE

Test results at the maximum load and

failure type of each test are summarized in

Table 6, and the final states of test specimen

are shown in Fig. 14 and Fig. 15. In almost

all the cases, final state results in shear

failure of pile except the cases of pile of

grade A without axial load which are determined by bending failure (break of PC

bars) , and also followed by tubular

deformation due to circular bending moment, if

axial load is large or diameter of pile is

large. This phenomenon should be noticed as a specific character of tubular section member

,

even if this kind of phenomenon is hard to occur in the case of piles in the ground

because of confining effect by surrounding

soil deposits.

As for footing, comparatively large cracks

develop in the case of short embedding length

but small or no cracks in the case of long

embedding length. And, if embedding part is

reinforced by spiral hoops, effect of

preventing cracks of footing is exhibited.

Figure 16 shows the comparison between

experimental values and analytical values of

shear strength of pile. Analytical values are calculated by the following equations, based

on the measured values of pile section and

strength of materials.

.Q cal cr -f1 /O + 2<t>0 ) g z V (4)

a

g

(5)

in which,

t : thickness of pile (cm)

Sq : geometrical moment of area of pile

O expressed by s = — (r 3 - r

3 ") 0 3

V 0 V

I : geometrical moment of inertia of pile

expressed by j = L ( r

* _ r

w

)

N : axial load (kg)

9

A

g

: equivalent area of pile section (cm 2

) o : effective prestress (kg/cm 2

) e c

c

: tensile strength of concrete assumed

as 7% of compressive strength (kg/cm 2

)

Tq : outside diameter of pile (cm)

r^ : inside diameter of pile (cm)

4>

‘ correction factor of shear force

assumed to be equal to 0.5

The mean value R of ratios of experimental

values to analytical values and standard

deviation a are 0.981 and 0.116, respectively,

as shown in Fig. 12. It should be noticed

that eq. (4) overestimates shear strength of

pile in some cases, because analytical values

are larger than experimental values. Y. Gotoh

and T. Shibata pointed out the accuracy of eq.

(4) is slightly inferior to the experimental

equation developed by them for estimation of

shear strength of reinforced concrete

column. 4 7

Table 6 shows also the ratios RBM between

experimental results and analytical results on the ultimate bending moment of pile. As for

M in phase 2 test series, all the parameters

are the measured values except that tensile

strength of concrete of pile is assumed as 7%

of compressive strength. In the process of

getting M in phase 1 test series, catalogued

values or initial condition of test are used

for the parameters such as yield stress of PC

bars, effective prestress and axial load

change during test, because those parameters

were not measured. It should be noted,

therefore, that the accuracy is different in

both test series and fairly high values of RBM

are observed especially in the case of pile

grade A without axial load condition. But, on

the whole, the values of RBM less than 1 are recognized in almost all the cases, i.e.

,

shear failure of pile precedes the occurrence

of final bending failure. This tendency is more remarkable for the case with axial load

157
condicion than the case without axial load

conditionThe importance of design philosophy to

avoid the preceding of shear failure of pile

will be increasing more and more in earthquake

resistant design of pile foundation.

Therefore, it is advisable to design leaving

sufficient surplus of shear strength of pile,

if the eq. (4) is adopted in design process.

8- CONCLUSIONS

Results of bending shear test for pile head

joint of prestressed high strength concrete

pile are discussed in this paper. The

conclusions applicable to practical design of

pile foundation derived from these tests are

as follows.

1)

The important influence factors on degree of fixation at pile head in the

allowable capacity design of pile foundation,

are the embedding length of pile into footing

and the condition with or without axial load,

as shown in Pig. 12. When the ratio of

embedding length to pile diameter a, changes

a from 0.2 to 1.0, a j_nCreases from 0.7 to 0.9,

and from 0.4 to 0.75 in average, under the

conditions with and without axial load,

respectively, forming increasing curves about

4.2 in width as axial load increases. If pile

is embedded to the length corresponding to

pile diameter, behavior of pile head joint

becomes similar to rigid joint. And if axial

load acts, degree of fixation at pile head

becomes large entirely. During an earthquake,

piles are acted by alternating axial load due

to rocking motion of superstructure from the

initial condition supporting long term load

due to the weight of superstructure.

Therefore, degree of fixation of piles located

at compression side becomes larger than piles

located at pullout side, and shear force is

apt to be concentrated to piles in compression

side. This effect of axial load will become

very important factor for design method of

pile foundation.

2) Condition if anchor steel bars into

footing are connected with pile or not, is

also an important factor on degree of

fixation. If anchor bars and pile unified,

becomes larger than the case that both are not

unified in the range shown in Fig. 12.

3) The ratio is nearly equal to or more

than 0.3- and is far from pin condition in any case of joint type used in this study. If

pile head is expected as pin condition, it is

necessary to develop a certain special joint

method separaterly from these joint types.

4) The final state of each test specimen,

except some of the piles in grade A without

axial load results in shear failure type.

Shear span ratios in these tests are scattering in the range from 1.43 to 3.20.

This result will be possible to effect

ratification of the conclusion of the

preceding paper

22 that the boundary of shear

span ratio at which shear failure of PHC pile

occurs, is nearly equal to 3 for the cases of

piles of grade B and C.

Although real shear span ratio of pile in

the ground under horizontal force is necessary

to investigate separately, this result proves

that the examination of shear strength of pile

is indispensable factor in earthquake

resistant design. And eq. (4) based on elastic theory is apt overstimate the real

shear strength of pile. Therefore,

improvement such as increase of lateral

reinforcement is important for this kind of

pile. At least, it is advisable to design

leaving sufficient surplus of shear strength

of pile, if eq. (4) is adopted in design

process.

158
ACKOWLEDGEMENTS

The experiments in this report were carried

by Building Research Institute under the

financial and technical support provided the

Committee of Pile Head Joint Method (Members

of Y. Itoh, K. Imaiznmi , K. Inbe, 0. Kobayashi, M. Hiraike, M. Fujisawa, N. Mastubara and the authors, phase 1 test series

in 1984 to 1985) and the Research Committee of

Pile Head Joint Method (Members of K. Ohshima,

J. Ohya, T. Okabe, A. Kubota, S. Hayama, M. Yasui and the authors, phase 2 test series in

1985 to 1986) establised in Japan Association

for Building Research. Promotion. The authors

would like to express their thanks to the

members of the Committees for the considerable

support.

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1982.7

27) T. Ozawa et al. : Strength and

Deformation Characteristics of Prestressed

High Strength Concrete Pile Connected with

Footing (in Japanese) , Summaries of Technical

Papers of Annual Meeting Architectural

Institute of Japan, pp. 1749-1750, 1982.10

28) T. Fukushima et al. : Experimental Study

on Bending Characteristics of Pile Head Joint

of PC Pile (in Japanese) , Summaries of

Technical Papers of Annual Meeting

Architectural Institute of Japan,

pp. 2145-2146, 1982.10

29) S. Yamato and M. Takagi : Bending

Characteristics of SC Pile Connected with

Footing (in Japanese) , Summaries of Technical

Papers of Annual Meeting Architectural

Institute of Japan, pp. 2147-2148 , 1982.10

30) M. Ohta et al. : Bending and Pullout

Tests of Pile Head Joint (in Japanese)

Summaries of Technical Papers of Annual

Meeting Architectural Institute of Japan,

pp. 2149-2152, 1982.10

31) H. Satoh et al. : Bending Property of

PRC Pile Connected with Footing (in Japanese)

Summaries of Technical Papers of the 37th

Annual Meeting Japan Society of Civil

Engineers, pp. 479-480, 1982

32) I. Yamamoto et al. : Ultimate Strength

of Joint between Precast Concrete Pile and

Footing (in Japanese), Jour, of Fujita

Technical Research Division, No. 19,

pp. 115-127, 1983.6

33) Y. Ohnishi et al. : Ultimate Strength of

Joint between Pile and Pile-cap on Lateral and

Axial Load (in Japanese) , Summaries of

160
Technical Papers of Annual Meeting

Architectural Institute of Japan,

pp. 2727-2730, 1983.9

34) K. Kokushoh et al. : Studies on the

Behavior of Pile-Footing Connections under

Axial and Lateral Forces (in Japanese)

,

Summaries of Technical Papers of Annual

Meeting Architectural Institute of Japan,

pp. 2731-2738, 1983.9

35) S. Yamato et al. : Studying on Earthquake- Induced Damage of Concrete Pile by

Horizontal Load Test with Axial Load under

Pile-Head Restrained (in Japanese) , Summaries

of Technical Papers of Annual Meeting

Architectural Institute of Japan,

pp. 2429-2430, 1984.10

36) T. Saitsu et al. : Proposal of a Joint

Method between Pile and Pile-cap (in

Japanese) , Summaries of Technical Papers of

Annual Meeting Architectural Institute of

Japan, pp. 2439-2440, 1984.10

37) Y. Ohnishi et al. : Ultimate Strength of

Joint between Pile and Pile-cap on Lateral

Load (in Japanese) , Summaries of Technical

Papers of Annual Meeting Architectural of

Japan, pp. 2441-2444 , 1984.10

38) H. Akiyama et al. : Experimental Studies

on the Steel Pipe Pile Footing Connections

under Lateral Forces, Part 1 to Part 3 (in

Japanese) , Summaries of Technical Papers of

Annual Meeting Architectural Institute of

Japan, pp. 2447-2450 , 1984.10

39) H. Akiyama et al. : Experimental Studies

on the Steel Pipe Pile Footing Connections

under Lateral Forces , Part 4 and Part 5 (in

Japanese), Summaries of Technical Papers of

Annual Meeting Architectural Institute of

Japan, pp. 1001-1004, 1985.10

40) K. Kokushoh et al. : Studies on Behavior

of Pile-Footing Connections with Steel Socket

under Axial and Lateral Forces (in Japanese)

,

Summaries of Technical Papers of Annual

Meeting Architectural Institute of Japan,

pp. 985-988, 1985.10

41) I. Yamamoto and M. Ohta : Strength

Properties and Failure Patterns on Joint

between Pile and Pile-cap (in Japanese)

,

Summaries of Technical Papers of Annual

Meeting Architectural Institute of Japan,

pp. 989-990, 1985.10

42) H. Ohno et al. : A Study on Behavior of

a Cone Shaped Steel Pile Head under Vertical

and Horizontal Loading (in Japanese)

,

Summaries of Technical Papers of Annual

Meeting Architectural Institute of Japan,

pp. 999-1000, 1985.10

43) S. Yoshida et al. : Horizontal Load Test

on High Strength PC (PHC) Pile (in Japanese)

,

JCI 7th Konference, pp. 469-472, 1985.5

44) A. Wada et al. : Study of the Stress

Transfer Mechanism in the Pile-Footing

Connection under Axial and Lateral Forces

(in Japanese) , Jour, of Structural and

Construction Engineering, Trans, of

Architectural Institute of Japan, No. 364,

pp. 111-120, 1986.6

45) I. Yamamoto : Pile Head Joint of

Precast Piles (in Japanese) , Materials of

"Problems on Aseismic Design of Pile

Foundation", Panel Discussion on Foundation

Engineering, Architectural Institute of Japan,

1985.10

46) T. Asama et al. : Design Method of Pile

Foundation by Considering Spring at Pile Head

and Horizontal Displacement (in Japanese)

,

Doken Shiryo, No. 1715, Public Research

Institute, Ministry of Construction, 78p.,

1981.8

47) Y. Gotoh and T. Shibata : Shear

Strength Estimation for Pretensioned Spun High

Strength Concrete Piles (in Japanese)

,

Summaries of Technical Papers of Annual

Meeting Architectural Institute of Japan,

pp. 983-984, 1985.10

161
Table 1 Comparison of degree of fixation for Infinite length pile

Kind of degree of

f ixacion boundary

condition

Degree of fixation References

. Equation Pin Rigid

Degree of fixation

to bending momenc

(Degree of restraint

for rocation)

a) Mo a - Usr Moy. a -0

r a “1

r 2), 3)

b) 6o»feMo

1 “r 1+EIS^c a ®Q

r a “Ir 12) ,15) ,16) ,18) .20)

25) ,26) ,29) ,39)

Degree of fixation

to displacement

(Displacement

factor)

c) Mo R -2-a

y o t R =2

y*> R -1

y o

7) , 8) ,13) ,17) ,24) , 28) ,30) ,32) ,33) ,41)

d) 0 o m kti o 1+2EIRA: V 1+EICA:

R -2

y o R »i

y o

6) ,12) ,15) ,16) ,25)

26) ,29) ,31)

tlo,0<>: bending moment and angle of rotation at pile head. M

°f

;

k

bending moment at pile head which is completely fixed. Inverse number of rotational spring at pile head.

Table 2 Properties of pile section (phase 1 series)

Pile Grade Standard Equivalent Design

B t M

cr H

u

I e

z

e cm

3 M

cr M

u cm cm Cm Cm cm Cm Cm

35 6 A 3.5 5.3 61540 3517 4.1 6.7

35 6 C 6.0 12.0 63970 3655 6.5 13.9

50 8 C 17.0 34.0 258800 10350 17.9 38.1

Table 3 Properties of test specimen (phase 1 series)

Test Pile rooting

Ho. p“ c

* c “c 35AX00 384 ou

| 241 1 2.45

35AYOO 384 64 207 | 2.52

35AZ00 903 65 173

' -1 m

Cl 35AX30 384 63 247 2.51

35AY30 S34 63 223 2.55

35A230 90 3 67 224 2.45

35AX60 384 60 225 2.71

35AT60 334 60 199 2.63

35A250 360 65 134 2 . 66 35CX0O 902 65 253 2.44

35CTOO 902 65 233 1.36

35CZOO 388 66 194 1.97

35c:go 387 62 ^4 cl O 1 . OM. 35CT30 387 65 212 1.36

33CZ2Q 888 69 170 1.53

35CX60 386 69 251 2.55

35CT60 3S6 65 234 2.30

35CZ5Q 388 63 225 2 . 12 50CX00 954 I 33

| 193 2.32

50CY00 954 1 34 137 2.41

30CZ00 901 1 38 1 175 2.71

1 5GCX60 95<* 34 177 cl«o o 50CT60 954 33 223 2.44

| 50CZ60 901 34 171 1 4-33

F :itg/c= : t:t= l :*L0 3 '<g /c=

c c

Table 4 Properties of test specimen (phase 2 series)

Thick- Equivalent Effective Design bending

Test ness moment iner. pre stress moment

No. t 1 0 M M

e e a c r

( cm ) (cm") ( k g / cm 2

) ( cm

( tm ) 10X ' 6 .77 6 5940 87.09 6.67 7.99

10Z ' 6.94 66 5 90 86.17 6.67 8 . 00

2 0 Z

' 6,42 64 5 20 95.33 6.91 8 . 20

2 0 Y

’ 6.15 6 3 3 20 94.91 .6.83 8.10

3 5 X

' 7.75 69 260 80 . 50 6.56 7.94

Standard 6 . 00 6 2620 80 . 00 6.17 7.42

Table 5 Properties of materials (phase 2 series)

Steel (kg/cm 2

) Concrete (kg/ cm

2

)

Diamecer (mm) 0

t 0

y

E

s *10 6 No. (day) F

c

E

c *10 5

PC bar 9.2 15255 13400 2.11 Pile 49 895 4.08

Fooc ing D13 6350 4455 2.00 F 1 ox

’ 35 228 1 .63

o 10Z* 35 226 1.61 Anchor D10 5762 3814 2.01

C 20Z

34 231 1.54

°

c

: Tension o : y Yield i 20Y 1 34 235 1.52

F Comp

E , s

: Young ' s modu.

g 35X

33 237 1.54

Table 6 Values and failure type at the aaxiauas loac nr- ?(t) o.(t) '•V

s=) 3SM* $SX** | Type

35A.X0O

-- 1 G 1

j I/O 10.32 3.31 i 3 1 35AY00 21.14 **11.97 14.32 0 . ^ c 3.30 2 47 - 35AZCO :i.o: -*16 . 65 3.51 1.50 -.95 S 35AX2Q 23.58 -18 .57 11.33 -1C . 36 1 . 19 1.32 ' S3 25AY20 25.36 -16.95 12.00 -3.34 1.20 2.02 3 1 35.AZ30 30.38 -20.70 10.39 -1-.35 1.44 1.93 S 35AX60 31.40 -20.53 n

* * 1.36 1.95 S

j 33AY50 30.74 -20.51 -11.95 Q .96 1.75 I S 33AZ60 34 . 15 -23.33 11.49 -16.47 2.31 : 3 !

25C3CCC 31.20 -17.32 19-96 -1.32 1 . 47 2.20 j 3 I 35CYOO 23.60 -13.16 13.52 -3 . IS 0.99 2.12 S j

1 35CZOO 33.00 -21.-3 14.95 -10.57 1.10 1.99 S

1 :=c:go ZZ.5B -20.71 16 . 10 -3.63 0.93 2.22 S i

| 35CS30 39.62 -19.16 12.08 -9.31 0.30 1.95 S

! 25CZ30 34.14 -22.50 14.93 -12.09 0.91 1.39 S

1

| 35C760 35.23 -22.90 16.04 -11.34 0.35 2.00 s? ! 35CY60 32.26 -20.33 15.16 -9.79 0.32 2.07 j S? !

i 25CZ6Q 38.48 -23.57 15.59 -15 . 01 0.34 1.75 ; Sr

1 iocrcoo 53 . 36 -39.06 40.42 -6.35 1-10 2.07 ! S

1 50C700 51. 36 -37.31 34.90 -10.33 0.95 1.35 S?

1 50CZD0 72.16 -45.6« 36.39 -17.69 1.00 1.52

1 3? 1

j 50CT60 74.16 —5.96 29 . 69 -15.35 0.33 1.73 3? i

i 30CY60 70.20 —4.05 36.39 -•

0.30 1.65 S?

' 30C250 33 . 6“ -5- . 39 33 . om -26.07 0 . 36 1.-3 S? : :cx' 33.53 -13. -6 L4 . 76 -6 o 17 0.94 • s 1

j ioz- 39.76 -13.35 14.33 -6 . 36 0.93 3.34 ; S :oz' 27.51 -19.45 10.73 -10. 12 0.71 -.77 | S :oy" 2* ,3S — 17 o 56 12.55 -7.06 0.32 3.30 ; 3

! 25S' 24 . 38 an 16.65 --•53 -.07 3.26 3 |

| M :Anaivzed nil—see oer.cinc taoacitv

I - i ucl uc ”"S3S“

| Max t M _ ».'!_) / (C r ’5 )| : Shear ssar. ratio S: Shear 3:3encing SB : Shear-oer.air.g S? : Shear and cite oer.cmg

162
6350 A 6350 C 6500 C

Fig. 3 Pile section (phase 1 series)

1

Hoop (03.2)

<9 50 1 4100

500

| <500~;750 A m . ifl'Vlu.lfljliU >\ <\ ;\ A ;\ i\ i\>\ !\ j\[

)

uv\nnnfmv\; \/TT\; \J \: \; \: 7T: \Y

i

B -350

/

'

PC bar 69.2-10

Fig. 5 Pile section (phase 2 series)

Embedding length 200 Anchor bar

X type Pile

I / lx - '

1/ rooting

\

J OIO

lt>JO J_=F=/ X A

250 100 (400)

(500)

Y type I 400

I OIO

<. irnO

f cnim

1

i S 1 400

100150

1 00,,Sneath

Z type

1

< 500> X * 100

Steel pipe

t-2.2 \

PC oar 6 7(49)

| 300 100 350 X (450)

(

) :Case of c500

Fig.i Detail of pile head joint

(Phase 1 series)

L:Load cell D:Dial gage for pile FD:Dial gage

for footing K: Inclinometer 1: Strain gage for

PC Dars fixed to tne floor SG: Strain gage on pile surface PG:Strain gage for PC bar in pile

f » f

* t SltTlCU

uo ; ST Dfi DS DK Sll CIO rp; —

100 or 350

Fig. 7 location of measurements (case of 20C)

Deformation of pile Strain on surface Strain of PC bar

Fig. 6 Detail of pile head joint (phase 2 series)

Fig. 9 Distribution of displacement and strain of pile

(case 35X')

163

Fig. 8 Relation between load and displacement

of pile (phase 2 series)
Degree

of

restraint

for

rotation

y _ (no) y e (TM)

Fig. 10 Relation between bending moment and displacement

ac loading point C

9 ( x 10

J ) P

e (*io' s ) ?

Fig. 11 Relation between bending moment and angle of rotation

at pile head

I

I

I

i

i

!

W. o

1.0

0.8

0.6

0.4

0.2

0

Fig.

Range of a

r without axial load

0.2

Ratio of

12 Liegree of

Phase N(t) i r r r l V Pile

First

0 0 A 0 A 30,60 • C

Second 30 © A <$> B

I I I L 0.4 0.6 0.8 1.0 1.2

embedding length to pile diameter

restraint for rocaclon ac pile head

164
J j A AUU

i / LiLA

35 AYOO 1 ( l

/ M f ^ 35AZOO | f 1

35AX30 |

( f

1

35 AY30 |

1 f

35AZ30 |

35AX60 |

Fig. 14 The final state of cracks of pile (phase 1 series)

Footing

a) Test snecxmen 10Z'

Pile Footing

b) Test specimen 20Y*

Fig. 16 Comparison between experimental and analytical

results of shear strength of pile

-> Test soecimen 35X'

9 Fig. 13 The final state of tracics of pile ^pnase 2 senes)

165
.

.

tfxes&a'fey® -4w-

ss>v^ysteift'w; $8$w

-
Schemes for Seismic Upgrading of a M^jor Hospital

BY

Joseph Vc Tyrrell, P e E

ABSTRACT

Design of the Oak Nolls Naval Hospital was begun in 1965 and construction was completed

in 1968. The design was according to the

then current criteria which was based on the

Uniform Building Code. Earthquake experience

in the 1960's created concern for the

adequacy of the design codes. The 1971 San

Fernando earthquake clearly demonstrated that

some existing U.S. structures did not provide

a desirable level of seismic safety.

Consequently, the U.S. Navy began evaluation

of facilities located in areas of high

seismic risk. The Oak Nolls Naval Hospital

was identified as requiring upgrading and

various mitigation schemes were proposed.

This paper discussed these schemes.

KEYWORDS: Retrofit; hospital safety; seismic

hazard mitigation; upgrading schemes.

1.

INTRODUCTION

In the 1960's the Alaska earthquake and other

events raised questions regarding the

adequacy of design codes. At the same time,

accumulating earthquake records indicated

that the seismicity of several areas in the

United States had been underestimated. In

1966, the first edition of the tri-service

design manual contained both increased design

criteria and increased estimates of

earthquake risk in several locations.

The Naval Facilities Engineering Command

had just begun appraising existing structures

when the 1971 San Fernando event occurred.

This confirmed that there were still

deficiencies in the design codes and that

certain existing structures were unsafe.

Several major hospitals were seriously

damaged by the 1971 event. The US Navy's

inventory of existing facilities includes a relatively small percentage of high rise or mid-rise structures but hospitals are typically in this category. Because of their

importance and the experience of earthquake

damage, hospitals were given a high priority

in the Navy's investigation.

Of the six naval hospitals investigated, all were found to be be deficient with respect to

earthquake safety. Two of these facilities

(Bremerton, WA. and San Diego, CA.) have been

replaced with new construction. Deficiencies

in the hospital at Charleston, SC were relatively minor and upgrading has been

completed at a cost of about $100,000.

Action has not been taken for the other three

hospitals. However, funding was approved for

the Oak Nolls hospital located at Oakland,

CA. and design completed. Unfortunately, the

bids were higher than the available funds and

at this time the next step remains to be

determined. This paper discusses the

existing Oak Nolls structure, the criteria

for upgrading, the mitigation schemes

considered, and probable reasons for the high

bids.

2.

EXISTING STRUCTURE

The Oak Nolls Naval Hospital is a nine story,

reinforced concrete moment frame with the

exterior columns formed as part of precast

concrete wall panels. The interior columns

and floors were cast in place. The lower

four floors are of irregular shape and extend

out on three sides from the tower. The tower

structure consists of four wings which are offset from the centerline of a central core

area (see Figs. 1 & 2). At some places there

are discontinuties in the exterior walls

where higher levels are supported from a cantilevered floor.

The Oak Nolls structure is supported on piles, some end bearing and some utilizing

skin friction. The tops of the piles are at

two different elevations depending upon the

location. The building was constructed

against a hillside. The lowest level is

completely below grade and the next higher

level is below grade on one side and at grade

on the opposite side.

3.

DESIGN CRITERIA

The Navy's criteria for lateral force in

connection with seismic upgrading of typical

buildings is that the capacity of the

structure be at least 75% of the demand

imposed by an earthquake with an 80%

probability of not being exceeded in 50

years. Under this loading, the structure

must not collapse.. This policy recognizes

the fact that there are a large number of

166

1 Naval Facilities Engineering Command,

Consultant in Earthquake Engineering
deficient structures and that it is not

economically feasible to provide funds to

fully correct all of the deficiencies.

The program focuses on structures that are liable to collapse. For essential

facilities like hospitals, the target is to

provide upgrading so that capacity equals

100% of the demand. A site seismicity

study was made and response spectra

established for the site (see Fig 3). The

preliminary structural investigations

utilized the spectra in connection with the

80% /50 year earthquake described above.

Analyses were based upon a lumped mass,

single degree of freedom model. The

structure was also analyzed for the code

forces required by Department of Defense

criteria. This latter criteria is based

upon the Uniform Building Code. The

hospital was found to be seriously

deficient for both loadings.

4.

CONDITIONS FOR UPGRADING

In addition to strengthening to provide

sufficient strength and ductility to meet

the design criteria, it was determined that

it was necessary to maintain full operation

of the hospital during construction. This

requirement had a major impact on the

selection of a structural scheme and

significant cost implications. It made it

necessary to relocate functions temporarily

while work in a given area was completed.

Control of noise, dust, and vibration also

became a significant issues.

A very detailed phasing plan was drawn up

in conjunction with the medical personnel

at the hospital. As a result of these

requirements, the construction time

required was increased from the normal

expectation of about 20 months to about 4

years.

5.

STRUCTURAL UPGRADING SCHEMES

THe firm of URS/BLUME, San Francisco, CA

performed most of the preliminary studies.

The following schemes were considered

initially:

a. Construct new concrete shear walls

normal to the existing walls for both the

tower section and the low roofed portion.

b. Add reinforced concrete to increase the

shear strength of the exterior walls, walls

around the elevator core, and a few

selected interior partitions.

c. Construct a very stiff new six story

structure in the southwest quadrant (no

existing construction in this area), and

subsequently a matching structure to

replace the existing construction in the

northeast quadrant.

The preliminary cost estimates were 10.5 to

11 million dollars (1974) for these schemes.

They all involved considerable internal

disruption. The third scheme would have

allowed relocation of functions from the

northeast area to the new southwest area,

thus facilitating work in the northeast

area. It appeared likely to take longer and

be somewhat more expensive. The other two

schemes involved exposing a large portion of

the occupants to close proximity to

construction at one time or another. They

also required some reshuffling of functions

within the existing space. None were entirely acceptable for the purpose of

keeping the facility operational during

construction. Consequently, it was decided

to explore other possibilities.

6.

SECOND GENERATION SCHEMES

The necessity to minimize interference with

operations had become the dominant factor

influencing design selection. To accomplish

this, attention was focused on reinforcement

placed at perimeter. This led to the

following schemes:

a. A variation of the first scheme with

concrete shear walls perpendicular to

exterior walls plus reinforcement of the

exterior walls.

b. "C" shaped concrete buttresses at the

ends of the wings (see Fig 4). The

buttresses were to be tied to the existing

structure by exterior beams extending along

the floors of the tower.

c. Braced steel buttresses at the ends of

the wings to be connected in a manner similar

to the concrete buttresses.

On closer examination it was apparent that

the first of these schemes still required

extensive structural work inside the building

below the fifth floor in order to carry the

reinforcing down to the foundations. Thus

this scheme was discarded. The other two

schemes also resulted in a similar condition

at two of the wings, however, there was much

less construction required within the

existing building.

Besides the difference in construction

techniques, the concrete and steel buttresses

also differed in design philosophy. The

existing structure was fairly stiff, although

much of the initial stiffness would be

supplied by elements not intended to take

such loading. The concrete buttress

envisioned utilization of a typical approach

- it would provide even greater stiffness and

thus take lateral loads before the vulnerable

members. The increase in stiffness would

167
result in a shorter period and consequently

the forces would increase according to the

position on the response spectrum.

The steel braced frames were planned to be

not quite as stiff as the existing

structure. Initially the existing

structure would resist most of the lateral

force but after a little yielding it would

gradually transfer to the new structure.

Under the forces associated with a large

earthquake, the stiff non-structure

elements (exterior walls and partitions)

would sustain some yielding. Under a great

earthquake there might be structural damage

in the original members. The new steel

braced frame would remain elastic at all

times.

7.

SELECTED SCHEME FOR STRENGTHENING

The braced steel buttresses were selected

because they cost less and were easier to

erect. While the concrete version might

have reduced the chance of damage, it was considered unrealistic to prevent any

damage because of details of construction

and the geometry of the existing

structure. The steel buttresses provide

the essential requirement for life safety.

The steel design required the new structure

to be more flexible than the existing, but

also to be stiff enough to quickly pick up

the lateral forces when yielding took place

in exiting elements. It was found to be

feasible to meet these requirements.

Additional detail of the steel concept is

shown in Figures 5 and 6.

The principal features of the were :

a. Vertical steel braced frames at the end

of each of the four wings.

b. Horizontal braced steel diaphragms

connected to the existing concrete floors

and roof and intended to transfer forces to

the new braced frames.

c. Diaphragm chord strengthening for all

floors above the fourth floor as well as both shear and chord strengthening of the

roof.

d. Drilled caisson foundations supporting

the new braced frames and providing for

transfer of lateral forces into the ground.

8.

ARCHITECTURAL CONSIDERATIONS

One problem with external modification was

the fear that the -architectural quality of

the building would be adversely affected.

To overcome this objection, it was planned to

provide some architectural treatment. Figure

7 shows one concept. Eventually, it was decided to use a simple, lightweight metal

cladding, treated to harmonize with the

existing concrete.

9. BASE ISOLATION

For some time the Navy has been considering

base isolation treatment to reduce the

lateral forces associated with earthquake

motions. This method appears to have

potential benefits for certain types of

construction. To date a project has not been

identified. However, proponents of several

proprietary systems suggested use for the Oak

Nolls hospital. They stated that it could be

installed with less disruption to operations

and lower cost than conventional

strengthening.

After study, this idea was discarded as impractical. A major reason for this

decision was the nature of the existing

structure. The building is unsymetrical in

the first four floors, the center of rigidity

is not near the center of mass, the length of

piles varies considerably from northeast to

southwest, the pile caps are at two different

elevations, and the structural details

provide very little ductility. The location

of the hospital, very close to a major fault,

posed the possibility of strong vertical

accelerations. Therefore, this application

would have taken base isolation into areas outside the framework of experience. Even if

the base isolation system worked perfectly,

strengthening of the superstructure would

have been necessary to avoid severe

overstress.

10. FINAL DESIGN AND BID

The final design was prepared by the firm of

H. J. Degenkolb Associates, Engineers, San

Francisco, CA. They reviewed and reconfirmed

the recommendations of the studies. However,

it was decided to make some changes since

extensive modifications for heating and

ventilating, fire protection, and medical

improvements had come into the project.

Because these items required interior work,

it was decided to place the ties and chord

strengthening members under the floors inside

of the building. These changes further

complicated construction and made the phasing

plan more difficult.

The bids were about were about 35% above the

estimate. This appeared to be attributable

to the complexity of the construction which

was rooted in the requirement to keep the

hospital fully operational during

construction. Since additional funds were

168
not available, the project was deferred. At

this time a definite plan of action has not

been developed.

11. SUMMARY AND CONCLUSIONS

A number of schemes were investigated and

several found to be possible structural

solutions for the Oak Nolls hospital. While

they were expensive, they were not nearly as costly as building a new hospital of

equivalent size and capability. The project

was not brought to a successful conclusion

because the difficulty and cost of

maintaining the hospital operational during

construction was under estimated. A suitable

upgrading for any given facility is strongly

influenced by the nature of its vulnerabil ity

and the function served.

169
U-S.SJ. HOSPITAL,

&A ICLA/sl ,Q , CALtFQfrNJA £ ! <S» U * S Z

170
SF O S

HOSPITAL.

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FIGURE

7

PERSPECTIVE

VIEW

-

PROPOSED

STRUCTURAL

SOLUTION
Theme III

Storm Surge and Tsunami

Experimental Investigation on the Damage of Offshore

Seawall due to the 1983 Nihonkai-Chubu Tsunami

BY

1 2 Katsutoshi Tanimoto and Hiroichi Tsuruya

ABSTRACT

A tsunami caused by the 1983 Nihonkai Chubu

Earthquake (M=7.7) struck the coasts along the

Japan Sea. A part of the seawall which was

under construction for a future reclaimed land

at the outside of Noshiro Port was collapsed

by the attack of tsunami.

In this study experiments were conducted on

the deformation of the tsunami during the

propagation on a continental shelf to the

coast. Tsunami forces acting on a vertical

wall were also tested under the action of

breaking bore to investigate the cause of the

damage of the seawalls.

The bottom topography of the sea from the

continental shelf to the coast was reproduced

in a 163m long wave channel.

From the experiment, the pressure distribution

of tsunami in a state of a breaking bore is

formulated and it is concluded that the

damages of the seawall in Noshiro could be

explained well by the present formula.

KEY WORDS: Damage due to Tsunami, Deformation

of Tsunami, Bore-type Tsunami, Tsunami Force,

Seawall

1. INTRODUCTION

On 26 May 1983, a tsunami, caused by a seismic

disturbance in the offshore of Akita

Prefecture at latitude 40°42' and longitude

138 °54', struck the coasts along the Japan

sea. The tsunami took about one hundred

person's lives and brought severe damages over wide area along the Japan Sea. Port

constructions such as seawalls and breakwaters

were also damaged in various places. For

example, the seawall which was under

construction for a future reclaimed land at

the outside of Noshiro Port was collapsed over

300 meters by the attack of tsunami. The

observed period of the tsunami ranged from ten

to twenty minutes. A remarkable aspect of

this tsunami was that it attacked the northern

coasts of Akita Prefecture where the bottom

slopes were relatively gentle, in a state of

breaking bores with undulations (solitons)

succeeding the bore front having a period of

about 10 seconds.

*Dr. of Eng., Chief of Breakwaters Laboratory,

Hydraulic Engineering Division, Port and

Harbour Research Institute, Ministry of

Transport, Japan.

2 Chief of Hydrodynamics Laboratory, Marine

Hydrodynamics Division, Port and Harbour

Research Institute, Ministry of Transport,

Japan.

Fig.1 Damage of seawall outside of Noshiro Port

177
Fig. 2 Standard cross section of seawall

in the completion (Section A-2)

Fukui et al. L

* investigated the wave pressure

of a bore-type tsunami acting on a seawall.

They have concluded from the experimental

results that both the dynamic pressure at the

collision and the continuing pressure after

the collision can be represented by linear

distributions having the maximum value at the

still water level.

On the other hand, Tanimoto 2

* has proposed the

formula of pressure on a vertical wall due to

non-breaking tsunami in a region of relatively

large water depth. The formula is a modified

one of Goda's formula for the wave pressure on

a vertical breakwater on the basis of

experimental results so that it can be applied

to long waves as tsunamis. In applying these

formulae, however, we must previously know how

the tsunami attacked to the seawall at Noshiro

Port together with its various parameters.

In this study, therefore, experiments were

conducted on the deformation of the tsunami

during the propagation on a continental shelf

to the coast. Tsunami force acting on a vertical wall was also tested under the action

of breaking bore to investigate the cause of

the damage of the seawalls.

2. DAMAGE Or SEAWALL

The seawall was under construction for a futurs rscl 2.i.nrsc l^nci oz Ncsrii^o i,c power plant at the outside of Noshiro Port.

Figure 1 shows the state of the collapse and

slide of caissons after the attack of the

tsunami. The front seawall facing to the

offshore is divided into three parts, namely,

A - 1 , A-2, and D according to the final crest

level of the seawall at the completion.

Except for the length 200m at the southern

part of D, caissons had already been settled.

Figure 2 shows the standard cross section of

the seawall at the completion in the part A-2.

The dimensions of the caisson are 20m long,

11m wide, and 8.5m deep. The total bottom

width including the footing is 12.5m. Only

the top levels of the parapet in the parts A-l

and D differ from that in the part A-2. The

Fig. 3 State of sliding of No. 68 caisso;

concrete cap and the wave dissipating coricr-ct blocks had been constructed to the level J< f>.( 'm for the caissons from No.l to No. 21, and <h.0ir from No. 22 to No. 35. The rubble backing i:ac been mounded to the level +2. 5m for the

caissons from No.l to No.3 2, and to the level'

-2.0m from No. 33 to No. 35. The caissons in

these sections were not damaged. For the

caissons from No. 35 to No. 50, neither the wove dissipating concrete blocks nor the rubb)

backing had been mounded. And there wcr. various stages of caissons such as that in

which sand was packed and only the capping was

placed, and that on which the top concrete war. constructed to the level +4.0m. The caissons

in this section were exceedingly damaged. The

caisson No. 39, for example, was moved about

38m to the shore side from the initial

position and overturned. The caisson Nc.-.O was moved to the shore side about 74m and this was the longest distance among ail the damaged

caissons

For the caissons from No. 51 to the outlet of

drainage, the concrete cap had beer,

constructed to the level t-r.Om, and the wave- dissipating concrete blocks and the rv ' ; backing (except for the caisson Mo. 51) i.t-f also been constructed. In this section re caisson has suffered the damage except f or

No. 51 caisson at the edge of this section.

For the caissons No.S I'VNo.S 5 in the southern

part from the outlet of drainage, the concrete

cap and the wave dissipating concrete block-,

had been constructed to the level + 4.0rn. Tne

rubble backing, however, had not; been mounded

and the caissons except for No. 85 was slided

over the distance from 0.15m to 21.5m. Figure

3 shows the state of the sliding of the

caisson No. 58 of which the sliding distance

was the maximum within this section.

178
From the fact in the field above-mentioned, it can be concluded that the sliding or the

collapse of a caisson came about mainly in the

region where the rubble backing had not been

mounded.

3. EXPERIMENTS ON TSUNAMI PRESSURE ACTING ON A VERTICAL WALL

The experiments were conducted in a concrete

flume 163m long, lm wide, and 1.5m deep. The

water depth in front of the wave paddle was kept constant at lm throughout the

experiments. The model of sea bottom is

started at the distance of 37m from the wave

paddle and composed of three slopes. The

slope of the most offshore side is 1 on 10 and

extended to the horizontal length of 5m.

Then, the second slope of 1 on 200 is

continued over the horizontal length of 95m

and connected to the shore slope of 1 on 20.

The scale of the model is 1/200 without

distortion. After this the size is described

with that in the prototype unless the special

notice is given. The offshore water depth on

the horizontal bottom corresponds to the depth

of 200m in the prototype.

The caisson having the width of 10m and depth

of 13m was set at the water depth of 9m. At

the front face of the caisson two pressure

sensors were equipped (below the still water

level 1.4m and 6.3m) to measure the variations

of the hydrodynamic pressures of the tsunami.

Water elevation was measured at the position

450m offshore from the caisson (station C,

water depth is 11.25m).

The periods of the tsunami used in the

experiment were 15, 10, and 8 minutes (these

periods correspond to 60, 40, and 35 seconds,

respectively, in the experiment). The

amplitude of tsunami was changed in nine

levels for each condition of period. The

offshore profile of the tsunami was sinusoidal

and the first wave was began to push. The

experimental case is summerized in Table 1. The symbol a 0 in the table represents the

amplitude of the first wave of the tsunami at

the uniform water depth 200m.

Figure 4 shows examples of the water surface

elevation (f|) and the wave pressure (p)

measured during first two waves. In the top

figure showing the time history of the water

surface elevation at the station C, we can observe that the tsunami approaching to the

caisson is greatly deformed into the bore

shape having undulations (solitons) from the

initial sinusoidal form in the offshore. It

is also observed that the reflected bore from

the caisson, which is propagating toward the

offshore, is superimposed on the incident bore

after about 1.5 minutes from passing of the

incident bore and that the water surface

elevation becomes about twice as much as that

Table 1 Experimental case

(Dimension for model test)

i r(s) L (m) 2<zo 2ao/L

60 187.79 9 types 3. 23xl0- s ~8. 52x10-5

1/200 40 125.17 // 5.91xl0- 5 ~2.40xl0-i

35 109.51 // 1.20x10-1-3.22x10-1

10 7(m)

0

-10

i = l/200 Tp-I0min e=350 Run 52

Station C

20

JO -

(tf/m

2

) Q

-

-10 - -20

p(Zo : -l.4m)

18

14 15

t (min)

(tf/m

z

)

p(z„=-6.3m)

20 C

0

- 0

-10

-20

18 19

2 3 4 5 6 7 J 14 15 16 17

t ( min

)

Fig. 4 Surface elevation n at station C and pressure p

acting on seawall

179
First wave

. Tv —

p. ia

_ I

Second wave

1 t

P, 0,

Fig. 5 Magnification of pressure signal

at 6.3m below still water level

of the leading bore after the overlapping. At

the front of the caisson,, the reflected bore

is to be superimposed on the incident bore

without any time difference. Therefore, in

the records of pressure, no separation of

incident and reflected bores is noticed.

The height of incident bore di s regarding

undulations is defined as the height from the

water level directly before its arrival. As

the first bore propagates on a still water,

the height a t is the height from the still

water level. The second bore, however, comes

when the water level is extremely lowered bv

the ebb of the first wave. The difference

between the water level and the still water

level is reprase n tec as A n » (Ah i <0) and the

Fig. 6 Vertical distribution of

bore pressure

bore is accompanied by solitons, the relation

p 0 =l.lp s is obtained.

Figure 6 shows the pressure distribution,

where w 0 is the specific weight of sea water,

a is the height of incident bore and z is the

height measured upward from, the water level

just before the oncoming bore (quasi "till

water). As the two points of measurement

locate below the initial still water level,

the data for the first wave distribute only in

Cue ^ a, c. i v c r c c i c u o z z . ~ z z c c ccr.c height a 2 of the second bore is measured from

the level Ah r

. Figure 5 shows the magnified records of

pressure measured at 6.3m below the still

water level. They correspond to the bottom

figure in Fig. 4. Similar to the water

elevation, the fluctuations with short periods

can be recognized. Especially, the impulsive

pressure is generated at the second wave. The

time cf duration, however, is very short.

Therefore, the net momentum of the impulsive

pressure is not dominant for the stability of

the caisson. Moreover, the impulsive pressure

does not appear at the elevation 1.4m below

the still water level and it can be thought

that the impulsive pressure will act only on a part of the vertical wall. From the above

consideration, the impulsive pressure is

disregarded. The intensity of pressure due to

the action of bore is defined as same way to

the height of bore. In the figure, the

maximum pressure except for the impulsive one

is denoted by p 0 , and the maximum pressure for

the smoothed record disregarding the

undulations is denoted by p s

. In this study,

p s is considered as the representative

pressure for the action of bore. When the

wave, on the other hand, ail the cats

distribute in the positive region of z as the

water level just before the second bore is

lowered and the pressure sensors are exposed

to the air. Although the data show some

scatter in the figure, the pressure above the

cuds i s t i 1 1 w d c s r Is*/ si ddcrsds^s — s tch*.

relative height increases on the average. The oressure below the quasi still water level can be considered to be constant from the

viewpoint of the total pressure, although the

pressure .slightly increases as the relative

height decreases. The straight line is c_ •n by the method of the estimation which will be

discussed in Chap. 4.

4 . METHOD OF TSUMAMI ~CP.CZ ESTIMATICM

From the experimental results obtained in

Chap. 3, the, distribution of bora pressure

acting on a vertical wall is proposed as snovn

in Fig. 7. The vertical distance rj* from the

quasi still water level and the constantpressure p x can be expressed as

v

* =3.0a I

( 1 ) p 1 = 2.2a/0 <z / ( 2 )

180
where, a x is the height of the incident bore

above the quasi still water level. Notice

that the quasi still water level here means

that just before the oncoming bore.

Therefore, for the second wave in the

experiment the water level lowered due to the

first ebb must be taken into consideration.

When the quasi still water depths h x and h 2 before and behind the caisson differ, the

difference of the hydrostatic pressure must be

considered. In the case of the figure (h x < h 2 ), the resultant hydrostatic pressure facing

to the offshore direction acts on the caisson.

At the bottom of the caisson an uplift acts to

the caisson such as P U = P]_ at the edge of the

caisson. The buoyancy must be corresponded to

the volume just before the attack of the bore

as shown in the figure.

The method in estimating a tsunami force

presented here is the same as that for the

non-breaking tsunami which has already been

„ 2 > proposed > if the wave height Hj is applied to

2a t •

5. INVESTIGATION OF THE CAUSE OF DAMAGE

(1) Critical height of bore against sliding

Figure 8 shows some typical cross sections

before the suffering. The cross sections of

the damaged caissons are sections III and IV

for which the rubble backing were not mounded.

The caissons of the section III was sliced

even if the wave dissipating concrete blocks

had been executed.

First of all, we examine that in what

condition the sliding of the caisson occurs on the basis of the formula of tsunami force

which has already been discussed in Chap. 4. In the investigation the quasi still water

depths before and behind the caisson are assumed to be equal (h 1 =h 2

) even for the

second wave because the ebb of the first wave continues for relatively long time. It is

also assumed that the tsunami force is not

affected by the existence of the wave

dissipating concrete blocks, since the tsunami

force resembles like a hydrostatic pressure.

In this situation, the critical height a _ of

the oncoming bore against sliding can be

expressed as

Fig. 7 Distribution of tsunami force

acting on a caisson

a F —

[iWt+Wj tan (d + $) + l.lwohr-IB

h* u B \ x2' 2zy ° /z/ (3)

where, h c

* = min{ h c , 3.0a c } , U the friction

coefficient between a caisson and a mound(=0.6), W 0 the weight of a caisson in sea water per unit length (tf/m), w 0 the specific

weight of sea water ( = 1.03 tf/m 3 ), h c the

height of a caisson from the quasi still water

level (m), h' the depth of the bottom of a caisson (m), B the width of a caisson (m), and

min{a,b} represents the smaller value of a and

b. The second term of the numerator in the

right-hand side of E q . ( 3 ) represents the

resistance force R against sliding by the

rubble backing, namely.

R = W0

' tan (d-rp) (4)

where, W'0 is the total effective weight of the

rubble backing above the sliding plane (tf/m),

9 the inclination of the sliding plane in

degrees, and b is given by the following

formula

^ = tan“1 /'; (5)

Section III

«£ 0

O.LtOOO

1

_J

7

-6.5 -

-3 5 / ^

Section IV

* 2.0 0 !_. i 0 00

lilZII

Fig. 3 Cross sections of seawalls

(under construction)

Fig. 9 Sliding plane of rubble backing

181
0 -I -2 -3 -4 -5 -6 -7

4A (m)

i i i i i i i [ 865 765 665 5.65 4.65 365 2.65 1.65

4, (m

Fig. 10 Critical wave height of

tsunami for sliding

manner, but it is larger than that against

sliding. Therefore, the results for the

overturning are not presented here.

(2) Inference of the damage condition

We now infer that in what condition the

caissons were damaged by applying the results

of the field investigation and the

experimental study on run-up heights 3 *.

From the records of the tide gages, we can conclude that the period of the tsunami was about ten minutes. Although an ebb can be

recognized at the beginning of the tsunami in

the record of the tide gage in Noshiro Port, we regard that the tsunami started from the

flood because the deviation of the ebb is vs rv small of the order of 0.2m.

According to the field investigation, the rur.- uo height at the coast around Noshiro Port was

6n,7m except near the special topography such

as the river mouth. The amplitude of the firsc

tsunami at the offshore of 200m deep, which

results in the run-up height of 6"W m on the

coast, is deduced as 1.2^1. 4m, on the basis of

the experimental results, although they are not shown in this paper. As already

described, the wave profile is graacly

deformed when it propagates on a continental

shelf. Again, on the basis of experimental

results, the height of first oncoming bore at

the water depth of 9m where the seawall exists

can be estimated as 2. 2^2. 6m.

where, u 2 is the friction coefficient between

the stones (=0.8). Figure 9 shows the sliding

plane through rubble backing. The resistance

force against sliding in Eq.(4) is given as m i p ^ irmm vs I'j® of P- -cir vsiri.cus ss of 0. which is detected by the trial and error

method. The minimum values for the section I and II in Fig. 8 were obtained when 9 equals

zero. The effective weight of. the rubble

backing is calculated by using the unit

weights of 1.8 tf/m 3 and 1.0 tf/m 3 for the

volumes above and below the water surface,

respectively.

Figure 10 shows the calculated result of a c for the section I^IV by using Eq. ( 3 ) . In the

calculation the astronomical tide level at the

attack of tsunami is fixed at -0.15m and the

quasi still water level just before the

oncoming bore Ah is assumed in the range 'from zero to -5m (the negative sign represents the

decrease of the water level). The weight of

the caisson per unit length in air is 251.3

tf/m for the section I^III, and 201.5 tf/m for

the section IV.

The critical heights a c against sliding are

5.3m, 4.5m, 3.5m, and 2.9m for the sections

I, II, III, and IV, respectively, when Ah=G.

These are the minimum values and the critical

height a c increases as the quasi still water

level decreases. The critical height against

the overturning can be obtained from same

These heights are greater than 1.94m which is

the height recorded at the tide station of

Noshiro Port. In comparing these results,

difference of the height between inside and

outside of harbor sheltered by breakwaters,

the response characteristic of the tide gage

well, and the effect of reflected waves.- \ According to the theoretical considerations

it is estimated that the ratio of the tsunami

heights between inside and ounside of the

harbor is about 0.5 and the damping ratio of

the tide gage well is about 0.8 for the firsa

wave. Consequently, the recorded tsunami

height of 1.94m is converted into 4.9m as the

he i ch t of tsunami outside the harbor. Since

this height cor responds to the sum of the

incident and reflected waves, the h e i g n t or incident tsunami can be reasonably considered

as the half of 4.9m, namely about 2.5m, w n i c

is nearly the same as the deduced height of

the first bore mentioned above.

As to the second wave, we assume that the

amolitude in the offshore is the same as the

first one. This assumption is reasonable

because of the results of the numerical

calculation for the 1983 tsunami generation

and propagation 5

.’ Then on the basis of the

experimental results 4

,* the quasi still water

level Ah lowered due to the ebb of the first

wave and the height a 2 of the second bore can

182
be estimated for the range of the offshore

amplitude from 1.2m to 1.4m as follows:

© Ah = -5.2(m), a 2 = 5.0(m),

and @ Ah = -5.8(m), a 2 = 6.0(m).

When these tsunamis attack the seawalls, their

stability can be judged as

1) For the action of the first bore;

Ah=0, a 1 = 2.3'V2.6<a c (Section IV)=2.9m.

Therefore, any section is not slided.

2) For the action of the second bore (J) ;

Ah=-5.2, a 2 = 5 .0>a c

( Section IV)=4.9m,

and <a c (Section III)=5.5m.

Therefore, only the section IV is slided.

3) For the action of the second bore @

Ah=-5.8, a2=“6.0>a c (Section IV )=5.1m,

>a c (Section III)=5.8m,

and <a c (section II )=7.5m.

Therefore, the sections IV and III without

rubble backing are slided, but section II with rubble backing is not slided.

In fact, the sections IV and III were slided.

Therefore, it would be preferable to adopt the

larger condition. It thus appears that the

first bore of about 2.6m high attacked the

seawalls but they were not slided. When the

water surface fell down about 5.8m on account

of the ebb by the first wave, the second bore

of about 6.0m high attacked and the seawalls

were damaged.

(3) Deformation of tsunami and its action

against seawall under the damage condition

Deformation of the oncoming tsunami and the

situation of the action to the seawall under

the damage condition are presented below,

although the tsunami height is slightly large

comparing with the inferred damage condition.

Figure 11 shows the water surface elevation

measured at various stations without setting

the model of the seawall. The water depth

becomes shallow from the top of the figure

which corresponds to the uniform depth

( h- 200 m ) to the bottom one (h = 0m) which

corresponds to the shore. The first wave of

the tsunami deforms and becomes bore as it

propagates to the coast. The site of the

seawall locates between the stations B(-8m)

and C

( -11.2 5m ) . At the station C which is

slightly offshore from the seawall, the

maximum water surface elevation of the first

wave arises slightly behind the tip and the

value is about 2.6m. At the station B which

183
Fig. 12 Water surface elevation before seawall and

wave pressures acting on vertical wall

time ( protocype ) time ( prototype

+ 4.0 01

rr i 00

f s 0 s + 4 0 m

.... + 2.0 ebb of first wave t2Cr—

-9.0

\\ -60m |

r*5.5s surge with foaming

t

3 ! 1,0s t s 1 1 0 s

/

3 I S 5 s

/

a I S 5s

t 3 27.5 s 27 5 s

t

3 5 3 .0 s r

3 23 Os

t 3 32 5 s / * 33 5 s

fTTTi1 —

Fig. 13 Action of first wave of

tsunami on seawall

Fig. 14 Action of second wave

of tsunami on seawall

184
Photo 1 Attack of second bore on seawall

is slightly onshore from the seawall, the

small undulations can be recognized at the tip

of the first wave. The amplitudes, however, are rather small and the maximum water

elevation is about 2.7m.

The second wave at the site of seawall comes

out as a distinguished bore when the water

surface is extremely lowered by the ebb of the

first wave. The lowered water surface level

at the station C just before the attack of the

second bore is about 5.4m, and 5.2m at the

station B. In the record of the second bore,

remarkable undulations can be recognized

especially at the station C. Among them the

second wave is the largest and the wave height

is 6m. After the smoothing of the undulations

the maximum height of the bore can be read as c.5m. On the other hand, the amplitudes of

undulations at the station 3 are small and the

maximum wave height is only about 1.4m. The

maximum height of the bore obtained from the

smoothed record of surface elevation is about

5.5m.

Figure 12 shows the water surface elevation at

the station C and pressure records measured at

the vertical wall when the seawall is settled

at the water depth of 9. Cm. Comparing with

Fig. 11, the reflected waves appear earlier

because of the existence of the wall.

Therefore, it is slightly difficult to

distinguish the incident and reflected waves.

If we read the heights of oncoming bore from

the record before the reflected bores, the

height of the first bore is 2.3m, the lowered

water surface level is 7.2m just before the

arrival of the second bore, and the smoothed

height of the second bore is about 6.5m. If we compare this result with that without the

seawall, the core heights are nearly the same,

but the decrease of the water surface just

before the second bore is greater when the

seawall exists. The reasons are that the

formation of a clapotis becomes remarkable by

the existence of a vertical wall and the point

of measurement is located near the loop.

The above discussion corresponds to the

situation that the vertical wall has infinite

length. Actually, the length is finite and as the wave length of a tsunami can be considered

to be very long, the water surface would not

be lowered as in the experiment. Namely, the

actual reduction of the water surface can be

considered as to exist between that of two

cases with and without vertical wall.

Taking the above discussion into

consideration, the reduction of the surface

water level Ah just before the attack of the

second bore in the estimation of the damage

state was given as the average of the two.

The large undulations can be recognized at the

second bore in the case without the seawall,

but they are not remarkable in the case with

the seawall. Although the undulations can also be recognized for the pressure, the

amplitudes of them are small. The maximum

pressure arises some time later from the

collision in this case.

Figures 13 and 14 shew the situation of the

attack of the first and the second waves to

the seawall, which are illustrated from the

successive photos taken by a motor-driver,

camera. The time is converted to the

prototype value. The overtopping to +4.3

crown level is started within 10 seconds after

the oncoming first bore is noticed around the

location of 50m in front of the seawall. The

height of the overtopping above the crown

level is from 1.0m to 1.5m and continues for

relatively long time. If the crown level was

+ 2.0m, it would be from 3.0m to 3.5m.

The action of the second wave to the seawall

is more rapid and severe. The time interval

from the detection of the oncoming bore with

foaming around the location of about 50m in

front of the seawall to the commencement of

overtopping is only 6 seconds. Moreover, the

remarkable feature is that the overtopping

abruptly happens from the extremely lowered

185
water Level by the ebb of the first wave. At

the very initial stage of the action, a relatively high green water is produced ( Photo 1). The succeeding overtopping height

is, however, not so high of the order of 1.0m

above the crown level of +4.0m.

The situation of the attack of the bore-type

tsunami on the seawall has been described

based on the experimental investigations.

Although the cross section and the water depth

in the experiments are slightly different from

the actual seawall for a reclaimed land of

Noshiro Electric Power Plant, the situation is

nearly the same. In the experiment, the water

surface level behind the seawall at the attack

of the second bore is nearly equal to the

initial still water level, because the

overtopped water cannot return to the offshore

side. However, the actual water level behind

the seawall would drop considerably comparing

with the experiment.

6. SUMMARY

We have described the investigation on the

damage of the seawall which was under

construction for a future reclaimed land at

the outside of Noshiro Port.

In the first place, experiments were conducted

to investigate the situation of the attack by

the bore-type tsunami and pressures

distribution of the tsunami. A formula in

estimating the tsunami pressure acting on a vertical wall was obtained.

From the experimental results and the field

survey, it appeared chat the first bore of

about 2.5m high" attacked the seawall but if was net slided, and when the water surface was fell down about 5.8m on account of the ebb by

the first wave, the second bore of about 5.0m

high attacked and the seawall was damaged at

last. One of the remarkable features of the

bore-type tsunami is that the time variation

is extremely rapid. For example, the second

bore attacks when the water level is graatlv

decreased by the ebb of the first wave and

shows an abrupt change such that the

overtopping begins only within 6 seconds after

the detection of the oncoming bore around the

locations of about 50m in front of the

seawall. Therefore, caissons of the saawalJ

were slided and damaged on account of the

great difference of the water levels between

before and behind the seawall, because

sufficient amount of water could not ante;/

behind the seawall due to the short time.

This paper is revised from that presented in

the Proceedings of the 31st Japanese

Conference on Coastal Engineering, 1984.

REFERENCES

1) Fukui, Y. et al.: Investigation on tsunami

(II), Proc. 9th Japan Conf. Coast. Eng.,

pp. 50^54, 1962, (in Japanese).

2) Tanimoto, K.: On the hydraulic aspects of

tsunami breakwaters in Japan, Proc. Inc.

Tsunami Syrnp., 1981, IUGG Tsunami

Commision, pp. 423^435, 1983.

3) Tanimito, K. et al.: Field and laboratory

investigations of the tsunami caused by

1983 Nihonkai Chubu Earthquake, Tech. Mote

Port and Harbour Res. Inst., No. 470, 299p.,

1983, (in Japanese).

4 ) Tsuruva , H. et al.: Experimental study on the deformation and run-up of tsunami in

shallow water, Proc. 31st Japan. Conf.

Coast. Eng., pp.237'\<241, 3934, (in

Japanese )

5) Tanimoto, K. et al.: Field survey and some

investigations on the tsunami damage

produced by 1983 Nihonkai Chucu Earthquake,

Proc. 31st Japan. Conf. Coast. Eng.,

00.252^256, 1984, (in Japanese).

6) Tanimoto, X. et al.: Experimental study on tsunami force and investigations of the

cause of seawall damages d u r : n g 1933

Nihonkai Chubu Earthquake, Proc. 31st

Japan. Conf. Coast. Eng., pp. 257^261, 1934,

( in Japanese )

186
A Summary of the Tsunami History of the United States

BY

Herbert Meyers 1 P„ Ac Lockridge, and J. F. Lander

ABSTRACT

The National Geophysical Data Center is

reviewing its global digital file of available

tsunami data for the first time with the ultimate goal of presenting U.S. data in a unified

catalog. While some of these data have appeared

in separate catalogs in the past, the unified

catalog will allow the special characteristics

of tsunamis in Alaska, Hawaii, and the West

Coast states to be described and compared. Some

of the early findings indicate that Alaska is

primarily a tsunami-generating region and

receives almost no impact from tsunamis

generated elsewhere. Hawaii and the West Coast

are primarily tsunami-receiving regions and have

had little impact from tsunamis generated by

local earthquakes. Most of the damage and fatalities from tsunamis in Hawaii in historical

times resulted from a single event, the 1946

tsunami originating in the Aleutians. Similarly

the most destructive tsunami for both Alaska and

the West Coast was the 1964 Gulf of Alaska tsunami .

KEYWORDS: Alaska; Hawaii; Tsunami; West Coast

1. INTRODUCTION

"Tsunami" is customarily defined as a series of waves generated by an impulsive disturbance in

the ocean. The definition includes water waves produced by ocean-bottom displacement from

earthquakes, landslides, or volcanoes. The

National Geophysical Data Center (NGDC) located

in Boulder, Colorado is in the process of com- piling and reviewing all available data about

tsunamis from local or remote sources which have

affected the U.S. NGDC ' s effort in compiling

tsunami data is complemented by the efforts of

two other components within the National Oceanic

and Atmospheric Administration: the Pacific

Marine Environmental Laboratory, a research

institution; and the National Weather Service,

which operates the Pacific and Alaska Tsunami

Warning Centers. The following tsunami parameters are included in the tsunami data base at

NGDC: (1) information about the cause (e.g.,

volcanic, earthquake, landslide), location of

source, depth and magnitude of earthquake where

appropriate, and (2) information about the

effects of the tsunami including maximum runup

height, tsunami magnitude or intens ity^ , local

runup, damage, deaths, arrival time, and travel

time for each location experiencing the tsunami.

2. SIGNIFICANT FACTS FOR ALASKAN TSUNAMIS

The Gulf of Alaska and the eastern and central

Aleutians provide perfect geologic conditions

for tsunami generation. The Pacific sea-floor

underthrusts the Alaska gulf and peninsula and

east and central Aleutian Islands producing

large magnitude thrust earthquakes - the kind

most frequently associated with tsunami generation. Alaska, like ether U.S. coastal areas prone to tsunami damage, has a relatively short

and incomplete history-. Although reports span

200 years (1788 to present), the record is probably incomplete up to 1900 or later due to low

population density in many areas and lack of

written tradition by the inhabitants (native

Americans, trappers, fishermen).

Since 1788, 34 tsunamis have been reported in

the Alaska-Aleu tian region. At least six of

these tsunamis were caused by landslides in the

bays of southeastern Alaska (Lituya and Yakutat

Bays). Deep fiords with steep sided glacier

valleys and active glaciers make these bays

dangerous source areas of giant local tsunamis.

These landslides have produced huge waves hundreds of meters high. Three tsunami events

in the eastern and central Aleutians have been

connected to volcanic eruptions in the area. Table 1 lists the 13 events that are classified

as destructive. All of them were caused by

local earthquakes. Note that the most significant number of fatalities and damage is from the

1964 tsunami (see Fig. 1).

^T sunami magnitude is defined by Cox and others

(1967) as m = log 2 H where H is the maximum runup

height in meters. Tsunami intensity is defined

by Solov'ev and Go (1974) as I = log2^2H.

Maximum runup height is maximum water height

above sea level for an event. Local runup is

maximum water height above sea level at a specific location for an event.

^•Chief, Solid Earth Geophysics Division, NOAA,

National Geophysical Data Center, 325 Broadway,

Boulder, Colorado 80303

187
More than one-third of the 27 local earthquakes

of magnitude 7.5 and greater produced tsunamis

in Alaska although most of them had runups of

less than one meter (see Fig. 2). Figure 3

shows that although Alaska has had some very

destructive events, there are greater risks in

other regions of the Pacific such as Japan,

South America and Southeast Asia. However,

Figure 4 shows that a greater percentage of

Alaskan earthquakes above 7.5 result in tsunamis

than in other regions except for the west coast

of South America. Alaska has experienced relatively little effect from tsunamis generated

elsewhere in the Pacific Basin. For example,

Alaska has experienced tsunamis from six earthquakes near Japan, but the runups were very

small (see Fig. 5). The largest amplitude from

such a remotely generated tsunami was 1.8 meters

recorded at Attu from the 1960 Chile tsunami.

Of the thirteen destructive tsunamis recorded

in Alaska, a few have generated massive tsunamis

in the Pacific Basin and have produced extensive

damage outside Alaska, particularly in Canada,

Hawaii and on the west coast of the United

States. Historically, Alaskan tsunamis have not

produced damaging waves in Japan, but western

Aleutian earthquakes are a potential threat. A

summary of the damage that occurred outside the

United States from Alaskan tsunamis is shown in

table 2. The April 1, 1946 tsunami, generated

in the east Aleutian Islands, produced some damage to boats and harbor areas in Chile. The

Central Aleutian Island tsunami of Mar. 9, 1957

caused minor damage in El Salvador; and reportedly damaged boats and houses on N.E. Honshu

Island, Japan. The Mar. 27, 1964 tsunami, that

originated in the Gulf of Alaska, caused $10

million damage in British Columbia, Canada, and

caused damage to the oyster and pearl harvest on

the Sanriku coast of Japan. Figure 6 shows the

locations throughout the Pacific Basin that

recorded the 1964 event.

Alaska has significant risks from earthquake

generated tsunamis, and from those generated by

volcanoes, landslides and icefalls. When

seismic gaps are closed by large earthquakes in

the Gulf of Alaska, along the Alaskan Peninsula,

and in the central Aleutians, massive destructive tsunamis may be generated. This area also

has the most apparent potential for producing

damaging tsunamis on the West Coast of the

United States.

3. SIGNIFICANT FACTS FOR HAWAIIAN TSUNAMIS

The earliest recorded tsunami in Hawaii occurred

in 1813. Since that date more than 70 tsunamis

have been observed in Hawaii. The Hawaiian

record is essentially as complete as possible

back to 1837. Unlike other regions of the

Pacific, Hawaii receives most of its tsunami

damage from distant earthquakes. Hawaii is

positioned over a "hot spot" that lies beneath

the Pacific plate. The volcanism that is

responsible for the formation of the islands

themselves is also responsible for the earthquakes that occur there. These smaller magnitude quakes (only one event recorded a magnitude

greater than 7.0) are not of the type that produce large tsunamis. In the last 100 years

there have been only seven tsunamis produced by

local earthquakes, two of which have caused

damage. Local earthquakes constitute only a minor tsunami risk (48 deaths) in comparison to

tsunamis generated by distant earthquakes (239

deaths )

Hawaii has received tsunami damage from 15

distant earthquakes. At least 63 such other

tsunamis generated in every quadrant of the

Pacific Basin have been recorded or observed in

Hawaii (Fig. 7). Eleven of these have achieved

the maximum runup recorded outside their

generating areas as they rolled up on the shores

of the islands in Hawaii. Figure 8 indicates

the maximum runup height recorded in Hawaii from

tsunamis generated in regions throughout the

Pacific. For example, Hawaii has experienced 12

tsunamis from earthquakes near Japan. The

largest runup was 9 meters. However, there are greater risks to Hawaii from earthquakes near Chile or Alaska. Only tsunamis generated in the

southwest Pacific whose energy is confined by

island chains and the small tsunamis generated

along the coasts of North and Central America

have not caused damage in Hawaii. Earthquakes

near Alaska, Chile and Kamchatka have caused

tsunami fatalities in Hawaii. Property losses

resulting from tsunamis in Hawaii have been

extensive and nearly 400 people have lost their

lives. The most damaging events were the April

1946 (see Fig. 9) Aleutian event and the May 1960

Chile event (Table 3). Of all the Pacific

regions, Hawaii is clearly the most vulnerable

to tsunami damage from outside sources.

4. SIGNIFICANT FACTS FOR WEST COAST TSUNAMIS

The Pacific Coast of the United States lies

along a plate boundary. However, the Pacific

plate is not being subducted beneath the con- tinental plate as in the areas of Alaska, the

Aleutians, and South America. Instead the two

plates are sliding past each other in a complex

system of faults, the most famous of which is

the San Andreas Fault. The epicenters of most

earthquakes lie on land. For these reasons, the

earthquakes on the Pacific Coast of the United

States do not produce tsunamis large enough to

travel across the ocean basin and be recorded

elsewhere. However, the historical records for

this area indicate that several earthquakes have

produced local tsunamis.

Among the Pacific coast states, California has

experienced the greatest number of tsunami

events, and is the only one to have had local

tsunamigenic (tsunami generating) earthquakes.

There were at least 12 incidents of unusual sea movement and flooding related to local earthquakes in the 1800' s. The validity of most of

these reports is questionable. Destructive

188
local tsunamis have not been generated in the

!

last one hundred years on the west coast of the

United States. The major earthquakes of this

century in California have generated only small

tsunamis

There were seven events which caused damage on

the west coast of the U.S. (see -Table 4). Five

of these were remotely-generated tsunamis that

caused more that $485 million (in 1980 dollars)

in property damage and 16 deaths on the west

coast. These events include: the tsunami produced by the Kamchatka earthquake of November 4,

1 1952 that damaged fishing boats and the harbor

at Crescent City; the March 9, 1957 tsunami

generated in the Aleutian Islands that smashed

ships and damaged docks in San Diego Bay; a magnitude 8.6 earthquake off the coast of Chile

on May 22, 1960 that caused damage in Crescent

City; the most damaging event by far, the March

28, 1964 tsunami generated in the Gulf of Alaska

that caused extensive damage on the coasts of

Washington and northern California (see Fig. 10)

and the November 1975 tsunami generated by an earthquake in Hawaii that caused minor damage to

Catalina Island, California.

Figure 11 shows the maximum runup height in

meters recorded on the U.S. West Coast from tsunamis generated in other regions. For example,

there were 6 tsunamis reaching the west coast of

the U.S. from earthquakes near Japan. None of

these produced high runups. There are greater

risks to the U.S. west coast from events in

Alaska and Chile.

On the east coast of the United States only

Puerto Rico and the Virgin Islands have recorded

damaging tsunami events. Five tsunamis have

been reported there. The event in 1918 killed more than 40 people and caused destruction in

Aguadilla, Mayaguez, and Punta Agujereada where

the runup reached 6 meters. A tsunami in 1868

in the Virgin Islands killed at least 7 people

and did extensive damage to ports and shipping.

5. SUMMARY

A summary of the available data for tsunamis on the shores of the United States is given in

table 7. There have been at least 600 deaths

and more than $500 million in damage in the past

200 years. Historical records of tsunami events

are valuable because they indicate the areas most likely to spawn tsunamis, the size of the

earthquakes most likely to generate tsunamis in

these areas, and areas most likely to incur the

heaviest amounts of damage from tsunamis once they are generated.

139
Figure 1. Tsunami damage to Seward port facilities from

the tsunami of March 27, 1964. The event resulted in more

than $280 million (in 1980 dollars) in property damage to

Alaska, California, and Hawaii.

ALASKA

EARTHQUAKE MAGNITUDE VS, NUMBER OF

TSUNAMI EVENTS 8 MAXIMUM RUNUP

20-

18-

16-

Number 14- of 12- Earth- 10-

Mas runnuM 30

7.0 »7.4 7.5 -7.8 Earthquake Magnitude

8.0 B Tsunanmigenic Earthquakes- - -

| | Non-tsunamigenic Earthquakes

/

RUNUP

< 0.0

Figure 2. Number of earthquake events of various magnitudes

generated on the Alaska coast, number of tsunami events

produced by these earthquakes, and maximum runup recorded

for events in each magnitude range.

190
A COMPARISON OF ALASKA, HAWAII, AND THE WEST COAST OF NORTH AMERICA WITH OTHER

TSUNAMI-GENERATING REGIONS IN THE PACIFIC

NUMBER OF LOCAL TSUNAMIS

in last IOO years

5 10 1 5 2 0 2 5 50 55

I r

1 t r 1 1 r— fn

40 4 5 50 55 —i

1 1 r~ 60 65 70 75 80 85

—i 1 1 1 r T~~ Hawaii

S. Pacific/New Zealand

New Guinea/Solomon Is.

Indonesia

Philippines

Japan 71no

Kamchatka/Kuril Is.

Alaska/Aleutian Is.

W. Coast N. America

W. Coast S. America

No Damage

CD Slight Damage

Moderate to Severe Damage

Figure 3. Comparison of the total number of undamaging events, the

total number of damaging events, and the number of events that caused

at least moderate damage.

NUMBER OF EARTHQUAKES

Since 1900

5 10 1 5 2 0 2 5 50 5 5 40 4 5 50 55 60 6 5 70 7 5 80 85

T 1 f ' 1 1 1 1 1 1 1 1 1 1 1

I 1 -O- Hawaii

S. Pocific/New Zealand New Guineo/Solomon Is.

Indonesia

I | Philippines

I 1 Japan

| ~] Kamchatka/Kuril Is.

| 1 Alaska/Aleutian Is.

1 ~| W. Coast N. America

[ 1 W. Coast S. America

I I 7.5 and Larger magnitude earthquake

I I 7.5 and Larger magnitude earthquake

that Generated Tsunamis

Figure 4. Numbers of earthquakes of magnitude greater than or equal to

7.5 that occurred in each region and number that generated tsunamis in

each region.

191
Max,

Runup

inMeters

ALASKA

Figure 5. Maximum runup in meters produced on the coasts of Alaska

by tsunamis generated in other regions. Numbers in parentheses are total numbers of earthquakes in each source region that generated

runups in Alaska.

PACIFIC BASIN TSUNAMIS

RESULTING FROM EARTHOUAKE OF MARCH 27, 1964

(ALL RECORDED RUN-UP HEIGHTS)

Figure 6. Pacific Basin tsunamis resulting from earthquake of March 27,

1964 (all recorded run-up heights).

192
Figure 7. Tsunamis recorded in Hawaii, 1900-1983.

(LOCAL)

Figure 8. Maximum runup in meters produced on the coasts of Hawaii

by tsunamis generated in various source regions of the Pacific

Ocean Basin. Numbers in parentheses are total numbers of earthquakes in each source region that generated runups in Hawaii.

&

( 17 )

CHILE

193
Figure 9. A major wave generated by the earthquake of April 1, 1946

in the Aleutian Islands striking the beachfront area at Hilo.

194
Figure lOa. View of normal conditions at intersection in Crescent

City, California.

Figure 10b. View of same area as above showing the damage

produced by the tsunami generated by the earthquake of

March 27, 1964 in the Gulf of Alaska.

195
WEST COAST

Figure 11. Maximum runup in meters produced on the West Coast of

the United States by tsunamis generated in various source regions

of the Pacific Ocean Basin. Numbers in parentheses are total

numbers of earthquakes in each source region that generated

runups on the West Coast.

196
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198
TABLE 4

SUMMARY OF TSUNAMI DAMAGE ON THE WEST COAST OF THE UNITED STATES

EVENT DATE SOURCE REGION LOCATION OF EFFECT DAMAGE

1812 12 21 S. CALIFORNIA, USA

1859 09 24 N. CALIFORNIA, USA

1946 04 01 E. ALEUTIAN ISLANDS

1957 03 09 CENTRAL ALEUTIAN IS.

1960 05 22 S. CHILE

1964 03 27 GULF OF ALASKA

1975 11 29 HAWAII

SANTA BARBARA, CA

HALF MOON BAY, CA

SANTA CRUZ, HALF MOON BAY, CA

LA JOLLA, SAN DIEGO, CA

WEST COAST

WASHINGTON

OREGON

CALIFORNIA

CATALINA I., CA

SHIP AT ANCHOR DAMAGED

SCHOONER DAMAGED

MINOR DAMAGE (1)*

MINOR DAMAGE

$500,000 DAMAGE

$5 MILLION DAMAGE

$0.6 MILLION DAMAGE (4)

$10.9 MILLION DAMAGE (11)

$2,000 DAMAGE

* Number of deaths in parentheses

TABLE 5

SUMMARY OF U.S. TSUNAMIS

• U.S.

EVENTS

NON-U.

EVENTS

EARLIEST OBSERVATION: 1788 1811

TOTAL TSUNAMIS

63 98

TOTAL TSUNAMIS WITH RUNUP GREATER THAN 1.0 M: 23 19

TOTAL DAMAGING TSUNAMIS: 26 13

TOTAL DEATHS

>490 130

TOTAL ESTAMATED DAMAGE IN MILLIONS OF U.S. 1980 DOLLARS: >$480 >$80

&

199

An Estimation of Tsunami Gonerated by Submarine Landslide

b y Sin— Iti Iwasakij& Hiroshi Takahashi

ABSTRACT

On the estimation of heights of tsunamis generated by

landslides the following was found through the laboratory exper iment. nuner i ca 1 computaions and linear

theory.

(1) In case of tsunamis generated by submarine land- slides the maximum wave generating efficiency

(wave energy generated by landslides / potential

energy loss in the water or the initial potential

energy ( net. submerged ) of the sliding body ) is about 10*.

(2) In case of tsunamis due to landslides plunging

into the water surface. 8 to 15* of kinematic en- ergy of the sliding body is transformed into the wave energy And. as indicated (1) . about 10* of the potential energy lost in the water is added to the wave energy.

KEY WORDS efficiency . landslide . tsunami

1. INTRODUCTION

The mam cause of tsunamis is submarine earthquakes,

but. other causes, landslides and volcanic eruptions . are also thought to generate tsunamis Espec i a I I y. submarine I ands I i de- i nduced tsunamis are not accompanied

with so-called foreshocks as in the case of earthquakes and volcanic eruptions And. tsunamis generated

by landslides plunging into the water body can occur

inland lakes and reservoirs where normally considered

to be safe for such a disasterous giant water waves. Predicting the wave heights for various dimensions and speeds of the slides for site hazard evaluation pur- poses is of clearly importance

On the estimation of heights of tsunamis generated by

landslides, one of the bases is WIEGEL ' s (1955) laboratory experiment One of the conclusions is ’ about 1*

of the initial potential energy ( net. submerged ) is transformed into the wave energy. " Based on this

conclusion . STRIEM&MILOH (1976) calculated the wave

height assuming the generated wave is one-dimensional

solitary wave Their submarine landslide model is shown

in Figl MURTY(1979) using same landslide model and wave system, also apllied WIEGELs conclusion to the submarine landslide induced tsunami occured on April

27. 1975. in Kitimat Inlet. Br 1 1 1 sh Columbia, and showed

good agreement with the observat i on. But. in calculating

the solitary wave energy. MURTY commited simple mistake

Inversely evaluated efficiency from the combination of

the observational data and the correct solitary wave energy i s 32. 2* This value is far from the WIEGEL s conclusion It is suggested that WIEGEL s result could

not be applied simply to all of the I ands I i de- i nduced

tsunami evaluation

In this paper, both submerged and not submerged land- slide induced tsunamis are investigated from the la- boratory exper i men t. numer i ca I computations and linear

theory And. an simple method for tsunami height estimation is presented

2. RECONSTRUCTION OF THE WIEGEL 'S EXPERIMENTAL DATA. The experimental set-up is shown in F i g. 2. Rectangu I ar box allowed to fall freely and wave heights were measured at the sta. 1 and sta. 2. The width of the model

basin was nearly equal to the width of the falling box The generated wave could be treated as one-d i mens i ona

wave. Parameters were the basin depth d. initial submergence Zo and bo.x weight W The range of parameters

are shown in upper leftside in Fig. 3. In Fig 3. the ordi- nate is the initial potential energy (net. submerged ) of the falling box (PE) and the abscissa is generated

wave enrgy (WE) WE was obtained by the equation (1) under the assumptions that the generated wave had deep water wave characteristics and the effect of dispersion was negligible at the sta. 1.

W E = p wg 2 H

2 T

2 / 1 6 rr ( 1 )

where g denotes the gravitational constant, p w the density of the water. H and T the maximum wave amplitude (trough to crest) and period of the wave measured

at the sta. 1. respectively.

In Fig. 3. 1* line of efficiency is added For large PE. the coincidence of the experimental data and 1* line

is rather good. but. for small PE. the efficiency is less than 1* for all of the experimental data. For explaining this tendency, the concept of Virtual Mass is i ntroduced. When a solid body moves in water in unsteady manner, the water around the body moves with

it. This water mass surounding the body is called Added

Mass (Virtual Mass ) Cons i der i ng the Virtual Mass, the available potential energy for wave generation is small compared with that estimated by WIEGEL. Virtual Mass depends on the shape and dimensions of the solid

body and density of the water. So that, the smaller the

PE. the smaller the ratio of available potential energy

to PE. It is well explained the discrepancy of the ex- perimental data and 1* line of efficiency for small PE

in Fig. 3. And. introducing the resistant coefficient Cd. the equation of motion for the velocity V of the falling

box i s

{ p + p v) • Vo dV/d t = ( p — p w) • Vo* g

+ (1/2) • pwV 2 S • Cd (2)

1. Research Worker, 1st Laboratory. Hiratsuka Branch,

National Research Center for Disaster Prevention.

Japan.

2. Dr. of Sc i ence D i r ec tor-Genera I . Nat i ona 1 Research

Center for Disaster Prevent ion. Japan

200
where p denotes the density of the falling box. p\

-Wv/Vo. Wv the Virtual Mass. Vo the volume of the box and

S the square measure of the bottom of the box. Wv is expressed from Fig. 4 taking the horizontal scale of the box L instead of 2a

, 2 x , Wv = Kl • - >r

1 L / 4 (3)

In the WIEGEL's exper i ment. Kl-1. 51. Equat i on (2) can be solved straightforwardly as follows

V = .fa / 0 • tanh ( Ja 0 t ) ( 4 )

a = ( p - p ») g / (p+pv)

Integrating ( 6) numer i ca I I y. the maximum wave generating efficiency is about 10% when the initial box submergence Zo- 0 . 0 . In the WIEGEL s experiaentai set-up.

the wave energy was calculated only one-side of the falling box. Then, to apply this result to Fig. 3. we must half the value of 10% The ealuculated efficiency is rather high compared with the experimental data It is indicated that the wave is not developed as expected

from the linear theory.

3. NONLINEAR NUMERICAL COMPUTATIONS USING THE BOUNDARY ELEMENT METHOD

In this section, the results of nonlinear computations

are compared with the linear theory. In the domain

shown in Fig. 6. using the Boudary Element Method, non- linear numerical computations were made All the physical quantities were nond i mens i ona I The basic

equation for the velocity potential <t> is

A <t> = 0 ; Q ( 8 )

0 — ( 1 / 2 )

• p w S * Cd/|(p>4-/0v) • Vo}

In the WIEGEL's experimental cond i t i ons. the falling

box reached the bottom of the basin immed i ate I y. Then,

the motion of the box can be treated as constant acceleration motion.

V = a • t

( 5 )

For convenience in later d i scuss i ons. phys i ca 1 quantities are written in nond imens i ona I form Using box horizontal scale L. grav i tat i ona I constant g and density

of the water p w. ve I oc i ty. energy a nd t ime are nond i mens i ona I i zed by ,/gL. p wgL^ and v/L/g respectively

Taking a as the abscissa in Fig. 5. WIEGEL’ s experimental data are reconstructed. For small a. the efficiency is less than 1 %. but for large a . there are upper limitations depend on Zo and d From the linear theoty (KAJIURA. 1963). the wave height

p when the box falls with constant acceleration is expressed in terms of time dependent Green s function

as foil ows j

r

r) = — ( a / 2 7T ) G t JJ0

z=o

d r d x o ( 6 )

s i n ff ( t — r ) cos m x

(coshim (d + Zs)}— cosh)m (d + Zs-AM))]

/cosh(md)dm (7)

where

cr 1 = m tanh(md) - 2 = ( x — x 0 ) 2

Zs = Z o— a r^/2 ; Zo< 0

where A denotes the ! ap I ac i an Boundary conditions are

<t>/^ t 4- ( 1 / 2 ) {(^ 0/5 x ) 2 +

( 3 <t> /<5 z ) } + 77 = 0 ; Sl± ( 9 )

~b <t> / ^ n = nz 1 /"$ 1 ; S lJt (10)

4> /ft n = 0 ; S 2_t (ID

"d / 3 n = n z • V ; S3± (12)

where nz denotes the z component of the unit outward

normal vector n Equations (9) and (10) are nonlinear

In this computations TL-30 0 and d— 1 6. 0 were fixed

Constant velocity fallings were assumed. The fall began

at r *0 and finished at r - r f . Var i ab I es were AM. V and ZoAccording to the linear theory, if the falling velocity is constant. 77 is expressed as follows with the aid

of Green’s function (7).

7? = - (V/ 2 tt)J

(Gt

to

Gt

y0 dxo

T-0 T-Xf

( 13 )

This integral is independent of V Then, wave height 77 is propotional to the constant falling velocity V Dependence to 77 of AM and/or Zo is not clear like

that.

Fig. 7-1 and Fig 7-2 are the comparisons of the nonlinear numerical computations and the linear theory

The ordinate is 77 max/V and the abscissa is Zo. Discrepancy between the computations and the linear

theory is clear for V£0 16 in Fig 7-1 and for all V in Fig7-2. In the WIEGEL’s exper iment. the falling body

thickness AM-1 00 Then, it is clear from the Fig. 7-2

201
that the low efficiency of the experimental data shown

in Fig 5 is due to the nonlinear effect.

4. TSUNAMI GENERATING EFFICIENCY DUE TO THE LANDSLIDES

PLUNGING INTO THE WATER BODY. SLINGERLAND & VO I GHT ( 1 982) presented a empirical

formula through the hydraulic studies.

+ log(r ) + (3 / 2 )

• log ( y ) (20)

Substituting (19) into (20), after some algebla. we get

(WE/Ek) = 0 . 2 6

log(77max/d) = — 1 „ 2 5 + 0, 7 1

• I o g ( K E ) (14)

where 77 max-max i mum wave amplitude at a standard distance r/d-4. 0 directly in front of a s I i de. d-aver age depth offshore from slide s i te. r-radi a I horizontal dis tance from the point of slide impact and KE-nondimensional slide kinematic energy Laboratory experimental data denoted by black dots shown in Fig. 8. Confidence limits at the 95* level are indicated by

dashed line. The back cal ucu 1 at ions of s I i de- i nduced

tsunamis occured in Lituya bay. Alaska and Disenchantment bay. Alaska are also shown in Fig. 8. Informations

of both slides are shown in Table. 1 Now. we consider the one dimensional propagat ion. The energy of solitary wave per unit width is expressed as

fo 1 I ows

E s - (8/3 p » e /77T

( 15 )

y = t) max/ d

Then, in the experimental condi t ion. the wave propagated

through a 180 arc. The total energy is

W E = E s • n r ( 16 )

And. slide kinematic energy is

Ek = (1/2) P s V o V f ( 17 )

where Vo denotes the volume of the slide and Vf the final velocity of the si ide Nondimensional form is

K E = E k/ (pd 3 gd ( 18 )

Accor d i ng to Fig. 8. there is little d i f f erence chang i ng

the coefficient of log(KE) from 0 71 to (2/3) in the equation (14)

I og ( 7 ) = - 1 . 2 5 +-(2 / 3) log(KE)

(19)

Substituting (15) into (16) yields

log (WE) = I og ( 8 t / 3 /T)+-log(pwg d

3

)

It is suggested from Fig. 8 that tsunami generating

efficiency is low for the Disenchantment bay slide

compared with that for the Lituya bay The reason of this difference can be considered as follows The

material of the Lituya bay slide was so 1 1 (

p

=2. 7g/cm*

)

but the Disenchantment bay slide was glacier (p-1. Og/

cm^) So that, for the case of Lituya bay. the slide

generated tsunami even in the water like submerged

landslide did. But. for the case of the Disenchantment

bay. the glacier did not generate tsunami in the water To confirm this specu I at ion, we consider the ideal slide whose KE is the same with that of the Disenchantment bay glacier and p is the same with that of the Lituya bay slide. The range of KE of the Disenchantment

bay glacier varies from 130 to 249 depending on the estimation difference of the final velocity In Fig. 8. A and B denotes the point of KE-130 and 249 respectively The efficiency in both cases evaluated by the present method is 15* and 8* And 10* of the potential energy loss in the water of the ideal slide transformed into the wave energy. Then, the total wave height

generated by ideal landslide denoted by rj max/d comes

to 1 30. The points A and B move to the points A

' and B

as shown in Fig. 8 This values considered to be reasonable compared with the experimental data. It is found

that to know the density of the slide is essential to evaluate tsunami height exactly.

5. CONCLUDING REMARKS

On the generation of tsunamis due to I ands i i de. var t- ous parameters are c3Verned. But. for the simple estimation of tsunami height, it is found that the following three parameters, the potential energy lost in the water, kinematic energy and density of the slide, are essential The tsunami generating efficiency became^

large as the initial submergence of the slide becomes

small The maximum generating efficiency is about 10* For the case of landslides plunging into the water

body. 8 to 15* of kinematic energy of the slide transformed into the wave energy REFERENCES

Ka ji ura. K. . The leading wave of a tsunami. Bui I of Earthq Inst. 41. 535-571. 1963. Murty. T S . Submar t ne slide-generated water waves in Kitimat In I et. Br 1 1 i sh Columbia. J Geophys Res

84. 7777-7779. 1979. Saunders. H. E. . Hydrodynami cs in ship design, Soc of Naval Architects and Marine Engineers. 1964

SI ingerland. R. 4 B. Voi ght. Eval uat i ng hazard of land- slide-induced water waves. J. of Waterways. Port. C oastaland Ocean Division. ASCE. 108. WW4. 504-512. 19 82

Striem. H. L. 4 T. Mi I oh. Tsunami s induced by submarine

slumpings off the coast of Israel. International

Hydrographic Rev. 53, 41-55. 1976

Wi ege 1. R. L. . Laboratory studies of gravity waves generated by the movement of a submerged body. Trans Amer Geophys Union. 36. 759-774. 1955

202
I

Fig. 1 A model of submarine land slide (after STRIEM &

MILOH. 1976)

4. Sel-up (or Series II

Fig. 2 WIEGEi/s experimental set-up.

203
u>6

— cr o <->

u.O

tom

>-

a:

cj UJ O 2

O Ui>

<

*

<

5

Xo

a:

Q.

a.<

8

204

Fig.

3

Relationship

between

energy

of

wave

disturbance

and

initial

potential

energy

(net.

submerged)

of

a

vertically

falling

box.

1%

line

of

efficiency

is

added

(after

WIEGEL.

1955)
or Rectangular

Section

Fig. 4 Added Mass Values for two

shape, (after Saunders, 1964)

Wv'M11

a/b Hi

O.l 2.Z3

o.z 1.90

0.5 1.7

I 1.51 z 1.36

5 1 . 2 ! 10 1.14 dimensional geometric

Fig. 5 Reconstruction of the WIEGEL/s experimental data

in terms of modified acceleration by the concept of

the Added Mass. 5# line of efficiency is obtained by

the linear theory.

205
206

Fig

6

Schematic

view

of

the

domain

for

numerical

compu

tat

i

ons.
o r^j • LD

cf

r^J • LD

cr

207

Fig7-1.

2

Comparisons

of

the

linear

theory

and

non

linear

numerical

computations.
j

dimensionless

)

Fig. 8 Relationship of maximum dimensionless wave

amplitudes (at r/d=4. 0) as function of slide dimension

less kinematic energy(after SLINGERLAND & VO I GHT) . a little portion is modified.

Vo ( c ) p (g/ea^) Vf (ca/sec) d (ca)

L1TUYA BAY

13

3 . 1X10 2 . 7 3

5 . 6 — 6 . 9X10 1 . 2X10^

DISENCHANTMENT BAY

/ 3 2 . 9X10 1 . 0 6. 0 — 8 . 3X10^ 3

8 . 0X10

Table. 1

208

A Computer Program for Producing Tsunami

Travel-Time Chart

BY

Masami Okada’ and Masajl Ichikawa 2

ABSTRACT

Tsunami travel time chart is used to predict

quickly the tsunami travel and arrival time

to the objective location or coast. Many

charts, prepared for the operational tsunami

warning system, were made by hand with much

time and labour. We have been developing he

software for producing tsunami travel time

charts which will be applied in the newly

computer system called Tokyo-EPOS. Earthquake

Phenomena Observations System, in JMA.

Using various data bases of bathymetry, the

travel time is calculated on the triangular

grid system on the basis of Huygens’ law.

The travel time chart is calculated firstly

on the fine mesh for small area near tide

station or wave source and the results of

preceding calculation are adopted as the

initial values of travel time for the

following computation with a rough mesh

covering a wide area. The method of

computations is described and a trial

production for the north Pacific ocean is

shown in this paper.

KEY WORDS : Tsunami, Tsunami Travel Time

Chart

1 . INTRODUCTION

Tsunami travel time charts are used by

tsunami waring systems to provide "real time"

predictions of the travel and arrival times

of a tsunami, when a large earthquake

occurred under the sea bottom. Each travel

time chart is centered on a hypothetical

center and shows the wave fronts emanating

from that center, for which warnings are required. The chart centered on a realistic

tsunami wave source is usually compiled after

the tsunami arrived along the coast and

observed at many tide stations.

In most cases the wave length of tsunami is so long that the wave velocity is given by

the square root of the product of the local

water depth and gravitational accelerations.

Travel time between any two points is given

by the integral of reciprocal velocity along

the wave ray path. Although there is no problem in hydraulics, it is impossible to

calculate theoretically the travel time for

the actual bathymetry, because the bathymetry

is too complicated. In practical

applications, these integrals are carried out

numerically.

In ray tracing method, the ray path should be

determined prior to the integral of travel

time. The amount of spreading of the ray

paths is also of interest since it gives a measure of the energy concentrations of

tsunami. Nevertheless this method cannot be

applied for the operational tsunami travel

time chart, because the shadow area in the

sea may come out in tsunami propagation map

behind islands and shoals. In our software,

the tsunami travel time chart is produced by

Huygens’ construction, which involve the

diffraction effect.

2 , BATHYMETRIC DATA BASE

We are able to use the ET0P05 world 5-minute

gridded Bathymetric-Topographic data base,

which was sent last year from the National

Geophysical Data Center, NOAA in USA under

the auspices of UJNR and through the courtesy

of Deputy Director. J.F. Lander. This data

base is very convenient to show the world

wide bathymetry, and we should also point out

that the values in the file indicate a local

average elevation and some of them near the

shore lines confuse sea with land. Thus we have revised many valued near shore lines in

the northwest Pacific, the region of 20-48®

and 1 20-1 52* E.

It is reasonable that the error in the

tsunami travel time chart depend on not only

the accuracy but also the grid size of the

bathymetric data base. We must pay attention

to the values in the shallow water in which

tsunami propagate slowly. In many cases of

calculating the tsunami travel time chart to

the tide station located on shore, more

fine/small gridded data of bathymetry are

1

: Chief. First Research Laboratory.

Seismology and Volcanology Division.

Meteorological Research Institute. Japan.

2 Dr. of Sci . . Former Head. Seismology and

Volcanology Divition. Meteorological Research

Institute. Japan.

209
required additionally to represent small

topography. We are now compiling the

1-minute gridded bathynetric data bases for

the seas adjacent to Japan Islands.

Futhermore we are going to prepare very fine

gridded data bases (about 500m*500m and about

100m*100m ) for about one hundred tide

stations on the basis of the bathymetric

chart so as to represent adequately the

bathymetry of the port and harbour in which

the tide station is located. An example of

bathymetric data base for small area is shown

in figure 1

3. METHOD

The travel time chart is compiled by using

Huygens’ construction on the basis of the

modified triangular grid system. This method

may be complicated and take much more time

and memories in computer calculation than the

ray tracing one. However the ray tracing

method is serious for the operational chart

in having the shadow zone behind islands in

tsunami propagation map.

The modified triangular grid on Mercator map

is shown in figure 2. The values of sea depth are given at apexes of small triangle’

and traval time is calculated at all points

(squares and circles in figure 2). Each

value of the depth is calculated by a simple

interpolation in two dimensions from the

bathymetic data bases. Considering the grid

system, we should remark the difference in

distance between the apparent path on the

grid and real one. For example the linear

length of AE In figure 2 is 0.9 percent

shorter than the one on the mesh, that is the

connection of AB, BC. CD and DE. The

relative discrepancy in distance between two

points on the grid is usually less than one

percent, except such cases as S to V that is

start and end points being close. The

computational area for tsunami travel time

chart is not a rectangle but a parallelogram

shown in figure 3.

The local depth in each small triangle is

regarded to be linear interpolation of nearly

values at three apexes, in calculating the

traval time. Then the ray path in a triangle

is a part of cycloid curve represented as

fol low.

x = a( 6-sin6)

y = a( 1 -C0S 6 )

The travel time from any point, such as an apex point A or a side point S in figure 2. to the surrounding closely points shown by

solid symbols are calculated analytically,

assuming that the sea bottom decline

uniformly in each small triangle. The travel

time T between any two points are given as

fol low.

T = min l£Tl

s

here S = ISi . Si is the cycloid path

connecting two points on a small triangle,

and £Ti is the travel time along Si. The

minimum is taken over all possible pathes S on the grid system. Actually we calculate

the propagating time from one point to others

on the same triangle in the order of arrival

time.

It is reasonable that the calculations is

carried out on the fine mesh for the coastal area near the tide station and on the rough

mesh for the ocean, because the long wave velocity is low in shallow water and the

bathymetry is generally more complicated in

the coastal area. Thus the travel time chart

is compiled on the basis of fine mesh for

small area near a tide station or the tsunami source area, and the results of preceding

calculations are adopted as the initial

values for following computation on a rough

mesh covering a wide region. The sckema of

computational region is shown In figure 3.

4. TRIAL PRODUCTION

The tsunami travel time chart to Okada tide

station, Izu-Osh! ma island, about 100km

south-southwest of Tokyo, is compiled as a trial production by a computer program, named

SEIJI in developing. The result of

calculation is printed graphically by a line

printer. as the present software. Is independent from XY-plotter. The chart for

the north Pacific ocean is shown in figure 4. Wave source point in calculation is not

located on the exact position of the tide

station but 250m northwest of the station,

because the position of tide station at which

sea depth is 0m is considered to be located

on land in calculating and no tsunami arrive

there. The locations of the centered point

and the tide station are shown by the cross +and close circle in figure 1, respectively.

The travel time at the hypothetical tsunami source point is supposed to be 50 seconds

from tide station, and the time shown in

figure 4 is that from tide station.

The computation consist of three step. In the first step, the tsunami travel time chart

for the sea on the north of Izu-Ohima island

shown in figure 5a was compiled on the basis

of small grid of £D=200m. Secondly the one

for the sea near the district of Kanto and

Tokai. central southeast part of Japan, was calculated on the basis of middle grid of

£D=2000m. whose covering area is shown in

figure 5b. The results of these calculation

is used in determination of initial values

210
for the last step in which the calculation

was carried out for the north Pacific ocean. Mesh size used in this calculation is 0.25

degree of longitude, and the total number of

calculating points is 1.752. 092.

For these calculation we used the computer in our institute, and it take about 30 minutes

in CPU time. We cannot compare the computer

made charts with the conventional hand-made

one. for the conventional tsunami chart from

Okada tide station has not been compiled yet.

In other case, two kind figures for Japan sea

tsunami agree considerably to each other as a whole.

211
Fig.l. Example of bathymetric data base of small area near Okada

tide station, Izu-Gshima. The cross indicates the

hypothetical center of the tsunami travel time chart to

Okada. Travel time at the point is considered to be

50 seconds from the tide station.

Fig. 2. Modified triangular grid system. Sea depth is given at

the point of square and travel time is calculate at all , points (squares and circles).

212
Fig. 3. Schematic figure of the computational regions. Tsunami

travel time chart to a tide station is firstly produced

for the small area on the basis of fine grid, then the

preceding results are used as the intial values of

computation for the wide region.

213
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Fi g . 4 Trial

Okada

AD is

production of the tsunami

tide station to the north

0.25 degree in longitude.

travel

Paci f i

time chart

ocean. Grid

f rom

s i ze

214
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215
Fig. 5 Examples of the computational region. Figures of (upper)

and (lower) show the regions used for the first and the

second step calculation of the tsunami 'travel time chart

to Okada. respectively. The one for the third step is

the nort/L Pac i f ic ocean

216
Numerical Simulation and Experiment on Tsunami Run-Up

by

Takaaki UDA*

Atsushi OMATA ** Yasuhisa YOKOYAMA* *

ABSTRACT

Some comparisons are conducted with

numerical simulations and experiments on tsunami run-up on the coast where coastal

structures or sand dune exist. The numerical

simulation model used in this study is quite

simple, based on the nonlinear shallow water

equation. Temporal and spatial distributions

of the water level and velocity are compared.

Consequently it is found that the simple

model used in this study has an ability to

simulate tsunami movement with enough

accuracy in practical use. In the case of

the permeable breakwater, which is usually

constructed in Japan, detached breakwater

can be modelled by the bed level change in one grid.

1. INTRODUCTION

In May, 1983 the coast along the Sea of Japan

was damaged severely by a large tsunami caused

by the Nihonkai-Chubu Earthquake. After the

tsunami various preventive works for future

tsunami hazard was proposed. In the planning

of these works numerical simulations were

carried out in order to predict the critical

height of the coastal revetment or the

innundated area. Generally there exist sand

dunes or shore protection facilities like as the coastal dike or the detached breakwaters

along the coastal Japan. Around these

structures fluid motion may be considered to

be very complicated and simulation model to

predict tsunami run-up should also become

sophisticated, while usually numerical

simulation must be done in wide region along

the coast. Therefore for the practical use numerical simulation model should be as simple

as possible. Accuracy of the simulation is

generally verified only by the comparison with

the run-up height of the previous tsunami

without the comparison of the distribution of

water level or current velocity around the

structures in detail based upon the field

data. Consequently the effects of the

existence of the detached breakwater or coastal dune along the coast are not taken

into consideration sufficiently. From this

point of view in this study some comparisons

are conducted with numerical simulations and

experiments on tsunami run-up on the coast

where coastal structure or sand dune exist.

The numerical simulation model used in this

study is quite simple, based on the nonlinear

shallow water equation. In the experiments

the distributions of water level and velocity

are measured with time and space. Temporal

and spatial distributions of the water level

and velocity are compared with the calculated

values.

2.

EXPERIMENTS

The experiments were carried out in a fixed-bed wave basin, 30 m long and 5 m wide

with 1/100 bed slope. There is a pneumatic

type wave generator at the offshore end of the

basin(Fig. 1). The sandy beach of uniform

slope, coastal dune, coastal revetment and

detached breakwater are modelled, and the

distributions of water level and velocity were measured with time and space. Locations of

various measuring points are shown in Fig. 2.

Each model is of uniform shape in the

longshore direction. In the case of detached

breakwater permeable and impermeable models are tested. The impermeable one is composed

of concrete blocks fixed to the bed to prevent

from scattering. The measurements of the

water level and the velocity are done by using

the capaci tance-type wave gauges and the

propel ler- type current meters set 5 mm above

the bottom, respectively.

The solitary wave generated at the offshore

end propagated on the bottom slope while

changing its profile. At the location 15 m shoreward from the generator the wave front

dispersed to form two or three short-period

waves and then they broke. Finally the

breaking bore was incident to the models.

3.

NUMERICAL SIMULATION

The numerical simulation model used in this

study is considerably simple, which solves

the nonlinear shallow water equation by the

leap-frog finite difference method using 5 cm grid and 0. 02 sec time intervals. The

fundamental equations are as follows:

!?

+

ax

=0 to

where A. : still water depth

t} : displacement above the S. W. L. D : total water depth (= A -+• 17 )

* Head, Coastal Engineering Division, Public

Works Research Institute, Ministry of

Construction

** Research Engineer, Coastal Engineering

Division, Public Works Research Institute,

Ministry of Construction

217
Tl : Manning's roughness coefficient

Gi : rate of discharge per unit width

Here A. takes a negative value on the dry

bed, that is the negative vertical distance

between the still water level and the ground.

For the initial condition wave profile

measured just in front of the generator was

used. The value of 71 is assumed to be 0.012,

and it was determined empirically in such a way that the difference between measured and

calculated run-up height in the case of 1/20

uniform slope is minimal. For the boundary

condition of the wave front the method by

Iwasaki et al. 1

’ was used. In their model the wave front is judged by whether value of D

is positive or negative. The rate of

discharge through the permeable detached

breakwater is calculated by

Ci)

where X : porosity ( = 0.5)

Jl : width of detached breakwater

f : drag coefficient (= 8. 8)

4. COMPARISON WITH THE EXPERIMENTAL RESULTS AND

THE NUMERICAL SIMULATIONS

The experimental results are compared with the

simulations in Fig. 3. The simultaneous

measurements of the water level and the

velocity was difficult because of the large

number of the measuring points, so that these

measurements were conducted in the repeated

experiments with the same physical condition.

In the simulation the mean wave profile

measured in front of the generator is used as the initial condition, and therefore this

condition is not always the same as the one when the run-up height was measured. This has

a considerable large effect on the results

because the initial condition is important in

the tsunami simulation. In spite of these

errors of the initial condition the

differences of the run-up height between the

experiments and the simulations are less than

i 09-6 and they agree considerably well.

The measured and calculated results of water

level and velocity are compared in Figs. 4

through 8 with regard to 1/20 uniform slope,

coastal dune, coastal revetment and detached

breakwater, respectively. The spatial

profiles of the water level are shown in

Fig. 9. Time in these figures is measured

right after the water level at x =»5m starts

rising. The reference level of water surface

is selected at the still water level in the

offshore region and at the dry bed on the

beach, respectively. In Figs. 4 through 8 the

velocity is expressed in the absolute value,

because uni -directional current meters were

used.

4. 1 1/20 uniform slope

The reflected wave in the experiment as shown

in Fig. 4 has many short-period waves. This

simulation can not predict these waves in

principle, because the fundamental equation

does not include the dispersion term. The

phase of the reflected waves in the simulation

is faster than that in the experiment,

although the cause of this difference is not

clearly understood. Both water level and

velocity profiles are generally in good

agreement, even if wave breaking and mixing at

the wave front are neglected. Furthermore the

spatial profiles in Fig. 9a) also show good

agreement each other.

4. 2 Coastal dune

The reflected wave in the offshore region as shown in Fig. 5 expresses the same result as the one in the uniform slope, while run-up

profiles of both water level and velocity

agree well. The calculated velocity at Pll on

top of the dune is higher than that observed,

so that it is considered that the reflected wave at P16 is generated earlier in the

simulation. This also causes the phase

difference of spatial profile shown in

Fig. 9b), that is, the profile at t =29 sec in

the simulation agrees with the one at t =30 sec in the experiment. These differences are considered to be due to the effects of drag

caused by the separation on top of the dune,

judging that the velocities at points

offshore from Pll located on top of the dune

are in good agreement as shown in Fig. 5 during

the run-up period. Furthermore in Fig. 9b) the

measured profile of the water surface around

on top of the dune is raised higher than that

in the simulation. There exist some small

differences. However, it is concluded that

the run-up motion around the coastal dune can

be simulated properly, even if present model

is fairly simple.

4. 3 Coastal revetment

In this case both water level and velocity are slightly higher on land as shown in Fig. 7. Except this smail differences the results

including spatial profile in Fig. 9c) show good

agreement between the experiment and the

simulation. In the simulation the revetment

is treated only as the bed level change in one

grid instead of considering the effect of the

revetment. Nevertheless it is considered that

water motion can be predicted by the model

properly, because the separation zone is

formed in front of the revetment

4. 4 Detached breakwater

The width of the detached breakwater in the

model experiment is narrow enough as 5 cm. Therefore the existence of the detached

breakwater can be expressed in terms of the

bed level change in one grid. The results of

impermeable and permeable detached breakwater

are discussed, dividedly.

The change of the water level and velocity in

218
the impermeable detached breakwater is shown

in Fig. 7. At the crown of detached breakwater

(PS) water level is higher and velocity is

smaller in the experiment compared with those

in simulation during run-up period. In

addition,, rapid current movement, keeping

shallow water depth, exists continuously at P6

and P7. The initial reflected wave from the

slope is generated at around P8 in the

simulation, while it is generated at about P15

in the experiment. It is thought that these

differences between the experiments and the

simulations are caused by errors in estimating

water level and velocity at the crown of the

detached breakwater during the run-up period.

The spatial profile of the water level is

shown in Fig. 9d). The difference of the water

level between the crown and just shoreward

point of the breakwater is large, and thin

curret develops on the slope at the initial

stage of the wave run-up. In calculating the

current over the detached breakwater it is

important to take into account the separation

of the current at the breakwater.

In case of the permeable detached breakwater

the type of the current over the breakwater is

different from the impermeable one. The

separation of the current is not remarkable at

3000 <0 500 "!“l

>000 2000 I

AIR

ROOU

i r

500 -1 r O^-GIASS WALL

X

'$I0E WALL

Njbsebvation slot

a) plane section

b) cross section (A-A’

)

Fig. 1 Wave basin

c«

the crown of the breakwater. As a result, it

is found that both water level and velocity

obtained in the experiment and the simulation

(see Fig. 8) have better agreement than that of

the impermeable case.

5. CONCLUSION

The numerical simulation was compared with the

experimental results on tsunami run-up in the case that some coastal structures and sand

dune existed on a coast. It is found that the

simple model used in this study has an ability

to simulate tsunami movement with enough

accuracy in practical use. In the case of the

detached breakwater it is important to taken

into consideration of the current over the

breakwater at the crown of the breakwater.

However, in the case of the permeable

breakwater, which is usually constructed in

Japan, detached breakwater can be modelled by

the bed level change in one grid.

REFERENCE

1) Iwasaki, T. and A. Mano : Numerical

simulation on 2 -dimensional tsunami run-up

using Eularian coordinates, Proc. 28th

Japanese Conf . on Coastal Eng. , pp. 250-254,

1979 (in Japanese).

a) 1/20 uniform slope

b) coastal dune

Fig. 2 Models and locations of measuring

Fig. 3 Comparison of run-up height

219
p I P2

P4P5

P6

P10

PI I

P12

P14

PI6

PI9

P2I

MEASURED

CALCULATED

o

a) water level

MEASURED

CALCULATED

P2

•Or

\ X-500 ca r — '

P3

»rJ X-lOOca

P5

•tol»C«

P6 /

^

1

P7 _

GH-4.0 ce — P8

M>r

P9 GH-3.7 ca ^

PI2 GH-9.3 ca ^

PI3

i°r

. f X—145 CM GH-10.7 cc

PI6 GH-M.5 Cfi —

PI8 ir

GH-17.0 ca

-15 -» -5 0 5 K) 13 20 ate a) water level

MEASURED

pi

tOOjC *XS,C calculated

P2

M) Or

\ X-500 ca

_

P4 WOr r—-

P5 ©Or

P6 (OOr

Q

\ X-2S ca PIO ^ X —85 ca, rv,—

Pit X—103 ca GH-5.2 ca

PI2 ^ X— 133 ca GH-6.7 ca

P 1

7 X- «c. GH-9.2 ca „

P16

•OOr

n\ X—235 ca GH-fl.7 ca /"- — , P19 '°5 X—3*0 ca GH-l5.3eo

P2I

«0,

n \ X—410 ea GH-20.S c«

b) velocity

uEASUREO

CALCULATED

P

_ / P2 •oor

r x-500 ca r— ——

P3

OOr

0 r x-ioo ca

r P5 •OO r ? Fx-50ce . r-C... - P6 •oo r 0

l . rs^

P7

•OOr0 LlZQ {9 _

P8 •OOr

Q

r X— 20 ca

( GH-4 Sea , P9

WOr

Q

\ X—49 ca GH-3.7 ca PI2

WOr

Q

h X—120 ca. GH-9.3 ca

, r~>-.

PI3

•oorJ

X—143 ea GH-10.7 ca PI6

wor q L 121.2.20 ca

PI8

«3r

0

\ »-;70c. SH-I(,gt« —

b) veloci ty

Fig. 4 Time series of water level and velocity

(1/20 uniform solpe)

Fig. 6 Time series of water level and velocity

(coastal revetment)

Fig. 5

MEASURED

P I P2

P 3 P5

P7

PIO

PI I

PI2

P14

P16

PIS

P2I

a) water level

p i P2

P3

P5

P7

<-100 I«

.A

P'0

”°

1» —Win . GH-8 3l».

PI. ~ot-- 120 ca

| GM-n.S ca, P12 »°-l45 cu GH“B3.0 cm

PI4 "Jj;-200 co GW—4.9 ce P'6 t

PI9 »--};><« CM-i 1 t» . P2I iTtt« gh° 9.7 c«

MEASURED

CALCULATED

c—

MEASURED

p i T -

calculated

P2 f

P3 x-ioo ca —

P4 ?»•» « —

P5 Si---—

P6

P7 T" —-

P8 "f «—50 cm CH-I.3c« —i_

PIO X—03 ca GH-4. 2 ca - ^

PI2 [ x—135 ce GH-8.7 ca

P15 ”rJ X—210 ca GH-10.3 ca

PI9 J X--320 ca OH- 16.0 cc

-13 -n -3 0 5 10 15 20 ate a) water level

PI

-.ca/ite ”5 r-~

MEASURED

calculated

P2 WOr^

X-300 ca r —

P 3

100,

nf x°iQ0 ca r\-— T— P4 •OOrpi nil, tm

. rv- v.

1 P5 WOrJ

x-30 ca

, GH-4. 3 ca rt ^

P6 °!!f »*J3 t« P7 "3- «. P8 J x— 30 ca GH-LSca

PIO

'“J »— 35 c« GH-4.2te —

PI2

WOr^ X— 135 ca GH-6.7 ce

PI5 *5 X—2IOca GH-10.9 ca

PI9

*00,J X— 320 ce GH-16 0 ca

-•9 10 -5 0 3 O 13 20 itc b) veloci ty

Time series of water level and velocity Fig. 7 Time

(coastal dune)

b) veloci ty

series of water level and velocity

(Detached breakwater : i mpermeabl e)

220
PIP2

P3

PAP5

P6P7

P 8 P9

PIO

PI I

PI2

PI3

MEASURED

P2 x-500<o

1

P3 »•«! < - rt-_— ^— r n PA X-55 <• tr —-- —

P5 n .. P6 "k=«c_., *

P7 »-» .

P8 • ^

P9 I'-1 '1 . PIO X~ 75 { G*-J.5c« _

Pll

•OOrJ

X—150 « GH-7.2c«

P12 1—225 ea Ot-n.O cm r..

b) veloci ty

Fig. 8 Time series of water level and velocity

(Detached breakwa ter : permeab le)

MEASURED

— MEASURED

MEASURED

MEASURED

Fig. 9 Spatial profiles of water level

221
'

_ .

.
Theme IV

U.S.-Japan Cooperative

Research Program

Effect of Number of Loading Cycles and Loading Velocity of

Reinforced Concrete Bridge Piers

Toshio

Kin j i

BY

1

2 Iwasaki , Kazuhiko Kawashima

3 4 Hasegawa , Tatsuhiko Koyaraa , 4 and Takeshi Yoshida

9

ABSTRACT

A series of dynamic loading tests with use of

large-scale reinforced concrete bridge piers

was conducted to investigate an effect of

number of load reversals with the same displacement amplitude as well as an effect

of loading velocity. For studying the number

of load reversals, seven specimens were subjected to a series of step-wise increasing

symmetric displacement with n cycles, in

which number of cycles n was varied from 3 to

10 as a parameter to be investigated. For studying the loading velocity, three

specimens were subjected to a series of

dynamic loading with velocity of 10 cm/sec

and 100 cm/ sec

It was understood from the study that

although deterioration of the specimen

progresses, as a general trand, in accordance

with increase of the number of inelastic

loading cycles, for specimens which failed in flexure, effect of the number of loading

cycles between 3 and 10 is less significant

at the loading desplacement smaller than or equal to the displacement at which the

maximum load was developed. It was also

found that deterioration of the specimen

subjected to dynamic loading of 10 cm/sec

takes place earlier by an amount of

approximately 1 8 than that subjected to dynamic loading of 100 cm/sec.

KEY W0RDS :Bridge Pier, Dynamic Loading Test,

Reinforced Concrete Structure, Seismic Design

1 . INTRODUCTION

In the past, number of highway bridge have

suffered damages due to extreme earthquakes.

Reinforced concrete bridge piers supporting

superstructures are quite susceptible to damage from strong ground motions. This fact

became evident during the Miyagi-ken-oki

Earthquake of 1978, when numerous reinforced

concrete bridge piers designed by the

traditional elastic approach suffered severe damages „ Such evidence clearly showed the

importance to study inelastic behavior with

energy dissipation of reinforced concrete

bridge piers under severe ground motions.

Various investigations have been made to study inelastic behavior of reinforced

concrete bridge piers subjected to strong

earthquakes, and specimens simulating

reinforced concrete bridge piers have been

tested under various loading conditions such

as static and dynamic cyclic loading,

2

)—

) earthquake-like loading . In almost all

the test, however, loads were step-wise

increased monotonically up to failure, in

which at each step several cycles of

symmetrical loading reversals with the same displacement amplitude were carried out. Although number of loading cycles at each

step may have a significant effect on the

test results, little is known about the

effect of load reversals on the mechanism of

inelastic behavior of reinforced concrete

bridge piers. Because structural response

amplitude also highly depends on the number

of response cycles , it is inevitable to study such effects in order to apply the

dynamic loading test data for practical

seismic design of reinforced bridge piers

which are to be subjected to earthquake

ground motion.

1 Dr. of Engr. .Director, Earthquake Disaster

Prevention Department, Public Works

Research Institute, Ministry of

Construction, Japan

2 Dr. of Engr. , Head, Earthquake Engineering

Division, Earthquake Disaster Prevention

Department, Public Works Research

Institute, Ministry of Construction, Japan

3 Research Engineer, ditto

4 Assistant Research Engineer, ditto

225
Furthermore, althougl the seismic effects to bridge piers are dynamic, in almost all the

past tests the specimens were subjected to static loading from a limitation of loading

facilities

This study presents a series of dynamic

loading test results of large-size reinforced

concrete bridge pier models subjected to different number of load reversals with the same displacement amplitude and different

loading velocities. Effect of number of

loading cycles and loading velocities on strength and energy dissipating capacity is

presented in terms of comparison of hysteresis loop, accumulated energy dissipation and

equivalent damping ratio.

2. TEST SPECIMENS AND EXPERIMENTAL SET-UP

2.1 Effect of Number of Loading Cycles

2.2 Effect of Loading Velocity

For studying the effect of loading velocity,

three reinforced concrete cantilever

specimens as shown in Fig. 2 were used. They

have the same characteristies on the cross section of 80 cm x 40 cm, effective pier

hight h of 2.4m, main reinforcement ratio of

0.87 % and tie reinforcement ratio of 0.08 %. Main reinforcements of 22 SD30 bars (deformed

bars) with diameter of 13 mm were placed on only two sides which sustained tension and

compression during loading, and no reinforcements were placed on the other two sides.

Tie reinforcements with use of 9 mm diameter

SR24 bars (plain bars) were placed at 20 cm intervals. They used portland cements with

uni-axial compression strength ranging from

2 311 and 385 kg/cm , and agregates with the

maximum grain size of 10 mm.

Seven large-size reinforced concrete specimens simulating cantilever piers of bridges

as shown in Fig. 1 and Table 1 were used for

studying zhe - number of load reversals.

Specimen number presented show those used

to identify specimens in a series of dynamic

loading tests conducted at the Public Works

1 1 ) Research Institute . The cantilever pier

was framed into a massive reinforced concrete

footing which was anchored to a test floor by means of post tensioned rods. The common characteristics over the seven specimens are

the cross section of 50 cm. x 50 cm, main

reinforcement ratio of 2.03 Z with use of 40

SD30 bars(deformed bar) with diameter of 13 mm, and tie reinforcement ratio of 0.1 S with

use of 9 mm diameter SR24 bars (plain bar)

placed at 25 cm intervals. The layout of tie

bars depends on the Japanese practice in

accordance with the Specifications for Design

of Highway Bridges. The tie bars were lapped

at least 60 mm. They used portland cement

with uni-axial compression strength ranging

2 from 319 to 340 Kg/cm , and agregates with

the maximum grain size of 10 mm. Effective

pier heights h from bottom to the center of

loading point were taken as parameter to be

investigated as 2.5 a (P-10, P19 and P-20),

1.752 m (P-17 and P-20) and 1 a (P-18 and P- 20) so that the shear-span ratio, which is

defined as a ratio between effective height h and effective width d (width- covering depth),

becomes 5-4, 3*8 and 2.2, respectively.

3. LOADING CONDITIONS

The specimens were loaded at the cantilever

tip as shown in Fig. 3 by means cf an electro-hydraulic dynamic actuator with

loading capacity of ±125 tf, max. stroke of

±125 mm and max. loading velocity of 1 m/sec.

Extensive electric instrumentation was used

to measure and record the basic deformation

parameters such as strains of main and tie

reinforcements as well as the displacement

and acceleration at the loading point.

Although a compression stress of about 5 to

20 Kg/cm is generated in the piers with

ordinary shape and height due to the dead

weight of superstructure, it was disregarded

in the test.

For studying the number of load reversals

specimens were subjected to a series of stepwise increasing symmetric displacement cycles

as shown in Fig. 4. At each step, n cycles of

loading with the same displacement amplitude

was carried out, in which number of cycles n was varied from 3 to 10 as a parameter to be investigated. The loading velocity was taken

as 25 cm/sec over each step. The step-wize

was determined by increasing the displacement

in each step by a displacement ductility

factor of one. The yield displacement

corresponding to one ductility factor was defined as the displacement at loading point

at which reinforcing bars at the extreme

226
tension fibre firstly reached yield strain.

A number of strain guages were placed on the two longitudinal reinforcing bars on each

side to the loading direction, and the

averaged strain of the two reinforcing bars , which were placed at the foot of pier, was used to determine the yield displacement. The

yield strain of reinforcement was assumed as 1800.U oased on a number of tensile test of

the specimens

It should be noted in the study of number of

load reversals that in specimen P-18, slips

between footing and test floor were developed

duing the 2 Sy loading due to inadequate

anchoring of the specimen to the floor.

Therefore, the 2 Sy loading was subsequently

carried out again, and these data were treated as those for 2 Sy loading.

and 3.8 failed clealy in flexure. By

increasing loading displacement, flexural

cracks were firstly formed on the side near

foot, and spalling-off of cover concrete at the critical flexural region followed.

Progress of significant flexural cracks

finally penetrated the critical section

causing break up of the concrete core, and

inelastic buckling of main reinforcement

either outwards between two adjacent tie bars

or sideways were progressively developed. It should be noted here that pull-out of main

reinforcement from footing was developed due

to loss of bond caused by load reversals in

the anchorage, which, in turn, caused

rotation at the foot and contributed to the

pier displacement significantly. Eventually

cut of main reinforcement was developed which

critically deteriorated the load and

stiffness

For studying the loading velocity, the same loading procedure as above was used, in which

number of cycles n was assumed as 10, and

loading velocity V was varied as 10 cm/sec

and 100 cm/sec.

Because the force detected by means of a load-cell equiped in the dynamic actuator

represents total force c applied to the

specimen including an inertia force

associated with masses of the specimen and

the loading jig, the lateral force P which

actually loaded to the specimen was calculated assuming that the inertia force

along the cantilever pier distributes in an inverted-triangle shape with the inertia

force at the foot being zero.

4. EFFECT OF NUMBER OF LOADING CYCLES

4 . 1 Failure Mode

Figs. 5,6 and 7 show failure modes of the

model piers with emphasis on cracks and

spalling-off of cover concrete as well as elongation, buckling and cut of reinforcements. Failure modes on the side

perpendicular and pararell to the loading

direction are presented here for the

specimens with shear-span ratio of 5*4 and

3.8, and those with shear-span ratio of 2.2,

respectively, so that progress of failure

mode be clearly observed.

The specimens with shear-span ratio of 5.4

In comparison, the specimens with shear-span

ratio of 2.2 failed in shear. By increasing

loading displacement, diagonal shear cracks

were firstly developed on the side, and break

up of the concrete core accompanying complete

spalling-off of the cover concrete brings to final degradation of the stiffness.

In all the abo ve -de s c i bed failure modes,

increase of the number of loading cycles n clearly accelerates failure in accordance

with increase of loading displacement. For

example, the specimen with shear-span raio of

5.4, which were subjected to three reversals

at each loading step, suffered spalling-off

of cover concrete only at a region 30 cm from

the foot even after the loading with

displacement amplitude of 8 Sy < while the same specimen subjected to ten reversals at each loading suffered much severe spallingoff of cover concrete at a region 40 cm from

the foot after the 5 <5 V loading.

Table 2 showed an effect of number of loading

cycles for flexural failure in terms of

number of longitudinal reinforcement which

were cut due to cyclic inelastic buckling. No longitudinal reinforcements were cut in the

specimens which failed in shear. It is

apparent that the number of loading cycles is of significant factor affecting the ultimate

failure mode of flexural cantilever piers

associated with cut of main reinforcement.

227
Fig. 8 shows strains induced in the tie bars

of the specimens with shear-span ratio of

2.2. Although failure of concrete made unable

to measure the strain of tie bars for larger

loading displacements, it is apparent that

increase of number of loading cycles brings

increase of strains developed in tie bars. It should be noted that tie bar at 50 era frora foot (mid-height of the pier), which were subjected to ten reversals at each step,

already yielded even at 2 loading

displacement.

4.2 Strength and Deformation Capacity

The behavior of the test speemens was firstly

studied from continuous plots of hysteresis

loop of the load vs. displacement at the

loading point. Fig. 9 shows comparisons of

the envelopes of the hysteresis loops.

some doubts as to whether the difference

observed in the comparison of envelops can be entirely attributed to the effect of number

of loading cycle, because of the slips

developed in specimen P-18 between footing

and test floor during 2 loading. As was described in the preceding chapter, as a consequence the P-18 specimen was totally

subjected to 20 load reversals with 2 5^

displacement, and this might cause appreciable deterioration of the specimen.

However, since progress of break up of the

concrete core accompanying yielding of tie

bars is pronounced in shear specimens, it is considered reasonable that effect of number

of loading cycle is more pronounced in shear

specimens than in flexural specimens although

the difference observed in Fig. 9 (c) is not

necessarily entirely caused by the effect of

number of loading cycles.

For the flexural specimens with shear-span

ratio of 5.4 and 3.8, effect of the number of

loading cycles varies in accordance with

loading displacement. For loading dispacement

smaller than or equal to 4 5^ for specimens

with h/d = 5=4 and 3 5 ^ for specimens with

h/d = 3.8, the envelopes of hysteresis loop

seem almost independent of the number of

loading cycles. It should be noted here that

the critical displacments 4 4 y (h/d = 5.4) and

3 <5 y (h/d =3-8) correspond to those at which

the maximum loads were achieved. This implies

that for flexural specimens the number of

loading cycles has less significant effect on the envelopes of hysteresis loop until the

maximum load is developed. However after

these displacements, a significant drop in

load can be observed in the envelopes in

accordance with increase of the number of

loading cycles. Especially, as was previously

discussed, drops in load associated with cut

of longitudinal reinforcements critically

depend on the number of loading cycles.

On the other hand, for the specimens with

shear-span ratio of 2.2, which failed in

shear, more significant changes of envelope

of hysteresis loop in accrdance with the

number of loading cycles are observed ,i.e.,

for loading displacement larger than

the envelope of the specimen subjected to ten

reversals is much smaller in load than that

of the specimen subjected to three reversals.

It should be, however, noted that there is

Table 3 summalizes the yield load P , max. load , yield displacement and ultimate

displacement j ^ for seven specimens. Only

averaged values developed between in positive

and in 'negative loading directions are presented in Table 3« For specimens with

shear-span ratio of 5-4 and 3*8, the max load

Pmax is almost free from the number of

loading cycles , which, in turn, shows that a ratio of the max. load to the yield load , P /P , is also almost independent of the

numben of loading cycles. For specimens with

shear-span ratio of 2.2, the max. load P^

when subjected to three inelastic revesals is

only 8% larger than that subjected to ten inelastic reversals, i.e., effect of number

of loading cycles' in terms of the max. load

P^ is not necessarily significant as anticipated from clear difference which was observed in the envelope of hysteresis loop

presented in Fig. 9 (c).

In comparison, the ultimate displacement {

of the specimens which failed in flexure

significantly varies in accordance with the

number of loading cycles, which, in turn,

affects the ductility factors, i.e., a ductility factor obtained from the tests

using three inelastic revesals for the same displacement amplitude is 36Z and 1 9 Z larger

than those from ten inelastic reversals tests

for specimens with shear-span ratio of 5-4

and 3-8, respectively. Therefore the number

of loading cycles have to be well examined

228
when the ductility factors obtained with

different test procedures are to be compared.

The ductility factor for specimens with

shear-span ratio of 2.2 does not, however,

show significant changes with respect to the

number of loading cycles.

4.3 Energy Dissipating Capability

An area surrounded by a hystersis loop as shown in Fig. 10 represents an amount of

energy dissipated during that loading

reversal, and hence, accumulating the energy

dissipation during each loading reversal,

total amount of energy dissipated in the

specimen during the loading test may be

calculated. Since the accumulated energy

dissipation so obtained is lager in those

specimens which experienced larger number of

loading cycles, direct comparisons of the

accumulated energy dissipation do not

represent the energy dissipation capability

of the specimens. Therefore, the accumulate

energy dissipations calculated for the

specimens which were subjected to three and

five cycles of inelastic reversals for each

loading step were multiplied by a factor of

10/3 and 10/5, respectively, so that the

accumulated energy dissipation be compared on the basis of the same number of loading

cycles of ten. Fig. 11 represents the accumulated energy dissipation thus obtained.

during the test for the specimens which were subjected to three load reversals is approximately 50Z larger than that for the specimens

which were subjected to ten load reversals.

Energy dissipation capability was also

studied in terms of an equivalent hysteretic

damping ratio h

Q which is defned as

in which W is an energy dissipated during

one load reversal and W is an elastic energy

as defined in Fig 10. Fig. 12 shows

variations of the equivalent hysteretic

damping ratio h

g with respect to the

displacement amplitude of cyclic loading. As

a general trend, the equivalent damping

ratios increase with increasing the loading

displacement, take a peak value of

approximately 0.2 at 4 <5 ^ (h/d = 5-4 and 3-8)

and 0.12 — 0.16 at 2~3 3 (h/d = 2.2), and

progressively decrease to a value of about

0.15 ( h/d = 5 • 4 and 3.8) and 0.1 (h/d = 2.2) . The

effect of number of loading cycles can be

observed in the decreasing path after h

g takes the peak value, i.e., h

Q decreases

promptly in the specimens which were subjected to larger number of loading cycles.

5 . EFFECT OF LOADING VELOCITY

5.1 Failure Mode

As anticipated from the previous discussions,

the accumulated energy dissipation significantly decreases with increasing the number of

loading cycles not only in the specimens

which failed in flexure but those in shear.

It should be noted, however, that such

decreases of the accumulated energy dissipation with respect to increase of the number

of loading cycles become apparent at the

loading displacement larger than about 4 5y>

3 5 , and 2 5 for the specimens with sheary y span ratio of 5-4, 3-8 and 2.2, respectively.

It should be reminded from Fig. 9 that those

loading displacements almost correspond to the loading displacement at which the maximum

load was developed. Therefore it may be

considered that effects of the number of

loading cycles on capability of energy dissipation are less significant until the maximum

load was developed, and that it progressively

becomes significant after the maximum load.

Total amount of accumulated energy dissipated

Fig. 13, 14 and 15 compare failure modes

observed on the side surface perpendicular to the loading direction. Progresses of the

flexural failure in the three specimens are essentially the same with those observed in

Fig. 5 and 6 , i.e, by increasing loading

displacement, flexural cracks were firsty

formed, and spalling-off of cover concrete at the critical flexural region followed.

Subsequently, significant flexural cracks

penetrated into concrete core, and even cuts

of main reinforcement were developed. It should be noted, however, that the spallingoff of cover concrete was initiated at 4 5 y for the specimen with loading velocity of 10 cm/sec, while it was initiated at 5 3 for

two specimens with loading velocity of 100

cm/ sec .

It was also observed that cuts of main

reinforcements were developed during 6 5 loading displacement for the specimen with

229
I

loading velocity of 10 cm/sec, while they

were developed during the 7 loading

displacement for the two specimens with

loading velocity of 100 cm/sec. Table 4 shows the effect of loading velocity in terms

of loading displacements where spalling-off

of cover concrete and cut of main

reinforcement were initiated. It is clear

from these evidence that failure is likely to develop promptly in specimen subjected to loading velocity of 10 cm/sec with an amount

of approximately 1 Sy as compared with the

specimens subjected to loading velocity of

100 cm/ sec

5.2 Strength and Deformation Capacity

Fig. 16 shows the hysteresis loops of the load

vs. displacement at the loading point, and

Fig. 17 shows the comparison of first excursion of hysteresis loops. The hysteresis

loops between three specimens show the same behavior for displacement smaller or equal to

4 ,5 , which implies that the effect of

loading velocity is less significant for that

displacement. However, beyond loading displacement of 5 8^ > specimen subjected to loading

velocity of 10 cm/sec shows appreciable drop

in load as compared with the specimens subjected to the loading velocity of 100 cm/sec.

The shape of hysteresis loop is also lean in

the specimen subjected to loading velocity of

10 cm/sec. Such a prompt deterioration of

the specimen with loading velocity of 10

cb/ sec is clearly brought by the prompt progress of flexural failure as described in the

preceding section.

Fig. 18 shows comparison of load vs. displacement relation of the specimen in terms of

envelopes of the hysteresis loop. As was auticjpated from the above discussion, the

maximum load was developed at 2 5 ^ on + side

and 4 5 ^ on - side for loading velocity of 10 cm/sec, while it was developed at 4 on + side and 4 — 5 Sy on - side for loading

velocity of 100 cm/sec. This shows that the

maximum loads are developed at larger loading

displacement in accordance with increase of

loading velocity, which implies that stable

load vs. displacement hysteresis can be

credited in the loading with higher loading

velocity

Table 5 summalizes the load and displacement

performance of the three specimens.

5 . 3 Energy Dissipating Capability

Figs. 19 and 20 show comparison of energy

dissipation and equivalent hysteretic damping

ratio h

e for the three specimens. It is

clear in these comparison that difference of

the energy dissipation and equivalent

hysteretic damping ratio in accordance with

loading velocity is less significant for

loading displacement smaller than or equal to

4 Sy- However beyond this displacement, the

energy dissipation and equivalent damping

ratio take larger value in loading velocity

of 100 cm/sec than in loading velocity of 10

cm/ sec

It should be noted here that the loading

displacement of 4 5y almost corresponds to the displacement where the load begins to decrease after taking the maximum load P .

6 . CONCLUDING REMARKS

In an attempt to study effects of number of

inelastic loading cycles and loading velocity

on strength and energy dissipating

characteristics of reinforced concrete bridge

piers, a series of dynamic loading tests with

use of ten large-size specimens was conducted

by varying the number of inelastic loading

cycles from three to ten and by varying the

loading velocity as 10 cm/sec and 100 cm/sec*

Shear-Span ratio of the specimens was 5.4,

3.8 and 2.2 for the study of number of

loading cycles, and 6.8 for the study of

loading velocity. Based on the test results

presented, the following conclusions may be

deduced :

1 ) Effect of number of loading cycles

i) As a general trend, deterioration of the

specimen progresses in accordance with

increase of the number of inelastic loading

cycles with regardless of flexural failure or shear failure

ii) However, for specimens with shear-span

ratio of 5-4 and 3-4 which failed in flexure,

effect of the number of loading cycles

between 3 and 10 is less significant for the

loading displacement smaller than or equal to the dispacement at which the maximum load was

230
developed. Within this loading displacement,

the envelope of hysteresis loop, acumulated

energy dissipation, equvalent hysteretic

damping ratio and the maximum load are less

significantly affected by the number of

inelastic loading cycles. However, over this

displacement, deteriorations of the specimen

becomes significant in accordance with number

of inelastic loading cycles. Especially,

deterioration of the specimen associated with

cut of main reinforcement due to cyclic

inelastic buckling critically depends on the

number of loading cycles.

iii) For the specimens with shear-span

ratio of 2.2, which failed in shear, the

effect of number of inelastic loading is much

significant as compared with the flexural

specimens. The envelope of hysteresis loop

shows significant drop in load at displacements larger than — 2 5^. •

2)

Effect of loading velocity

i) Deterioration of the specimens in terms

of loading displacements at which spallingoff of cover concrete and the maximum load P

u were developed takes place earlier in

specimen subjected to loading velocity of 10 cra/sec than those subjected to loading

velocity of 100 cm/sec with an amount of

approximately 1 5^..

ii) Effect of loading velocity on energy

dissipating capability in terms of

accumulated energy dissipation and equivalent

damping ratio are less significant for

loading displacement smaller than or equal to

4 Sy

However beyond this displacement,

energy dissipation capability is appreciably

larger in the specimens subjected to loading velocity of 100 cm/sec than in the

specimen subjected to loading velocity of 10

cm / s e c .

REFERENCES

1) Okubo, T. , Ohashi, M. , Iwasaki.T. , Kawashima, K. and Tokida, K. : Damage

Features of Civil Engineering Structures due

to The Miyagi-ken-oki Earthquake of 1978,

11th Joint Meeting, Wind and Seismic Effect,

UJNR, Tsukuba, Japan, 1979.

2) Park, R. and Blakeley, R.W.G. : Seismic

Design of Bridges, RRU Bulletin 43. National

Roads Board, 1979

3) Yamamoto, T., Ishibashi, T., Otsubo, M.,

and Kobayashi, S. : Experimental studies on Seismic Resistance of A Pier with

Reinforcement Terminated in A Tension Zone,

Proc. of JSCE , Vol.348, V-1, 1984.

4) Ohta, M. : A Study on Earthquake Resistant

Design for Reinforced Concrete Bridge Piers

of Single- Column Type, Report of the Public

Works Research Institute, Vol. 153, 1980 (in

J apane s e )

.

5) Ikeda, S., Eya , S. and Yamaguchi, T. : Behavior of Reinforced Concrete Cylindrical

Columns under Reversed Cyclic Loadings,

Trans, of the Japan Concrete Institute, 1986.

6) Mutsuyoshi, H., Machida, A. and Tsuruta,

K. : Predicting the Nonlinear Earthquake

Response of Reinforced Concrete Structures in

Consideration of Strain Rate Effect, Proc. of

Japan Society of Civil Engineers, vol. 366,

V-4 , 1986.

7) Park, R., Kent, D.C. and Sompson, R.A. : Reinforced Concrete Members with Cyclic

Loading, Proc. of ASCE, vol. 98, N o . 9 S T 7

,

1972 . 8) Krawinkler, H. and Popov, E. : Hysteresis

Behavior of Reinforced Concrete Rectangular

and T-Beam, 6 WCEE

9) Gulkam, P. and Sozen, M. : Inelastic

Response of Reinforced Concrete Structures to Earthquake Motions, ACI Journal, 1974.

10) Kawashima, K. and Aizawa, K. : Earthquake

Response Spectra Taking Account of Number of

Response Cycles, Earthquake Engineering

and Structural Dynamics, Vol. 14, 1986.

11) Kawashima, K., Hasegawa, K. Koyaraa, T. and Yoshida, T. : Experimental Investigation

on Dynamic Strength and Ductility of

Reinforced Concrete Bridge Piers, Part I, No. 2232 of the Technical Memorandum, Public

Works Research Institute, Ministry of

Construction, 1985.

231
Table-1

Specifications

of

Test

Specimens

for

Studying

Table-2

Number

of

Longitudinal

Reinforcement

Cut

Number

of

Load

Reversals

During

Cyclic

Loading

232
0=500, 0=

500 5 2 S 9 £ MC

°H

T33

-PCO3O P

CQ *H 3 o

CDO E rH

•H (0 O >

<ra,

m

co c

•H

-P T3

cq a)

CD o

E-<

J

cv

ti

"H

S9f = 5E-a = 0 ”

"OOSjC"

a:<

as z o uj z

CC o O u_ 5 2

cc

COtT'’ XS2=a S ZSZ ‘ ? ZS9 ?

CNJ

(Nl Cl. a o a ^ 03

<T03

CNJ

r\j

a.o'

a.

CQ rH3

CQ30) >

<D as3oJ

<Po30) -Oe35JJD 3

°H

>s T33

-Pco3o

Cm

CQ33EHO3

co- co

-p

CQ3

Esi •H

(*?*) 1N31M3D VldSlQ

tu CO3

-P33E

•H33X

to

si

•H

tc-i

233

!•

Reaction

Floor

Fig. 4

Step-wise

Increasing

Symmetric

Loading

Displacement

with n

Reversals
2Sy 3Sy 65y (a) Number of Loading cycle n=10

mirrtnt u

mm

75 y 85y

-U

h—

hn ;—t

11,1

K~ T:

r— i ! r 1 -i—rn

—

! q

=4-

i i i

25 y

TT ++

1 Tx

TT

1 44

35y 4Sy S5y 6a r (b) Number of Loading cycle n=3

Fig. 5 Failure Mode for Specimen with h/d=5-4

i i

25y

1 ml ** all T

j 4 Ui_i—

~t~

^ 1

jijjf nsc

[ tsu

tttt

1M

3Sy 4Sy 5iy 6Sy

(a) Number of Loading cycle n=10

75 y

75 y

mim

65y

-XS5y

T

^-7,-

25 y

-a"

35y 4Sy

! ! t

55y

XI

SI! ms

65y (b) Number of Loading cycle n=3

6 Failure Mode for Specimen with h/d=3.8

75 y 8«Sy

234
LOAD

(T)

LOAD

(T)

75 y Sy 2Sy

25y

3Sy

(a) Number of Loading cycle n=10

5 r

3Sy 4§y 5Sy (b) Number of Loading cycle n=3

Fig. 7 Failure Mode for Specimen with h/d=2.2

J ^jvLLa 5 PIP

5 5

^ i

7/ 5JM1 {fu

45y 55y 6Sy

Strain of Tie Bar Strain of Tie Bar

at S3 and S4UIO j at Si and S2UIO 5

)

—1—1—l

1—1—1—1—l—1

n = 10 n= 5 —

n --3 ‘

1 1 1 1 1 1 1 1 1 1 i 20 i i i 1 1 1 1 1 1 1 1 1 1 1 1 i 1 1 -10 -8 -6 -4 -2 0 2 4 6 8 10 DISPLACEMENT ( y

)

(a ) h / d = 5 4

-10 -8 -6 -4 -2 0 2 4 6 8 10 DISPLACEMEN ( xSy )

( b ) h / d = 3 8

DISPLACEMENT ( xSy ) (c) h/d = 2. 2

Fig. 9 Comparison of Envelope of Hysteresis Loops

235
FORCE

*-0 <x> «

W J.) A9H3N3 Q3i\nniMnOO\/

CL •£= co . 5 ^

o

CO

— CO

UJS po

LJ ,, CP

<c a

c\j Q.co

ao

‘-'0 X

CO _

H-^ 'T

LlJ 2 >p

Lei CP< -O _J '' CM Q. -C co .

C

o

•H

-P

06CL •H

coCO •HQ

>»

CU3 c.©

CW

T3©P

H) pH3

E

3

O

O

<cCmO

O

CO *H

c6CXE

OO

C©

-H

Ph

IN 1) A9H3N3 Q3lViniNn00\/

-p o

C -H © -P

(—I as 03 OS >

•H CU3 3 C

CP -H w a,E

«m a3 O Q

C O

O -H *H P

P ©

•H Sh C ©

•H P

c_j co © >s Q 3C O

M

&L

ouva ONidiAiva

OliVa ONIdlAIVQ

Oliva ONIdlAIVQ

236

Fig.

12

Comparison

of

Equivalent

Hysteretic

Damping

Ratio
LOAD

(T

ON

)

2Sy 3Sy 4Sy 55y 6Sy Fig. 13 Failure Mode for Specimen with Loading Velocity V=10cm/sec (P — 37)

Fig. 14 Failure Mode for Specimen with Loading Velocity V=100cm/sec (P—38)

•

i

-

* •

. .

, - . L ^ - -1

. .

.

j

i

. . . ...

h

_____

=s y Fig. 15 Failure Mode for Specimen with Loading Velocity V=100cm/sec (P — 39)

(a) V=1 Ocm/sec ( P — 37)

ior

-10

-100 -80 -60 -40 -20 0 20 40 60 80 100 DISPLACEMENT (MM)

(b) V=1 GOcm/sec ( p — 38)

Comparison of Hysteresis

I O' i > -100 -80-60 -40 -20 0 20 4C 60 80 CC DISPLACEMENT (MM)

(c) V=100cm/sec (p — 39)

Loops

237

Fig . 1
C

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LlI

LlI

<J> <

CLcnQ

C i ) avcn

CmO

03a,o

as>c w O .H w co •H

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00

oo

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03CUOo

P3

03 •H

03

a)g

0) -p

03>> X

CmOCo

•r—

03

u,3OXw

-p

03G

•H

P"-i CmOCo

03 “HG

a3aBOo

r--

00

•H 8 m ouvy SNidHva

0II3a3iSIH

-p

o3 Od

ooc

•Hae

cd Q

-P

03G

03 -P

03

-Pc

03 i—

cd>

•H3

CmOGO

03 •Hg

cdaEoo

C\f00

•H

bn

(ui-il A983N3 03880SSV

238

Fig.

1.9

Comparison

of

Energy

Dissipation
I

Study of In Situ Testing for Evaluation of Liquefaction

Resistance and Occurrence

Jeffrey A. Farrarl

ABSTRACT

Analysis of penetration testing for liquefaction

resistance evaluations performed at field liquefaction sites in Japan is described. The study

reviews results of standard penetration, cone . penetration, and vibrocone tests performed in

October and November of 1983 at six field study

sites which exhibited evidence of liquefaction

in Akita and Niigata, Japan. Standard penetration test data with energy measurements and cor- rections to account for sampler dimensions are analyzed using N1 and Japanese simplified

methods, and factors of safety against liquefaction are compared. Mechanical cone penetration

test data are analyzed according to methods proposed by Seed and Olsen. A method to evaluate

liquefaction resistance from cone penetration

testing by using the state parameter is developed and evaluated. Flat plate dilatometer

tests were performed at three liquefaction sites

in the Hachiro-Gata Lagoon area of Japan.

Dilatometer tests were performed using quasi

static and dynamic advancement techniques, and

ability to predict liquefaction occurrence is

evaluated. Vibrocone testing and its ability to

quantify cyclic strength is evaluated. Findings

of the report indicate additional case study

data from silty sand sites and with differing

loading conditions are required to further proof

test relationships.

KEYWORDS: Cone Penetration; In Situ Testing;

Standard Penetration; Vibrocone

1. INTRODUCTION

The purpose of this report is to investigate

various in situ testing techniques for evaluation of liquefaction occurrence in Japan. The

current report is an extension of U.S. -Japan

cooperative studies initiated in 1983 with

fieldwork. This report focuses on the cone penetration (CPT) and flat plate dilatometer

(DMT) tests but also addresses the use of standard penetration (SPT) and vibrocone penetration

(VCPT) testing. Data obtained in a comprehensive field study performed in 1983 at six study

sites are reviewed, and a field study with the

DMT was performed. Results of these studies

have developed valuable case history information

on various methods of geotechnical testing and

will improve our ability to accurately predict occurrence of earthquake liquefaction.

From the results of the 1983 field study,

several important papers have been published to

date. However, there is yet no evaluation of

the CPT test from the data collected. Several

researchers have recently published papers on the use of CPT for liquefaction resistance evaluation, but lack case history validation.

Results of the U.S. -Japan 1983 field data can be

used to check these methods. Another new in

situ test, the DMT, has recently become very

popular among geotechnical groups. Laboratory

chamber testing in sands has indicated that the

instrument may have good potential for eva- luation of liquefaction. To date, only one field liquefaction study site in the U.S. has

been investigated. In this study of the DMT,

three locations in the Hachiro-Gata Lagoon were investigated by quasistatic and dynamic advancement techniques.

1 . 1 Background Information

A comprehensive field study was performed to

investigate in situ testing techniques in

October and November 1983 in Akita and Niigata

prefectures. Three sites from each prefecture

were selected for study, with two sites showing

evidence of liquefaction and one site showing

none. As shown on Fig. 1, the Niigata sites were affected by the 1964 Niigata earthquake of

Richter magnitude 7.5 and Akita sites were affected by the 1983 Ni honkai -Chubu earthquake

of Richter magnitude 7.7. According to strong

motion seismographs in areas nearby study sites,

maximum ground accelerations of 0.17 and 0.16 g were used in analyses of data for Akita and

Niigata sites, respectively.

Study sites in Akita were labeled sites A, B,

and C. The sites are all located in the Hachiro

Gata Lagoon area as shown on Fig. 2. The

Hachiro-Gata area is reclaimed land surrounded

by protective dikes. Site A is located on the

approach to Gomyoko bridge at dike station WC

15+553. This location experienced serious lateral spreading of the approach fill. Site B is

located at dike station FD 6+900 where there was

a slope failure of the dike and numerous sand

boils. Site C is located at dike station WC

0+580 where no damage was experienced.

Study sites in the Niigata area were labeled D,

E, and F (Fig. 3). Site D is the Kawagi shi -Cho

apartment building site which is famous for the

rotational movement of apartment buildings.

Ishihara and Koga (1) have documented soil con- ditions at the site through SPT, CPT, and undisturbed sampling. Site E is on the left bank of

the Showa Ohashi bridge where extreme damage to

the bridge and surrounding ground mass was experienced. Site F is on the right bank of the

Showa-Ohashi bridge site where no damage

occurred. Analysis of the Showa-Ohashi bridge

damage was performed by Iwasaki , et al . (2).

1 Supervisory Civil Engineer, U.S. Department of

the Interior, Bureau of Reclamation, Division

of Research and Laboratory Services,

Geotechnical Branch, Denver, Colorado.

239
The field study at these study sites was probably one of the most intensive of its kind to

date. A listing of the testing performed at

each study site is given in Table 1. Some of

the tests performed included SPT with energy,

measurements, CPT, Swedish dynamic cone, Swedish

weight sounding, piezocone, vibrocone, and other

dynamic cone tests. Borings and soundings were generally spaced at distances of 2 m from one another. Soil types present at the sites are summarized on figure 4. The sites are almost

completely comprised of clean fine to medium

sands

Following the field investigations, 0Y0

Corporation compiled the sounding and laboratory

data obtained from the study in a report dated

March 1984. Presentation of SPT energy measurements and vibrocone results were made by Sasaki,

et al . (3), at the meeting of the U.S. -Japan

Panel on Wind and Seismic Effects (UJNR) in May

1984. Results of SPT measurements were documented by Kovacs, et al . (4), in National Bureau

of Standards report dated August 1984. With

additional data on SPT energy made available,

Seed reevaluated the SPT data base to refine the

N1 method (5). At the UJNR meetings of May of

1985 and 1986, Sasaki, et al . (6, 7), presented

papers refining the vibrocone test. On

August 17-20, 1985, the UJNR group sponsored a workshop on in situ testing with results soon to

be published. Ohya and Iwasaki (8) presented a paper which concentrated on the dynamic penetration testing and its correlation with liquefaction resistance. In this paper, relationships

were proposed for a ratio of static penetration

from the CPT to dynamic penetration from the

Swedish dynamic cone or automatic ram sounding

test to predict liquefaction.

Several approaches to liquefaction resistance

evaluations from CPT testing have evolved over

the past 10 years. Many initial proposals were

not very refined and sometimes depended on extensive site-specific correlation with the SPT

test. In 1985, both Robertson (9) and Ishihara

(10) proposed relationships with normalized cone

tip resistance and cyclic strength. Both methods were felt to only be applicable to clean,

low compressibility sands as an increase in

fines or more compressible minerals such as micas would greatly reduce cone resistance. At

the ASCE specialty conference on in situ

testing, Seed (11) proposed a similar relationship based on refined SPT-CPT correlation.

Only two case history sets of information were available to validate the proposed method: The

method included correlations to account for

fines content.

Some researchers have argued that the cone

appeared, to be too insensitive to stress-fabri

effects which are beneficial to liquefaction

resistance. Olsen (12) presented a method which

incorporates normalized tip and sleeve

resistance to evaluate liquefaction strength.

This method was developed for the electric cone

with a sleeve directly behind the tip. The difficulty with this method is that sleeve data

must be accurately determined. In cohesionless

soils, sleeve resistance is typically one-tenth

of the tip resistance; and additional accuracy

of less rugged, independent load-cell electric cones is required for the method.

Researchers working on the Beaufort Sea drilling

islands developed a correlation of state parameter to normalized cone resistance to study

failures. The concept of the state parameter

was presented by Been and Jeffries (13) in 1985.

The state parameter measures the distance or position of a sand in given void ratio stress

condition to a reference steady-state line. The

steady-state line is the same as defined by

Castro and Poulos and originally proposed in the

critical void ratio concept of Casagrande. The

advantage to this approach is that many significant engineering parameters such as static

stress - strain behavior can be correlated to

the state parameter. In addition, positive

values of the state parameter are indicative of

collapse potential of sand structure under

existing stress conditions resulting in very

large strain development in loading events such

as earthquakes. Recently, Been, et al . (14),

have determined a correlation of normalized cone tip resistance with state parameter by determining steady-state lines for sands used in cone chamber tests.

In this report, each of the concepts for CPT

evaluation of liquefaction will be studied.

Depending on the results, more in-depth studies

will be made. Conclusions regarding the general

applicability of the methods will be made.

An additional study was performed with the flat

plate dilatometer (DMT) testing device to gain

valuable case history experience with the

instrument. The Public Works Research Institute

arranged for a field study with this device

which took place in the Hachiro Gata Lagoon area during October 29 through November 1, 1986. The

DMT test has been discussed by Marchetti (15)

who developed the test in the mid-1970's. In 1982, Marchetti (16) discussed the possible use of the test for liquefaction evaluations.

Robertson (17) proposed a correlation of the

horizontal stress index to cyclic stress ratio.

Based on discussions at the UJNR workshop on in

situ testing and with users and manuf acturers of

the DMT, the differences between dynamic and

quasi-static insertion were also evaluated to see if sensitive sands could be identified.

2. DISCUSSION

2 . 1 SPT Simplified Methods

Two methods of evaluation of liquefaction resistance using the SPT find widespread usage. In Japan, Iwasaki and Tatsuoka, et al . (18), have

developed a method whose basis is in a large

quantity of undisturbed sample testing to

240
develop a correlation between cyclic strength

for 20 cycles of equivalent loading ( R

20 ) » SPT

N value normalized for overburden pressure

effects, and particle size expressed in terms of mean particle diameter (D 50 ).

In the U.S. Seed (5) and coworkers developed a method which separates liquefaction and nonliquefaction from observed field performance data

on a plot of cyclic stress ratio and SPT N1

value normalized to a stress level of 1 ton/ft?.

The method also distinguishes for fines content

in sands. Neither method is applicable for lowplasticity silts or gravelly soils.

When studying previous liquefaction sites, the

most difficult task is to determine the actual zones of liquefaction. Since SPT testing is the

most widely accepted method of evaluating liquefaction, it will for the basis for testing cone penetration and other methods for their ability

to predict liquefaction. As a result, the SPT

analysis must be carefully performed to evaluate

zones of liquefaction.

Results of SPT analysis using both SPT techniques by Ohya and Iwasaki ( 8 ) are shown on Fig. 5. On this figure,' liquefied and critical

zones where factor of safety is close to 1 are shown. Inspection of these plots indicate there

is good agreement among the methods in their

predictions. One will also notice, however,

that the two methods diverge at high and low

factors of safety. This trend has been identified in the past by others. The divergence at

large factors of safety could possibly be

explained by loosening of undisturbed samples of

dense sands during the sampling process, thereby

reducing the R 1 20 strength.

The analysis of factor of safety by the Seed

N]_ 6 o method was reperformed to incorporate

energy corrections and corrections for sampler

barrels in accordance with the 1984 report.

Calculations for this analysis are summarized on Table 2. The boring logs used for this analysis

were reported by 0Y0 Corporation in March 1984.

To estimate the induced cyclic stress ratio at

the depth of interest, the simplified equation

proposed by Seed was used. To simplify calculations, a single unit weight of 120 lb/ft^ was

used and the reduction factor rq was the equation proposed by Iwasaki (18). SPT N values

obtained with the ASTM sampler were increased to

account for larger inside barrel diameter by

using the recommendation by Kovacs (4). The

factor of 20 percent is probably too large for

low blow count soils, and a factor of 10 percent

is probably more appropriate. After conversion

for sampler size, the N value was then adjusted

to an energy level of 60 percent using the

equation proposed by Seed and Kovacs (4, 5). At

each field study site, only several energy data

points were available at depths of greater than

30 ft. Where energy data were available,' they

were directly applied to the N value obtained;

and an averaye value was used for the remainder

of the boring. At site E, energy data were not

available; but the same operator and drill rig

previously used at site D were again used. At

site F, no data could be assumed; therefore,

Japan average energy was used (4). Corrections

for rate and borehole size, as recommended by

Seed (5), were not used as it was felt that they were purely speculative and probably overemphasi zed

After conversion to Ngg, the value was then nor- malized to an overburden pressure of 1 ton/ft 2 using the following equation for the factor C^;

C N = (ay) n = (aj )' 0,5 Equation 1

with ay (vertical effective stress) in ton/ft^

The graphic C

N values proposed by Seed (1984) were not used^as it requires prejudging the relative density of the deposit. The values of C^

using the above equation are median points of

those from the two curves recommended by Seed.

After obtaining values of N^gO along with fines

content information, cyclic strength was determined by visual examination of the go chart.

If fines were present, increased strength was interpolated from the curves for various ranges

of fines contents. In some cases, the cyclic

strength was estimated as infinite following

concepts of limiting shear strain potential.

The resulting change in factor of safety from

both SPT methods is shown oh figures 6 and 7.

Factors of safety from the Japanese method were

not recomputed but scaled graphically from the

report by Ohya and Iwasaki. It can seen that,

at lower factors of safety, the agreement is now better between the two methods in most cases.

This can be attributed to the higher energy

level (68 percent) delivered in Japan. Correction to the Ngg value results in an increase of

the N value. Remaining differences at a low

factor of safety can be attributed to slightly

higher cyclic strength observed in the Japan

undisturbed sample test data due to consolidation of loose sands.

2.2 Analysis of 1983 Field Data Using the QC1

Method

The method presented by Seed (11) is refined

correlation of cyclic stress ratio to normalized

cone penetration resistance (Qcl). The effects

of particle size on the correlation were inferred by refining the Qc/N versus D 50 data

base to incorporate energy consideration on the

SPT value and then using the SPT data base to

develop the relationship. In this approach, the

N value should be normalized to the 60-percent

energy level to check the Oc/Ngg versus D 50

relationship. Using Oc /Ngo and D 50 , the Ctl

method could be extended to silty sands by

inferring fines content from D 50 ; and dividing

lines separating occurrence of liquefaction were developed. The chart for clean sands is very
similar to the charts proposed by Robertson and

Ishihara (9,10). Seed compared this approach to case history data at the Niigata Kawagi shi -Cho

apartment site [Ishihara and Koga (1)] and the

Imperial Valley study sites in the U.S. [Youd

(19)]. Additional field performance data from

this study will assist in validating this

method

In the 1985 UJNR workshop on in situ testing,

Ohya and Iwasaki presented evidence that the

ratio of Ct to dynamic blow count N<j from the

Swedish dynamic penetration test could be used

to predict occurrence of liquefaction. Larger

ratios of Q-/Nd had indicated increasing possibility of liquefaction. In this study, the

Qc/ n 60 ratio was also studied for this effect.

A comparison of Qc/N60 versus D 50 particle size

is shown on figure 8 . On the figure, points

which did or did not liquefy according to SPT

Ni 60 analysis are appropriately marked. From

this plot, there is no consistent trend in

liquefaction occurrence; and there is wide

scatter among the data.

Begemann friction mantle cone (ISSMFE M2

penetrometer) tests were performed at each of

the study sites. Tip resistance data, Qc , were normalized for overburden pressure effects using

the same correction factor as used in the

N x 60 analysis. It should be noted that this

normalization proposed by Seed (11) is not in

accordance with findings presented by

Jamiolkowski (20), who recommended the n exponent of equation 1 to be -0.72 with ay in

kg/cm2. As with the SPT analysis, a single unit

weight of 120 1 b/ ft 3 is used with all stress

computations

A dependent analysis of occurrence of liquefaction was performed to evaluate the C^i method.

At each SPT depth interval, C^i values were averaged. Plots of Qci and cyclic stress ratio were made on the proposed chart, and points

which did or did not liquefy from SPT prediction

could be marked. A summary of dependent data is

given in Table 3, which also shows the average

friction ratio in the SPT interval. All of the

values are less than 2.5 percent, which is the common borderline dividing sands and sand-siltclay mixtures using the Begemann friction mantle

cone. For electrical cones, this value will be

different. The Qci method should not be used

for liquefaction assessment using the Begemann cone if higher friction ratios are obtained.

A summary chart, Fig. 9, shows that liquefaction occurrence is accurately predicted at the sites.

From the individual charts for each site the

cyclic strength was determined by visual inspection, taking into account the percentage of

fines and D 50 particle size. The cyclic

strength and factor of safety for the 1 data

are summarized in Table 4 and factor of safety

is plotted on Figs. 6 and 7 for comparison to

SPT methods. The 0q \ factor of safety compares

very well with SPT methods. It is interesting

to note that, at site A, liquefaction is predicted to greater depth using the Qci method.

This may possibly reflect on the greater uncer- tainty of overburden pressure corrections for

the SPT at shallow depths.

Results of the study of the Qc i method indicate

that the method is very suitable for evaluation

of liquefaction resistance for clean sands.

Since no sample is obtained in the test, the

friction cone should always be used to check

friction ratio. Further field evidence is

required to evaluate silty sands and sand deposits which have been subject to larger cyclic

stress ratios. Additional study is also

required for sands of varying compressibilities

due to varying mineralogies.

2.3 Evaluation of Olsen's CPT Method of

Liquefaction Assessment

A method of liquefaction resistance evaluation

using normalized cone and friction sleeve data

from the CPT test was proposed by Olsen (1984).

By incorporating normalized sleeve resistance,

sensitive soils can be detected. The CPT tip

resistance is relatively insensitive to stress

history effects, such as overconsolidation, and some aging effects, such as slight cementation.

The sleeve is more sensitive to effects, such as overconsol idation.

Olsen developed a soil characterization chart

that identifies sensitive, normally consolidated, and overconsolidated behavior. The

advantage of this method is that it may also

identify sensitive silts and silt-sand mixtures

which are potentially liquefiable.

Normalization of the tip resistance uses the same form as previous sections of this report;

but the exponent n in equation 1 varies with

soil type, and an iterative process is used.

Sleeve friction is normalized by dividing by the

effective overburden pressure, since it is a measure of remolded strength. Using SPT-CPT and

laboratory cyclic stress ratios, stress ratio

contours of liquefaction resistance strength were drawn on the characteri zation chart.

The relationship developed by Olsen was developed for electric cone testing with friction

sleeves located immediately behind the tip.

The sleeve friction measured by the Begemann

friction mantle cone is located further back

behind the cone tip. The sleeve has a shoulder

or bearing surface at its base and, therefore, measures higher levels of sleeve friction.

DeRuiter (21) states that the electric friction

is, on the average, only one-half of Begemann

mechanical friction values. In electric cone testing with subtracti on-type cones, the

sleeve friction can be in error as much as

± 0.3 ton/ft 2

, making it difficult to obtain

accurate friction readings.

For this study, the Begemann cone data were nor- malized for a qualitative comparison with

242
Olsen's chart to evaluate if use of such a chart

was feasible for mechanical cones. The exponent

n to obtain the correction factor to normalize

Qc was taken as -0.72, since practically all of

the soils were clean sands. Friction sleeve

data were not reduced to be more comparable to

electric cone friction values. Normalized tip

and sleeve resistance data points are summarized

in Table 3. A summary plot with data points for

all six sites is shown on the soil characterization chart on Fig. 10. On Fig. 11, data

from site B are plotted on the characteri zati on chart, which contains the contours of cyclic

strength. Beside the data points on this

figure, the SPT N^g cyclic strength values are shown

From review of the figures, it is evident that

the Begemann friction values are higher than the

electrical data. It is possible that the cyclic

stress ratio contours and characterization zones could be shifted to the right to make Olsen's

method applicable to the Begemann cone. The

general trend of liquefiable data points lying

on the sensitive side of the normally consolidated zone confirm the concepts of this method.

Dense dilative sands fall in the overconsolidated zone.

positive value of state parameter, are contractive in shear. The contractive condition indicates potential collapse of structure with

extremely large deformations. Contractive soils are rare in natural deposits; but if identified,

they can exhibit extensive deformation upon

earthquake or additional static loading. Been

and Jeffries (13) have related the state parameter to angle of phase transformation, drained

peak angle of shearing resistance, dilation

rate, and decrease in peak to steady-state friction angle.

In a currently unpublished paper. Been, et al . (14), have determined the steady-state lines for

all of the sands used in CPT chamber testing and

have found a unique relationship to Qc nor- malized to mean effective confining pressure and

the state parameter in clean sands. The initial

correlation developed is shown on Fig. 13.

Additional study is needed to determine the

relationship for silty sands.

In this study, it was desired to calculate the

state parameter for qualitative comparison to

liquefaction resistance. The equation proposed

by Been, et al . (14), to determine the state

parameter from Qc is as follows:

Overall, conclusions regarding this approach are that the method in concept is fundamentally

correct. Possibilities exist to modify this

relationship for mechanical cones. The benefit

of this approach is that it may be able to identify sensitive silts and sand-silt mixtures and

eventually apply cyclic strength estimates. For

now, if sensitive sands with fines or silts are identified by the method, laboratory testing may

be required. Disadvantages to the method are

its log-log relationship, which makes resolution

difficult. Also, sleeve friction measurement is more sensitive and subject to error.

Since accurate conversions of mechanical to

electrical cone data cannot be made, electrical

cone data are required from case history sites

in clean sands, sands with fines, and silts to

val idate the method .

2.4 Evaluation of the State Parameter for

Liquefaction Assessment

ii> = -.092 In Equation 2

where Ij and I x

' are the first total and effective stress invariants (mean normal stress). In equation 2, stress is expressed in kPa. Calculation of state parameter requires knowledge of

the value of Kg (ratio of in situ vertical to

horizontal stress). Based on results of DMT

tests to be presented later in this paper, a Kg

value of 0.6 was assumed for liquefied sites and

1.0 for nonliquefied sites. The state parameter

was calculated for all Qc values.

Average state parameter values in SPT depth

intervals are summarized in Table 4. Inspection

of these data reveals extremely low state parameter values in a low cone bearing zone from 5.0

to 6.5 m in depth at site B. Liquefaction sites

D and E had more moderate values of the state

parameter

The state parameter (ii>) proposed by Been and

Jefferies (13) is simply a convenient method to

develop a common referencing system which can quantify static stress - strain behavior of

sands and eventually sand-silt mixtures. The

concept of the state parameter is shown on Fig. 12. The unique reference line is the

steady-state line which is fairly easily

measured on remolded contractive specimens. The

state of the soil is expressed as a function of mean normal stress and void ratio. Soils, whose

position is the below the steady-state line with

a negative value of state parameter, will exhibit dilatant behavior in shear. Soils, whose

state lies above the steady-state line with a

The state parameter for the zone present in site

B is very close to being a contractive value;

but if the equation for state parameter is examined, it is found that extremely low values of

tip resistance would be required to indicate a contractive condition. In fact, the tip

resistance may approach zero since the sand

structure collapses away from the cone.

Results of this analysis indicate that moderately dilative sands are subject to level ground

liquefaction and resulting excessive deformations. In order to develop a correlation in

terms of the state parameter, the Ct]_ versus

243
cyclic strength relationship was used. Qcl values for clean sands from the chart proposed by

Seed were converted to state parameter values by

assuming Kq = 0.5 with typical sand unit weight

and ground-water location at the ground surface.

Then, state parameter points calculated in SPT

intervals were plotted with liquefaction occurrence as predicted by the SPT N^gg method noted.

The results are shown on Fig. 14.

The results indicate that the majority of the

liquefied data points fall in the proper position, with exception of two near-surface data

points at sites B and C and a point at site D at

7.3 m with high cone bearing. At site C, there

was no evidence of liquefaction. As discussed

earlier, shallow blow counts from the SPT are more difficult to evaluate due to uncertainty of

pressure normalization. Also, erratic energy

transmission is a problem with the SPT at

shallow depths. It is highly probable that the cone can better estimate near-surface liquefaction occurrence.

The effect of deriving a state parameter versus cyclic strength correlation based on Kg = 0.5

conditions will have a similar effect as the

method, which disregards varying K0

in pressure

correction of the N value. In equation 2, the

state parameter is most sensitive to the Qc

value and less sensitive to mean normal stress,

since the tip resistance is typically two orders

of magnitude larger than mean normal stress

levejs. At higher Kq values, the dominant

increase in 0C will cause a more negative value

of state parameter to be estimated if Kq = 0.5

conditions are assumed. This will result in

predicted increase in cyclic strength, which is

appropriate for liquefaction studies. However,

the Kq = 0.5 assumption would overestimate the

dilational tendency for static stress - strain

behavior predictions.

In conclusion, there seems to be a relationship

between the state parameter and occurrence of

liquefaction. Further study should be performed

to evaluate the effects of higher Kq values on the state parameter and its ability to predict

liquefaction. Since it is difficult to determine field Kq values, a Kq - 0.5 assumption is

appropriate for young alluvial soils which are typical subjects of liquefaction studies. The

shortcoming of the method is that rel ationshi ps

still need to be developed for silty sands and

silts. Additional studies are required to test

sand in a contractive condition and sands which

contain fines. The benefit of this approach for

clean sand at this time is that static stress - strain behavior of the soil can also be estimated. As with the Qcl method, there is no sample obtained, so the method should only be

applied with friction ratios of less than

2.5 percent and gradation of the sand should be

checked to be sure the fines content is less

than 10 percent.

2.5 Analysis of FI at Plate Dilatometer Test

Results

The flat plate dilatometer test (DMT) was introduced by Marchetti in the mid 1970's. Marchetti

(15) presented a summary of the equipment and

data reduction procedures in the ASCE

Geotechnical Journal . The DMT consists of a flat blade-shaped penetrometer with a 60-mm circular expandable membrane located on the face of

the blade. The blade is usually quasistatically advanced at 20-mm/s at 20-cm intervals, while thrust pressure at the surface is

recorded. At each 20-cm increment, pressure is

released from the rods and the membrane is

expanded. Two readings are obtained at each

i nterval :

- First, an internal pressure to lift the

membrane into an approximately planar

condition

- Second, an internal pressure to extend the

membrane approximately 1.1 mm into the

soi 1

The two readings are obtained in about

30 seconds. After subtraction of inertial pres- sures required to expand the membrane freely in

air, values of Pg and Pi are obtained from the

first and second readings, respectively. Based

on the two readings, three index parameters are determined as follows:

Material or Deposit Index, Id I d

=

Lateral Stress Index, Kd K , =

Dilatometer Modulus, Ed E. = 34.7 (P. - P_)

d 10

The material index is used to determine the soil

type. It represents the normalized difference

between 1-mm expansion pressure. Pi, and contact

pressure, Pg, which is the measure of soil

rigidity. Since soil classification is based

only on rigidity, some errors in estimated soil

type may exist. The horizontal stress index is

related to lateral earth pressure at rest. The

dilatometer modulus is a numerical solution for

(E/1 - y2) when expansion is modeled as loading

of a circular area in an elastic half space.

The modulus obtained must be empirically related

to soil modul us .si nee drainage conditions and

disturbance vary.

Dilatometer parameters have been related to Kq,

a', Su , P'c, E 25 (sand), and E (clays). The

soil friction angle in sands is now based primarily on thrust data and bearing capacity analysis [Schmertmann (22)]. Based on initial

244
chamber tests in normally consolidated and

prestressed sands, Marchetti (16) felt that the

instrument was more sensitive to prestress than

the CPT and proposed it as a viable method for

evaluation of liquefaction potential. Marchetti

proposed that kq values of less than 1.5 would

indicate sands at relative densities of less

than 60 percent and that resistance to liquefaction in terms of cyclic strength may be roughly

proportional to Kq/10. A most recent paper

showing additional chamber test results in sands

was presented by Baldi, et al . (23). Robertson

(17), after consideration of chamber test data

and a relationship between relative density and

liquefaction resistance proposed by Christian

and Swiger (24), proposed a correlation between

the Kq and liquefaction resistance. After

attending the 1985 UJNR workshop on in situ

testing, DMT users felt there may be some relationship of liquefaction to the change in Kq

with static and dynamic advancement. In this

study, both the effects of static and dynamic

driving and Robertson's method will be analyzed.

On October 29 through November 1, 1986, the

Public Works Research Institute, Ministry of

Construction, Japan, performed a series of nine

soundings using DMT and CPT testing in the

Hachiro Gata Lagoon area. DMT equipment was supplied by the Bureau of Reclamation for this

study. Six soundings were located in sites B and C of the 1983 field study. An additional

site A' was located near the Gomyoko Bridge at

dike station WC 13+620 where the dike had

settled up to 1 m, and approximately 6 m of

banking soil were available for testing. CPT

and SPT data were available for site A

1 from

previous investigations. At each site, a cone penetration test using a Dutch mantle cone (ISSMFE penetrometer Ml) was performed to evaluate any change in soil conditions since the

dikes were rehabilitated after 1983. The

rebuilding did not include subsurface

densi fi cation

At each site, dynamic and static DMT tests were performed to evaluate effects on the Kq index.

Static DMT tests were advanced with a portable

cone penetration machine which used anchors to

develop additional reaction. Thrust pressure to

quasi -statically advance the DMT was measured at

the surface with CPT hydraulic pressure gauges.

At site B, the pressure gauges malfunctioned and

a second sounding with the DMT blade was performed. The Swedish automatic ram sounding

equipment was used to dynamically advance the

DMT. This device uses a chain cam mechanism to

lift and drop the hammer so that consistent

energy is applied. The 63.5-kg hammer is

dropped from 50 cm. The DMT blade was driven

with CPT rods for 20 cm, and the number of blows were recorded for each increment. On the two

dynamic soundings at sites B and C, the hammering action loosened the wire contact at the

base of the sensing disk at about 7 and 5 m,

respecti vely ; and the soundings were terminated.

No other specific equipment problems occurred.

The completed sounding numbers, test site, test

type, and depths are listed below:

Sounding

No. Site Test type Depth

86-1-2 B Quasi -static 15 m

86-1-3 B Dynamic 7 m

86-2-2 A' Quasi -static 8 m

86-2-3 A' Dynami

7 m

86-3-2 C Quasi -static 5 m

86-3-3 C Dynamic 4 m

DMT data were reduced using IBM PC Basic program

DILLY4 distributed by GPE, Inc., the U.S.

distributor of the DMT. The program provides

tabular output of all parameters and a graphic

output of strength, stress, and modulus values.

Computer data reduction used methods recommended

by Schmertmann (22) for plane strain friction

angle and Kq. For the dynamic soundings, the

thrust data from the adjacent static tests were

input instead of trying to convert the number of

blows to thrust.

Summary graphs which contain pertinent data from

both quasi-static and dynamic tests on one graph

for each site are shown on Figs. 15, 16, and 17.

Cone penetration data from the 1983 field studies and the current study are plotted in terms

of C^i for sites B and C are shown on Figs. 18

and 19. From Fig. 18, it can be seen that 0c i has increased to a depth of 5 m since 1983.

Below 5 m, the same zone of low cone bearing

sands exists with little increase in C^. At

site C, it can be seen that Cti has increased

over a depth of 2.5 to 5 m.

At site A', the DMT summary graph (Fig. 16)

indicated an interbedded deposit of silty sands

and sands with highly erratic data. Cone tip

resistance data also indicated soils with fines.

Data from site A

1 were not included in the analysis of liquefaction resistance. It will be

included when additional gradation and cone data

from the previous borings can be analyzed.

The effects of dynamic driving of the DMT in

loose and dense sands can easily be distinguished from the summary graphs. Dynamic driving causes an increase in material index and

friction angle and a decrease in horizontal

stress index and modulus. Of interest for liquefaction potential is the decrease in Kq . At

site B, dynamic driving resulted in reducing Kq

to below 1 in critical liquefied zones. At

site C, the reduction in Kq is also apparent;

but values remained above 2. In an attempt to

quantify this reduction, the ratio of dynamic to

static Kq were analyzed. At site C in dense

sand, a 60 percent reduction of Kq occurred at

245
3 to 4 m in depth. At site B, an 80-percent

reduction in Kd occurred at a depth of 5 to 6 m

in loose sand. However, a 90-percent reduction

in Kq occurred at 4 to 5 m at site B where nor- malized cone bearing was significantly higher.

Additional checks of dynamic and static Kq

ratios resulted in no correlation to liquefaction resistance. The only conclusion that can be drawn is that, if dynamic Kq is less than 1,

liquefaction seems highly likely.

The quasi -static Kq values were then compared to

Robertson's method. Kq values were averaged in

the 1983 SPT intervals. Using the SPT factor of

safety as a guide, values in zones which had

liquefied at site B could be plotted on the proposed relationship by Robertson on Fig. 19.

These data confirm a limiting Kq value of 4.2 at

cyclic stress ratio of 0.18. At site C, the Kq

value varied considerably in the SPT interval;

and a bar was used to show the nonliquefied

zone. Since 0c j_ had increased in 1986, the Q^i cyclic strength was used to show increased

strength of the deposit. This approach indicates the Kq versus cyclic strength line may

increase more sharply than proposed by

Robertson. Additional field performance data

will be needed to confirm the Kq relationship

at higher cyclic stress ratios.

In conclusion, the field tests at Hachi ro Gata

have resulted in a better understanding of the

DMT in sands. Major changes can occur between

static and dynamic DMT tests to the point where

dynamic driving should be discouraged in loose

and dense sands alike. It is doubtful that a useful correlation between dynamic and static Kq

values and liquefaction resistance can be developed. Robertson's proposed correlation works

well, but additional information will be needed

at higher levels of cyclic loading and in dense

sands at low levels of cyclic loading.

Since material index is not necessarily a good

indicator of soil type, there may also be questions regarding fines contents in the soils

tested

2.6 Comments on the Static and Dynamic

Penetration Resistance

So far on this project, there have been four

types of static dynamic comparisons of penetration resistance attempted. The ratio of cone resistance to SPT resistance was shown on Fig. 10 to be highly erratic. Likewise, Kq

ratios obtained in this recent study do not show

consistent trends. Ohya and Iwasaki (1985)

showed some success in correlating CPT Qq to the

Swedish dynamic penetration Nq value. However,

there is also considerable scatter among the

data; and the relationship could not be correlated to quantify cyclic strength. Sasaki and

others at the Public Works Research Institute

have shown the vibrocone to have good success,

at least in a qualitative sense. If any method

of a static to dynamic comparison methods stands

a chance of success, it seems to be the

vibrocone.

Recent efforts to quantify the decrease in cone penetration resistance, D, from the VCPT with

cyclic strength are summarized by Sasaki, et al

(7). In Table 4, the most recent version of the

equation expressing cyclic strength ( R 1 20 ) ver ~ sus decrease in cone resistance and effective

overburden pressure was used to calculate

strength. The D value was obtained for the

study sites from plots shown in Sasaki, et al . (3). A review of Table 4 indicates that

liquefaction would not have been predicted at

any of the study site SPT intervals.

Review of the decrease in resistance diagrams

indicates a general trend of reduction of D with

depth. In current field use, the force on the cone tip is being measured at the ground sur- face. As a result, friction buildup on the rods

is influencing the penetration resistances. The

recent modification to incorporate effective

overburden pressure into the strength equation

may compensate for this effect; but it is

possible that friction buildup occurs at different rates for the static and dynamic soundings which are performed independently. If

this is the case, the static and dynamic com- ponents of D may have to be adjusted separately.

Static Qc can be compared to CPT Cfc data and

friction can be evaluated; but the only solution

to evaluating the effect on vibratory Qq is to

locate a load cell within the tip or just above

the vibratory section.

A second problem with the correlation is the use of cyclic strength data from undisturbed samples

obtained the field. The sampling process has a tendency to density loose sand. Trimming and

reapplication of confining stresses may further

density the sand. The converse effect can occur

with dense sands. As a result, there is a general flattening of the cyclic strength versus

D curve. On Fig. 21, the two equations proposed

for the vibrocone are plotted on a cyclic

strength versus D diagram. Also plotted on the

diagram are the SPT N^gO cyclic strength data

and D value summarized in Table 4. In the area

of 100- to 60-percent reduction (1 - D = 0.0 to

0.4), SPT strength data for a given D value seem reasonable. At smaller values of D or as 1 - D approaches 1, there is considerable scatter.

The point of lower cyclic strength in this area originate from deeper zones at liquefied sites

where friction has reduced the magnitude of

decrease. If the friction problem is solved, it

may be possible to develop the anticipated trend

line shown on the figure.

To summarize, two recommendations can be made to

improve VCPT performance. First, the effects of

rod friction must be omitted or determined

accurately. Second, either of the three

approaches which follow could be taken to

develop the cyclic strength correlation; laboratory chamber studies of sands of known cyclic
strength, field cyclic strength determinations

on sands which are frozen prior to sampling; or correlations to accepted in situ testing techniques such as Iwasaki - Tatsuoka, N^gy, or Qcl

approaches. In order to extend the prediction

into sands containing fines, it may be helpful

to add a friction sleeve to the static penetrometer system to assist in identifying these

materi al s

3.

CONCLUSIONS

Comparisons of liquefaction resistance evaluation techniques using the SPT, CPT, DMT, and

vibrocone have been performed with the data

obtained from the 1983 U.S. -Japan cooperative

field study sites and a field study performed in

1986. Pertinent conclusions regarding the methods considered along with recommendations for

further study are as follows:

a. SPT data were reanalyzed to incorporate

energy and sampler corrections using the N^gg

method proposed by Seed (5) and compared to

the Iwasaki, et.al. (18), method. The

methods agree very well with only slight differences at low factor of safety. At a large

factor of safety, the methods diverge due to

the limiting shear strain concept of the Nigg

method

b. CPT data were analyzed using the Cti

method proposed by Seed (11). Results of an SPT dependent analysis and independent Cfci factor of safety comparisons show the method

was very reliable and suitable for use in

liquefaction analysis. Additional field performance comparisons in silty sands and sands

subjected to higher levels of cyclic loading

are required.

c. CPT data were analyzed for a qualitative

comparison to the method proposed by Olsen

(12). The method in concept is fundamentally

correct, but modifications would be required

to develop the relationship for mechanical

cones

d. CPT data were analyzed and a method

developed to evaluate liquefaction resistance

using the state parameter proposed by Been

and Jeffries (1985). The method should have

comparable accuracy as the Qcl method and has

the additional benefit of estimating static

stress - strain behavior in shear.

Additional study will be required to evaluate

prestress effects and extend the correlation

to silty sands.

e. A series of six DMT soundings were performed by the Public Works Research Institute

at the Hachi ro Gata Lagoon area during

October and November of 1986. Comparisons of

static and dynamic K<j values did not indicate

a trend correlated to liquefaction

resistance. Static Kq values were compared

to the method proposed by Robertson (1984)

with favorable results, additional studies of

silty sands and sands subjected to higher

levels of cyclic loading are required.

f. Of all static-dynamic penetration techniques, the vibrocone test has the best

potential for accurate prediction of

liquefaction resistance. Equipment or calculation methods need to be improved to over- come rod friction effects. The correlation

of penetration decrease to cyclic strength

requires further study.

4. ACKNOWLEDGMENT

This study was made possible by a science award

for foreign specialists by the Science and

Technology Agency, Japan. The host agency was the Public Works Research Institute, Ministry of

Construction, Japan. Special thanks are extended to Drs. Sasaki and Iwasaki for their

interest and assistance with this study. Also,

thanks are extended to Matsumoto and Kondo of

the Ground Vibration Division, PWRI, for

assistance in field DMT testing and data

reduction

The Bureau of Reclamation approved my detail and

supplied DMT testing equipment to perform this

study.

5. REFERENCES

(1) Ishihara, K. , and Y. Koga, "Case Studies of

Liquefaction in the 1964 Niigata Earthquake,"

Journal of the Japanese Society for Soil

Mechanics and Foundation Engineering , vol . 21,

No. 3, September 1981.

(2) Iwasaki, T. , K. Tokida, and S. Yoshida,

"Investigation on Seismic Resistance of Grounds

at Showa Bridge Damaged During the Niigata

Earthquake," Technical Memorandum No. 1591,

Public Works Research Institute, Ministry of

Construction, Japan.

(3) Sasaki, Y., K. Kutara, T. Iwasaki, H. Miki,

Y. Itoh, K. Kaminaga, W. D. Kovacs, L. A. Salomone, J. A. Farrar, "U.S. -Japan Cooperative

Research on In-Situ Testing Procedures For

Assessing Soil Liquefaction (No. 1),"

Proceedings of the Sixteenth Joint Meeting,

U.S. -Japan Panel on Wind and Seismic Effects,

UJNR , Washington, D.C., May 15-18, 1984.

(4) Kovacs, W. D., and L. A. Salomone, "Field

Evaluation of SPT Energy Equipment, and Methods

in Japan Compared with the SPT in the United

States," NBSIR 84-2910, National Bureau of

Standards, Center for Building Technology,

Gaithersburg, Maryland, August 1984.

(5) Seed, H. B., K. Tokimatsu, L. F. Harder,

and R. M. Chung, "The Influence of SPT

Procedures in Soil Liquefaction Resistance

Evaluations," Report No. UCB/EERC-84/15

,

College of Engineering, University of

California, Berkeley, California, October 1984.

247
(6) Sasaki, Y., Y. Koga, Y. Itoh, T. Shimazu,

and M. Kbndo, "In-Situ Test for Assessing

Liquefaction Potential Using Vibratory Cone

Penetrometer Proceedings of the Seventeenth

Joint Meeting, U.S. -Japan Panel on Wind and

Seismic Effects, UJNR, Tsukuba, Japan,

May 21-24, 1985.

(7) Koga, Y., Y. Sasaki, Y. Itoh, and 0.

Matsuo, "Field Investigation of Seismically

Damaged Sites by Use of Vibratory Cone

Penetrometer," Proceeding of the Eighteenth

Joint Meeting, U.S. -Japan Panel on Wind and

Seismic Effects, UJNR, Washington D.C.,

May 12-15, 1986.

(8) Ohya, S., T. Iwasaki , and M. Wakamatsu,

"Comparative Study of Various Penetration Tests

in Ground that Underwent Liquefaction During the

1983 Ni honkai -Chubu and 1964 Niigata

Earthquakes," Proceedings of the First

U.S. -Japan Workshop on In Situ Testing Methods

for Evaluation of Soil Liquefaction

Susceptibility, San Francisco, California,

August 17-20, 1985 (unpublished).

(9) Robertson, P. K. , and R. G. Campanella,

"Liquefaction Potential of Sands Using the CPT,"

Journal of the Geotechnical Engineering

Pi vi si on , American Society of Civil Engineers,

vol . Ill , No. 3, March 1985.

(10) Ishihara, K. , "Stability of Natural

Deposits During Earthquakes," Proceedings,

Eleventh Internati onal Confrence on Soil

Mechanics and Foundation Engineering, vol . 1,

August 1985.

(11) Seed, H. B., and D. DeAlba, "Use of SPT

and CPT Tests for Evaluating the Liquefaction

Resistance of Sands," ASCE specialty conference

on the Use of In Situ Tests in Geotechnical

Engineering, Blacksburg, Virginia, June 1986.

(12) Olsen, R. S., "Liquefaction Analysis Using

the Cone Penetrometer Test (CPT)," Proceedings

of the Eighth World Conference on Earthquake

Engineering, San Francisco, California, July

1984.

(13) Been, K. , and M. G. Jeffries, "A State

Parameter for Sands," Geotechnique, vol. 35,

No. 2, 1985.

(14) Been, K. , J.H.A. Crooks, D. E. Becker, and

M. G. Jeffries, "State Parameter Interpretation

of the Cone Penetration Test in Sands," submitted for publication in Geotechnique, March

1985.

(16) Marchetti , S., "Detection of Liquefiable

Sand Layers By Means of Quasi-Static Penetration

Tests," Proceedings of the Second European

Sysposium on Penetration Testing, Amsterdam,

May 24-27, 1981.

(17) Robertson, P. K. , and R. G. Campanella,

"The Flat Plate Dilatometer Test for

Liquefaction Assesment," Soil Mechanics Series

No. 79, Department of Civil Engineering,

University of British Columbia, Vancouver, 1984.

(18) Tatsuoka, F., S. Iwasaki, K. Tokida,

S. Yasuda, M. Hirose, T. Imai , and M. Kon-no,

"Standard Penetration Tests and Soil

Liquefaction Potential Evaluation," Soils and

Foundations, JSSMFE, vol. 20, No 4., December

1980.

(19) Youd, Te L. , and M. J. Bennett,

"Liquefaction Sites, Imperial Valley,

California," Journal of the Geotechnical

Engineering Division , American Society of Civil

Engineers, vol. 109, No. 3, March 1983.

(20) Jamiol kowski , M., G. Baldi, R. Bellotti,

V. Ghionna, and E. Pasqualini, "Penetration

Resistance and Liquefaction of Sands,"

Proceedings of the Eleventh International

Conference on Soil Mechanics and Foundation

Engineering, San Francisco, California, 1985.

(21) deRuiter, J., "Electric Penetrometers for

Site Investigations," Journal of the Soil

Mechanics and Foundations Division , American

Society of Civil Engineers, vol. 97, No. SM2,

February 1971.

(22) Schmertmann, J., "A Method for Determining

the Friction Angle in Sands from the Marchetti

Dilatometer Test," Proceedings of the Second

European Symposium for Penetration Testing,

vol. 2, Amsterdam, May 1981.

(23) Baldi, G., R. Bellotti, V. Ghionna,

M. Jamiol kowski , S. Marchetti, and E. Pasqualini, "Flat Plate Dilatometer Tests in

Calibration Chambers," Use of In Situ Tests in

Geotechnical Engineering , Geotechnical Special

Publication No. 6, American Society of Civil

Engineers, June 1986.

(24) Christian, J. T., and W. F. Swiger,

"Statistics of Liquefaction and SPT Results,"

Journal of the Geotechnical Engineering

Division, American Society of Civil Engi neers

vol . 101, No. GT-11 , 1975.

(15)

Marchetti, S., "In Situ Tests by the Flat

Dilatometer," Proceedings of the Geotechnical

Division, American Society of Civil Engineers,

vol. 106, No. GTS, March 1980.

248
Items of Field Investigation

RaglOB 1

i Akita (Lagoon Uacmro)

1 Niigata

Site No. i A 8 C 0 i e ! F

j

i Soars ae Front w«ac Kawaglahl 1 Show* Oheehl bridge

bridge levee levee -efto left slda - eigne side

j

Occurrence of Liquefaction ye* yee no yea yas 1 no

Soring

No. 1 (a) 1

i Os.20.0 Os-20.0 Os.20.0 Os-20.0 Os.20.0 Os.20.0

No. 2 (a) | Os.20.0 Ch.20.0 0s.20.0 Os-20.0

1

s.?.r.

No. 1 (polnca) i 10 10 10 10 10

i 10

(JIS saaplar)

No.(points) j 8 4 9 4

S.F.T.

No. 1 (polnca) 1 10 10 10 10 10 10

(ASTM saaplar)

No.(polnca) ; 8

, 4 8 4. - 1 .

1

Sand Saopling

1 (polQCS) , 5 5 2 ’ '

Piezocone () 0^9.7 0*20.0 |

i Os.12.

0s.20.0 Os. 18.0 Os. 10.

Automatic Ras Sounding (a) j Os-14,4 0s,20.0 I CH20.0 Os.20.0 Os.20,0 Os.20.0

Dutell Con* (a) 0^6.2 (HU. 8 l 0s*. 2 CH10.4 Os, 12.0 CK6.2

Vibratory Con* (a)

j 0s*.

1 (H20.0 ! Os.7.4 - 0*16.

Os.13.2 Os.7.2

Swadlab Sounding (a) 0s. 12.0 Os-13.0 0*L3.0 Os.20.0 Os. 13.

Os.1.0

Dynaaie Cona Penetration Taat (a) ! 0*1.19.0 0*i.20.0

j 0s. 13.

Os-19.

Os.17.2 Os.16.3

?S Logging (a) Os.21.0 0s.20.0 I Os.20.0 (H20.0 Os.20.0

Magnetic Suacepclbilicy

Logging (a) - 8.CHL3.0|

1 1.3S.1.8 i 3.4S.21.2 3.3s.20.0

Deaalcy Logging (a) Qs.20.0 3.0^13.81 1.CH20.3I 0.3S.20.3

, Os.21 . 4 3.0S.20.1

Caliper Logging (a) Os.20,8 !

1 0-.20.3 !

1 Os.20.6 0--20.9 Os.20.8

Electric Logging (a) 3.3S-20.0 8.3^14.3 I 4.3S.19.8 1 1.2S.20.5 3.4S.20.3

| 3. 3s. 19.0

Claaalflcatloe Taat (polnta) 2J 23 22 ! 26 i

i *

1 -

Cyclic Trlaxlal Teat (aarlaa) 3

1 3 ! I

j 4 ! - -

Saaonanc Colun Teat

and Cyclic torsional Taat

( sarlea) j

1 2

I 2

1

i

i

2 1

1 - -

TABLE 1
SUMMARY

OF

SPT

N160

CALCULATIONS -

SITE AHACHXRO

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TABLE 2

SHEET
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SPT

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CALCULATIONS -

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CO 1 fe 1 CM in CM 00 in o in o 2 HI a, i 9—J CM CM CM co co 9-H CM

S3 1I1

OS !1 00 CM 00 o- 00 CO CM CO w Ei* 1 • • • © o • CM O • • 2 2 a* ! rf* OS CO O CO OS CO

f—

m i CM co CO CM CM j ii s i Ei* 1 CO C- OS eo O CO CM CO 2 H a. i r**4 CO co CM CM CM CM

CO ffi i < ii

co

w ii CM CO CO CM in ‘•f O' co 00 CO *9* 00 00 CM cm in 3 O > i fH h CM 9-^ 2 O W X 1 H N M i r =«• co ii o • ii co 00 00 00 00 co CM OS CO in in e- in oo co 00 m a t eo CM CM CM CM CM CM 9—4 9-

9-H ^ 9^

a

a iii O o 0 0 0 O

O o o o O o o o o o O

O

O

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£* • i © o • o • © e • o • o © e e O e • « c • cu a i 9-

CM CO in CO o 00 OS o c—5 CM co «* in CO e» oe 05 o

Ed i ^4 ?**4 9-4 9—J 9—» m-4 9^ ’^—5 9—4 CM Q 1|

252 t

Am
SUMMARY

OF

SPT

N160

CALCULATIONS -

SITE DKAWAGISHI

-CHONIIGATA

X 11 O H a I *H c—

X CM

X

X o X o e»X e- P" CM

XX

i-i

a o o 1 CM CM <-• o o *H 9-4 *H 1-H

O 9-4

O 1-H 1-H CM CM

X

-3 2 XX 1 • X o U « * E- i o o o o o O o o 0 0 0 O O o o O O o • o

>•

X 2 X 1 A

O f- £ 1 73 11 O 73 11 Q M X O 1 *H M* 10

XX c- p- P“

X

X

X

X

X

X

X

XX

X

X 'r

-3 -3

X 1—1 1 *H ^H 1-H *H 1-H *4 i-H i-H i-H ^H 1-H 1-H 1-H *H i-H i-H

—

H

1-H 1-H

WO

X 6- 1 M > H < 1 o o o o o o o o O o o o O o o o O

O o o

Cb

a

X

X 11 o X 11 X CO o cX i-H CM

OX i-H e**

X P- o X

X o o X

X

XX 1 *H 1-H 1-H 9-4 i-H *H —H i-H i-H 1-H CM CM CM CM

*h

X 1 1 z 1111 CM e~ i0 ««r CO CM «-H

XX CMXX

X

X

X

X -<1" CM o i z 1 o X X X X X • P- p- • o I 3 1-H *H i-H 1-H *H • o 11

i“H o o o o o o o o o

73 11 • 73 1 CO CM CM *H 1-H o X

XX

X

XX

XX *9

X

X CM

EM

X 1 CM CO *9*

X

X cX

X

X o 1-H CMX *3"X

XXX o X X X 1 X E-> > 73 111 o o O o o o o o o 9—4 1-H 1-H *H *H i-H 9-4 *H *4 9-H CM O X 11 X c~ *H *3* X e- X X X X CM X X c- CM X p- 1—4 x X 1 z X 1 00 o X X X • o • X 1-H C"- O X *H X X X X X p“ I1 X fH X 9-4 *H *H i-H i-H CM CM X X > 11 a 1 Tj* V '1* Tj> X X *9 X X CM X X Q X 1 x X X X X X X X X X X X X X X X X X X X O H * 1 X Z 1 X 11

73

X I1 XX o X

X 69

XX Z M X 1 f—4 i-H *H CM X X X 111 X 111 X X X X X X CM

X

X

X 1 o z z X I X X X o X o X X •«r

l-l

X 1 *-4 *H i-H CM

X

X t1 £ X 11 CO e» c»

X

XX

X

X Z H X 1 9-4 CM CM

73

X 1 < 111

73X 11 x CM CM H*

XX CM

X *H

X CM CM *»•

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XX 3 O > 1 hH 1-4 Z O X X 1 H4 CM l-l 1 r 73 11 o • 11 CM X X O X CM ** X 'T —H X X C— p- X X X CM X

10 s 1 X CO CO COX 1-HX CM CM 'T CM CM CM CM CM CM CM

a

a 111 o o o o o o o o o o o o o O

O o o O

O o X

E* • 111 CO CO CO CO CO

X

XXXXXX

X

X

X

X

X

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X

x

a 1 9-4 CM CO

X

X c—X

X o CM

X

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X pX

X o X 1 1-H *4 i-H iH 1-H i*H ^H 9^ i-H ^H CM a 111

mill .9.gT,

TABLt 2

SHEET 4

OF
SUMMARY

OF

SPT

N16Q

CALCULATIONS -

SITE ESHOWA

BRIDGE -

NIIGATA

X

O EQ CM

a c®

a

a cm

a <—i o CM c®aOO <-• o w O o O 9-4 9—

i—p

—

9—4 7*4 9-4 9=4 9-4 v—

a CMa CM CM O' 2 a X U3 • ©

O « 9-4

H

©

OO

O©a

©© 0 0 0 © o 0 0 0 o o o

>• OS 2 a /V

A o £

03 s o 03 t Q M 03

O 9=4 co

a

a

a

aa

a

a

aaa

a

a

aaa O' O'

-3

Ja 1—

© 9*4 9*4

H 9*4 9—0 9^ 9-4 9-4 9—2 9—1 9*4 r=4 9-4 9-4 9=4 9*4 9—0 9=4

W

O os E- o • ® o © o « w >= &* < o o o o o o o o O O O o O o O o o O

O

St. o 03

a o a o»

a CO o CM

a <s=

a p*4

a o a

a

aa c®

aa

a e®

a

a CM 9-4 9—3 9-4 1—* ®"4 9—4 9-4 9—4a CM CM CM CM 9-4 -h CM

^4 83 1 2 i CO e®

a

a

a

aa o*

a lO 9=4 £-**> O' CMa c®

a

a *4 05 ! 2 <71 a o> a CM f*4 o o a a a a a a c® e~ c® c*- e® a O

9*4 <9-3 t-4 9-4 »—

9=4 o o o o o o o o o o o o

03 • 03

a

a

a

a o* o«

a

a CM i—4

O o a

a

a c® c®

aa

H

Wa CM

a o*a

a c®a

a

O

h cm

a O' O'

a

a c®

a

a o

QS

X 03 « • ©

© c • a h E-> o o o o o o o o 9—

r"4 9=4 r-4 *4 9—4 9-4 9-4 9=4 i-< CM > 03 O a CM a 9=4 e® a ^4 a i-*f e® O' C® CM a c® a a O' a a e®

to

a

©O® 2 es o» CM s®

a O' CM O' o a o o o a O'a O'

a

a l i—5 ^4 c—<• 9-4 9=4 O' a CM a a CM cm a

>*O O' o> o* o* O' O' O' O’ O' O' O' O' O' O' O' O' O' e 02 a a a a a a a a a a a a a a a a a a a a o w *

OS 2M

03a

a CM

a CM o o a c®

a

a 2 M a p—

c 9-4 p-4 •<—4 a CM a CM

'“SX

02

a CM CM CM o'

a

a O'

a

aa • • © 2 2 a C® a a o' a a CM CM M a ^4 9-4 CM a CM a Js a a a f-4 cm a O' c® a ^3 2 S- a 9=4 9^ CM CM 9H a

03

X <

03a o a

a

a CMa 9-4 CO

a 5 4 6 a O' O'a

aa o'

a 3 O > CO a 2 O a x

t-t cm

H

r 4k 03 o • a CM o a

a CMO CM o a «—* <“1 o o a

a o a CM

O

in

a CM CM 9H CM CM

a O' O’ O'

a o' o*

a

a CM CM

a CM CM CM

a

a

© • o o o o O o O O ooooooooo o o o X CO co CO CO a a a a a 3 3 3 a a a a a a a a H • • © * & a

9**} CM CO o'

a

a c*

a

a

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H

Na O’a

a c®

aa o a 9*4 9=4 94 9”*J 9=4 9-4 9-4 ^4 CM o

254

TABLE 2

SHEET
SUMMARY

OF

SPT

N160

CALCULATIONS -

SITE FSHOWA

BRIDGE -

NIIGATA 1 X 1 O E- Q 1 00 CM 3*

X p»X

X 9—3

H o O

O 1 o X

X

X

X r-3

XX

X

JZtfi

X 1 X • X • X

XXX

XX • X

O

W ^ E- 1 ® o e o • • • • ©

©© o ©©

©

© o > OS 2 W 1 A A A A XS A A A A O H Z 1 !

CO 11 O CO 1 OHMO 1 CM CO

X

XXX

x

X

XXX

X

XX

X «9* *p

-3 -3W

m 1 • t-H o-H —H 9“

9—3 9—3 9*3 ?—} 9—3 c—3 9=3 9*3

H

O

X E- 1 o M >< E-> < 1 O o o o o o O o o o o © o o ©

©©

©©

Ch CJ co

x 11 o X 11 to pX o 'P

X c* o -<r

X

X

X

X

X

X *p

X

X

X

X

X 1 X CO CM CO

XX

XXXX

—

X

X

X 1—3

X

X 1 z 11 » 11 p- 05 co o o pX PX

XX

X 9*3

XX

X 1-3

X z 1 05 to •«r CO CM t-H o o X X X X X X p- r— p- p- p- X o 111 —H •^3 t=*3 r-3 r-3 o o o o © © o © © © © o

CO 1 • CO 1 CO 05 05

X

X p- eX

XX rpXX

X r-3 r-3 o ©

X

EW

X 1 CM CO

X

X pX

X o —HX

X 3*X

X pXX

© o X X CO 1 W E-* E- 1 o o o o o o o o —*3 i—

r-3 r-3 —3

H

X

X > CO 11 o X 11 X *r o CO X X c~ x p* X X *9* X X X X X X

CD

X 1 CM z X 1 co 00 *“3

X

X o X *“3

X *3

XX

X

X 1t CM CM CO X rr X X X X ^-3 X X <*9* X 3*

>* 1 a 1 00 00 00

X

X

X 00XXXXXXXXXXXXX

a

x 1 X

XXXXXXXX

X

X

X

XXXXX

X

X

X

O

W

X 1 X z 1 Cl] 11

COX 11 X co

X CM

X •*p

X pX *p Z M X 1 CO CM X X X X r-3 X X X X 111 X 11 X o X ** X o X

X

XX 1 • 9—3 • o • X • • • z z X 1 X X X X X ©

HH

X 1 X

XX

XX

x 11 z X 11 *s< o X o pX

XX Z E- X 1 ?"3 X X X X X X

X

CO

X 1 < 11

CO

X 11 ?-3 CM *r *!•

X

XXXX

X

X

X p- p- eXX

X 3 o > 1 r-3 X X z o X X 1 l-t CM 1 Z' ** co 11 o • 11 o o X P- X X X p* X »r v—

X X x s 1 o o CM CM CM X X X X X X X X X X X 1—3 X a a 111 o O 0 0 0 O o o o o o o o o o o © © © o X 11 CO CO CO CO CO CO X X X X X X X X X X X X X X H • 1 x a 1 CM co X X X X o X X X X p- X X o X 1 *-3 —

—3 9^ 1-3

X

a 111

255

TABLE 2

SHEET
TABLE 3 SUMMARY OF CONE PENETRATION CALCULATIONS - ALL SITES

SITE DBPTB

a.

FIELD CYCLIC

STRESS RATIO

8PT

N 1 60 STRENGTH

FRICTION

RATIO

S

MINUS

NO. 200 SIBVB

1

050

at.

QC/N60

RATIO

QCl

AVG. n-0.5

TSP

CYCLIC

3TRENGTH

QCl HBTUOD

QCl

AVG. n=0 . 7 2 TSP

FSN SLEEVE

FRICTION

TSP

STATE PARAMETER

A 1.3 0.13 0.04 1.9 10 45 0.10 64 1.44

A 2.3 0.14 0.09 1.2 6 0.22 4 44 0.11 58 0.70 -0.11

A 3.3 0.18 0.32 4.8 21 0.20 2 27 0.13 33 1.35

A 4.3 0.18 0.33 1.0 10 0.26 3 60 0.17 71 1.70 -0.13

A 5.3 O'. 19 0.35 0.5 13 0.20 5 90 0.29 95 3.20 -0.18

B 1.3 0.12 0.32 1.5 12 0.28 5 121 >.5 169 1.57

B 2.3 0.15 0.10 1.2 11 0.28 12 108 0.35 139 -0.20

B 3.3 0.16 0.22 1.1 3 0.40 5 90 0.20 109 2.15 -0.16

B 4.3 0.17 0.12 1.1 5 0.20 2 20 0.06 24 0.47 -0.06

B 5.3 0.17 0.13 0.6 3 0.24 2 27 0.06 30 0.20 -0.05

B 6.3 0.18 0.12 2.3 2 19 0.05 20 0.34 -0.04

B 7.3 0.18 0.18 1.5 3 44 0.11 48 0.82 -0.09

B 8.3 0.18 0.18 1.5 6 0.22 4 65 0.16 67 0.58 -0.11

B 9.3 0.18 0.17 1.6 5 0.24 2 28 0.08 29 0.42 -0.06

B 10.3 0.18 0.16 1.0 7 0.23 3 44 0.11 43 0.38 -0.08

B 11.3 0.18 0.12 1.0 5 0.28 5 54 0.14 52 0.63 -0.08

C 3.3 0.12 0.06 1.3 3 0.28 10 59 0.13 69 0.44 -0.11

C 4.3 0.13 >.5 1.7 3 0.28 3 97 0.21 109 2.18 -0.14

C 5.3 0.14 0.23 1.9 2 0.28 4 90 0.22 98 2.36 -0.11

C 6.3 0.15 0.26 5 0.23 5 120 0.33 127 -0.14

D 1.3 0.11 0.27 0.9 3 0.32 7 120 0.30 170 2.70

D 2.3 0.14 0.21 0.3 4 0.36 10 129 0.33 169 1.64 -0.20

D 3.3 0.15 0.11 1.7 2 0.36 5 54 0.12 67 1.02 -0.12

D 4.3 0.16 0.07 1.4 2 0.40 6 45 0.10 53 0.58 -0.11

D 5.3 0.16 0.08 0.8 4 0.36 7 56 0.12 64 0.72 -0.12

D 6.3 0.16 0.12 2.2 3 0.42 5 51 0.11 56 1.69 -0.13

D 7.3 0.17 0.13 0.9 3 0.34 6 74 0.17 79 0.99 -0.13

D 8.3 0.17 0.13 1.9 12 0.18 5 46 0.16 48 0.56 -0.10

D 9.3 0.17 0.10 1.2 3 0.34 6 50 0.11 51 0.89 -0.09

B 10.3 0.15 0.13 0.6 1 0.41 6 63 0.13 63 0.31 -0.10

B 2.3 0.1 0.11 1.0 9 0.22 8 61 0.17 79 1.06 -0.13

B 3.3 0.14 NA 1.4 39 0.10 13 40 0.23 56 0.87

B 4.3 0.15 0.12 1.0 5 0.25 4 41 0.12 47 0,48 -0.10

B 5.3 0.15 0.13 1.3 2 0.29 5 55 0.13 62 0.79 -0.12

B 6.3 0.16 0.17 1.4 3 0.52 4 59 0.11 64 1.06 -0.11

B 7.3 0.16 0.15 1.4 1 0.40 3 45 0.10 52 0.79 -0.09

B 8.3 0.16 0.16 1.3 3 0.42 4 64 0.12 67 0.79 -0.12

g 9.3 0.16 0.12 1.6 3 0.40 5 50 0.11 52 0.82 -0.10

B 10.3 0.16 0.16 2.1 5 0.38 4 50 0.11 49 0.55 -0.08

B 11.3 0.16 0.11 1.6 4 0.41 4 43 0.10 41 0.53 -0.08

? 2.3 0.12 0.08 2.2 2 0.40 15 102 0.23 129 3.35

P 5.3 0.15 >.5 1.3 5 0.26 3 95 0.30 105 1.50 -0.15

P 5.3 0.15 0.28 1.3 4 0.24 5 141 >.5 150 1.70

256
TABLE 4 SUMMARY OF FACTOR OF SAFETY AND STRENGTH DATA SHEET 1 OF 3

1 SITE DEPTH

i.

PIBLD CYCLIC

STRESS

RATIO

H160

BPP

CYCLIC STRBHGTH

N160 MBTHOD

PS

HI 60

CYCLIC

STRBGTH

I4T MBTHOD

PS

IIT

QC1

TSP

CYCLIC STRBHGTH

QC1 MBTHOD

P3

QC1

VIBRO

COMB

D

X

STRBHGTH

RL20 VIBRO COMB

STATB : parahbtbr;

A 2.3 0.17 9 0.09 0.5 0.18 1.1 44 0.11 0.6 0.70 0.32 -0.11 1 A 3.3 0.18 20 0.32 1.8 0.23 1.3 27 0.13 0.7 0.30 0.34

A 4.3 0.18 23 0.33 1.8 0.23 1.3 60 0.17 0.9 0.20 0.35 -0.13 1 A 5.3 0.19 24 0.35 1.8 0.27 1.4 90 0.29 1.5 0.20 0.34 -o.i8 : A S.3 0.19 22 0.28 1.5 0.23 1.2

A 7.3 0.19 40 >.5 0.31 1.6

A 8.3 0.18 31 >.5 0.28 1.6

A 9.3 0.18 51 >.5 0.35 1.9

A 10.3 0.13 61 >.5 0.36 2.0

A 11.3 0.18 55 >.5 0.36 2.0

A 12.3 0.18 32 >.5 0.29 1.6 •

a 2.3 0.15 6 0.10 0.7 0.12 0.8 108 0.35 2.3 0.90 0.31 -0.2 !

B 3.3 0.16 20 0.22 1.4 0.17 1.1 90 0.20 1.3 0.50 0.32 -0.16 :

B 4.3 0.17 11 0.12 0.7 0.16 0.9 20 0.06 0.4 0.85 0.28 -0.06 ; B 5.3 0.18 12 0.13 0.7 0.16 0.9 27 0.06 0.3 0.80 0.27 -0.05 i B 6.3 0.18 11 0.12 0.7 0.15 0.8 19 0.05 0.3 0.80 0.25 -0.04 !

B 7.3 0.18 17 0.18 1.0 0.19 1.1 44 0.11 0.6 0.40 0.29 -0.09 ;

B 8.3 0.18 17 0.18 1.0 0.19 1.1 65 0.16 0.9 0.80 0.22 -0.11 !

B 9.3 0.18 15 0.17 0.9 0.18 1.0 28 0.08 0.4 0.70 0.22 -0.06 ! B 10.3 0.18 13 0.16 0.9 0.17 0.9 44 0.11 0.6 0.75 0.20 -0.08 !

B 11.3 0.17 11 0.12 0.7 0.15 0.9 54 0.14 0.8 0.75 0.20 -0.08 ;

B 12.3 0.17 14 0.16 0.9 0.15 0.9 0.75 0.19

B 13.3 0.17 11 0.12 0.7 0.14 0.8 0.75 0.18

C 3.3 0.12 6 0.06 0.5 0.12 1.0 59 0.13 1.1 0.30 0.32 -o.ii : C 4.3 0.13 29 ) .5 0.22 1.7 97 0.21 1.6 -0.40 0.50 -0.14 ! c 5.3 0.14 21 0.23 1.6 0.2 1.4 90 0.22 1.6 0.10 0.35 -o.u : c 6.3 0.15 22 0.26 1.7 0.22 1.5 120 0.33 2.2 0.05 0.42 -0.14 ! c 7.3 0.15 27 0.32 2.1 0.24 1.6

c 8.3 0.15 24 0.28 1.9 0.23 1.5

c 9.3 0.15 32 >.5 0.27 1.8

c 10.3 0.16 51 >.5 0.34 2.1

c 11.3 0.16 25 0.28 1.8 0.26 1.6

c 12.3 0.16 38 ) .5 0.3 1.9

c 13.3 0.16 29 ) .5 0.26 1.6

c 14.3 0.15 23 0.27 1.8 0.22 1.5

c 15.3 0.15 33 >.5 0.28 1.9

c 16.3 0.15 27 0.35 2.3 0.24 1.6

c 17.3 0.15 14 0.20 1.3 0.22 1.5

c 18.3 0.15 22 >.5 0.25 1.7

c 19.3 0.14 18 0.24 1.7 0.21 1.5

c 20.3 0.14 24 >.5 0.0
TABLE 4 SUMMARY OF FACTOR OF SAFETY AND STRENGTH DATA SHEET 2 OF 3

SITE DEPTH FIELD Ml SO CYCLIC ; PS CYCLIC PS QC1 CYCLIC PS VIBBO STRENGTH STATE CYCLIC STEBHGTH STBBGTH 3TRBMGTH COME RL20 PARAMETER

STEE3S H160 ! H 1 60 I IT IAT QC1 QC1 D VIBRO

R. RATIO BPP METHOD ! METHOD TSP MBTHOD X COMB

D 2.

0.14 13 0.21 1.5 0.17 1.2 129 0.33 2.4 1.00 0.31 -0.20

D 3.3 0.15 10 0.11 0.7 0.13 0.9 54 0.12 0.8 1.00 0.30 -0.12

D 4.3 0.16 7 0.07 0.4 0.11 0.7 45 0.10 0.6 0.90 0.28 -0.11

D 5.3 0.16 8 0.08 0.5 0.12 0.8 56 0.12 0.8 0.40 0.30 -0.12

D 5.3 0.16 11 0.12 0.8 0.13 0.8 51 0.11 0.7 0.80 0.25 -0.13

D 7.3 0.17 12 0.13 0.8 0.16 0,9 74 0.17 1.0 0.40 0.28 -0.13

D 8.3 0.17 10 0.13 0.8 0.17 1,0 46 0,16 0.9 0.30 0.30 -0.10

D 9.3 0.17 9 0.10 0.6 0.13 0.8 50 0.11 0.6 0.30 0.30 -0.09

D 10.3 0.16 11 0.13 0.8 0.12 0.8 63 0.13 0.8 0.10 0.38 -0.10

D 11.3 0.16 7 0.07 0.4 0.13 0.8 0.40 0.26

D 12.3 0.16 9 0.10 0.6 0.14 0.9 0.20 0,33

D 13.3 0.16 7 0.07 0.4 0.11 0.7 0.40 0.26

D 14.3 0.16 10 0.13 0.8 0,15 0.9 0,10 0.39

D 15.3 0.16 15 0.17 1.1 0.18 1,1 0.00 0.45

D 16.3 0.15 15 0.17 1.1 0.17 1.1 0.10 0.40

D 17.3 0.15 20 0.22 1.5 0.20 1.3

D 18.3 0.15 20 0.28 1.9 0.20 1.3

D 19.3 0.15 29 >.5 0.25 1.7

D 20.3 0.14 26 0.39 2.8 0.22 1.6

8 2.3 0.13 8 0.11 0.8 0.18 1.4 61 0.17 1.3 1.00 0.31 -0.13

8 4.3 0.15 10 0.12 0.8 0,16 l.l 41 0.12 0.8 1.00 0.27 -0.10

8 5.3 0.16 12 0.13 0.8 0.16 1.0 55 0.13 0.8 0.95 0.25 -0.12

8 6.3 0.16 16 0.17 1.1 0.14 0.9 59 0.11 0.7 0.90 0.24 -0.11

8 7.3 0.16 14 0.15 0.9 0.15 0,9 45 0.10 0.6 0.90 0.22 -0.09

B 8.3 0.16 15 0.16 1.0 0.15 0.9 64 0.12 0.8 0.90 0.19 -0.12

8 9.3 0.16 11 0.12 0.8 0.13 0.8 50 0.11 0.7 0.80 0.21 -0.10

B 10.3 0.16 15 0.16 1.0 0.15 0.9 50 0.11 0.7 0.75 0.20 -0.08

8 11.3 0.16 10 0.11 0.7 0.13 0.8 43 0.10 0.6 0.70 0.20 -0.08

B 12.3 0.16 9 0.10 0.6 0,12 0.8 0.60 0.21

8 13.3 0.16 35 >.5 0.26 1.6 0.70 0.28

8 14.3 0.15 25 0.32 2.1 0.17 1.1 0.40 0.26

8 15.3 0.15 23 0.27 1.8 0.22 1.5

8 16.3 0.15 27 0.36 2.4 0.23 1.5

B 17.3 0.15 29 0.40 2.7 0.24 1.6

B 18.3 0.15 18 0.20 1.3 0.18 1.2

8 19.3 0.14 19 0.21 1.5 0.21 1.5

8 20.3 0.14 27 0.40 2.9 0.25 1.8

258
TABLE 4 SUMMARY OF FACTOR OF SAFETY AND STRENGTH DATA SHEET 3 OF 3

! SITB DEPTH FIELD N160 CYCLIC PS CYCLIC PS QC1 CYCLIC PS VIBRO STRENGTH STATE CYCLIC STRBNGTH STREGTH STRENGTH CONE RL20 PARAHBTBR

STRESS Ml 60 N160 I4T I4T QC1 QCl D VIBRO

1 . RATIO BPP METHOD METHOD TSP MBTHOD X CONE

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! P 14.3 0.15 28 0.37 2.5 0.22 1.5

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259
Figure I Location of test sites and the

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FIGURE 21
U.Sc-Japan Coordinated Earthquake Research Program on

Masonry Buildings, Seismic Behavior of Three-Story

Full Scale Block Planar Frame Under Cyclic Lateral Loading

ABSTRACT

by

Shin Okamoto

Masaomi

1 O ?!« T ' 1

, Yutaka Yamazaki , Takashi Kaminosono

Teshigawara

'

4 and Hisahiro Hiraishi D

In order to know the total behavior of structure

which consists of reinforced masonry (RM) walls, RM beams, and reinforced concrete (RC) slabs, one reinforced clay block masonry specimen was tested

under cyclic lateral loading. The specimen was a three story and two bay frame structure which was

a part of a five story RM apartment house

building. Loading applied to the specimen was

planned so as to reproduce the same stress

condition as that of the prototype apartment

house building structure under the earthquake

motion. Main purpose of this test is to get basic

data for the design guideline of RM buildings and

for a full-scale five story RM building structure

which will be tested in 1987 at Building Research

Institute, Tsukuba, JAPAN.

The objectives of this test, design procedure of

the test specimen, loading method, and test

results are presented and discussed in this

paper. Conclusions obtained from this planar

frame test are as follows;

1) The specimen showed larger shear capacity than

that requested by the tentative design

gui deline.

2) The beams with short span should be designed

to prevent from shear failure at small

deflection by decreasing those depth or amount

of reinforcement.

3) Joint steel arranged in the beam showed the

effect on preventing the faceshell from

falling down.

4) The specimen was still elastic and had only

minute flexural cracks within the temporary

shear stress level.

5) Short beam was forced into 1.6 to 2.0 times

drift angle as long as story drift angle by

long wall (Wall -A).

6) Transverse wall was effective totally on the

strength at the maximum shear stress level.

KEY WORDS; Reinforced clay block masonry,

Walls, Beam, Transverse wall,

Reinforced concrete slab,

Full scale planar Frame,

Static lateral loading test

INTRODUCTION

U.S. -Japan Coordinated Research Program on Reinforced Masonry(RM) Building has been started

since 1984 . The targets of Japanese side are not only to know behaviors of masonry elements

and RM structures, but also to establish a design guideline of medium-rise RM buildings up to five stories. Structural elements of

building such as walls, beams, and wall-beam

subassemblages had been tested in 1985 /_ ' and

many useful informations concerning the

structural behavior of elements were obtained. On the basis of those test results, a five story

apartment house building was designed as a prototype building structure 7

). In order to know

the total behavior of structure which consists

of RM walls, RM beams, and reinforced concrete

(RC) slabs, one reinforced clay block masonry

specimen was tested in 1986 at Building Research

Institute(BRI) , Tsukuba, Japan. The specimen was

a three story and two bay frame structure which

was a part of a five story RM apartment house

building. Loading applied to the specimen was planned so as to reproduce the same stress

condition as that of the prototype apartment

house building structure under the earthquake

motion

.

OBJECTIVES

Test on the three story reinforced clay block

masonry frame was planned on the basis of the

test results of RM walls, RM beams, and RM wallbeam subassemblages. Main objective of this

test is to know the structural behaviors of a full-scale RM planar frame. And to know behaviors

of structural elements which compose the test

frame is also important objective. These elements

are listed as follows;

1) wide wall with transverse wall

2) narrow wall

3) outer wall with varying axial load

4) short beam which depth or reinforcement is

decreased

5) lon-g beam

6) wall-beam joint panel

7) small openings in wall and wall-beam joint

panel

8) special reinf orcement(spiral reinforcement,

joint steel)

Construction technique was also checked while the

specimen was under construction. The test results

*1 Director,

Research Planning and Information Department,

Building Research Institute (BRI),

Ministry of Construction (MOC), JAPAN

Tatehara 1, Ohho, Tsukuba, Ibaraki 305, JAPAN

*2 Head, Large Scale Structure Testing Division,

Production Department, BRI, MOC, JAPAN

*3 Senior Research Engineer,

Large Scale Structure Testing Division

Production Department, BRI, MOC, JAPAN

*4 Research Engineer,

Large Scale Structure Testing Division

Production Department, BRI, MOC, JAPAN

*5 Head, StructureDivision,

Structural Engineering Department, BRI, MOC,

JAPAN

281
are considered to be helpful to establish the

design guideline of buildings and to make a plan

of the test of a full scale five story RM

building structure which will be tested in 1987 at BRI.

SPECIMEN

Prototype Structure

A prototype structure of a five story apartment

house building was designed using concrete block

units in accordance with the tentative design

guideline of RM buildings. The specimen described

in this paper was a part of this prototype

structure. Plan of the prototype structure is

illustrated in Fig.l. Thickness of wall is

19cm(7.5inches). Ratios of total wall length to floor area of the structure are 15.2cm/m 2 (0.556inch/ft 2

) in longitudinal direction

(loading direction), and 18.4cm/m 2 (0.637

inch/ft 2

) in transverse direction. A part of

the Frame-Y3 was selected as the specimen,

because the Frame-Y3 had various types of wall

and beam. The tested portion is also shown in

Fig.l.

Details of Specimen

Figure 2 shows plan, elevation, and reinforcement

details of the specimen. The specimen constructed

by reinforced clay block masonry was a three

story structure and 2.8m (9.2feet) in each story

height. There were three walls, those were 1)

wide wall (Wall-A), 2) narrow wall (Wall-B), 3)

outer wall (Wall-C). There were two spans. One

span between Wall-A and Wall-B was short, and the

other span between Wall-B and Wall-C was long.

Therefore, the specimen had short beam (Eeam-A)

and long beam (Beam-B) with 2m(6.6ft) wide RC

slab at each of the first and second stories. All

of walls and beams were 19cm (7.5inches) in

thickness. The specimen had a RC beam for loading

at the top floor level. But dimensions and

reinforcement details of the specimen was

slightly different from those of the prototype

structure, because of differences in the modulus

of masonry units and some modification in details

of the short beams. Walls and beams were as follows

1) Wall-A was 3.8m(12.5ft) long and with

2m(6.6ft) transverse wall at the center of the

wal 1

2) Wall-B was l.lm(3.6ft) long and with small

openings

3) Wall-C was l.lm(3.6ft) long and with 2m(6.6ft)

transverse wall at one edge of the wall.

4) Short beam(Beam-A) at the 1st story was

1 .0m( 3.3ft) span, 0.8m(2.6ft) depth, and with

small flexural reinforcement. Short beam at

the 2nd story 1.0m(3.3ft) span, 0.5m(1.64ft)

depth, and with small flexural reinforcement.

5) Long beams(Beam-B) were 1.9m(6.2ft) span, and

0.8m( 2 . 6 f t) depth.

Small openings for piping were arranged in Wall-B

and in joint panels of Wall-A and(Beam-A). Spiral

reinforcing bars were placed around the flexural

reinforcing bars at the bottom of the 1st story

wall, where the flexural reinforcing bars were

spliced, in order to prevent buckling and bondslippage of reinforcing bars in early stage.

Material Properties

Concrete with small size aggregate which diameter was smaller than 10mm(0.4inch) was casted into

walls and beams in order to get a good

compactibility. Super plasticizer was also used

for getting good compactibility. The value of

slump became 25cm from 18cm by adding super

plasticizer. Concrete with normal size aggregate,

that was 25mm(linch) was casted into slabs and

loading beam. Prism compressive strength and

compressive strength of slab concrete are

tabulated in table 1 and 2. Six kinds of deformed

reinforcing bar were used in the specimen. Those were 10mm diameter for slab, 13mm for shear

reinforcement in walls and beams, 16mm for

vertical reinforcement in walls and horizontal

reinforcement in beams, 19mm for flexural

reinforcement in walls and beams, and 22mm and

25mm for foundation and loading beam. Mechanical

properties of reinforcing bars are listed in

table 3.

DESIGN OF EXPERIMENT

Loading Method

Test setup is shown in Fig. 3. Horizontal load was applied to the top of the Wall-A by four

actuators. In pushing d i r e c t i o n ( n eg a t i v

loading), horizontal load would be distributed by

loading beam, and in pulling direction(positive

loading), that would be done by PC bars which

were set in the loading block. This specimen was

the model of lower three story parts of five

story building. Vertical load were applied to the

top of each wall by eight center hole jacks in

order to adjust vertical stress at the bottom

end of walls to that of walls in the five story

building. Weight per unit floor area was assumed

1.2 ton per square me t ers( 2261 b / f

2

). Wall-A

should be considered to locate the place where

boundary beams were attached at the both ends. So

vertical loading was controlled so that the

vertical stress at the bottom of Wall-A was

constant. Wall-B had also boundary beams at both

sides. For this reason, the applying vertical

load to Wall-B was kept constant. Outer wall

(Wall-C) had boundary beams at its one side, so, varying axial load due to shear force of boundary

beams would act on the Wall-C. The range of

varying load tp apply during this test was determined on the ultimate strength of upper

beams which were supposed to be connected to the

Wall-C in five story building. Overturning

moments were adjusted to that of five story

building subjected to external forces with Ai

distribution. Figure 4 shows the horizontal

loading hysteresis. Up to the mean shear stress

of 10kg/cm 2

( 142.2psi) which was the value divided

total horizontal load by wall area in loading

direction; i.e. ,11, 343cm 2

, experiment was

controlled by the value of horizontal load. And

after the drift angle of 2nd floor to basement

exceeded 1 .Ox 1

- ^ r a d i a n , experiment was

controlled by the value of this drift angle of

2nd floor to basement.

Measuremen t Horizontal and vertical load were measured by

load cells which were attached to actuators and

centerhole jacks. Horizontal and vertical

282
displacement were measured by strain gage type

transducers which accuracy was from BO^strain/mm

to 1000^ strain/mm. Strain of reinforcements and

concrete surface were measured by strain gages.

TEST RESULTS

Figure 5 and 6 show the crack pattern and the

yielding stage of reinforcing bar at several

shear stress levels or deformation levels

respectively. In those figures deformation was described as the drift angle of 2nd floor to basement. Development of crack and yielding

stage of reinforcing bars were as follows;

1) There were no cracks in the specimen up to the shear stress of 4kg/cm^(56.9psi) that was considered as temporary design stress.

2) Up to the shear stress of 6kg/cm“(85.3psi),

flexural cracks occurred at the bottom of IF and 2F Wall-A and at the end of 2F and 3F

beams. Reinforcing bars at the 3F short beam

and 2F, 3F slab close to Wall-A reached its

yielding strain.

3) Flexural cracks of all beams and 3F Wall-C

occurred till the shear stress of

8kg/cm^( 1 13 .8psi)

.

4) Reinforcing bars at the bottom of Wall-A and

short beam except for 2F lower part reached

its yielding strain till the shear stress of

10kg/cm^( 142.2psi)

.

5) Shear cracks were observed at the 3F part of

Wall-C and at the 2F short beam until the

shear stress reached 1 2kg/cm^(l 70.7psi). Up to this shear stress level, reinforcing bars at all beams except for 2F upper part of long

beam reached those yielding strain.

Reinforcing bars of Wall-A reached those

yielding strain up to the bottom of 2nd floor

slab. Reinforcing bars at the bottom of each

story close to long beam in Wall-B and at the

exterior part in Wall-C also yielded.

6) Until the drift angle of 2.5x10“^ radian,

shear cracks at the IF part of Wall-A, 3F

short beam, at the 2F and the 3F part of long

beam and at the joint panel zone occurred.

7) Until the drift angle of 5.0x10“ radian, the

3F part of Wall-B failed in shear and vertical

crack at the 3F and the 2F parts of Wall-C and

the IF part of Wall-A occurred along the

flexural reinforcing bars. The reason of shear

failure in Wall-B is considered that loading

beam attached to the top of Wall-B restrained

the rotation of Wall-B. Maximum horizontal

load that was 176ton ( 1 5.5kg/cm =220.5psi) was recorded at this drift angle. On the

yielding distribution of reinforcing bars,

yielding hinges were made at the places that were the bottom of all walls, the end of all

beams and both ends of the 2F part of Wall-B

in negative direction. But in positive

direction, yielding hinges were made at the

both end of the 2F and the 3F part of

Wall-C instead of at the long beams end close

to Wall-C.

_ T 8) When the drift angle reached almost 10.0x10

radian, Wall-A failed in compression at its

bottom part and 2F short beam failed in shear.

9) Over the drift angle of 10. Ox 10“ J rad ian and

the faceshell or grout concrete in the 3F part

of Wall-C, the 2F short beam, and the

compressive failure zone of Wall-A fell down.

But it was observed that joint steel arranged

in half part of the 2F short beam prevented

those faceshells from falling down.

The relationship of first story horizontal drift

of Wall-A vs. horizontal load is shown in Fig. 7. Until the mean shear stress was under

lOkg/cm^C 142.2psi) , the hysteresis loop showed

almost such as elastic that had small hysteresis area and pointed to the same restoration force

recorded before. From the drift angle of 2.5x10“

^radian the restoration force deteriorated with

every repeated loading at the same drift angle.

Maximum force was recorded at the drift angle of

5. Ox 10“^radian. As above mentioned, from the

drift angle of 5. Ox 10“^radian to that of 10x10“

~radian the 2F beam with short span failed in

shear and Wall-A failed in compression at its

critical section, however, the restoration force

of specimen kept the mean shear stress over

12kg/cm^(170.7psi). This shear stress was almost

the same level that was requested by the design

guideline. In Fig. 7, the analytical results of

both static and dynamic are indicated. 7)

DISCUSSION OF TEST RESULTS

Horizontal Stiffness

Change of the story horizontal stiffness with

increase of the first story drift angle is shown

in Fig. 8. The stiffness obtained from elastic

analysis is also indicated in Fig. 8. The

stiffness obtained from elastic analysis agrees

with those obtained from the test at the mean

shear stress of 4kg/cm^(56.9psi) which value was considered as temporary shear stress. Therefore,

the specimen was still elastic within the

temporary design force. In the case of the first

and second stories, the decrease of the stiffness

changes at the shear stress of 12

kg/cm^( 170.7psi). The reason is considered that

the specimen yielded totally at this shear stress

level. The decrease of the stiffness also

changes, at the first story drift angle of 5x10

^radian when the specimen indicated the maximum

shear stress. The decrease of the stiffness of

the third story started faster then those of the

first and second stories, because cracks of

loading beam occurred in early stage at slits of

slabs

.

Equivalent Damping Ratio

Change of the equivalent damping ratio obtained

from the hysteresis loop of the specimen with

increase of the first story drift angle is shown

in Fig. 9. The equivalent damping ratio was approximately in the range of 2% to 4% at the

shear stress of 10kg/cm z when the first story

angle was 0.73xl0~^radian. The equivalent damping

ratio increased with increase of the first story

drift angle. At the first story drift angle of

5x10“^ radian when the specimen indicated the

maximum strength, the equivalent damping ratio

increased up to the range of 10% to 25%.

Drift Angle of Beams

Drift angle of Beam-B, long beam between Wall-B

and Wall-C, was almost equal to the story drift

angle. On the other hand, drift angle of Beam-A,

short beam between Wall-A and Wall-B, was 1.6 to

283
2.0 times as larger as the story drift angle. It is considered that Beam-A was forced into large

deformation by the long wall(Wall-A).

Behavior of Wall-A

Figure 10 shows the deformation mode of Wall-A at several loading stages. It is evident that the

deformation mode of Wall-A was flexural. Until

the first story drift angle reached 5x 10"^radian,

Wall-A deformed only with the rotation at its

bottom. The specimen reached its maximum strength

when the first story drift angle was about 5 x 10”

radian, and at that time third story of Wall-B

failed in shear. After reaching maximum strength,

the deformation mode of Wall-A changed; i.e., the

Wall-A deformed in the third story more than in

the other stories. In order to study the

components of deformation, the strain

distribution of reinforcing bar, (its location is

shown in Fig. 11(b)) and the curvature

distribution of Wall-A are shown in Fig. 11(d).

The curvature at the lowest part of wall was very

large. But those of other parts were very small.

Tensile strain of the reinforcing bar was also

very large, i.e., 1.5% at the first story drift

angle of 1.2x10”^ radian ( 1 2k g/c m 1 70.7 psi ). The curvature at the lowest part of wall was

15x10 - k(l/mm) and 50x10”-^ (1/mm) at the drift

angle of 2. 5x10”-^ radian and 5. Ox 1 0"

r a d i a

respectively. The relationship of the rotation

angle due to this curvature vs. story drift angle

is shown in Fig. 11(c). Sixty percent of drift

angle was represented by the rotation angle of

the lowest part of wall at the drift angle of

5.0xl0“^radian. This is match to the fact that

the strain of reinforcing bar was not so large in

upper part of wall. Because the compressive

strain capacity of brick masonry was too small to let the tension bar be strain hardening range.

It is seemed that the pulling out of reinforcing

bar occurred at the first story drift angle of

1.6x10”^ radian. The lap joint was made at the

critical section and this lap joint was

reinforced by spiral bars, and the fixing length

to foundation was 1 m. It is considered that

reinforcing bar was pull out from foundation

because of better fix characteristic in lap joint

part. After pulling out of reinforcing bar, (A-3

in Fig. 11(b)), the strain of reinforcing bars in

inner part of wall and at the top edge of lap

joint became larger.

Effect of Transverse Wall

Figure 11(a) shows the strain distribution of

reinforcing bars. The strain was very large at

the critical, section. The reinforcing bar set in

the edge of wall did not work so effectively up

to the drift angle of 1.2x10’

radian( 12kg/cm =1 70.7psi ) . This reinforcing bar.

however, worked after the drift angle of 2.5x10”

radian. It is observed that transverse wall of T- shape was effective totallv from the drift angle

of 0.73x10” radian( lOkg/cm^), and the transverse

wall of Cross-type did after 1 . 2 x 1 0

” J radian(12kg/cm^).

Analysis

On next three cases, maximum horizontal strength

of this specimen was calculated on condition that

the same shear force act on the every story of

each three walls.

Case A: ignoring the effect of axial force of

beams

Case B: considering the effect of axial force of

beams

Case C: add the case B to the effect of hardening

of beam reinforcing bars.

The effect of hardening was considered by the way

that yielding strength of reinforcing bar was assumed to be 1.2 times as much as its natural

yielding strength. In case B and Case C, it was supposed that axial load of beam acted only on loading beam(RF beam). The reason why the

hardening of only beam reinforcing bars was considered in the case C was based on the fact

that the test result did not show a large value

of strain of the wall reinforcing bars(Fig. 11(b)) as above mentioned. Results of analysis

tabulates in Table 4. The results of analysis

showed the smaller value than that of test

results. The strength obtained by case C was smaller than test results by 40ton(23%) in

pulling direction and 30 ton(17%) in pushing

direction respectively. The shear failure of

third story of Wall-B is considered to be one of

the main reasons of the difference between

analysis value and test results. The shear force

carrying capacity of Wall-B was 29.6 ton. But

shear strength capacity of the third story of

Wall-B will be 52ton2). If the third story of

Wall-B would carry this shear f orce(52ton), the

maximum strength by case C would be over 156 ton in both direction, and the difference between

test results and analysis(Case—C) would be within

ten percent. More detailed and accurate analysis

on ultimate strength of this test will be

necessary, but this hand calculation shows

followings

1) The influence of axial load of beam due to the

loading method did not observed on the total

ultimate strength. But in order to know the

influence to each member, more accurate

analysis should be done.

2) It should be needed to evaluate the effect of

failure of the third story of Wall-B on total

strength capacity of the specimen.

CONCLUSIONS

Conclusions of the reinforced clay block three

story planar frame test are as follows;

1) Shear stress at the drift angle of 10x10”-'

radian was over the value that was requested

by the tentative design guideline.

2) The beams with short span may be designed to prevent from shear failure at the small

deflection range by decreasing those depth or amount of reinforcement.

3) Joint steel arranged in the beam with short

span had the effect on preventing the

faceshell from falling down.

4) The specimen was still elastic and had only

minute flexural cracks within the temporary

shear stress level.

5) Equivalents damping ratio increased

approximately from three percent at low shear

stress level to more than ten percent at maximum shear stress.

6) Short beam was forced into 1.6 to 2.0 times

drift angle as long as story drift angle by

284
long wall (Wall-A).

7) In large deformation range, story drift angle

was mainly represented by the rotation at the

lowest part of wall,

8) Transverse wall was effective totally on the strength at the maximum shear stress

* level

.

9) It should be needed to evaluate the effect of

failure of the third story of Wall-B on total

strength capacity of the specimen.

ACKNOWLEDGMENT

In carrying out the experiment described in this

paper, the authors are pleased to acknowledge the

considerable assistance of Messrs.

Kouji Saitoh: Nihon Renga Seizou Co. Ltd.,

Masanori Yokoshi: Matsumoto Renga Co. Ltd.,

Kiyoshi Ichijoh: Hokkaido Nozai Kougyo Co. Ltd.,

Tadao Sankawa: Nishitani Tougyo Co. Ltd.,

Masashi Yamamoto: Tobishima Corp. Co. Ltd.,

Hiroto Katoh, Hiroshi Isoishi: Building Research

Insti tute

,

Akio Nakaoka, Takaaki Nishi: Hasegawa Komuten

Co . Ltd .

,

Mototaka Matsuno: Hazamagumi Co. Ltd.,

Mitsuyasu Itou, Tatsunori Ryoh, Nobuhiko Yokoi,

Nobuo Yokoyama: Kunishiro Taika Kougyo Co. Ltd.,

and

Mr. Kyle Woodward: UCSD, USA

The authors also would like to express their

appreciation to TECCMAR, PROCMAR, and BLDCMAR

members for their contribution to coordinate and

to promote this U.S. -Japan research program.

REFERENCES

[1] S.Okamoto and J. Noland, "U.S. -Japan

Coordinated Program on Masonry Research",

Proceedings of the Third Conference on Dynamic Response of Structures, EM Div.,

ASCE, 1986, pp. 55-70

[2] T.Okada, H.Hiraishi, T.kaminosono and

M.Teshigawara , "Flexural Behavior of

Reinforced Masonry Walls and Beams",

Proceedings of the Third Conference on Dynamic Response of Structures, EM Div.,

ASCE, 1986, pp. 481-488

[3] T.kaminosono, H. Isoishi, Y.Yamaguchi and

R.Kawai, "Seismic Capacity of Reinforced

Masonry Walls Including Effects of Axial

Stress", Proceedings of the 4th Canadian

Masonry Symposium, 1986, pp. 163-173

[4] M.Fu-jisawa, T.Kawashima and Y.Yamaguchi,

"Seismic Capacity of Reinforced Masonry Walls

Effects of Shear Span Ratio", Proceeding of

the 4th Canadian Masonry Symposium, 1986,

p'p .377-395

[5] M.Teshigawara, H. Isoishi and M. Yamamoto,

"Seismic Behavior of Masonry Beams",

Proceedings of the 4th Canadian Masonry

Symposium, 1986, pp. 759-773

[6] M.Teshigawara, H. Isoishi, A. Nakaoka and

T.Terada, "Erect of Transverse Wall Attached

to Reinforced Concrete Masonry Wall",

Proceedings of the 4th Canadian Masonry

Symposium, 1986, pp. 774-782

[7] Y.Yamazaki and M.Teshigawara, "Earthquake

Response of Five Story Reinforced Concrete

Masonry Test Building," Proceedings of the

Third Conference on Dynamic Response of

Structures, EM Div., ASCE, 1986, pp. 55-70

285
Table 1. Prism Compressive Strength

Story Strength[kg/cm]

1 F 273

2 F 272

3 F 231

Young mpdulas is evaluated by Ea.l).

E=2.1xl05 x/F'm/200 --- Eq.l)

E=239ton/cm 2

Table 2. Slab Concrete Compressive Strength

Story Strength[kg/cm ] 2 F 327“

3 F 314

R F 319

Table 3. Mechanical Properties of Reinforcing Bars

Bar

[mm]

Material

[SD]

Area

[era 2

]

Yield Stress

[kg/mm2

] 10 30 0.71 39

13 30 1.27 37

16 35 1.99 39

19 35 2.87 40

22 35. 3.87 41

25 35 5.07 40

Table 4. Results of Preliminary Analysis

Case Load Wall [ ton

2

A B C [ ton ] A + 81 23 10 114

- 74 31 24 129

B + 81 25 11 117

- 78 35 26 139

C

4* 103 37 15 155

- 76 49 31 155

286
~

1

3.390

S00

.

4.20U-

,

4

500

95 5.400 2.600 5.400

X 1 *1

(Scale in mm)

Fig.1. Plan of Prototype Building

Fig. 2 b ) Cl ay Brick Unit

30 m

Fig. 2 a) PI an , El evati on , and Reinforcement

of Test Specimens

3505 r MS

- 2000 - -2150- 3250

1

1 595, ,545 1545| r

795 |-, 295

Fig. 3. Test Setup

287

I

790 I

i

500
t tkt/ciji')

"t : Mean shear stress i.e. Q/sectional wall area in loading direction

Rl: The first story drift angle of Wall-A.

Fig. 4. Horizontal Loading Hysteresis

h

( p

L

-

c »«

!/' /

\ '] Nv( ..

X =12 , (kg/cm*)

L.

i - i •ApaktL -s

V "

a

— *v

R, =5*10“*

: s

. .j

' j

’SVl

(

pam

/

*

-!fc—« r< V E brV^KV

\ _

7^ -V J

Rt =10*10“*

(radian)

Fig. 5. Crack Pattern

288
T = 4 (kg/cms

) = 6

= 8

* = 1 0

= 12

R i = 1/200

= I / I 00

> 1/100

UNYIELD DISC0NECTI0N

Fig. 6. Yielding Stage of Reinforcing Bars

• Result of Earthquake Response Analysis

[ Spring for sway was 2000 ton/cm.

Ground Motion was "Hachinohe EW

1 '

which max acceleration was 450 Gal.

*l:mean shear stress at the ultimate

stage obtain by static farame analysis

*2:mean shear stress of the first story

when base shear corresponded to 1G. *3 : wall 1 ength ratio

Fig. 7. First Story Drift of Wall-A V.S. Lateral Force Relationship

289
Stiffness

(ton/cm>

Line shows the first cycle of each loading stage

Fig, 8, Change of Story Stiffness Fig. 9. Change of Equivalent Damping Ratio

290
=12

Kg/cm'

t

=10

kg/cm

BAR BAR

C-2 C-1 A-2 A-l

v->

WALL-C ° WALL-A

O Location of spiral reinforcement

Fig. 11 b) Notation of Reinforcing Bars

Ro/R

o o o

o o

o

o •

•

®

• •

• •

10 '" 10' 5 10' 2 Drift Angte (rad.) o flexural deformation during story hight

• Rotation Angle at the lowest part

Fig. 11 a) Strain Distribution of Reinforcing

Bars in Transeverse Wall

Fig. 11 c) Ratio of Flexural Drift

to Story Drift

LOAD

( 1 /mm)

2 xlO fc 1000;/

HVsIl

1!l()J i;o DISTRIBUTION

( I /mm

)

r>x!o* mono//

10.7)

DIS TRIBUTION

Fig. 11 d) Strain of Reinforcing Bars and Curvature Distribution (Wall-A)

9

291
itm

.

> spy i iihjhbi -

.
U.S. Coordinated Program for Masonry Building

Research— 1986/87

by

James L. Noland 1

ABSTRACT

The U.S. Coordinated Program for Masonry

Building Research is an integrated set

of twenty-five specific research tasks

which address topics ranging from material properties to system behavior to

design methodology. Work began in late

1985 and is currently in progress. Two

new research tasks have been added to

the program and the schedule has been

slightly revised.

KEYWORDS: Masonry; Research; Tests;

Structure; Design; Analysis

1. INTRODUCTION

The U.S. Coordinated Program for Masonry

Building Research is the U.S. part of

the third in a series of joint U.S.-

Japan research projects conducted under

the auspices of the UJNR Panel on Wind

and Seismic Effects. This report is a brief review of the work and activities

of the U.S. side during the period of

May 1986 to May 1987.

2. OVERVIEW OF U.S. PROGRAM

A general description of the U.S. Coordinated Program for Masonry Building

Research was presented at the UJNR meeting of May 1986. The essential ingredients of the program are repeated here

for convenience.

2.1 Basic Concept

The U.S. program currently consists of

twenty-five separate, but coordinated

research tasks. Two additional research

tasks were initiated in the period

covered by this report, i.e., response

of two-story reinforced concrete masonry

walls to in-plane fully reversed lateral

loads and shake table testing of threestory one-quarter scale reinforced

concrete masonry buildings. A complete

list of research tasks is presented in

Table 1.

3.0 SCHEDULE

Task 1.1 was initiated in February 1985

and completed in August 1985 as a precursor to the majority of tasks which

began in the fall of 1985.

The program schedule as of March 1987 is

shown in Figure 1. It has not yet been

revised to include new Task 3.1(c).

Work on that task will begin in June

1987 and last for approximately two

years

.

Schedule revisions include:

1) a slight extension of Task

3.1(a)

2) a new start-date for Task

3.2(b)

3) a slight extension of Task

3.2(a)

4) a slight reduction of time for

Task 3.2 (bl)

5) an advanced start-date and extended end-date for

Task 4.2

6) an approximately 7 month extension of Tasks 5.1 and 5.2

7) a slight extension of Task 6.2

8) inclusion of new Task 7.1

9) a six month extension of Task

8 .

1

10) a six month postponement of

both start and end dates of

Task 8 .

1

11) a nine-month advance of the

start date of the second phase

of Task 9.1

12) a six-month postponement of

Task 9.3

13) a six-month advance and a sixmonth extension of Task 9.4

14) a two-year advance and a sixmonth extension of Task 10.1

15) a six-month extension of Task

li.l

Note that the U.S. program now extends

to January 1992.

4.0 ORGANIZATIONAL CHANGES

Gerald Dalrymple has replaced Alan • Yorkdale on the Industry Observers and

Industry Participation Panels. Mr.

Dalrymple is a senior Engineer with the

Brick Institute of America.

Dr. Charles Scribner of the NBS has replaced Dr. Leyendecker as liaison with

UJNR.

1 Atkinson-Noland & Associates, Inc.,

2619 Spruce S t .

,

292
5.0

MEETINGS

In addition to numerous meetings within

the U.S. of various subgroups of TCCMAR/U.S. the following major meetings

have been held:

1) May 1986 - Hart, Kariotis, Ewing,

Hamid and Noland met with Okamoto,

Yamzaki, Teshigawara, Kaminosono, Baba

and Senbu at BRI to review progress,

data and experimental procedures.

2) July 1986 - Regular meeting of

TCCMAR/U . S

3) September 1986 - Special meeting of

TCCMAR/U. S

4) September 1986 - Second Joint U.S.-

Japan Technical Meeting (JTCCMAR)

5) February 1987 - Regular meeting of

TCCMAR/U. S

grammetry. Those measurements will augment those obtained from strain gages

and LVDTs

7.2

Electronic Mail

Electronic mail has been used on a trial

basis by several researchers to exchange

messages. The system uses a VAX computer at the University of Illinois

(courtesy of Dr. M. Sozen) as the "post

office". This form of communication

will be used by all of TCCMAR/U. S. in

the near future.

8.0

RESEARCHERS EXCHANGE

Exchange of researchers between Japan

and the United States is an important

component of the joint research which

fosters communication and understanding.

The next meeting of TCCMAR/U. S. will be

in August 1987 and the next JTCCMAR

meeting will be in October 1987 in

J apan

6. PROGRESS

The approximate percentages of completion of the individual research tasks as

of May 31, 1987 are as follows:

Task Percent Complete

1.1 100

1.2(a) 40

I. 2(b) 30

2.1 30

2.2 30

2.3 10

2.4 -0-

3.2(a) 35

3.2(b) 15

2.1(c) -0-

3.2(a) 45

3 . 2 ( b 1 ) 10

3 . 2 ( b2 ) 30

4.2 100

1/5.2 60

6.2 80

7.1 10

8.1 70

8.2 - 0

- 9.1 50

9.2 20

9.3 -0-

9.4 -0-

10.1 - 0

- II. 1 30

7.0 MISCELLANEOUS

7.1 Photogramme try

Dr. M. Teshigawara of BRI was the first

exchange researcher. He spent approximately four months in the U.S. residing

in Boulder, Colorado. In addition to

monitoring the work at the University of

Colorado, he visited several masonryrelated organizations throughout the

U. S . .

Dr. K. Woodward of the Universi ty of

California- San Diego spent two weeks in

June 1986 at BRI. He was primarily interested in laboratory equipment and

procedures used at BRI for large specimen tests.

Dr. O. Senbu of BRI spent one month in

the U.S. in August-September 1986 observing the research work at the University of Colorado, participating in the

Second JTCCMAR meeting in Keystone,

Colorado and later visiting the researchers at the University of

California in San Diego.

Dr. F. Seible of UCSD has applied for a Japanese government research grant for

foreign researchers. If successful, he

plans to be in residence at BRI during

the last half of 1987.

9.0 CONCLUSIONS

The U.S. program is progressing well

after a fitful start which unfortunately

was well after the program in Japan

started. Now that data is becoming

available from the U.S. research, more

frequent meetings of the U.S. researchers are planned and comparisons

with Japanese results will be possible.

Displacements at several times in the

load histories of in-plane experimental

specimens, i.e., shear walls and floor

diaphragms are being obtained by photo293
JAN

as

JUNE AUG JAN

*85 85 86

10 1.1 PRELIMNARY STUOlES

1.2 MATERIAL MOOELS

io

2.1 FORCE -DISPLACEMENT MOOEL

2 2 STRAIN ANALYSIS MOOEL

2.3 OVNAMC RESPONSE MOOEL

2.4 DYNAMIC PROPERTCS

OF MASONRY SYSTEMS

10

11(A) IN-PLANE WALLS (STORY HEIGHT)

11(b) SEQUENTIAL DISPLACEMENT- N-PLANE WALLS

12(A) OUT-OF-PLANE WALLS

(STATIOCONC. MAS.)

3-2(bt) OUT-OF -PLANE WALLS (DYNAMCCONC. MAS ) 12(b2) OUT-OF-PLANE WALLS

(STATC A DYNAMIC-CLAY MAS.)

4.0

4.1 WALL-TO-WALL INTERSECTIONS

4.2 FLOOR- TO-WALL INTERSECTIONS

10

1! CONCRETE PLANK 01APHRAMS

S.2 EXISTING OIAPHRAM OATA 1

10

11 GROUTING PROCEDURE

12 REINFORCEMENT BONO A SPLCES

7.0 7.1 SMALL SCALE MOOELS

10

11 LIMIT STATE DESIGN METHODOLOGY

1.2 NUMERICAL RELIABILITY INDICES

9.1 DESIGN Of RESEARCH BU LONG

9.0 9.2 FACILITY PREPARATION

9 3 FULL SCALE TEST PLAN

9.4 FULL SCALE TEST

10.0

10.1 DESIGN RECOMMENDATIONS A CRITERIA DEVELOPMENT

11.0 11.1 COORONATION

TCCMAR MEETINGS

JTCCMAR MEETINGS

JUNE

86

JUNE

87

JUNE

88

JAN

09

JUNE

89

JAN

•90 JUNE

90

JAN

91

JUNE

91

JAN

92

AA —

AA —

A —

A —

—

JAAA1—4

1-

- 4-—

4

AA—I

AL.

i A-

—

hD

I n

A-<

A

- A --

! c ^ — PROPOSAL SU8UITTAL FIGURE 1 - U.S. PROGRAM SCHEDULE

294
Research Tasks

Category Task

1 .

1 .

1.2(a)

1.2(b)

2 .

2 .

2.2

2 .

2 .

3 .

3.1(a)

3.1(b)

3.1(c)

3.2(a)

3.2 (bl)

3.2 (b2)

4 .

4 .

4 .

5 .5.1

5.2

6 .6.1

6.2

7.0 7.1

8 .8 .

8 .

9 .

9.1

9.2

9 .

9.4

10 .

10.1

11.0 11 .

Table 1 - U.S. Coordinated Program for Masonry Building Research

Task Title

Preliminary Material Studies

Material Models- Concrete Masonry

Material Models- Brick Masonry

Force-Displacement Models

Finite Element Models

Dynamic Response Models of Shear Walls

Dynamic Properties of Masonry Systems

In-Plane Walls, Story Height

Sequential Displacement - In-Plane Walls (3 story)

In-Plane Walls, Two Story

Out-of-Plane Walls (Static-Concrete Masonry)

Out-of-Plane Walls (Dynamic-Concrete- Masonry)

Out-of-Plane Walls (Static & Dynamic - Clay Masonry)

Deleted

Floor- to-Wall Intersections

Concrete Plank Floor Diaphragms- In-Plane Behavior

Survey of Existing Diaphragm Data

Deleted

Reinforcement Bond & Splices

Shake Table Studies of Scale Model Buildings

Limit State Design Methodology

Numerical Reliability Indices

f Design of Research Building

Test Facility Preparation (for full-scale test)

Full Scale Test Plan

Full Scale Test

Design Recommendations & Criteria Development

Coordination

295
Manuscripts Authored

for Panel Meeting But

Not Presented Orally

Seismic Investigation and Improvement of

Electrical Utility Systems

BY

Joseph V. Tyrrell, P.B.^

ABSTRACT

Electrical utility systems were investigated

at naval bases location in areas of high

seismic risk in the United States. Electrical

systems within buildings as well as lifelines

are addressed. The earthquake safety and

reliability of commercially supplied power was

not a part of the investigations although it

is a key ingredient to performance after an earthquake.

1. INTRODUCTION

The Naval Facilities Engineering Command has

been conducting earthquake safety

investigations of naval shore facilities

located in areas of high seismic risk. These

investigations consist of identification of

site hazards (faults, tsunami, etc.),

establishing site response spectra based upon

a site seismicity study, And investigation of

lifelines and structures. The program

involves 55 naval bases or activities and as

of April 1987 is about 75% completed.

2. NATURE OF THE THREAT

Earthquake damage to electrical systems may

result from electrical causes (short circuits,

surges, overloads, etc.), structural failures

(supports, housing, adjacent structures,

etc.), site hazards (tsunami, liquefaction,

etc.), and secondary effects such as fire.

Prior to the San Fernando earthquake of 1971,

little attention to earthquake design was given to electrical systems. After 1971,

design codes were upgraded and some of the

lessons learned about electrical systems were

incorporated. Facilities constructed before

1973 warrant investigation and some that were constructed since may be deficient depending

on the criteria used and the designer.

The typical naval base purchases most of its

electric power from a commercial utility

company, but has some on-base generating

capacity. With regarded to manufactured

electrical equipment, earthquake safe

equipment is available for most items if

specified. Because of the high rate of

replacement of electrical equipment this

aspect of the problem will diminish fairly

rapidly in future years. The Navy

specifications cover new equipment

satisfactori ly. For these reasons, an equipment replacement program does not seem

to be necessary. However, many deficiencies

in existing electrical systems do warrant

attention including equipment supports and

anchorage.

There appears to be no feasible way to

complete prevent earthquake damage to a utility system. A large earthquake will very

likely cause some disruption of electrical

service regardless of the design code and

criteria used. The chance of disruption can

be minimized for specific facilities such as hospitals by use of standby emergency

generators and special care in the details of

the system. In a city or base wide system,

redundancy and multiple sources will limit

loss of power. Normal, good engineering

practice provides reasonable protection for

surviving system components in the event of

a problem of an electrical nature. Present

criteria for both structural and electrical

design seems adequate, but there are no standards for performance the earthquake

survivability of electrical systems.

The bulk of potential problems with existing

electrical systems fall within the realm of

structural engineering. Fortunately,

structural criteria is available and design

and analysis methods well known. However,

even if all of the system components are structurally adequate, survivability is not

assured because of the consequences of damage

to structures housing the system or to

adjacent structures.

Secondary losses may result from earthquake

damage to an electrical utility. These

include fire, injury from electric shock,

release of hazardous materials, loss of

production, and loss due of rescue and

recovery being impeded by lack of power.

Many transformers in use at naval bases

contain PCB's, and could cause contamination. These transformers are being

replaced but it take several years before the

hazard is eliminated.

^ Naval Facilities Engineering Command,

Consultant in Earthquake Engineering

299
3. SITE HAZARDS

All facilities in active seismic areas are potentially subject to damage due to ground

shaking. The motions induced in electrical

systems may cause shorting and various degrees

of damage due to swinging wires,

displacements, and deflections in structures.

It may also cause structural damage to the

structures which are part of or adjacent to

the system. Other site hazards are more localized and include faulting, liquefaction,

soil slides and instability, tsunami or seiche, and the hazards resulting from

adjacent facilities.

4. IDENTIFICATION OF HAZARDS

Base maps showing the roads and structures

were useful in assessing in conjunction with

discussions with local public works office

personnel.. Utility maps show the power

distribution system. The system layout can be

checked against known or suspected site

hazards. The need for redundancy, looping, or entry at additional points can be considered.

In this connection the use and positioning of

emergency generators is appropriate. If

necessary, a follow-up investigation of site

hazards is made.

A through examination of the system components

is then made. The easiest and most effective ’ technique was found to be a visual

examination. Where major buildings are a part

of the system, it is necessary to determine if

a structural analysis is needed. The Navy has

developed a method for rapid preliminary

evaluation. It was found to be useful to

employ a set of design guides for inspecting

the components, determining mitigations, and

estimating cost. Figures 1, 2, & 3 are typical examples. All costs are in 1975

dollars.

5.

RESULTS OF INVESTIGATIONS

Problems with anchorage and bracing were

common and found in many locations.

Conductors running between two points rigidly

anchored were found and differential movement

would surely cause a break. Some structures,

particularly older masonry buildings, were

found to be inadequate. Typical conditions of

system components are described belov/.

5.1.

GENERATING PLANTS

The boilers and generators themselves were

adequate, however, a few were not proper

anchored. Ancillary equipment such as heat

exchangers and condensate tanks were found to

lack proper anchorage and sometimes needed

bracing. Conductors and piping frequently

were not detailed to accommodate differential

displacements. Some piping supports did not

provide any lateral resistance. About 50% of

the buildings required some structural

upgrading.

5.2. SUBSTATIONS

About 75% of the transformers and other

equipment was unanchored and vulnerable to

sliding or overturning. About 25% required

bracing. A few dead end structures were

found to be unstable.

5.3. UTILITY POLES’

Most poles at the activities investigated were wood but some reinforced concrete or

steel was found. The typical utility pole

with one transformer is satisfactory;

however, they vary in height and loading

conditions. It is probably worthwhile to

check a few. Some poles are utilized for

multiple purposes including telephone lines

and street lights. Others may have as many

as four transformers mounted on them. Where

there are heavy loads and/or changes in line

direction, poles are sometimes guyed. An

experienced structural engineer can tell when

these conditions require analysis. In the

past the Navy experienced some earthquake

damage to poles on Guam.

5.4.

STREET LIGHTS

Where street light poles were single purpose

types, no problems were found. It may be

prudent to check a typical unit.

5.5.

SWITCHGEAR AND CONTROL PANELS

At least one third were found to be

unanchored and most were unstable under

earthquake loading and required bracing. As

with many of the other deficiencies found,

remedial measures were simple and inexpensive.

5.6.

PUMPS AND PUMP MOTORS

Most were found to be adequately supported

and secured; however, ancillary equipment and

piping frequently needed improvement. Where

vibration isolators are used under pumps or equipment, their effect must be considered

and their anchorage checked. Lateral

restraint is needed without interfering with

the function of the isolator.

300
5.7.

BATTERY RACKS

Racks are typically unanchored and batteries

were sitting on shelves with no lip or restraint of any kind. Even a slight shake

could cause disruption. 5.8.

TELEPHONE EXCHANGES

Switchboards and distribution panels were usually anchored but found to require bracing

at the top of the frames.

5.9.

IN-BUILDING LIGHT FIXTURES

Several different types of fixtures were

common. The support and mounting were the

principle concerns. In general, chain hung

fixtures and those on swivel pendants were

satisfactory. Some were support by rigid

pendants or light gage vertical members, or by

light gage suspended ceiling supports. Most

of these latter types were vulnerable and it was considered necessary to add safety chains

from the fixture to a competent structural

support above.

6. CONCLUSIONS

Existing electrical systems were found to be

vulnerable to earthquake damage which would

cause power outages and other damage. A great

deal of improvement can be accomplished by

simple, inexpensive measures such as anchorage, bracing, and reinforcement of

supports. Certain major structures such as generating stations are vulnerable - especially older masonry buildings.

Structural improvements of buildings may be

simple and inexpensive (e.g. providing X bracing to a steel frame) or they may be

complex and costly. In some cases it may be

economical to rebuild rather than retrofit a structure.

Since it is not feasible to provide a completely earthquake proof system, it is

necessary to plan and prepare for disaster

recovery. Looping the system and providing

more than one feed point are desirable steps

in minimizing loss of power. The use of

standby emergency generators should be

considered. Flexible connections should be

provided and slack allowed between rigidly

anchored points. Dispersed and/or on-site

water supply is desirable.

301
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303
FIGURE 3

304
Wind Tunnel Test on Behavior of Rooftiles

under Strong Wind

BY

* i *2

Hisashi Okada 1 and Tatsuo Murota

SUMMARY

A series of wind tunnel tests was

conducted on the fluttering and scattering

behavior of "hikkake-san-gawara", a hook type

rooftiles, which are used widely for roofings

of Japanese wooden dwellings. 1/20 scale

models of two-storied dwellings of a rectangular plan and 1/20 miniature rooftile

models made by brass cast were used for the

test. The boundary layer realized in the wind

tunnel corresponded to that on rural

districts.

Following results were obtained:

(1) The spatial distribution of rooftile

scattering was non-symmetric. Rooftiles on the right side roofs were susceptible to

scattering

.

(2) Locations of fastening in the customary

tile-roofing works was effective to prevent

rooftile scattering in strong winds of less

than about 40m/s.

(3) Based on the consideration on the wind

resistant mechanism in hikkake-san-gawara

roofing system, the value of wind force

coefficient for the first fluttering of

rooftiles was obtained.

KEY WORDS : Wind Tunnel Test, Wind Resistant

Mechanism of Rooftiles, Fluttering and

Scattering of Rooftiles, Wind Force

Coefficient

1^ INTRODUCTION

Roofs of timber construction dwellings

in Japan are mostly covered by clay tiles

called as "hikkake-san-gawara" which are set on roof sheathings by hooking their upper

edges to crosspieces nailed on sheathing

boards. This type of rooftiles has often been

damaged by strong winds. In order to prevent

those damages, tile-roofing workmen fasten

rooftiles with nails in some special

locations known empirically.

In 1986 Building Research Institute was requested by Japan Clay-Tile Roofing

Constructor Association to make wind tunnel

tests to get information on the behavior of

clay-tile roofings in winds and also on the

effect of nail fastenings.

This paper describes the results of the

wind tunnel tests.

2^ DESCRIPTION OF WIND TUNNEL TESTS

2 .

1

Wind tunnel

The tests were conducted using the

turbulent boundary layer wind tunnel of

Building Research Institute, the dimensions

and performance of which are listed in Table

1 .

2 .

2

Similarity Law

2.2.1 Similarity of air flow

Mean wind velocity profiles in turbulent

boundary layer can be described by the

following equation.

U* Z U(Z)= —— ln( —— ) (1)

K. Zo

in which, U(Z):mean wind velocity at the

level Z U* : friction velocity

K :Karman's constant

Zo :roughness parameter

In order to derive similarity for the mean wind velocity profile, the following

similarity equation which is called "Jensen's

similarity law" must be satisfied.

)p < Zo

( 2 )

in which, Z is height and/or length scale and

suffix p and m refer to the quantity for

full-scale and model, respectively.

Eq.(2) can be modified as follows:

(Z)m

_ (Zo)m

(Z)p (Zo)p

The term on the left hand side of eq.(3)

is a geometric parameter' related to the ratio

of model scale to full-scale.

2.2.2

Similarity of forces on rooftiles

The wind force(F) which acts on a rooftile can be expressed as the product of

velocity pressure(q), wind force

coef f icient(C) and area of rooftile(A).

F=C * q

* A (4)

in which, q=l/2pU^

p=air density

U=wind velocity

On the other hand, the force of

resistance(R) against the removal of a rooftile which is not fastened to sheathing

boards is approximately the same as the

weight of the rooftile. The force(R) can be

expressed as follows:

R=w* A*cos0 (5)

in which, w=weight of a roof tile per

unit area 0=pitch of roof

For the similitude of force, the

following condition must be satisfied:

*1 Senior Research Officer, Building

Aerodynamics Division, Structural Engineering

Department, Building Research Institute,

Ministry of Construction

*2 Director, Structural Engineering

Department, ditto

305
( F/R

p=( F/R

m ( 6 ) Substituting eqs.(4) and (5) into

eq.(6), the following equation can be

obtained

<-n£-)M—£-)* (7)

Um “m

This equation is very important to

convert the wind velocity in wind tunnel

tests into the one in full-scale phenomena.

2.3

MODELS

2.3.1 Dwelling Models

Two 1/20 scale models two-storied

dwellings were used for wind tunnel tests.

Gable and hip roofs of 4.5/10 pitch were

selected to be tested as shown in Figs.l and

2, respectively. Crosspieces are attached to roof surfaces in a direction parallel to the

edge lines of eaves at a space of the

effective length of rooftile as shown in

Fig. 4.

2.3.2 Roof-Tile Models

1/20 scale models of hikkake-san-gawara

which are identified as clay rooftiles of

JIS 53A were made by casting brass into

moulds. A set of JIS 53A clay rooftiles

includes rooftiles of 12 different shapes

(Photo. 1 and 2). The shape of rooftil.es which are used most in number in tile roofing is

shown in Fig. 3. Here after we will call this

rooftile as a common tile.

Common tiles are used at locations

except ridges, verges and eaves. In this wind

tunnel test, we casted 1496 and 1756 pieces

of miniature rooftiles for a gable roof and a hip roof model, respectively. About 80%(gable

roof) and 75%(hip roof) of them were common

tiles.

Common tiles have small projections

called "Hikkake" on upper edges of their

under sides. This projection act as a hook on setting a common tile on crosspieces(Fig,4).

Figs. 5 and 6 show rooftile arrangement

for gable and hip roof, respectively. In case

of actual tile-roofing works, rooftiles set

along ridges or edges of roofs are not only

hooked on crosspieces but also fastened by

nails or copper wires. In the wind tunnel

tests, two tile-fastening conditions were tested for comparison. One is the case

similar to actual works stated above and the

other is the case where all rooftiles are put on the sheathing without fastening. In the

former case rooftiles fastened to the

sheathing by adhesive tapes at places shown

in Figs. 5 and 6. Adhesive tape fastening was effective up to the wind speed of 40m/s at

height of the roof ridge.

In Table 2 weights of miniature

rooftiles are shown. The weight of a common

tile was 1.22gw, which was 1/2,220 of the

full-scale tile(2.7kgw). The value of w /w

m

in eq.(7) was 5.55 for common tiles.

2.4 Test Series

Table 3 describes eighteen cases of the

wind tunnel tests.

Wind directions were measured in such a way as shown in Figs. 5 and 6. In most cases the roof models was

exposed to wind from a certain direction and

the wind speed was increased step by step.

On the other hand, in case of test No. 8, 13

and 18, the roof model was rotated slowly and

continuously on the turn-table and was

exposed to wind of a constant speed, shown on the column "wind speed" in Table 3.

2.5 Wind Tunnel Flow

The wind tunnel flow was arranged to be

similar to natural winds by arranging

roughness blocks and spires set on the floor

of wind tunnel. Mean wind velocity and

turbulent intensity profiles of the wind

tunnel flow at the site of roof models after

arrangement are shown in Fig. 7. The value of

the roughness parameter Zo in the wind tunnel

flow was about 1.3cm. It corresponds to 26cm

in full-scale because the size reduction

ratio of the model is 1/20. According to

Table 4 the wind tunnel flow is similar to

those on rural districts. Fig. 8 shows power

spectrum of turbulence at the height of 40cm

above tunnel floor. This height corresponds

to the level of roof ridges.

3. TEST RESULT

The behavior of rooftiles was observed

at each step of wind speeds and recorded by

video tape recorders. Locations of rooftiles

which moved in wind are shown in Figs. 10 to

25 with reference to the wind speeds. In

these figures the movements of rooftiles are classified into two types;

1) fluttering: remarkable fluttering at or near the initial position. The distance of

displacement is less than the width or breadth of a clay tile.

2) scattering; movement with the larger

distance of displacement.

The values of wind speed in the figures

correspond to those for full-scale phenomena

at height of eaves and they mean that the

movements of rooftiles shown there occurred

at a wind speed less than the value in

parentheses( ) or at a wind speed larger than one in ( ) and less than another without

parentheses.

The minimum wind speed at which the

first remarkable movement of a rooftile was

observed in listed on Table 5 for each test. The details of rooftile movements will

be described in the following two sections,

where the words "right-side roof" or "left

side roof" will be used without noticing that

they are seen from the windward.

3. 1 Movement of rooftiles

1) In case of roofs with no fastened

rooftiles

(1) Gable roof, wind direction 0° (Fig. 10)

When the wind speed exceeded 25m/s,

rooftiles near the edge of the windward verge

began to flutter. Some of them scattered at

306
the wind speed more than 30m/s to leeward

direction on the right-side roof surfaces and

on the contrary to wind ward direction on the

left-side roof surface. Tiles on the windward

ridge were demolished at the same time.

The increase in the number of scattered

tiles was more rapid on the right side roof

surface, where most rooftiles in five or six rows near the eaves were blown off when the

wind speed exceeded 40m/s.

(2) Gable roof, wind direction 90° (Fig. 11)

Rooftiles in two rows on the windward

eaves began to flutter when the wind speed

exceeded 25m/s, and tiles in the first row

scattered before the wind speed came to

30m/s. No other scattering was observed after

that. At about 40m/s two rows of tiles on the

windward right-side verge and a few rooftiles

on the leeward left-side verge were blown

off.

(3) Gable roof, wind direction -45° (Fig. 12)

When the wind speed exceeded 30m/s, the

first fluttering was observed along eaves and

verge near the windward corner. The

fluttering of eaves rooftiles extended

leeward and all of them fluttered or a few of

them scattered at 35m/s. No other scattering was observed until the wind speed became

40m/s after that. At 40m/s rooftiles on the

ridge and rooftiles on the verge of leeward

roof scattered.

(4) Gable roof, wind direction 45° (Fig. 13)

When the wind speed exceeded 25m/s,

fluttering occurred on rooftiles on the

windward verge and eaves of a windward roof.

At a little higher than 30m/s a row on windward eaves, a few row on the windward

verge of the windward roof and a row on the

windward verge of the leeward roof scattered.

Area of scattering on the windward roof

spread with the increase of wind speed and

most of the windward rooftiles were blown off

at 40m/s.

(5) Hip roof, wind direction 0° (Fig. 14)

At a little higher than 25m/s, two rows

of rooftiles near eaves on the windward roof

began to flutter and the first row was blown

off at about 30m/s. At a little lower than

35m/s, scattered rooftiles on windward eaves

of the right-side roof parallel to the wind

direction and also rooftiles on windward

corner ridges.

Scatter of rooftiles of the right-side

roof spread leeward with the increase of wind

speed

.

(6) Hip roof, wind direction 90° (Fig. 15)

In this case the movement of rooftiles

was similar to that in the preceding case.

(7) Hip roof, wind direction -45° (Fig. 16)

At a wind speed of a little lower than

30m/s, two rows of rooftiles on eaves of the

right-side windward roof began to flutter and

they scattered soon after that. No other

scattering occurred until 40m/s, when all the

rooftiles of the right-side windward roof

scattered suddenly. Several rooftiles on the

left-side windward roof also scattered at the same time.

2) In case of roofs with fastened rooftiles

(1)

In the tests No. 10, 12, 15, 16 and 17

No remarkable movement of rooftiles was

observed

.

(2) Gable and hip roof, wind direction 0°

(Figs. 17, 21)

The first fluttering was observed at

about 30m/s on rooftiles on the windward part

of roofs parallel to the wind direction and a few of them scattered at a wind speed of a little lower than 40m/s.

(3) Gable roof, wind direction -45° (Fig. 19)

At 35m/s a few rooftiles on the 2nd and

3rd row in the windward verge of the windward

roof began to flutter, but they didn't

scatter at a wind speed more than 40m/s. In

this case rooftiles on the first row are

fastened

.

3.2 Non-svmme tricity of scattering and effect

of fastening

It will be noted that even if the shapes

of right and left side roofs are symmetric,

the areas of rooftile scattering on right and

left side roofs are not symmetric. In general

the area of scattering on the right side roof

is far larger than that on the left side

roof. The directions of scattering are also

different on the right and left side roofs:

on the right-side roof, rooftiles scattered

leeward but on the left-side roof they

scattered to the opposite direction.

The non-symmetric feature of scattering

is considered to be caused by the nonsymmetricity of tile-roofing works shown in

Fig. 9; the left and bottom edge of a rooftile are put on the adjacent tiles but the right

and upper edge are put under another

rooftiles.

Comparing the area of rooftile

scattering shown in Figs. 10 to 16 with those

in Figs. 17 to 22, the fastening is very

effective for preventing scattering and

fluttering of rooftiles. It is considered

that the locations of fastening selected

empirically in the customary roofing works

will be effective to prevent scattering in

strong winds of less than 40m/s.

3.3 Comparison of scattering in winds of

fixed direction with that in winds of

non-f ixed direction

Figs. 23 to 25 show the area of

scattering after the test No. 8, 13 and 18 in

which the wind speed was constant but the

wind direction changed continuously. The

areas of scattering in these tests are

generally larger than those in the tests for

fixed wind direction. The difference is

considered to be reasonable, because

rooftiles in the test No. 8, 13 and 18 were

exposed to more extra wind effects than those

in the other tests.

3.4 Force coefficient and wind resistant

mechanism of rooftiles

Wind resistant mechanism of the

"hikkake-san-ga wara" roofing system can be

modeled as shown in Fig. 9, where forces f

^ is sum of wind force, weight of a rooftile and

fastening resistance if the tile is fastened.

Supposing that the i-th rooftile is

307
lifted by winds, force f^ must satisfy the

equation:

f i >(

3

)f i+l +(

3

) 2

f i+2+(

3

) 3

f i+3+

“ " ' (8)

In general it is not unreasonable to suppose

that the wind force acting on tile i is

approximately the same as those acting on the

adjacent tiles because the size of tiles is

very small compared with that of roofs and

the spatial change of wind pressure is not so large in a small area of roof surfaces.

Therefore, in the case that rooftiles

are only hooked on crosspieces and not

fastened to sheathings, force f

i can be

supposed to be approximately equal to

f i+ j

( j = 1 , 2 , 3 , *•* ). Then the eq.(8) can be

simplified as f^>0. This condition leads to

the wind force coefficient as follows:

C=

W W

l/2pU 2 A

~ 1/16U2 A

(9)

where, W=weight of rooftile

A=effective area of a rooftile

If the (i+k)th rooftile is fastened and

the sum of wind force and weight of rooftiles

is the same for i-th and i+j( j = l ,2,***,k)th

r oo f t i les , th e fastening resistance Ff

necessary for preventing the scatter of the

i-th rooftile can be obtained as follows:

F

f >f i^-.M^)^/a/3)k do)

j=l 3

The first fluttering wind speeds of

rooftiles obtained in the tests were

approximately 25ra/s in all cases. The

corresponding wind force coefficient is

obtained by eq.(9);

C=

2. 7kgw

1/16x25 2 x( 0. 235x0.265)

= 1.11 (ID

where, 0. 235=ef fective breadth of rooftiles

in mm

0. 265=ef fective length of rooftiles

in mm

References

1) N. Frank and M. Jensen "Model-Scale Tests

in Turbulent Wind, Part 1" , The Danish

Technical Press,1963

2)

J. Couhnian "Adiabatic Atmospheric Boundary

Layer : A Review and Analysis of Data from

the Period 1880-1972" (Atmospheric

Environment,Vol.9,Pergamon Press, 1975

A

4^ CONCLUSION

Model scale tests were carried out in a boundary layer wind tunnel on the behavior of

"hikkake-san-ga wara" roofing system. The

following conclusion were obtained from the

tests:

(1) The spatial distribution of rooftile

scattering was non-symmetr ic. Rooftiles on the right side roofs were susceptible to

scattering.

(2) The fastening of rooftiles was very

effective for preventing scattering and

fluttering of rooftiles. It is observed that

the locations of fastening in the customary

tile-roofing works was effective to prevent

rooftile scattering in strong winds of less

than about 40m/s.

(3) The wind resistant mechanism of hikkakesan-gawara was taken into consideration and

the wind force coefficient for the first

fluttering of rooftiles was obtained.
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*4-4 u c •H -3 0) ' 4-1 4-1 co 0) o 3 3 »H y y y i 0) C 00 •H E- •H iH H H "3 o

o 4-1 >4—

)

4-1 •H •H ; -H 3 c

-3 H CO U o 4-> 4-1 4-> y y

<o CO c C0 c 1 _ ! 5- C -fH u 0J o CO y y y y

01 _x 4-1 CO E 01 CO CO CO CO E u c c E > u. U *3 “3

•rH O o CO O 3 y y •H •H 1 Q 3 CJ cc y 0) ( > >

1 u

Wind

Speed

1

COEO

'T

! | CO: CO E E

CN CN

<r i

c

o y

co •H y

y H i« •H u CO c c

U y y o o o y i- •o •H oo •H 4-J 4-1 4-1 Q O O m in o o m co O O in m co O o m in co

4-1 <r vj- cti sr jj O' j-j O' <r -u

CO -3 i 1 o 1 o 1 o

01 C u u u H •H

coc

•H co y y CO CD 01 CO c y <u 3 CO CO y co co 3 c c > “3 U > *3 H

oi HJ o o CU *H 0) co -h y

fH CO c c y i- > y t- >

-3 CU

C0 Cj- 8- 3

3.

CD -C y y

00 tH

3. 40 CL >H co •H CU •H o CO r" oo JZ o

2

o fH CN CO nJ 5 6 7 8 O' O—, _H—1 CN CO <r o io

P-l mo^4 ^-4

j

309
Fig.

6

Arrangement

of

rooftiles

(Hip

Roof)

Fig.

2

Hip

roof

model

Fig.

4

Hooking

mechanism

of

(dotted

rooftiles

are

fastened

in

cases

(scale

in

mm)

rooftiles

to

crosspieces

0f

Ce

sts

on

the

effect

of

fastening)
Height

Z

(cm)

Turbulent

intensity <7U/U (%)

0 5 10 - 15 20

Mean wind

velocity U (Z) /U (100)

Fig. 7 Mean wind velocity and turbulent

intensity profile of wind tunnel flow

Fig. 8 Power spectrum of turbulence

of wind tunnel flow

Fig. 9 Overlapping of rooftiles and

wind resistant mechanism model

Photo. 1 Shapes of miniature rooftiles (1) Photo. 2 Shapes of miniature rooftiles (2)

311
(a)U=27m/s

(25m/s)

(a)U=29m/

(27m/s)

(b)U=34m/s

(32ra/s)

(b)U=34m/s

(32m/s)

(c)U=41m/s

(40m/ s) A

(c )U=41m/

(40m/ s)

Fig. 10 Distribution of movement of rooftiles

( Gable Roof without fastening

, Wind Direction =0 degree )

Fig. 11 Distribution of movement of rooftiles

( Gable Roof without fastening

s Wind Direction =90 degree )

scattered in the

preceding test but

is not repaired

fluttered in

the preceding

steps

fluttered in

this step

scattered in

the preceding

steps

scattered in

this step

312
A

(a)U=32m/s

(30m/ s)

A

A

(38m/s)

(27m/s)

(b)U=34m/s

(32m/s)

Fig. 12 Distribution of movement of rooftiles

( Gable Roof without fastening

, Wind Direction =-45 degree)

Fig. 13 Distribution of movement of rooftiles

( Gable Roof without fastening

, Wind Direction =45 degree )

scattered in the preceding test but

is not repaired

fluttered in

the preceding

steps

fluttered in this step

scattered - in scattered in

the preceding this step

steps

313
(a)U=27ra/s

(25ra/s) A

(a)U=29m/s

(27m/ s)

(b)U=34m/s

(32m/ s)

(b)U=34m/s

(32m/s)

Fig. 14 Distribution of movement of rooftiles

( Hip Roof without fastening

, Wind Direction =0 degree )

Fig. 15 Distribution of movement of rooftiles

( Hip Roof without fastening

, Wind Direction =90 degree)

scattered in the

preceding test but

is not repaired

fluttered in

the preceding

steps

fluttered in

this step

scattered in

the preceding

steps

scattered in

this step

314
U=36m/

s

(35m/s)

A

r>

(a)U=32m/s

( 29m/s)

A

A

(c)U=40m/

s

(38m/ s)

Fig. 16 Distribution of movement of rooftiles

( Hip Roof without fastening

, Wind Direction =-45 degree )

Fig. 17 Distribution of movement of rooftiles

( Gable Roof with fastening

, Wind Direction =0 degree )

Fig. 18 Distribution of movement of rooftiles

( Gable Roof with fastening

, Wind Direction =90 degree ) A

U=41m/s

(39m/ s)

Fig. 19 Distribution of movement of rooftiles

( Gable Roof with fastening

, Wind Direction = -45 degree )

scattered in the preceding test but

is not repaired

fluttered in the preceding

steps

scattered in

this step

scattered in the preceding

steps

fluttered in this step

315
U=42m/s

(39m./ s)

Fig. 20 Distribution of movement of rooftiles

( Gable Roof with fastening

, Wind Direction =45 degree )

Fig. 23 Distribution of movement of rooftiles

( Gable Roof without fastening

, Rotation , U=40m/s )

U=40m/s

(38m/ s)

Fig. 21 Distribution of movement of rooftiles

( Hip Roof with fastening

, Wind Direction =0 degree )

Fig. 24 Distribution of movement of rooftiles

( Gable Roof with fastening

, Rotation ,U=42m/s )

U=43m/

(41m/s)

Fig. 22 Distribution of movement of rooftiles

( Hip Roof with fastening

, Wind Direction =90 degree )

Fig. 25 Distribution of movement of rooftiles

( Hip Roof with fastening

, Rotation ,U=42m/s )

scattered in the

preceding test but

is not repaired

fluttered in the preceding

steps

fluttered in

this step

scattered in

the preceding

steps

scattered in this step

316
Seismic Microzonation Using Urban Ground Information

BY

1 2 Shinkichi Kishi , Toshikazu Morohoshi

Keiichi Otani^ and Hiroshi Takahashi4

ABSTRACT

The data of about 30,000 borehole

records centering around the

Metropolitan area has been stored in

the database. The data were produced

at the fundamental survey for

construction of various structures.

Regional distribution of the ground

properties such as thickness of soft

ground, vibrational characteristics

of ground, liquefaction potential of

sand layer, etc. are grasped from

the database.

In addition, using the character istics

of earthquake motions in the basement

and in the sedimentary layer estimated

by the model constructed from the

observed data at network of strongmotions including three deep

boreholes, regional distribution of

the size of surface ground motions

were estimated.

of the deep boreholes in the preTertiary basement and installed at

the strong-motion observation points

on the ground. A tentative estimation

procedure were already proposed^

.

In this study a whole ground

structure is divided into seismic

basement and sedimentary layer, and

furthermore the sedimentary layer is

divided into engineering bedrock and

surface ground. In this paper a method is described to estimate the

regional distribution of the size of

surface ground motions in the study

area of eastern neighbor of Tokyo

Metropolis against an assumed

earthquake, using the amplification

characteristics of engineering bedrock

estimated from the above stated

procedure and of surface ground

estimated from the database and the

observed data of microtremors.

KEY WORDS: Database, Urban ground,

Borehole record. Microtremors, Ground

motion. Seismic microzonation.

1 . INTRODUCTION

Through the Niigata Earthquake in 1964

and the off-Tokachi Earthquake in

1968, because of the relations between

the damage and the ground properties,

development of the database system of

borehole records in urban areas and

development of its application methods

for prediction of seismic damage have

been conducted in NRCDP (National

Research Center for Disaster

Prevention) . Some of the results are

practically utilized in a few

municipal authorities and at present

the data of 20,000 borings in Tokyo

metropolitan area and of 10,000

borings in Sizuoka prefecture are

stored in the database"'.

On the otherhand, in NRCDP, a research

project to estimate the average

amplification character ist ics of the

thick sedimentary layer for the short

period of S-waves in a specific region

of the Metropolitan area has been

under way using the data of

accelerometers installed at the bottom

2. SUMMARY OF THE DATABASE

The data of depth of layer, soil name,

N-value, etc. on the borehole records

mainly produced at the fundamental

survey for construction of various

structures are stored in the database.

Location of boring was determined by

the mesh with one side about 100 meter

long

.

2 . 1 Input System

The mnemonic code tables for nonnumerical data, such as soil name and

soil color, were designed in order to

make the input work easier. A part of

soil code is shown in Fig.1.

Recently two input works of coding and

punching were put together by making

most of the microcomputer system, and

Head, Information Processing Labo

.

4th Research Division, National

Research Center for Disaster

2 Prevention, Japan.

^Researcher, Infomation Processing

^Labo. 4th Research Division, NRCDP.

Head, Earthquake Engineering Labo.

^2nd Research Division, NRCDP, Japan.

Director General, NRCDP, Japan.

317
as a result the time required in the

work has been sharply shortened.

2.2 Accumulation System

Fig. 2 shows how the boring data are

arranged in computer. Each item is

coded to binary type and packed in a word

Retrieval by location index is basic

in microzoning. The location of

boring is given by the mesh with one

side about 100 meter long. This mesh

size is equal to the size of dividing

side and length of the 1:25,000

topographic map, published by

Geographical Survey Institute of Japan,

into 100 equal parts, i.e. 4.5" in

longitude and 3.0" in latitude, and is

convenient to overlay with other

information in practical use. With the expansion of the area of

accumulated data, geographical filing

has been required. In the database,

a unit file is constituted on the

basis of the 25,000 topographic map.

2.3 Retrieval System

As a typical system there is a manmachine interaction system for the

retrieval and display of the soil

profile using XY-digitizer and CRT

(Cathode Ray Tube) . The system

configuration is shown in Fig. 3. In

this system, soil profile at requested

point or line on the map indicated by

the input-pen is drawn on the CRT, as

is shown in Fig. 4. In the picture

figures on the left signify the depth

of layer and those on the right

signify N-value.

3. OBSERVATION OF MICROTREMORS AND

REGIONAL DISTRIBUTION OF

VIBRATIONAL CHARACTERISTICS OF

THE SURFACE GROUND

As a means to measure the vibrational

characteristics of the surface ground,

observation of microtremors were

carried out at several tens of sites

in the study area. While a method to

output the regional distribution of

vibrational characteristics of the

ground from the database is described.

Here we define the layer with N-value

more than 50 as an engineering bedrock

and presume the upper layer i.e.

surface ground as a single layer of

which elastic property is determined

from the average of N-value.

3.1 Predominant Period of the Ground

obtained from the Observation of

Microtremors

The observation was carried out for

3-5 minutes at every point. Used

apparatuses were 1Hz velocity

horizontal component seismographs,

amplifiers and data recorder. The

most noiseless part for 10 seconds

in the record were input to the

spectrum analyser and a period which

showed a maximum response was read

as the predominant period of the

ground

3.2 A Formula to estimate Vs from

N-value

In the observation point of microtremors, when we suppose the surface

ground as a single layer, the velocity

of S-wave Vs is obtained from the Law

of quater wavelength

Vs = 4H /T

where H and T are the thickness and

the predominant period of the layer,

respectively

On the other hand, calculating the

average of N-value N in the surface

ground from the borehole record of

the observation point, and supposing

a relation

Vs=aN b

, and determining the coefficients by

the method of least squares, we got

a formula

3.3

Regional Distribution of the

Vibrational Characteristics of the

Ground

Predominant period of surface ground

is estimated from the formula

T=4H/Vs

where H and Vs are determined

automatically from N-value on the

borehole record in the database.

The above stated results are shown in

Fig. 5 with contours on the topographic

map. Solid line and broken line show

the contours of the predominant periods

obtained from the observation and

estimated from the database,

respectively. Black dots show the

observation points of microtremors

with measured predominant per iods ( sec )

Borehole records in Fig. 4 are the data

of the observation point A-D and each

line on the right side sgnifies the

boundary of surface ground and

engineering bedrock.

318
In Fig. 5, outlines of the contours of

measured and estimated predominant

periods coincide with each other.

While the contours of 1 sec coincide

well, the contour of measured period

of 0.5 sec is seemed to lie on the

contour of estimated period of 0.75

sec. On the other hand the boring

data are densely distributed than the

observation points of microtremors

and therefore discordance between both

contours are seen in the region where

could not be interpolated by the

measured data.

4 . SEISMIC MICROZONATION

Seismic microzonation is to evaluate

the characterist ics of the earthquake

motions in each mesh in the objective

region as large as a municipality,

considering the difference of

erathquake motions due to the locality

of the ground condition, and to apply

the result to make the various kinds

of maps predicting the seismic damage.

In this study we divide a whole ground

structure into seismic basement and

sedimentary layer. The seismic

basement is the pre-Tertiary layer and

the sedimentary layer is a general

term for Tertiary and Quaternary layer.

Moreover we divide the sedimantary

layer into engineering bedrock and

surface ground. The surface ground

is the alluvial layer with N-value

less than 50 and the engineering

bedrock is equivalent to Tokyo-Reki-So

with N-value more than 50.

4.1 Characteristics of the Basement

Motions

Characteristics of the basement motions

is expressed by the zero-damped

velocity response spectrum SVi(T;M,R)

(gal-s) of the earthquake motions in

the pre-Tertiary basement. This

characteristics is given by the model

SVi ( T ; M , R ) = 1 0 a (T) M-b (T) lo 9-| 0

R+C (T)

here M and R are the JMA magnitude and

hypocentral distance in kilometers.

The coefficients a, b and c are

distinguished by the location of

epicenter against the volcanic front

of Japan. In this study the

earthquakes were limited to the one

whose epicenters were located out of

the front.

4.2 Average Amplification

Characteristics of the Sedimentary

Layer

The average amplification

characteristics of the sedimentary

layer at the standard observation

point on the ground is defined as follows

,

1 N

G(T)=^ X SVo ( T ; Mn , Rn ) /SVi ( T ; Mn , Rn

)

n = 1 where SVo(T;Mn,Rn) is the zero-damped

velocity response spectrum obtained

from the observed data on the ground

for n'th earthquake (total N)

.

4.3 Average Amplification

Characteristics of the Engineering

Bedrock

The average amplification

characteristics of the engineering

bedrock toward the basement motions

is given by dividing G(T) by Go(T)

which signifies the amplification

characteristics of the surface

ground at the standard observation

point

Ge (T) =G (T) /Go (T)

In practice Ge(T) is given by the

approximation

Ge (T)» 2 Kn- ( log

1

Q

T) n

where n=0~4, and the coefficients Kn

are determined by the method of least

squares

.

Now the above stated character istics

were calculated for the data of 7 earthquakes2 observed at the standard

point (Urayasu city hall) between

1980 and 1984.

Fig. 6 shows the estimated zero-damped

velocity response spectra in the preTertiary basement. Fig. 7 shows the

average amplification characteristics

of the sedimentary layer at the

standard observation point (broken

line) and the amplification

characteristics of the surface ground

determined from N-value (solid line).

Fig. 8 shows the average amplification

characteristics of the engineering

bedrock toward the basement motions.

While borehole record of the standard

observation point is shown in Fig. 9.

4.4 Characteristics of the Earthquake

Motions at the Objective Point

The characteristics of the earthquake

motions at m'th objective point in the

study area is given in the zero-damped

velocity response spectra

SVm ( T ; M , R ) = Gm ( T ) • Ge (T) -SVi ( T ; M , R

)

319
Where Gm(T) is the amplification

characteristics of the surface ground

at the objective point determined

from the boring data. In this study,

as a scale for microzonat ion to

evaluate the earthquake motions, we

adopt the spectral intensity (gal-s 2

) proposed by Housner

C T2 SIm=\ SVm (T ; M , R) dT

JT 1 This is not a momentary scale as a maximum acceleration but is more

stable scale.

4.5 Estimation of Regional

Distribution of Spectral Intensity

Although the spectral intensity is

provided for the integral of zerodamped velocity response spectra from

T -j = 0 . Is to T 2

= 2.5s, here we set

T-|=0.15s because of the disturbance

of Ge (T) in the short period range

and T2=4.0 s in order to take account

of the predominant vibration of the

ground widely.

Here notice that the amplification

characteristics of the surface ground

at the objective points correspond to

the velocity structure mentiond in the

previous chapter, chat is, a single

layer with Vs determined by the

average of N-value. And the Q-value

and the density of the surface ground

are uniformly given a value 10 and a value 1.65, respectively.

Fig. 10 shows an example of the

distribution of the spectral intensity

of earthquake motions in the study

area estimated through the procedures

described in this paper. Magnitude and

hypocentral distance of the assumed

earthquake is respectively 7.0 and

100 km. Estimated spectral intensity

is displayed in each mesh with the

number corresponding to each rank of

the intensity. Contours are drawn by

hand for the intensity 200 and 240.

Looking wholly at the figure a tendency is found that the spectral

intensities become larger from

northern region (the upper part) to

southern reclaimed land (the lower

part). The tendency appears similarly

in the distribution of the predominant

period of the surface ground in Fig.

and generally speaking the tendency

corresponds to the distribution of

the thickness of soft ground.

The values of the spectral intensities

displayed in Fig. 10 show not only the

regional relation on the surface, but

also the result of amplification of

earthquake motions from the basement

to the surface in the absolute scale.

In this connection, comparing that

the spectral intensity distributed on the surface are about 200, that of

the engineering bedrock was about 82

and that of the pre-Tertiary basement

was about 17. Namely, the spectral

intensity of the earthquake motion

has been amplified 4 times between

the pre-Tertiary basement and the

engineering bedrock, 2.5 times between

the engineering bedrock and the ground

surface, and 10 times between the preTertiary basement and the ground

surface

5. CONCLUDING REMARKS

The unique procedures of seismic

microzonat ion synthesizing each

technique developed in NRCDP , such as the model to estimate the

characteristics of earthquake motions

in the pre-Tertiary basement, the

observation of strong-motions to

estimate the amplification

characteristics of the sedimentary

layers and the database system of

borehole records to evaluate the

characteristics of earthquake motions

on the ground at every 100 meter's

meshes, may provide basic and useful

data for the regional prediction of

seismic damage.

REFFERENCES

1) S.Kishi: Apllication Method of

Columnar Sections of Soil intended

for the Prediction of Seismic Damage

(in Japanese) , Journal of the

Japanese Society of Soil Mechanics

and Foundation' Engineer ing , Vol.22,

No. 3, pp. 152-160, 1982.

2) K.Otani and S.Kinoshita: Spectral

Characteristics of Ground Motion in

the Tokyo Metropolitan Area, The

18th Joint Meeting U.S. -Japan Panel

on Wind and Seismic Effects, 1986.

320
Soil Name Card MT LP CRT

Surface SUR 0

Humus Y 1 Y r

Shell Q 2 Q Q

Pumice P 3 P A

Boulder B 4 B O

Gravel G 5 0 o

Sand

gravel DG 6 0.

•o

• Sand D 7 •

©•

© Fine sand F 8 •

9

9

• Medium

sand M 9 :

9

99 Silt T 10 - mi 1

1

1

1

1

11

1

Clay C 11 = —

Sandstone DN 12 I

.

Siltstone TN 13 IMudstone MN 14 1 =

MT : Magnetic Tape

LP : Line Printer

CRT: Cathode Ray Tube

Fig.1 Table of soil code

Records of

Soil Profile Soil Profile

Boring number

Longitude

Latitude

Height

Depth of bottom

Water lebel

Year/Month

Operator

Depth of layer

Soil name

Soil color

Inclusion

Soil name

Condition

Soil name

Condition

Soil name

Condition

Gravel size

Depth of test

Blows/30cm

1

i

i

1

•

»

I

^Man^

° Boring Site

• Designated Point

Fig. 2 Data format in computer

Fig. 3 Man-machine interaction

system

321
A B C D

pnu M

IIE

. C.053Q

i\\z

1 53.33.51 .0

C 5?26

t-jr-if. 3c.5b

5 5. 3d. 5J..0

1 3y.56.4j.fl

c eioe

533937 893?

35.39.10.0

139.54.51.0

Bl 10

II

II

F.H

G £ 1 0

3 i

I U7 1 1 37

55- -11 -3;.-0

1 J9 .55.19.

i . n0

4 l .10

40.07 41.03

Fig „ 4 Columnar sections of soil at the observation

points of microtremors

322
9

Fig. 5 Predominant period of

observation points of

(eastern neighbor

surface layer and

microtremors

of Tokyo)
(s-]d6) DJ^Dads asuodsaj A\no}af\

Fig.

6

Estimated

zero-damped

velocity

Fig.

7

Average

amplification

characteristics

of

response

spectra

in

the

sedimentary

layer

and

of

surface

layer

pre-Tertiary

basement

at

Urayasu

city

hall
100

J 1 L o o

rr ld

o

325

Fig.

8

Average

amplification

characteristics

Fig.

9

Soil

profile

at

Urayasu

city

hall

of

engineering

bedrock
123‘3 1

' 115*55' 115*36' 11S*3r

Fig. 1

Regional distribution of spectral intensity
Simplified Evaluation Procedure of Seismic Stability

of Embankments on Liquefiable Ground

by

Yasuyuki Koga 1 and Osamu Matsuo2

ABSTRACT

This paper presents a simplified procedure which

evaluates the seismic stability of embankments constructed on sandy liquefiable ground without any complicated analysis. The procedure was made on the basis of 38 field data of the embankments which were damaged or not damaged by the earthquakes in Japan

in the past several decades and the seismic stability

analyses which were performed by the sliding method

considering seismic force. The main concepts of the procedure are “critical N-value”, “safety number” and

“average safety number”. The critical N-value was de- fined as an Af-value of the ground which distinguish the seismic stability of embankment, the safety number as the ratio of the measured V-value to the critical AJ-value,

and the average safety number as the average of the safe- ty numbers calculated from the ground surface to a certain depth. The seismic stability of the embankment

can finally be estimated by comparison of the minimum

average safety number with unity.

Key words: seismic stability, embankment, liquefaction.

1. INTRODUCTION

The major earthquakes that occurred in the past

several decades in Japan caused serious damages to many

embankments on liquefiable sandy ground. With this background, it has recently become necessary that the

seismic safety evaluation of earth structures such as river and road embankments be made. It is desirable that comprehensive stability analyses are applied to each

embankment section on the basis of detailed soil data

obtained from in-situ and laboratory tests. This is, however, almost impossible and impracticable from the viewpoint of cost performance because the total length of

such embankments to be assessed is usually quite long.

Therefore, it is required that a simplified procedure to evaluate the seismic stability of embankments be devel- oped which can roughly select potentially dangerous

sections of the whole embankments.

These kinds of procedures have been proposed to date by several official agencies in Japan 1 ). However, it was considered from examinations that there is yet room

for improvement in them on the mechanical rationality

and the usability.

This paper presents a newly developed simplified

procedure for seismic stability evaluation of embankments.

2. OUTLINE OF THE PROCEDURE

In this chapter, we will briefly describe a developing

process and characteristics of the procedure in order to make it easy to comprehend.

Prior to developing the procedure, the authors considered the characteristics of the procedure as follows. It

is expected that the procedure should be used to that the

seismic stability can be evaluated using only soil data

previously obtained, in order to evaluate point by point over a long embankment. On the other hand, it has been

revealed that earthquake damages to river and road em- bankments is mainly due to the subsurface conditions. 2) Therefore, in order to expect the evaluation accuracy on

the procedure, the subsurface field data are indispensable.There are several types of methods of seismic stability

evaluation of embankments as follows: fl) empirical

method based on field performance data, (2) pseudostatic sliding surface method, and (3) method by

dynamic response analysis. 11 Method (1) tends to be

influenced by the field data which were used in producing the method, and not necessarily to correspond to mechanical rationality. On the contrary, methods (2)

and (3) are not substantially appropriate as a “simplified

method”.

As a result of above consideration, the authors de- cided to make a simplified procedure which is a combination of methods (1) and (2) but not need a seismic

stability analysis.

On the basis of the above mentioned considerations,

the procedure was made according to the steps shown in Fig. 1. Firstly, details of the stability analysis method

were determined so that it can distinguish failed and not

failed embankments as much as possible, by applying the

analysis method to them. Next, a comprehensive para- metric study was performed on hypothetical embankment-ground models with various combinations of em- bankment-ground conditions and design seismicity.

From this study, a “critical A'-value” of sandy layer was summarized for various sandy soil types and design

seismicity considered. This value can be used as a basic index to evaluate whether an embankment overlying on

it is seismically stable or not. Then, the “average safety

number”; a factor to directly evaluate the seismic stabil- ity of an embankment was proposed.

Finally, the evaluation procedure mentioned above was converted to a table for seismic stability evaluation

of embankments in order to make it easy to use.

In conclusion, this procedure is such that all the re- quired information are soil type and jV-value of the subsurface ground and design seismicity.

3.

DEVELOPMENT OF THE PROCEDURE

Developing process of the procedure will be described

in the followings according to the flow chart shown in Fig. 1

1 Head, Soil Dynamics Division, Construction Method and

Equipment Department, Public Works Research Institute, Ministry of Construction, Japan.

2 Senior Research Engineer, Soil Dynamics Division, Construction Method and Equipment Department, Public Works

Research Institute, Ministry of Construction, Japan.

327
3.1 Case Records of Seismically Damaged Embankments

Among case records of seismically damaged embankments during past earthquakes such records were

collected that the damages were clearly described, the

borings data and SPT-./V-value data were obtained, and

also the subsurface ground mainly consists of sandy soil. Total number of the embankments was 38; 3 of the

Niigata Earthquake of 1964, 14 of the Miyagi-ken-oki

Earthquake of 1978, 21 of the Nihon-kai-chubu Earthquake of 1 983.

They all were river embankments with a height of 2 —

7m, and settlements caused by earthquakes ranged from

0 to 2 m.

->.2 Seismic Stability Analysis

3.2.1 Description of Analysis

The circular sliding method considering seismic effect was used in the seismic stability analysis of the embankments. Modified-Fellenius method expressed by effective

stress strength was adopted. In calculating a factor of

safety, the effect of earthquake was taken into account

in two ways. One is that seismic effect is considered as a seismic force (Eq.(l)), the other as a decrease of shear

strength due to a generation of excess pore-water

pressure in a sandy layer of the ground during earthquake (Eq.(2));

Z[c' -us - b) cos

S1 _ I (W- sina+kh -W- y/R)

shear strength parameters = o° and C u —qu /2 = 0.2 + N/40 (kqf/cn?) (Ohsaki’s formula). On the other hand,

the properties of embankment material were given as Tt = 1.8 tf/nf. C = 2.0 tf/n? and <f) = 25° regardless of the

soil type.

Seismic coefficient value k

was determined from

the maximum acceleration at ground surface, asmax

(gal), which were estimated based on the ground motion

records taken in the neighborhood. It follows: = asmax //

k ^ = 0.65 x k s where g is the acceleration of gravity (= 980 gals) and

ks

is the peak (instant) seismic coefficient of ground sur- face.To estimate the dynamic shear strength ratio, R, of

the sandy soils, the two empirical methods, by the Specifications for Highway Bridges in Japan (SHBmethod)4) and by Tokimatsu and Yoshimi (T-Ymethod) 5 ) were used. It is well known that for denser

sand the former method estimates the ratio R smaller

than the latter one. In addition, the excess pore-water

pressure ratio u^/a'y was given by

ud./<*v ~ F.

7 (Fl ^ 1 ) (5) ud /a'v =1 (fl < 1) J where FL

is the liquefaction resistance factor which is a ratio of the shear strength ratio R to the shear stress ratio during earthquake L. 6 )

Fs 2 _ 2 { C' • £+ { W— + • b } cos <» * tan <f>' I (VT sin «)

( 2)

where, Fs : factor of safety

C\ <t>' effective stress shear strength

parameters

W : weight of slice u0 hydrostatic pressure u

excess pore-water pressure generated

by cyclic shear stress during an earthquake

kh horizontal seismic coefficient

£ : length of sliding arc of slice b : width of slice a : angle of sliding arc to the horizontal

plane

y

vertical distance between the center of circle and-the center of gravity

of slice R radius of sliding circle (see Fig. 2)

For each embankment two factors of safety were

calculated by Eqs. (1) and (2), and the smaller one was

finally adopted.

The procedure to compute the factor of safety is shown in Fig. 3. In the figure the following Meyerhof’s

formula was used to relate the N-value to the relative density £>* .

£>* = 21 J N/((Jy + 0 . 7 )

( 3 )

in which o'vis effective vertical stress in kqf/cm2. 3) Mechanical and physical properties of sandy soils were assumed as in Table 1, and those of clayey soils were assumed as follows: unit weight f t =\.btf/rrf ,

3.2.2

Results and Discussions

Fig. 4 shows the relationship between the factors of

safety given by the analyses and the observed settlements

for 38 embankments. Figs, (a) and (b) correspond to the

SHB-method and the T-Y-method respectively.

In either method, there is a general tendency that the

factor of safety Fs decreases as. the settlement D in- creases, although the data plots are considerably scat- tered. When Fs

is equal to Fs2 (see Eq (2)), the value

Fsis small and it can be supposed that there would have

been a build-up of pore-water pressure within the sand

layer, possibly leading to the liquefaction. By comparing

the data of Fs = Fs2 in the two figures it can be seen

that the factors of safety in Fig. 4(a) are smaller than

those in Fig. 4(b). It reflects that the SHB-method of

Fig. 4(a) tends to estimate the dynamic shear strength

ratio R smaller. Now, let us examine quantitatively as to which

method has resulted in a better agreement between the

factors of safety and the settlements of the damaged

embankments. It may well be approved to define “compatible” and “not-compatible” as shown in Fig. 5. In the

figure the horizontal line, £> = £> 0

is supposed to divide

data with significant damages and with insignificant or without damages. Therefore, £) 0 may be called “critical settlement”. Similarly, the vertical line, F<.= Fs 0 , in the figure is supposed to distinguish the two data groups

appropriately; the value of Fs

may be called “critical factor of safety.” It might be appropriate that the value

of D o be determined from such engineering judgment as the maximum settlement under which the damaged em- bankment can keep its original function. On the other

hand, the value of F$0 is usually assumed as Fso = l-0 However, this is reasonable only if the seismic stability

analysis method and parameters applied are considered

as mechanically reasonable. Thus, we should not hold

328
that Fs0 = 1.0 in this case, because the rationality of the

analysis method described above has not been verified. An index parameter expressed as Eq.(6) was used to evaluate and compare the degree of compatibility of the

results as shown in Fig. 5.

Na „ nb Na + Nc NB +ND

( 6 )

where, Ic = index parameter of compatibility,

Nt = number of data plotted in region i. Subscript A,B,C,D = region shown in Fig. 5.

Parameter Ic

is a function of Fs0 and Z)0 , and the

value of Fs o and Do were determined so that the value

of Ic takes a maximum. The results are summarized in Table 2. It is seen from the table that the index para- meter of compatibility I c

is larger in (b) compared with

that in (a), where Dq- 20 cm and Fs0 = 0.8. The resultant Do'val ue of 20 cm is considered to be

reasonable for river and road embankments as a safer value to keep their original functions. In addition, the

resultant Fs0 -value of 0.8 coincide well with a result by

the authors 7) that the critical factor of safety computed

by Eq.(2) is smaller than unity. Also shown in a right

column of Table 2 is a percentage of compatibility pr that is commonly used ans is given by:

From the model studies a set of critical .V -values was compiled and its sensitivity to such factors was ex- amined.

3.3.2 Results

Only the final result is described below as the process

to derive the critical .V-value was rather complicated.

For the standard model with an embankment height of

4m and an ground water level of lm below the surface

as in Fig. 6, the critical ADvalue was given approximately

as:

, C - k s ~ D x 2 NCr = (3.03 + 0.37 • z) x ——- (8) r v .4 • k s - B

in which Z is a depth from ground surface and A, B, C

and D are constants relevant to soil type (exact values

of the constants will be shown in Table 3). Fig. 7 illustrates an example of the results. It is evident in the figure that the critical A'-value increases

with design seismicity , depth Z and coarser grained

sandy soil. As for another factors, it was confirmed that

as the embankment height and the depth of ground

water level increases the critical A/-value increases: however, they were not so sensitive as the above factors.

3.4 Average Safety Number

p, (I) na + nb +nc + nd

In this case, this Pc was not directly used as a measure to compare the propriety of the two methods

described above. It is because the case records studied

herein are not necessarily a subset of the population by

a random sampling but a subset which has a deviation

towards the set of significantly damaged embankments.

Therefore, it was presumed that, if we use Eq. (7) to compare the proprieties of the two methods, it might

lead to a statistically wrong judgment. On the contrary

Eq. (6) would lead to a correct judgment even in this case. In this point of view, it is concluded from the

result shown in Table 2 that the T-Y-methos is more

adequate than the SHB-method to evaluate the dynamic

shear strength ratio R of sandy soil, and only the T-Y- method will be used below.

3.3 Critical N-value

In this part a “critical jV-value” is proposed, which

will be a basic factor to judge the seismic stability of an embankment by compared with a measure N-value of

the sandy deposit at a depth. The critical 7V-value of the sandy deposit is determined corresponding to the soil type, the design earthquake intensity and the depth, and

if a measured .V-value is lower than the critical N-value

the deposit is considered to make the overlying embankment seismically unsafe, and vice versa.

3.3.1 Model Studies

In order to make up the critical N-value, a set of

model studies were performed with the hypothetical

embankment-ground model illustrated in Fig. 6. The

studies consisted of more than a hundred cases of

seismic stability analyses with the model and various

combinations of factors shown in Fig. 6 which are known. to affect the seismic stability of embankment.

The critical N-value was derived from the above

studies under the condition that the soil type of the ground was uniform over the whole depth. But the

actual ground consists of various soil layers. Therefore,

it is not necessarily easy to judge the seismic stability of an embankment only by comparing measured N-value

and critical N-value at each depth. Fig. 8 illustrates this. In the case of (a) and (b) in the figure it is easy to judge

as “safe” and “unsafe” respectively, because the

measured A/-values are greater or smaller than the critical N-values over the whole depth. On the other hand, it is difficult in the case of (c) to judge, although it is a com- mon case. Thus, in order to enable to judge the seismic

stability of an embankment in a quantitative manner

even in such case as (c), an “average safety number” was

contrived as a measure for judgment.

3.4.1 Definitions

The safety number Ns and the average safety number

Ns were defined as: Ns =N/NCr (9) Ws =

( Ns )/z (10)

3.4.2 Studies

A comprehensive studies were performed on how to determine the average safety number ,VS as defined by

Eq. (10), considering such factors as:

( 1 ) The maximum of safety number Ns (2) Safety number Ns of non-sandy soils such as

silt and clay ( ; for such soils critical N-value

,Ncr, could not be determined because they are not susceptible to liquefaction)

(3) Depth within which the safety number is com

sidered

(4) Corrections for the embankment height and the ground water level

329
The manners to calculate Ns were examined for 17

cases of combinations of these factors. Thus, the best way to calculate Ns so that Ns can represent a factor

of safety computed by the stability analysis was found

out. Comparisons of the cases were made based on

Eq. (6).

3.4.3 Results

As a result, the best way to calculate Ns was given as follows:

(1) The value Ns Should be calculated as an arith- metic mean of Ns of every 1m below water table up to the depth where Ns gets minimum. This is given by:

=

C( ^ Ars )/z] Ql) 6 1=1

(2) For silt or clay, Ns should be equal to 2.0 re- gardless of its N-value.

(3) The upper limit of Ns for sandy soil should be

3.0. _

(4) By using Ns as calculated as above, the seismic

stability of an embankment will be judged as follows:(Ns ) ml n < 0.8 : “unsafe”

0.

8 ^ (Ms)min < 1-0 : “possibly unsafe”

i a < t \r \ : “safe”

1.

u s v - Jrnm

Figs. 9 and 10 show the relationship between the

average safety number and the factor of safety and that between the average safety number and the observed

settlement, respectively. In both figures also shown is the percentage of compatibility Pc as calculated by Eq.

(7). From these results it can be concluded that the

average safety number calculated by the method de- scribed above is in good consistency with the factor of

safety and the observed settlement from practical point

of view.

4. SIMPLIFIED EVALUATION TABLE OF

SEISMIC STABILITY OF EMBANKMENT

For the purpose of making it easy to apply the simplified method to evaluate the seismic stability of embankment described in the previous chapter, Table 3 was

provided. This table is so designed that by filling in the

banks in the order of the number we can automatically

get the result. In addition, it whould be noted that the

conditions of application described in the upper part of

the table must be kept, which were duduced in the

developing process of the method.

5. SUMMARY

A simplified procedure to evaluate the seismic stabili- ty of embankments was proposed on the basis of comprehensive seismic stability analyses of embankments

that experienced severe earthquakes in the past. The

procedure is such that the seismic stability can be

evaluated by judging whether the minimum average

safety number is greater or smaller than unity. Finall>

it was shown that the judgments by using the procedure

agreed fairly well with the past field performance data.

References

1) Sasaki, Y., Koga, Y. and Taniguchi, E.: Earthquake

resistant design of earth structures, Technical Note oi Soil Dynamics Division No. 34, Soil Dynamics Divi- sion, Public Works Research Institute, April, 1984.

2) Public Works Research Institute: Report on the dis- aster caused by the Nihon-kai-chubu earthquake ot 1983, Report of PWRI, Vol. 165, March, 1985 (in

Japanese).

3) Meyerhof, G.G.: Discussion, Proc. 4th ICSMFE, Vol

3, p. 1 10, 1957.

4) Japan Road Association: Specifications for Highway

Bridges, Part V, Earthquake Resistant Design, pp

16~20 and pp. 69~75, May, 1980 (in Japanese).

5) Tokimatsu, K. and Yoshimi. Y.: Empirical correlation

of soil liquefaction based on SPT N-value and fines content, Soils and Foundations, Vol. 23, No. 4, pp

56~74, JSSMFE, 1983.

6) Japan Road Association: Technical Guideline for Designing of Common Utility Duct, 1986 (in Japa

nese).

7) Koga, Y. and Matsuo, O.: Considerations on dynamic

strengths of soil used in seismic stability analysis

Proceedings of the 21st Annual Convention of

JSSMFE, pp. 867~870, June, 1986 (in Japanese).

330
Table 1 Properties for Sandy Soils

(a) Mechanical properties

SPT N-value 0'(°) c* (tf/m 2

) N< 10 30 0

10< N< 30 35 0

30 < N 40 0

(b) Physical properties

Soil type 7r (tf/m 3

) 7sat (tf/m 3) D so (mm)

Sandy silt 1.6 1.8 0.04

Fine grained

sand

1.75 1.95 0.15

Medium

sand

1.8 2.0 0.35

Coarse

grained sand

1.8 2.0 0.6

Table 2 Summary of Compatibilities

Evaluation method of dynamic

shear strength ratio R

Do (cm) Fso

Na Nb

Ic * Rc ** Na+Nc Nb+Nd

(a) “Specifications for Highway

Bridges” method 30 0.2

116

0.49 0.71

1 1 + 10 16+1

(b) Tokimatsu-Yoshimi method 20 0.8

11 16

0.53 0.71

1 1+3 16+8

*) seeEq.(6), **) see Eq. (7)

331
Table 3 Table for Simplified Evaluation of Seismic Stability of Embankment on Sandy Deposit

Site:

i Max. Seismic Coef. Ks = Embankment Hight H = (m) Ground Water Depth Hw = (m)

< Conditions of Application > (1) The ground mainly consists of sandy deposits.

(2) Max. seismic coefficient of ground surface, Ks, is 0.1 ~ 0.3. (3) Embankment height, H, is 2 ~ 6m.

(4) Ground water level below ground surface, Hw, is 0 ~ 2m.

Depth

from

G.L. (m)

2 Depth

from W.L. L (m)

3

Soil Type

4Standardized

Critical N-value (Nct )

5Correction Factor6 Corrected

Critical N-value

7Measured

N-value (N)

8Safety

Number

(Ns)

OAverage Safety

Number

(Ns)

1 3.4

2 3.8

3 4.1

4 4.5

5 4.9

6 5.2

7 5.6

8 6.0

9 6.4

10 6.7 < Remarks > t

2

3

Ks = Asmax/g (Asmax • max. ground acc., g: acc. of gravity)

Kh = 0.65 x Ks (Kh: design seismic coefficient)

L is to be measured from the ground water level Soil types and the corresponding mean grain size (D so ) are as follows:

Soil type Clay, silt Sandy Silt Silty fine-grained sand Fine sand Medium sand Coarse sand

D50 (mm) 0.04 0.07 0.15 0.35 0.60

5 Correction factor Sandy silt

Silty fine-grained sand : (

Fine sand

Medium sand

Coarse sand

65Ks

4 1 Ks — 5.3'

. 72Ks - 10.4, 41 Ks - 4.3 ;

. 8 5 Ks - 10.5. 41 Ks — 2.7'

.1 38Ks - 6.3. 4 1 Ks + 6.6' 244Ks - 1.7. 4 1 Ks + 24.3

f

=

) 2 =

)2 =

) 2 =

) 2 =

6 (Corrected Critical N-value) = (Standardized N-value) x (Correction Factor)

8 (Safety Number Ns) = (Measured N-value)/(Corrected Critical N-value). For sandy soils, if calculated Ns is greater than 3.0, Ns should be equal to 3.0. For clay and silt (except for sandy silt). Ns = 2.0 For depth above the ground water level, calculation of Ns is not needed. Ns =i (Ns ) "b L

10 Minimum of average safety number : Ns

‘ = min (Ns)

L

10Minimum of Average Safety Number

Ns

' Judgment Final Result 1

Ns’ < 0.8 Unstable

Sr- nzf> 0.8 < Ns' < 1.0 Possibly unstable

1.0 < N s

' Stable

332
Fig. 1 Flow Chart of Producing Simplified Procedure to Evaluate

Seismic Stability of Embankment

333
Fig. 2 Key Sketch for the Circular Sliding Method in the

Seismic Stability Analysis

Fig. 3 Flow Chart to Compute Factor of Safety

334
Observed

settlement

:D

(m)

Observed

Settlement

!

D

(m)

(a) Dynamic shear strength ratio R of sandy soil evaluated by the

"Specifications for Highway Bridges method" was applied

fb) Dynamic shear strength ratio R of sandy soil evaluated by the Tokimatsu-Yoshimi method" was applied

Fig. 4 Relationships Between Observed Settlements and Computed

Factors of Safety

335
Fig. 5 Key Sketch for Evaluating Degree of Compatibility Between

Observed Settlements and Factors of Safety

J~77?77F77F

Uniform

ground

Sandy silt Silty fine-grained sand

Fine grained sand

Medium grained sand

Coarse grained sand

Ground water level IG.L. — — 1 , — 3 , — 5m

Relative density i Dr =20, 40, 60, 80%

'r

cf) The underlined indicates the standard value in a set of calculation.

Fig. 6 Hypothetical Analysis Model

336
Critical N-value I Ncr

0 5 10 10 20

Fig. 7 Illustration of Critical N-value

N- value N- value

10 20 30

N- value

a) iV>iVcr

through the depth

b) N< Ncr (c) N^Ncr

through the depth

Fig. 8 Examples on Comparison of Observed and Critical N-value

337
Observed

settlement

-

Fig. 9 Correlation Between Minimum Average Safety Number and

Factor of Safety

Fig. 10 Correlation Between Minimum Average Safety Number

and Observed Settlement

338
Strong Motion Observation System of PWRI

Toshio IWASAKI*, Yasushi SASAKI** and Tetsuro KUWABARA***

SUMMARY

This report presents the Strong Motion

Observation System of PWRI (Public Works

Research Institute) as a continued report of

the PWRI's Dense Instrument Array Observation

System previously presented at the 18th Joint

Meeting

This report will cover the observation

stations where PWRI is collecting data, data

base system in which strong motion

observation data collected up to now are recorded, digitizing method for analog

records, and the situation of data

utilization up to now.

KEY WORDS: Civil Engineering Structure,

Earthquake Resistant Design, Strong Motion

Observation, Strong Earthquake Motion Record,

Data Base

1. INTRODUCTION

Strong motion acceler ographs have been

installed on civil engineering structures

related to the Ministry of Construction at

291 places and the total number of accelerograph is 698 as of March, 1987. Strong

motion earthquake records of 7,992 components

from 669 earthquakes have been obtained after

1965 when the real strong motion observation was started. Among these records, 1,102

components have been digitized and being used

for various kinds of analyses.

The strong motion observation means, in

a narrow sense, the measurements of the

vibration waveforms and amplitudes (normally

acceleration) of strong earthquakes by means

of strong motion accelerographs installed on the ground and structures and also the

primary analyses of the characteristics of

the magnitude of their acceleration amplitude

and vibration period. In a broad sense, the

strong motion observation will include

secondary analyses in which correlations are examined among the earthquake motion

characteristics such as the characteristics

of magnitude of acceleration amplitude and

vibration period obtained from primary

analyses, the characteristics of earthquake

mechanism including the magnitude of

earthquake and focal depth, the propagation

passage characteristics such as epicentral

distance, topography and ground conditions,

and the structural characteristics by type

and scale of structures and by kind of

structural materials. In a country like

Japan where the earthquake activities are

very vigorous, the strong motion observation

will greatly contribute to the rationalization of the earthquake resistant design of

civil engineering structures, to the safe and

economical construction of ever-diversifying

and scale-increasing structures and also to

supply basic data for formulating earthquake

disaster prevention measures by region and by

topography and ground conditions.

This report will present PWRI's Strong

Motion Observation System as a continued

report of "Dense Instrument Array System by

PWRI for Observing Strong Earthquake Motion"

previously presented at the 18th Joint

Meeting. This report will cover the

observation stations where PWRI is collecting

data, data base system in which strong motion

observation data collected up to now are recorded, digitizing method for analog

records, and the situation of data

utilization up to now.

2. PR0GRESS0F STRONG MOTION OBSERVATION BY

THE MINISTRY OF CONSTRUCTION

The Strong Motion Observation by the

Ministry of Construction is carried out by

PWRI for civil engineering structures and

also by Building Research Institute for

buildings respectively. Fig. 1 shows the

outline of PWRI's Strong Motion Observation

System. As the strong motion observation for

civil engineering structures, a SMAC-A type

strong motion accelerograph was first

installed at the Sarutani Dam within the

jurisdiction of the Kinki Regional

Construction Bureau in 1957. And the strong

motion accelerographs were installed

thereafter only at 10 places until 1963.

However, the acceleration records obtained by

a strong motion accelerograph installed in an apartment house in Niigata City were greatly

useful for investigating the earthquake

damages caused by the Niigata Earthquake

occurred in June 1964. Thereafter, the

necessity of strong motion observation was widely recognized by the public. Then,

discussions were made between the river

Bureau and Road Bureau of the Ministry of

Construction and PWRI on the installation of

accelerographs, and a plan for installing

SMAC type strong motion accelerographs was formulated. This plan has been steadily

carried out by the Ministry of Construction,

Regional Construction Bureaus, Hokkaido

Development Bureau, Public Corporations,

local self-governing bodies, and PWRI. And

Strong Motion Observation is being performed

at 291 civil engineering facilities

throughout the nation in March 1987.

* Dr. Eng., Director, Earthquake Disaster

^Prevention Department, PWRI, MOC

Dr. Eng., Head, Ground Vibration Division, Earthquake Disaster Prevention

**,*Dept. , PWRI, MOC

Research Engineer, Ground Vibration

Division, Earthquake Disaster Prevention

Dept., PWRI, MOC

339
Fig, 2 shows the progress of the number

of places where accelerographs were installed

by type of facility. The number of places

where accelerographs were installed increased

after the Niigata Earthquake in 1964. In the

decade from 1965 to 1975, the strong motion

accelerographs were installed mainly on bridges among road facilities and more slowly

on dams. In the decade from 1975 to 1985,

the installation of strong motion accelerographs on road facilities was made more

slowly compared to the previous decade, but

the number of accelerographs installed on

river facilities increased after the Miyagiken-oki Earthquake in 1978. In addition,

after the establishment of the Construction

Technology Evaluation System in 1981, "the

Development of Newly Developed MOC Type

Strong Motion Accelerograph for Measuring the

Acceleration of Earthquake Motion" was

carried out, and many MOC type strong motion

accelerographs were installed on dam

facilities and road facilities after 1982.

After 1975, the average annual increase in

the number of places where accelerographs

were installed was 14 and the average annual

increase in the number of accelerographs

installed was 33.

3. PRESENT SITUATION OF STRONG MOTION OBSERV ATION

3. 1 Outline of Observation Stations

Table 1 shows the number of observation

stations and the number of strong motion

accelerographs (or number of components) by

facility and by the type of accelerograph.

And Fig. 3 shows the stations where the

strong motion observation is being performed

by road, river and dam facilities as of March

1987. Proportion of the observation stations

by facility is 57% for road facilities, 12%

for river facilities, 29% for dam facilities,

and 2% for sewerage facilities. And the

proportion of observation stations by type of

strong motion accelerographs is 61% for SMAC

type, 23% for electromagnetic type and 16%

for MOC type.

3.2 Maintenance and Inspection of Strong

Motion Accelerographs

Public Works Research Institute (PWRI)

performs periodic inspection and calibration

once a year for strong motion accelerographs

installed by the Ministry of Construction and

a part of strong motion accelerographs

installed by prefectures. In FY 1986, PWRI

inspected and calibrated 169 SMAC type strong

motion accelerographs installed in 83 places,

186 components of electromagnetic type strong

motion accelerographs installed in 22 places,

and 33 MOC Type strong motion accelerographs

installed in 13 places.. By the adjustment

and replacement of parts, it became possible

to regain the normal conditions for more than

80% of SMAC type strong motion

accelerographs. And among the main items of

adjustment and replacement the replacements

of consumables such as silica gel and

batteries are more than 50%.

3.3

Digitization of Strong Motion Records

After the occurrence of an earthquake,

the person in charge of strong motion

observation at each observation station will

confirm the presence of strong motion records

and, if records have been obtained, will

recover and mail the records to PWRI.

The records obtained from SMAC type

strong motion accelerographs will be then

digitized by the procedure shown in Fig. 4. The digitization is performed by reading the

earthquake motion records and time records

with a digitizer. Records made as digital

values are subjected to Zero-Line correction,

Arc correction and Time axis correction and

then are turned into data with the intervals

of 1/100 second. Then, the Fourier

transformed data are applied to band-pass

filters determined based on the accuracy of

digitization and instrument character-istics.

That is, if the acceleration record by strong

motion accelerograph is a 0 (t) , then its

Fourier transform F 0 (l)can be given by

F 0 ( / ) =

'COa 0 (t) • e~

2

" Mt • dt (3.1)

Then, corrected acceleration a(t) based on filter Fc(f) considering the accuracy of

digitization and instrument characteristics

of strong motion accelerograph can be given

by the following formula:

a (t)

where,

aoF„ ( / )

• F c ( 1 ) U

e

2 " ‘ *

• d f (3.2)

0

I - I 1

exp ( i <t> ( I ))

I ,«-/ .. R( M

r , f

, exp ( i 4> ( / )

F ‘ ( 1 >* R( 7 )

/ „n— I exp( i [ I )

R( S )

0

OS / < / i.

I I I S / is I I s

I i S 1 u

f u i s ^ s Lu

f > f 111

(3.3) where, / , u and f u , are cut-off frequencies

at low frequency side and high frequency side

respectively, and f tl and / uu are the frequencies when the magnification of correction

filter corresponding to { lu and J u , respectively becomes zero. Also, for correction

factors R ( / ) and <t> ( / )for amplitude and phase,

the following equations are used by

considering the effect of the air damper

employed in SMAC type strong motion accelerograph working as spring.

R( / )«

1 4- (U/N) 2

/(l-U 2

) 2 + { (l-U 2 )U/N+2hU> 2 (3.4)

<f> ( / ) »tan" 1 r 2MJ

( 1-U 2

) { 1+ (U/N) 2 H- 2MJ U/N

(3.5)

340
In the above equations, U=///„,N=/ a // n ,

( n is the natural frequency of Dendulura,

and / „ is constant. Fig. 5 shows 1 /R( /) and

0(f). f (U = l/3Hz and f u , = 12Hz are used here

as cut-off fruencies by taking account of the

accuracy of digitization and the accuracy of

strong motion accelerograph.

3.4 Present Situation of Strong Motion Observation Records

As stated previously, strong motion

records of 7,992 components have been

obtained from 669 earthquakes occurred after

1965 when the Strong Motion Observation was actually started, and 1,102 components of

them have been digitized. Fig. 6 shows the

annual progress of the number of recorded and

digitized components and the number of

earthquakes observed. In 1968, records were obtained from many earthquakes such as the

Matsushiro Earthquake Swarm, Ebino

Earthquake, Hyuganada Earthquake, and

Tokachi-oki Earthquake. Also in 1978 and

1983, the number of earthquake records

increased as a result of the occurrence of

the Miyagi-ken-oki Earthquake and the

Nihonkai-chubu Earthquake. After 1975, the

annual average number of earthquakes observed

is 20, from which 97 records or 291

components were obtained, and 16 records or

48 components of them were digitized.

Fig. 7 shows the number of record

components classified by magnitude,

epicentral distance and maximum acceleration

for 1,007 components for which particular

earthquakes were identified among 1,102

digitized components. From this figure, it can be known that 115 components, which is

52% of 223 components with the maximum

acceleration of greater than 100 gal, were

obtained at the places with the epicentral

distance of less than 50 km where the

magnitude was 5.0 ^ 2 6.9 and that fewer

records were obtained at the places with

epicentral distance of less than 100 km where

the magnitude was M^7.0.

The largest maximum acceleration on the

ground is 616 gal recorded by a SMAC-B2 type

strong motion accelerograph installed on the

ground near the Itajima Bridge with an epicentral distance of 20 km during an

earthquake (M=6.6, D=20 km) occurred near the

coast of the western part of Ehime Prefecture

in 1986. And the largest maximum

acceleration on the structure is 802 gal

recorded by a SMAC-B2 type strong motion

accelerograph installed on the pier of the

Kaihoku Bridge with the epicentral distance

of 85 km during the Miyagi Offshore

Earthquake ((M=6.7, D=50 km) in February

1978.

Among these digitized records, the

records larger than 30 gal on the ground and

larger than 100 gal on the structure are

published as Technical Note of PWRI and as of

March 1987 12 volumes were issued.

4. STRONG MOTION OBSERVATION DATA BASE SYSTEM

4.1 Strong Motion Observation Data Base

Strong motion observation data is

classified into three categories, data

related to observation stations, data related

to observed earthquakes and data related to

strong motion records. And it is also

classified into two categories based on data

type, numerical or character data and figure

data. For the purpose of retrieving these

data quickly, Strong Motion Observation Data

Base was developed in the Optical Disc Filing

System (0LFIS) introduced by PWRI in FY 1985.

This system is built in the large computer in

PWRI, and numerical and character data are

registered as relational data base, and data

retrieval, addition and updating can be

easily made by using software such as RDB1

(Relational Data Base System), ACE3 (End User

Language) and 0LFIS. Outline of 0LFIS is

shown in Fig. 8. Strong Motion Observation Data Base

comprises 7 kinds of numerical and character

data tables and 4 kinds of graphic data.

Various kinds of retrievals are possible for

numerical and character data, and those shown

in Fig. 7 in the foregoing Chapter are the

results of retrieval with plural conditions

attached to the Digitized Strong Motion

Records Table and Earthquake Items Table.

Data can be easily added to the data base

from TSS terminal in the case of a small

amount of data and by batch job using 0LFIS

software in the case of a large amount of

data. And data update can be made on display

of TSS terminal.

The following is outline of numerical

and character tables.

1) Earthquake Items Table

Earthquake items determined by Japan

Meteorological Agency (J.M.A.), for which

strong motion records were obtained, are registered

2) Original Strong Motion Records Items Table

Maximum acceleration values directly

read out from strong motion records are

registered.

3) Digitized Strong Motion Records Items

Table

Maximum acceleration values after

digitized and corrected are registered.

4) A Large Amount of Acceleration Values

Table

All acceleration values after digitized

and corrected are registered.

5) Strong Motion Observation Station Items

Table

Data such as strong motion accelerograph items and management body items at

observation stations are registered.

6) J.M.A. Observation Station Items Table

J.M.A. observation station items are

341
registered

7) J.M.A. Seismic Intensity Table

J.M.A. seismic intensities, for the

earthquake strong motion records were

obtained, are registered.

4.2 StronR Motion Observation Data Retrieval and Graphic Processing System

Strong Motion Observation Data Retrieval

and Graphic Processing System (SMDB86) was

developed in order to retrieve data in Strong

Motion Observation Data Base and to quickly

and easily obtain list output and graphic

output. Fig. 9 indicates the outline of the

proces-sing system. This system is for

producing graph and table outputs in

interactive mode from graphic display

terminals and has the following four

functions

1) Strong Motion Observation Station Retrieval

and Plotting Module

This has the function of producing the

location maps of strong motion observation

stations and the lists of strong motion

observation points near epicenter in order

to identify observation stations near the

epicenter and to confirm where to make

contact after the occurrence of earthquake.

2) Prompt Report Preparation Module on Strong

Motion Observation Records

This has the function of producing table

of prompt Report on Strong-Motion Records,

location figure of strong motion

observation stations obtained records

shown in Fig. 10 and figure of J.M.A.

seismic intensity distribution. This

module works about one week after the

occurrence of earthquake by using seismic

intensity in each area and the maximum

value read out from strong motion records

transmitted.

3) Primary Analysis Module for Strong Motion

Records

This has the function of performing

retrieval by earthquake or by observation

station by using data in the strong motion

observation data base.

a) Retrieval by earthquake No.

When an earthquake No. is input, the

lists of earthquake, record, and

observation station items can be

obtained. In addition, location figure

of strong motion observation stations

obtained records, figure of J.M.A.

seismic intensity distribution, figure

of strong motion acceleration and figure

of acceleration response spectrum ratio are prepared.

b) Retrieval by observation station No.

When an observation station No. is

input, the lists of observation station,

record and earthquake items can be

obtained. In addition, figure of

observed earthquake epicenters as shown

in Fig. 11 can be obtained. Another

function as same as the retrieval stated

in Para, a) is also available.

4)

Technical Note Table and Figure Preparation Module

This has the function of preparing the

lists of earthquake, record and

observation station items as well as location figure of strong motion

observation stations obtained records,

figure of J.M.A. seismic intensity distribution, figure of strong motion acceleration, figure of acceleration response

spectrum ratio, and list of digitized and

corrected acceleration values by means of

batch job after the input of earthquake

occurrence term or earthquake No.

5. Situation of Utilizing Strong Motion

Records

Strong motion records, for which digitization and instrument correction were

completed, are individually utilized for

input earthquake motions during the design of

civil engineering structures or examination

of analysis method by comparing analyzed

values to observed values or input earthquake

motions during shaking table tests. Moreover,

numerous strong motion records are being

utilized for using by statistical analyses,

the relations between maximum acceleration,

duration of earthquake motions or repetition

characteristics and magnitude of earthquake,

epicentral distance or ground condition.

Achievements obtained from the above are

reflected upon "Specifications for High-way

Bridges, Part V, Earthquake Resistant Design"

and other various standards and guides. The

following is some of the results obtained

from statistical analyses.

5.1 Development of attenuation formula of

maximum horizontal acceleration

Attenuation formulas by magnitude,

epicentral distance and ground condition were determined by using 197 sets of horizontal

acceleration records of strong motions which

were observed at 67 strong motion observation

stations (on the ground) located across the

nation and operated by the Ministry of

Construction and the Ministry of Transport.

a; c

( 987 .4 X 10°

2 1 6M

232 .5 X 10°

31 3M '

403 .8X 10°' 26SM

365 .3 X 10° ’ 268M X

(Group 1) X ( A +30)*' - 218 (Group 2) (Group 3)

( A + 30)*'

1 82 (A1 1 ground)

(5.1)

where

a max: Presumed value of maximum

horizontal acceleration (gal)

M : J.M.A. Mgnitude

A : Epicentral distance (km)

Group 1 : G.C. Group 1 set forth in

"Specifications for High-way

Bridges, Part V, Earthquake

Resistant Design"

342
Group 2 : G.C. Group 2 & 3 set forth in

"Earthquake Resistant Design"

Group 3 : G.C. Group 4 set forth in

"Earthquake Resistant Design"

5.2 Development of attenuation formula of

maximum vertical acceleration

Attenuation formulas by magnitude,

epicentral distance and ground condition were obtained by using the vertical acceleration

of strong motion records from 119 components

observed at 53 observation points located

across the nation and operated by the

Ministry of Construction and the Ministry of

Transport.

' 1 17 .0 X 10°

2S8M (Group 1) 88.2X 10°’ 297M 'X ( A + 30)" 1 " l9 ° (Group 2) 13.5X io°- 402M

, (Group 3)

58 .8x 10° • 292M X ( A+30)-‘ •” s(All ground)

(5.2)

where, all the symbols are the same as equation (5.1)

5.3 Development of attenuation equation of

acceleration response spectrum

It is known that the shape of

acceleration response spectrum greatly varies

depending on the kind of ground. Shown in

Fig. 12 a) and b), the acceleration response

spectrum ratio diagram of horizontal

acceleration records obtained on the Ground

Condition Group 1 (G.C. Group 1 set forth in

Specifications for High-way Bridges) and

Ground Condition Group 3 (G.C. Group 4 set

forth in Specification for High-way Bridges).

By using horizontal acceleration of 20 197

sets of strong motion records explained in

Paragraph 5.1, PWRI has developed equations

for estimating horizontal acceleration

response spectrum by magnitude, epicentral

distance, ground condition and natural

period

S a (M, A ,T* ,GC, ,h=0 .05) = <2 (T* ,GCi ) x 10 l,< T *- CCi ' "

X(A+30)-'-‘ 79 (5.3)

where

a(Tk, GCi), b(Tk, GCi): Coefficients

each Natural Period and each

Ground Condition

Sa : Estimated value of horizontal

acceleration response spectrum

(gal)

M : J.M.A. Magnitude

A : Epicentral distance (km)

GCi : Ground Condition (same as that

of equation (5.1))

When estimating a response spectrum with the

damping factor of other than 0.05, the

following equation is used:

Sa (M, A ,T* ,GCj ,h) *S a (M, A ,T* ,GC, ,h=0 .05)

6. SUMMARY

Main items explained in the foregoing

chapters will be summarized below.

1) Twenty-two years have passed after the

real start of the Strong Motion Observation

and, as of March 1987, observation by 698

strong motion accelerographs is being carried

out at civil engineering facilities operated

by the Ministry of Construction in 291

stations across the country.

2) After 1975, increases in the number of

strong motion observation stations and in the

number of strong motion accelerographs are

respectively 14 and 33 in average per year.

3) Proportion of the number of observation

station by facility to the whole is 57% for

road facilities, 12% for river facilities,

29% for dam facilities and 2% for sewerage

facilities. And the proportion of the number

of accelerographs by type to the whole is 61%

for SMAC type, 23% for electromagnetic type

and 16% for newly developed MOC type.

4) In FY 1986, periodic inspection and

calibration once a year were performed for

169 SMAC type strong motion accelerographs in

83 places, for 186 components of

electromagnetic type strong motion

accelerographs in 22 places and 33 MOC type

strong motion accelerographs in 13 places.

5) From 1965 to December 1986, the records

of 7,992 components were obtained from 669

earthquakes, and digitization was made for

1,102 components.

6) From annual average number of earthquakes

of 20 after 1975, 97 records or 291

components have been obtained, and about 16%

of them or 16 records or 48 components have

been digitized.

7) Number of digitized strong motion records

observed within the epicentral distance of

100 km of earthquakes with M^7.0 is small.

8) The largest maximum horizontal

acceleration on the ground is 616 gal, and on

the structure is 802 gal.

9) Diversified data retrieval and graphic

processing has become possible as a result of

the arrangement of Strong Motion Observation

Data Base and the development of Strong

Motion Observation Data Retrieval and Graphic

Processing System (SMDB86).

10) By utilizing the strong motion records,

many research achievements have been realized

such as the development of formulas for

estimating the maximum acceleration and

acceleration response spectrum.

343
ACKNOWLEDGMENT: Japanese), Technical Memorandum of Public

Works Research Institute, No* 2038, December

Persons in charge of strong motion observe- 1983.

tion who work in the Regional Construction

Bureaus of the Ministry of Construction,

Hokkaido Development Bureau, various Public

Corporations, local self-governing bodies and

so forth have greatly contributed to the

strong motion observation. The authors would

like to express here the deepest gratitude to these persons for their steady efforts made

up to now.

REFERENCES

Chapter 1 1) E. Kuribayashi, T. Iwasaki and S. Wakabayashi: Strong Motion Observation and

the Utilization of Its Records (in Japanese),

Civil Engineering Journal, Vol. 15, No.l,

pp28-33, Jan. 1972.

Chapter 2 2) Public Works Research Institute, Ministry

of Construction: Strong Motion Observation

for Civil Engineering Structures (in

Japanese), Technical Memorandum of Public

Works Research Institute, No. 1847, July

1982.

Chapter 3

3) Public Works Research Institute, Ministry

of Construction: Methods of Estimating

Maximum Earthquake Motions and Earthquaxe

Response Spectra (Part 4)—Re-examination of

Attenuation Formulas for Maximum Acceleration,

Velocity, Displacement, and Acceleration

Response Spectrura--(in Japanese), Technical

Memorandum of Public Works Research, No. 1993,

March 1983.

4) K. Kawashima, Y. Takagi and K. Aizdwa:

Accuracy of Digitization of Strong-Motion

Records Obtained by SMAC-Accelerograph (in

Japanese), Proc. of Japan Society Civil

Engineers, No. 323, pp67-75, 1982.

5) K. Kawashima, Y. Takagi and K. Aizawa:

Procedure of Instrument Correction and

Displacement Calculation for SMAC-B2

Accele rograph Records with Considering

Accuracy of Digitization (in Japanese), Proc.

of Japan Society Civil Engineers, No. 325,

pp35-44, 1982.

6) System Division, PWRI, Ministry of

Construction: Manual of Optical Disc Filing

System (in Japanese), June 1986

Chapter 5 3) Same as Paper 3) stated above.

7) Public Works Research Institute, Ministry

of Construction: Methods of Estimating

Maximum Earthquake Motions and Earthquake

Response Spectra (part 5)--Developraent of

Correction Formulas Based on Attenuation

Constants for Earthquake Response Acceleration Spectra —(in Japanese), Technical

Memorandum of Public Works Research

Institute, No. 2001, July 1983.

8) Public Works Research Institute, Ministry

of Construction: Methods of Estimating

Maximum Earthquake Motions and Earthquake

Response Spectra (Part 6

) --A t t enua t i on Formulas for Maximum Vertical Acceleration

and Acceleration Response Spectra —(in

344
ig.

1

Outline

of

Strong

Motion

Observation

System

in

Public

Works

Research

Institute
Fig. 2 Annual Transition of the Number of Strong

Motion Observation Site for Each Facility

Table 1 Situation of Strong Motion Accelerograph

Installation for Public Works Facilities

As of March 31, 1987

Accelerograph Type

Facility Number of Site SMAC Type Electromagnetic

Type

Newly Developed

MOC Type

Number of Site Number of Site Number of Site

(Number of (Number of (Number of

Accelerograph) Component) Accelerograph)

Road Facility

(Bridge, Tunnel, Ground us

29 13 where Construction is

arranged)

(242) ( 365) (53)

River Facility

25 (Embankment, Weir, Flood 36

.

5 b Gate

(50) (37) (14)

Dam Facility

34 24 32 28

(53) (255) (59)

Sewerage Facility

(Pipe, Disposition 6

6 (13)

0 0 Facility)

Total

291

178 66 47

(358) (657) (126)

346
Note the Numaer is P W R I Station Number Fig 3 Location of Strong Motion Observation Site

for Each Facility

347
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348
108 Obasrved Earthquake Number

70

- SO

50 1er A

o

Lki

40 »u

l/t -OO

o

30 S

Sz

20

J 10

A.O. 1965 57 69 71 73 75 77 79 31 33 35 37

Matsushlra S»arm

Maior

Earthquake

O

i 3‘

— 5 3 Z Z

Fig. 6 Annual Transition of the Number of Recorded

Component, Digitised Component, and Observed

Earthquake

1 100 qat -

30 qat — 100 qai. ~ 30 qal

Total Number 1,007 Components

Epicentrql Distance A I km I

Fig. 7 Digitized Component Number Calssified by Earthquake Magnitude, Epicentral Distance and Maximum

Acceleration

1 300

— 200

- 100

349
01 FIS

f!0n l Relotionol Oats Bose System

Fig. 8 Outline of Optical Disc Filing System (OLFIS)

Fig. 9 Outline of Strong Motion Observation Data

Retrieval and Graphic Processing System (SMDB 86)

350
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Fig. 10 Location Figure of Strong

Motion Observation Stations Obtained Records

Fig. 11 Figure of Observed

Earthquake Epicenters

a) Ground Condition Group 1 b) Ground Condition Group 3

Fig. 12 Characteristics of Acceleration Response Spectrum

351
.
Evaluation of Economical Losses Induced by the

Nihonkai-Chubu Earthquake of 1983

By

1 2 Toshio Iwasaki , Kazuhiko Kawashima , 3 4 Naomi Qbinata , and Takashi Kanoh

ABSTRACT

Effect of earthquake are not restricted to

damages of structures but it covers much

larger aspects including economic effects.

This paper presents a procedure proposed

aiming to evaluate direct and indirect losses

caused by an earthquake. The decrease of

products was counted from three sources,

i.e., decrease of products due to damage of

facilities for production, due to damage of

transportation facilities, and due to damage

of materials and goods. Effect of investment

for restoration was also considered in the

analysis. Based on direct loss associated

with these decrease of products, second

spread effects were estimated through interindustry relation analysis ( input-output

analysis )

The procedure was applied to the losses by

the Nihonkai-chubu earthquake of 1983, and

the estimated gross prefectural products

agreed well with those actually developed

after the earthquake. It was also concluded

that further improvements are also needed to take account of actual economical activities

immediately after the earthquake.

KEY WORDS: Indirect Losses, Economical

Effects, Nihonkai-chubu Earthquake, Interindustry Relation Analysis

1 . INTRODUCTION

Damages induced by earthquakes have mostly

been evaluated in terms of damages associated

with failure of facilities and/or structures.

Effects of earthqake are , however , not

restricted to damages associated with failure

of facilities but they cover much larger

aspects including economic effects such as decrease of products due to suspension of

production, decrease of consumption due to

low social activities, etc. Losses associated

with such effects are likely to be much

larger than those counted only from failure

of facilities. However, mechanism of

economical losses induced by earthquakes is

so complicated that little is known about

these effects.

This paper presents a procedure proposed for

aiming to evaluate direct and indirect losses

caused by an earthquake as well as an example

of application of the procedure for the

Nihonkai-chubu, Japan, earthquake of 1983.

2. EFFECTS OF AN EARTHQUAKE ON ECONOMICAL

LOSSES

'When an earthquake developed damages at some

of cities, towns and villages, the region

within which damages took plase may be

defined. Economical losses associated with

that earthquake may then be evaluated on the

basic of an amount of decrease of products

within that region. When an amount of

production G (t) within the region after the

earthquake can be defined, G (t) may be

evaluated as

G ( t ) = G (t) - G , ( t ) + G (t) (1

o n d r in which G (t) represents products at the

fiscal year of t within the region, which was supposedly achieved if the earthquake did not

take place, and G^(t) and G (t) represent a decrease of products within- the region

associated with the earthquake damages, and

an increase of products within the region

associated with investments for restoration,

respectively.

Consequently, it may be possible to evaluate

G (t) provided appropriate evaluation of

G°(t), G^(t) and G

r (t) be made. It should be

noted here that provided one prefecture can

be regarded as the region within which

economical losses are to be evaluated, the

products G (t) can be obtained in a form of

gross prefectural products after a couple of

years later than the earthquake. However the

gross prefectural products only describes

final products achieved in the prefecture at

the specific fiscal year, and precise

background informations on economical effects

of earthquake can not be traced. Therefore it

is required to develop appropriate analytical

method with use of a relation presented by

Eq.(1) for identifying economical effects of

the earthquake. It should be noted that the

accuracy of analytical model be checked by

comparing products G (t) predicted by Eq.(1)

with the actual products determined

statistically

There are many aspects which can be regarded

as effects induced by an earthquake, and

economical effects may be classified as shown

in Table 1, in which losses of facilities,

properties, goods and materials can generally

be evaluated by damage survey conducted by

each body of the industry and regional

1) Dr. of Eng., Director, Earthquake

Disaster Prevention Department, Public

Works Research Institute, Ministry of

Construction, Japan

2) Dr. of Eng., Head, Earthquake Engineering

Division, Earthquake Disaster Prevention

Department, Public Works Research

Institute, Japan

3) Research Engineer, Earthquake Engineering

Division, ditto

4) Assistant Research Engineer, Earthquake

Engineering Division, ditto

352
government. Although surveys for secondary

effects caused by spread of direct losses of

facilities, properties, goods and materials

are not generally made, they include ;

1

decrease of products due to damage of

facilities for production

2)

decrease of products due to damage of

transportation facilities including

transportation media

3)

decrease of products due to damage of

materials and goods

Therefore by appropriately evaluating the

decrease of products induced by the ' above

described three types of damages, economical

effects of an earthquake may be identified in

accordance with a flow chart shown in Fig.1.

Reduction factor a , which represents a degree

of reduction of products due to the

earthquake, is defined for each industry as

a = DP / PA ( 2 )

this study.

When the facilities for production and the

number of labor can be considered as independent , Eq.(3) is seperated into two

independent functions as

Y = a +

P K (4) Y = a + 7 K (5)

2) Unit Product Method

When the products per unit facilities for

production or unit number of labor, which are designated here as a unit products, can be

determined, the products after the earthquake

may be evaluated by multiplying the unit

products and facilities for production and/or

number of labor available after the

earthquake. Because unit products are generally time dependent, regression analysis

was made to obtain the unit product at a specific fiscal year concerned.

in which DP and PA represent a decrease of

products due to earthquake and total products

supposedly developed if the earthquake did

not take place, respectively.

3. EVALUATION METHOD FOR ECONOMICAL LOSSES

INDUCED BY AN EARTHQUAKE

3-1 Evaluation Method for Decrease of

Products Associated with Damage of Facilities

for Production

Seismic damage of facilities and/or

properties for production and sales, such as damages of factories, shops, machineries,

public utilities, etc., induces a decrease

and even suspension of production. Such a decrease of production and/or products was assumed here to be estimated by substituting

products P^ actually developed after the

earthquake from products pP which were supposedly developed if the earthquake did

not take place. It was also assumed here that

the products pP actually developed after the

earthquake be evaluated from industrial

structure models. The following two models

were used for such purpose.

1 ) Production Function Model

An amount of products is generally specified

by a number of labors and an amount of

properties for production. Therefore, once an equation representing the products in terms

of labor and facilities for production ,i.e.,

production function model, is formulated, the

products developed after the earthquake may

be evaluated by specifying the nummber of

labors and the amount of facilities for

production available after the earthquake.

The Cobb-Douglas type equation in the form of

Eq.(3) is sometimes used for such purpose,

Y = a K? L

r (3)

in which, Y, K and L represent products,

private capital stock, and number of labor,

respectively, and a , /? and T represent

constant to be determined by regression

^analysis. The amount of facilities for

production was considered as the private

capital stock for the analysis of products in

The priority for selecting the products with

use of the above-described procedures was assumed as an order of Eq.(3), Eq.(4) — Eq.(5)

and unit products method.

3.2 Evaluation Method for Decrease of

Products Associated with Damage of

Transportation Facilities

Seismic damages cf transportation facilities

as well as transportation media cause a decrease of the products. Such a decrease of

products was assumed here to be estimated as

a decrease of trafic volumes which were defined as a differense between the trafic

volumes supposedlly developed if the

earthquake did not take place and those

actually developed after the earthquake. The

reduction factor was determined in

accordance with Eq.(2).

3.3 Evaluation Method for Decrease of

Products Associated with Damage of Materials

and Goods

Decrease of products is also developed due to damages of materials and manufactured or partially manufactured goods for which the

productions depend on.

It should be noted here that an amount of

loss due to damage of materials and goods are generally determined for each industry by

private or regional government bodies.

Therefore, the reduction factor was obtained

by dividing the loss due to damage of

materials and goods by the products given in

the inter-industry relations table ( inputoutput table )

3=4 Evaluated Method of Decrease of

Productions

Decrease of products was analyzed by the

input-output analysis ( analysis of interindustry relations ) with use of the

reduction factors associated with damage of

facilities for production, damage of

transportation facilities, and damage of

materials and goods. It should be noted here

353
that although the reduction factor of

products can be estimated independently from

the three types of damage, the products for

i-th industry are, however, likely to be

influenced by the most critical damage

between the three damages. Therefore,

decrease of products for the i-th industry

JX^ may be evaluated as

JX.

1 X.

1 x max

l

/

( 6 )

in which a . ’, ay? and represent the

reduction factor of produces associated with

the damage of facilities for production, the

damage of transportation facilities, and

damage of materials and goods, respectively.

X. represents the supposed products of i-th

industry unless the earthquake occured.

Besides such decrease of products,

productions of other industries, which did

not suffered damages but depend on the

products of i-th industry, are also

influenced. Such an effect is designated in

this study as a first spread effect, and it was analyzed by an input-output analysis.

The input-output table has a form as shown in

Table-2, in which I. and X. represent the ith industry and the products of i-th

X. . , f. and V. are

ij industry, respectively,

the followings;

X^ . : products used from I. to I

‘J or

f.V

i When an products

of

out

the

of

amounts paid from I. to I. final demand of I.

1 ^ value added of I

amount of a. 1) X. of i-th industry became

use due to earthquake damage, X. ., f. and tf. of the input-output table on i-ttrline change

to (1-«.)X.., (l-cr.)f. and (l-ff.)V.,

respectively1

^ It should fee noted here that even if j-th industry did not suffere damages

from the earthquake, the product of j-th

industry decreases from X. to (1-a.)X..

Simiraly, when an amount of«^ (0sJ a^t) if

the products X^ in k-th industry suffered

damages, it also makes the products of j-th

industry decrease from X . to ( 1 - « ^

)X . Therefore when the decreases of products

arise in industries i and j simultaneously,

both gives influences to the production of jth industry. Although interactions of the

decrease of products between various

industries are significantly complex, it was assumed in this study that the decrease of

products in j-th industry is specified by

either I. or I., which has larger reduction

factor nr. 1 Therefore the decrease of products

JX. in j-th industry due to the first spread

ef:fe ct was estimated as

JX,

J X . x max ( a . J i ij

0 ) (7)

It should be noted here that such a decrease

of products JX., then, decreases the values

added such hs commercial interests and

salaries, i.e., JX. was to be used to develop

products in a form"' of consumption expenditure

and capital formation if the earthquake did

not occure. Such a effec.t due to decrease of

products JX. is designated in this study as a second spread effect. Determination of the

second spread effect follows the procedure

which will be described in 3*6.

3«5 Evaluation of Products Which was Supporsedly Developed If the Earthquake Did

Not Take Place

Products, which was supporsedly developed

unless the earthquake occured, were determined by a regression analysis of the

products in the last several years before the

earthquake

3.6 Evaluation of Increase Due to Investment

for Restoration

Repair and restoration after the earthquake

accompany expenditures to purchase materials

and goods for restoration. Such expenditures

give positive impacts on various industries,

which, in turn, develop the spread effeets to the other industries. The spread effects

caused by such chains were analyzed by the

input-output analysis.

The input-output table presented in Table 2 may be written as

A X + F = X (8)

where, A_, X. and X represent input coefficient

matrix, output (products) vector and final

demand vector, respectively. From Eq.(8),

output vector X can be obtained as X = ( I - A. ) -1 F (9)

in wh^ch I represents unit matrix, and ( 1_ -

_A )” is generally refered as the Leontief

inverse matrix table. The procedure to calculate an increase due to investment

associated with restoration consists of;

1 ) Evaluation of a first spread effect

(T) Form final demand vector F_ by studying

restoration investments

(2) Calculate output (products) X developed

by F. Q) Calculate value added X V by

multiplying T vector of ratio of value

added V to X

1

2) Evaluation of a second spread effect

© Determine final demand vector F. associated with value added calculated

byQ)

(5) Calculate output (products) X developed

by (4) with use of Eq.(9)

© It erate Q), © and © until converge

3) Estimation of investment multiplier

© Calculate production inducement

multiplier by dividing total products

developed by the restoration

investments

© Calculate income inducement multiplier

by dividing total income inducement by

the restoration investment

© Calculate tax by assuming tax ratio

4. APPLICATION TO THE NIHONKAI-CHUBU

EARTHQUAKE OF 1983

354
4.1 Reduction of Products Due to Seismic

Damages

unless the Nihonkai-chubu earthquake occured

are also estimated as

Total amount of direct damage in the Akita

prefacture due to the Nihonkai-chubu

earthquake which occured in 1983 at sea bottom of Japan Sea with magnitude of 7.7 in

Richter scale, was estimated as 148.2 billion

yen in the nominal price of 1983, which is

approximately 6.9% of the gross prefectural

products of 2163.5 billion yen. Decrease of

products in the Akita prefecture was estimated in accordance with the proposed

procedure presented in the preceding chapter.

1 ) Decrease of products due to damage of

facilities for production

Direct damages of the facilities and/or

properties for production were estimated by

private and regional government bodies after

the earthquake, and the data for agriculture,

forestry, fisheries, manufacturing

industries, wholesale and retail were

available. Among these damages, decrease of

products associated with damages of forestry

was disregarded in this analysis because of

difficulty to evaluate products vs. properties relation.

Reduction factor of agricultural products was firstly estimated by formulating a production

function model assuming the cultivated area

as the private capitai stock in nq. j

However, because accuraly of such a production function model for the

agricultural products was poor, it was then

decided to estimate the reduction factor

from the unit product model, i.e.,

agricultural products per unit cultivated

area was estimated by regressing the data

between 1970 and 1981 as

U = 6.81 - 0.105 x Y ( R=0.95 ) (10)

YE = U x 162,329

= 01753 x 162,329

= 122,293 ( million yen ) (13)

Therefore substituting Eqs.(12) and (13) into

Eq.(2), one obtains the reduction factor a of

agricultural products as YE58 Y^g

YE

= 1.9 %. :u)

58

Fisheries facilities suffered severe damages

by tsunami with an amount of 4,195 million

yen. Because accuracy of production function

models with use of Eqs.(3)~ (5) taking

account of number of fisher and fisheries

capital was not sufficient, the reduction

factor of fisheries products was estimated

based on fisheries products per unit

fisheries capital stock as 26.5 %.

For products of manufacturing industries,

production function model was formulated from

the data of 1982 ( 57 in Showa ) as

Y = 8.03K°' 519

L°‘

481

(R = 0.87) (15)

in which Y, K and L represent price of

forwarding goods in million yen, fixed assets

r r. V 3 7i clpd ^ of' "1 o

^ respectively. Because seperate regression

analyses predicted K and L in 1983 as K = 179,939 million yen and L = 87,698 persons,

the products of manufacturing industries YE58

which would supposedly achieved in 1983 ( 58

in Showa ) if the earthquake did not take

place may be estimated as

YE = 8.03 x 179,939°‘-519 x'87,6980 ’ 481 = 1,022,741 (million yen) (16)

in which U and Y represent the agricultural

products per unit cultivated area in million

yen, and year in Showa, respectively, and R represents a regression coefficient.

Therefore, the agricultural products per unit

cultivated area in 1983 ( 53 in Showa ( fiscal year ) ) were estimated from Eq.(10)

as

U co = 6.81 - 0.105 x 58 = 0.753 (ID

bo

On the other hand, total cultivated area in

1983 in the Akita prefecture was estimated as 162,329 ha, and the cultivated area of 3,139

ha was reported to be damaged by the

Nihonkai-chubu earthquake. Therefore when it

is assumed that the area which suffered

damages by the earthquake could not be used

for creating agricultural products in 1983 ( 58 in Showa ) , agricultural products at this

year Y co may be estimated as

5o

Y = U x ( 162,329 - 3,139 ) = 01753 x 159,190

= 119,928 ( million yen ), (12)

The agricultural products YE^q which would

possibly developed in 1983 ( 58 in Showa )

On the other hand, damage on facilities ' of

manufacturing industries, which suffered

damages from the Nihonkai-chubu earthquake,

was reported as 2,925 million yen. Therefore,

the products of manufacturing industries Y^g

in 1983 ( 58 in Showa ) was estimated as

Y =8.03 x (179, 939-2, 925)°‘ 519

x 87,698

0 ” 481

= 1,014,081 ( million yen ) (17)

It should be noted in Eq .(17) that loss of

number of labor due to death and/or injured

was disregarded. The reduction factor was, then, estimated for manufacturing industries

from Eq.(2) as

q , YE?8 ~ 158

. 0.85 % (18)

IE

58

Reduction factor for wholesale and retail was estimated by formulating a production

function model with use of the products and

the capitals, in which capitals for wholesale

and retail were estimated from capitals per

unit number of employee.

The reduction factor of products associated

with damage of facilities and/or properties

for productions was thus estimated for each

355
industry as shown in Table 4- It should be

noted here that Table 4 includes the

reduction factor of products associated with

damage of transportation facilities,

materials and goods, which will be presented

later

2)

Reduction factor of products associated

with damage of transportation facilities

Restoration of road was made very quickly

after the Nihonkai-chubu earthquake because

the damage was relatively limited. Therefore,

effect of road damage was disregarded here,

and reduction factor of product was estimated only for railways and ports.

In the railways, decrease and suspention of

transportation were developed at Oh-u MainLine, Oga Line and Gonoh Line for 10 20

days, and this was reported to reach

approximately 1.7 % of the annual total

transportation for passengers. It should be

noted, however, that because freight

transportation was alternatively made by

trucks instead of railway, effect of

suspension and decrease of railway

transportation is regarded practically

limited. Therefore effects of railway , damage

were taken into account only for third

industries by disregarding their effects for

first and second industries. The reduction

factor was assumed as 1.7 %.

Ports and harbours had suffered one of the

most severe damages due to the Nihonkai-chubu

earthquake, and it took one year and five

months for their restoration.

Handling volume of the materials which depend

on ship transportation such as crude oil,

cement, heavy oil, petroleum goods was substantially decreased in the Akita

prefecture. However, such a decrease of

landing volume was not necessarily considered

to bring direct decrease of products which

depend on ship transportation; most of cement

was landed at ports located in adjacent

prefectures and transported to the Akita

prefecture by trucks, heavy oil and petroleum

goods were landed with use of exclusive

wharves, which did not suffere damages, etc.

Alternative transportation could not be made

only for crude oil, which were used at the

Tohoku Electric Power Station. Fig. 2 shows

landing volume of crude oil in the Akita

prefecture, and it is seen that volume is

only 450 thousands ton in 1983 ( 58 in Showa

) , which is approximately 45 % decrease as compared with the volume of 820 thousand tons

which would supposedly expected if the

earthquake did not take place. Therefore, it was assumed that damage of ports only

developed an decrease of electricity with an amount of 45 %. It should be, however, noted

that because shortage of electricity in the

Akita prefecture was supplied from

surrounding prefectures, decrease of products

due to shortage of electricity was not

developed. Therefore, the reduction factor of

45 % was considered only for production of

electricity.

The reduction factor of products thus

estimated is shown in Table 4.

3) Reduction factor of products associated

with damage of materials and goods

Decrease of products associated with damage

of materials and goods was evaluated for

damage of agriculture, livestock, forestry, commerce and industry which were shown in

Table 3. Dividing these decrease of products

by amounts of products ( output ) of the

input-output table, reduction factor of

products was evaluated for each industry as shown in Table 4. It is seen in Table 4 that

although the reduction factors are generally small, those for agriculture and commerce are appreciably larger than for

other industries.

4) Critical reduction factor of products

The most critical reduction factor of

products among the three types of damages,

i.e., damage of facilities for production,

transportation facilities, and materials and

goods, was evaluated as shown in Table 4. It

is apparent that the reduction factor was especially high for the ship transportation

in the electricity industries ( « = 45 %) and

for fisheries ( a = 26.3 % )• 5) Analysis of decrease of products

The input-output table of the latest version

of 1980 was used for input-output analysis to calculate decrease of products. It should be

noted that the change of inter-industry

structures developed between 1980 and 1983

(the Nihonkai-chubu earthquake of 1983)

cannot be considered in this analysis. Large

classifications with 14 categories and small

classifications with 1 64 categories were used

for input and output, respectively. It should

be noted, however, that only for a category

of transportation small classifications were

used for input to identify ship

transportation. Table 5 shows decreases of

products predicted by the procedure described

in the preceding chapter using the reduction

factors presented in Table 4* The output was compiled into semi-large classifications with

31 categories. It is seen in Table 5 that

total amount of decrease of products is

predicted as 248.1 billion yen, which

corresponds to approximately 1 .7 times of the

direct damage of 148.2 billion yen ( refer to Table 3 ) , and that the most significant

decrease of products occured in service ( 95.4 billion yen ) followed by electricity,

gas and water industries ( 730 billion yen ).

Then, the second spread effects were calculated. Table 6 summalizes the second

spread effects thus calculated as well as

the first spread effect and direct damages

due to the Nihonkai-chubu earthquake. A ratio

of the total damages caused by the earthquake

to the gross prefectural products of the

Akita prefecture was estimated as 11.0 % and

10.2 % on the bases of input-output table

(1/0 base) and value added, respectively.

4.2 Products Supposedly Developed Unless The

Earthquake Occrued

Fig. 3 shows variation of gross prefectural

products between 1975 ( 50 in Showa ) and

356
1982 ( 57 in Showa ). The gross prefectural

products Y at the year of y ( in Showa ) are obtained from regression analysis as

Y = -1,680,580 + 68,575 x y (million yen)

(19)

Extrapolating the products" with use of

Eq .

( 1 9 ) » one obtains the gross prefectural

products in 1983 as 2296 . 8 billion yen.

4.3 Increase of Products due to Investment

for Restoration

Increase of products due to investment for

restoration was analyzed by the input-output

analysis seperating the restorations of

public works ( 44-3 billion yen ) and those

for housings ( 11.9 billion yen ). It was found from the analysis that 22 billion yen

of the total restoration of 44.3 billion yen

for public works was paid to the industries,

and that the rest ( 22.3 billion yen ) was spent for value added. According to such

investments, products of 89.2 billion yen

were developed including the original

investments, which, in turn, developed

secondary spread effects. Including all these

effects, it was found that the investment for

restoration of public works developed

products of 122.3 billion yen as well as value added of 40.9 billion yen.

The same analysis was made for the

investments for restoration of housings ( 11.9 billion yen ), and it was found that the

investments developed products of 23*3

billion yen and secondary spread effects of

6.4 billion yen.

4*4 Effect of the Nihonkai-chubu Earthquake

on Economy in the Akita Prefecture

Table 7 summalizes the decreases of products

by the earthquake as well as the increase of

products due to investments for restoration.

It was predicted that the • Nihonkai-chubu

earthquake caused decrease of 87.5 billion

yen on the basis of value added. Because the

products supposedly developed unless , the

earthquake occured were estimated as 2296.8

billion yen in 4.2, it was considered that

the gross prefectural products would be

2209.3 billion yen.

On the other hand, the gross prefectual

products of the Akita prefecture in 1983 was reported in 1985. Table 8 compares the abovedescribed predicted products with the actual

products. It is seen in Table 8 that the

gross prefectual products in 1983 was 227.3

billion yen, which is 63.5 billion yen

larger than the above prediction. Precise

comparisons depending on each industry are ;

1

When a ratio R is defined as the actual

gross prefectual products divided by the

predicted gross prefectual products, the

ratio R takes a value between 0.89 and 1 .22

excluding four industries, i.e., forestries(R

= 1.46), minings(R = 1.43), electricity, gas

and water(R = 0.58), and service and others(R

= 0.78).

2) In case of fisheries industry, a decrease

of products predicted as 27.1%, which is one of the most significant decrease of products,

is fairly in good agreement with the actual

decrease of 25.5%.

3) A ratio R is quite large in forestry and

minings while it is very small in service.

The reason for such discrepancies may be

attributed to the fact that the products for

these industries varied significantly between

1980 and 1983, i.e., although the latest

version of input-output table of 1980 was

used for the analysis of indirect losses of

the Nihonkai-chubu earthquake of 1983, it

does not consider changes of production

structure developed between 1980 and 1983- It is considered that the products of forestry

and minings had been decreasing between 1980

and 1983, while those of service had been

increasing

4) The actual products in electricity, gas

and water industry are approximately two

times larger than those predicted. This is

regarded as a difference of assumption in

this analysis that although the products of

electricity was assumed to decrease in

accordance with decrease of landing volume of

crude oil at the port, actual products did

not decrease because the electricity was supplied from a net-work covering the whole

Tohoku region.

5. CONCLUDING REMARKS

For aiming to develop a complehensive

procedure which considers socio-economic

effects of earthquake from more wide views,

proposed was a procedure which evaluates

effect of earthquake on economic loss at a specific region, and applicability of the

procedure was studied for the Nihonkai-chubu

earthquake of 1983. The features and

tentative conclusions deduced from this study

are the followings.

1 ) To evaluate economical loss due to

earthquake, proposed was a procedure which

considers not only decrease of products with

use of input-output analysis from three

sources, i.e., damage of producing

facilities, damage of transportation

facilities, and damage of materials and

goods, but increase of products associated

with investment for restoration. This is an extention of the procedure, which is widely

used to evaluate an positive effect for

constructing public works, to seismic

effects. The proposed procedure has, however,

several limitations, such that it uses an input-output table which could only be used

for economical activities at the normal time,

it cannot be used to evaluate economical

damage at a very limited area, etc., and

further improvements are required.

2) Reduction factor of products, which is

defined as the decrease of products due to the Nihonkai-chubu earthquake divided by the

products supposedly developed if the

earthquake did not take place, takes

significantly high value of 45 % in ship

transportation and of 26.3 % in fisheries as

357
compared with reduction . factor in other

industries such as 1.9% in agriculture and

0.8 % in manufacturing industry. Reduction

factor of products in road transportation was very limited. This is considered to be

attributed from the fact that functions of

ports had been suspended for a long time due

to damage of the Akita port, and tsunami

caused significant damage to marine

facilities, while restoration of road

facilities could be made promptly.

3) Although total direct damage in the Akita

prefecture caused by the Nihonkai-chubu

earthquake was 148.2 billion yen, decrease of

products associated with such direct damages

was estimated as 3 68.6 billion yen, which is

2.5 times larger than the total direct

damages and consists of 248.1 billion yen by

decrease of production, 112.2 billion yen by

decrease of consumption and 8.3 billion yen

by damage of materials and goods.

Consequently, considering total direct

damages of 148.2 billion yen and decrease of

products of 368.6 billion yen, it is

considered that the Nihonkai-chubu earthquake

developed a loss of 508.5 billion yen in the

Akita prefecture, which corresponds 11.0 % of

the gross prefectural products of 4639.6

billion yen ( I/O base products ) in 1983.

4) Substituting increases of products

developed by investment for restoration, the

decreases of products in the Akita prefecture

was estimated as 216.6 billion yen on the I/O

base and 87.5 billion yen on the value added

base. Consequently, because gross prefectural

products in 198 3, which was supposedly

developed if the earthquake did not take

place, was estimated as 2296.8 billion yen,

the gross prefectural products actually

developed in 1983 was predicted to be as 2209-3 billion yen. In comparison, actual

gross prefectural products in 1983 were reported by the Akita prefecture as 2272.8

billion yen, which is 63.5 billion yen larger

than the above prediction.

REFERENCE

1 ) Akita Prefecture : Record of the Nihonkaichubu Earthquake of 1983, 1983

2) Kawashima, K. , Obinata, N. and Kanoh, T. : Estimation of Socio-economic Damage

induced by Earthquakes Part 1 Case Study

for the 1983 Nihonkai-chubu Earthquake,

Tecnhical Memorandum, No. 2267, Public

Works. Research Institute, Ministry of

Construction, 198-5 3) Katayama, T. and Moriyoshi, S. : Nihonkaichubu Earthquake and Noshiro City Case

Study of Effect of Earthquake on Regional

Social Activities, Report NO. 84-02(5),

Institute of Industrial Science, University

of Tokyo, 1984.

4) Kawashima, K. , Obinata, N. and Kanoh, T. : Effects of Socio-economic Damage Induced

by Earthquake A Case study for the 1983

Nihonkai-chubu Earthquake, Civil

Engineering Journal, Vol.29-3, 1987.

358
Table-1 Classification of Earthquake Effects

Classification Objective and/or Form of Damage Effects of Loss and/or Damage

Human Life Death, Injury Loss of man power. Mental effects

Facilities for Living House, Educacional Facilities, Hospitals,

etc

Functional loss for Daily Human Activities

Facilities for Production Cultivate Field, Non-Residential House

(Factory, Shops), etc. Decrease of Production, Loss of Products

Life-line Facilities Electricity, Telecommunication, Gas, Water and Sewage, Trash

Decrease and/or Loss of Capability for Produc tion

Transportation Facilities Road, Bridge, Port, Rail-way Decrease of Production due to Decrease

of Transportation Capability for Material

and Products

' Decrease of Production due to Decrease of Transportation Capability for Passengers

Public Works Excluding

Transportation Facilities

River, Sabo-works, Slope failure Damage associated with Failure

Transportation Media Trucks, Buses, Trains, Ships Decrease of Production due to Decrease of Transportation Capability

Manufactured Goods,

Partially Manufactured

Goods

Loss of Manufactured and/or Partially

Manufactured Goods Decrease and/or Loss of Value of Goods, Decrease of Products due to Loss of Manufactured and/or Partially Manufacture

Goods L_

Table-2 Inter-industry Relation Table (Input-output Table)

Purchase

Sell

Intermediate Demand Final

Demand

Products

II 1

2 U

* - —

Intermediate II Sales 12

Ii Xii '••• xir fi Xi

Xii fi X

i

Value Added v i

Products xi xr

359
Table-3 Damages by the Nihonkai-chubu Earthquake of 1983

(million yen)

Type Classification Damage

(million yen)

Damage of Agricultural Facilities 24,477

Facilities Fishery Port Facilities 1,486

Marine Facilities 4,195

Forestry Facilities 5,557

Commercial Facilities 2,925

Industrial Facilities 4,918

Other Commercial and Industrial

Facilities

2,263

Public Works 59,253

Other Public Works 3,958 House 26,665

Educational Facilities 1,733

Damage of Agriculture 2,411

Materials and

Goods Lives tock 7 Forestry 400 Commerce 1,344

Industry 3,437

Other Commerce and Industry 708 Rail Ways 2,200

Others 301

Total 148,238

Table-4 Reduction Factor of Products [%]

Classification

rr© Due to Damage of Facilities for Production

( 2) Due to Damage

of Transportation

Facilities

(3) Due to Damage

of Materials

and Goods

(4) Most Critical

Reduction

Factor Between (D , (2) and (3)

1. Agriculture 1.93 0 0.79 1.93

2. Forestries 0 0 0.48 0.48

3. Fisheries 26.27 0 0 26.27

4. Mining 0 0 0 0

5. Manufacturing 0.85 0 0.12 0.85

6. Construction 0 0 0 0

7. Electricity, Gas, Water 0 0 0 0

8. Commercial 0.01 0 0.75 0.75

9. Finance and Insurance 0 0 0 0 10. Landed Estate 0 0 0 0 Road 0 0 0 0

1 1 Tl - rrj Rail

^ 0 1.7 1 ’ 0 1.7 1 ’ xx. iranspor ta- tion and Ship 0 45.0 2) 0 45.02) Communication —z,

:

Communication 0 0 0 0

12. Service 0 0 0 0

13. Official Business 0 0 0 0 14. Unclassif icable 0 0 0 0

Note: 1) Only for products for third industry

2) Only for products for electricity

360
Table-5 Direct Loss due to Reduction of Products

Classification

(T) Original

Products

(billion yen)

( 2) Reduction of

Products

(billion yen)

(3) Decreasing

Rate (

2)/(D

[ Z ]

1 , Cultivated Agriculture 253.6 4.9 1.93

2 , Livestock 52.9 1.02 1.93

3 . Forestry 82.8 1.04 1.25

4 . Fisheries 11.5 3.02 26.27

5 . Mining 91.6 0.78 0.85

6 . Food 181.7 18.55 10.21

7 . Texture and Clothes 100.5 1.81 1.80

8 . Wood and Wooden Goods 236.8 2.36 1.00

9 . Pulp and Paper 48.4 0.7 1.45

10 . Printing and Publishing 20.2 0.17 0.85

11 . Leather and Rubber Goods 1.3 0.02 1 . 64

12 . Chemical Products 35.1 0.3 0.86

13 . Petroleum and Coal Goods 43.5 0.42 0.97

14 . Ceramics 47.6 0.4 0.85

15 . Iron and Steel 13.9 0.12 0.85

16 . Iron and Non-iron Goods 146.5 1.24 0.85

17 . General Machinery 44.3 0.38 0.85

18 . Electric Machinery 149.6 1.27 0.85

19 . Transportation Machinery 62.9 0.53 0.85

20 . Precision Machinery 14.8 G .13 0 . 84 21 . Other Manufactures 12.6 3.31 26.27

22 . Construction of Buildings 356.3 5.28 1.48

23 . Construction of Public Works 246.9 4.77 1.93

24 . Electricity, Gas, Water 193.0 73.02 37.82

25 . Commerce 460.2 7.82 1.70

26 . Finance and Insurance 114.5 1.95 1.70

27 . Landed Estate 264.1 4.22 1.60

28 . Transportation and Communication 307.5 5.23 1.70

29 . Service 641.4 95.4 14.87

30 . Official Business 145.2 7.32 5.04

31 . Unclassif icable 34.4 0.66 1.93

32 . Total 4 , 415.6 248.14 5.62

Table-6 Predicted Reduction of Products due to

the Nihonkai-chubu Earthquake of 1983

Direct Damage

(billion yen)

Damage Associated with Decrease

of Products (billion yen) Total Damage

(billion yen)

Ratio to Gross Prefectural Products

[%] Direct Effect of Consumption

(2nd Spread Effect)

Total

1/0 Base 148.2 248.1 112.2 360.3 508.5 11.4

Supplementary Value

Base (Gross Prefectural

Products Base)

73.2 120.6 39.9 160.5 233.9 10.2

361
Table-7 Effect of the Nihonkai-chubu Earthquake of 1983

on Economy of the Akita Prefecture

I/O Base Supplementaly

Value Base

(T) Decrease of Products Damage of Materials and Goods 8.3

L) 4.1

by the Earthquake

(billion yen) Direct Decrease associated with 248.1 120.6 decrease of products

Decrease of Consumption 112.2 39.9

Total 368.6 164.6

(2) Increase of Products First Effect 89.2 51.4 due to Investment for Restoration of Public

Works Second Effect 33.1 11 .

(billion yen)

Total 122.3 63.1

( 3) Increase of Products First Effect 23.3 11.7 due to Investment

for Restoration of Second Effect 6.4 2.3

Housing

(billion yen)

Total 29.7 14.0

Decrease of Products ( (T) + (D + (3) ) (billion yen) 216.6 87.5

Ratio to Gross Prefectural Products (%) 4.7 3.8

Note: 1) Total of damage of products and goods presented in Table 3.

Table-8 Comparison of Predicted and Actual Products

Developed in the Akita Prefecture in 1983

(Million yen in the price of 1983 fiscal year)

Prediction on Products of Akita Prefecture in 1983 Fiscal Year

Reduction of Products (D Products (Z) Predicted (7) Actual Classif ication Decrease of Products due

to Seismic

Damage

(2) Increase of Products due

• to Investment

for Restoration

of Public Works

(3) Increase of Products due

to Investment

for Restoration

of Housings

_

© Sub-total ©-©-©

Supposedly

Developed

Unless the Earthquake Occurred

Products © *© -©Products Developed

in 1983

Ratio ©/©

1 . Agriculture 5,372 174 30 5,168 205,363 200,195 201,578 0.99

2. Forestries 1,743 435 275 1,033 46,248 45,215 30,904 1.46

3. Fisheries 1,899 27 6 L , 866 7,320 5,454 5,193 1.05 - Minings 1,396 1,458 184 -246 58,540 58,786 41,018 1.43

5Manufacturing 10,391 2,902 721 6,768 341,085 334,317 322,208 1.04

6Construction 7,845 45,670 10,039 -47,864 263,632 311 ,496 244,481 1.27

7

Electricity, Gas, Water

25,929 487 98 25,344 70,509 45,165 78,327 0.58

3. Commercial 14, *21 2,701 554 11,166 333,730 322,564 334,056 0.97

9. Finance and Insurance

* ,451 1,457 317 2,677 88,167 85,490 91,728 0.93

10 Landed Estate 4,471 627 144 3,700 232,495 228,795 226,156 1.01

- 1 Transportation and Coonunication

5,013 1,676 356 2,981 110,476 107,495 104,986 1.02

22 Service 31,612 5,563 1,220 74,829 539,195 464,366 592,214 0.78

Total 164 , 544 63,156 13,944 87,422 2,296,760 2,209,338 2,272,849 0.97

362
LANDING

VOLUME

OF

CRUDE

OIL

AT

AKITA

PORT

[THOUSAND

TON]

Fig.1 Evaluation of Economical Effects of Earthquake

FISCAL YEAR Y IN SHOWA

rig. 2 Variation of Landing Volume of Crude Oil

at the Akita Port

2400 —1

1 1——11

1 1

r 1

r—\ 2 GPR SUPPOSEDLY DEVELOPED

£ 2 300 UNLESS THE NIH0NKAI - CHUBU ——x

- EARTHQUAKE 0CCURED

2 200

S

0 X

2100 - / 0 -

1— 0/

0 2000 -

-

crj 1900 - 9^o \

- <

£ 1800 -

- OUJ gross prefectural products

£ 1700 ^ = -16.806.580 + 68.575 x Y [million yen] - CL

( Y - FISCAL YEAR IN SHOWA)

| 1600 -

- O

I 1 1 1 • 1 —1 11

1 DUO

50 51 52 53 54 55 56 57 58

(1975) (1977) (1979) (1981 ) (1983) FISCAL YEAR Y IN SHOWA

Fig. 3 Variation of Gross Prefectural Products

between 1975 and 1982

363
Analysis of Strong Motion Earthquake Records

on January 9, 1987 12by

^ Nobuyoshi Yamaguchi , Sadaiku Hattori , Shinsuke Nakata , Hatsukazu Mizuno^, Masanori Iiba~*

ABSTRACT

Strong motion networks of Building Research Institute observed strong motions from

the magnitude 6.6 Iwateken-Chubu ( middle of

Iwate Pref.) earthquake of January 9, 1987.

Peak acceleration on the first basement of a building in Hachinohe city was 197 gals and

561 gals on the fifth floor.

A slight building damage occurred during

this earthquake. The damaged building had a experience of a moderate damage during the

1968 Tokachi-Oki earthquake. The building had

been repaired and strengthened after the

1968 earthquake, and it had damage again.

BRI have installed accelerographs in another

building beside the damaged building, and

measured strong motions during this

earthquake.

The analysis for these records clarifies

the periods of them, and the behavior of the

observed building. The building damage in

1968 and 1987 is examined. It is suggested

the strengthening of the building after the

1968 earthquake is insufficient.

KEY WORDS : Strong motion record,

Earthquake input,

Seismic response of building,

Earthquake damage

1.

INTRODUCTION

Building Research Institute have observed strong motions of the ground and

buildings since 1957, in order to clarify

building behavior and strong ground motions

during severe earthquakes. For example, BRI

succeeded liquefacted ground motions during

the 1964 Niigata earthquake, and the world

record building responses ( 1.06 G ) during

the 1978 Miyagiken-Oki earthquake. The object

of this paper are to analyse strong motion

records and to evaluate strengthening

effects of a building based on the records.

2.

EARTHQUAKE DATA AND OBSERVED RECORDS

Table 1 shows the earthquake data during

the Iwateken-Chubu earthquake of January 9,

1987.1) Figure 1 shows isoseismal map of

the earthquake. 2) Table 2 summarized the

observed records by BRI network. ' Although

this earthquake was comparatively large among

the earthquakes occurred in inland of Japan,

the earthquake damage was slight on the

whole due to its deep hypocentre.

3. OBSERVATION SITES

Photo 1 shows Atago junior high school

in Miyako city which was the closest site to the epicentre among the BRI stations. The

accelerograph of BRI was installed on the

west end of the first floor in the school

building. The site was cut for construction

of the school, and its ground condition is

fairly good.

Photo 2 presents Hachinohe city office

which is 77.2 km apart from the epicentre.

Hachinohe city office consists of the old

city office completed in 1960, the annex building in 1969 and the new city office in

1980. The old city office and the annex building are jointed by expansion joints

shown in photo 3. Photo 4 shows the annex building where the accelerographs were

installed on the first basement and on the

fifth floor. Figure 2 shows the soil profile

of the site. The layers from the surface to -9m are consist of clay loam and sandy loam

whose N values are less than 5. There are

the layers of gravel in the depth of -9m

whose N values are over 50.

4DAMAGE OF THE 1968 EARTHQUAKE

It was reported the old city office in

Hachinohe had been damaged during the 1968

Tokachi-Oki earthquake. Figures 3-5 show

the damaged penthouse and frames. Figures 6 - 8 show plan views of the old city office

and shear or bending damage of the columns in

1968. The old city office is reinforced

concrete building supported by precast

concrete piles (length: 9m, 5m with

basement). ' The building was repaired and

strengthened after the 1968 earthquake.

The damaged penthouse was removed from the

building, and walls for strengthening were installed around the east stairs of the

building.

1: Geotechnical Engineering Division, Structural Engineering Department, Building Research Institute, Ministry of Construction,

Japan

2: Dr. of Science, Special Senior Scientist,

International Institute of Seismology and

Earthquake Engineering, USEE, BRI

3: Dr. of Eng. Head of Building Engineering

Division, USEE, BRI

4: Head of Housing Construction Division,

Production Department, BRI

5: Research Engineer, Earthquake Information

Division, IISEE, BRI

364
5. INSTRUMENTATION

The accelerograph was SMAC-M which

includes three components of force balance

type transducers and a FM recording system

using magnetic cassette tapes. The records ( * marked on table 2 ) were digitized every

1/200 or 1/100 seconds, and were made a tentative base line correction by least

square fit method. The results are called

uncorrected accelerograms. Table 2 summarized the maximum accelerations of them.

But the values without * marks are calculated

from the half length between positive and

negative peak values. A final base line

correction is carried out for the records of

Hachinohe city office, and the velocities and

displacements are calculated by the method of

M.D. Trifunac.^ Table 3 presents the cut-off

and roll-off termination frequencies for the

Ormsby filter used in the method.

6.

ANALYSIS OF STRONG MOTIONS

Figure 9-11 show uncorrected accelerograms of Miyako IF, Hachinohe B1F and 5F.

Figure 12 - 16 show corrected accelerograms,

velocities and displacements of Hachinohe B1F

and 5F. Figure 17 - 22 show elastic response

spectra of the records of Miyako IF and

Hachinohe B1F. Figure 23 - 25 show Fourier

amplitude spectra of the records of Hachinohe

5F.

Figure 10 shows the maximum accelerations are on pulses, and a duration of the

principle motion is about 10 seconds.

Figure 18 shows frequency domain of EW components and periods of the motions are about

0.07 seconds (14 Hz) and 0.15 seconds (7 Hz).

These periods are rather short among ground

motions of typical buildings in Japan. The

soil profile and short epicentral distances

are considered to affect their

characteristics. Although the maximum

accelerations in Miyako were large, there was

no remarkable damage around Miyako city

because of their short periods.

Figure 21 shows the periods of EW component of Hachinohe B1F are around 0.16 - 1.0

seconds (1-6 Hz). Figures 12 - 13 show the

maximum velocities of NS and EW components

are about 10 kines, and a duration of the

principle motion is about 10 seconds. Figure

23 presents the period of NS component of

Hachinohe 5F is about 3 Hz and that of EW

component is around 2 - 2.5 Hz. Figure 16

indicates the maximum velocity of EW component reaches about 38 kines, and the maximum

displacement is about 3 cm. The results

indicate the annex building deformed 1/500

of the height by adopting the assumption of

fixed base. This deformation suggests

structural damage of the annex building would

not arise during this earthquake. In fact,

any structural damage of the annex building

is not observed.

7. DAMAGE OF BUILDING

The old city office in Hachinohe had

damage of the fixed windows during this

earthquake. The building have been jointed

to the annex building with expansion joints.

A glass glazing method of the old city

office using stiffened putty material is

consider to cause this window damage. The

fixed windows of the annex building using PVC

glazing channels had no damage during this

earthquake.

The old city office also had slight

structural damage in spite of strengthening

after the 1968 earthquake. Photo 5 shows the

column besides the courtyard, which had

slight shear-bending damage due to the

spandrel walls. Photo 6 presents the cracks

on the walls which had been added for

strengthening after the 1968 earthquake.

Photo 7 shows the damage of expansion joints

by the collusion of two buildings. Photo 8

indicates the repaired column by the injection of epoxy adhesive to the cracks.

8. CONCLUSIONS

Strong motions were observed beside

the damaged building, Since a height of the

damaged building is about 15 m, the natural

periods of that will be about 0.3 seconds.

The maximum response accelerations of the

damaged building is evaluated 500 - 600 gals

for EW direction and 300 - 400 gals for NS

direction based on the response spectra of

the ground motions. This result indicates the

loads acting to the building are moderate.

Since any structural damage is not observed

around Hachinohe city except this case, it

is considered this building needs more

strengthening. Further inspections for this

problem are required.

9. ACKNOWLEDGMENT

Strong motion observations in Miyako and

Hachinohe are performed in cooperation with

Atago junior high school of Miyako city and

Hachinohe city office. The survey for the

old city office in Hachinohe was assisted by

the staff of Hachinohe city. The authors

would like to express great thanks to them

and Dr.Tetsuo Kubo and Mr.Hideo Tsukagoshi

for making the programs of analysis.

REFERENCES

1)

. Announcement of Japan Meteorological Agency

2)

.The Mainichi News papers of Jan. 10 1987

3)

. 'Strong Motion Earthquake Observation News,

the Earthquake of Jan. 9 1987 at the Middle

of Iwate Pref.', Kenchiku Gijutsu No. 427

4)

. 'The Reconnaissance Reports for the

disasters of the Tokachi-Oki Earthquake of

1968’, AIJ, Dec. 1968

5)

.M.D.Trif unac.'Low Frequency Digitization

Errors and a New Method for Zero Baseline

Correction of Strong-Motion Accelerograms',

EERL 70-07, CIT

365
Table 1 Earthquake Data

Date & Time Jan. 9 , 1987 15 : 14 (JST)

Location Middle of IWATE Prefecture

Epicentre 39 51 1 N

141 47* E

71 km depth

Magnitude 6.6 ( JMA)

Intensity 5 : Oh-funato, Morioka

4 Hachinohe, Sakata, Ishinomaki, Sendai

3 i Mutsu, Aomori, Akita* Shin jo, Yamagata,

Fukushima, Ehirakawa* Onahama, Mito,

Tokyo, Yokohama, Urakawa, Hiroo, Obihiro

Kushiro

Table 2 Observation Results

Observation Sites Max. Acceleration (gal)

station floor instrument. Epicentral

Distance

component note

NS EW UD

MIYAKO

Atago junior

high school

IF SMAC-M 27.6 km 186

(N03W)

388

(S87W)

219

(UP)

HACHINOHE

City office

B1F ft 77.2 149

(S16E)

197

(N74E)

89

(UP)

H 5F M ff 312

(N16W)

561

(S74W)

218

(UP)

SENDAI

Tohoku univ.

IF ft 195.2 31

(N68W)

27

(S22W)

20

(UP)

H 9F ft f T 103

(N68W)

80

(S22W)

35

(UP)

366
Table 3 Cut-off & Roll-off Termination Frequency for filter

Record L

f T L

f C H

f C H

f T

^

HACHINOHE B1F 0.05 0.1 25 27

!» 5F 0.05 0.1 25 27

g. 1 Isoseismal Intensity Map Fig. 2 Soil Profile by Boring

( HACHINOHE City office)

367
Fig. 3 North Side Elevation of Old City Office

and Damaged Windows in 1968 (Ref. 4)

-tlOO ^

HI

Fig. 5 Section of 4th-Frame of Old City Office

and the Damage of Columns in 1968 (Ref. 4)

368
Annex

Building

Fig. 7 Second Floor Plan of Old City Office — Dotted lines show the added walles for

and the Damage in 1968 (Ref. 4) strengthening after the 1963 earthquake.

code damage

B-l small bending cracks

B-2 large bending cracks

B-3 very large bending cracks

S-l small shear cracks

S-2 large shear cracks

S-3 very large shear cracks

369
“Oc

03 U-

>-* CQ

370

Tint'

(SECI

Fig.

10

Uncofrected

Accelprograins

of

HACHJNOHF

B1F
tUb!U?M»/

’81.

I.

9/NU

Of

nll*HO/H«CMINOKC-Sr/NI6U

.'COBKEC

NOIivy 3 l 3 : 3 v

(3NIVI UI3013A

T3C

03 &U

ij

•H U3 u x

0 O

»H 2

0) »—

> aso

-o *<

01 2

JJ

CO 4J

pH (0 U

UJ tn "

4-> C

CO U

CJ c

o

- a

E E

CO O

u U

OQO C/3

i-i 2

<n

0J u

u<

*T30U

Qj f—

U CL U (0 O *H a o

00

•HCh NQUVH31333V ii

3033 A AN 3 U 33 V-)dSia

T3c

CO

OJ >

"O

0)

a*uH

CO U

O

• CL E E

co o

i- O

00O 3

Lj CjJ 0)

I u-»

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U

U J-J •c c

<D -o E

CD CD XJ U

u co

OJ -H U CL

i* (0 o *H O Q

00

•Hu<

371

nent

at

HACHIN0HE

5F
ruxmun

pseudo

vt'iocirr

mum

h

maximum

pseudo

velociit

ELASTIC RESPONSE SPECTRUM

E861E!N40/'87. 1. 9/nlO- OF IUA1E PREF . /n I T AKO - 1 F/M/ /t'NCOR.

N3 COnPUNENT

dapping - o. 3. s. io ano :o penctpi of critical

ELASTIC RE5PONSE SPECTRUM

E80 10 ! N4/ • 87. 1 . 9/NU (JF n 11 A KO/H ACH I NOME -0 1 F/ S 1 5E /UNCOR.

Nl COnPONENT

dapping - a. 3. L io aro 30 percent of critical

3

ER. I CD isec.i

2 . 17 Response Spectrum ( N’S, MIYAKO IF ) Elastic response spectrum

'??-!. r/^ID. CF U

T £ cs £

? . / .*1 1 7 A ^ C - 1 F / “ / /UNCOR.

£- CC“f*CS;N’'

- C. 2. 3. rc anO 2C ‘•EACEmT OF CRITICAL

: ' -

- '-v ! \ : \ /7~~? CCCH^—X ,N\\, \ v:y- V ./ /AAJ • \ i ~V v\\ i V^rXi )<NV \ //AX

x'V'V. \/ :\HV \Vv ‘N*. \'X>-

ELASTIC RESPONSE S ? E C T R l

1

r

'J’. it,/ ‘27. i. S/NL OF - ! T AKC/HACh ! \'GI-!E -5 J.c /r,74E /UNCOR.

Eu CCfiPOsEsT

C*flPi*C - C. 2. 5. IO anQ 20 RCACEmT 9* CR! TICal

3

! " ~

' ,^~T7~TT , • ~ N '• \ s V, V/V ^ y 7 //I/ J

0-01 0.02 0- 00 0.1 0.2 0. 0 1-0 2.0 9 . 0 10. 0 PERIOD (SEC.

Fig. 18 Response Spectrum ( EW, MIYAKO IF )

elastic response spectrum

E86TEIU40/ '87. 1. 9/rtta. OF 1UATE PREF. /n! TAKO- lF/fl/ /UNCOR.

UO COnPONENT

OahRING - a. 2. 5 . 10 AMO 20 PERCENT Of CRITICAL

Fig. 21 Response Spectrum ( EW, HACHINOHE B1F )

elastic response spectrum

E86T01U4/ *07. 1. 9/NU OF PI I T AKO/MACH I NOM£ -8 1 F/UP

UO COnroNCNT

OArrPlMQ o o. 3 . 5 . 10 anQ 20 PERCENT Of CRITICAL

Fig. 19 Response Spectrum ( UD, MIYAKO IF ) fig. 2 2 Response Spectrum ( UD , HACHINOHE B1F )

372
FOURIER AMPLITUOE SPECTRUM OF ACCELERATION

E 06 TO 2 N 4

/

87 . 1

. 9 /NU OF M

T AKO/HACH I NOHE -SF/N

6 U /UNCOR.

NS COMPONENT

Fig. 23 Fourier Amplitude Spectrum of NS Comp, at HACH1NOHE 5F

FOURIER AMPLITUOE SPECTRUM OF ACCELERATION

E 86 T 02 E 4 /

87 . 1 . 9 /NU OF nlTAKO/HACHINOHE- 5 F/S 74 U /UNCOR.

EU COMPONENT

600 r -i—|

1 1—i1—|—i—

4001-

2001-

10 IS FREO'JENCT - HZ Fig. 24 Fourier Amplitude Spectrum of EW Comp, at HACHINOHE 5F

FOURIER AMPLITUOE SPECTRUM OF ACCELERATION

E 65 T 02 U 4 /

87 . 1 . 9 /NU OF M I TAKO/H ACH I NOHE -Sr /UP /UNCOR.

jj 2 CC".*Cn£sT

Fig. 25 Fourier Amplitude Spectrum of UD Comp, at HACHINOHE 5F

373
Photo 1 Atago Junior High School

in MIYAKO

Photo 3 Expansion Joint between Old

and Annex Building of City Office

Photo 5 Damage of Column by the Spandrel

Walls during the Earthquake of 1937

I Photo 7 Damage of Expansion Joint by the Collision

of Boch Buildings during the Earthquke of 1987

*uj*ccxo« tr hhoto 4 Annex Building of HACHINOHE

City Office

Photo 6 Damage of Walls during the Earthquke of 1957

( These walls are added after 1968 )

Photo 8 Repaired Column after the Earthquke of 19S

( The same column as photo 5 )

374

Appendix:

Ten Task Committee

Reports

Report of Task Committee on

(A) STRONG-MOTION INSTRUMENT ARRAYS AND DATA

Date

May 14, 1987

Place

Public Works Research Institute, Ministry of Construction

Toyosato-machi , Tsukuba-gun, Ibaraki-ken 305, Japan

Attendees: Japan Side - H. Tsuchida (Chairman)

S. Hattori

M. Katsumata

K. Ohtani

Y. Sasaki

S. Noda

T. Kuwabara (Observer)

K. Tamura (Observer)

( PHRI

(3RD

(MRI)

(NRCDP)

(PWRI)

(PHRI)

(PWRI)

(PWRI)

U.S. Side - A.G. Brady (Chairman)

A.G. Franklin

R.R. Ledzian

H. Meyers

( USGS

(WES)

( USBR

)

( NOAA

I . Activities and Principal Accomplishments to Date

1) Catalogs of strong-motion earthquake records observed both in the

United States and Japan are being exchanged. U.S. data are

published in the "Seismic Engineering Program Report," and the

Japanese data are published in "Strong-Motion Earthquake Records

in Japan." Many other publications were also exchanged.

2) Preparation was made for the workshop on strong-motion earthquake

observation, which is expected to be held in August 1987. (See No.

10, following)

3) A progress report with ideas for consideration in a policy for

international dissemination of strong-motion earthquake records

has been prepared and presented to the Task Committee Meeting and,

subsequently, to the Panel. (See No. 12, following)

II . Future Programs

1. After an earthquake which has caused damage to structures or an earthquake during which maximum acceleration exceeding about

O.lg has been recorded, the task committee of the country where

the earthquake has taken place will provide a list of the

strong-motion earthquake records for the counterpart of the

task committee. The list contains maximum component

acceleration of each record. If there is such a list compiled

by any organization, the list mentioned above may be replaced

by i t

377
2 . Every year the task committee will exchange catalogs of the

strong-motion earthquake records during the previous year. The

catalog contains maximum component accelerations and wave forms

of major records. If there is such a catalog compiled by any

organization, the catalog mentioned above may be replaced by

it.

3 The Task Committee makes appropriate arrangements to provide

digitizable copies of records when they are requested. In

addition, arrangements will be made to provide information on the characteristics of the site, instrumentation, and

structures at the locations where such records are obtained.

4. When the organizations taking part in the Panel publish reports

on strong-motion earthquake records, these organizations will

distribute copies of the reports to the other organizations of

the Panel interested in them. The Task Committee exchanges

lists of the organizations and their Panel representatives who

wish to receive the reports.

5. The Task Committee will continue to exchange digitized data on

all the major strong-motion earthquake records recovered in

both countries, under the conditions mutually acceptable to the

agencies involved in acquisition of the data.

6. The Task Committee will exchange technical information

concerning seismometers and accelerometers to be developed, and

the deployment and installation of instruments with emphasis on

bore-hole instruments, and the design of seismic arrays.

7. The Task Committee plans to assist and cooperate, where

possible, in the following areas:

a. Assistance and cooperation with governmental organizations

in other seismic areas, in order to promote high quality

strong earthquake motion observations in all seismically

active areas of the world.

b. Assistance and cooperation with any international effort to

record strong ground motion close to the source of a large

magnitude shock.

c. Exchange of scientists, engineers, and appropriate

equipment in a timely fashion for observations following

large earthquakes in either country. As the largest and

most interesting aftershocks occur immediately following

the main earthquake, recovery of the most useful aftershock

data requires that, the response to major earthquakes be

initiated as rapidly as logistics permit.

378
8

The Task Committee feels the UJNR Panel on Wind and Seismic

Effects should play a major cooperative role in the

implementation of relevant parts of the Resolution of the

International Workshop on Strong-Motion Earthquake Instrument

Arrays held May 1978, in Honolulu, Hawaii . The exchange of

complete information on all aspects of the program, as it

develops particularly in Japan and the United States, will be

carried out in the manner of our standard exchange when

appropriate. Particularly with unique data from arrays, or deep bore-hole instruments in the U.S. and Japan, Task

Committee Chairmen will retain the responsibility of relaying

information on their existence and subsequently of transferring

the digital data from such institutions as the USGS , CDMG, USC,

and UCSD in the U.S.; and PWRI, SRI, PHRI, MR I , and NRCDP in

Japan.

9.

The Task Committee recommends collaborative studies involving

strong-motion instrumentation and interpretation of resulting

strong-motion data sets for purposes of earthquake hazard

reduction.

10. The Task Committee will hold a workshop on Strong-Motion

Earthquake Observation in the San Francisco area in August

1987. The details of the workshop program will be discussed

between, both Task-Committee chairmen and include array and

network planning, recording instrumentation, digitizing

hardware, and the digitization of representative records.

11. The Task Committee recognizes that another workshop on

Processing of Seismic Strong Motion Records to be held in

Vancouver in August 1987 is significantly relevant and would

complement the Panel's Workshop. The Task Committee urges mat

Panel members contribute to this workshop.

12. The Panel discussed the occasional, but extended,

unavailability of existing strong-motion data of engineering

significance from very large earthquakes outside the U.S. and

Japan. Based on input from the earthquake engineering data

community, and guidance from the Panel on Wind and Seismic

Effects, the Task Committee will draft a policy for

international dissemination of strong-motion earthquake records

for consideration by the Panel.

379

PROGRESS REPORT

WITH IDEAS FOR CONSIDERATION OF

A POLICY FOR INTERNATIONAL DISSEMINATION

OF

STRONG-MOTION EARTHQUAKE RECORDS

May 15, 1987

Task Committee on (A) Strong-Motion Instrument Arrays and Data

U.S. -Japan Panel on Wind and Seismic Effects, UJNR

1 . Introduction

At the 18th Joint Panel Meeting held in May 1986, in

Washington D. C., it was pointed out that some strong-motion

records of significantly large earthquakes which had taken place

in countries other than U.S. and Japan are still not available.

It was also pointed out that there were many matters to be

considered in establishing a dissemination rule which would be

accepted by both the disseminating group and the users group.

The Panel assigned the Task Committee-A to study this subject and

propose a policy for for international dissemination of strongmotion earthquake records.

The Task Committee started its activity on this subject

immediately after the 18th Joint Panel Meeting and has determined

1) necessary information and data to study the policy, 2)

necessary considerations from vi ewpo ints of the disseminators

and the users, and 3) tentative contents of the final report.

Then the task committee started to collect the data, the

information, and the viewpoints.

This is a progress report to present the direction of

the activity and achievement so far.

2. Tentative Contents of Final Report

The Task Committee-A de t erm ined tentatively the contents of

the final report, included as an Appendix to this report.

The Task Committee is presently engaged in collecting information

and data and discussions to complete the report along the Hues

of the proposed contents.

3. Considerations for Establishing Policy

People related to strong-motion earthquake records can be

divided into two groups. The first group consists of people

engaged in the installation and operation of instruments, data

processing, and dissemination of the records. This group will

be called, tentatively in this report, the disseminators group.

The other group consists of people who are using the records in

practice or research. This gr'oup will be called the users

380
group. Some people may belong to both groups.

The conditions for record dissemination desired by the

disseminators group may not harmonize with those of the users

group. If the policy for the records dissemination fails to

encourage the disseminators group or at least to be accepted by

them, people of the group will be discouraged in continuing and

expanding their efforts. This may cause contraction of the

observation activity and as a result less opportunity for the

users group to obtain the strong motion records of future

earthquakes. On the other hand, if the policy considers too much

the conditions requested by the disseminators group, people of

the users group may meet inconveniences or impossibilities to

obtain the records. This may cause inefficiency in utilizing the

strong motion records and progress in earthquake engineering.

Therefore, the policy should consider the conditions requested by

the disseminators and users groups in appropriate balance.

The Task Committee is collecting the possible conditions

from both groups and evaluat ing them. The conditions collected

by this time are presented in the following subsections. The

conditions are just "collected conditions" and are not "selected

conditions for the policy." So, some of them may be

contradictory to each other or overlapping each other, at this

moment

3.1 Conditions, from Users’ Viewpoint

The users hope that:

1) After an significantly large earthquake, the records are

made available as many as possible and as soon as possible.

2) At least, the waveforms and the peak values are made

available immediately after an significantly large

earthquake

3) Whether a field investigation team on seismic damage has had

information or not on the ground motions which caused the

damage makes big difference in fruitfulness of the

investigation. Therefore, copies of the records are

available to all the field investigation teams at an

appropriate time.

4) For planning and design of restoration of damaged

structures, the strong motion records are very important.

High priority is given for such users.

5) Data bases of the strong motion earthquake records are

provided and they can be accessed through the telephone

cab 1 e

6) When responsibility to the damage is questioned to the

designer or the constructor, the strong motion earthquake

record is very important information for the judgement. The

record should be made aval labe to the people in concern.

381
< ) magnitude to

program, the

the possible aftershock of the largest

organize rescue activities and preparedness

strong motion earthquake record is very

helpful

8) Organizations such as World Data Center, NOA A is

es tabl i shed

9) Instrument characteristics are made available together with

the strong motion earthquake records.

10)

Information is attached to the record on the data processing

details, such as procedures of the base-line correction, the

instrument correction, the interpolation of data, etc.

3.2 Conditions from Disseminators’ Viewpoint

The disseminators consider that:

1) Some of the strong motion networks are operated by

researchers and their final tasks are to produce research

reports which contribute to the scientific progress and to

the practical application, not to disseminate the records.

For such cases, the priority for the network operator to

use the records for his research purpose has to be

guaranteed. Otherwise, the records are made public after

certain periods from the recovery. (Some researchers may

analyze the records in a very similar way which the operator

has had in his mind, when the operator is very busy to

process and disseminate the records.)

2) Following case may happen. A researcher operator obtained

some significant records and he analyzed and published a

report. Then, he made his records public. However, he had

intention to make some statistical analysis on all his

records, when the records would reach at volume of his

satisfaction. Unfortunately, other researchers made a

similar statistical analysis and presented an empirical

formula such as a magni tude-epicentral distance-peak

acceleration relationship, while the record owner is waiting

for additional records. The researcher operator lost

oppo r tun ity to present his own f ormu la although he has his

own records

3) Observation, processing, and dissemination of the strong

motion records requires high knowledge and time consuming

efforts of the disseminators. This effort of the high

knowlegable people is not sufficiently respected by the

users. The user’s recognition on this point has to be

improved for promoting the wide and quick circulation of the

strong motion earthquake records.

4) The disseminators willingly make their records public at the

382
earliest possibility, if the priority for the researcher

operator to make his report on his records and the disseminators'*

efforts are adequately recognized.

5) The disseminators wish to know all the users who used their

records to ensure future information distributions such as

correction of the records or the instrument characteristics.

6) The researcher operators are not working full time for the

strong motion records. Their time is limited. Therefore,

the records from the network operated by the researcheroperator can not be made public immediately. Strengthening

of personnel and expenditure is essential.

7) Records of structural response should be made public after

careful consideration. Some researchers who are not familiar

with such structures may make an inadequate interpretation

and judge the structure unsafe, although it will be judged

safe by the researchers familiar with such structure. Inadequate

interpretation and its public release may cause social problems.

Before the public release, the agreement of the structure's

owner should be confirmed.

8) Some researchers are applying corrections to the records

obtained by others and making a data base. Even if such a case,

the priority and recognition of the original records should be

guaranteed

9) Stronger requests to make the records public are made to the

government agencies than to the private organizations. The

records owned by private organizations are as useful as the

government agencies. The request should be made equally to

the government agencies and to the private organizations.

10)

The users should give the original record owner credit for

his effort in an appropriate wav, such that the source of the

records is clearly described in a reference list of the report.

o

383
Appendix

Task Committee Report

on

Policy for International Dissemination

of

Strong-Motion Earthquake Records

Contents (proposed)

1 . Introduction

2. Current Dissemination of Records

2.1 U . S . Records

a. General

b. USGS Records

c. CDMG Records

e. University Records

d. Records Funded by NSF

e. Others

f. Guidelines in U.S.

2.2 Japanese Records

a. General

b. NRCDP Records

c. PHRI Records

e. PWRI Records

f. BRI Records

g. University Records

h. Others

i. Guidelines in U.S.

2.3 Records of Mexican Earthquake in 1984

2.4 Records of Chilean Earthquake in 1984

2.5 Records of Other Major Earthquakes

a. Records of Yugoslavia

b. Records of USSR

c. Records of People's Republic of China

d. Others

3. Data Base of Strong-Motion Earthquake Records

(Record coverage, operator, users, conditon for use,

e tc . ) 3.1 Data Bases in U.S.

3.2 Data Bases in Japan

3 . 3 Others

4. Considerations for Establishing the Policy

4.1 Considerations from Users' Viewpoint

charge

384
CJ

CO Considerations from Network Operators’ Viewpoint

Considerations to be included in the Policy

4.3.1 With respect to researchers engaged in

observation, processing, and dissemination

4.3.2 Priority of record owners in making research

reports

5. Proposed Policy

6 . Problems for Further Study

385
Report of Task Committee on

(B) LARGE-SCALE TESTING PROGRAM

Date : May 13, 1987

Place : Public Works Research Institute, Ministry of Construction

Toyosato-machi , Tsukubagun, Ibarakiken , 305 Japan

Japan Side - K. Ohtani (Chairman) ( NRCDP

S. Noda ( PHRI

S. Okamoto (BRI)

H. Yamanouchi (BRI)

T. Murota (BRI)

Y. Yamazaki (Observer) (BRI)

U. S. Side - H. S. Lew (Chairman) ( NBS

R. J. Smith (COE)

I

Activities and Principal Accomplishments to Date

The main activities related to this task committee during the

past year are as follows:

1. The Joint Technical Coordinating Committee on the Masonry

Research (JTCCMAR) was held in Keystone, Colorado, in

September, 1986. Coordinated research projects on masonry

structures were discussed.

2. Active exchanges of U.S. and Japanese researchers have

taken place under the Joint Program of Masonry Structures

and have been very useful.

3. Testing of a full-scale three-story masonry planar frame

was carried out at the BRI in 1986.

4. Construction of a full-scale five-story masonry building

has been initiated at the BRI.

5. Research grants have been provided by the National Science

Foundation to U.S. research institution. Currently 25

individual research projects are in progress and being

coordinated by the U.S. TCCMAR.

II . Future Programs

The status of the U.S. -Japan Joint Program of Masonry Structures

was reviewed. The Task Committee adopted the following

resolutions

1. To assure effective exchange of information on masonry

structures, Task Committee recommends continuation and

expansion of personnel exchange under the coordinated

research program.

386
2 . Task Committee discussed topics of future joint research

projects. Task Committee will facilitate continued

deliberation of the following topics between appropriate

U.S. and Japan organizations.

a. Seismic performance of precast concrete construction.

b. Seismic performance of composite and mixed

constructions

c. Application of high strength construction materials,

such as concrete and steel, to structures in high

seismic zones.

d. Application of base isolation and controlled damping

systems to critical structures located in high

seismic and wind zones.

3. The Meeting of the Joint Technical Coordinating Committee

on Steel Structures Research is scheduled to be held in

Tokyo, July 1987 to summarize the Joint Research Project on

steel structures.

4. The 3rd meeting of the Joint Technical Coordinating

Committee on Masonry Research is scheduled for Fall, 1987

in Japan.

5. Testing of a full-scale five-story masonry building will

begin at the BRI, November, 1987.

6. Task Committee (B), Large-Scale Testing Program and Task

Committee (D), Evaluation of Performance of Structures

should maintain closer liaison to facilitate the planning

of instrumentation and analysis for full-scale test

structures

7. Exchange of information on large-scale testing facilities

and large-scale testing programs should be encouraged.

387
Report of Task Committee on

(C) REPAIR AND RETROFIT OF EXISTING STRUCTURES

Date: May 14, 1987

Place: Public Works Research Institute, Ministry of Construction

Toyosato-machi , Tsukuba-gun, Ibaraki 305, Japan

Attendees: Japan Side - M. Hirosawa (Chairman) (5RI)

M. Shinohara (FWRI)

S. Nakata (Observer) (BRI)

U. S. Side - H.S. Lew (Temporary Chairman) (N3S)

I

Activities and Principal Accomplishments to Date

1. Both U.S. and Japan investigation teams provided technical

assistance to repair and retrofit of structures damaged during

the 1985 Mexico Earthquake, and related reports have been

published.

2. Research grants have been provided by the National Science

Foundation to U.S. institutes to carry out research projects on

repair and retrofit technology.

3. Manual of Repair Methods for Civil Engineering Structures

Damaged by Earthquakes was published by the Ministry of

Construction.

4. The fourth workshop on repair and retrofit of structures was

held on May 8 and 9, 1987 at the BRI. The outlines of the

workshop and its conclusions are in Appendix.

5. Japanese side provided the U.S. side the Japanese handbook on

repair and retrofit of reinforced concrete buildings.

II . Future Programs

1. Continue and expand the exchange of reports, books and special

publications related to repair and retrofit of existing

structures

2. As a result of successful completion of the fourth workshop on repair and retrofit, the fifth workshop is proposed. Possible

date might be in 1988 UJNR meeting to be held in U.S.

3. Compile information on specific projects such as evaluation of

particular designs and cost of completed projects.

388
4 e Devote increased attention to developing methods to assess

nonstructural items in existing structures which could pose

severe hazards and/or cause loss of service. Emphasis should

be placed to develop methods for retrofit or repair of

important nonstructural elements.

5. Carry out research to evaluate the potential structural

benefits and relative economics of using base isolation and

supplemental damping devices for retrofitting existing

structures. Such devices may be of particular interest fcr

historical or monumental structures in high seismic zone.

6. Maintain liaison with task committee D, "Evaluation of

Performance of Structures" and J, "Wind and Earthquake

Engineering for Transportation Systems" to exchange

information relative to repair and retrofit.

389
Append i x

Workshop on Repair and Retrofit of Existing Structure

Hay 0-9. BRI

\c t i v i ties

5 sessions- 18 presentations

General Discussion on issues

Concluding Remarks

1.

Similarities of. current techniques for repair and strengthening indicate

the opportunities for fruitful exchange of ideas. Commonality of techniques

and solutions make such exchange and coordination of activities particularly

beneficial in the area of repair and strengthening activities.

2.

Presentations emphasized concerns related to performance of elements and

connections, need for quality control, cost of construction.

3.

There appears to be a need in both countries to stimulate technical

discussions on repair and strengthening. This workshop provided a means for

initiating further discussions within each country and the desirability of

similar joint meetings in the future.

390
Areas of future research activi ty

1. Most pressing need is for experimental verification of repair and

strengthening techniques in the following areas,

a ) Use of new materials

b ) Development of innovative techniques

c ) Evaluation of foundation effects

d ) Influence of member response on overall structure response

e ) Implementation of analytical techniques calibrated from experimental

resul ts

2.. Need to continue efforts for common understanding of performance,

design, and construction data. Documentation of data and exchange of

information should be carried out to permit development of a data base

in area of repair and strengthening.

3. There is a need for studies related to damage control for prevention of

environment hazards and for maintenance of operations in existing

s tructures

4. Discussions between researchers and designers should be continued for

maximum benefit from activities underway in each country.

a ) Benchmark structures could be used for comparison of techniques and

design procedures.

b ) Typical methods used need to be discussed in detail.

Examples - bracing systems, infill walls, jacketing columns. It is

proposed that future workshops consider a particular topic in detail

and prepare a state-of-the-art report for use by designers and

researchers

391
(D) EVALUATION OF STRUCTURAL PERFORMANCE

Date

May 14,1987

Place

Public Works Research Institute, Ministry of Construction

Toyosato-machi , Tsukuba-gun, Ibaraki 305, Japan

Attendees: Japan Side - S. Okamoto (Chairman) (BRI)

(BRI)

( PWP.I

)

(BRI)

(BRI)

Y. Ishiyama

M. Okahara

Ho Yamanouchi

Y. Yamazaki (Observer)

U. S. Side - N.J. Raufaste (Temporary

R. J. Smith

R.N. White (Observer)

M. Miyahara (Observer)

Chairman) (NBS)

( ARMY

( Cornell Univ .

( NSF , Tokyo)

I . Activities and Principal Accomplishments to Date

Task Committee D organized three U.S. -Japan workshops on the

evaluation of the performance of existing buildings for resistance to

earthquakes. The First Workshop on "Seismic Performance of Existing

Buildings" was held in Japan in 1983. The Second Workshop was held

in the U.S. in 1984. The Third and Final Workshop was held in Japan

in 1985, and concentrated on comparisons of evaluations of benchmark

structures and on updating information on methodologies by each

Comparisons reveal that the various methods are suitably accurate for

the expected level of evaluation for buildings with good arrangement

of lateral strength elements, and that a general idea of expected

damage level can be estimated. However, additional evaluations are

recommended to validate and improve the methods to make them more

useful, especially for buildings of unusual geometry or with special

secondary elements, and to include effects of soil- structure

interaction

The proceedings of the three workshops were published in three

volumes* (1983, 1984, 1985) and were summarized in two papers: (1)

"Performance of Existing Buildings: Results of Benchmark Structure

Analyses Performed in U.S. -Japan Workshops", by R.N. White and P.

Gergely, Proc. of Third ASCE EMD Conference on Dynamic Response of

Structures, Los Angeles, April 1986, and (2) "Seismic Performance of

Existing Buildings", by P. Gergely, R.N. White, and G.R. Fuller,

Proc. of Third U.S. National 'Conference on Earthquake Engineering,

Charleston, S.C., August 1986.

* Available from R.N. White, Cornell University, Ithaca, N.Y. 14853

country

392
A status report was presented on development of ATC-14 report,

"Methods for Evaluating the Seismic Strength of Existing Buildings,"

and ATC-20, "Development of Procedures for Post-Earthquake Safety

Evaluation of Buildings" was described.

Discussion concluded with a presentation on the status of various

related projects, objectives, and possible future activities to be

pursued

II . Future Programs

The Task Committee reconfirmed its objectives and activities at the

May 14, 1987 meeting and agreed to the following:

1. Encourage the continued exchange of literature, analyses, and

research reports relative to assessment of seismic resistance of

buildings

2. Promote the dissemination of research findings to the other task

committee chairmen and to the design and construction community.

3. Expand the catalog of benchmark existing buildings in each country

that have been evaluated and instrumented.

4. Continue to evaluate other buildings in Japan, U.S. and other

countries where data are available using Japanese and American

methods of analysis including provisions of ATC-14.

5. Improve screening and evaluation methods for determination of

seismic resistance of existing buildings, and identify research

needs to improve evaluation technology.

6. Continue to plan joint activities with Task Committees B and C by

encouraging summer meetings of the three chairmen in each country.

7. Participating government agencies in UJNR with missions in

evaluating structural performance will be contacted by the Task

Committee chairmen in both countries for continued support for the

Task Committee activities.

393
Report of the Task Committee on

(E) NATURAL HAZARD ASSESSMENT AND MITIGATION THROUGH LAND USE PROGRAMS

Date : May 13, 1987

Place: Public Works Research Institute, Ministry of Construction,

Toyosato-machi , Tsukuba-Gun , Ibaraki-ken, 305 Japan

Attendess

Japan Side - K. Kawashima (Chairman) (PWRI)

S. Hattori (BRI)

M. Nagaoka (GSI)

Y. Sugimura (BRI)

U. S. Side - A. G. Brady (Temporary Chairman) (USGS)

L.R.L. Wang

(Old Dominion University)

I

Activities and Principal Accomplishments to Date

1. The Geographical Survey Institute (GSI) of Japan has been

developing mapping procedures for natural hazard prediction.

The Public Works Research Institute (PWRI) of Japan has revised

a procedure for assessing direct and indirect losses from

earthquake disasters, and the direct and indirect losses were

evaluated for the Miyagi-ken-oki eathquake of 1978 and the

Nihon-kai chubu earthquake of 1983.

The Building Research Institute (BRI) of Japan is developing

automatic computer programs for analyzing the expected ground

intensity at an arbitrary sire, and has made analyses on ground

response characteristics.

The National Research Center for Disaster Prevention of Japan

has been developing a seismic microzoning method using an urban

ground database, and applying satellite remote sensing

technology to natural disaster prevention.

2. The U.S. Geological Survey conducts a broad based program in

natural hazards reduction. This program includes assessment

of: (1) earthquake ground motion; (2) soil liquefaction

potential; (3) landslide potential; and (4) earthquake economic

losses

3. Several papers pertaining to Task Committee (E) were included

in the proceedings of the 19th Joint Panel Meeting:

a. Theme II "Earthquake Engineering"

b. Theme III "Storm Surge and Tsunamis"

II . Future Programs

1. Both sides will continue to exchange technical information on the following subjects:

394
a. Evaluation of seismic hazard and mapping procedures.

b. Evaluation of damage potential of structures, soil

liquefaction potential and tsunami hazard.

Co Seismic response characteristics under various ground and

topographic conditions.

d. Planning methodologies for earthquake disaster mitigation in

large populated cities.

e. Evaluation of socio-economic losses from earthquake

disasters

2. Both sides will continue to promote coordinated research and

exchange of specialists in natural hazard assessment and

mitigation through land use programs.

3. The U.S. chairman will furnish to task committee members future

reports on natural hazard assessment and mitigation through

land use programs being coordinated or sponsored by the Federal

Emergency Management Agency (FEMA), the Interagency

Coordinating Committee of the National Earthquake Hazards

Reduction Program (ICC/NEHRP), the U.S. Geological Survey

(USGS), the National Science Foundation ( NSF ) , the National

Academy of Science (NAS) and National Academy of Engineering

(NAE)

The Japanese chairman will perform his similar responsibilities

for reports from National Research Center for Disaster

Prevention (NRCDP), Geographical Survey Institute (GSI), Public

Works Research Institute (PWRI), Building Research Institute

(BRI), and Port and Harbor Research Institute (PHRI).

4. Each member of the Task Committee shall provide to his chairman

in time for a report at the uGcn Joint Meeting in 1988, a one

or two page description of his contributions to a projected

workshop in May 1989. The contributions are to come from the

subjects in paragraph II-l. above, and are to include an outline of research and expected results (in the form of

publications, etc.) The two chairmen will compile these brief

descriptions into a report to the Panel at the 20th Joint

Meeting in May 1988.

395
Report of Task Committee on

(F) DISASTER PREVENTION METHODS FOR LIFELINE SYSTEMS

Date : May 14, 1987

Place: Public Works Research Institute, Ministry of Construction

Toyosato-machi , Tsukuba-gun , Ibaraki 305, Japan

Attendees : Japan Side - T. Iwasaki (Chairman) (PWRI)

K. Kawashima (PWRI)

H. Sugita (Observer) (PWRI)

U. Sc Side - Jo V. Tyrrell (Temporary Chairman) (NFEC)

L. R.L. Wang (Old Dominion University)

J.H. GATES (Cal Trans)

I

Activities and Principal Accomplishments to Date

1. In August, 1986 Task Committee Co-Chairman, Dr. T. Iwasaki

(PWRI) attended the 3rd U.S. National Conference on

Earthquake Engineering, Charleston, South Carolina. A

planning conference was held during his visit and was

attended by Dr. Iwasaki, Dr. S. C. Liu, Dr. Leon Wang, Dr.

Masanobu Shinozuka, and others. The possibility of a U.S.-

Japan Workshop on Lifeline Systems was discussed.

2. A technical report by PWRI entitled, "The Manual of Repair

Methods for Civil Engineering Structures Damaged by

Earthquakes (in Japanese)" was made available to NSF. In

the Manual procedures of damage surveys and repair works

for civil engineering structures including buried sewer

pipelines are described. An outline of the results of this

research project was introduced at the 18th UJNR Joint

Meeting of the Panel on Wind and Seismic Effects. PWRI

authorized NSF for its translation into English and for

distribution to the U.S. professional community.

Currently, the document is being considered for translation

by NSF.

3. Complete copies of the Manual report (Japanese version)

were distributed to concerned individuals of U.S. and Japan

during the 19th UJNR Joint Meeting. To introduce the

Manual training courses were held in Tokyo, Sendai, and

Nagoya in April, 1987, in which about 1,000 practicing

engineers participated.

4. PWRI is also conducting a dense instrument array

measurement program at PWRI campus and in Shizuoka

Prefecture to identify ground motions applicable to seismic

design of buried pipeline structures. An outline of the

dense instrument array program was presented at the 18th

UJNR Joint Meeting.

396
The Ministry of Construction is promoting the construction

of cable box systems (CAB systems) under urban highways, in

order to move aerial cables (for electricity and

telecommunication) into underground cable boxes, and has

prepared Technical Standards for CAB Systems (Draft) . In

conjunction with these activities, PWRI has proposed a seismic design procedure of CAB Systems and conducted

dynamic experiments for examining the seismic safety of CAB

systems. An outline of the Technical Standards for CAB

systems (Draft) and the dynamic experiments was reported

during the 13th UJNR Joint Meeting.

5. In U.S., a study program to assess the design and operating

standards for lifeline systems has been initiated by the

Building Seismic Safety Council (BSSC) under the support of

the Federal Emergency Managment Agency (FEMA) , with the

broad participation of ASCE members and other

professionals. A workshop on abatement of earthquake

hazards to lifelines was held in November, 1986 in Denver,

Colorado, and the proceedings will be available and

distributed soon.

6. NSF has continued the support of a number of research

projects dealing with lifeline earthquake engineering.

Among them, experimental results and analytical results

were presented during the 19th UJNR Joint Meeting,

The strong motion array project (SMART) and the soil

response measurement project have been continuously

operated and new data has been recorded. The data recorded

from these projects is being continuously exchanged between

the two countries.

II . Future Programs

1. The members of the Task Committee will facilitate the

continued exchange for technical information and

cooperation in research on specialized problems listed as follows

a. Survey reports of seismic damage to lifeline systems

and seismic observation results

b. Seismic design procedures for lifeline systems

c. Procedures and instrumentation to detect and inspect

damage and the behavior of lifeline structures during

earthquakes

d. Repair and retrofit methodology for lifeline systems

e. Estimation of reliability of lifelines

f. Management and public education on importance of

lifeline systems

g. Investigation of needs for large-scale testing of

lifeline systems

397
2

It is recommended to continue U.S. -Japan coordinated

research projcts on field observations and analyses on

seismic behavior of buried pipelines, experiments and

analyses on dynamic characteristics of joint behavior,

effects of soil liquefaction on buried pipes, soil-buried

pipes interaction, seismic reliability of lifeline systems,

and seismic design methods. As for field observation, the

Japan side will concentrate on seismic behavior of

pipelines buried in soft soils and in boundary layers

between hard and soft soils, whereas the U.S. side will

focus on dynamic behavior of pipelines buried in and near

active faults.

3. To promote U.S. -Japan coordinated research programs in

earthquake resistance of lifeline facilities, it is

recommended to plan a U.S. -Japan Workshop on lifeline

earthquake enginering in 1989. Time, location and topics

of this Workshop will be discussed between relevant

individuals of both countries. The details of this

Workshop will be finalized during the 20th UJNR Joint

Meeting to be held in Washington, DC, in May 1988.

4. The Task Committee recognizes the importance and potential

impact to the engineering practice of the Manual of Repair

Methods for Civil Engineering Structures Damaged by

Earthquakes (Draft) (in Japanese) prepared by PWRI, and

recommends that the current translation of the document

into English be given a high priority and the document be

made available for wide distribution as soon as possible.

5. Drs . T. Iwasaki and K. Kawashima of PWRI and Dr. Leon Wang

of Old Dominion University are proposing a joint research

project on seismic component design, analysis and

experimental observation of buried pipelines including the

effects of pipe-structure . interaction , and seismically

induced soil liquefaction.

6. Dr. T.D. O'Rourke of Cornell University is visiting Japan

in the beginning of June 1987 to discuss with Japanese

experts a possible coordinated research project for

studying the effects of large ground movements on buried

pipelines

398

Report of Task Committee on

(G) WIND CHARACTERISTICS AND STRUCTURAL RESPONSE

Date : May 14, 1987

Place: Public Works Research Institute, Ministry of Construction

Toyosato-machi , Tsukuba-gun, Ibaraki-ken 305, Japan

Attendees : Japan Side - K. Kurashige (Chairman) (MRI)

T. Fuji tan i (MRI)

T. Murota (BRI)

N. Narita (PWRI)

K. Yokoyama (PWRI)

U. S. Side - C.F. Galambos (Temporary Chairman )( FHWA

I

Activities and Principal Accomplishments to Date

Noting that:

a. High winds frequently cause loss of life and

extensive property damage.

b. The knowledge derived from the exchange of high wind

date and information on wind effects can be useful in

reducing loss of life and property damage,

c. Insufficient knowledge exists on the effects of high

winds on structures and methods for modeling these

effect, and

d. The needs for wind data for various applications and

new projects are increasing.

The Task Committee carried out the following activities:

1. The following document was exchanged:

Report of Observation at the Meteorological Observation

Tower, No. 2, Meteorological Research Institute, Tsukuba

Science City, Japan.

2. Measurements of high winds on 20 to 35m high towers at

Kashima coast were completed on March, 1987. A new wind

observation project is under preparation by Meteorological

Research Institute at another coast near the point stated

above

3. National Science Foundation funded a study to obtain

comprehensive measurements of surface pressures and

internal pressures acting on a full-scale low-rise building

situated in open terrain near Lubbock, Texas.

4 „ Concerning high wind effects, particular investigations

continued in the U.S.A. on various projects, such as wind

loading and strength prediction of glass cladding and wind

loading and response of transmission towers.

399
5.

Research needs were discussed as follows:

a. Comparison of design methodology for wind resistance

of structures in both countries.

b. Comparison of wind tunnel test techniques using a common prototype to develop test models and

techniques

c. Comparison of design philosophies, codes and practice

in both countries.

d. Better definition of the effect of internal pressure

in buildings especially with large internal open

spaces

e. Further development of actively controlled damping

systems for full-scale buildings.

f. Assessment of the influence of material deterioration

on wind effects.

g. Development of fundamental wind data in offshore

areas

I I . Future Programs

1. Exchange observation reports of high winds.

2. Encourage interaction between meteorologists and engineers

to identify the types of wind data required for x* v-l t x' e use

in establishing extreme wind distributions; in determining

wind loading on buildings and structures; in understanding

the urban wind climate; and in considering design issues

involved with the wind generation of energy.

3. Encourage the establishment of minimum requirements for the

simulation of atmospheric boundary layers in wind tunnels

and the exchange of boundary layer wind tunnel test

results, including comparisons with corresponding fullscale situations.

4. Exchange information on criteria, techniques and

instrumentation for structural modeling in boundary layer

wind tunnels. Also encourage the exchange of information

of measurements made on actual structures and on

comparisons with predictions from wind tunnel studies of

these structures.

5. Expand the exchange of engineers and meteorologists, and

encourage the mutual use of available research facilities

in the two countries.

6. Encourage the development of work plans for the pursuit of

the research needs listed in 1.5.

7. Plan a workshop in the coming year, with details to be

worked out between the respective U.S. and Japanese Task

Committee chairmen.

400
Report of Task Committee on

(H) SOIL BEHAVIOR AND STABILITY DURING EARTHQUAKES

Date : May 13, 1987

Place Public Works Research Institute, Ministry of Construction

Toyosato-machi , Tsukuba-gun , Ibaraki 305, Japan

Attendees: Japan Side - Y. Sasaki (Chairman) (PWRI)

Y. Koga (PWRI)

H. Tsuchida (PHRI)

0 . Matsuo (Observer) (PWRI)

H. Matsumoto (Observer) (PWRI)

U. S. Side - A. G. Franklin (Chairman) (WES)

R. R. Ledzian (USBR)

I. Activities and Accomplishments to Date

Exchange of relevant documents on soil behavior during

earthquakes was made throughout the year. During the

Nineteenth Joint Meeting of the Panel, seven papers were

submitted from the Japanese side and one paper was

submitted from the U.S. side.

2. Under the auspices of the UJNR, the following researchers

were exchanged between the U.S. and Japanese sides:

a. Mr. Jeffrey A. Farrar of the USBR was hosted by the

PWRI in October and November 1986.

b. Mr. Takashi Tsuchida of the PHRI was hosted by WES

from November 1985 to November 1986.

c. Dr. Randall W. Jibson of the USGS was hosted by the

PWRI from January 1987 to March 1987.

d. Mr. Yoshitaka Hachiya of the PHRI is being hosted by

the WES for one year beginning in March 1987.

I I . Future Programs

1. Proceedings of the 1985 workshop on In-Situ Testing for

Evaluation of Soil Liquefaction Susceptibility will be

published

2. The committee recommends that a workshop on remedial

measures to mitigate or prevent damage due to liquefaction

failures be held in conjunction with the twentieth Joint

Meeting, as a first step in a future cooperative research

program

401
3 . An exchange of available documents on current status and

recently completed research concerning remedial measures to

mitigate or prevent damage due to liquefaction will be made

during the summer of 1987, in preparation for the

recommended workshop. This exchange will be coordinated by

the U.S. and Japanese chairmen.

4. When requested, the Task Committee will assist in

arrangements for visits to earth embankments including fill

dams and foundations which have been subjected to

significant ground motions.

5. The Task Committee encourages the maximum exchange of

information on the following areas of earthquake

engineering which are currently of intense interest to both

U.S. and Japanese engineers.

a. Liquefaction potential of soils

b. In-Situ testing methods for evaluation of soil

liquefaction potential

c. Seismic stability of embankment structures and

natural slopes.

d. Remedial treatment of embankments and foundations to

improve seismic stability

e. Seismically induced permanent displacements in earth

structures

f. Seismically induced settlements of soils and

foundations

g. Earthquake effects on semi-buried concrete structures

h. Earthquake effects on retaining walls

i. Centrifuge testing facilities and methods

j . Laboratory testing of models of geotechnical

structures

k. Field performance data showing full-scale seismic

response of earth dams

402
Report of Task Committee on

(I) STORM SURGE AND TSUNAMI

Date

May 13, 1987

Place: Public Works Research Institute, Ministry of Construction

Toyosato-machi , Tsukuba-gun, Ibaraki 305, Japan

Attendees

Japan Side - T. Uda (Chairman) (PWRI)

K. Tanimoto ( PHRI

K. Kurashige (MRI)

M. Okada (Observer) (MR I)

U. S. Side - H. Meyers (Temporary Chairman) (NOAA)

J. Tyrrell. ( NFEC

I

Activities and Principal Accomplishements to Date

1. The U.S. and Japanese sides have continued to colaborate on

tsunami propagation studies. Mr. Okada of MRI/JMA has used the

bathymetric data received from NGDC/NOAA to produce preliminary

propagation charts. These results were reported as part of the

19th Joint Meeting.

2. Mr. Okada has sent to NGDC/NOAA a copy of this computer model

for tsunami propagation.

1 1 . Future Programs

1. The Task Committee supports the continuation of exchange of

data and information on Tsunami and Storm Surge. The Committee

also supports exchange of researchers whenever possible.

2. The Task Committee recommends a Second Tsunami and Storm Surge

Workshop to be held in U.S. in connection with the 20th Panel

Meeting. Details of the workshop will be arranged by the Task

Committee Chairmen with the approval of the Panel Chairmen.

Specifically the workshop will address computational and

experimental methods in tsunami and storm surge.

3. Mr. Okada will use his model and data to produce a Pacific-wide

tsunami travel chart. This work is very valuable for the U.S.

and other pacific countries because it will result in an improvement in tsunami warnings.

4. NEDL will test Mr. Okada ' s model and compare results with

actual observations in the NOAA data base.

5. The U.S. side (NOAA) will give to the Japan side the report and

documentation of the Storm Surge Model, SLOSH (Sea, Land, and

Overland Surges from Hurricanes).

403
Report of Task Committee on

(J) WIND AND EARTHQUAKE ENGINEERING FOR TRANSPORTATION SYSTEMS

Dat e : May 13, 1987

Place : Public Works Research Institute, Ministry of Construction

Toyosato-machi , Tsukuba-gun, Ibaraki-ken 305, Japan

Attendees: Japan Side - H. Shinohara (Chairman) (PWRI)

T. Iwasaki (PWRI)

K. Yokoyama (PW'RI)

H. Sato (Observer) (PWRI)

U. S. Side - C.F. Galambos (Chairman) (FHWA)

J. H Gates (State of California)

I

Activities and Principal Accomplishments to Date

The major accomplishments of Task Committee (J) during the past year

were as follows:

1. Summaries of wind and earthquake research on bridges conducted

by both countries were presented to the 19th Joint Meeting, and

exchange of information was made.

2. The coordinated research study on the seismic performance of

bridge piers and columns was begun in 1983, and the following

experiments were conducted.

a. Performance of reinforced concrete piers and columns

subjected to dynamic cyclic loading

b. Model tests on the failure of reinforced concrete piers

c. Testing full-scale concrete columns and associated

small-scale model tests

d. Behavior of concrete filled steel tubes

3. Mr. Koichi Minosaku, researcher of Bridge Division, PWRI, MOC,

is a guest researcher at the Center for Building Technology,

NBS, U.S.A. for 10 months beginning in September, 1986, to

perform cooperative experimental research on the seismic

performance of bridge piers and columns. This assignment is a part of the personnel exchange program of the UJNR Panel.

4. The Third U.S. -Japan Bridge Workshop was held in Tsukuba

Science City, in May 8-9, 1987 with cooperation from

organizations concerned in both countries. Fourteen U.S. and

thirty-three Japanese members participated in the workshop.

Exchange of technical information was made and comprehensive

discussion were held on future cooperative programs. The

proceedings will be made and distributed to the participants of

the both countries. Some of the subjects were:

404
a. Dynamic analysis of cable-stayed bridges

b. Reliability in bridge design

c. Earthquake resistant design of highway bridges and

earthquake motion

d« Repair and retrofit of existing bridges

e. Wind resistant design of suspension bridges and

cable-stayed bridges

f. Mitigation of vibration of highway bridges

g. Seismic behavior of RC bridge piers

h. Diaphram wall foundations

II . Future Programs

As a result of comprehensive discussions, the Task Committee (J)

hereby resolves to carry forth with the following programs:

1. Continue with the coordinated experimental research study on

the seismic performance of bridge piers and columns, and review

the progress of the study at the intermediate stage.

2. Develop a coordinated research study, both theoretical and

experimental, on the seismic, aeroelastic, and aerodynamic

response of cable-supported bridges. Reciprocal wind tunnel

testing for bridges should be made in the institutes concerned

to evaluate and compare the wind tunnel testing methods. A

specific proposal for the project is to be developed by the

co-chairmen

3. Encourage the continued exchange of information and research

personnel through the Panel on Wind and Seismic Effects of

UJNR

4. The Task Committee endorses the initiation of a coordinated and

expanded program of research on remedial measures to mitigate

or prevent damage to existing bridges by the natural hazards of

earthquakes and strong winds.

5. The Task Committee proposes the fourth Bridge Workshop to

promote cooperative research on wind and seismic-resistant

design of long span bridges, retrofit and strengthening of

existing bridges. The Workshop will be held in U.S. in

conjunction with the 20th Joint Meeting of the Panel in 1988.

Details of the Workshop should be determined based on

discussion between the Task Committee Chairmen. Emphasis of

the workshop should be on practical applications. The workshop

is to be followed by a technical tour.

405
6 . Develop a coordinated research study to evaluate and compare

the seismic design criteria for bridges used in Japan and the

United States. Several bridges should be identified by each

group. Bridges identified by one group should be redesigned

for seismic loads only by other group using their particular

seismic criteria. The study should identify the differences

and similarities between the two seismic criteria and attempt

to identify common areas of interest. A specific proposal for

the project is to be developed by the co-chairmen.

406
NBS-1 14A irev. 2 -ec) U.S. DEPT. OF COMM.

BIBLIOGRAPHIC DATA

SHEET (See instructions

)

1. PUBLICATION OR

REPORT NO.

•88-3703

2. Performing Organ. Report No. 3. Publication Date

JANUARY 1988

4. TITLE AND SUBTITLE

Wind and Seismic. Ef f ects--Proceedings of the 19th Joint Meeting of the

U.S. - Japan Cooperative Program in Natural Resources Panel on Wind and Seismic

Effects

5. AUTHOR(S)

Noel J. Raufaste, Editor

6. PERFORMING ORGANIZATION (If joint or other than NBS, see instructions)

NATIONAL BUREAU OF STANDARDS

U.S. DEPARTMENT OF COMMERCE

GAITHERSBURG, MD 20899

7. Contract/Grant No.

8 . Type of Report & Period Covered

Final

9.

SPONSORING ORGANIZATION NAME AND COMPLETE ADDRESS (Street. City, State , ZIP)

10.

SUPPLEMENTARY NOTES

Document describes a computer program; SF-185, FlPS Software Summary, is attached.

11.

ABSTRACT (A 200-word or less factual summary of most significant information. I f document includes a si gn i fi cant

bi bl iography or literature survey, mention it here)

The Nineteenth Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects was

held at the Public Works Research Institute, Tsukuba, Japan, from May 12-15, 1987.

This publication, the proceedings of the Joint Meeting, includes the program, list of

members, panel resoulut ions , Panel charter, task committee reports, and technical

papers

.

Subjects covered in the papers presented to the panel include: (1) wind engineering,

(2) earthquake engineering, (3) storm surge and tsunami, (4) U.S. - Japan Cooperative

Research Program, and (5) results of recent task committee workshops.

12.

KEY WORDS (Six to twelve entries; alphabetical order; capi to li ze only proper names; and separate key words by semicolon s) accelerograph; codes: design criteria; disaster; earthquakes; earthquake hazards;

geotechnical engineering; ground failure: liquefaction; pipeline; seismicity; solids;

standards; structural engineering; structural response; tsunami; wind loads; winds

13. AVAILABILITY

Unlimited

| | For Official Distribution. Do Not Release to NTIS

j~ Order From Superintendent of Documents, U.S. Government Printing Office, Washington, D.C.

20402.

XX Order From National Technical Information Service (NTIS), Springfield, VA. 22161

14. NO. OF

PRINTED PAGES

427

15. Price

$36.95

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