Wind and Seismic Effects

Proceedings of the 23rd Joint Meeting

NIST SP 820

N4T L INST. OF STAND & TECH R.IX.

-QC

100

• U57

VIENT OF COMMERCE

#820 ite of Standards and Technology

1991
I
Wind and

Seismic

NIST SP 820

PROCEEDINGS OF

THE 23RD JOINT

MEETING OF

THE U.S.-JAPAN

COOPERATIVE PROGRAM

IN NATURAL RESOURCES

PANEL ON WIND AND

SEISMIC EFFECTS

Building and Fire Research Laboratory

National Institute of Standards and Technology

Gaithersburg, MD 20899

U.S. DEPARTMENT OF COMMERCE

Robert A. Mosbacher, Secretary

National Institute of Standards and Technology

John W. Lyons, Director

0CIOO

"2.

Effects

Issued September 1991

Noel J. Raufaste,

EDITOR
National Institute of Standards and Technology Special Publication 820

Natl. Inst. Stand. Technol. Spec. Publ. 820, 636 pages (Sept. 1991)

CODEN: NSPUE2

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1991

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, DC 20402-9325
PREFACE

BACKGROUND

Responding to the need for improved engineering and scientific practices through exchange

of technical data and information, research personnel, and research equipment, the United

States and JAPAN in 1961 created the U.S. -JAPAN Cooperative Science Program. Three

collateral programs comprise the Cooperative Science Program. The U.S. -JAPAN

Cooperative Program in Natural Resources (UJNR), one of the three, was created in January

1964. The objective of UJNR is to exchange information on research results and exchange

scientists and engineers in the area of natural resources for the benefit of both countries.

UJNR is composed of 17 Panels each responsible for specific technical subjects.

The Panel on Wind and Seismic Effects was established in 1969. Seventeen U.S. and six

Japanese agencies participate with representatives of private sector organizations, to develop

and exchange technologies aimed at reducing damages from high winds, earthquakes, storm

surge, and tsunamis. This work is produced through collaboration between U.S. and

Japanese member researchers working in 1 1 task committees. Each committee focuses on

specific technical issues, e.g., earthquake strong motion data. The Panel provides the

vehicle to exchange technical data and information on design and construction of civil

engineering lifelines, buildings, and water front structures, and to exchange high wind and

seismic measurement records. Annual meetings alternate between JAPAN and the U.S. (odd

numbered years in JAPAN; even numbered years in the U.S.). These one-week technical

meetings provide the forum to discuss ongoing research and research results; one-week

technical study tours follow the meetings.

The National Institute of Standards and Technology (NIST) provides the U.S. -side chairman

and secretariat. The Public Works Research Institute (PWRI), JAPAN, provides the JAPAN

Side chairman and secretariat.

Cooperative research is performed through formal Panel Programs. In 1981 cooperative

research in Large-Scale Testing was started under the auspices of the Panel. Also in 1981,

joint research on Reinforced Concrete Structures was initiated. Full-scale testing was

performed at the Building Research Institute (BRI), one of the six Japanese member

organizations, with supporting tests in JAPAN and in the U.S. Two years later a joint

research program on Steel Structures was initiated. Full-scale testing again was led by BRI

with supporting tests in the U.S. and JAPAN. The U.S. -JAPAN coordinated program for

Masonry Building Research was started in 1985. A U.S. -JAPAN coordinated program on

Precast Seismic Structural Systems was initiated in 1991.

Task Committee meetings, exchanges of data and information through technical presentations

at annual Panel meetings, exchanges of guest researchers, visits to respective research

laboratories and informal interactions between Panel meetings, joint workshops and seminars,

and joint cooperative research programs all contribute to the development and effective

delivery of knowledge that has influenced design and construction practices in both countries.

iii
Direct communication between counterpart country organizations is the cornerstone of the

Panel. Effective information exchanges and exchanges of personnel and equipment have

strengthened domestic programs of both countries. There are opportunities for experts in

various technical fields to get to know their foreign counterparts, conduct informal

exchanges, bring their respective views to the frontiers of knowledge, and advance

knowledge of their specialties.

The Panel's results have supported improvements in practices in both countries. They have:

1. created and exchanged digitized earthquake records for use as the basis of

research for Japanese and U.S. geotechnics and structures;

2. produced data that advanced U.S. design and construction of bridge columns;

3. produced large-scale testing data that advanced seismic design standards for

buildings;

4. created a database comparing Japanese and U.S. standard penetration tests to improve seismic design criteria for soil liquefaction;

5. created databases on storm surge and shore line interaction and on tsunamis

and tsunami warning systems for use by designers to verify mathematical

models of tsunamis and storm surge.

HIGHLIGHTS OF THE 23rd JOINT PANEL MEETING

The 23rd annual U.S. -JAPAN Joint Panel meeting was held at the Public Works Research

Institute, Tsukuba, JAPAN, during 14-17 May 1991. Eighteen U.S. members participated13 from Federal agencies and 5 from universities. Forty-two papers were authored~22 by

U.S. members and 17 by Japanese. Thirty-one papers were presented orally. The papers were organized into seven themes: Wind Engineering; Earthquake Engineering; Storm Surge

and Tsunamis; Performance of Nonstructural Systems; U.S. -JAPAN Cooperative Research

Program; International Decade for Natural Disaster Reduction; Summaries of the Task

Committee Workshops conducted during the past year; and Reports (oral presentation only).

Some highlights of the meeting include:

0 Vice Minister, Dr. MITANAT, Ministry of Construction (MOC) recognized

the importance of the work of the Panel. The panel activities was reported at the 13th Plenary Meeting of the U.S. -JAPAN Cooperative Program in Natural

Resources (UJNR) held on 10 May 1991, Tokyo, where progress of the

Program's 17 Panels was discussed.

iv
o A Press Briefing was held during the Panel's Opening Session to celebrate

NIST's gift to PWRI of a cutting from its Sir Isaac Newton Apple Tree. The

ceremony was covered by the national and local press.

o Useful information was gained about JAPAN'S public works projects and civil

engineering research and their application of research into practice:

discussion on new technologies to reduce building damages from

earthquakes e.g., response control systems; PWRI is leading a 2 year

base isolation program working with 30 private organizations and

leading a 5 year effort in active control also working with 30 private

organizations;

better understanding of Japan's building structural code requirements

and regulations;

predicting liquefaction potential using geomorphological mapping;

discussion on investigation of the 16 July 1990 Philippine Earthquake

by the Japanese team;

o The Panel developed a catalogue, State-of-Art and Recent Trend on

Engineering Technology for Reducing Wind and Earthquake Effects containing

title, keywords, and author index of over 1200 Panel papers published during

its 23 year life. Over 1000 copies will be distributed to developing countries

in support of the International Decade for Natural Disaster Reduction

(IDNDR).

0 An 11 -Panel Task Committee was established on Wind and Earthquake

Engineering for Offshore and Coastal Facilities.

o Six of the Panel's 11 Task Committee workshops are scheduled to be held

during the coming year.

o A Press Briefing concluded the technical sessions with national coverage of the

Panel's activities.

TECHNICAL SITE VISITS

Following the Joint Technical Meeting, the Panel members participated in a conference on

the International Decade for Natural Disaster Reduction (IDNDR) and visited construction

projects, disaster areas, and laboratories. These visits occurred during 18-23 May.

Highlights include:

IDNDR one-day symposium, "A Focus on Disaster Prevention 100 years After the Nobi

Earthquake" was held in Gifu City north of Nagoya in recognition of their disastrous

V
earthquake of November 1891. Six presentations were made: three from the U.S. -side Panel

(Hurricane Hugo, Loma Prieta Earthquake, and U.S Federal Agency Programs in Disaster

Countermeasures) and three from the JAPAN-side Panel (Nobi Earthquake, Effective

Communication to Local Population During A Disaster, and Disaster Preparedness in Gifu).

The Japanese organizers will publish a proceedings of the conference in Fall 1991. This

event received extensive press coverage. >

The Japanese Government Headquarters for IDNDR under the Prime Minister's Office took

the lead in developing a basic policy for the promotion of IDNDR in JAPAN. The MOC is charged with carrying out the policy because its administration is involved in reducing the

effects of natural disasters. The Panel on Wind and Seismic Effects, with MOC and other

Japanese government agencies, supported the idea to hold the IDNDR symposium to enhance

public interest and understanding in natural disaster prevention.

The Building Research Institute (BRI), MOC, described their research being conducted at

its several laboratories. At the Building Element Laboratory, tests on the seismic resistance

of beam-column connections of glued laminated timber construction were reviewed. A full- scale test of base isolation and other methods of improving seismic resistance were

demonstrated at the Field Test Site. At the Large Structural Test Laboratory, tests on

viscously damped shear walls were demonstrated in their four-story steel frame building, and

earlier tests conducted in the laboratory as part of the U.S. -JAPAN Collaborative Program

were discussed. The Structural Test Laboratory was visited where various tests on high

strength concrete materials and members were being conducted.

The Kumagaigumi Company Institute of Construction Technology (KICT), described

their Structural Dynamic Testing Laboratory, the Ocean Hydraulic Laboratory, and the

Environmental Testing Laboratory. For the first, tests on a visco-elastic damped shear wall were demonstrated and a videotape described the development of seismic response control

systems and other research involving earthquake resistance of tall buildings. In the second

laboratory two open channels and a test basin were observed. Their 40 m x 20 m x 1.5 m

(131 X 66 X 5 ft) deep basin is nearing completion and will be able to individually control

wave generation in each of its two open channel 0.5 m (1.6 ft) wide basins and to model

tidal conditions and coastal currents. In the Environmental Laboratory's super-clean room,

protocol measurement methods are developed for electronic components.

The Public Works Research Institute (PWRI), MOC, described its Structural Engineering

Laboratory's 26.4 MN (6 million lb) universal testing machine and specimens of deep steel

beams and columns that had been tested. At the Vibration Laboratory the problem of

liquefaction was discussed with a vivid model demonstration. At the Earthquake Engineering

Laboratory base isolation of bridges was the prime subject. Various types of bridge bearings were shown and a videotape described their use. PWRI's Boundary Layer Wind Tunnel is the largest in the world. Its test section is 41 m x 4 m x 30 m (135 x 13 x 98 ft). The

laboratory was built to test a 1:100 scale model of the 4 km Akashi-Kaikyo suspension

bridge. The Honsho-Shikoku Bridge Authority paid for the construction of this facility. A

variety of models of other major bridges will be tested in this facility during the coming 10

years.

vi
Tokyo Metropolitan Disaster Prevention Center (MDPC), is located in the new Tokyo

Metropolitan Government City Hall, Shinjuku-ku, Tokyo. The 55-story building is 243 m

(797 ft) high and 44 m (144 ft) wide. MDPC has centralized its command and control for

disaster response and relief operations. MDPC occupies three floors. Its mainframe

computer is isolated from the building by base isolators. MDPC provides the command and

control for the Tokyo metropolitan area. Disaster information is analyzed for damage

assessment and prediction of severity. Their map information system provides for two-way

communication to analyze the severity of the event and to communicate to the 64 disaster

relief centers within Tokyo. Construction of the City Hall was completed in April 1991

costing 180 billion yen ($1.3 billion). The MDPC facility cost 18 billion yen ($130 million).

Chubu High Speed Surface Transportation (CHSST) Corporation, Nagoya, manages this

experimental magnetic levitation, two-passenger car train ~ a candidate mass transportation

system. CHSST's objectives are to produce a low-noise and vibration and hydrocarbon free

propulsion system. The passenger car testing on their 1,5 km test track commenced in May

1991 and will continue through May 1993. CHSST estimates the cost per km for the twin

rail system will be $7.4 million.

Toyota Motor Corporation and Toyota Home, Toyota, is located near Nagoya. Included

in this visit was a review of some advanced automotive systems development, a tour of one

automotive production plant, and discussion about Toyota's residential home subsidiary -

Toyota Home. Toyota claims 5% of its automotive sales are returned to perform R&D. A

greater amount of research soon will be performed in their Belgium and U.S. laboratories.

Toyota's prefabricated home business is based on its automotive technology. About 85% of

the home is fabricated at Toyota's plant; 3-8 weeks are required to construct the home onsite (subject of weather conditions).

Second Nunobiki Road Tunnel, Kobe, is a 3 km 90-degree curved tunnel under

construction to reduce traffic congestion which has increased since the completion of the first tunnel in 1969. Construction started in 1989 with expected completion in March 1993. The

tunnel is cut from rock having compressive strength of 1000-2000 kg/cm^. The slot drilling

method is used for a 300 m (984 ft) section adjacent to the bullet train and the existing

highway tunnel. Progress is about 70 cm (2.3 ft) per 24-hour day. Blasting is performed in

the remaining 2.7 km (1.7 mi) tunnel. The cost of the tunnel is projected at $32 million.

The tunnel is being constructed by the Kobe City Public Corporation.

Akashi-Kaikyo Bridge, Kobe, is a 4 km (2.5 mi), 3-span, 2-hinged truss suspension bridge

with center span of 1,990 m (6,529 ft); each end span is 960 m (3,150 ft). It will be the

world's longest bridge when completed in 1998. The bridge under construction by the

Honshu-Shikoku Bridge Authority will connect Awaji Island to Honshu Island crossing the 4

km (2.5 mi) Akashi Straits. This bridge is part of the overall major Honshu-Shikoku

highway project to connect Honshu Island and Shioku Island by automobiles and trains.

There are three major bridge crossings; this one is the eastern-most crossing. The central

crossing was the first completed link, occurring two years ago.

vii
About 200,000 tons of steel will be used in the superstructure and about 1.42 million of

concrete will be used for the cable anchorage substructure. The anchorage superstructure on

the Honshu side was excavated using construction equipment to control tolerances within

about 15 cm (6 in). The two tower caissons are 85 m (279 ft) in diameter and 64 m (210 ft) deep. They were towed to the site and sunk by filling its outer lining with water and then

the center with 525,000 m^ of concrete.

The bridge is designed to survive an earthquake of 8.5 M and 10 minute average wind

velocity of 46 m/s. The main towers will rise 300 m (985 ft) above sea level. The bridge

cables are about 1.1m (3.6 ft) in diameter, consisting of 37,000 steel wires 5 mm in

diameter.

The cost of the bridge is $3 billion. Nineteen corporations are working on the

superstructure; five corporations are working on the towers.

Higashi-Kobe Bridge, Kobe, is a 3-span continuous cable-stayed bridge under construction

by the Hashin Expressway Public Corporation (HERP). The bridge is part of the Bayside

Route that connects Osaka and Kobe. Its center span is 485 m (1,591 ft) and end spans of

200 m (656 ft) each. Based on center span length, this 885 m (2,904 ft) bridge is the

world's longest cable-stayed bridge.

Its two-column rectangular towers are 145 m (476 ft) tall constructed of welded steel

sections. A square notch cutout was made at each tower comer to reduce wind-induced

vibration from vortex shedding which can damage the structure as demonstrated by wind

tunnel model testing. Wind tunnel model tests indicated a need to dampen the tower from

wind-induced vibration during construction. Mass-tuned oil dampers were installed at computed heights of the tower during construction and moved upward during its erection.

Twelve cable stays are attached to each tower in a harp pattern. The harp pattern cable

configuration was selected because it reduces the compressive vertical forces and bending

moments (about a horizontal axis perpendicular to the longitudinal axis of the bridge) on each

bridge tower column as compared to the fan pattern commonly used. The approximately 125

mm (5 in) diameter cables are covered with a 20 mm (0.8 in) thick polyethylene tube for

corrosion protection. These covers are designed with ribs parallel to the cable axis to reduce

the probability of wind and rain induced vibrations. Each of the 96 cables consists of 241 to 301 7 mm diameter wires. The towers are fabricated on a 35 m x 32 m x 26.5 m (115 x

105 X 87 ft) deep pneumatic concrete caisson.

The bridge has two decks each for three lanes of traffic. The lower bridge deck is 36.5 m

(120 ft) above the mean water level. . Its main girders consist of 9 m (30 ft) deep "Warren

Trusses." The corresponding dimensions of the boxed section diagonals are 1.0 x 0.7 m (3.3

X 2.3 ft). The vertical downward force component at the intersection of the truss diagonals on the top deck at the bridge is balanced by the tensile force from the cable connecting the

top deck of the bridge and the tower. The main girders and their tower supports at the tops

of the caisson are pin-supported in the longitudinal direction to lengthen the fundamental

period of the bridge to 4.4 sec. Mass tuned oil-dampers are installed at each end of the

bridge to reduce longitudinal displacements from earthquake ground motions.

viii
Construction started in 1989 with completion scheduled for 1993. Total cost for the project

is projected at 34 billion yen ($250 million).

Port Island, is a man-made island in Kobe Harbor constructed from fill material excavated

from mountains 25 km north of Kobe and transported to the harbor by conveyors passing

through a tunnel cut in the mountain. The 436 ha (1077 acres) island was constructed during

1966 through 1981. Various social and cultural facilities are being constructed on the center

of the island. The residential buildings on the island have a planned population of 16,000.

There are five container cargo berths on each side of the island. The island is currently

being extended to double its size. Also, an airport for Kobe is planned as a separate island

further into the bay as a new Port Island extension. The construction cost of Port Island is 558 billion yen ($4.1 billion). Kobe City and the Ministry of Transport have construction

responsibilities for the island.

Rokko Island, is adjacent to Port Island. It too, is man-made constructed from rocks

excavated from the nearby Rokko Mountains. The island has an area of 580 ha (1432 acres)

and container and liner cargo berths surround its perimeter with a residential area in the

center of the island. It has a planned population of about 30,000. Its construction began in

1972 and is expected to be completed in 1993. Kobe City and the Ministry of Transport are

responsible for its construction. The estimated construction cost is 1,240 billion yen ($9.1

billion).

Kansai International Airport, a man-made island located in the southeast part of Osaka Bay

about 4 km (2.5 mi) off the coast of Osaka, has been under construction since 1987.

Completion is expected in the summer of 1994. This site was selected to minimize noise

pollution from large numbers of aircraft required to serve this region. The existing Osaka

International Airport is seriously restricted in operation because of noise. It will be the first fully developed off-shore airport and the first around-the-clock airport ever constructed in

Japan.

The airport is a rectangular island 4.4 km x 1.25 km (2.7 x 0.8 mi) covering an area of 511

ha (1262 acres). Fill materials, imported by ship from mountains on the nearby mainland

and from island sites is crushed near the source, delivered to push barges on the sea coast by

belt conveyors and barged to the reclamation site. Poor strata under the island requires

designing for settlement of up to 11 m (36 ft). The island, located in water varying from 16-

19 m (52 to 62 ft) in depth, requires over 180 million m^ (235 million yd') of fill material.

To design against liquefaction from potential seismic events and to control future settlement,

sand drain piles each 40 cm (1.3 ft) in diameter spaced 2,5 m (8.2 ft) apart are driven into

the alluvial clay layer. Over 11 km (6.8 mi) of sea walls form the periphery of the island.

The access bridge connecting the island to the mainland is a single truss construction except

near the ends where spans are steel-box girders. Spans vary in length up to 150 m (492 ft) at the navigation lanes. The bridge has six lanes for autos and a two-rail system designed for

120 km/hr (75 mph) train*

Noel J. RAUFASTE, Secretary

U.S. -Side Panel on Wind and Seismic Effects

ix
ABSTRACT

This publication is the proceedings of the 23rd Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects. The meeting was held at the Public Works Research Institute,

Tsukuba, Japan, during May 14-17, 1991. The proceedings include the program, list of

members, panel resolutions, task committee reports, and 42 technical papers.

The papers were presented under seven themes: (I) - Wind Engineering, (II) - Earthquake

Engineering, (III) - Storm Surge and Tsunamis, (IV) - Joint Cooperative Research Program,

(V) - Performance of Nonstructural Systems, (VI) - International Decade for Natural Disaster

Reduction, and (VII) - Summaries of Task Committee Workshop Reports (oral presentations

only).

KEYWORDS: accelerograph; bridges; buildings; concrete; design criteria; disasters; disaster

reduction; earthquakes; geotechnical engineering; ground failures; lifelines; liquefaction;

masonry; repair and retrofit; risk assessment; seismic; soils; standards; storm surge;

structural engineering; tsunamis; and wind loads.

X
CONTENTS

PAGE

PREFACE iii ABSTRACT x AGENDA FOR 23im1 JOINT UJNR MEETING xvii

LIST OF PANEL MEMBERS xxv

LIST OF TASK COMMITTEE MEMBERS XLi RESOLUTIONS xLiii

THEME I: Wind Engineering

Probabilistic Performance Goals for Wind and Tornado Design 3

James R. Hill, James R. McDonald, and Robert C. Murray

Analysis of Tornado in Mobara City on December 11, 1990 15 H. Niino, T. Fujitani, T. Murota, N. Yamaguchi, and H. Okada

Hurricane Hugo in Perspective 29

Peter R. Sparks

THEME II: Earthquake Engineering

The Woridwide Earthquake Risk Management (WWERM) Program 41

S. T. Algermissen, Paul. R. Krumpe, and Walter W. Hays

Influence of Embankments on the Liquefaction of Groimd 47

Yasuyuki Koga, Jun-ichi Koseki, Takao Shimazu, and

Osamu Matsuo

Maximum Ground Acceleration on Soft Qay Sites 57

Tingley K. Lew

Broadband Strong-Motion Seismic Observation Network

Operated by Japan Meteorological Agency 75 Masahiro Yamamoto, Iwao Sasakawa, Akira Nagai, Takeshi Kakishita,

Akihiko Wakayama, Koichl Uhira, and Masaaki Seino ^

Seafloor Seismic Network Offshore Southern California. 87

Charles E. Smith

3d
CONTENTS

PAGE

Distribution of Liquefaction in the Saitama Plain by the 1923 Kanto Earthquake and Geomorphological Maps 99

Tadao Dohi and Takekazu Akagiri

Damage to Civil Engineering Facilities From the July 1990 Philippine Earthquake Ill Yasushi Sasaki and Michio Okahara

Recorded Structural Response During the M7.1

Loma Prieta Earthquake, 143 A. Gerald Brady

Monotonic and Cyclic Loading Tests of RockfiU Materials 149

Norihisa Matsumoto, Nario Yasuda, Masahiko Ohkubo,

and Ryo-ichi Yoshioka

Full-Scale Structural Applications of Active Control Systems 159

T. T. Soong, A. M. Reinhom, and M. Shinoziaka

Base-Isolated Buildings and Seismic Safety Evaluation

in Japan 169

Yoshikazu Kitagawa and Mitsumasa Midorikawa

Dynamic Modeling of Soil-Structure Systems for Control

Applications 179 H. A. Smith and R. I. Borja

Current Research Efforts in Japan for Passive and Active Control of Highway Bridges Against Earthquake 187

Kazuhiko Kawashima, Kinji Hasegawa, Shigeki Unjoh,

Hiroyukl Nagashima, and Hideyuki Shimizu

Status of U.S. Research on Structural Control Systems 211

S. C. Liu, H. J. Lagorio, and K. P. Chong

Vibration of Large-Scale Shaking Table Foundation 217

Chiikahiro Minowa, Nobuyuki Ogawa, Tadashi Mikoshiba,

Keiichi Ohtani, and Norio Oyagi

Safety and Economic Beneflts Realized From Upgrading

the Lawrence Berkeley Laboratory 225

Donald G. Eagling and James R. Hill

Structural Regulations of Buildings and Development of Technology in Japan 239

Yuji Ohashi and Hiroyuki Yamanouchi

xii
CONTENTS

PAGE

Seismic Hazard Assessment of Liquefaction Potential

at Mormon Island Auxiliary Dam, California, USA 247

David W. Sykora, Joseph P. Koester, and Mary E. Hynes

System IdentiHcation of Spirally Reinforced Circular Bridge

Columns Subjected to Cyclic Lateral Loading 269

William C. Stone and Andrew W. Taylor

Simplified Procedure for Inspecting Earthquake Resistant

Potential of Hi^iway Bridges Against Soil Liquefaction 295

Ken-ichi Tokida, Jiro Fukui, and Hideo Matsumoto

Evaluation and Repair of M^or Water Resources Structures for Improved Safety and Seismic Stability 309

William Roper

Shaking Table Tests for Hydrodynamic Pressure Acting on Double Cylindrical Caisson Type Breakwater in the Deep Sea 319

Tatsuo Uwabe, Makoto Osada, and Setsuo Noda

THEME nL Storm Surge and Tsunamis

Subaerial and Submarine Landslide Generated Tsunamis 335

James F. Lander, Patricia A. Lockridge, and Herbert Meyers

A Study on the Effect of Tsunami Breakwater

by Numerical Simulation 345

Chiaki Goto and Katsutoslu Tanimoto

THEME IV: Joint Cooperative Research Program

The U.S.-PRESSS Program 353

M. J. Nigel Priestley

U.S.-Japan Cooperative Research Program

Precast Seismic Structural System 357

Shin Okamoto, Shinsuke Nakata, Takashi Kaminosono,

Hitoshi Shiohara, and Mizuo Inukai

xiii
CONTENTS

PAGE

THEME V: Performance of Nonstructural Systems

Current Efforts to Improve the Performance of Nonstructural Systems Based on Experience From the Loma Prieta Earthquake 391

Robert Reitherman

State-of-the-Art Report on Seismic Design of Building Equipment

and Nonstructural Components in Japan 397

Masaya Hirosawa, Hatsukazu Mizuno, and Mitsumasa Midorikawa

THEME VI: International Decade for Natxu:al Disaster Reduction

Damages and Lessons From the Loma Prieta Earthquake* 415

Stephen A. Mahin

Damages and Lessons Learned From Hurricane Hugo* 435

Peter R. Sparks

United States Federal Agency Programs in Disaster

Countermeasures* 451

Paul F. Krumpe and Kenneth F. Sullivan

Manuscripts Authored for Panel Meeting but Not Presented Orally

Boundary Layer Wind Tunnel Study on Full Aeroelastic Long Span Bridge Model 467

Koichi Yokoyama, Tomoki Kanazaki, and Masahiko Yasuda

Comparison of Responses of a Select Number of Buildings

to the 10/17/1989 Loma Prieta (California) Earthquake

and Low-Level Amplitude Test Results 475

M. Celebi, L. T. Phan, and R. D. Marshall

Evaluation of Penetration Resistance for Liquifaction Assessment at Jackson Lake Dam 501

Jeffrey A. Farrar and Nario Yasuda

Torsional Resistance of Multi-Column Foundation 527

Yukitake Shioi, Michio Okahara, Jiro Fukui,

and Syoji Takagi

* U.S. papers presented at the IDNDR Seminar in Gifu City, Japan on May 20, 1991.

xiv
CONTENTS

PAGE

Nondestnictiye Evaluation of Civil Structures and Materials 541

Ken P. Chong and John B. Scalzi

NSF Supported Research in Concrete Structures and Materials 549

Ken P. Chong, Shih-Chi Liu, and Henry J. Lagorio

Development of Water Level and Wave Height

Design Data 557

Martin C. Miller, William E. Roper, Leon E. Borgman,

and Joannes J. Westerink

U.S. Coordinated Program for Masonry Building Research: Sixth Year Status 569

James L. Noland

Appendix: Task Committees A-J Reports 579

XV

AGENDA FOR 23rd JOINT UJNR MEETING

Tuesday. May 14. 1991 - 10:30-12:00

Conference Hall (8th Floor) , Public Works Research Institute

Call to order by Dr. Takashi lijima, Secretary-General, Japan

Panel, Director of Planning and Research Administration Department,

Public Works Research Institute

Remarks by Dr. Hiroshi Mitani, Vice-Minister of Construction for

Engineering-Affairs, Ministry of Construction

Remarks by Dr. Edward M. Malloy, Counsellor for Scientific and

Technological Affairs, Embassy of the United States of America

Remarks by Dr. Tetsuhisa Shirakawa, Director, International Affairs

Division, Science and Technology Promotion Bureau, Science and

Technology Agency

Remarks by Dr. Toshio Iwasaki, Chairman, Japan Panel, DirectorGeneral, Public Works Research Institute, Ministry of Construction

Remarks by Dr. H.S. Lew, Chairman (Acting Chairman), U.S. Panel,

Chief Structures Division, Building and Fire Research Laboratory,

National Institute of Standards and Technology, Department of

Commerce

Introduction of Japan Members by Japan Panel Chairman and U.S.

Members by U.S. Panel Chairman

Election of Joint Meeting Chairman

Adoption of Agenda

Closure by Dr. Takashi lijima

Adjourn

Commemorative Planting

Group Photograph

Lunch - 12:00-13:00 (Hosted by the Geographical Survey Institute)

Technical Session - 13:00-14:20 (International Conference Room 8F)

* Speaker

Session 1 - Theme I - WIND ENGINEERING

Chairman: T. Iwasaki

xvii
13:00-13:20 Probabilistic Performance Goals for Wind and

Tornado Design

James R. Hill*, James R. McDonald, Robert C. Murray

13:20-13:40 Analysis of Tornado in Mobara City on December 11,

1990

H. Niino*, T. Fujitani, T. Murota, N. Yamaguchi, H.

Okada

13:40-14:00 Hurricane Hugo in Perspective

Peter R. Sparks*

14:00-14:20 Discussion

14:20-14:30 Break

Task Committee Meetings - 14:30-17:00

C—Repair and Retrofit of Existing Structures (Meeting Room 531)

D—Evaluation of Performance of Structures (Meeting Room 531)

E —Ground Motion and Seismic Design Forces (Meeting Room 532)

H—Soil Behavior and Stability During Earthquakes (Meeting Room

627)

J--Wind and Earthquake Engineering for Transportation Systems

(Meeting Room 628)

Preliminary Meeting

K—Wind and Earthquake Engineering for Offshore and Coastal

Structures (Meeting Room 629)

Adjourn

Reception (Hosted by Vice-Minister for Engineering-Affairs of the

Ministry of Construction)

Wednesday y May 15

Technical Session - 9:00-12:00

Session 2 - Theme II - EARTHQUAKE ENGINEERING

Chairman: H.S. Lew

9:00- 9:20 The Worldwide Earthquake Risk Management (WWERM)

Program

S.T. Algermissen*, Paul R. Krumpe, Walter W. Hays

9:20- 9:40 Influence of Embankments on the Liquefaction of

Ground

Y. Koga*, J. Koseki, T. Shimazu, O. Matsuo

9:40-10:00 Maximum Ground Acceleration on Soft Clay Sites

Tingley K. Lew*

xviii
10:00-10:20 Discussion

10:20-10:40 Break

Session 3 - Theme II - EARTHQUAKE ENGINEERING

Chairman: H.S. Lew

10:40-11:00 Broadband Strong-Motion Seismic Observation Network

Operated by Japan Meteorological Agency

M. Yamamoto*, I. Sasakawa, A. Nagai, T. Kakishita,

A. Wakayama, K. Uhira, M. Seino

11:00-11:20 Seafloor Seismic Network Offshore Southern

California

Charles E. Smith*

11:20-11:40 Distribution of Liquefaction in the Saitama Plain

. . by the 1923 Kanto Earthquake and Geomorphological

Maps

T. Dohi, T. Akagiri

11:40-12:00 Discussion

Lunch - 12:00-13:00 (Hosted by the Port and Harbour Research

Institute and the Meteorological Research Institute)

Technical Session - 13:00-15:20

Session 4 - Theme II - EARTHQUAKE ENGINEERING

Chairman: T. Iwasaki

13:00-13:20 Damage to Civil Engineering Facilities from the

July 1990 Philippine Earthquake

Y. Sasaki*, M. Okahara

13:20-13:40 Recorded Structural Response During the M7 . 1 Loma

Prieta Earthquake Comments

A. Gerald Brady*

13:40-14:00 Monotonic and Cyclic Loading Tests of Rockfill

Materials

N. Matsumoto, N. Yasuda*, M. Ohkubo, R. Yoshioka

14:00-14:20 Discussion

Session 5 - Reports of Workshops

Chairman: T. Iwasaki

14:20-14:30 (H) Workshop on Remedial Treatment of Potentially

Liquefiable Soils

K. Tokida* - , 14:30-14:40 (I) 2nd Tsunami Workshop

J. Lander* - 14:40-14:50 (J) 1st Workshop on Seismic Retrofit of Bridges

K. Kawashima*
14:50-15:00 (J) 7th Bridge Workshop - M. Fujiwara*

15:00-15:20 Discussion ; 15:20-15:30 Break

Task Committee Meetings - 15:3 0-17:00

A—Strong-Motion Instrumentation Arrays and Data (Meeting Room 531)

B—Large-scale Testing Program (Meeting Room 532)

G—Passive, Active, and Hybrid Control Systems (Meeting Room 627)

F—Disaster Prevention Methods for Lifeline Systems (Meeting Room

628)

I —Storm Surge and Tsunamis (Meeting Room 629)

Adjourn ^

'

Thursday. May 16

Technical Session - 9:00-12:00

Session 6 - Theme II - EARTHQUAKE ENGINEERING

Chairman: H.S. Lew

9:00- 9:20 Full-Scale Structural Applications of Active

Control Systems

T.T. Soong, A.M. Reinhorn, M. Shinozuka*

9:20- 9:40 Base-Isolated Buildings and Seismic Safety

Evaluation in Japan

Y. Kitagawa*, M. Midorikawa

9:40-10:00 Dynamic Modeling of Soil-Structure Systems for

Control Applications

H.A. Smith*, R.I. Borja

10:00-10:20 Discussion

10:20-104 0 Break

Session 7 - Theme II - EARTHQUAKE ENGINEERING

Chairman: H.S. Lew

10:40-11:00 Current Research Efforts in Japan for Passive and

Active Control of Highway Bridges Against

Earthquake

K. Kawashima*, K. Hasegawa, S. Unjoh, H. Nagashima,

H. Shimizu

XX
11:00-11:20 Status of U.S. Research on Structural Control

Systems

S.C. Liu*, H.J. Lagorio, K.P. Chong

11:20-11:40 Vibration of Large-scale Shaking Table Foundation

C. Minowa*, N. Ogawa, T. Mikoshiba, K. Ohtani, N.

Oyagi

11:40-12:00 Discussion

Lunch - 12:00-13:00 (Hosted by Building Research Institute)

Technical Session - 13:00-17:00

Session 8 - Theme III - STORM SURGE AND TSUNAMIS

Chairman: T. Iwasaki

13:00-13:20 Subaerial and Submarine Landslide Generated

Tsunamis

James F. Lander*, Patricia A. Lockridge, Herbert

Meyers

13:20-13:40 A Study on the Effect of Tsunami Breakwater by

Numerical Simulation

C. Goto*, K. Tanimoto

13:40-13:55 Discussion

Session 9 - Theme IV - U.S. -JAPAN JOINT COOPERATIVE RESEARCH

PROGRAM

Chairman: T. Iwasaki

13:55-14:15 The U.S.-PRESSS Program

M.J. Nigel Priestley (H.S. Lew*)

14:15-14:35 U.S. -Japan Cooperative Research Program Precast

Seismic Structural System

, S. Okamoto, S. Nakata*, T. Kaminosono, H. Shiohara,

M. Inukai

14:35-14:50 Discussion ' ; 14:50-15:10 Break

Session 10 - Theme II - EARTHQUAKE ENGINEERING

Chairman: T. Iwasaki

15:10-15:30 Safety and Economic Benefits Realized from

Upgrading the Lawrence Berkeley Laboratory

Donald Eagling*, James R. Hill

15:30-15:50 Structural Regulations of Buildings and Development

of Technology in Japan

Y. Ohashi*, H. Yamanouchi

xxi
15:50-16:10 Seismic Hazard Assessment of Liquefaction Potential

at Mormon Island Auxiliary Dam, CA, U.S.A.

David W, Sykora, Joseph P. Koester, Mary E. Hynes

(Gus Franklin*)

16:10-16:30 System Identification of Spirally Reinforced

Circular Bridge Columns Subjected to Cyclic Lateral

Loading

W.C. Stone, A.W. Taylor (H.S. Lew*)

16:30-17:00 Discussion

Adjourn

Dinner (Hosted by Director General of Public Works Research

Institute)

Friday. May 17

Technical Session - 9:00-12:00

Session 11 - Theme II - EARTHQUAKE ENGINEERING

9:00- 9:20 Simplified Procedure for Inspecting Earthquake

Resistant Potential of Highway Bridge Against Soil

Liquefaction

K. Tokida*, J. Fukui, H. Matsumoto

9:20- 9:40 Evaluation and Repair of Major Water Resources

Structures for Improved Safety and Seismic

Stability

William Roper*

9:40-10:00 Shaking Table Tests for Hydrodynamic Pressure

Acting on Double Cylindrical Caisson Type

Breakwater in the Deep Sea

T. Uwabe*, M. Osada, S. Noda

10:00-10:20 Discussion

10:20-10:40 Break

Session 12 - Theme V - PERFORMANCE OP NONSTRUCTURAL SYSTEMS

Chairman: H.S. Lew

10:40-11:00 Current Efforts to Improve the Performance of

Nonstructural Systems Based on Experience from the

Loma Prieta Earthquake

Robert Reitherman (Stephen Mahin*)

11:00-11:20 State-of-the-Art Report on Seismic Design of

Building Equipment and Nonstructural Components in

Japan

M. Hirosawa*, H. Mizuno, M. Midorikawa

11:20-11:40 Discussion

xxii
Lunch - 11:40 - 12:40 (Hosted by the National Research Institute

for Earth Science and Disaster

Prevention)

Reports on Activities Related to IDNDR - 12:40-13:00

Chairman: T. Iwasaki

Reports of Task Committees - 13:00-15:00

A—Strong-Motion Instrumentation Arrays and Data

B—Large-scale Testing Programs

C—Repair and Retrofit of Existing Structures

D—Evaluation of Performance of Structures

E—Ground Motion and Seismic Design Forces

F—Disaster Prevention Methods for Lifeline Systems

G—Passive, Active, and Hybrid Control Systems

H—Soil Behavior and Stability During Earthquakes

I —Storm Surge and Tsunamis

J—Wind and Earthquake Engineering for Transportation Systems

15:00-15:20 Break

Adoption of Final Resolutions 15:20-16:10

Chairman: T. Iwasaki

16:10-16:20 Break

Closing Session 16:20-16:40

Call to Order by Dr. Takashi lijima, Secretary-General, Japan

Panel, Director, Planning and Research Administration Department,

Public Works Research Institute

Closing Remarks by Dr. H.S. Lew, Chairman (Acting Chairman) , U.S.

Panel, Chief, Structures Division, Building and Fire Research

Laboratory, National Institute of Standards and Technology,

Department of Commerce

Closing Remarks by Dr. Toshio Iwasaki, Chairman, Japan Panel,

Director-General, Public Works Research Institute

Closure by Dr. Takashi lijima

Press Conference 16:50-17:20 (Conference Hall - 8F)

xxiii

LIST OF PANEL MEMBERS

UNITED STATES-SIDE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1991

Dr. Richard N. Wright

Chairman

Director, Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 2 0899

301.975.5900 FTS 879.5900 FAX 301.975.4032

Mr. Noel J. Raufaste

Secretary-General

Head, Cooperative Research Programs

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 2 0899

301.975.5904 FTS 879.5904 FAX 301.975.4032

Dr. S. T. Algermissen

Research Geophysicist

Hazard and Risk Assessment Group

MS 966, POB 25046

Denver Federal Center

Geological Survey

U.S. Department of the Interior

Denver, CO 80225

303.236.1611 FTS 776.1611 FAX 303.236.0618

Dr. Celso S. Barrientos

Supeirvisory Physical Scientist

INESDIS-E/RA28

Suitland Professional Center, Room 315

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

5001 Silver Hill Road

Suitland, MD 20746

301.763.4178 FAX 301.420.1883

Dr. Eddie N. Bernard

Director, Pacific Marine Environmental Laboratory

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

7600 Sand Point Way, NE - BIN C15700/Building 3

Seattle, WA 98115-0070

206.526.6239 FTS 392.6800 FAX 206.526.6815

XXV ,
Dr. Roger D. Borcherdt

Branch of Engineering Seismology and Geology

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94 02 5

415.329.5619 FTS 459.5619 FAX 415.329.5163

Dr. A. Gerald Brady

Research Civil Engineer

Branch of Engineering Seismology and Geology

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5664 FTS 459.5664 FAX 415.329.5163

Mr. H. Lee Butler

Chief, Research Division

Coastal Engineering Research Center

U.S. Army Engineer Waterways Experiment Station

Office CEWES-CR

P.O. Box 631

Vicksburg, MS 39180-0631

601.634.2405 FTS 542.2405 FAX 601.634.4314

Dr. Ken P. Chong

Program Director, Structures and Building Systems

Directorate of Engineering

National Science Foundation

1800 G Street, N.W. , Room 1108

Washington, DC 20550

202.357.9542 FAX 202.357.7636

Mr. James D. Cooper

Deputy Chief, Structures Division, HNR-10

Federal Highway Administration

U.S. Department of Transportation

63 00 Georgetown Pike

McLean, VA 22101

703.285.2447 FAX 703.285.2379

Mr. Robert D. Dikkers

Assistant Chief, Structures Division

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301.975.5863 FTS 879.5863 FAX 301.975.4032

xxvi
Dr. A. G. Franklin

Chief, Earthquake Engineering & Geophysics Division

CEWES-GH Geotechnical Laboratory

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39180-3909

601.634.2658 FTS 542.2658 FAX 601.634.3139

Mr. G. Robert Fuller

Acting Director, Manufactured Housing and Construction

Standards Division

Room 9152 HSMM

U.S. Department of Housing and Urban Development

Washington, DC 20410-8000

202.708.2210 FAX 202.708.0299

Mr. James H. Gates

Caltrans - Structures

P. O. Box 942874

Sacramento, CA 94274

916.445.1439 FAX 916.445.0574

Mr. Peter E. Gurvin

Director, Building Design and Engineering Division

Foreign Building Operations

Building SA-6, Room 335

U.S. Department of State

Washington, DC 2 052 0

703.875.6117 FAX 703.875.6204

Dr. Walter W. Hays

Deputy for Research Applications

Office of Earthquake, Volcanoes and Engineering

905 National Center

Geological Survey

U.S. Department of the Interior

Reston, VA 2 2 092

703.648.6711 FTS 959.6717 FAX 703.648.6717

Mr. James R. Hill

Manager, Natural Hazards Mitigation Programs

Office of Safety, Health and Quality Assurance

U.S. Department of Energy

Washington, DC 20585

301.353.4508 FAX 301.353.5285

Mr. Larry C. Hultengren

Senior Structural Engineer

Office of Civil/Structural Engineering

Foreign Building Operations

U.S. Department of State

Washington, DC 20520

703.875.6194 FAX 703.875.6204

xxvii
Dr. William B. Joyner

Geophysicist

Branch of Ground Motion & Faulting

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5640 FTS 459.5640 FAX 415.329.5163

Mr. Roger M. Kenneally

Structural Engineer

Structural and Seismic Engineering Branch

Seismic Section

Office of Nuclear Regulatory Research

U.S. Nuclear Regulatory Commission

Washington, DC 20555

301.492.3893 FAX 301.443.7836

Mr. Richard W. Kramer

Technical Review Staff, D-3210

Bureau of Reclamation /

U. S. Department of the Interior

P.O. Box 25007

Denver, CO 80225

303.236.8539 FTS 776.8539 FAX 202.647.5269

Mr. Paul F. Krumpe

Physical Scientist

Office of U.S. Foreign Disaster Assistance

Agency for International Development

U.S. Department of State

Washington, DC 20523

202.647.9758 FAX 202.647.5269

Dr. Henry J. Lagorio

Program Director, Architectural and Mechanical Systems

Division of Biological and Critical Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550 ; 202.357.9780 FAX 202.357.9803

Dr. H.S. Lew

Chief, Structures Division

Building and Fire Research Laboratory

National Institute of Standards and Technology

U.S. Department of Commerce

Gaithersburg, MD 20899

301.975.6061 FTS 879.6061 FAX 301.975.4032

xxviii
Mr. Tingley K. Lew

Research Structural Engineer

Structures Division, Code L51, Bldg. 560, Rm. 278B

Naval Civil Engineering Laboratory

Port Hueneme, CA 93 043

805.982.1234 FAX 805.982.1418

Dr. Shih-Chi Liu

Program Director, Structural Systems

Division of Biological and Critical Systems

National Science Foundation

1800 G Street, N.W. , Room 1132

Washington, DC 20550

202.357.9780 FAX 202.357.7636

Mr. Robert B. MacDonald

Manager, Geology Branch

Engineering and Research, Code D-3 610 A

Bureau of Reclamation

U.S. Department of the Interior

Building 67, Denver Federal Center

Denver, CO 80225

303.236.6904 FTS 776.8643 FAX 303.236.6763

Dr. Wayne N. Marchant

Chief, Research and Laboratory Services Division

P. O. Box 25007, Denver Federal Center

Code D-3700

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303.236.5983 FTS 776.5983 FAX 303.236.7664

Dr. Francis G. McLean

Chief, Geotechnical Engineering and Embankment

Dams Branch

P.O. Box 25007

Bureau of Reclamation

U.S. Department of the Interior

Denver, CO 80225

303.236.3854 FTS 776.3854 FAX 303.236.6763

Mr. Herbert Meyers

Chief, Earth Geophysics

National Geophysical Data Center

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

325 Broadway

Boulder, CO 80303-3328

303.497.6521 FTS 320.6521 FAX 303.497.6513

xxix
Mr. Howard D. Nickerson

Earthquake Engineering and Weapons Specialist

Naval Facilities Engineering Command

Hoffman Building #2, Room 12S63, Code 04B2

200 Stovall Street

Alexandria, VA 22 3 32

703.325.0044 FAX 703.325.1916

Dr. William E. Roper

Assistant Director

Research and Development (Civil Works)

U.S. Army Corps of Engineers

20 Massachusetts Avenue, N.W

Washington, DC 20314-1000

202.272.0257 FAX 202.272.0907

Dr. Erwin L. Schaffer

Assistant Director, Wood Products Research

Forest Products Laboratory

Forest Service

U.S. Department of Agriculture

One Gifford Pinchot Drive

Madison, WI 53705-2398

608.264.5672 FTS 364.5672 FAX 608.231.9592

Mr. Charles E. Smith

Research Program Manager

Offshore Minerals Management

Technology Assessment and Research Branch

Minerals Management Service

U.S. Department of the Interior

381 Elden Street, MS 4800

Herndon, VA 22070-4817

703.787.1559 FAX 703.787.1010

Dr. T. T. Soong

Samuel P. Capen Professor

Department of Civil Engineering

National Center for Earthquake Engineering Research

State University of New York at Buffalo

212 Ketter Hall

Buffalo, NY 14260

716.636.2469 FAX 716.636.3733

Mr. Kenneth F. Sullivan

Lifelines Project Manager

Earthquakes and Natural Hazards Division

Federal Emergency Management Agency

Washington, DC 20472

202.646.2889 FAX 202.646.3104

XXX
Mr. Edward F. Younger

Director, Structural Engineering Service

Code 088C1

Office of Facilities

Department of Veterans Affairs

810 Vermont Avenue, N.W.

Washington, DC 20420

202.233.2864 FTS 373.2864 FAX 202.233.7854

xxxi
ALTERNATE MEMBERS

Dr. Clifford J, Astill

Program Director

Division of Fundamental Research for Critical

Engineering Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9500 FAX 202.357.9803

Dr. Mehmet K. Celebi

Research Civil Engineer

Branch of Engineering Seismology and Geology

Geological Survey

U.S. Department of the Interior

345 Middlefield Road, MS 977

Menlo Park, CA 94025

415.329.5623 FTS 459.5623 FAX 415.329.5163

Mr. Michael Changery

Chief, Global Analysis Branch

National Climatic Data Center

National Oceanic and Atmospheric Administration

U. S. Department of Commerce

Federal Building

Ashville, NC 28801

704.259.0765 FTS 672.0765 FAX 704.259.0246

Mr . C . Y . Chen

Senior Civil/Geotechnical Engineer

Foreign Building Operations

A/FBO/PE/CSB, SA-6, Rm. 327

Washington, DC 20520

703.875.6207 FAX 703.875.6204

Mr. Vincent P. Chiarito

Research Structural Engineer

Structural Mechanics Division

Structures Laboratory

U.S. Army Engineer Waterways Experiment Station

3909 Halls Ferry Road

Vicksburg, MS 39108-6199

601.634.2714 FAX 601.634.2309

Dr. James F. Costello

Senior Structural Engineer

Structural and Seismic Engineering Branch

Structural Section

Office of Nuclear Regulatory Research

U.S. Nuclear Regulatory Commission

Mail Stop 007NL

Washington, DC 20555

301.492.3818 FAX 301.492.3696

xxxii
Mr. Lucian G. Guthrie

Structural Engineer

Office of Chief of Engineers

HQUSACE (CEEC-ED)

U.S. Department of the Army

2 0 Massachusetts Avenue, N.W.

Washington, DC 20314-1000

202.272.8673 FAX 202.272.1485

Dr. James R. Houston

Chief, Coastal Engineering Research Center

US Army Engineer Waterways Experiment Station

P.O. Box 631

Vicksburg, MS 39180-0631

601.634.2000 FAX 601.634.2655

Mr. James Lander

Geophysicist

Cooperative Institute for Research in

Environmental Sciences

University of Colorado

Campus Box 449, Room 152 RL3

3100 Marine Street

Boulder, CO 80309

303.497.6446 FTS 320.6446 FAX 303-497.6513

Mr. Robert R. Ledzian

Senior Staff Assistant for Research

Office of Liaison - Engineering Research, Code 3020

Bureau of Reclamation

U.S. Department of the Interior

18th & C Streets, N.W.

Washington, DC 20240

202.208.3432 FAX 202.208.6252

Mr. Ronald J. Morony

Director, Innovative Technology and Special

Projects Division

Office of Policy Development and Research

U.S. Department of Housing and Urban Development

Washington, DC 20410-6000

202.755.0640 FAX 202.755.0299

Dr. J. Eleonora Sabadell

Director, Natural and Manmade Hazards Mitigation Program

Directorate of Engineering

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9780 FAX 202.357.9803

xxxiii
Dr. John B. Scalzi

Program Director, Structures and Building Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

202.357.9542 FAX 202.357.7636

xxxiv
JAPAN-SIDE PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1991

Dr. Toshio Iwasaki

Chairman

Director-General

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 3 05

Tel. 0298-64-2211

Dr. Takashi lijima

Secretary-General

Director, Planning and Research Administration

Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tadao Dohi

Head, Geographic Department

Geographical Survey Institute

Ministry of Construction

1, Kitazato, Tsukuba-shi

Ibaraki-ken 3 05

Tel. 0298-64-1111

Mr. Tokunosuke Fujitani

Chief, The Second Research Laboratory

Applied Meteorology Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1 Nagamine, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-7111

Mr. Kazuhiro Nishikawa

Head, Bridge Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction . 1, Asahi, Tsukuba-shi,

Ibaraki-ken 3 05

Tel. 0298-64-2211

XXXV
Dr. Masaya Hirosawa

Deputy Director General

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Kazuhiko Kawashima * Head, Earthquake Engineering Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Yoshikazu Kitagawa

Director

Testing and Evaluation Department

Building Research Institute

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Mr. Yasuyuki Koga

Head, Soil Dynamics Division

Construction Method and Equipment Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tadahiko Fujisawa

Head, Fill Type Dam Division

Dam Department

Public Works Research Institute

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tatsuo Murota

Director, Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

xxxvi
Dr. Shinsuke Nakata

Head, Structural Division

International Institute of Seismology and Earthquake

Engineering

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 3 05

Tel. 0298-64-2151

Dr. Setsuo Noda

Director, Structural Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka,

Kanagawa-ken 2 39

Tel. 0468-41-5410

Mr. Keiichi Ohtani

Head, Earthquake Disaster Prevention Laboratory

National Research Institute for Earth Science and

Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tsukuba-shi,

Ibaraki-ken 3 05

Tel. 0298-51-1611

Mr. Shin Ohtsuka

Head, Typhoon Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi

Ibaraki-ken 3 05

Tel. 0298-51-7111

Mr. Michio Okahara

Head, Foundation Engineering Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Shin Okamoto

Director, International Institute of Seismology and

Earthquake Engineering

Building Research Institute

Ministry of Construction

1, Tatehata, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

xxxvii
Dr. Norio Oyagi

Director, Third Research Division

National Research Institute for Earth Science

and Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-1611

Dr. Yasushi Sasaki

Director, Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Masaaki Seine

Head, Seismology and Volcanology Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-51-7111

Mr. Minoru Fujiwara

Director, Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Katsutoshi Tanimoto

Director, Hydraulic Engineering Division

Port and Harbour Research Institute

Ministry of Transport - 3-1-1, Nagase, Yokosuka,

Kanagawa-ken 239

Tel. 0468-41-5410

Mr. Ken-ichi Tokida

Head, Ground Vibration Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

xxxviii
Dr. Takaaki Uda

Head, Coastal Engineering Division

River Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi, , Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Tatsuo Uwabe

Chief, Disaster Prevention Laboratory

Structural Engineering Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi

Kanagawa-ken 239

Tel. 0468-44-5030

Dr. Hiroyuki Yamanouchi

Head, Structural Dynamics Division

Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Yutaka Yamazaki

Associate Director for Composite Structures Research

Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 3 05

Tel. 0298-64-2151

Mr. Koichi Yokoyama * Head, Structure Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

* Secretary

XXXIX

LIST OF TASK COMMITTEE MEMBERS

September 1991

LIST OF TASK COMMITTEE MEMBERS

1991

Task Committee US Side Japanese Side

Strong-Motion Instrumentation A.G. Brady* S. Noda*

Arrays and Data • jyi . K

.

ceieJDi S . Hattori

A.F. Franklin M. Ichikawa

W.B. Joyner Y. Ishiyama

R.B. MacDonald K. Ohtani

F.G. McLean Y. Sasaki

H. Meyers H. Yamanouchi

ija.rge oCa.xe lesumg Jrrograins rl . o . Lew* J\

.

un uanx ^ C.E. Smith Y. Koga

S. Noda

S. Okamoto

K. Yokoyama

H. Yamanouchi

Repair and Retrofit of K.P. Chong* M. Hirosawa*

Existing Structures G.E. Freeland T. Iwasaki

J .Ptcic^y "i o q

H.S. Lew H. Shinohara

R.J. Morony

D.H. Oh

J.B. Scalzi

Evaluation of Structural G.R. Fuller* S. Okamoto*

Performance S.A. Asar T. Hanafusa

M.K. Celebi K. Masamura

J.B. Scalzi T. Murota

M. Okahara

H. Shinohara

H. Yamanouchi

K. Yokoyama

Ground Motion and Seismic S.T. Algermissen* K. Kawashima*

Design Forces Programs C.S. Barrientos S. Hattori

A.G. Brady T. Iwasaki

G.R. Fuller M. Nagaoka

P.F. Krumpe Y. Sugimura

T. Uwabe

XLI
Task Committee-continued US Side Japanese Side

F. Disaster Prevention Methods R.D. Dikkers* Y. Sasaki*

for Lifeline Systems M.K. Celebi K. Kawashima

T n rp nuro ua

P.F. Krumpe s. Noda

T. Liu K. Ohtani

J.B. Scalzi M. Shinozuka

J.S. Spencer T. Uwabe

K.F. Sullivan

S. Wu

G. Passive, Active, and Hybrid S.C. Liu* Y. Kitagawa*

Control Systems C. Biglow

L • C • Hultengren

A. Lin

T.T. Soong

K.F. Sullivan

H

»

Soil Behavior and Stability A.G. Franklin* K. Tokida*

During Earthquakes C.Y. Chen Y. Koga

F.G. McLean N. Matsumoto

C.E. Smith M. Okahara

Y. Sugimura

H. Tsuchida

I. Storm Surge and Tsunamis H. Meyers* T. Uda*

C.J. Astill M. Ichikawa

C.S. Barrientos K. Ku Ta sh 1 cf ^ E.N. Bernard K. Masamura

L. Butler M. Okada

C.F. Jelesnianski N. Oyagi

P.F. Krumpe H. Takahashi

J. Lander K. Tanimoto

W.E. Roper

J. Wind and Earthquake J.D. Cooper* S. Saeki*

Engineering for A.G. Franklin M. Fuj iwara

Transportation Systems J.H. Gates T. Iwasaki

J.B. Scalzi K. Kawashima

M. Okahara

Y. Sasaki

K. Yokoyama

I

* Chairman I

XLII

I
RESOLUTIONS

RESOLUTIONS OF THE TWENTY-THIRD JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS (UJNR)

Public Works Research Institute — '^ Ministry of Construction

1, Asahi , Tsukuba-shi, Ibaraki 305

May 14-17, 1991

The following resolutions are hereby adopted:

1. The Twenty-third Joint Panel Meeting provided an opportunity

to exchange valuable technical information which was beneficial to both countries. In view of the importance of

cooperative programs on the subject of wind and seismic

effects, the continuation of Joint Panel Meetings is

considered essential.

2. The following activities have been conducted since the

Twenty-second Joint Meeting:

a. Guest researchers from both countries performed joint

research that advanced the state of earthquake

engineering.

b. Technical documents, research reports, and proceedings of

workshops were exchanged.

c. The Japanese MOC documents on base isolation systems for

buildings were translated into English by the U.S. -side

Panel. They were reviewed by both sides and approved for

publication and dissemination by NIST and NTIS.

d. Five workshops were held:

a) 2nd Tsunami Workshop, Task Committee (I), at

Honolulu, Hawaii, Nov. 5-7, 1990.

' b) 1st Meeting of Joint Technical Coordinating

Committee on Precast Seismic Structural Systems,

Task Committee (B), at San Diego, CA, Nov. 29-Dec.

1, 1990.

c) 1st Workshop on Seismic Retrofit of Bridges, Task

Committee (J), at Tsukuba, Dec. 17-22, 1990.

d) Workshop on Remedial Treatment of Potentially

Liquefiable Soils, Task Committee (H) , at Tsukuba,

Jan. 22-25, 1991.

e) 7th Bridge Workshop, Task Committee (J), at Tsukuba,

May 8-11, 1991.

XLIII
3. A two day seminar where Panel members participated was held in

Charleston, South Carolina during the Panel's study tours on May 21 - 22, 1990 on the subject of lessons learned from

Hurricane Hugo and repair and retrofit of historic buildings.

The seminar was hosted by the Charleston Historic Foundation

and The Citadel.

4. The Panel will continue to seek methods to contribute to the

International Decade of Natural Disaster Reduction (IDNDR)

such as exchanging Proceedings of joint Panel Meetings and

Task Committee Workshops with their respective Country's

National Committees of IDNDR. The Panel compiled a bibliography of papers presented at the past 23 Joint Panel

meetings and T/C workshops, and will disseminate to various

countries suffering from strong winds and earthquakes.

5. The Panel recognizes the importance of research on improving

the design and construction of offshore and coastal

facilities. The Panel approves the establishment of a new

Task Committee (T/C "K" ) on Wind and Earthquake Engineering

for Offshore and Coastal Facilities. Its Mission Statement

will be developed and discussed at the respective country's

domestic meetings.

6. The Panel recognizes the importance of the work by both sides

in the US-Japan Joint Research Program on Precast Seismic

Structural Systems (PRESSS).

7. The Panel accepts each Task Committee's report developed

during the Twenty-third Joint Meeting. Each report presents

objectives, scope of work, accomplishments, future plans,

information exchange, impacts and barriers.

8. The Panel endorses the following proposed Task Committee

workshops during the coming year.

a. Task Committee (A) plans a workshop on Strong Motion

Record Processing for Long Periods during the spring of

1992.

b. Task Committee (B) proposes holding a workshop to develop

a joint research program on Composite, Mixed, and Hybrid

Structures during the spring of 1992.

c. Task Committee (B) will conduct its 2nd meeting of the

Joint Technical Coordinating Committee on Precast Seismic

Structural Systems (PRESSS) in Japan, Oct 30-Nov. 1,

1991.

d. The Committee (E) has scheduled a workshop on Hazard/Risk

Assessment and Design Earthquake Loading in Tsukuba,

Japan during 28-29 May 1991.

XLIV
e. Task Committee (F) has scheduled the Fourth Workshop on Earthquake Disaster Prevention for Lifeline Systems

during August 19-21, 1991 in Los Angeles.

f. Task Committee (G) has scheduled a workshop on the Design

of Bridge Base Isolation at the National Center for

Earthquake Engineering Research during September 1-5,

1991.

g. Task Committee (J) will hold its Eighth Bridge Workshop,

scheduled in the U.S. just prior to the 24th UJNR Joint

Meeting.

Scheduling for the workshops and planning meeting shall be

performed by the U.S. and Japan chairmen of the respective

Task Committee with concurrence of the Joint Panel chairmen.

Results of each activity shall be presented at the next Joint

Panel Meeting.

The Panel recognizes the importance of continued exchange of

personnel, technical information, research results, and

recorded data that lead to mitigating losses from strong winds

and earthquakes. The Panel also recognizes the importance of

using available large-scale testing facilities in both

countries. Thus, these activities should continue to be

strengthened and expanded and, as appropriate, share Task

Committee activities at other meetings that have technical

interests in the Task Committee activities. To facilitate

these exchanges, the Panel will provide official endorsement.

The Twenty-Fourth Joint Meeting of the UJNR Panel on Wind and

Seismic Effects will be held at NIST, Gaithersburg MD, U.S.A.,

May 1992. Specific dates, program, and itinerary will be

proposed by the U.S. Panel with concurrence of the Japan

Panel.

XLV

Theme I

Wind Engineering

i

ii

I

Probabilistic Performance Goals for Wind and Tornado Design

by

James R. Hill*

James R. McDonald**

Robert C. Murray***

ABSTRACT

Department of Energy, (DOE), facilities across the United States, use wind and

tornado design and evaluation criteria based on probabilistic performance goals. In

addition, other programs such as Advanced

Light Water Reactors, New Production

Reactors, and Individual Plant Examinations

for External Events for commercial nuclear

power plants utilize design and evaluation

criteria based on probabilistic performance

goals. The use of probabilistic performance

goals is a departure from design practice for

commercial nuclear power plants which have

traditionally been designed utilizing a deterministic specification of wind and

tornado loading combined with deterministic

response evaluation methods and permissible

behavior limits. Approaches which utilize

probabilistic wind and tornado hazard curves

for specification of loading and deterministic

response evaluation methods and permissible

behavior limits are discussed in this paper.

Through the use of such design/evaluation

approaches, it may be demonstrated that there

is high likelihood that probabilistic

performance goals can be achieved.

1. INTRODUCTION

Use of deterministic design and evaluation

criteria to meet probabilistic performance

goals is the approach used by DOE facilities across the United States as given in UCRL15910, "Design and Evaluation Guidelines

for Department of Energy Facilities Subjected

to Natural Phenomena Hazards." An

overview of the approach is presented in this

paper. References 1-14 provide related

background material.

UCRL-15910 wind and tornado provisions use deterministic evaluation criteria with the

hazard annual probability of exceedance

specified to get design/evaluation wind

speeds. The evaluation of response and

structure or equipment capacities are performed using deterministic methods which

are familiar to engineers. Wind speeds are developed from probabilistic hazard curves

are used to meet performance goals for

various building occupancies.

2. PERFORMANCE GOALS

Performance goals may be expressed in terms

of annual probability of exceedance of some

level of damage. Levels of damage can

include that damage beyond which occupants

are endangered, beyond which hazardous

materials cannot be confined, beyond which a

facility cannot safely shutdown, or beyond

which a facility cannot continue its mission.

The performance goal for the safety of

Advanced Light Water Reactors is on the

order of IX 10'^ annual probability of

exceedance for external events induced core damage. Performance goals upon which

design/evaluation guidelines are based for

DOE facilities in UCRL-15910, are presented

in Table 1.

For each performance goal there are separate

wind and tornado design and evaluation

criteria. DOE management categorize

facilities or individual structures, systems, or components depending on the cost, mission

importance, or hazard to people or the

* U. S. Department of Energy, Office of Safety

Appraisals, Washington, D.C., 20585

** Texas Tech University

^ Lubbock, Texas 79409

***Lawrence Livermore National Laboratory

Livermore, CA 94550 . :

3
environment. For UCRL- 159 10, a DOE

natural phenomena hazards panel selected

reasonable and achievable performance goals,

bounded by current practice.

• For ordinary facilities, the

performance goals are consistent with

design according to conventional

building code provisions.

• For high hazard facilities, the

performance goals are comparable to performance reached by nuclear

power plants as measured by

probabilistic risk assessments

(PRAs).

3. PERFORMANCE GOAL

ACHIEVEMENT

Structure/equipment performance is a function of: (1) the likelihood of hazard occurrence and (2) the strength of the

structure or equipment item. Therefore,

design and evaluation criteria have been

developed to attain performance goals by: (1)

specification of hazard probability for

definition of wind and tornado loadings and

(2) specification of response evaluation

methods, acceptance criteria, and design

detailing requirements with controlled levels

of conservatism. Acceptable performance can only be reached by consistent

specification of all design or evaluation

criteria elements as shown below:

Loading

Spacification Raspons*

J Evaluation o

uAccaptable

Lsvels ) Design

Oataiis

UCRL- 159 10 wind and tornado design and

evaluation guidelines contain the following

provisions:

Lateral Force Provisions

Damage Control Provisions

Detailing Provisions

Quality Assurance and Peer Review

Provisions

Wind/tornado performance depends on the

level of the hazard and on conservatism in the

response evaluation and acceptance criteria.

For example, a performance goal of 1X10"^

can be achieved either by: (1) a conservative

evaluation/acceptance approach for a more

frequent hazard probability, such as IX 10"^ or by: (2) a median centered

evaluation/acceptance approach coupled with

a less frequent IXIO''^ hazard probability.

UCRL- 159 10 uses the former approach

because conservative evaluation/acceptance

approaches are well established, extensively

documented, and commonly practiced.

4. DESIGN AND EVALUATION

CRITERIA FOR WIND LOAD

A uniform approach to wind load

determination that is applicable to the design

of new facilities and the evaluation of existing ones has been developed for DOE. A

uniform treatment of wind loads is

recommended to accommodate extreme,

hurricane, and tornado winds. Buildings or

facilities are first assigned appropriate usage

categories. Criteria are recommended such

that the performance goals for each category can be achieved. Procedures according to ASCE 7-88 (formerly ANSI A58.1) are recommended for determining wind loads

produced by straight, hurricane, and tornado

winds. The extreme wind/tornado hazard

models developed for DOE sites were used to establish site-specific criteria for each of

some 25 DOE sites.

The performance goals established for

General Use and Important or Low Hazard

usage categories are met by conventional

building codes or standards. These criteria

do not account for the possibility of tornado

winds, because wind speeds associated with

extreme winds typically are greater than those

for tornadoes at exceedance probabilities

greater than approximately IXIO""^. For this

reason, tornado design criteria are specified

only for buildings and facilities in Moderate

and High Hazard categories, where hazard

4
exceedance probabilities are less than ' • 1X10-4

The traditional approach for establishing

tornado design criteria is to select extremely

low exceedance probabilities. For example,

the exceedance probability for design of

commercial nuclear power plants is IX lO'^.

There are reasons for departing from this

traditional approach. The low exceedance

value for commercial nuclear power plants

was established circa 1960 when very little was known about tornadoes from an

engineering perspective. Much has been

learned about tornadoes since that time. Use

of a low hazard probability is inconsistent

with the practice relating to other natural

hazards, such as earthquakes. There are many uncertainties in tornado hazard

probability assessment, but they are not

significantly greater than the uncertainties in

earthquake probability assessment. The

strongest argument against using low

probability criteria is that a relatively short

period of record (37 years) must be

extrapolated to extremely small exceedance

probabilities. For these reasons, an

alternative approach was used in the

UCRL-15910guideUnes.

4. 1 Establishment of Wind and Tornado

Hazard Annual Probabilities

The rationale for establishing tornado criteria

is described below. Figure 1 shows the

tornado and straight wind hazard curves for

two DOE sites (SLAC and ORNL). The

wind speed at the intersection of the tornado

and straight wind curves is defined for

purposes of this discussion as the transition

wind speed. An exceedance probability is associated with each transition wind speed.

If the exceedance probability of the transition

wind speed is less than 1X10'^ per year,

tornadoes are not a viable threat to the site,

because straight wind speeds are higher than

tornadoes for a given return period. Thus,

from Figure 1, tornadoes need not be

considered at SLAC, but should be

considered at ORNL.

Best-estimate wind speeds were developed

for each DOE site, along with the transition

wind speed. Those sites with transition wind

speed exceedance probabilities greater than

10'^ should be designed for tornadoes;

others should be designed for extreme winds

or hurricanes.

The tornado wind speed is obtained by

selecting the wind speed associated with an exceedance probability of 2X10"^ per year.

The value of 2X10-^ is the largest one that can be used and still represent a point on the

tornado hazard curve. For example, the

tornado wind speed for the ORNL is 130

mph (peak gust at 10m).

A comparison of the slopes of the tornado

hazard curves for the DOE sites reveals that

the slopes are essentially the same even though the transition wind speeds are

different. The criteria required to meet the

performance goals of Moderate and High

Hazard facilities can be met by using

multipliers that are equivalent to an

importance factor in the ASCE 7-88 design

procedure. The multipliers are specified in

lieu of two different exceedance probabiUties

for Moderate and High Hazard facihties. The

value of the importance factor is selected to achieve lower probability of tornado damage

for High Hazard facilities compared to Moderate Hazard facilities. The importance

factors are then chosen to meet the

performance goals.

In general, design criteria for each usage

category include:

1 . Annual hazard exceedance probability.

2. Importance factor.

3. Wind generated missile parameters for

Moderate and High Hazard facilities. 4. Tornado parameters for Moderate and

High Hazard facilities, if appUcable.

The criteria are formulated in such a way that a uniform approach for determining design

wind loads, per ASCE 7-88, can also be used

for extreme, hurricane, and tornado winds.

In order to apply the ASCE 7-88 procedure,

tornado gust wind speeds must be converted

to fastest-mile. Appropriate gust response

factors and velocity pressure exposure

5
coefficients are utilized in the process of

determining wind loads. Appropriate

exposure categories also are considered in the

wind load calculations. Open terrain

(Exposure C) should always be assumed for

tornado winds, regardless of the actual terrain

conditions.

4.2 CRITERIA FOR DESIGN OF

FACILITIES

The criteria presented herein are consistent

with the performance goals described earlier.

Buildings or facilities in each category have a

different role and represent different levels of

hazard to people and the environment. In addition, the degree of wind hazard varies

geographically. Facilities in the same usage

category, but at different geographical

locations, will have different criteria specified

to achieve the same performance goal.

The minimum wind design criteria for each of

the four usage categories are summarized in

Table 2. The recommended basic wind

speeds for extreme wind, hurricanes, and

tomadoes are contained in Table 3. All wind

speeds are fastest-mile. Minimum

recommended basic wind speeds are noted in

the table.

4.2.1 General Use Facilities

The performance goals for General Use

facilities are consistent with objectives of

ASCE 7-88 Building Class I, Ordinary

Structures. The wind-force resisting

structural system should not collapse under

design load. Survival without collapse

implies that occupants should be able o find an area of relative safety inside the building.

Breach of the building envelope is acceptable,

since confinement is not essential. Flow of

air through the building and water damage are acceptable. Severe damage, including total

loss, is acceptable, so long as the structure

does not collapse.

The ASCE 7-88 standard calls for the basic

wind speed to be based on an exceedance

probability of 0.02 per year. The importance

factor for this class of building is 1.0. For

those sites within 100 miles of the Gulf of

Mexico or Atlantic coastlines, a slightly

higher importance factor is recommended to account for hurricanes.

Terrain surrounding the facilities should be

classified as Exposure B, C, or D, as appropriate. Gust response factors and

velocity pressure exposure factors should be

used according to rules of the ASCE 7-88

procedures.

Wind pressures are calculated on the walls

and roofs of enclosed buildings by

appropriate pressure coefficients specified in

the ASCE 7-88 standard. Openings, either of

necessity or created by wind forces or

missiles, result in intemal pressures that can

increase wind forces on components and

cladding. The worst cases of combined

intemal and external pressures should be

considered, as required by the standard.

Structures in the General Use category may

be designed by either allowable stress design

(ASD) or strength design (SD) as appropriate

for the material used in construction. Except

when applicable codes provide otherwise,

plausible load combinations shall be

considered to determine the most unfavorable

effect on the building, foundation, or

structural member being considered. When

using ASD methods, allowable stresses

appropriate for the building material shall be

used with the following combinations that

involve wind:

(a) DL + W

(b) 0.75(DL + W + LL)

where

DL = dead load

LL = live load

W = wind load

When using SD methods, the following load

combinations that involve wind are recommended:

(a) U = 0.9DL+ 1.3W

(b) U = 1.2DL-h0.5LL+ 1.3W

The SD method requires that the strength

provided be greater than or equal to the

strength required to carry the factored loads.

6
Appropriate strength reduction factors must

be applied to the nominal strength of the

material being used.

4.2.2 Important or Lx)w Hazard Facilities

Important or Low Hazard facilities are equivalent to essential facilities (Class II), as defined in ASCE 7-88. The structure's main

wind-force resisting structural systems shall

not collapse at design wind speeds.

Complete integrity of the building envelope is not required because no significant quantities

of toxic or radioactive materials are present.

However, breach of the building envelope

may not be acceptable if wind or water

interfere with the facility function. If loss of

facility function is caused by water damage to sensitive equipment, collapsed interior

partitions, or excessive damage to HVAC

ducts and equipment, then loss of cladding

and missile perforation at the design wind

speeds must be prevented - r . An annual wind speed exceedance probability

of 0.02 is specified, but the importance factor

for important or Low Hazard category

structures is 1.07. For those sites located

within 100 miles of the Gulf of Mexico or Atlantic coastlines, a slightly higher

importance factor is used to account for

hurricane winds.

Once the design wind speeds are established

and the importance factors applied, the

determination of wind loads on Important or Low Hazard category structures is identical to that described for General Use category

structures. Facilities in this category may be

designed by ASD or SD methods, as appropriate, for the construction material.

The load combinations described for General

Use structures are the same for Important or Low Hazard structures. However, greater

attention should be paid to connections and

anchorages for main members and

components, such that the integrity of the

structure is maintained.

4.2.3 Moderate Hazard Facilities

The performance goal for Moderate Hazard

facilities requires more rigorous criteria than

is provided by standards or model building

codes. For some DOE sites, tornadoes must

be considered.

Extreme Winds and Hurricanes

For those sites where tornadoes are not a

viable threat, the recommended basic wind

speed is based on an annual exceedance

probability of IX 10" 3. The importance

factor is 1.0. For those sites located within

100 miles of the Gulf of Mexico or Atiantic

coastiines, a slightiy higher importance factor

is specified to account for hurricanes.

Once the basic wind speeds are established

and the importance factors applied,

determination of Moderate Hazard category

wind loads is identical to that described for

the General Use category. Facilities in this

category may be designed by ASD or SD

methods, as appropriate, for the material

being used in construction. Plausible load

combinations shall be considered to

determine the most unfavorable effect on the

building, foundation, or structural member

being considered. When using ASD,

allowable stresses appropriate for the

building material shall be used with the

following wind load combinations:

(a) U = 0.9(DL + W)

(b) U = 0.67(DL + W + LL)

The SD load combinations recommended for

the Moderate Hazard category are:

(a) U = DL + 1.3W

(b) U= 1.1DL-^0.5LL+ 1.2W

Greater attention should be paid to connections and anchorages for main

members and components, such that the

integrity of the structure is maintained.

A minimum missile criteria is specified to account for objects or debris that could be

picked up by extreme winds, hurricane

winds, or weak tornadoes. A 2X4-in. timber

plank weighing 15 lbs. is the specified

missile. Its impact speed is 50 mph at a maximum height of 30 ft above ground level.

The missile will break glass; it will perforate

7
sheet metal siding, wood siding up to 3/4 in. thick, or form board. The missile could pass

through a window or a weak exterior wall

and cause personal injury or damage to

interior contents of a building. The specified

missile will not perforate unreinforced

concrete masonry or brick veneer walls or

other more substantial walls.

Tornadoes

For those sites requiring design for

tornadoes, the criteria are based on sitespecific studies. The basic wind speed is associated with an annual hazard probability

of exceedance of 2X10'^. The wind speed

obtained from the tornado hazard model is converted to fastest-mile. The importance

factor for the Moderate Hazard category is 1.0.

With the wind speed converted to fastest-mile

wind and an importance factor of 1.0,

equations in the ASCE standard should be

used to obtain design wind pressures on the

structure. Exposure Category C should

always be used with tornado winds

regardless of the actual terrain roughness.

The velocity pressure exposure coefficient

and the gust response factor are obtained

from appropriate tables in the ASCE

standard. External pressure coefficients are used to obtain tornado wind pressures on

various surfaces of the structure. A

distinction is made between the main windforce resisting system and components and

cladding.

If the building is not specifically sealed to maintain an internal negative pressure for

confinement of hazardous materials, or, if openings greater than one square foot per

1000 cubic foot of volume are present, or, if openings of this size can be created by

missile perforation, then the effects of

internal pressure should be considered

according to ASCE procedures. On the other

hand, if the building is sealed, then

atmospheric pressure change (APC)

pressures associated with the tornado should

be considered.

APC pressure is half its maximum value at

the radius of maximum wind speed in a

tornado. Thus, critical tornado loading will

be one-half maximum APC pressure plus

maximum tornado wind pressure. A loading

condition of APC alone can occur on the roof

of a buried tank or sand filter, if the roof is exposed at the ground surface. APC

pressure always acts outward. The effect of

rate of pressure change on ventilation

systems should be analyzed to assure that it does not interrupt function or processes

carried out in the facility. Procedures and

computer codes are available for such

analyses.

Plausible load combinations shall be

considered to determine the most unfavorable

effect on the building, foundation, or

structural member being considered. When

using ASD methods, allowable stresses

appropriate for the building materials shall be

used with the following load combinations

that involve wind.

(a) U = 0.75(DL + Wt)

(b) U = 0.625(DL + Wt LL)

The SD load combinations recommended for

the Moderate Hazard category are:

(a) U = DL Wt

(b) U = DL-HLL-^Wt

where

Wt = tornado loading, including APC, as

appropriate.

Two missiles are specified as minimum

criteria for this usage category. The 2X4-in.

timber plank weighing 15 lbs. is assumed to

travel in a horizontal direction at a speed up to 100 mph. The horizontal speed is effective

up to a height of 1 50 ft above ground level.

If carried to a great height by the tornado

winds, the timber plank could achieve a terminal vertical speed of 70 mph in falling to the ground. The horizontal and vertical

speeds are assumed to be uncoupled and they

should not be combined. The missile will

perforate most conventional wall and roof

cladding except reinforced masonry or

concrete. The cells of concrete masonry

8
walls must be filled with grout to prevent

perforation by the timber missile. The

second missile is a 3-in. diameter standard

steel pipe, which weighs 75 lbs. It can

achieve a horizontal impact speed of 50 mph

and a vertical speed of 35 mph. Its horizontal

speed could be effective to heights of 75 ft above ground level. The missile will

perforate conventional metal siding,

sandwich panels, wood and metal decking on

roofs, and gypsum panels. In addition, it will perforate unreinforced concrete masonry

and brick veneer walls, reinforced concrete

masonry walls less than 8 in. thick, and

reinforced concrete walls less than 6 in. thick. Although wind pressure, APC, and

missile impact loads can act simultaneously in a tornado, the missile impact loads can be

treated independently for design and

evaluation purposes.

4.2.4 High Hazard Facilities

The performance goal can be achieved for

this category if the main wind-force resisting

members do not collapse, structural

components do not fail, and the building

envelope is not breached at the design wind

loads. Loss of cladding, broken windows,

collapsed doors, or significant missile

perforations must be prevented. Strong air

flow through the building or water damage

cannot be tolerated.

Extreme Winds and Hurricanes

For those sites that do not require specific

design for tornado resistance, the

recommended basic wind speed is based on

an annual hazard exceedance probability of

IX 10'"^. The importance factor is 1.0. The

wind speed is fastest-mile at an anemometer

height of 10 meters above ground level.

Once the basic wind speeds are established

and the importance factors applied,

determination of High Hazard facility wind

loads is identical to that described for the

General Use category. Facilities in this

category may be designed by ASD or SD

methods, as appropriate, for the material

being used in construction. Recommended

wind load combinations are the same as for

Moderate Hazard facilities. Greater attention

should be paid to connections and anchorages

for main members and components, such that

the integrity of the strucnare is maintained.

The missile criteria is the same as for the

Moderate Hazard category, except that the

maximum height achieved by the missile is 50 ft, instead of 30 ft.

Tomadoes ,

For those sites requiring design for tornado

resistance, the criteria is based on sitespecific studies. The recommended basic

wind speed is associated with an annual

hazard probability of exceedance of 2X10'^

(the same as the Moderate Hazard category).

The wind speed obtained from the tornado

hazard model is converted to fastest-mile.

The importance factor for the High Hazard

category is 1.35.

With the wind speed expressed as fastestmile and an importance factor of 1.35,

equations should be used to obtain design

wind pressures on the structure. Exposure

Category C should always be used with

tornado winds regardless of actual terrain

roughness. The velocity pressure exposure

coefficient and the gust response factor are obtained from appropriate tables in the ASCE

standard. External pressure coefficients are used to obtain tornado wind pressures on

various surfaces of the structure. A

distinction is made between the main windforce resisting system and components and

cladding in determining wind pressures.

If the building is sealed to confine hazardous

materials, the wind and APC load

combinations specified for the Moderate

Hazard usage category also should be used

for this category. The effects of rate of

pressure change on ventilating systems

should be analyzed. Recommended tornado

wind load combinations for Moderate Hazard

facilities also apply to High Hazard facilities.

Three missiles are specified as minimum

criteria for this usage category. The 2x4-in.

timber plank weighs 15 lbs. and is assumed

to travel in a horizontal direction at speeds up

9
to 150 mph. The horizontal missile is effective to a maximum height of 200 ft above ground level. If carried to a great

height by the tornado winds, it could achieve

a terminal speed in the vertical direction of

100 mph. The horizontal and vertical speeds

are uncoupled and should not be combined.

The missile will perforate most conventional

wall and roof cladding except reinforced

masonry and concrete. Each cell of the

concrete masonry shall contain a 1/2-in.-

diameter rebar and be grouted to prevent

perforation by the missile. The second

missile is a 3-in. -diameter standard steel pipe,

which weighs 75 lbs. It can achieve a horizontal impact speed of 75 mph and a

vertical speed of 50 mph. The horizontal

speed could be effective at heights up to 100

ft above ground level. This missile will

perforate unreinforced concrete masonry and

brick veneer walls, reinforced concrete

masonry walls less than 12 in. thick, and

reinforced concrete walls less than 8 in. thick. The third missile is a 3000-lb

automobile that is assumed to roll and tumble on the ground and achieve an impact speed of

25 mph. Impact of an automobile can cause excessive structural response to columns,

walls, and frames. Impact analyses should

be performed to determine specific effects.

Collapse of columns, walls, or frames may

lead to further progressive collapse.

5. SUMMARY AND CONCLUSIONS

UCRL-15910 is an example of deterministic

design/evaluation criteria developed to achieve probabilistic performance goals.

UCRL-15910 also covers seismic and flood

as well as wind and tornado criteria as discussed in this paper. The criteria

developed is consistent with the consensus

standard ASCE 7-88. Conservatism is specified which is sufficient to achieve the

performance goals. This conservatism

increases from general use to high hazard

usage category.

The UCRL-15910 wind/tornado design and

evaluation guidelines follow the philosophy

of: (1) gradual reduction in hazard annual

exceedance probability; and (2) gradual

increase in conservatism of evaluation

procedure as one goes from a General Use to

a High Hazard facility. Four separate sets of

design/evaluation criteria have been presented

in UCRL-15910, each with a different

performance goal. In all these criteria,

loading is selected from hazard curves on a

probabilistic basis but response evaluation

methods and acceptable behavior limits are deterministic approaches with which design

engineers are familiar.

6. REFERENCES

1. Advanced Light Water Reactor

Requirements Document, Volume 1

,

ALWR Policy and Summary of TopTier Requirements, Final Draft for

Review, 1990, Electric Power

Research Institute, Palo Alto,

California.

2. Individual Plant Examination of

External Events (IPEEE) for Severe

Accident Vulnerabilities - lOCFR

50.54 (f), (Generic Letter No. 88-20

- Supplement 4, Draft for

Comment), 1990, U.S. Nuclear

Regulatory Commission.

3. Kennedy, R.P., S.A. Short, J.R.

McDonald, M.W. McCann, R.C.

Murray, and J.R. Hill, June 1990,

Design and Evaluation Guidelines

For Department of Energy Facilities

Subjected to Natural Phenomena

Hazards, Lawrence Livermore

National Laboratory Report UCRL15910, 1990.

4. Regulatory Guide 1.76, U.S.

Nuclear Regulatory Commission.

5. Safety Analysis Reports for Nuclear

Power Plants, LWR Edition, 1989,

NUREG-0800, U.S. Nuclear

Regulatory Commission.

6. Uniform Building Code. 1988

Edition, International Conference of

Building Officials, Whittier,

California.

10
7. Coats, D.W. and R.C. Murray,

Natural Phenomena Hazards

Modeling Project: Extreme

Wind/Tornado Hazard Models for

Department of Energy sites. Report

UCRL-53526 Rev. 1, Lawrence

Livermore National Laboratory,

Livermore, California, 1985.

8. McDonald, J.R., Structural Details

for Wind Design, Lawrence

Livermore National Laboratory,

Report UCRL-21131, 1988.

9. U.S. Department of Energy, General

Design Criteria, DOE or Dev

6430.1 A, Washington, D.C., 1989.

10. Minimum Design Loads for Building

and Other Structures, ASCE 7-88

(formerly ANSI A58.1), American

Society of Civil Engineers, July

1990.

1 1 . Evaluation of External Hazards to Nuclear Power Plants in the United

States, NUREG/CR-5042, UCID21223, Lawrence Livermore

National Laboratory, Livermore,

California, 1988.

12. McDonald, J.R., Extreme winds and

Tornadoes: An Overview, Lawrence

Livermore National Laboratory,

Report UCRL-15745, 1985.

13. McDonald, J.R., Extreme Winds

and Tornadoes: Design and

Evaluation of Buildings and

Structures, Lawrence Livermore

National Laboratory, Report UCRL15747, 1985.

14. ANS 2.3, Standard for Estimating

Tornado and Extreme Wind

Characteristics at Nuclear Power

Plant Sites, ANSI Standard

published by the American Nuclear

Society, La Grange Park, Illinois,

1983.

11
Table 1. Performance Goals for Each Usage Category

Usage

Category

Performance Goal

Description

Performance Goal Median

Annual Probability of Exceedance

General

Use

Maintain Occupant Safety - 10 ' of the onset of major component^' damage

to the extent that occupants are endangered

Important or Low Hazard

Occupant Safety, Continued Operation

with Minimal Interruption - 5x10-* of component damage to the extent that

the component carwiot perform its function

Mooerate

Hazard

Occupant Safety, Continued Function,

Hazard Confinement - 1 0 * of component damage to the extent that the

component cannot perform its function

High

Hazard

Occupant Safety, Continued Function.

Very High Confidence of Hazard Confinement

~ 1 0 ' of component damage to the extent that the

component cannot perform its function

Reactor

Facilities Beyond the scope of UCRL-15910

Component refers to structure, equipment, or distribution system.

Table 2. Summary of Minimum Wind Design Criteria

Building Category General Use Important or Low Hazard

Moderate Hazard High Hazard

Annual Probability

of Exceedance

2x10-2 2x10-2 1x10-3 1x10-4 w

ind

Importance

Factor

1.0 1.07 1.0 1.0

Missile Criteria 2x4 timber plank 15 lb @

50 mph (horiz.); max.

height 30 ft.

2x4 timber plank 15 lb @

50 mph (horiz.); max.

height 50 ft.

Annual Hazard

Probability

of Exceedance

2x10-5 2x10-5

Importance Factor 1 = 1.0 1 = 1.35 APC 40 psf @ 20 psf/sec 125 psf @ 50 psf/sec

T

o

rn

ad

o

Missile Criteria

2x4 timber plank 15 lb @

100 mph (horiz.); max.

height 150 ft; 70 mph (vert.)

3 in. dia. std. steel pipe, 75

lb @ 50 mph (horiz.); max.

height 75 ft, 35 mph (vert.)

2x4 timber plank 15 lb @

150 mph (horiz.), max,

height 200 ft; 100 mph

(vert.)

3 in. dia. std. steel pipe, 75

lb @ 75 mph (horiz.); max.

height 100 ft, 50 mph (vert.)

3,000 lb automobile @ 25

mph, rolls and tumbles

12
Table 3. Recommended Basic Wind Speeds for DOE Sites

Fastest-Mile Wind Speeds at 10m Height

Building Category

General

Use

Important

or Low

Hazard

Moderate Hazard High Hazard

Wind Wind Wind Tornado Wind Tornado

DOE PROJECT SITES 2x10-2 2x10-2 1x10-3 2x10-5 1x10-4 2x10-5 Bendix Plant, MO 72 72 - 144 - 144

Los Alamos National Scientific Laboratory, NM

77 77 93 - 107 -

Mound Laboratory, OH 73 73 1 wO 1 oo

Pantex Plant, TX 78 78

Rocky Flats Plant, CO 109 109 I'M 1 D 1 Sandia National Laboratories,

Albuquerque, NM

78 78 93 107

Sandia National Laboratories,

Uvermore, CA

72 72 9v 1 1 1 1 o

Pinellas Plant, FL 93 93 130 - 150 - Argonne National Laboratory-East, IL 70(1) 70(1) - 142 -- 142 Argonne National Laboratory-West, ID 70(1) 70(1) 83 - 95 -

Brookhaven National Laboratory, NY 70(1) 70(1) - 95(2) - 95(2)

Princeton Plasma Physics Laboratory, NJ 70(1) 70(1) 103 103

Idaho National Engineering Laboratory 70(1) 70(1) 84 95

Feed Materials Production Center, OH 70(1) 70(1) 139 139

Oak Ridge National Laboratory, X-10, K-25,

and Y-12. TN

70(1) 70(1) 113 113

Paducah Gaseous Diffusion Plant, KY 70(1) 70(1) — 144 — 144 ronsmouin uaseous urrrusion riani, un 7n(1) 7n(1) 110 110

njevaua lesi sue, inv 70 70 87 100

Hanford Project Site WA 70(1) 70(1) 80(1) 90(1) Lawrence Berkeley Laboratory, CA 72 72 95 111 Lawrence Uvermore National Laboratory, CA

72 72 96 113

Lawrence Uvermore Nationad Laboratory, Site 300, CA

80 80 104 125

Energy Technology and Engineering

Center, CA

70(1) 70(1) 95(2) 95(2)

Stanford Linear Accelerator Center, CA 72 72 95 112

Savannah River Plant, SO 78 78 137 137

NOTES:

(1) Minimum extreme wind speed. . -

(2) Minimum tornado speed.

(3) Although extreme winds govern at Rocky Flats, it is recommended that facilities be designed for the tornado missile

criteria. APC need not be considered.

13
Figure 1. Straight Wind and Tornado Regions of Wind Hazard Curves

14
Analysis of Tornado in Mobara City on December 11, 1990

H NiinoS T Fujitani^, T Murota^, N Yamaguchi^ and H Okada^

ABSTRACT

On the evening of December 11, 1990, a tornado hit the Mobara City in the central part of the Boso Peninsula, about 55km

southeast of Tokyo. The tornado produced

one of the severest damages in the Japanese

records. One person died, 73 persons were

injured, 82 houses completely destroyed,

161 houses partially destroyed and 1504

houses damaged. The characteristics of the

tornado as revealed from damage survey are reported.

KEYWORDS :tornado, damage survey,

estimated wind speed, translational speed.

1. INTRODUCTION

In Japan there have been enough records

of strong winds associated with typhoons

and extratropical cyclones to establish wind

resistant design codes based on their statistical properties. Since tornadoes occur

less frequently over narrow areas and last

rather shortly, however, only sparse data

which directly record strong winds associated with them exist. Accordingly, damage survey has been one of the effective

means to study characteristics of tornadoes;

in fact, their general characteristics such as dimensions, translational velocity and maximum wind, have been reasonably estimated

through damage surveys.

Though the immediate damage due to

tornadoes may be confined to a small area, the probability that such a damage will lead

to the secondary damage to much wider

area seems to be increasing as the urban

functions become sophisticated and complex. Constructions which may be liable to

this probability include nuclear power plants,

chemical factories, bullet trains and ultralong-span bridges. Thus, it seems neces- sary to establish wind resistant design codes

which consider strong winds due to tornadoes as soon as possible. It seems also nec- essary to establish a warning system similar

to the tornado warning in the United States.

Since tornadoes are less frequent phenomena, it is important to accumulate data on their characteristics before any statistical analysis is done. Whenever a tornado

occurs, it is an urgent task to make a quick

damage survey and collect every detum re- lated to the tornado.

On the evening of December 11, 1990, a tornado hit the Mobara City which is located in the center of the Boso Peninsula, 55km southeast of Tokyo and produced one of the severest damage in the

Japanese record. One person died, 73 per- sons were injured, 82 houses completely destroyed, 161 houses partially destroyed and

1504 houses damaged. Scientists of Meteorological Research Institute(MRI) and

Building Research Institute(BRI) started a

detailed investigation on the tornado on December 12. In the following, the characteristics of Japanese tornadoes are introduced

briefly, and then the characteristics of the

Mobara Tornado as revealed from the damage survey are reported.

^ Physical Meteorology Division, Meteorological Research Institute.

^ Applied Meteorology Division, Meteorological Research Institute.

^ The Third Division, Building Research

Institute.

15
2. CHARACTERISTICS OF JAPANESE TORNADOES

The characteristics of Japanese tornadoes described in this chapter are extracted

from an excellent technical report by Mitsuta^^ who summarized the statistics on the

tornadoes in Japan from 1961 to 1982:

1) The number of tornadoes during 22 years

is 396. This gives an average number of 18

per year, which is much smaller than 771

per year in the United States^^.

2) There is a distinct regional difference in

the occurence of tornadoes. The Pacific side

of the Japan Island experiences more tornadoes than the Japan Sea side does. Among

43 prefectures, the Kagoshima Prefecture,

the southest prefecture of the Kyushu Is- land, experiences tornadoes most frequently

(37 tornadoes during 22 years). The number

density per 10^A;m^ per year is maximum for

the Okinawa Prefecture, the southest prefecture of Japan, and is 6.3. This number is larger than that for the Oklahoma State in

the United States.

3) Tornadoes occur most frequently in September, and next in August. These two

months account about 38 percent of the total occurence of tornadoes. March has tor- nadoes least frequently. September and August are the months when the Japan Islands are hit by typhoons most frequently. Since

tornadoes are often accompanied with typhoons, this may explain the reason why

these two months have such frequent oc- curence of tornaxioes.

4) About three fourths of tornadoes move

toward directions between east and north.

Only few estimations for the translational

speed of tornadoes exist. They give the av- erage value of about 10m/s(36km/h).

5) 0n average twenty eight houses are totally

destroyed, 0.5 persons are killed and 20.6

persons are injured per year.

6) No wind record at the radius of maximum wind exists. Wind speeds of more

than 60m/s at Hachijou Island in 1964 and

more than 80m/s at Tokyo in 1978 have

been recorded near tornado paths. Wind

speed of more than lOOm/s have been estimated from a damage to a construction in

Toyohashi City in 1969.

3.MOBARA TORNADO

3.1. Meteorological conditions

From 1800 to 1930JST on the evening

of December 11, 1990, damages due to

heavy winds were observed at Mobara, Futtsu, Kimitsu, Kamogawa, Choshi Cities

and Maruyama Town in the Chiba Prefecture and hails over much wider region.

Among the cities, the Mobara City in the

central part of the Boso Peninsula experienced the severest damage which has been

concluded to be due to a tornado because of

the following reasons:

1) A funnel cloud was recorded on a video

film which was taken from 1km distance.

The video shows anti-clockwise rotation of

the funnel cloud.

2) The wind directions estimated from damage survey at more than 60 locations show

the existence of a convergent flow associated

with a cyclonic vortex.

3) A sudden pressure drop of about 9hPa

was recorded at the Chosei High School near

the damage path (Sect. 3. 1.2). A number

of habitants in the damage path reported

symptoms such as singing in the ears and

nosebleeding which seem to be caused by a sudden and large pressure drop.

It is not possible to tell if all the damages due to the heavy winds in the Chiba

Prefecture were caused by tornadoes, since

we have not made a damage survey over

the cities except for the Mobara City. It

is noted, however, that a funnel cloud was

eyewitnessed in the Kamogawa City.

3.1.1. Synoptic condition

16
An extratropical cyclone of 1004hPa

was located in the east of the Korea Peninsula at 0900JST in the morning of December 11. It moved eastward while developing

and reached the Noto Peninsula at 1500. A

cold front extended southeastward from the

center of the cyclone. Another cyclone was

generated on the cold front near the boundary between the Aichi and Miye Prefectures

at about 1500 and moved eastward along the

south coast of the Honshu Island while developing. It reached the boundary between

the Shizuoka and Kanagawa Prefectures at 1800, then the boundary between the Tokyo

and Saitama Prefectures at 2100.

Figure 2 shows the surface weather map

at 1800, about 1 hour prior to the tornadogenesis. The cyclone is located at the west

part of the Kanto Plain. The cold front is extending toward the Izu Peninsula. According to the Tokyo Radar of the Japan

Meteorological Agency (JMA), a group of

cumulonimbuses were located about 150km

east of the cold front. Two strong cells were

located at the south edge of the cumulonimbuses. These two cells moved northeastward

and generated the heavy winds and tornadoes in the Chiba Prefecture. Thus, the Mobara Tornado was generated in the warm

sector of the extratropical cyclone. The re- sults of the analysis of the mesoscale features which led to the tornadogenesis are described by Niino et al."^^.

3.1.2. Meteorological records

No meteorological records were obtained in the damage path of the tornado.

However, several records were found near

the path. These records* will be described

in the following briefly. The analysis of

these records is still in progress and its re- sults will be reported elsewhere"*^.

l)Wind speed and direction

Maximum wind speed of 30.8m/s was

recorded at 1910 by an anemometer in the

farm field of the Mobara Agricultural High

School ( at the height of 10m at Point D) lo- cated about 900m east of the tornado path.

The wind direction was SE before 1850,

then changed to N~E, and was SE right

before the tornado cut the electric power

supply. The anemometer at the Mobara

Transforming Station of the Tokyo Electric

Power Company(Point F), about 700m east

of the path, recorded the maximum speed

of 29.5m/s at about 1920. The variation of

the wind direction at Point F is quite similar

to that at Point D. The wind direction before 1840 was ESE, then showed large fluctuations, and was SE when the maximum

speed was observed. The anemometer of the

Environmental Division, the Chiba Prefecture, at Point B showed that the wind direction right before the electricity was stopped

was NE. The wind directions at these three

points are consistent with the cycloninc ro- tation of the tornado as revealed from the

video image.

2) Pressure

Pressure records were obtained at Point

C, D and E in Fig. 3. A sudden pressure

drop of about 9hPa was recorded at about

1900 at the Chosei High School (Point C),

about 400m east of the path. This pres- sure drop is considered to be due to the

tornado itself. At the Mobara Agricultural

High School (Point D), where the wind data

were recorded, a pressure drop of 2hPa was

recorded at about 1900, and at the Fujimi

Junior High School(Point E), about 2000m

west of the path, a drop of 2.5hPa was

recorded. These pressure drops are considered to be due to a mesocyclone which

spawned the tornado.

* The time described in section 3.1. will

be based on the time recorded on barographs, anemograph and others. Since no

caribration of the time was done, the description of time will involve some ambiguity.

17
3. 2. Summary of the damage

The damages due to the Mobara Tornado were summarized by the Mobara City

on January 17, 1991, and is cis follows:

1) Damage to human : 1 fatality and 73 injuries.

2) Damage to houses : 82 houses completely

destroyed, 161 houses partially destroyed

and 1,504 houses damaged.

3) Sum of damage in Japanese currency : 3,700 million yen.

The damages to lifelines summarized

by the Fire Fighting and Discister Prevention Section, the Chiba Prefecture, on December 17, 1990, are:

1) Stoppage of electric power: 13,000 houses.

2) Interruption of telephone: 1,000 circuits.

These damages were one of the severest caused by a tornado in Japan after the

World War II and the Disaster Relief Act

retroactive to December 11 was invoked to

the Mobara City at 1700 on December 12.

3. 3. Damage distribution

Figure 3 shows the distribution of damages to houses based on the data provided

by the Mobara City. Solid circles show

completely destroyed houses, solid triangles

partially destroyed houses and open circles

damaged houses. As will be described in

Section 3.4.1., it has been found that the

tornado moved northeastward. The first damage started near the intersection, where

the old Route 128 meets Route 409, at Midori-Machi in the south part of the City.

The width of the damage path increases

northward along the JR Sotobo Line. The

damage path includes the area around the

JR Shin-Mobara Station and extends to the

Ko-bayashi Area in the north part of the

City. The length of the damage path is about 6.5km and its average width is about

500m ( The maximum width is 1200m).

Most of the severe damages to houses

are concentrated in the Takashi Area in the

center of the City. Figure 4 shows the damages in the Takashi Area photographed from

a helicopter of the Kyodo-Tsushin Company

on the morning of December 12. If the distribution of the damages near the Mobara

City Gymnasium north of the Takashi Area

is examined more carefully, one notices that

the distribution has an eastward protuber- ance of about 500m. It is also noteworthy that several houses in the east part of

the protuburence were totally destroyed or

severely damaged.

3.4. Characteristics of the tornado

In the following the characteristics of

the tornado as revealed from the damage

survey are described.

3. 4.1. Tornado path

Figure 5 shows the spatial distribution

of wind directions obtained from the damage survey on December 12. The wind

directions were estimated from the directions to which road signs, poles, trees and

crops on farms were leaning or fell down.

It is seen that the wind directions on the

east side of the dashed lines are S~SE and

those on the west side W~NE. Thus, the

wind direction changes markedly axzross the

dashed lines, which are essentially conver- gence lines. These lines are located in the

northwest side of the heavily damaged area

in Fig. 3. These facts show that the center of the tornado vortex moved along the

dashed lines and the tornado had a cyclonic

rotation in accordance with the analysis of

the video image and wind records.

The tornado moved straightforwardly

north-northeastward from Midori-Machi,

where the first damage was noted, to the

Mobara City Gymnasium. The tornado

path derived from the distributions of wind

directions and damages, however, becomes

obscure near the Gymnasium. The path becomes evident again near the Route 128 in

18
the north of the Gymnasium and extends

north-northeastward. The tornado crossed

the JR Sotobo Line right south of the ShinMobara Station and moved further straightforward to the Kobayashi Area.

It is noted that the tornado path seems

to jump near the Gymnasium in addition to

the eastward protuburence of the damage

distribution. Two possible reasons may be

considered: Firstly, the Gymansium whose

dimensions are 77m in width, 94m in length

and 21m in height, could have blocked the

inflow near the surface to the tornado vortex. This may have resulted in the generation of two vortices both in the east and

northwest sides of the Gymnasium. Secondly, either the intensification of the circulation of the mesocyclone or the increase

of surface roughness may have led to the

change of the structure of the tornado vortex. Laboratory experiments^^ show that

in either case a thin laminar vortex can

be transformed into a thick turbulent vor- tex or multiple vortices. In any case, it is evident that some structural change of the

tornado vortex occured in the middle of its path. Further investigation concerning the cause of the structural change is currently

in progress.

3.4.2.Translational velocity

The principle for estimating the translational velocity of the tornado is remarkably simple: One would only need to know

the times at which two different places first experienced damage due to the tornado. In

practice, however, to obtain accurate estimates for the times is extremely difficult.

In the previous studies, the times have been

estimated within an accuracy of 1 minute

at least. In the present study, we have an- alyzed the time data recorded by a single

computer at a control center of a security

company. The computer records the times

and places at which alarm signals due to

damages to houses or stoppage of power

supply are given. The accuracy of the time

data is 1 second.

In addition to the accurate estimates

described above, another estimate was made.

The latter was based on the eyewitness evidence provided by a habitant and her 'playback' simulation in addition to the time

data of stoppage of electric power supply.

1) Estimation based on the data of the security company

The earliest alarm signal, which notified a stoppage of electric power supply

came from the Hayano Area in the south

part of the Mobara City at 2 seconds before

1913. Three signals which notified direct

damages to buildings and houses came from

the Takashi Area in the center of the City

at 01, 07 and 10 seconds after 1915. The

distances between the point in the Hayano

Area and the three points in the Takashi

Area are about 2000m. Accurate distances

and the time differences give the estimate

for the translational speed of 14.8,16.5 and

17.8m/s whose average is 16.3 m/s.

2) Estimation based on the time of stoppage

of electric power supply and eyewitness ev- idence provided by the habitant

The data of the security company gave

estimates for the translational velocity in

the southern half of the tornado path. Since

electric power supply was stopped when the

tornado hit the center of the City and cut

the high voltage electricity cable(6.6kV),

the time data of the security company were

not available for the northern half of the

path. The time for the stoppage of power

supply was recorded by the Tokyo Electric

Power Company and is between 1915 and

1916.

According to a habitant near the JR

Shin-Mobara Station, severe wind due to

the tornado occured after the stoppage of

electric power supply. She experienced the

stoppage of power supply on the first floor.

>\
She watched outside through windows and

then went upstairs to find a flashlamp. The

severe wind started when she found the

flashlamp. A 'playback' simulation by herself gave a time interval of 90~100 sec- onds between the stoppage of power supply and the severe wind. The distance of

1600m between the high tension cable and

her house together with the time interval

give an estimate for the translational velocity of 16.2'-18.0m/s.

The estimates for the translational velocity derived from two different method

agree reasonably well and it may be con- cluded that the tornado moved north-northeastward at the speed of 16m/s. The translational velocity of the tornado is considered to be closely related to the movement

of the mesocyclone which spawns the tor- nado. The translational velocity of the

mesocyclone obtained by the single Doppler

radar at Meteorological Research Institute

was 14.9m/s^^ which again agrees well with

that of the tornaxio.

3.4.3. Estimation of maximum wind

speed

For the purpose of disaster prevention, it is very important to accumulate

enough data on maxumum winds of tornadoes. Though direct measurements of the

maximum wind due to the Mobara Tornado

do not exist, estimates of maximum wind

were made ba^ed on damages to constructions. It was assumed that the constructions were damaged by the pressure forces

due to winds which were steady and uniform

in the vertical direction. It is noted that the

estimated wind speeds are additions of ro- tational winds and translational velocity of

the tornado vortex. Most of the constructions used for the estimation of the wind

speed are located near the ground. Therefore, it should be remembered that the es- timated wind speeds are affected by local

conditions such as a presence of buildings near the constructions. The estimated wind

speeds as well as the types of the constructions used for the estimations are shown in

Table 1. The locations of the constructions

together with the estimated wind directions are indicated by arrows in Fig. 6. The

largest wind speed was derived from a distortion of a road sign and is 78m/s. Thus,

it can be concluded that the Fujita scale of

the Mobara Tornado is at least F3.

4. SUMMARY

The characteristics of the tornado which

hit the Mobara City, the Chiba Prefecture, on the evening of December 11, 1990, were

studied based on the damage survey and

meteorological records. The following re- sults were obtained:

1) The tornado was generated in the warm

sector of a extratropical cyclone.

2) The tornado produced one of the severest damage in the Japanese record: 1 person died, 73 persons were injured, 82

houses completely destroyed, 161 partially

destroyed and 1504 damaged.

3) The damage path is 6.5km in length and

500~600m in average width. The maximum

width is 1200m.

4) The analysis of the damage distribution

and the tornado path indicates that some

structural change of the tornado vortex oc- cured at the middle of its path.

5) The analysis of the times at which the ini- tial damages ocurred at several buildings or houses gave a translational speed of the tor- nado of 16m/s.

6) The maximum wind speed estimated from

damages to constructions was 78m/s; the

Fujita scale of the Mobara Tornado is at

least F3.

20
ACKNOWLEDGEMENT

The authors express their hearty thanks

to the habitants of the Mobara City who an- swered their questionnaire in the intervals

of rehabilitation work, and to the Prefectural Office of Chiba, the City Office of Mobara, the Observatories of the Japan Meteorological Agency, the Mobara Agricultural High School, the Chosei High School,

The Fujimi Junior High School, the Tokyo

Electric Power Company and the information media who provided with various forms

of data concerning the tornado. They are

also grateful to Messrs. Hisao Ohno, Hiroshi Nircisawa, Osamu Suzuki, Nobuyuki

Kinoshita and Dr. Izuru Takayabu, Meteorological Reserach Institute(MRI), and Mr.

Hisashi Nagamatsu, Building Research In- stitute(BRI), who made the damage survey

in corporation with them. The present investigation wcis made as a part of the Cooperative Project entitled "An Investigation on the Characteristics of the Air Flows Associated with the Tornado in the Mobara

City" between MRI and BRI.

REFERENCES

1) Mitsuta, Y.(Ed.), 1983: Studies on Wind

Disasters Caused by Tatsumaki(Tornadoes

and Waterspout) and Severe Local Storms

in Japan, (in Japanese with English Abstract), Final Report of the Special Research Project for Natural Disaster Sponsored by the Ministry of Education, 124pp.

2) Fergusson, E.W., F.P. Ostby and P.W.

Leftwich, Jr., 1989: The Tornado Season of

1986, Mon. Wea. Rev., 117 , 221-230.

3) Niino, H., O. Suzuki, T. Fujitani, H. Nirasawa, H. Ohno, LTakayabu, N. Kinoshita,

T.Murota and N.Yamaguchi, 1991: An Observational Study of the Mobara Tornado,

Proceedings of the Third Tornado Symposium, Norman, Okalahoma, April 2-5, 1991(

to appear).

4) Niino, H., T. Fujitani, Y. Yamaguchi and

T.Murota,1991: On the Characteristics of

the Mobara Tornado, December 11, 1990.,

(in Japanese), J. Japan Soc. Wind Eng.,

(to be sumitted).

5) Church, C.R., J. T. Snow, G.L.Baker and

E.M.Agee, 1979: Characteristics of Tornadolike Vortices as a Function of Swirl Ratio: A

Laboratory Investigation, J.Aimos. Sci.,

36, 1755-1776. .

21
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22
Fig. l;Locations of cities in the Chiba Prefecture.

Surface map

Fig. 2:Surface weather map at 1800JST on December li, 1990,

23
0 1km

Fig. 3:Daraage distribution in the Mobara City.

24
Fig. 4:Aerial Photograph of the damaged area (the Takashi Area of the

Hobara Ci ty) (Courtesy of Kyodo Tsush i n Company),

25
Pig. 5 : Di 3 tr i bu t ion of estimated wind directions and the tornado path.

26
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27

par t ia 1

1 y destroyed

Hurricane Hugo in Perspective

by

Peter R. Sparks*

ABSTRACT

Wind and surge conditions in

Hurricane Hugo are reviewed and

their relationship to the extent and

nature of damage is discussed. It is

shown that in areas where the gust

wind speed exceeded 50 m/s,

10%-20% of the buildings were

destroyed or seriously damaged. At

40 m/s the figure was about 6%

and at 30 m/s only 1% were

seriously damaged. Surge damage,

although locally severe, only

accounted for about 10% of the

overall damage.

In general damage was related to

the local conditions not to the

relative severity. Thus the wind

damage in Charleston, where the

wind conditions had a recurrence

interval of 50 years, was similar to

that in Sumter where the same

wind speed had a recurrence

interval of more than 200 years. It is suggested that this was due to

the fact that more than 95% of

structural systems and 99% of

building enclosures had been

selected without direct reference

to design wind conditions which

had been based on appropriate

recurrence intervals. For recently

constructed buildings design flood

levels did have a bearing on the

structural form.

Certain types of buildings and

components performed very

poorly, primarily mobile homes,

unreinforced masonry buildings,

older single family dwellings close

to the ocean and some newer ones

with masonry foundations, most

roofing materials and some wall

cladding components.

The severe disruption of the

electricity supply system,

beginning when the wind reached

about 30 m/s, and its effect on the

community is discussed and

contrasted with the much better

performance of the telephone

system.

It is concluded that the extensive

damage resulted in general not

from unprecidented wind and

surge conditions, but from the

failure to use available knowledge

on wind and surge effects in the

selection of building and utility

systems.

KEYWORDS: buildings; damage;

Hurricane Hugo; storm tides;

tropical cyclones; utilities.

1. INTRODUCTION

By the time Hurricane Hugo

crossed the coast of South Carolina

just after midnight on September

22, 1989, it had already caused

over $2 billion worth of damage in

the Caribbean. In its passage

through the states of North and

South Carolina it caused further

damage estimated to be nearly $7

billion, making it the most

expensive storm in the United

States history.

* Department of Civil Engineering,

Clemson University, Clemson, SC

29634-0911

29
At landfall Hugo had a central

pressure of 934mb and an eye

about 50 km in diameter. Winds

circulated in a counterclockwise

direction about the eye, but the

high forward speed produced a

highly asymmetric wind field,

increasing speeds to the right of

the track and reducing them to the

left. The highest winds and worst

storm surge occurred where the

northeastern part of the eye-wall

crossed the coast about 35 km

northeast of the city of Charleston

at Bulls Bay. The high forward

speed (13 m/s-18 m/s) and a track

almost normal to the coast resulted

in very unusual wind conditions in

the interior of South Carolina and

the western part of North Carolina.

2. WIND CONDITIONS

A number of anemometers

measured wind conditions during

the passage of Hugo. Unfortunately

only one of these did so under

standard meteorological conditions

and no anemometers survived in

the area of highest winds.

Figure 1 shows the probable

maximum gust wind speeds at a

height of 10 m in the most exposed

locations, based on corrected

observations in Charleston, Myrtle

Beach, Beaufort, Sumter, Camden,

Florence, Charlotte, and Hickory

plus the use of a theoretical wind

field model.

3. STORM TIDES

Continuous records of water level

were obtained near the Customs

House in Charleston and at Winyah

Bay near Georgetown. Additional

still water levels were obtained by

surveys of tide marks along the

coast.

In the Charleston area tides ranged

from 3 m above mean sea level in

Charleston Harbor to 4.5 m at the

north end of the Isle of Palms. At

the south end of Bulls Bay the tide

reached 6 m and at the north end

of the bay, in the fishing village of

McLellanville, it was 4.2 m to

4.8 m. In the resort areas along

the northern coast, including

Myrtle Beach, tides were 3 m to

4 m above mean sea level. The

normal tidal range is less than 2 m

and much of the occupied land is

only 2.5 m to 4 m above mean sea

level.

In the areas that experienced the

eye of the storm, Charleston, Folly

Beach, Sullivans Island and the Isle

of Palms, the offshore winds before

the eye held the surge back until

the winds dropped as the eye

passed. The water level then rose

rapidly, remained up while the

trailing eyewall passed and then

dropped as the winds subsided.

To the northeast of the eye the

surge rose more steadily, reaching

its peak at about the time of the

highest winds. The surge began to

drop as the winds diminished and

moved around to the south.

To the southwest of the eye the

offshore winds suppressed the

surge and little flooding was

experienced.

4. SEVERITY OF THE STORM

Based on its central pressure, Hugo

was the 10th most intense storm to

strike the United States this

century, being about the same

strength as the 1926 storm

30
in Miami and Hurricane Hazel

which struck the Carolinas in 1954.

The hurricane produced the

highest storm tide recorded on the

Atlantic coast this century, but a

similar tide was recorded at

Savannah Beach near the South

Carolina-Georgia border in 1893

and higher tides might have been

experienced in the Charleston area

in the storm of 1752. However, in

the southern part of Bulls Bay the

tide was a very rare event which

would be expected to be exceeded

on average only once in over 500

years. In McLellanville the mean

recurrence interval was about 300

years and at the northern end of

the Isle of Palms about 200 years.

On Sullivans Island, in Charleston,

and along most of the northern

coast the mean recurrence interval

was approximately 100 years and

about 80 years at Folly Beach.

In the Charleston area the winds

had a mean recurrence interval of

about 50 years, but similar

conditions had not been

experienced since 1911. They had

however also been experienced in

1893, since records began in 1879.

Based on the description of

damage, the wind conditions in

Hugo were almost certainly

exceeded in 1752 and possibly in a

number of other storms in the 300

year history of the city.

In the Bulls Bay area the wind

conditions probably had a mean

recurrence interval of about 100

years, possibly as high as 200 years.

It was about 60 years in

McLellanville and less than 20

years in the Myrtle Beach area. The combination of an intense

storm and high forward motion

produced rare wind conditions in

the interior of South Carolina and

in the Charlotte area of North

Carolina. At Sumter, 150 km from

the coast, the mean recurrence

interval was probably between 200

and 300 years. The very gusty wind

conditions in Charlotte, 300 km

from the point of landfall, may have

had a recurrence interval more

than 100 years.

5. BUILDING PERFORMANCE

Since the early 1960s engineers

have recommended that buildings

and other structures be designed

for conditions with a mean

recurrence interval of 50 to 100

years. Well designed structures

usually incorporate factors of safety,

so serious structural damage should

not be expected in conditions with

a mean recurrence interval less

than 500-1000 years. Such

conditions existed only for surge

action in Bulls Bay and probably did

not exist an)where for wind action.

Nevertheless the damage was

widespread. The extent of damage

appeared to be related to the local

conditions, not the relative severity

of those conditions, nor was

damage confined to buildings

constructed before appropriate

design criteria were available. In

retrospect this was not surprising.

Most of the damage occurred in

South Carolina where the building

stock consists of approximately

75% wood-framed single-family

dwellings, 15% mobile homes, and

about 10% other types of buildings.

These other types, however

comprise about 35% of the value of

construction.

31
Single-family dwellings are usually

constructed on traditional lines,

sometimes influenced by

prescriptive building code

requirements. These requirements

generally do not adequately reflect

the need to resist wind and water

effects.

Since 1976 mobile homes have

been constructed to Federal

Government standards. The wind

loading requirements in these are

much lower than those generally

set for conventional construction

and reflect a very poor

understanding of wind effects on

buildings.

In South Carolina buildings over

450 m2 or three or more stories in

height must be designed by a

registered architect or engineer.

Even these buildings often contain

components selected on the basis

of empirical design procedures and

product data which do not

adequately consider the effects of

wind (or surge).

Under these circumstances it is

unlikely that, of all of the buildings

affected by Hurricane Hugo, more

than 5% had structural systems

designed to resist appropriate wind

and surge loads. Probably less than

1% of the buildings had roofing

and cladding systems specifically

designed for appropriate wind

conditions. Design wind speeds

based on recurrence intervals

therefore had very little bearing on

the wind resistance of most

buildings.

Design flood levels had more of a

bearing on coastal construction

because the availability of insurance

through the government-run

National Flood Insurance Program

was tied to satisfying floor elevation

requirements based on the 100

year flood level. Unfortunately the

flood maps were not very accurate

and buildings were often

constructed without consideration

of the forces generated by the

storm surge and the associated

wind loads.

Table 1 gives the insured losses for

buildings and their contents. The

number of insurance claims for

each type of building is in almost

exactly the same proportion as its

relative contribution to the building

stock and the amount of damage is

in the same proportion as its

relative value. Although the flood

damage was widely reported and

many buildings were destroyed,

flood losses were only about 10% of

the total property losses and

accounted for only 4% of the

insurance claims.

Table 2 gives the Red Cross

estimate of the number of housing

units damaged. This was made

shortly after the storm,

subsequently insurance claims were

made on nearly 3 times as many

properties as were initially

observed to be damaged.

Ultimately claims were made on

30% of buildings in South Carolina

but only 3% were seriously

damaged or destroyed. The vast

majority of damage was due to

relatively minor wind damage

followed by extensive rain damage.

The distribution of damage was

closely related to gust wind speeds,

the proportion of mobile homes,

and, for surge damage, proximity to

the ocean. In some rural areas

where 20%-30% of the population

32
lives in mobile homes, 70%-80% of

the buildings completely destroyed

were of that t3^e.

On the islands near Charleston and

on the shores of Bulls Bay, nearly all

buildings experienced some wind

or surge damage and in some

coastal communities nearly 50% of

the ocean-front buildings were

completely destroyed.

In counties where the wind gusted

to more than 50 m/s, at least 80%

of the buildings experienced some

damage and 10%-20% were

destroyed or experienced extensive

damage. Damage rates diminished

as the wind speed dropped. At

40 m/s about 6% of buildings were

destroyed or seriously damaged. At

30 m/s less than 1% received

significant damage.

Certain types of buildings and some

building components performed

poorly. As mentioned earlier

mobile homes had a high damage

rate, but so did single-story

unreinforced-masonry buildings,

older pre-engineered metal

buildings, older single-family

dwellings close to the ocean and

some newer ones using masonry

foundations, mostly roofing

materials and some wall cladding

elements. The poor resistance of

many of these buildings could be

traced to poor building code

requirements and inappropriate

design and construction practices.

6. PERFORMANCE OF UTILITIES

The electric utility companies had

constructed transmission and

distribution systems v/hich were

very vulnerable to wind damage.

The decision to use wood poles for

high voltage transmission systems

and overhead local distribution

systems was probably made on

economic grounds. Although

interruption of power supply is

common in the region in severe

thunderstorms and had been

experienced in earlier hurricanes,

the extreme vulnerability of the

system had not been appreciated by

the utility companies nor their

customers. It had certainly not

been appreciated that loss of

electric power for an extended

period of time would have such an

effect on the basic essentials of

life - the ability to obtain food and

water and the provision of

wastewater facilities.

Approximately 1.5 million people,

living in areas where the wind

speed exceeded 30 m/s, lost

electric power. The utility

companies made tremendous

efforts to restore power, but the

extent of the damage was so great

that in many cases the system

could not be repaired, but had to be

rebuilt, usually to the previously

existing standards. Eight days after

the storm only 23% of the

customers in the Charleston area

had power and in some rural areas

it was two to three weeks before

power was restored.

The telephone companies had

attempted to reduce the

vulnerability of their systems by

burying cables. In most areas there

was very little interruption of

telephone service. Other forms of

communication, including

television and radio broadcasts and

even two way radio

communications of emergency

services, were seriously affected by

33
loss of electric power and the

collapse of transmission towers.

7. CONCLUSIONS AND LESSONS

LEARNED

Hurricane Hugo taught South

Carolina and the nation some very

expensive lessons. The potential

for disaster had been recognized

but the political and economic

systems made it difficult to remedy

the situation. It is worth examining

what went wrong.

The state had chosen a laissez-faire

system of building control. The

power to adopt and the

responsibility for enforcement of

building codes had been delegated

to local jurisdictions. This resulted

in considerable variation in the

quality of control, from some of the

best in the country to no control at

all. The model building code adopted

showed a very poor understanding

of the wind and surge effects on

structures and permitted forms of

construction in wood and masonry

which were inconsistent with the

loading requirements. Lacking

proper deemed- to- comply

documents, the performance

requirements of the code were

unenforceable for non-engineered

structures. Using the permitted

forms of construction in the code

led to structural systems unsuitable

for a hurricane, tornado, and

earthquake prone area such as

South Carolina.

In poor rural areas mobile homes

were the only affordable form of

housing for many people. These

homes were built to inadequate

standards and when delivered to

site were often not properly

anchored.

Despite the fact that cladding

failures can pose a threat to public

safety and even compromise the

structural integrity of a building,

cladding systems had been

specified by those without adequate

training in structural design. The

professional licensing laws had

permitted this.

In recent years owners had often

been unwilling to pay for adequate

professional inspection during

construction. The quality of

construction had suffered.

Compounded by poor building code

requirements, the use of

inappropriate design professionals

and inadequate material testing

standards, this had led to cladding

systems and, in some cases,

structural systems with very low

wind resistance.

The insurance industry had been

willing to insure property with

little regard for its wind resistance.

Premiums were often based on the

building's fire resistance. Thus

there was little or no incentive for

the owner to improve wind

resistance in order to meet

insurance requirements.

The electric utility companies had

been permitted to construct

transmission and distribution

systems which were very vulnerable

to damage by a hurricane. The

public was unaware of how fragile

the system was or how dependent

they were on a continuous supply of

electricity.

Finally, there was the lack of

experience. For 30 years prior to

34
Hugo, during which time the

hurricane-prone areas of the state

expanded rapidly, no major

hurricanes hit the South Carolina

coast.

From an engineering standpoint

the dominating lesson is clear and

the solution is straightforward.

Design and construction practices

must reflect the loads which are

known to occur in conditions with

reasonable recurrence intervals.

The prospects for achieving this

and correcting past mistakes are

not good.

The building stock changes slowly

and the fragmented building

control system in the United States

is ill-equipped to impose improved

construction standards. Based on

the success of the National Flood

Insurance Program, the best

prospect for improved construction

is by insurance industry action.

That industry certainly learned its

lesson in Hugo when it paid out the

equivalent of all property insurance

premiums collected in South

Carolina in the last 5 years.

Not all lessons learned were

concerned with what went wrong.

Properly designed buildings

performed well and the telephone

service was a fine example of what

can be achieved with good

planning. Perhaps the most

valuable lesson learned was that

modern technology can be used

very effectively to monitor storms

and warn coastal residents of

impending danger.

In 1893 a storm of similar strength

to Hugo drowned 1000-2000

people in South Carolina. Prior to

Hugo's landfall over 250.000

people were evacuated from the

coastal area. Only a handful of

people were killed, none by

drowning in their homes. It is a

poor reflection on the construction

industry that many evacuees

returned to buildings that were in

considerably worse condition than

their counterparts after the 1893

storm.

8. ACKNOWLEDGEMENTS

This study of the effects of

Hurricane Hugo was made possible

by grants from the National Science

Foundation, South Carolina Sea

Grant Consortium, South

Commission on Higher Education,

the Federal Emergency

Management Agency and Clemson

University. This support is

gratefully acknowledged.

35
Table 1 . Insured Damage to Buildings and Contents in South Carolina

Number of Claims Total Amount

($ million)

Wind

Conventional dwellings 287456 (74%) 1423 (55%)

Mobile homes 57808 (15%) 172 (7%)

Commercial property 41282 (11%) 997 (38%)

Total wind damage 386546 2592

Flood

All types 15739 320

Table 2. Number of Dwelling Units Destroyed or Damaged in South Carolina

Destroyed Major Minor * Total

Damage Damage

Single family dwellings 3783 18146 57698 79627

Apartments 313 2647 8948 11908

Mobile homes 5200 5976 9063 20239

Total 9296 26769 75709 111774

Externally observable damage only, excludes damage to outbuildings,

fences, decks, porches etc.

36
Figure 1. Probable Maximum Gust Wind Speeds (m/s)

37

Theme II

Earthquake Engineering

The Worldwide Earthquake Risk Management (WWERM) Program

by

S.T. Algermissen^ Paul R. Krumpe^

and Walter W. Hays-*

ABSTRACT strategies.

The Worldwide Earthquake Risk Management

(WWERM) Program was initiated as a cooperative effort of the Agency for

International Development (AID), U.S. Office

of Foreign Disaster Assistance (OFDA) , and

the U.S. Geological Survey in late 1989.

The initial objective of the program is to

develop an action plan and conduct a twoyear pilot program to demonstrate the

utility, importance, feasibility, and need

for uniform global earthquake hazard mapping

and risk assessment/management. Three areas

have been selected for the pilot studies;

Chile, Indonesia, and Morocco. The WWERM

program is a project of the International

Decade for Natural Disaster Reduction

(IDNDR)

KEYWORDS: Earthquake; hazard;

international; risk

1. INTRODUCTION

The Worldwide Earthquake Risk Management

(WWERM) Program was developed as an outgrowth of an Executive Briefing on "Strategic Planning to Reduce Economic

Impacts of Earthquake Hazards throughout the

World" held at the National Academy of

Sciences, Washington, D.C., on March 8 and

9, 1988 (Williams and Hays, 1988) . This

Executive Briefing established the need for

a more quantitative evaluation of the

potential for earthquake disasters and for

the assessment of the economic impact and

casualty loss potential of such disasters.

In addition, the need to assess the longrange economic losses associated with

destructive earthquakes was also outlined.

2. OBJECTIVES OF THE PROGRAM

The initial objectives of the program have

been to develop an action plan and conduct

a two-year pilot program in each of three

geographic regions to demonstrate the

utility, importance, feasibility, and need

for uniform global earthquake hazard mapping

and risk assessment/management, in

preparation for and as a program linked to

the Decade for Natural Disaster Reduction.

The three geographic regions are Morocco,

Chile, and Indonesia. The principal

objective in these three areas is to

demonstrate the feasibility and utility of

defining the earthquake hazard through the

development of probabilistic ground-motion

maps and the estimation of future earthquake

economic losses in selected pilot studies.

The demonstration of the feasibility of

these types of studies will lead to a broader program that will form the basis for

risk management, lifesaving preparedness

planning, and the development of mitigation

Additional goals are to develop a five-year

implementation plan with OFDA and to obtain

financial commitments (co-funding) from

USAID missions, host governments,

international organizations, and the private

sector. These kinds of commitments will

support the worldwide scope of the WWERM

program and allow it to be fully integrated

into and be a major contribution to the

International Decade of Natural Hazard

Reduction

3. RATIONALE FOR THE PROGRAM

Earthquake risk management (the reduction of

both the economic and life loss resulting

from earthquakes) requires the following:

(1) The quantitative assessment of the

hazard; and (2) the evaluation of the risk

(loss) in terms of the expected hazard. No

efficient plan for earthquake risk

management is possible unless it is based on

an assessment of both the spatial and

temporal nature and magnitude of the hazard

and risk. As an example, large earthquake

ground motion in the Atacama Desert of

Northern Chile represents a real and

considerable earthquake hazard but

negligible risk since structures and

population at risk are very minimal.

Conversely, moderate earthquake ground

motion in Santiago, the capital, would

result in significant economic loss and

casualties

The examples cited are obvious, but on a global scale the evaluation of expected

earthquake losses is a difficult scientific

and technical problem of considerable

subtlety. The evaluation of the magnitude

and spatial distribution of expected losses

is, however, an essential prelude to successful risk management because it: (1) allows the always inadequate resources for

disaster mitigation to be applied to areas

of catastrophe potential on a rational

basis; and (2) identifies areas of

catastrophe potential that may not be easily

recognized because of the combination of

unusual ground motion (and related

geological hazards, such as landsliding and

liquefaction) with large exposures of poorly

reinforced structures with high occupancy.

^U.S. Geological Survey, Denver Federal

Center, Box 25046, MS 966, Denver, CO 80225

^U.S. Office of Foreign Disaster Assistance,

2201 C Street, N.W., Room 1261A, Washington,

D.C. 20523

^U.S. Geological Survey, National Center,

MS 905, 12201 Sunrise Valley Drive, Reston,

VA 22092

41
Obvious examples are the Mexico City

disaster of 1985 and the 1970 earthquake in

Peru (which killed approximately 60, 000

people) . This project represents a pioneering

international effort to quantitatively

evaluate earthquake hazard and risk and,

therefore, to provide an improved technical

base for risk management.

4. PROGRESS

4.1 Indonesia

Visits were made to Jakarta and Bandung in

June 1990 by S. T. Algermissen (Project

Chief), Paul C. Thenhaus (Geologist), and

E. V. Leyendecker (Engineer). Through the

interest and encouragement of OFDA and AIDIndonesia, the visit and meetings culminated

in a project proposal to AID-Indonesia

outlining a project design for earthquake

risk management training in Indonesia.

Recommendations in this proposal are for a collaborative hazard and risk assessment of

the Gorontalo area in Sulawesi Province,

Indonesia, involving two future training

trips by USGS staff followed by a technical

workshop upon project completion that would

document hazard evaluation methods,

products, and uses in mitigating the

earthquake hazard and risk. Post-workshop

project evaluation is also included.

Additional funding for the expanded program

in Indonesia was supplied both by OFDA and

AID-Indonesia

.

The area selected for the case history is

the City of Gorontalo in Sulawesi. The city

of Gorontalo and the surrounding area in

Sulawesi have several advantages as a pilot

study area in Indonesia: (1) The area is

highly seismic (Figure 1) ; (2) a damaging

earthquake occurred near there on April 18,

1990 and damage data are available for the

earthquake; and (3) data are available in

the Gorontalo area. The surficial geology is an important factor in altering the levels

of ground motion and, consequently, an important element in estimating future

damage

.

Earthquake damage and geological data for

Gorontalo and the surrounding area will be

obtained from scientists at the Geological

Research and Development Center at Bandung.

These damage data will aid USGS in

developing vulnerability (loss to structures

as a function of ground shaking)

relationships for Gorontalo. The Geological

Survey is developing a probability ground

shaking and earthquake loss model for the case history area.

In summary, sufficient data are available

for the Gorontalo area such that it provides

an excellent location for the demonstration

of seismic hazard and risk (loss)

assessment. The steps in the hazard and

risk study are:

(1) Development of a probabilistic model

for the computation of the expected

earthquake ground motion in Sulawesi

(specifically the Gorontalo area) . This

model is developed from an analysis of the

earthquake catalog, seismotectonic maps and

seismic wave attenuation data developed

under the OFDA-USGS-SEASEE program. The

effects of local surficial geology on ground

shaking in the Gorontalo area will also be

incorporated into the model.

(2) Compute the expected ground motion in

the study area for periods of time of

interest of, for example, 10, 50, and 250

years. For example, the expected ground

motion in a 50-year period is used in

seismic design provisions of model building

codes in the United States. These maps are routinely prepared by the U.S. Geological

Survey (see, for example, Algermissen and

others, 1990) .

(3) Use the ground-motion maps in (2) above

to calculate the expected future earthquake

losses in the Gorontalo area for various

time period of interest.

4.2 Chile

A visit to Santiago, Chile, was made by

S. T. Algermissen in late June 1990 to discuss the WWERM Project with possible

Chilean collaborators. The development of

suitable earthquake hazard evaluations

(probabilistic ground-motion maps) is

somewhat eased in Chile because a concurrent

project is underway in Chile (an AID-OFDAUSGS cooperative project) to provide a hazard evaluation. Thus, the emphasis in

discussions in Chile has been on the

selection of a suitable area for risk

(earthquake loss) assessment.

Considerable data on earthquake losses and

vulnerability exist in central Chile as a result of the studies of Dr. Joaquin Monge

of the University of Chile, of the damage

resulting from the large, M^=l .8 central

Chile earthquake of 1985. The area chosen

for the pilot study of hazard and risk

assessment in Chile is the community of

Maipu (Figure 2) . In the 19th century,

Maipu was an agricultural community, but it

is now largely a residential area in the

southwestern part of the greater Santiago

metropolitan complex. Nearly all of the

information necessary for a seismic risk

study is already available: (1) An accurate

population count; (2) the spatial

distribution of structures with regard to construction type; and (3) data on damage

from past earthquakes. Considerable

additional information is available on damage resulting from the 1985 earthquake,

nature of the surficial geologic materials,

etc. An example of the housing data

available for Maipu is shown in Table 1.

A technical workshop on the project, with

co-sponsorship by the United Nations

Disaster Relief Organization (UNDRO) is

planned for Santiago, Chile, in October

1991

.

4 . 3 Morocco

Morocco is a highly seismic country that has

experienced many damaging earthquakes

(Figure 3) . A cooperative program of

42
seismic hazard and risk assessment is being

planned through Dr. Oriss Ben Sari, Director

of the Center for Technical and Scientific

Research (CTSR) in Rabat. CTSR will supply

data on the distribution of population,

classes of buildings, and the nature of the

surficial geology in the study area. CTSR

already has an active seismological network

research program and a 22-station

seismological network. They have nearly

completed a seismotectonic map of the entire

country.

5 , SUMMARY

The preliminary phase of the WWERM program

is proceeding on schedule. It is hoped that

the scope of the project can be expanded

sufficiently to provide meaningful

comparative measures worldwide, such that

these comparisons provide an improved

quantitative basis for earthquake disaster

management and contribute substantially to the International Decade for Natural

Disaster Reduction.

6. REFERENCES

(1) Algermissen, S.T., Perkins, D.M.,

Thenhaus, P.C., Hanson, S.L., and Bender,

B.L., 1990, Probabilistic Earthquake

Acceleration and Velocity Maps for the

United States and Puerto Rico, U.S.

Geological Survey Misc. Field Studies Map,

MF-2120.

(2) Williams, Mary E. and Hays, Walter W. (1988) Strategic Planning to Reduce Economic

Impacts of Earthquake Hazards throughout the

World, Proceedings of the Executive

Briefing, National Academy of Sciences,

Washington, D.C., March 8-9, 1988, U.S.

Geological Survey, Open-File Report 88-361,

149 pps and Appendices.
90° 95° lOo' 105° ' 110° 115° 120° 125° 130° 135* Mo" M5°e

Figure 1. Seismicity of Indonesia showing the major shallow earthquakes

1897-1984 and the location of the selected study area. Note the

high rate of seismicity in the Gorontalo area.

Figure 2. Location map showing the project area in Chile (Maipu)

44
TlPO DE COMUNA

VIVIENDA MAIPU

URBANO RURAL TOTAL URBANO (TOTAL)

rarilCUIai. 24 .042 1.291 25 .333 13.822

casa 20 .693 1.219 21 .912 10.095

departamento 1 343 1 .343 1.296

mejora,emergencia 1 882 1 .882 2.056

viv. u8 conveniiiio 103 - 1 03 332

callampa 6 - 6 1 8 rancho,ruca o choza 6 3 63 " mdvil -

1 1 6 otro tipo 1 5 a 2 3 1 9

1 6

Colectiva: 25 4 29

residencial, pensi6n

hotel, motel, posada ° instituci6n 1 5 3 1 8 1 4 otro tipo 1 0 1 1 1 2 TOTAL 24 067 1 .295 25 362 13.838

Table 1. Distribution of buildings by construction type in the

communities of Maipu and Cerrillos in the metropolitan

area of Santiago, Chile
j
Influence of Embankments on the Liquefaction of Ground

BY

Yasuyiiki Koga' , Jun-ichi Koseki^, Takao Shijnazu'^ and Osamu Matsuo^

ABSTRACT

Of the influence of the embankments on the

liquefaction of sand deposit the characteristics

of the excess pore water pressure are examined

through the results of shaking table tests of

I model embankment on a liquefiable ground.

Emphasis is laid on the maximum pore pressure and

i the relationship between liquefaction resistance

factor and pore pressure generation in the free

field and below the embankment.

! KEY WORDS tLiquefaction, Embankment,

Model shaking table test

1. INTRODUCTION

It has been well known since the Niigata

earthquake that the liquefaction of ground might

cause large damages to many kinds of structures,

j As a consequence of many studies on liquefaction

\ several methodologies to assess the liquefaction

I potential have been developed and some of them

have already been adopted in technical codes.

However, most of them are only applicable to predict the behaviour of level free field

ignoring the effect of structures to the ground.

And the effect of the structure on the

liquefaction of ground is not yet clarified.

We sometimes confront a situation to evaluate the stability of the embankment on liquefiable ground during earthquakes. In that

case the liquefaction potential or the excess

pore water pressures of the ground near

embankments are needed for the stability

analyses. The aim of this paper is to distinguish their characteristics in the free

field and the ground below the embankment based

on model tests.

2. REUTED PREVIOUS STUDIES

2.1 Lique f action characteristics of the level

ground

There have already been substantial

knowledge about liquefaction of level ground. In order to represent the stability of the ground

against liquefaction, the liquefaction resistance

factor Fl. is defined as below(l).

in which R is the dynamic shear resistance

ratio, L is the shear stress ratio during

earthquakes.

Seed et al. formulates the excess pore

pressure in the ground (see Fig.l)(2):

2 . _J N

\^/^"

* = sin M— (2)

in which Uo = excess pore pressure generated by

cyclic loading, a'© = initial effective

confining stress, Nu = number of cycles required

to cause liquefaction under uniform stress conditions, N = number of loading cycles, a

= experimental constant.

In Japan the excess pore pressure due to earthquakes is often formulated by use of Fl in

Eq.(l) based on the following assumption.

Assumption: the dynamic shear resistance ratio

R and the number of loading cycles N are in a linear relationship when both are represented in

logarithmic scale, that is,

R = a N ^ (3)

in which a, b = experimental constant.

If the shear stress ratio during earthquakes

L , whose number of cycles to cause liquefaction

is Nl, , is applied by N cycles(Fig.2) , it

follows from Eq.(3)

L = aNL'' (4)

Combining Eqs.(3) and (4) leads to

'Head, Soil Dynamics Division, Construction

Equipment and Method Department, Public Works

Research Institute, Ministry of Construction,

Japan.

^Researcher, ditto.

^Assistant, ditto.

^Assistant Director, Economic Cooperation Bureau,

Ministry of Foreign Affairs, Japan. Formerly,

Senior Researcher, Soil Dynamics Division,

Construction Equipment and Method Department,

Public Works Research Institute, Ministry of

Construction, Japan.

47
This equation is rewritten as

Substituting this into Eq.(2) leads to the

following relationship between the excess pore

pressure and the liquefaction resistance factor

~ = F^'--*^ (5)

a o IN 1-

Take the technical guideline for designing

of common utility duct (3) for instance, the

equation

= F:.-^ (6) a o

is proposed.

2.2 Evaluation method of the influence of an embankment on the ground liquefaction and the

stability of the embankment

The effects of the embankments on liquefiable grounds are; 1) increase of the

overburden pressure, 2) increase of the irertia

force, 3) additional initial shear stress,

4) spatial difference of lateral confining

condition.

Stability analysis methods of the

embankments on liquefiable grounds during

earthquakes are; 1) slip surface analysis method,

2) local safety factor method through FEM, 3)

permanent deformation analysis method through

FEM. Of these the method 1) is mainly adopted in

technical codes in Japan. The resistance of the

ground for the stability analysis might be

expressed either in terms of total stress on the

basis of accumulated strain due to cyclic loading

or in terms of effective stress through excess pore pressure caused by cyclic loading. The

latter formulation is frequently used in

Japanese technical codes(4).

Fig. 3 shows a typical procedure in Japan for

the stability analysis of the embankments on liquefiable grounds.

The equation below is often used for the

stability analysis of an embankment shown in

Fig. 4(4):

_ £ {c' -l+CW-Uo-b-Ue-b) -cosg •tan<^ '}

£ (W-sina+kh-W-h/R) ^ '

in which c', .^ ' = cohesion and shear resistance

angle of soil,

W = total weight of a slice,

1 = length of the arc of a slice bottom,

b = width of a slice.

Uo = pore pressure in ordinary condition,

Ue = excess pore presure caused by

earthquake loading,

kh = design seismic coefficient,

R = radius of a slip circle,

h = height of the center of slip circle

from the center of gravity of a slice

(see Fig. 5)

The excess pore pressure Ue is calculated

from Eq.(5). That is, a horizontal overburden is

assumed for the soil below the embankment

ignoring its shape effect, then the liquefaction

resistance factor F^ is calculated to estimate

the excess pore pressure in the ground.

2.3 Liquefaction potential from cyclic loading

tests in the laboratory

The characteristics of sands have been

examined by cyclic triaxial tests or cyclic

torsion shear tests. The liquefaction

characteristics of level ground has been

simulated by isotropically consolidated undrained

cyclic loading of triaxial specimens since the

beginning of the study of liquefaction(5) . When

cyclic torsional shear tests are conducted,

isotropically or anisotropically consolidated

specimens are applied undrained cyclic torsional

loading under the condition of confined axial

displacement (6) . When the ground near structures

or the soil in a slope is concerned, cyclic axial

loading is given to anisotropically consolidated

triaxial specimens (7,8) . When cyclic torsional

shear tests are conducted, undrained cyclic

torsional shear is applied after some initial

torsional shear is given to isotropically or anisotropically consolidated specimens. In one case the axial displacement is confined during

cyclic loading (9) while in another the axial

displacement is not confined with constant

confining stresses ( 10) . Such different loading

methods and displacement confining conditions

give different progressive deformation and pore

pressure generation of soils. As one of the

examples, Finn et al. formulated the pore

pressure generation under undrained cyclic

loading of anisotropically consolidated triaxial

specimens as follows(Fig.6) (8)

:

1/ a =^ 4—L- .sin-'ff^] -ll

(8)

in which Ucy = pore pressure, a 30 = lateral

confining stress, N = number of cycles, Nao = number of cycles to develope a pore pressure

equal to 50% of the confining stress 030. a

= parameter depending on the consolidation ratio

Ko~ a 10 / <J 3o •

48
2.4 Liquefaction characteristics in the ground near a structure from model shaking table tests

Yoshirai and Tokimatsu performed shaking

table tests of rigid structures on sand layers

and observed the change of pore pressure near a structure and settlement behaviour(Fig.7) (11)

.

The writers conducted shaking table tests of

embankments on liquefiable grounds, which

exhibited different behaviour of pore pressures

in the free field and below the embankment

(Figs.8,9)(12).

3. EXPERIMENTAL INVESTIGATION

3 . 1 Outline

The difference of the pore pressure in the

free field and below an embankment was precisely

examined from model shaking table test results.

Investigated were shaking table tests on the

model shown in Fig. 10. The spatial distribution

of pore pressure was first examined and then the

acceleration and pore pressure time history were

analyzed on the basis of the cumulative damage

theory to obtain the relationships between the

liquefaction resistance factor and the pore

pressure ratio ru.

3.2 Outline of model tests Three test cases are analyzed in this paper.

Model embankments were placed on saturated sand

layers as shown in Fig. 10. The relative densities

of the ground were 73-79%. Sinusoidal excitation

with different frequencies were applied as shown

in Table 1. In respective test cases the

magnitude of the excitation was increased

stepwise, however, only the first cases that the

liquefaction of the ground first occurred were

analyzed.

3.3 Method of data arrangement and calculation

Fig. 11 shows the procedure of the data

arrangement and calculation.

1) Calculation of pore pressure ratio

The time history of pore pressure ratio is

calculated from time history of the pore pressure

by dividing by the effective overburden pressure

ovo'» The pressure ovo' can be calculated from

Eq.(9) one dimensionally by use of the unit

weight of the model and the depth of pore

pressure gauges. However, as the installed depth

of the gauges might have some error it was

estimated from the record of pore pressure when

liquefaction surely occurred as shown in Fig. 12.

avo' = y ' • z + • H (9)

in which y ' = effective unit weight, z = depth

from the ground surface, = total unit weight,

H = height of embankment.

2) Arrangement of the distribution of excess pore

pressure ratio r u The distribution of the pore pressure ratio ru was arranged at every specified time.

3) Calculation of dynamic shear stress ratio

T d/ a vo

The dynamic shear stress ratio ra/avo' was calculated from Eq. (10) (see Fig. 13).

<J vo a vo I g in which ovo' = initial effective overburden

pressure, Ai = response acceleration at the i-th

layer, hi = thickness of the i-th layer, y i = unit weight of the i-th layer, g = acceleration

of gravity.

4) Calculation of the number of cycles required

to cause liquefaction

The number of cycles when the pore pressure

ratio r u = 1 is counted for every pore pressure

gauge in the free field.

5) Assumption of the liquefaction resistance

curve

The shape of the liquefaction resistance

curve, which is required to apply the cumulative

damage theory, was assumed to be given by

Eq.(ll). This curve was determined referring to the liquefaction resistance curve of Toyoura sand

obtained from cyclic torsion shear tests as shown

in Fig. 14.

T a/ an.o' ^ 0.3 ; Ta/a^o' - 0.3 = -1.25 log (Nc/N')

r a / o mo * < 0.3;

Ta/c„o' - 0.3 = -0.11 log (Nc/N') (11)

Eq.(ll) can be written as

Nc/N' = f ( T a/omo') (12)

in which f = equation to give the relation

between cyclic stress ratio and normalized number

of cycles, Nc = number of cycles to cause liquefaction when uniform cyclic stress ratio

Ta/omo' is applied, N' = parameter to locate

the liquefaction resistance curve, a mo' = effective mean stress, which is calclated

assuming Ko = 0.5.

6) Evaluation of liquefaction resistance curve

The parameter N' was evaluated from the

following procedure.

In order to apply the cumulative damage theory

to an irregular train of shear stress ratio, it

is divided into every half cycle then the

cumulative damage up to the i-th half cycle is

given by the equation below,

49
D = i; 0.5

(13) i(Nc)i

in which (Nc)i = number of cycles to cause liquefaction when the i-th half cycle stress

ratio (rd/amo')i is repeatedly applied.

Substituting Eq.(12) into Eq.(13) leads to the

equation below.

D =

0.5

N' 1 f{(Td/omo')l} (14)

A value of D >1.0 indicates a liquefied

condition and D<1.0 a non-liquefied condition,

therefore following equations are obtained for

respective conditions.

Liquefaction condition:

0.5 N ' ^ 2 (15)

(16)

I f{(td/amo')l}

Non-liquefaction condition:

N'> 2:777-^4 m

Substituting the shear stress ratios up to the

cycles to liquefaction, which was obtained in 4),

into Eq.(i5) and the shear stress ratios up to one half cycle before the above cycles to liquefaction into Eq.(16), we can estimate the

possible range of the parameter N ' which locates

the liquefaction resistance curve. A possible

range of N' is estimated by conducting the same procedures for every pore pressure gauge, the

average N' of the ground can be evaluated.

7) Calculation of equivalent cyclic shear stress ratio L

When the parameter N' to locate the

liquefaction resistance curve is obtained, the

cumulative damage D up to the i-th half cycles

is computed from the above Eq.(14). Then the

equivalent cyclic shear stress ratio L is

determined as the amplitude of uniform cyclic

shear stress ratios of 20 cycles to give the same cumulative damage.

The relationship between the cumulative damage

D due to the shear stress ratios up to the i-th

half cycles and the number of cycles to cause liquefaction is given by

''^ 0.5 D = 2-^^ =20/N,

Thus

Ni. = 20/D

(17)

(18)

The equivalent shear stress ratio L

corresponding to is computed by the following

equations obtained from Eq.(ll)

Nl^n' ; L=0.3-1.25 (logN^-logN')

Nl.>N' ; L=0.3-0.11 (logNL.-logN') (19)

8) Evaluation of dynamic shear resistance

ratio R

Substituting = 20 into Eq.(19), the

dynamic shear resistance ratio R to cause liquefaction in 20 cycles is computed.

9) Computation of liquefaction resistance

factor Fu

The liquefaction resistance factor Fl is

computed for every half cycle using the

equivalent shear stress ratio L derived in 7)

and the dynamic shear resistance R derived in

8) by

F,. = R

(20)

Note that R and L are both determined for

uniform 20 cycles.

10) Arrangement of the relationship of

liquefaction resistance factor Fl and excess

pore pressure ratio r u The relationship between the liquefaction

resistance factor F^ and the excess pore

pressure ratio r ^ is arranged for every pore

pressure data.

3.4 Results and discussion

Fig. 15 shows examples of time histories of the

dynamic shear stress ratio td/ovo' , the

equivalent shear stress ratio L , the

liquefaction resistance factor F^ , the pore

pressure ratio r u of the variables described

above at 4 typical locations. The parameter N'

was assumed 18 because it was estimated about 18 for all cases from Eqs.(15) (16). The reference

liquefaction curve is determined by substituting

N' =18 into Eq.(ll). Fig. 16 shows the

distribution of excess pore pressure ratio r u at the 5-th cycle of excitation and right after the

excitation. It is shown in all the cases that the excess pore pressure ratio r u builds up faster

in the free field than below the embankment

although the shape of distribution is a little

different respectively. The liquefaction

resistance factor Fl, is larger below the

embankment than in the free field because of the

overburden effect of embankment. It is also shown

that while the excess pore pressure ratio r u in

the free field builds up to 1, the excess pore

pressure does not below the embankment. The

50
buildup is smaller below the crest than below the

slope and the increase of pore pressure ratio r u is smaller in the shallower portion of the ground

than in the deeper portion. This can be

attributed to the different lateral confining

conditions in the ground due to the embankment

that a complete liquefaction does not occur below

the embankment while it does in the free field.

Fig. 17 shows the relationship between the

liquefaction resistance factor Fl and the excess pore pressure ratio r u • Such relationships are obtained by combining the time histories of r u and Fl. in Fig. 15. Figs. 17 a)d)g) shows the

result in the free field, Figs. 17 b)e)h) that

below the crest and Figs. 17 c)f)i) that below the

slope toe. The following can be seen from these

figures.

i| l)The pore pressure ratios r u are distributed

in the range from F^"^ to F^""^. The influence

of the installed depth of pore pressure gauge

could not be seen.

1 2) Below the crest of the embankment the pore

I pressure ratio showed a smaller tendency than

, those in the free field against the same Fl.. The

I pore pressure ratio did not reach unity even when

Fl becomes below 1. Regarding to the effect of

the installed depth of gauges, the pore pressure

j ratio exhibits lower value at the deeper

I location.

I 3)The pore pressure ratio below the slope toe was

less than that in the free field against the same Fu. However, it exceeded one when F^ reached

' one. This can be reduced to that the overburden

j pressure was calculated one-dimensionally

.

, 4) The effect of the excitation frequency on the

relation between Fl and ru was small in the

free field and below the embankment.

j 4. CONCLUSION

1) The liquefaction resistance factor F^ is

j larger and the increase of the pore pressure is

j slower below the embankment than in the free

I field because of the overburden effect of

embankment

.

2) The pore pressure ratio below the embankment

I does not develope up to 1 while that in the free

field can develope up to 1. 3) The pore pressure ratio below the embankment is

|; smaller than that in the free field when compared

jj with the same F^. Below the embankment the pore

j pressure ratio is smaller than 1 even when F^ is

less than 1. 4) When the stability of an embankment on a

' liquefiable deposit is considered the stability

factor is estimated genarally smaller if the pore

pressure below the embankment is computed only by

use of the relationship between Fu and for

the free field.

REFARENCES

1. Tatsuoka,F. et al.(1980): Standard Penetration

Tests and Soil Liquefaction, Soils & Foundations, Vol.20, No. 4, pp. 95-111.

2. Seed,H.B. et al.(1976): Pore-Pressure Changes

During Soil Liquefaction, Jour. of Geotech.

Eng. Div., ASCE, Vol.102, No.GT4, pp. 323-346.

3. Japan Road Association( 1986) : Technical

Guideline for Designing of Common Utility

Duct (in Japanese), pp. 65-66.

4. Japan Road Association( 1986) : Technical

Guideline for Highway Earthwork —

Countermeasures for Soft Ground(in Japanese),

pp. 81-88, pp. 280-291.

5. Seed,H.B. and Lee, K.L. (1966) : Liquefaction of

Saturated Sands During Cyclic Loading, Jour. of

Soil Mech. and Foundations Div., Proc. of

ASCE, Vol.92, N0.SM6, pp. 105-134.

6. Ishihara,K. and Yasuda,S. (1975) : Sand

Liquefaction of Sand Deposits During

Earthquake, Soils & Foundations, Vol.17, No. 3, pp. 1-17.

7. Lee, K.L. and Seed, H. B. ( 1967) : Dynamic Strength

of Anisotropiccally Consolidated Sand, Jour,

of Soil Mech. and Foundation Div., ASCE,

Vol.93, No. 5, pp. 169-190.

8. Finn,W.D.L. et al.(1978): Cyclic Pore

Pressures under Anisotropic Conditions,

Earthquake Engineering and Soil Dynamics,

Proc. of ASCE Geotech. Eng. Div. Specialty

Conf., Vol.1, pp. 457-470.

9. Tatsuoka,F. et al.(1982): Cyclic Undrained

Stress-Strain Behavior of Dense Sands by

Torsional Simple Shear Test, Soils & Foundations, Vol.22, No. 2, pp. 55-70.

10. Watanabe,N. et al.(1991): Permanent Deformation Characteristics under Static and Cyclic

Shear stress Loading(in Japanese), 26th Annual

Meeting of Japanese Society of Soil Mech. & Foundations Eng.

11. Yoshimi,Y. and Tokimatsu,K. (1977) : Settlement

of Buildings on Saturated Sand During

Earthquakes, Soils & Foundations, Vol.17,

No.l, pp. 24-38.

12. Koga,Y. and Matsuo.O. (1990) : Shaking Table

Tests of Embankments Resting on Liquefiable

Ground, Soils & Foundations, Vol.30, No. 4, pp. 162-174.

13.Shiraazu,T. et al.(1991): Characteristics of Pore Water Pressure in the Foundation of

Embankment during Earthquakes (in Japanese),

26th Annual Meeting of Japanese Society of

Soil Mech. & Foundations Eng.

51
Table 1 Analyzed Test Cases

Case

Excitation Condition Excitation

Wave form Frequency Duration No. of Cycles Acceleration

1 Sinusoidal 2.5Hz 8 sec 20 190 gal

2 Sinusoidal lO.OHz 2 sec 20 315 gal

3 Sinusoidal 5.0Hz 4 sec 20 208 gal

START

0 4 0 6

Cyclic Rotio, N/Ny 08

Fig.l Rate of Pore -Water Pressure Buildup

in Cyclic Simple Shear Tests (2)

J

bCO

.2 R

« . 05 L M

CO4)

in

N N 1 Number of Cycles log N

Fig. 2 Relation between Stress Ratio

and Number of Cycles

' - - Sand Layer - -

Fig. 4 Embankment on Li que fi able Ground

Design Earthquake Ground Condition Embankment Condition

Cyclic Shear

Stress Ratio L

Liquefaction

Resistance Ratio R

Liquefaction Resistance Factor F,.= R/L

Ft,~ue/ a Relation

Excess Pore

Pressure ue

Stability Safety Factor Fs

Fig.3 Procedure for Stability Analysis of

Einbankments on Liquefiable Ground

^ slip surface

Fig.5 Stability Calculation by Slice Method

52
0 0.5 I 1.5 2 2.5 3 35 4 4.5 5 Cycle N/NjQ

Fig. 6 Normalized Curves from Pore Pressure

Build-up Equation (8)

(0) S/D'O V. ^(b) S/P.O.I •/.

^(C) S/P. 1. 1%

WWW/

(d) S/D-2.7%

S=8ettlenient of structure D=depth of sand deposit

Numbers on contour line

indicate : ^JU/ovi

Fig. 7 Contours of Maximum Pore Pressure Buildup

Ratio, u / o ZD (%) for Test A 20 (11)

Fig.8 Contour of Excess Porewater Pressure Ratic

r u = u / o 'vo just after the Shaking (Test

1 at 220 gal Shaking ;t=9 sec) (12)

• P12 'PIT

. 0

(c) Test 2 (T

location :

—1—1——1—'—'—"—•

2 qal shaking)

PI 2 m—

^

M

(d) Test 2 (17 location :

2 qal shaking) PIT

0 0.6 -0.6 Shear strain : T (») 0 0.6 Shear strain : T ( »

)

Fig.9 Shear Stress-Strain Relationships in the Models (12)

75 r

^0 i^20 AO -Z5_

J i

bentonite layer (thickness = S mm)

O 0

• • + +

• • o o

• •

0 O 0 •P21 •P22' 'P^.*

0 O 0

•P3 •Pi* 'K'

0•P26

0.P17

O•P8

O•P27

°P18

•P9

o

—cr - - (unit in cm)

o Acceleromeler + Shear strain gauge

• P.W.R lransc*jcer I Displacement meter

Fig. 10 Analyzed Model Embankments

53
Model shaking test

Acceleration record Pore pressure record

3) Calculation of dynamic shear stress ratio

1) Calculation of pore pressure ratio

A) Calculation of the number of

cycles to liquefaction

5) Assumption of the liquefaction resistance curve

Cumulative damage theory

6) Evaluation of liquefaction resistance curve

2) Arrangement of the

distribution of pore

pressure ratio r u

7) Calculation of equivalent

cyclic shear stress ratio L

8) Evaluation of dynamic shear resistance ratio R

9) Computation of liquefaction resistance factor

10) Arrangement of the relationship

of liquefaction resistance factor F I, and pore pressure ratio

Fig. 11 Procedure of Data Arrangement and Calculation

u

Fig. 12 Estimation of the Effective Overburden

Pressure from Pore Pressure Record

O Accel erometer

• Pore pressure gauge sz

r 3.t -t--rj- shear stress ratio at P

h< b^A4 tj _ r t (Ai hi tAa hg )+ r a « t (A3 h3 4A4 h4

)

-«—-W-p —a uc ^(rt(hi+h2)+(rs.t-l)(h3th4);

Fig.l3 Calculation of Dynamic Shear Stress Ratio

2 5 10 20 50 100 200 500

Number of Cycles N

Fig. 14 Reference Liquefaction Resistance Curve from

Cyclic Torsion Shear Test of Toyoura Sand

54
•P21 •P26

.P3. .P8

b) Free field( Case 3, P8) d) Below the embankment( Case3, P3)

Fig.l5 Time Histories of x ^1 a L, FL, ru

55^
a) Case 1 b) Case 2

2 sec(5cycles) 0.5 sec(5cycles)

If

1 ^''^

8 sec(20cycles)

7t

2 eec(20cycles)

c) Case 3

1 eec(5cycles)

4 Bec(20cycles)

Fig. 16 Distribution of Excess Pore Pressure Ratio

I.o

a

0) a.

0 P8 •PS

D P17 • PIS A P26 » P27

"1

•^f • ,

0- 05

0P3 •pt

0P12 "Pis

»PJ1 »P22

Ou 05

1

•P6

• PI J »P24

'P'^^^

Liquefaction Resistance Factor Fi. Liquefaction Resistance Factor Fl, Liquefaction Resistance Factor Fi, a) Casel Free field b) Casel Below the crest c) Casel Below the slope toe

oP6 «P9

o PI7 Pie 4 P2S » P27

1

• \ *

\ o\

Fl ' -

\ *\

Liquefaction Resistance Factor Fl

d) Case2 Free field

o

0)

I- Q.Q5

(U

I.O

0.

0P3

DPI2 'Pia

aP2l »P22

Liquefaction Resistance Factor Fl

e) Case2 Below the crest

o

u

Q- 05

• P6

PIS »P»

- » ) ^!

\ \^'--

Fc'" \ A

Liquefaction Resistance Factor Fl. 0 Case2 Below the slope toe

0)

la

aj

v. Q- 05

0)

t.o

cu

<

OPg (PS

OPt7 PIS 4P26 AP27

f —

Fl-' '

1 o ~

^-.5

<5.

CUQ5

01

oP3 •P4

0PI2 "PIS

»P2I *P22

•A

L^A

Fu---

Liquefaction Resistance Factor Fl

g) Case3 Free field

Liquefaction Resistance Factor Fl

h) CaseS Below the crest

u

a. 05

•

F

• PB

P15 aPX

4

Fl-'-- f

Liquefaction Resistance Factor Fl

i) CaseS Below the slope toe

Fig. 17 Relationship between Liquefaction Resistance Factor F i_ and Excess Fore Pressure Ratio r

56
Maximum Ground Acceleration on Soft Clay Sites

by

Tingley K. Lew*

ABSTRACT

During the 7.1 magnitude 1989 Loma Prieta northern

California earthquake, relatively large horizontal

maximum ground accelerations (MGAs) were re- corded on soft clay (young San Francisco Bay mud)

sites. Some of these MGAs were as much as three to four times larger than those recorded nearby on rock

about the same distance from the fault rupture. These

large MGAs were primarily caused by resonance between the surface clay deposits and the dominant

frequency components of the incident waves. An

equation is presented for determining the upperbound

MGA on soft clay sites. The upperbound MGA on

soft clay sites is controlled by the soft clay shear

strength (t^,), density, and layer thickness. The theoretical results were compared with the Loma Prieta earthquake MGA data from young San Francisco Bay

mud sites. With a couple of exceptions, the t^^ =

1,000 psf (47.9 kPa) theoretical curve defines the upperbound MGA data for soft clay sites in the Loma

Prieta earthquake.

The equation presented can be used to compute the upperbound MGA on a soft clay site from the soft clay density, shear strength, and layer thickness in- cluding that of the surficial fill. For sites where only

the soft clay layer thickness is known, the approximate upperbound site MGA can be determined from

the MGA versus layer thickness curve presented for T^,= 1,000 psf (47.9 kPa).

KEYWORDS: Earthquake; shear strength; layer thickness; maximum ground acceleration; soft clay; strongmotion; young bay mud.

1.0 INTRODUCTION

The expected earthquake ground motions at a site are needed for the design of new facilities and the esti- mation of potential earthquake damage in natural

disaster preparedness planning. One of the major

parameters used in the design of new facilities and

the prediction of potential earthquake damage to existing facilities is the horizontal maximum ground acceleration (MGA). Generally, the frequency content

of the earthquake ground motion is considered indi- rectly by sp)ecifying the shapes of the design spectra.

Duration of the ground motion is considered implicitly in the design of new facilities by the indirect specifications of ductility requirements (e.g., the R^

values and structural detailing requirements in the 1990 Recommended Lateral Force Requirement by

the Structural Engineers Association of California). In the estimation of potential earthquake damage to existing facilities, the assumed available ductility is based on the structural detailing of the facility, past

earthquake performance of similar types of construction, and engineering judgment.

The MGA, duration, and frequency content of earth- quake ground motion at a site depend on the source mechanism, path of the incident waves, and the local

soil conditions at the site. Generally, we do not know

the source mechanism nor the path of the incident waves from the source to the site for future earth- quakes relative to a given site. We can only estimate

the effects that the local soil conditions may have on

the incident waves.

The accelerations of the incident earthquake waves at the bedrock of a site can be greatly amplified by

resonance of the site soil deposits. Resonance occurs when the characteristic shear wave periods of the surface soil strata are near one of the dominant periods of the incident waves (Ref 1 through 5). Thus,

structural damage at the site is intensified. Even

greater earthquake damage occurs if the fundamental

period of the structure is near one of the characteristic shear wave periods for the site. Double resonance of

the site and structure relative to the dominant periods

of incident earthquake waves occurs under such a

condition. This phenomenon was demonstrated by

the structural damage at Mexico City about 300 km

from the epicenter in the 1985 Mexico earthquake

(Ref 6).

The aim of this paper is to develop an analytical

relationship for determining the upper bound hori- zontal ground acceleration on soft clay sites.

'Naval Civil Engineering Laboratory

Port Hueneme, CA 93043

57
Currently, the design MGA for a site is estimated

from empirical attenuation relationships derived mainly

iTom recorded strong-motion data in the western United

States. Definitive soil condition data (e.g., depth of

the soil layers, soil type, shear strength, and shear wave velocities) are not available for most of the strong-motion recording sites in the United States. Hence, the attenuation relationships generally differ- entiate only between soil and rock (Ref T). In addition, the majority of the strong-motion data used to develop the attenuation relationships comes from

earthquakes and recording sites in southern California. Most of these recording sites are located in deep

alluvium, generally much thicker than those in the San Francisco Bay area. On deep alluvium sites, amplification of the incident waves traveling up to the surface would be offset by the attenuation if the sediment section were relatively thick. The lack of

adequate strong-motion data from soft clay sites pre- vents the establishment of a separate, statistically

significant data set for analysis.

The heaviest damaged areas in the 1906 San Francisco earthquake are on sites with soft clay soil (young

bay mud, YBM) (Ref 8), Figure 1. During the 7.1 magnitude 1989 Loma Prieta northern California

earthquake, many strong-motion records were obtained from soft clay sites as well as stiff soil and rock

sites in the San Francisco Bay area (Ref 9 and 10). A

plot of the MGA verses the closest distance to fault rupture is shown in Figure 2? The median and 84

percentile curves using Campbell's attenuation equation (Ref 1 1) is given for reference. Each data point

shown in the figure is the average of the MGA in the two orthogonal horizontal directions. With a few

exceptions, most of the data for stiff soil and rock

sites are within the median plus one standard deviation (o) curve. By contrast, most of the data for soft

soil sites are above the (median + a) curve, with some

of them apparently larger than (median + 2a). There

appears to be an upper limit for MGA in soft soil sites of about 0.3 g, at least for this earthquake. It is apparent that the incident waves at soft clay sites (YBM) were greatly amplified by the relatively soft clay layers near the surface. For example, the re- corded MGA on soft clay at the Naval Station Activity (NSA) Treasure Island is 0.16 g. The corresponding recorded MGA component on rock at nearby

Verba Buena Island is only 0.06 g. Thus, the apparent amplification factor by the soft clay layer at NSA

Treasure Island is 2.67. At some other soft clay sites, the dynamic amplification factor relative to adjacent

rock sites is as large as about 4.0. Dynamic amplification factors between 4 and 8 are determined from

low level aftershock recordings for soft clay sites in the Loma Prieta earthquake.

2.0 PROPERTIES OF YOUNG SAN FRANCISCO

BAY MUD

The San Francisco Bay is bounded primarily by marshlands, alluvial deposits, and the coastal ranges. The

major geological units in the San Francisco Bay are shown in Figure 1. These units are classified as bedrock, alluvium, and YBM, Reference 8.

1. Bed rock consists of standstone, siltstone, and greenstone of the Franciscan formation. Bedrock at YBM

sites is generally located at between about 200 and

600 feet (61.0 and 182.9 m) below the surface.

2. Alluvium consists of late Quatenary (10,000 yr to

3 my b.p.) flood plain deposits of silt and clay, inter layered with alluvial fans and stream bed deposits of sand and gravel.

3. Young bay mud consists of Holocene age (10,000

yr b.p.) sediments, soft, water-saturated, organic-rich

silts and clays, occasionally interlayered with sand

deposits. These sediments are derived from sus- pended materials brought into San Francisco Bay by

rivers draining the Central Valley and streams drain- ing the Southern Bay area. Typical YBM properties

are shown in Table 1. The thicknesses of YBM vary between zero and 120 feet (36.6 m) for the sites shown. YBM densities vary between 90.5 and 113

pcf (1,450 and 1,810 kg/m^) with a mean of 101.7 pcf

(1,629 kg/m') and coefficient variation of 0.053. Their

shear strengths vary between 100 and 1,500 psf (4.79

and 71.8 kPa) with a mean of 583 psf (27.9 kPa) and o of 230 psf ( 1 1 .0 kPa). The water contents of YBM

vary between 35 and 120%, with the majority being

greater than 50%.

Seed and Sun (Ref 12) reported that the shear modulus of YBM decreases with increasing shear strain. Figure 3. By contrast, the damping ratio of YBM

increases with shear strain. The rate of shear modulus reduction with increasing strain for YBM is sig-

^A summary of the attenuation relationships com- monly used in the U.S. is presented in this reference. 'For emphasis, no attempt was made in the figure to differentiate the data from the U.S. Geological Sur- vey (USGS) and the California Strong Motion Instru- mentation Program (CSMIP).

58
nificantly less than that for typical sands, but is generally in good agreement with other clays. The

damping characteristics are well within the range for typical clays. Figure 3.

Seed and Sun presented a summary of the shear wave

velocity profiles of YBM at six San Francisco Bay

shore sites measured by various investigators. Figure

4. The former reported that the shear wave velocity

is essentially constant, 250 fps (76.2 m/s), for the upper 30 feet (9.1 m). Thereafter, the velocity ap- pears to increase gradually to about 500 fps (152.4 mJ

s) at 60 feet depth. However, we noted that the shear wave velocity at some sites may remain constant to depths of 60 feet or greater (e.g., the San Mateo site shown in Figure 4). Although no shear wave velocity

profile is currently available for the NSA Treasure

Island strong-motion recording site, the standard

penetration values for the YBM at the site remain

near zero for depths of between 45 and 100 feet (13.7

and 30.5 m) as for the upper 20 feet (6.1 m) of sandy

hydraulic fill over 15 feet (4.6 m) of loose sandy fill. This suggests that the shear wave velocity is rela- tively low to the l(X)-foot depth.

Young Bay mud tends to flow laterally toward the direction of minimum potential energy when not confined during earthquake ground shaking (e.g., into sloughs, channels, or excavations). Although YBM

does not liquefy during earthquakes, thin, clean, dis- continuous, loose sand layers within the YBM generally have a high potential for liquefaction. Water

saturated, loose silty sand and/or sandy fill on top of

the YBM also have a relatively high potential for liquefaction during earthquakes."* Differential settle- ment, one of the major causes of stmctural and underground utility damage in earthquakes, generally ac- companies soil lateral flow or spreading, liquefaction, and/or vibratory compaction.

3.0 THEORY

Based on the strong-motion data available before the 1989 Loma Prieta earthquake, seismologists tended

to find good agreement between site specific dynamic amplification factors determined from weak

(10-' to 10-3 g) and strong (0.04 - 0.20 g) motion.

Reference 13. Accordingly, the maximum ground

acceleration for a soft clay sites on the San Francisco

Bay shore can be determined by modifying the site rock acceleration by the site dynamic amplification

factor (DAF). For example, the estimated rock MGA

at the Naval Air Station (NAS) Alameda from a 7.0

magnitude about 7.5 km east of the site on the Hayward fault is 0.35 g. The DAF for the sites from the Loma Prieta earthquake is 3.25. The DAF was esti- mated from the 0.26 g MGA recorded at the sites with

15 feet (4.6 m) of medium dense sand fill over 28 feet

(8.5 m) of YBM and the 0.08 g MGA recorded on

rock or rock-like material at the Strawberry Canyon,

Berkeley station about the same distance from the nearest ruptured surface. Hence, the MGA at NAS

Alameda would be 1.14 g from a 7.0 magnitude

earthquake on the Hayward fault. Physically, this is not probable because at the relative low shear strength

of the YBM, Table 1.

In engineering mechanics, the strength of a given

material defines the conditions that produce failure. Failure takes place when a material is subjected to a

certain stress state, usually called strength. Failure of

the material occurs by fracture or yielding (plastic

flow). In the fracture mode, the material disinte- grates at certain stresses or strains. In the yielding

mode, the material strain increases by relatively large

amounts at constant stress or with relatively small

stress increments. Soft clay fails in the yielding/

plastic flow mode whereby relatively large strain increases accompanies each stress increment.

By definition then, no known material is able to transmit stresses that are greater than its inherent

strength. The incident earthquake horizontal shear waves at the site bedrock generally propagates verti- cally, or nearly so, upward to the surface. When

these shear waves encounter the soft clay layer, the

relatively low shear strength of the layer limits the amplitudes (accelerations) of the waves consistent

with its shear strength. Thus, the shear strength of the

soft clay limits the maximum acceleration that can be

transmitted to the ground surface. When the accelerations of these waves exceed the shear strength of

the soft clay, the clay yields and, thereby, limits the

stress transmitted to its yield sfrength. In addition,

energy is dissipated by the soft clay layer as it yields.

This further limits the stress amplitudes of subse- quent waves transmitted to the ground surface.

"Liquefaction potential for saturated, loose sandy fills can be mitigated by ground improvement techniques,

such as vibrafloation, compaction piles, and gravel

columns. Such techniques appear to have successfully prevented liquefaction in localized areas on NSA Treasure Island during the Loma Prieta earth- quake.

59
The approach used in the following theoretical devel- opment is similar to that used by Seed and Idriss (Ref

14) in their development of a simplified procedure

for determining soil liquefaction.

First we assume that the soft clay layer (column) of

thickness, h, behaves as a rigid body with maximum

acceleration a^^^ on top, and maximum rigid body

shear stress (t ) on the bottom as shovm in Figure 5a. ^ men r ^ The layer thickness, h, may include thin interlayers

of sand and a relatively thick layer of loose silty sand

and/or fill on the surface. It is assumed that these

layers do not have a significant effect on the overall dynamic response of the soft clay layer as a whole.

The maximum shear stress at depth h for the rigid soil column:

max r yh ax8

(1)

where y = wet unit weight of soft clay (FL'^)

h = soft clay layer thickness (L)

^max ^ maximum ground acceleration (FT^)

g = gravitational constant (FT"^)

Because the soft clay column is deformable, the actual shear stress at depth h increases at a decreasing

rate with depth as shown in Figure 5b. We account

for this by modifying (t^), by a reduction factor r^, Figure 5c:

^^ma)^ d

" M ^W^r (2)

The reduction factors used in this study are given in Table 2. These values correspond to the average

values given by Seed and Idriss (Ref 14). By substi- tuting Equation (2) into Equation (1), we get

max d

(3)

ult max d

(5)

where MGA = maximum ground acceleration

in g units

= shear strength of soft clay

By substituting Equations (4) and (5) into Equation

(3) and solving for MGA, we get

MGA =

'ult (6)

Thus, the upperbound maximum ground acceleration on a soft clay site is a function of the soft clay shear

strength and an inverse function of the soft clay

density and thickness.

According to the unit weights and shear strengths

given in Table 1 for the young San Francisco Bay

mud, it is judged that the following values are representative:

y = lOOpcf (1.602 kg/m^)

T^, = 300-1,000 psf (14.4 - 47.9 kPa)

The above values were substituted into Equation (6) for soft clay layer thicknesses t)etween 10 and 100

feet (3.05 and 30.5 m). The MGAs for shear shengths

of 300, 600, and 1,000 psf (14.4, 28.7, and 47.9 kPa)

are plotted in Figure 6. From the curves in the figure,

it is apparent that MGA on soft clay sites increases

with the soft clay shear strength. For a given shear

strength, the site MGA decreases with increasing

layer thickness due to the increase in soil inertia force with layer thickness. The site MGA decreases rapidly with increasing layer thickness for thicknesses

between 10 and 40 feet (3.05 and 12.2 m). For layer

thickness greater than 40 feet (12.2 m), the site MGA

is relatively insensitive to increases in layer thickness. The increase in inertia force with layer thick- ness in this thickness range offsets the reduction in shear stress in the soil column.

Furthermore, we define

MGA =

ax

and

(4)

4.0 COMPARISON WITH EARTHQUAKE DATA

The theoretical results are compared with the maximum ground acceleration data from young San Francisco Bay mud sites in the 1989 Loma Prieta earth60
quake in Figure 7. Data points for sites with known

soft clay layer thickness are represented by the solid symbols in the figure. Because we have the soft clay

layer thicknesses for only 14 of the 20 young San

Francisco Bay mud sites shown on Tables 3 and 4, the approximate thicknesses for the other sites were computed from the dominant site periods estimated

from the strong-motion records by using the following equation.

h = - (7)

where h = layer thickness (L)

T = fundamental shear wave velocity (T)

= shear wave velocity at small

strains (LT ')

conventional probabilistic stmctural and dynamic

analysis of the Church with a known level of damage.

The site MGA estimate was based on the assumption

that all the damage to the Church was caused by the main shock. There were many smaller magnitude

aftershocks following the main shock. However, the

relative contribution of these aftershocks to the overall damage of the Church is unknown. Thus, the

resulting estimate is considered an upperbound value

by Elton and Marciano (Ref 15). The estimated mean MGA is 0.33 g with a coefficient of variation (COV)

of 0.24. In Figure 7, the range of probable MGAs at the site is represented by the vertical bar (between

0.17 and 0.33 g). We judged that the most probable

MGA at the site is about 0.25 g, the solid symbol.

The 1886 Charleston, SC earthquake data follows the general trend of the MGA data recorded in young San

Francisco Bay mud sites during the 1989 Loma Prieta earthquake.

R = reduction factor to account for the decrease in with increasing

shear strain^

For simplicity, = 250 fps (76.2 m/s) and R = 0.67 were used for all the other sites. For sites with two

distinct dominant periods, the computed layer thick- nesses for the two periods were summed to obtain the

total soft clay layer thickness.

If the proposed theory is valid, all the data points

should be below the MGA curve corresponding to the upperbound shear strength of the soft clay (YBM).

With a couple of exceptions, all the data points are below the curve for t^, = 1,000 psf (47.9 kPa), Figure

7. The choosing of t^, = 1,(XX) psf was fortuitous. However, the trend of the data definitely follows the asymptotic portion of the curves. The grouping of

the data between the 300 psf and 1,(XX) psf curves for

soil thicknesses between about 10 and 100 feet (1.98

and 19.8 m) suggests that the dynamic shear strength

at the YBM sites is generally between 3(X) and 1 ,000

psf (9.66 and 47.9 kPa).

As a further check on the validity of the theory

developed, the estimated ground acceleration near St. Michaels Church, Charleston during the 1886 Charleston, South Carolina earthquake (Ref 15) is plotted

in Figure 7. The soft clay layer thickness near the site

is about 52 feet (15.8 m). Reference 16. The earth- quake had an estimated magnitude between 6.5 and

7.0 with its epicenter about 15 miles (24 km) north- west of Charleston. The site MGA was estimated by

5.0 CONCLUSIONS

Based on theory presented and the comparison be- tween the theoretical results with strong-motion MGA

data from young San Francisco Bay mud (soft clay)

sites in the 1989 Loma Prieta earthquake, the follow- ing conclusions are drawn:

1. The upperbound MGA on soft clay sites is controlled by the soft clay shear strength (t^,), density, and thickness.

2. With a couple of exceptions, the t^^ =

1,000 psf (47.9 pKa) theoretical curve defines the up- perbound the MGA data for soft clay sites in the Loma Prieta earthquake.

3. The trend of the Loma Prieta earthquake

data follows the trend of the theoretical curves.

4. The equation presented can be used to compute the upperbound MGA on a soft clay site from the soft clay density, shear strength, and layer

thickness including that of the surficial fill.

5. For sites where only the soft clay layer

thickness is known, the approximate upperbound site MGA can be determined from the MGA versus soft

5V^ = V^G/p, where G = shear modulus and p = mass

density. See Figure 3 for reduction of G with increas- ing shearing strain.

61
clay layer thickness curve for t^, = 1,000 psf (47.9 kPa)

in Figure 6.

6.0 RECOMMENDATIONS

The MGA data from 20 strong-motion recording sta- tions in the Loma Prieta earthquake were used in the comparison with the theoretical results. The soft clay

layer thicknesses are known for only 14 sites. The

approximate layer thicknesses for the other six sites were estimated from the dominant periods in the strong-motion accelerograms for the sites. It is rec- ommended that the soil layer thicknesses and soil properties, such as shear strength, shear wave velocities, and densities be determined for all the soft clay

strong-motion recording sites where they are not

available. Comparison of the theory presented with

recorded MGAs on soft clay sites from other earth- quakes is also recommended and encouraged.

7.0 ACKNOWLEDGMENTS

The continual support by Mr. H. Nickerson (NAVFAC

Code 042B) and Mr. J. Cecilio (NAVFAC Chief Engineer) is appreciated. Mr. D. M. Boore (USGS

geophysicist) and Mr. R. B. Darragh (CSMIP, seis- mologist) provided the author with the soil thickness data for the USGS and CSMIP soft clay recording

sites, respectively.

This paper was sponsored by the Naval Facilities Engineering Command.

8.0 REFERENCES

1. H.B. Seed and I.M. Idriss. "Analysis of Ground

Motions at Union Bay, Seattle During Earthquake

and Distant Nuclear Blasts." Bulletin of the Seismological Society of America, 60:1, 125-136, Feb 1970.

2. R.D. Brocherdt. "Effects of Local Geology on Ground Motion Near San Francisco Bay." Bulletin

of the Seismological Society of America, 60:1, 29-

61, Feb 1970.

3. W.B. Joyner, R.E. Warrick, and A.A. Oliver, III. "Analysis of Seismograms From a Downhole Array

in Sediments Near San Francisco Bay." Bulletin of

the Seismological Society of America, 6:3, 937-958,

1976.

4. S.H. Scale and R.J. Archuleta. "Site Amplification and Attenuation of Strong Ground Motions."

Bulletin of the Seismological Society of America,

76:6, 1673-1699, 1989.

5. M.D. Trifunac. "How to Model Amplification of

Strong Earthquake Motions by Local Soil and Geological Site Conditions." Earthquake Engineering

and Structural Dynamics, No. 19, 833-846, 1990.

6. E. Rosenblueth. "The Mexican Earthquake: A

First Hand Report," American Society of Civil Engineers, Civil Engineering, 56:1, 38-40, 1986.

7. W.B. Joyner and D.M. Boore. "Measurement,

Characterization, and Prediction of Strong Ground

Motion" in Earthquake Engineering and Soil Dynamics II - Recent Advances in Ground Motion Evaluation. Geotechnical Special Publication No. 20,

American Society of Civil Engineers, 43-102, 1988.

8. R.D. Borcherdt, W.B. Joyner, R.E. Warrick, and

J.F. Gibbs. "Response of Local Geological Units to Ground Shaking." U.S. Geological Survey Professional Paper 941-A, edited by R.D. Borcherdt, A52-

A67, 1975.

9. R, Maley, A. Acosta, F. Ellis, E. Etheredge, L. Foote, D. Johnson, R. Porcella, M. Salsman, and J. Switzer. "U.S. Geological Survey Strong- Motion

Records From Northern California (Loma Prieta)

Earthquake of October 17, 1989." U.S. Geological

Survey Open File Report 89-568, Menlo Park, CA,

85 pp. 1989.

10. A. Shakal, M. Huang, M. Richie, C. Ventura, T. Cao, R. Sherburne, M. Savage, R. Darragh, and C.

Peterson. "CSMIP Strong-Motion Records From the Santa Cruz Mountain (Loma Prieta) California, Earth- quake of 17 October 1989." California Division of

Mines and Geology Report OSMS 89-06, Sacramento,

CA, 195 pp, 1989.

11. K.W.Campbell. "Near-Source Attenuation of

Peak Horizontal Acceleration." Bulletin of the Seis- mological Society of America, 71:6, 2039 - 2070,

1981.

12. H.B. Seed and J.I. Sun. "Implications of Site Effects in the Mexico City Earthquake of Sept. 19,

1985 for Earthquake Resistant Design Criteria in the San Francisco Bay Area of California." Report No.

UCB/EERC - 89/03, Earthquake Engineering Re62
search Center, University of California, Berkeley,

CA, Mar 1989.

13. K. Aki. "Local Site Effects on Strong Ground

Motion" in Earthquake Engineering and Soil Dynamics II - Recent Advanced in Ground Motion Evaluation. Geotechnical Special Publication No. 20,

American Society of Civil Engineers, New York,

NY, 103-155, 1988.

14. H.B. Seed and I.M. Idriss. "Ground Motion and

Liquefaction During Earthquakes." Earthquake Engineering Research Institute, El Cerrito, CA, 1982.

15. D.J. Elton and E.A, Marciano, "Ground Acceleration Near St. Michael's Church During the 1886

Charleston, S.C. Earthquake." Earthquake Spectra,

6:1,81-103, 1990.

16. D.J. Elton and J.R. Martin. "Dynamic Site Periods in Charleston, S.C." Earthquake Spectra,

5:4, 703-734, 1989.

63
JS o o o o o

+>N1

O o o o o o o O o o

tn +• o o o o eg o o o to Kl

re c tA 00 to

0) 0) 1 1 1 1 1 1 1 1 1 1 j: o O o o O o o o o o

(A +> 1-4 o o o to o o o o o o N H I-I I-i Csl >*• •* >* M to Ifl 0) c CO

4J 0 o ff\

Q

J

a Oh 1—t

ui 1 •H 1 1 1 1 1 1 1 <a 1 1 1 1 1 §

rt

Q •H ^a so Csl

a E • •

K> 1—1 N

+> o

b> 1-1 r-l

M r*f t-«

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vO 1-1 CVJ CO ro i-i 1—1 rH Os o

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t2

so so so

re 0 M 10 10 to (0 CO o> s u MnD •o c 3 m 10 C +> -* o O to o o o o u In 1 1 1 CvJ CsJ so so o to Csl 1 1 10 1 1 rt o o o i-i A (—1 »—i

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V)

v

n

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u +>

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4* CU re i-i re re

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B V)

n rH

H

B 9 •rt 0 re M

J3

VI VI o i 0) a. M a a. 8 < M .3 X 0 M W u. tS X s

64
Table 2. Maximum Shear Stress Reduction Factor for

Deformable Body Relative to Rigid Body , r .

Depth h rd

-

( Tmax d

(ft) ( Tmax

)r

0 1 00

5 0 99

JO 0 98

15 0 97 ZO 0 96

_ 50 0 92 40 0 85

50 0 75 60 0 68

70 0 61 80 0 55

90 0 55

100 0.50

65
n

0)n

U

u U)

0) (I) c +> 1 1 1 ui CO o> 1 1 o

so Csl Csl so 0

VI

<aM

r-t i-i

u

ro ro HI 0) >j- »> so o Os -J- IS. o +> 3 B p 0 +> u M r—i

1^ fs. ui CsJ 1—I

so

H

U J]

u Ul >!•

M 1 Ul Ul ^ Hn

TJ 0)

O

0)

n o Os

O Ul Csl a- o>

+>

M

10 (U

-—^

Ml r-H »> CO r<i r1 to Csl so CO

g s -P s K1 1-1 csl Csl 3- Csl to to • It-I N 1—1 N. \ N \ \ \ N N. V.

+>

u vt ••0 ^a- Kl Ul * Csl to rH o

lU jB t—<

OS K1 1—1 o CO OS rH rsi

it-i

H »—1

B—1

rH

D

CA

0) •o CO in N1 Ul so ts. to

(0 0)

M ss N. rs. o o U) ts. CO Os 3 +>

(T (0 <u 5 u o o O o o o i—1

-' o o o d

.§

•p

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Ui UJ o o O o O o o o o o o nHB

(40 T3 <: »> CO CQ «—1 o 1—1 so

144 lU

a Cfl Csl Si f-4

M r-1 r—1

1—1 rH ri

•rt -p t) s 1-1 u « o o o o o o O o o o O

O e 3 N (31 N •s \ s \ \ \ \ s s \ U (4 •H o- CVI HI csl Csl Csl Csl csl OS CO o

(13 Csl CslQQ

B 14 <U 0 , . U Ul £ X o o o o o o o o o o o o

<u j3

+•u0B )H I-I iH

u

u 14 ^4 i-H •H •H •H a) d) d) d) •rt n) 0 0 Q 0 0 > •a •a > > T3 > TJ 0 TJ

+>

n

M 3 V) <n 0 3 3 0 3 0 3 0 3 (4 3

d) 0) E E E E E E E H -p iH p -P +> -p 1-1 I-I r-t r-l -P 0 >, Uh 1-1 I-I >, >, r-l >, rH >, rH >, <f4 >, a. I/) 8 .55 H 5

•r( •H

H

(0 ul CQ Ul Ul U- U> i55 ul in

oma

(0 (0 +> » n

a

fO

O •sCO COO rH so

H Csl N» CO CO Q

0) 1 0) 1 3 rH

f-l S (B 1 ID r-l 0) •P C

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N

U 'H 1 -P < 3 0) 3 g

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B

hB

H -P (A Oa E (4 o +> •P t) d)

B

n (0 0 n 0 3 > d) cn > rH

H 0 h u X 0 p 0 •< d> < m

•H (U 0 0 X

H o u >, H' 4> > ^ < z 0 0) o 'P 0 (4 0 'iH (4 (0u

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u

B 1 H o (A

u d) o u DU

h

B

J >, •p > >,

m

p 14 (4 3 (4 (4 d> 6 3 (0 •H -P •o (1) M H -P •H § •P 0 U. p (4 u 0 0) •H iti iH H 0 (a U +> u X

"O

w (0 ;4 +> iHu ji r-l t4 E u

B •P >, (QB

u

S. < i-i (0

B •r( j3 n)

u (Q

B d) ro

B 3 0 H 1-4 < > (4 0) > U ;4 U 0 rH tn l4 (0 a.

•t-l 0 0) s >i U- U. s dl U.

u 14 E U iH (31 Ui 0 (31 E -p U tft a CO 01

(0

B TI U4 "0

B f) Ul d) \0 Ul o 4*

•p

J iH <u 1

r-l 3 0 Kl

<i I o 1—1 dl

w (rt S oa < OS 03 tn u. K1 l§ so Ul so ^ Cvl > n 0

•H o Ul Ul csl so Os to Ul o P d o c csl o> so to o Csl o<

(0 z o Csl sO U) so -3- r—1

CSJ Ul

p r—1

1—1 l-f 1—«

1—1 P4

—1rH i—1

\n

66
el0)n

h (4

(U <u ^ 0 0 K1

•—1

(M vi- 5 u ~ V f-t

ThiM

0)n

uM

ram 0) a)

—

m CO CO V* NO K» so vo

+> 3 >i c +> 0 +> ^ i»4 o> o> CVJ to Ul UI H U :3 0 - ro LT) Ul VO Kl -3- <

Thi

(4

0) o CO Ov so fH Ul M

+>

u in so r>. rH 0 ui CO

rem 0) ~ CsJ CvJ CVJ vt -* Ml CSI Csl C 4< V. \ \ \ •s \ ri ^ >^ LO 0 so Ul 0 ui UI rH

•*> 0 - so m •rt O Ml tvl 0 K1 0 ui

III Th D 10 IS. 0 00

0) Vl NO vO 0 o> so so

+•n 0) 0 u o 0 0 r-i 0 UI 0 0

4*

H 4) \ \ \ \ \ CvJ

N .§ •H (4 0) LO 0 Ifl r-l Ul VO

(/)

M fO CVJ K1 CSI 10

uu

Q 0 Q

Q

"O < ^ 9> ,^ so r—1

ID T3 (!)

M CvJ CvJ f—(

»—1

Csl -H 0) m U o 0 0 0 0 g N s. \ s. s. s. N. s. U H — CO 0 IH SO sO

+•

m

N to csi CvJ r-l r-l r-i Csl VI c) 0

UlX o 0 o 0 0 0 u t4 (4 (4 a 0) 0) 0) (U > > > TJ T3 > 0 > T3 > T> > TJ

5? 0 0 S 0 3 3 0 3 0 3 0 3 0 3

4) 0 £ E g

g E E E

+> rH iH iH iH rH iH rH rH

•rt 0 l-( T3 p-t iH >i >-i rH iH >i rH >i rH

,25 •H rt

n i-l KS

HH (0H

n 'H

II .2 y_ ffl 1 1

.a .s

+> ^

VO Cvl Ov Ul

-0 r*.

a -

>_ s

(4Q tJ Q Q) Q) 1 41 (4 (4 M H g

X c 1 M

4^ (4 COH (0 c Q 4) 0 z )^ (4 (4 4i ^ r-l (4 •H 4) Q (U M C •H c c [-1 4» CC Q flu 44 1 'H 0 0 H ID n 1 Ui l4 3 CvJ T3 0 .a 0 TJ rH < ^ LU rH M 01 iH M 01 (0 8 <; •v -W 'H r-4 •H 'H I-H z TJ ^ 1 0 3 u (0 0 3

(J s i-t 4^ XI c •H c Xi 0) n (Q TJ r-t (0 U a tl c u C C < 03 u bl u 0) 3 0 •0 rH

OJ 0) u. 0 nj 0 u. 411 *H c

+>

u •H iH Li 0 H 'H +>

m

(4 0 c 4* fl >-i in c c 4h 0)

m iH 0 j: (0 Ml ra Mh +> M

u> Ql 0 in 0 M c0 fO CVJ 0 CVJ .a

•H

M -0

M I-H -J

+> 0 CM UI p—1 n z s s s s s z a0

•H

4»

(0 4*ins0

•HHcih

4>nBW

01

(4 c 3 0

*> •H

a. 4» 3 0s

+>

rH

a 3 0

re

u

<M +»V) 0f

+> > n

01M0

rHU

67
1

I

122°22'30" 121"»52'30"

Bay mud Bedrock

Recording

Alluvium Station

Figure 1 . Geological conditions in the vicinity of San Francisco Bay

(after Brocherdt, et al., Ref 8).

68
{ I I I 1—I « 1 r
' I I I I I I I I I IM| I i I I 1 I I i I I I I I I I III

Typicol damping ratio for cloys

Relationship uaad

' ' ' t t HI 10 10 lO"* 10

Shear Strain (%)

10

Figure 3. Variation of normalized shear modulus and damping ratio with shear strain for you San Francisco Bay mud (after Seed and Sun, Ref 12).

70
20 -

40

60

a.

80

100

120

500

Shear Wave Velocity, V^ (fps)

1000 1500 2000

r-jjT—I—I—I—I—I—

r

1—I—I—I—I—I—I—1—

r

d

\

\\

\

\

\\

\\

\San Mateo (Bridgeway Park)

Ravenswood

Redwood Point

San Mateo (Audubon School)

Coyote Hill Ennbarcadero Center

140 I I L J ' '

' 500 1000 1500 2000

Figure 4. Measured shear wave velocity profiles for six young bay mude sites on the San Francisco Bay shore (after Seed and Sun, Ref 12).

71
72


Broadband Strong-Motion Seismic Observation Network

Operated by Japan Meteorological Agency

operated by Japan Meteorological Agency 12 3 4 Masahiro Yamamoto , Iwao Sasakawa , Akira Nagai , Takeshi Kakishita . 5 6 7 Akihiko Wakayama , Koichi Uhira , Masaaki Seino

SUMMARY

The JMA has deployed a strong motion seismic observation network of newly developed seismometers, the

digital 87 type electromagnetic strong motion seismographs, installed in 74 stations in Oct 1989.

The seismometer has an electromagnetic force feed- back which operates over the frequency band 0.005 to lOHz and the dynamic range 0.03 to 980 gal.

The analysis of data from two earthquakes with magnitude 7.1 and 6.5 is presented in this paper. From the

latter earthquake, strong motion seismogram at near field

(epicentral distance: 13.3km) was obtained without clip

(max. a=330 gal, max. v=27 kine and max, d=9.4 cm).

The focal process analysis based on the new seismographs is also presented.

For the data obtained in the frequency band less than

0.05 ^^z, however, the computed displacement seismograph becomes noisy due to the limitation of minimum

resolution to 30 mgal.

KEY WORDS: Electromagnetic Strong Motion seis- mograph(ESM), Earthquake of near E off NE Japan in 1989, Earthquake of near Izu-Oshima Is. in 1990,

Source process

1. FOREWORD

Since 1950, the mechanical t)'pe analogue strong motion seismographs(MSM) have been operated at 1 14

stations for the local large earthquake observation by

JMA. The MSM has a defect that, in frequency band less than 0.2H2, frequency is proportional to displacement

and the instrument is mechanically clipped only at 3 cm

of displacement JMA has a responsibility for the quick

tsunami warning service. Therefore, it is very important

to observe the large amplitude and low frequency seis- mic waves from tsunamigenic earthquakes. The mechanical strong motion seismograph has a limitation for

that purpose. Therefore, the Meteorological Research Institute (MRI) of JMA had developed a new prototype

strong motion seismometer in 1983 . JMA has started an operation of it in Oct. 1989.

In this paper, we first describe outline of the new

strong motion seismometer and then the station

distribution. Subsequently, the merits and some remaining problems will also be presented based on the result of wave form analysis from two earthquakes which oc- curred in E off NE Japan (M: 7.1) and near Izu-Oshima

Is. (M: 6.5).

2. AN OUTLINE OF THE ELECTROMAGNETIC

STRONG MOTION SEISMOGRAPH {ESM) AND

THE STATION DISTRIBUTION

JMA, in cooperation with MRI, has started the operation of the ESM at 74 out of 114 seismic stations in OCT, 1989. This seismograph operates over the fre- quency band 0.005 to lOHz and dynamic range 0.03 to 980 gal as shown in Fig. 1. As compared with the MSM as shown in the same

Fig. 1, the ESM has the broadband in frequency and high

dynamic range. The ESM is composed of the measurement (MP) and

the processing (PP) parts as shown in Fig. 2. The feed1 Deputy Hcid, Seismology and Volcanology Division, Japan Metcorologica] Agency

2 Senior Teehnica] Staff, Seismology and Volcanology Division, Japan Meteorological Agency

3 Technical Staff, Seismology and Volcanology Division, Japan Meuorological Agency

4 Technical Staff, Seismology and Volcanology Division, Japan Meteorological Agency

5 Research Staff, Seismology and Volcanology Research Division, Meteorological Research Irtstitute 6 Research Staff, Seismology and Volcanology Research Division, Meteorological Research Institute

7 Chief, Seismology and Volcanology Research Division, Meteorological Research Institute

75
back operates over the frequency band 0 to 400 Hz and

d>Tiamic range 1 mgal to 1000 gal. In order to minimize

the noise, three 16 bits ADC are implemented with over- sampling design, for each of the three components.

Digitized 50 sampies/s in MP are transmitted to PP. The same data are also telemetered to relevant regional

Tsunami warning centers and also stored on floppy disk

as accelerogram at each station. The distribution of the 74 stations that are equipped with the ESM is shown in

Fig. 3.

3. WAVE FORM COMPARISON BETWEEN MSM

AND ESM

For monitoring the proper operation of the ESM, two

seismograms from the ESM and the MSM recorded at

the same station are compared. The ESM seismogram

is filtered with the filter that has the same characteristics

as the MSM.

The earthquakes which are used for the comparison

are as follows.

Fig. 4 shows the seismic activities which occurred in E off NE Japan during the period of Oct 27 to Nov. 10, 1989. The main shock took place on Nov. 2 with a magnitude of 7.1 accompanying a tsunami (max. height is 92

cm from peak to trough). The epicentral distribution

(magnitude 4.0 or more) and the cross section along A-B

line are shown in Fig. 4.1 and Fig. 4.2, respectively. The

focal mechanism of main shock shows a reverse type

fault reflecting the E-W compressional stress field as shown in Fig. 4.3 .

Fig. 5 shows the seismic activities near Izu-Oshima

Is. which include an earthquake of magnitude 6.5 (the

main shock) on Feb. 20, 1990. Some disasters such as road collapse and small Tsunami (peak to trough is 32

cm) were reported

Epicentral distribution of the main and after shocks

(magnitude 2.5 and more) is shown in Fig. 5.1. The focal mechanism of the .main shock shows a strike slip fault

3) reflecting the NW-SE compressional stress field . They

are projected on two vertical planes in Fig. 5.3 . Two seismograms, from each of the two seismographs (MSM and ESM) are shown in Fig. 6 from these

two earthquakes. Left and right seismograms corre- spond to the MSM and the ESM respectively.

Fig. 6.1 is the seismograms recorded at Hachinohe

(HAC in Fig. 3. epicentral distance of 150km) from the earthquake of E off NE Japan (M: 7.1). Fig 6.2 is the seismograms recorded at Oshima (epicentral distance of

13 km) and Tokyo (epicentral distance of 1 13 km).

Note tliat the polarities of EW and UD components

are reverse between and £5M.

The seismogram of ESM at Oshima recorded the event satisfactorily, on the other hand, that of MSM is

saturated.

4. SPECTRAL ANALYSIS

Three seismograms recorded at Hachinohe from the E off NE Japan is shown in Fig. 7.1 after filtering with

bandwidth of 0.01 to lOHz for acceleration and 0.05 to 10 Hz for velocity and displacement.

The maximum traced amplitude on this record be- comes about three times as large as the record taken by

theA/5M.

Acceleration spectrum density for EW component of

the period 20 sec. before and after maximum amplitude

is shown on the left of Fig. 7.2. Noise spectrum density with the minimum digitized

resolution of 30 mgal is estimated about 0.1 cm/sec at

0.05 Hz and 0.2 cm/sec at 0.025 Hz. The noise might

overshadow the signal in the frequency range less than 0.0 25 Hz.

Displacement spectrum density after the low cut filtering at 0.05 Hz is shown on the right of Fig. 7.2. Three seismogram records at Oshima from the Izu Oshima Is (M: 6.5) is shown in Fig. 8.1 after filtering

with bandwidth of 0.01 to 10 Hz for acceleration and

0.05 to 10 Hz for velocity and displacement as with Fig.

7.1. Recorded maximum acceleration, velocity and dis- placement are 330 gal, 27 kine and 9.4 cm, respectively.

Acceleration spectrum density for NS component of

the period 20 sec before and after maximum amplitude

is shown on the left of Fig. 8.2.

In this case also, some considerations may become

necessary with regard to the noise level for the records

76
obtained in the low frequency range less than 0.025Hz.

Displacement spectrum density after the low cut filtering at 0.05 Hz is shown on the right of Fig, 7.2. From these two analyses, it becomes clear that an appropriate filter needs to be chosen in case of the wave

analysis in low frequency range.

5. AN OUTLINE OF THE SOURCE PROCESS

OF 1990 NEAR IZU-OSHIMA EARTHQUAKE

The source process of the near Izu-Oshima earth- quake in 1990 was studied with ESM displacement data obtained at Oshima(OSH), Ajiro(AJI), Irozaki(NGT)

and Mishima(MIS), whose locations were shown in

FigS.l. These data were filtered for removing frequency

components higher than 0.3Hz in order to reduce the ambiguity due to inhomogeneous crustal structure near the

surface. Synthetic seismograms were calculated with the

velocity structure model of Yoshii et al. (1985) . The non-linear inversion method proposed by

Takeo (1987) was adopted to deduce the dislocation distribution and rupture propagation on the assumed

fault plane.

In the present study, the main-shock fault was esti- mated fit)m the after shock distribution to be 20km long

and 12km wide (from the depth of 1,5km to 13,5km)

with strike of N352° and dip on 76". The fault plane was

divided into 15 sections and double couple source was

arranged at the center of each section. The source pro- cess of the earthquake was assumed to be expressed by

an integration of double couple sources having the same

source lime function. The synthetic seismograms calculated with the fi- nal solution were compared with the observed seismograms as shown in Fig. 9.1, In this case, only pure suike

slip was taken into consideration. Rapture propagation

and seismic moment distribution over the fault are shown in Figs. 9.2 and 9.3, respectively. They suggest

that the rapture front propagated at the speed of 2-

3km/sec from north to south and the total seismic mo17 ment of 5x10 Nm, most of which was released at the

central part of the fault. These results show that major

dislocation took place in the central part of the fault with

time delay of a few seconds after the onset of rapture and

dislocation was small at shallower portion of the faulL

References

1) Seismology and Volcanology Research Division>lRI: Strong-Motion Seismograph model 83 for the Japan Meteorological Agency Network, Techn. Rep.

MRI, 7. 130p, 1983

2) Sendai District Meteorological Observatory, JMA:

Seismic Activity in the Region far E off Sanriku

(Oct.27-Nov.l8,1989), Rep.Coordinat. Committee

EarthqPredict.. 44, pp38^3, 1990

3) Earthquake Prediction Information Division JMA: On

the Earthquake (M6.5) near the Izu-Oshima Island, February 20, 1990, Rep.Coordinat. Committee

Earthq.Predict..44,ppl47-152 ,1990

4) Ishikawa,Y., K.Matsumura, H.Yokoyama, and

H.Maisumoto: SEIS-PC -- its outline --, Geological Data Processing, 10, pp 19-34, 1985

5) Yosii,T., S.Asano, S.Kubota, Y.Sasaki, H.Okada,

T.Masuda, T.Moriya, and H.Murakami: Crustal Structure in Izu Peninsula, Central Japan, As Derived from

Explosion Seismic Observations. 2.Ito-Matsuzaki Profile, J.Phys.Eanh, 33, pp435-452, 1985

6) Takeo, M.: An Inversion Method to Analyze Uie Rupture Processes of Earthquakes Using Near-Field Seismograms, BuU. Seism. Sos. Am., 77, pp490-513, 1987

77
O.OlHz O.lHz IHz lOHz lOOHz

Frequency —

Fig. 1 Observatinal Domains in frequency and dynamic ranges

of Mectianical Strong Motion seismograph (MSM)

and Elecromagnetic Strong Motion seismograh{ESM).

78
sAccelero-

:i;3rnetex

I X

i z

Signal Pr<

I LPF

JOBS Cth A/D

Conv. LPF

lOBI 4th A/D

Conv.

Li

" J LPF U A/D

Conv.

Trigger Processor

Data Memory

Communication

port

1L_ IL _li

RS422X2

Clock

Measurement part

Communication

port

{Displacement)

Radio

Receiver

Keyboad

ClocJc IT IT

Digital

Filter

(Integrator)

CRT Display

Plotter

Interface Processor FDD

D/A

Converter

Analog Output

(Displacement)

ft

x(Acce eration)

Processing part

Fig. 2 Block diagram of ESM.

79

1989 10/27 60:60 — 1989 11/10 24:00

SCeR£-9l Fig. 4-2

Fig. 4-3

Fig. 4 Outline on the earthquake of E off NE Japan (M: 7.1) in 1989.

Fig. 4-1 Epicentral distribution. Solid circles show the location of the ESM.

HCH is used for the analysis in this report.

Fig. 4-2 Focal distribution projected on cross section A - B in Fig. 4-1.

Fig. 4-3 Focal mechanism solution plotted on the upper hemisphere.

Solid and open circles show the compressional and dilatational first motion respectivery.

81
19?9 93/29 eeiea — 1999 ez/je 24

1

NGT

« o

5 o

* o

39 -J HO^ N-1B7

• up o doa/n

Fig. 5-1 Fig. 5-2

1990 02/29 00:00 - 1990 02/28 24:00 : H 2.5-9.0 N=104 MAP : SAGfiM I B A

KM

2 6 -

00 o

•00 00 © 000 O 00 o

o . O O o O

o O o O 000 oOooo OoOoo

00 O 000 o coo o O o CO OCOOO 00 00

o 000

O

O CD O O 00 O

o 00 o <jO

Coo o O

o

o

O o

ccQP o

C»3'0 o CE© o o

coCEto 00

o CO 000 O o O CO 00

o<53 CO O

o ai3O

00 cocjo o oO O ° - O ^3 Oo

o

X:Y = 1:1

Fig. 5 Outline on the earthquake of near Izu-Oshima Is. (M: 6.5) in 1990.

Fig. 5-1 Epicentral distribution. Solid circles show the location of ESM.

OSH is the Oshima station used in this study.

Fig. 5-2 Focal mechanism solution.

Fig. 5-3 Focal distribution projected on cross sections A - B (left) in Fig. 5-1 and the direction perpendicular to it(right)

82
HACHINOHE

Fig. 6-1

OSHIMA

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Fig. 6-2

6 Comparisons of seismograms. MSM (left) and ESM (right).

ESM's seismogram is filtered out in orded to be the same MSM's characters.

Fig. 6-1 Seismograms of HAC from the earthquake of E off NE Japan in 1 989.

Fig. 6-2 Seismograms of Oshima and Tokyo from the earthquake of near Izu-Oshima is. in IS

83
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Fig. 7-1 Seismograms of acceleration, velocity and displacement.

Fig. 7-2 Spectrum density of acceleration and displacement from the EW component at HAC.

Fig. 7 Seismograms at HAC and spectrum analysis.

Fig. 7-1 Seismograms of acceleration, velocity and displacement.

Fig. 7-2 Spectrum density of acceleration and displacement from the EW component at HAC.

84
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Fig. 8-1 Seismograms of acceleration, velocity and displacement.

Fig. 8-2 Spectrum density of acceleration and displacement from the NS component at Oshima.

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85
Seafloor Seismic Network Offshore Southern California

by

Charles E. Smith*

ABSTRACT

Designing offshore facilities to resist earthquakes pose perplexities that differ from those

problems encountered in designing similar facilities for other environmental loads, such as winds, waves, currents, and ice floes. Normal

operational loads occur frequently over the life- time of a structure, and therefore, design information can be accumulated quickly. Because of

this, design codes and other standards of practice

have evolved fairly rapidly to address these nor- mal types of loads. In contrast, earthquakes are

rare events and data on the response of seafloor

sediments to earthquake-induced ground motions are scarce, thereby introducing significant

uncertainty for analyzing seismic-hazard as- pects of offshore operations. To reduce this

uncertainty, a research program has been undertaken by the Minerals Management Service,

jointly with industry, to develop and deploy

instrumentation to measure seaifloor seismic

motions and to gather data to assess the effeas

of large seismic events predicted for the future.

The primary objectives of this paper are to pres- ent an overview of the Seafloor Earthquake

Measurement System (SEMS) Program and to describe the installation of a SEMS network

offshore southern California.

KEYWORDS: Earthquake hazards, offshore

j platforms, offshore stmctures, seafloor seismic

! response, seismic design criteria.

1. INTRODUCnON

Both Government and industry recognize the

need to acquire design and verification information on seismic activity offshore, particularly at locations of current or future oil and gas operations. Industry needs seismic data to design

facilities that are stmcturally adequate and cost

effective, and the Federal Government requires

the information to fulfill its regulatory responsibilities to ensure safe, pollution-free operations.

The necessity for the data by each group may at

first seem diverse; however, on further examination the needs are foimd to be compatible and

complementary.

Design criteria developed by the oil and gas

industry for offshore structures, as presented in the American Petroleum Institute Recommended Practice 2A (API, 1989), are well established and sound, provided that adequate

information about the environment is available

to make the necessary engineering decisions.

Climatic and oceanographic information for

most offshore areas is available or can be obtained from military, commercial, or private entities that have conducted activities in those areas for decades. The data, although more

complete for some regions than others, nevertheless provide some guidance for offshore engineering.

Even in the most seismic areas of the world, the occurrences of very strong ground motions at a given site is a rare evenL In fact, the infrequency

of recorded strong seismic motions has tended

to foster a passive attitude toward the design of

Research Program Manager

Technology Assessment and Research

Branch

Minerals Management Service

Hemdon, Virginia 22070-4817

87
offshore structures. Seismic data are not convenient by-products of other activities that have

aided in the collection of climatic and oceanographic information. Therefore, an independent, concerted, and sustained effort must be

taken to obtain and quantify earthquake data for use by engineers.

2. OFFSHORE EARTHQUAKES

In the design of offshore facilities, the need for

offshore seismic data to complement earthquake

data obtained from onshore locations has been

well documented (Reece, Ryerson, and McNeil,

1981. NAS, 1976, Henyey et al, 1979). In the

United States, much of the effort has focused on

the areas offshore southem California and, to a lesser extent, offshore Alaska. Subjects of

major importance relate to area seismicity, soil response, energy transfer between the sediment

and platfomi foundation^ overall platform re- sponse, and the consequoKe to operations re- sulting from the response (Smith, 1990). In the

past, estimates of anticipated earthquake ground

motions for offshore structures were made totally firom onshore records because of the lack

of reliable long-term measurements offshore.

Although certain design tasks are common to both onshore and offshore structures, offshore

facilities involve many differences that make

these tasks unique and are the subject of much

interest to engineers. These differences include,

for example, the following aspects: a) attenua- tion factors may be different in saturated

seafloor soils, especially if the soils are soft and/or gassy; b) wave reflection by the water

column may alter site-specific wave forms; c)

offshore stmctures are often much larger than

onshore facilities and have their mass concentrated at the very top of the structure; d) the

presence of water changes the characteristics of

earthquake ground motions and introduces new

forces that result from the added mass effects of

the water, e) earthquakes may cause mud slides

that can travel very long distances on the

seafloor and exert large loads against structures and pipelines; f) earthquakes may give rise to seaquakes, which involve the propagation of

"P" waves through the water column and which

can result in damage to floating facilities. Finally, the locaticHi of offshore epicenters may

have some uncertainty associated with them be- cause the seismographic nets from which they were determined are sited shoreward of an off- shore event (Reece, Ryerson, and McNeil, 1981,

Poulos. 1988).

To eliminate many of the uncertainties associated with offshore earthquake ground motions,

the Minerals Management Service (MMS),

jointly with the oil and gas industry, is sponsoring the installation of a seismic monitoring net- work offshore southem Califomia. The purpose

of this program is to obtain high quality seafloor

seismic data and correlate that information with

measured and predicted platform response data. Scientific and engineering analyses of the vari- ous forms and magnitudes of quakes will be

conducted and compared with onshore data with

a view to extrapolating the data to larger events which arc required for design purposes.

3. RESPONSE METHODOLOGY

The validity of a methodology used for prediaing the response of offshore stmctures to earth- quakes is central to the design of those

structures. In this context, validity refers to the

ability of the methodology to accurately calculate the true platform response. However, it may

not be possible to totally verify the validity of

the methodology over the complete range of

parameters of interest Validity issues can be

partially resolved through empirical approaches

wherein actual field measurements are compared with calculated responses. This approach

is illustrated in figure 1 . The figure highlights

bifurcation of the analytical domain into calculated versus measured response. With this ap- proach, an earthquake can be measured with

land-based instruments. Then these measure- ments can be used as input to a soil response

model which, in tum, calculates the soil re- sponse in the vicinity of a seafloor Instrumentation system. The calculated soil response can

then be compared with the measured response

to determine the validity of the calculation for that specific earthquake. If there is disagreement between the calculated and measured re- sponse, instrumentation performance and/or

model validity is questionable. Similar compar88
isons can, in principle, be made between calculated and measured response to study the validity of other elements of the composite evaluative

methodology.

Although there is an extensive seismographic

networic onshore in southern California, the cov- erage offshore has been neglected. The knowledge of the uncertainties that exist in using

onshore seismic data for the design of offshore

facilities, the importance of the southern California area for future energy needs, and the high

probability of future large earthquakes in that area contributed to the decision to develop ac- curate and reliable seafloor monitoring insuiiments. These instruments are being installed to form an offshore network to gather data to supplement onshore information. To close the data

loop, offshore structures located near the monitoring sites have been instrumented to obtain

structural response information for comparison

with predicted values.

4. SEMS SEAFLOOR INSTRUMENTATION

In 1977, MMS (then the Conservation Division

of the U.S. Geological Survey), the Department

of Energy, and several major oil companies

contracted with Sandia National Laboratories

for the development of the Seafloor Earthquake

Measurement Systems (SEMS). The objective was to develop a totally self-contained unit that could be installed at remote locations and that would allow the retrieval of data at any time

during the operational lifetime of the system.

The principle design considerations for the

SEMS unit were that it have low power con- sumption, a broad range sensing capability, an

efficient use of data storage, and simplicity of

installation and data retrieval. From 1979 to 1980, four prototype units (referred to as SEMS

I) were placed in the Santa Barbara channel

(Ryerson, 1981). These initial prototype units

only had an operadonal life of less than 1 year. However, the preliminary data and analyses

showed that there may, as postulated, be fundamental differences between onshore and off- shore ground motions during seismic events

(Reece, Ryerson, and McNeil, 1981).

In 1985, a second generation prototype system,

known as SEMS IT, was developed. The project was conducted as a joint effort with Shell Oil

Company, and the unit was installed in 240 feet (72 m) of water 10 miles (16 km) offshore Long

Beach, near Shell's platforms Elly and Ellen in the Beta Field. The SEMS II unit had improved

operating specifications and was designed to have a system life of 4 years. The system re- corded two significant earthquakes during the month of July 1986. The first event, known as the Palm Springs Earthquake (M=6.0), occurred on July 8 and had an epicentral distance of 91

miles (145 km) from the SEMS monitoring site. The second event, the Oceanside Earthquake

(M=5.8), occurred on July 13 and had an epicentral distance of 46 miles (74 km). An analysis

of the data from these two earthquakes indicated

that the seabed responded comparably to on- shore areas in terms of its vibrational frequency

but with a significant decrease in vertical sedi- ment motion as recorded by onshore instrumentation (Sleefe and Engi, 1987).

Both the SEMS I and SEMS n units used a detection algorithm based on the arrival of the vertical component of the earthquake signal.

Because of this technique, the units suffered

from a high false-trigger rate—as many as 1,000

per month. The excessive triggering rate, as well as other software problems, resulted in the SEMS n unit being removed after a 2 1/2-year

service life.

Based on experience gained with the first two

prototypes and using new component and bat- tery technology, a more sensitive and reliable system, dubbed SEMS III, was developed. The

SEMS in, like its predecessors, was designed to operate as an autonomous unit It consists of a seafloor electronics package called the Data

Gathering Subsystem (DAGS) and a shipboard

interrogation unit The DAGS connects to a seismic probe embedded approximately 8 feet (2.4 m) in the seafloor, and it records and stores seismic signals and transfers digital data on command to a shipboard receiver using underwater acoustic telemetry. The surface system,

referred to as the Buoy Repeater System, consists of acoustic transmitting and receiving cx)mponents and a user-friendly computer. The

computer is used for retrieving and analyzing

89!
data from the DAGS and altering operating parameters. The operational concept is shown as SEMS Unit No. 1 in figure 2.

The increased sensitivity of the system results from its new microprocessor software, which

includes an improved multiaxis detection algorithm (Sleefe, 1990). The microprocessor continuously monitors accelerometer signals and

stores data when seafloor motions reach a pre- determined value. The triaxially configured ac- celerometer package (seismic probe) allows it to detect movement in three directions to over- come a major disadvantage of previous prototypes. The previous systems were designed and

built before it was realized that the predominate

seafloor response would be in the horizontal

direction. In addition, the SEMS in unit provides improved lower noise electronics and

lower power consumption to accommodate an

8-year life systenL

5. OFFSHORE SEISMIC NETWORK

In July 1989, a SEMS JE unit was installed in the Beta Field in 2 10 feet (64 m) of water. This

effort was accomplished with the cooperation of

Shell Oil Company. The unit was placed near

the previous location of the SEMS n prototype

as shown in figure 3. This system was designed

to operate as shown in the diagram referred to as SEMS Unit No. 1 in figure 2. The major differ- ence in the installation procedure for the SEMS

in unit, as compared to previous systems, was

that the seismic probe (accelerometer package) was emplaced by means of a vibrocoring drill that vibrated the probe into the seafloor sediment. The soil coupling between the probe and

sediment is greatly improved by embedding the

probe at least 8 feet (2.4 m) below the mud line. A complete detail of the development and instal- lation of the Long Beach unit is given by Sleefe

(1990).

Initial results from the Long Beach SEMS unit

have been very encouraging. The unit has accurately triggered on several small local earthquakes and has provided detailed data for study.

Table 1 provides a listing of earthquakes and

aftershocks recorded during the first 4 months

of 1990.

Figure 4 shows the location of these earthquakes

relative to some of the known major faults. It is clear that the SEMS HI unit has the capability to detect and record earthquakes occurring both

onshore as well as offshore. The quality of the

data is very good as demonstrated by the recording of the Upland California earthquake of February 28, 1990 (M=5.5), as shown in figure 5. A

preliminary analysis of the data recorded thus

far confirms the early findings of a reduction in the vertical seismic response. A statistical analysis has shown that offshore peak vertical accelerations are nearly an order of magnitude

weaker than the corresponding onshore measurements (Sleefe, 1990).

Because of the success of the SEMS HI unit at the Long Beach site, two additional units were

planned to form an offshore data-gathering net- work. This network will provide engineers with

data necessary to verify, and if required, modily

existing response models used in the design of

offshore facilities.

The second unit in the network was installed

offshore Point Pedemales near UNOCAL'S platfomi Irene as shown in figure 6. As a joint

project with UNOCAL, the unit was placed at a water depth of 240 feet (74 m) and was located some 500 feet (150 m) from the base of the steel jacket Unlike the Long Beach unit, the Point

Pedemales instrument is not completely auton- omous, but is hard-wired to the platform as shown in figure 2. The DAGS was located on

the deck of the platform. The accelerometer

probe was emplaced by using a vibrocorer drill in the same manner as with the Long Beach unit. This second unit was connected to the DAGS by

running a weighted cable along the seafloor and

through a specially designed riser to the platform deck. The major advantage for using the

unit in this way is that the DAGS could be

connected directly to a telephone modem and

therefore, interrogated at any time. Also, since

the unit is powered from the platform and, if necessary, with standby batteries, the system

will have a greater extended life span.

90
The platform is also instrumented with a series of accelerometers to measure its seismic re- sponse. Data taken at the base of the platform

(where the response is influenced by the soil- structure interaction) and at the platform deck can be compared with the "free-field" response

measured by the SEMS probe. Analysis of measured data from these three locations, especially

after a moderately large earthquake, will improve knowledge of the way energy is trans- ferred from the seafloor sediments, through the

piles, and into the platform.

To complete the offshore network, future plans

call for the installation of a third SEMS unit

offshore Ventura in the spring of 1991. It is proposed to locate the unit near UNOCAL'S

platform Gilda in 205 feet (62 m) of water as shown in figure 7. It will be hard-wired to the

platform in the same manner as the Point Pedernales unit. However, the exaa type of system

to be used has not been decided at this time.

Consideration is being given to making the unit

totally autonomous and placing it at a deeper

water site, 600- to 8(X)-foot (180 to 240 m) water

depths.

With the installation of the third SEMS unit, an

offshore array will be formed that will be capable of recording all significant earthquakes af- fecting southern California. Figure 8 is a composite for the location of the units, and the

locus of circles shown indicates the sensitivity

of the units to earthquakes of a given magnitude.

This illustration shows a high probability that a small to mediimi earthquake will be recorded by

at least one or two units, and larger events by

possibly all three. The acquisition of additional

seismic data, as well as the monitoring of sediment responses at multiple seafloor sites, will

provide a useful framework to assess current design methodologies. With this information,

which may take several years to obtain and

analyze, engineering efforts can be improved to ensure a safe design within economic reason.

6. CONCLUSION

Seismic activity poses a potential hazard for offshore structures in terms of ground accelerations, rupturing along fault planes and triggering

of large mud slides. Currently, design methods

for offshore structures are based on premises

that ignore differences between the responses of

onshore and offshore sediments. However, pre- liminary analyses of seismic response data taken

from seafloor instruments have indicated several interesting discrepancies. Because of the

environmental sensitivity of offshore oil and gas

operations, issues relating to seafloor response,

as well as platform response, need to be more

fully investigated, both experimentally and analytically. To this end, a reliable and sensitive

SEMS unit has been developed to measure and

collea seismic seafloor responses. These units

are being deployed to form an offshore network

to gather data for comparison with onshore re- sults and to verify current design procedures.

Data of this nature will gready enhance the

ability to understand the response of, and to develop design guidelines for, offshore struc- tures located in seismically active areas.

7. REFERENCES

API 1989. API Recommended Practice

2A(RP2A), Recommended Practice for Planning. Desi gning, and Constructing Fixed Offshore Platforms. 18th Edition, American

Petroleum Institute, Washington, D.C.

Henyey, TJ-., Teng, T.I., McRaney, J.K., and

Manov,D.V. 1979. A seafloor seismic monitoring netwo± around an offshore oil field platform and recording of the August 13, 1978,

Santa Barbara earthquake, OTC 3613, Proceed

ing<;. 11th Annual Offshore Technology Cnnfprence. Houston, Texas.

NAS 1976. Seafloor Engineering: National

Needs and Research Requirements. National

Academy of Sciences, Washington, D.C.

Poulos, H.G. 1988. Marine Geotechnics. Unwin Hyman Ltd., London, England.

Reece, E.W., Ryerson, DJE. and, McNeil, R.L.

1981. Long-term measurements of ground mo91
tions offshore, Proceedings. International Confprpnre on Recent Advances in Geotechnical

Farthquake Hngineering and Soil Dynamics. University of Missouri-Rolla, Rolla, Missouri.

Ryerson, D.E. 1981. Seafloor Earthquake Measurement Sy stems. Volumes I. IT. and TTT. Report

No. SAND 81-1810, Sandia National Laboratories, Albuquerque, New Mexico.

Sleefe, G.E. 1990. The long-term measurement

of strong-motion earthquakes offshore southem

California, OTC 6336, Proceedings. 22nd Annual OffshofR Technology Conference. Houston, Texas.

Sleefe, G.E. and Engi, D. 1987. Seafloor re- sponse for two southem California earthquakes.

Proceeding s. 1987 Society for Experimental

Mechanics Spring Conference on Experimental

Mechanics . Houston, Texas.

Smith, C.E. 1990. Seismic design considerations for offshore oil and gas stmctures, Proceedings of the 21st Joint Meeting of the

U S /Japan Cooperative Program in Namral Resources—Panel on Wind and Seismic Effeas . NIST SP776, National Institute of Standards

and Technology, Gaithersburg, Maryland.
Date Location

Epicentral

Magnitude (Ml Distance (km^

Jan. 17, 1990 Gardena 3.3 35

Feb. 26, 1990 Upland 5.5 74

Feb. 28, 1990 Upland 4.8 74

Mar. 10, 1990 Offshore Newport Beach 2.6 11

Apr. 4, 1990 Offshore Oceanside 4.0 73

Apr. 17, 1990 Upland 4.6 70

Table 1 . Earthquakes recorded by the Long Beach SEMS unit

OFFSHORE PLATFORM RESPONSE TO EARTHQUAKES

MEASURED AND CALCULATED

Figure 1. Methodological basis for offshore platform design

93
94
121° 120° 119° 118° 117° 116° 115°

O =3.0<Mc<4.0

O = 4.0<Mc< 5.0

Q= 5.0<Mc< 6.0

A = SEMS Location

Figure 4. Significant earthquakes recorded by the SEMS III unit

95
Acceleration in milli-g

lU milli-g

Vertical ComponenI

y

- Component

Instrument 80km from Epicenter

Raw Data

10 20 30 40 50 60 70 80

Time in Seconds

Figure 5. SEMS recording of Upland, California, earthquake (2/28/90. M=5.5)

90

Source MMS PAC Sum Figure 6. Location of the Point Pedernales SEMS unit

96
120°45' 120"30' ^20'' 119*30' 119° 118'30' 118°

120^30' 120° 1 19°30' 1 19'' 1 18°30' 1 18°

liMS/ 10/90

Soorce MMS PAC Summary Imlex 1989

Figure 8. SEMS offshore data-gathering network

97

Distribution of Liquefaction in the Saitama Plain

by the 1923 Kanto Earthquake and Geomorphological Maps

Tadao DOHl', Takekazu AKAGIRI

Abstract

The 1923 Kanto earthquake caused

intensive damages in the Kanto district including Tokyo. It caused

geomorphological accidents, which

were collapses, slides and cracks in

many places in the Kanto district.

And so many cracks were caused in the

Saitama lowland in the center of the

Kanto plain. Many houses were

destroyed and agricultural land had

been collapsed.

The authors researched the

aspects of geomorphological accidents

and how to use existing geomorphological maps for prediction of

liquefaction potential. And they con- cluded that the reports described

very precisely and correctly the

aspects of geomorphological accidents

and that the cracks and others were typical liquefaction. And it became

clear that existing geomorphological

maps are also available for prediction of liquefaction potential.

Keywords: Liquefaction peotential.

The 1923 Kanto earthquake. Natural

levee. Land condition map series

Liquefaction

Introduction

The Kanto earthquake, one of the

big earthquake, caused intensive

damages in the Kanto district including Tokyo in 1923. This earthquake

caused geomorphological accidents,

which were collapses, slides and

cracks in many places in the Kanto

district. Among them, so many cracks

were caused in many places in the

Saitama lowland in the central part

of the Kanto plain. And many houses were destroyed and agricultural land

had been collapsed.

Geomorphological mapping is

available to predict liquefaction

potential, but it is not clear how to use existing geomorphological maps

compiled by GSI. National Land Agency

and other organizations.

The authors tried to clarify the

aspects of geomorphological accidents, especially liquefaction and

how to use existing geomorphological

maps for prediction of liquefaction

potential.

1. Method

They considered to clarify the

aspects of geomorphological accidents

caused by the earthquake, especially

on cracks, eruption of sand and water

based on interview, fieldcheck, land

condition maps and reports of the

earthquake.

The interview with farmers was done with information of past

landscapes using former topographic

maps. We selected farmers who were older than 80 years old in good

health, and who had lived there con- tinuously and could remember the

aspects very clearly.

l:Director, Geographical Department,

Geographcal Survey Institute.

Ministry of Construction

2:Head. Second Geographical Division,

ditto

99
They compared with results of

field survey^ interview, land condition maps and the reports, and con- sidered liquefaction potential and

how to use existing geomorphological

maps.

2. The 1923 Kanto Earthquake

The hypocenter of the earthquake

was 35. r N. 139.5° E in the Sagami

bay. The maximum amplitude in Tokyo

was 14-20cm. and the magnitude was

M7. 9. Finally, decease and missing

142,000. complete and partial

destruction of houses 254.000.

destruction by fire 447.000. Geomorphological accidents, which are collapses, slides, cracks, upheaval of

ground and so on, were caused in many

places in the Kanto district. The

ground raised up about 2m in the Boso

peninsula and part of Kanagawa

prefecture. Tokyo and its surroundings subsided. Sea floor of the

Sagami bay raised in the north and

subsided in the south part. Tsunami

hit the coastal area of the Kanto

district about 12m high at Atami.

3. Source map

There are many reports and maps

showing aspects of the earthquake

damages which are building damage,

geomorphological accidents, tsunami,

vibration of the earthquake and so

on, especially, in the reports by the

Shinsai Yobo Chousakai and Geological

Survey. The reports and maps are the

most useful for us to follow geomorphological accidents. In the map,

distribution of cracks is shown. The

aspects were drawn on the topographic

map on the scale of 1:50,000 by GSl

at that time and reduced scale into

1:100.000. According to the map, long

and many cracks were caused in villages facing rivers from north to

south.

3-1. Aspects at Hanada in Koshigawa

town

According to the report, the

outline is shown as in the followings

in this area. In Hanada. many

cracks were caused from east to west

in dry and paddy fields. They were

about 900m long, 2m to 9m wide. Much

water squirted with sand. Sand

deposited on the ground surface 15cm

thick and much sand flowed down into

the former river course thick. Large

scales of crack were 900m long. 60cm

wide and Im deep. And ground surface

waved by sand deposits and many belt

like collaspes. Small cracks were 20m

long, Im wide, Im deep on the dry

fields in many places Sand came

out from the bottom of the well 3m

deep. (Geological Survey : 1924)

.

Interviewing with old peoples,

it was made sure that the results of

interview were the same as the report

by GS(1924) and that much sand flowed

horizontally. And boiled sand was

coarse sand with greyey blue. And it came from the layer under ground

water level. Sand layer of the dry

fields moved and flowed into the

paddy fields in the former river

course to fill it about 3 feet thick.

It took many days to take off sand by

many farmers.

in the dry fields, where sand

erupted out. product of crops was

very poor for 20 to 25 years since

the earthquake and kinds of crops to

grow well were limited to a few

kinds.

3-2. Nakasone in Yoshikawa town

Cracks, sand boils were seen in

Nakasone. Nakasone is located on a belt like landform. facing the river,

where is im to 1.5m higher than the

delta surface, and where the Nakagawa

- river flows down from north to south.

100
And the river runs in the west and

villages are located in the east on the same landforin. Three linear

cracks were caused along the village

with erupted sand. Each of them con- tinues about 1km long north to south

intermittently. But, the width of

cracks were only 15cm.

3-3.Kawafuji in Yoshikawa town

The site is located at the point

where the Furutone river meanders and

point bars are formed by flooding

water. A few long cracks were distributed continuously from the

southern area. In addition to these

long cracks, seven or eight cracks

with several hundreds of meters long were caused in parallel to the point

bars in the slip off slope. The

aspects of damage were also very intensive with cracks, erupted sand and

water in dry fields. And it is clear

that liquefaction was caused very intesnsively in this area.

3-4. Kawakubo in Kasukabe town

Kawakubo is located along the

Furutone river and well developed

natural levees and riverine dunes.

According to the document, the outline is shown in the followings.

In housing area and dry fields

near the Furutone river, cracks were caused intensively with erupting

water, sand, black clay, decayed

tree-tips and pumice gravels. Water

erupted out and reached about one

foot from the ground level and inundated the surroundings about 15cm

deep. House floors were submerged and

sand deposited 4 inches thick. Also

irrigation canal was buried with

sand.

Live fence in the Mr. ISHIKAWA

Kizo' s yard was moved partially by

the earthquake as showing the figure.

This is a typical case of horizontal

movement by liquefaction. The length

of the movement was about 72m long

and moved 1.5m toward east. And the

live fence was changed from straight

line into curved arc-like form. In Moto-shinjuku, hens" egg size gravels

erupted out with much water from the

aquifer in gravel layer 3m deep.

Thus, it is clear that horizontal

movement by liquefaction was caused

in Kawakubo.

3-5. Liquefaction

As mentioned above, based on

field survey of landforms and interview, it was made sure that cracks

and eruption of water and sand were

caused. These phenomena were typical

liquefaction. And these were seen in

many places very clearly which were Kawakubo, Hanada and Kawafuji at

least based on our survey and documents.

4. Geomorphological maps for applied

purposes (Land condition map series)

Land condition map series is one

of thematic maps published by GSI

showing geomorphological classification, distribution of ground

level and facilities rlating to

natural hazards onto one sheet map

with document at another side.

Geomorphological classification is a method to classify ground surface

into landform units based on geomorphological process, form, material

and time when the landform was

formed.

At the beginning of the classification work, the whole area is

classified into large geomorphological units which are mountainous area,

hi 1 ly-and-tableland and lowland. The

large classification unit is classified into smaller landform units.

Small units are, for example, river

terrace, fan, delta, coastal dune.

101
artificially changed landforms and so

on. Detailed units are natural levee,

former river course, back marsh and so many m i c r o 1 a nd f o r m s in the

lowland.

Fluvial proces transports much

material to the lower basin, deposits

material and have formed various

kinds of landforms in the plain.

Natural levee is a iandform which

river floods formed, it consists of

fine sand and silty material, and

slightly higher than the surrounding.

Watching the cross section of natural

levee, it looks like a lens located

onto the delta surface. Usually the

surface of natural levee is rather

dry and is used as dry fields and

villages are distributed on it.

Riverine dunes are formed by aelion

process on natural levee.

An undercut slope is the site

where river water runs at the bank of

the outside of the meander course and

steep cliffs are easily formed, and

erosion is intensive. And slip off

slope is another side, that is. the

inside of the meander. Point bar is

formed there, which is a waving sandy

Iandform.

In Kawafuji and Hanada. the inside of meander consists point bar.

Sandy material deposits very thin in

parallel. The slip off slope slightly

waves gently.

4. Former edition of topographic maps

GSI supplies xerox copies of

former editions of topographic maps

to the public to know pas landscapes.

These maps are out of print and each one copy is kept by GSI.

Topographic maps were used as base maps of the crack distribution

map. The maps are classified into

former edition of topographic maps

in the present. And we caught general

aspects of the surroundings of the

study area around 1923 using former

maps. It is about 70 years since the

earthquake. And landscapes changed

remarkably. For the reason former

edition maps are indispensable to un- derstand content of the interview

correctly. While, former editions of

topographic maps give much historical

information which is difficult to

know using aerial photographs.

In Nakasone. thin natural levee

is formed. Cracks were caused in its

eastern fringe. But, the size of the

present natural levee is larger than

the one around the 1923 earthquake,

because the it had been banked up

using abandoned soil supplied by the

excavation of river course. It is 1 to 1.5m higher than the original

one. And another side of natural

levee facing delta surface was also

banked 2 feet or 3 feet for housing

spot by spot.

Thus, based on former edition of

topographic maps, reports, interview

and field survey, it became clear

that the original natural levees are usually smaller than the present ones

shown in the land condition maps.

6. Characteristics of liquefied sites

shown by geomorphological maps

Many cracks and eruption of

sand and water were seen in the study

area. Based on the research, it was made sure that most of these

phenomena were liquefaction. These

were caused on the natural levees and

point bars and the edge of well

developed sandy landforms. The distribution coincided with past cases which other researchers reported.

But, various long cracks were caused

in undercut slopes in addition to

slip off slopes. Naturally, in banked

landforms. Natural levees cover

saturated soft ground.

Liquefaction can be caused the

102
most easily in the fringe of natural

levees with banked sites and sandy

landforms.

In the classification of natural

levees based on aerial photo-interpretation for the land condition map

series, boundaries of the natural

levees are shown using the most outer

fringe lines of natural levees including sandy banked parts.

Natural levees with banked parts were interpreted together as natural

levees. The fringe of the natural

levee consists of the same material

as the original natural levees in

many cases and it is difficult to

distinguish original ones and banked

zone, if there is nobodey who knows

the landscapes before construction.

Geomorpho 1 og i ca 1 map series

(land condition map series by GSI)

has been established for prediction

of flooding more than thirty years

ago based on the experience of Isewan

typhoon which hit Japan in 1958 and

killed more than 5,000 peoples. And,

the map also can be available to predict liquefaction potential because

both flooding and liquefaction potential depend on the distribution of

m i c ro- 1 andf or ms in the lowland.

Naturally, it is necessary to know

the difference of characteristics between flooding and liquefaction.

8. Conclusions

(l)Many reports on the 1923 Kanto

earthquake show various cracks and

squirt of water and sand. The authors

followed the contents based on interview with old farmers selectively in

the study area. As a result, the con- tents of the reports are correctly

and precisedly described on

earthquake damages. It became clear

that cracks with water were typical

liquefaction based on literature and

field check. Liquefaction was caused

in broad area with sand boils, collapses, horizontal movement of sand

with water.

(2) In the study area, liquefied sites were distributed in the fringing zone

of riverine dunes, dry and paddy

fields in low natural levees and dry

fields in slip off slope. Sandy

micro-landforms are very thin and

look like a weak cover on the

saturated soft delta surface.

(3) In many places, present natural

levees are the results which were

banked up using abandoned soil by the

excavation of the channel around 1923

earthquake onto the original natural

levees. There are many small landform

units which are difficult to interpret using aerial photographs because

of partial artificial change.

(4) Existing geomorpho 1 og i ca 1 maps

(land condition maps) are also available for prediction of liquefaction

potential in addition to specialists'

work. It is necessary to read the

maps tracing the inner boundaries of

natural levees. It is natural that

artificially changed landforms are

easily liquefied.

References

Geological Survey 1 ns t i tu t e (1 82 4 )

:

Kanto Jishin Chousa Houkoku Vol.1

(The Report of Field Survey on the

Kan to-Ear thquake Damage) (GS1924)

Geographical Survey Institute

(1917) :Topographical Map of

KASUKABE on the scale of 1:50,000

Geographical Survey Institute

(1972):Land Condition Map, Noda on the seal e of 1:25, 000

Ministry of Cons turuc t i on (1976):

1:25,000 Simplified Geomo rpho 1 og i

cal Map Series of the Kanto District

Shinsai Yobo Chousaka i (1 92 5 )

:

Shinsai Yobo Chousakai Houkoku

Vol.100. (The Report of the Office

for Prevention of Earthquake

Damage 1925)

103
tsunami m^X^^ ^

seismic intensity Vn rajtsaf^^*

hypocenter

1 9 2 3 ^M^itiiS^Sitlli^EI

M«tt!lMligSfB^^l ^. ttfiHU^Sm : 1924

Index map of the 1923 Kanto earthqualie

(Shinsai Yobo CliousakaiRepor t No. 100 =

Office for Prevention of Earthqualce

Damages. 1925)

Moved live fence by liquefaction

on the sandy landform at Kawakubo

KASUKABE Town. Saitatna pref.

(Geological Survey, J apan. 1924)

Figure 2.

104
I

105
KOSH I GAYA C. )l ^ 7

HANADA

crack

10 0, 000

Figure 4. Distribution map of Cracks in the lowland

in the southern area of Saitama pref.

Geolo.ica. Survey Ins 1

1 tu t e ( 1 92 4 ) El - 4 1^^5,5®% ,C fc |^ S tftS ^

106
107
Figure 6. Topographical map around 1923

paddy field

1:50. 000

dry field

Geographical Survey Institute

( 1917) :Topographi cal Map of

KASUKABE on the scale of 1:50,000

108
(1:25. OOOcolor) x 0.

7

109

Damage to Civil Engineering Facilities From the July 1990 Philippine Earthquake

. by

Yasushi SASAKI*^

[ABSTRACT]

A major earthquake occurred In the

Philippines on July 16, 1990. The northern

part of the Luzon Island suffered from

serious damages. The Japanese Government

dispatched the Technical Expert Team of the

Japan Disaster Relief Team to provide

guidance and advice on emergency

restoration and post-earthquake measures

and "Report of Expert Team of Japan

Disaster Relief Team on the Earthquake In Philippines of July 16. 1990" was finally

compiled by the Team.

This paper summarizes this report and

Includes the data obtained by subsequent

survey. Furthermore, damaged pile

foundation of Magsaysay bridge is analyzed

to evaluate the lateral resistance of pile

foundations and the seismic force.

[KEYWORDS]

Earthquake; Liquefaction; Damage; Road

Side-slopes; Bridge; Pile; River facility; Dam.

1. Introduction

An earthquake of Magnitude 7.8 on Richter

scale hit the central part of Luzon Island

at around 16:26 ( local time) on July 16, 1990. The epicenter of this earthquake was reported to be some 130 km from Manila in the north north east direction and was

located about 40 km from Cabanatuan in the

north east direction.

Considering the severity of the

earthquake damage In the Philippines, the

Japanese Government dispatched the Japan

Disaster Relief ( JDR ) Team ( consisting of

Medical team. Rescue team and a Technical

Expert team) In response to the demand from

the Philippines Government.

The Medical team of 2 medical doctors, 4 nurses, and 3 coordinators was dispatched

to the Philippines on the day after the

earthquake, and returned to Japan on July

24 after cooperating in activities at hospitals In Baguio city. The Rescue team of

11 fire fighters, 11 police officers, and 4 coordinators was dispatched on July 18 and came back on July 26 after carrying out cooperative work in Baguio.

Following to these dispatches, the

Japanese Government also decided to dispatch the Technical Expert team of 4 civil engineers, 2 building engineers, 1 geologist, and 1 coordinator. The Technical

and Michio OKAHARA*=

Expert team departed from Japan on July 29, two weeks after the event, and came back to Japan on August 7 after staying In the

Philippines for 10 days. The team compiled

their findings and recommendations gained

from their field reconnaissance of the

damaged sites into a preliminary report, and

made a briefing of the report to the

engineers of the Philippines government

before they left from Manila.

The Technical Expert team members drafted

the final report soon after their return to Japan. The final report of 360 pages, which

Is written in English, contains a detailed

description about the types and

distribution of damage, the urgent

temporary repair methods of the damaged

infrastructures, and also . the mid- and

long-term rehabilitation works and

countermeasures against further disasters.

In order to hold the lecturing seminar on

the report on March 7, 1991, the Japanese

government again dispatched three experts

Including one of the authors of this paper

to the Philippines. These experts again

visited the severely damaged Baguio and

Dagupan areas for two days before the

seminar.

This paper summaries the final report of

the JDR Technical Expert team and mentions

briefly the current status of the

rehabilitation works of the damaged civil

engineering structures. The damage and

repair works of the building structures are

to be written in a separate paper.

2. Outline of the earthquake characteristics

and the damage

Seismic Activity in Philippine

The Republic of the Philippines and Japan

are situated In the Clrcum-Paclflc Seismic

zone. Both countries suffer from Intense

seismic activity and have experienced

severe damage due to a large number of

destructive earthquakes.

The earthquakes in the Philippines are

classified Into (1) earthquakes caused by

subductlon of plates at such deep sea

• 1 Director, Earthquake Disaster Prevention

Department. P. W.R.I. , Ministry of

Construction

• 2 Head. Foundation Engineering Division,

Structure and Bridge Department, P.W.R.I., Ministry of Construction

111
trenches as the Philippine trench, the East Luzon trench and the Manila trench, and (2) earthquakes caused by dislocation of active

faults such as the Philippine Fault.

These trenches and the major active faults are shown In Flgure-2. 1. Notable destructive earthquakes In recent years are shown In Table 2.1. The earthquake

damage In Mindanao Is mostly caused by

tsunamis. The Ragay Earthquake of magnitude

7.0 In southern Luzon in 1973 had

accompanied left-lateral strlke-sllp

displacement as much as 3.2m on the ground

surface ( Allen, 1975).

Large earthquakes of magnitude greater

than 6.8 have occurred repeatedly along the

Philippine fault, even In 20th century.

Flgure-2.2 shows the epicenters of these

earthquakes along the Philippine fault and

the lengths of their causative faults

estimated from the earthquake magnitudes. As shown In this figure, a destructive

earthquake has not occurred In the central

part of Luzon Lsland, where the earthquake

of July 16, 1990 took place, since 19th

century. The shaded areas In this figure are

the seismic gap along the Philippine fault ( Nakata et.al ).

Although the Philippines has very high

seismic activity and experienced many

destructive earthquakes repeatedly In the

past, only 15 seismic observation stations are operated by the Philippine Institute of

Volcanology and Seismology (PHIVOLCS).

Furthermore, data from these observation

stations are not yet centralized enough to gain sufficient accuracy for the

determination of the location of epicenters.

Strong motion observation, which Is Indispensable for developing seismic design

codes for structures. Is not being carried out sufficiently. Five strong motion

accelerometers had been Installed In the

Philippines but none of them was In operation at the time of the earthquake

because there was a lack of recording

paper. Consequently, no strong motion

record was obtained for this earthquake.

The Earthquake and the Distribution of

Seismic Intensity

The parameters of this event are as follows: Time and Date; 16:26, July 16, 1990 ( Local

time)

Magnitude; 7.8 In RIchter scale ( by USGS)

Epicenter: 15.658' N, 121.227- E ( by USGS)

Focal Depth; 25km (by USGS)

The distribution of seismic Intensity In Rossl-Forrel scale Is reported by PHIVOLCS

as shown In FIgure-2.3.

The Causative Fault

The epicenter of this earthquake Is located on the Philippine fault, and some

120km long segment of this fault Is considered to be broken by this event. The

maximum lateral gap which appeared on the

ground surface was about 5 meters. The northernmost tip of the fault

movement Is confirmed to be at Cayapa

village near Imugan In a mountainous area, which Is east of Bagulo, and the

southernmost was around Bantug. The

epicenter was located one third of the

broken fault length from the southernmost

point. An epicenter Is the origin of the fault

breakage, thus It Is considered that the

earthquake shaking was caused by the fault

rupturing 80 km northwards and 40 km

southwards along the Philippine fault. The fault movement traced along the fault

is shown in Flgure-2.4 ( Nakata et.al).

Outline of Damage

This earthquake Is firstly characterized

by a wide variety of damage over a wide

area. The damage by this earthquake Is mostly

distributed In a triangular area surrounded

by Bagulo, Dagupan and San Jose. This area

is In the north west side of the fault. The second characteristic is that there

was little damage near the epicenter while

the damage in areas far from the epicenter,

such as Bagulo and Dagupan was more

serious. The third characteristic showed

that the damage can be classified into two

categories; 1) structural damage by strong

ground shaking, 2) geotechnical damage.

Of these two types damage, 2) geotechnical

damage Includes both a) settlement of

buildings, collapse of bridges, and the

damage to pavement due to the soil liquefaction, and b) damage to roads In mountainous areas due to road-side slope

failures.

The largest number of casualties and

Injured were reported In Bagulo City in Benguet Province. Most were caused by the

collapse of medium size to high rise

buildings such as hotels. Also, collapse

and fire which occurred simultaneously at the Export Product Area caused a large loss

of lives. A large number of passengers on a

bus passing on the Kennon Road were

trapped In landslides and reported to be

dead.

Nuevavizcaya Province Is located in the

east of Benguet Province where numerous

landslides occurred on steep slopes in mountainous regions.

Most of the casualties In Cabanatuan City were caused by the collapse of the

112
Christian College. Collapse of a building In the public market was also reported.

Although this city was close to the

epicenter. other damage was not

substantial.

In the cities along the Llngayen Gulf,

Including Dagupan city, severe damage took

place due to soil liquefaction. In Dagupan

City, the ground surface sunk about 1 to

1.5m over a wide area, and a large number of

buildings settled and/or Inclined.

Damaged bridges were found over a wide

area. From the 39 bridges examined during

the field reconnaissance by the JDR

Technical Expert team, 28 bridges were

evaluated to be damaged from a view point

of load bearing capacity In the following

ways; a) Fallen brIdges-6 (Including 1 temporary bridge), b) Seriously damaged-6,

c) Partially damaged-8, and d) Slightly

damaged-8.

In addition, failure to embankments near

bridge approaches were commonly observed

even though their strength were not

affected.

Damage to roads was observed at many

places, especially on the Japan Philippines

Friendship Highway which Is the main route

In the northern part of Luzon Island, as well as on the Kennon Road, the Marcos

Highway and the NaguUIan Road to Bagulo.

Among these roads only the Nagulllan road

was made passable for one way traffic after

the earthquake. Others were left closed even though Intense restoration work was In progress at the time of reconnaissance,

resulting In serious obstacles to dally life as well as to rescue and restoration

activities. The earthquake occurred In the

rainy season, so there weis a high

possibility of secondary disasters, such as slope failures, even after the Immediate

restoration work had been done.

River facilities were damaged over a long

distance, especially In the Agno River

system, and sinking, cracking and breakage

of levees, and damage to parapet and

revetment structures was observed.

Although there was little damage along the

Panpanga river, severe damage to the levee was observed on the west side of Mt. Arayat.

The damage to dams was also serious.

Cracking, sinking and sliding at top of the

embankments were observed at the Anbuklao

Dam and BInga Dam In the Agno River System,

and the Pantabangan Dam and the Maslway

Dam of the Panpanga River System.

Restoration Works

According to the final report by the

Philippine government, the damage

statistics are as follows: casualties 1,642;

Injured 1.495; missing 178; totally collapsed

houses 16,467; partly collapsed houses

26,646; total amount of damage to Infrastructures 14 billion pesos ( about 70 billion yen, about 500 million dollars)

(Ikeda).

By referring to the report by the JDR

Technical Expert team, the report by the

Asian Development Bank and the report by

the World Bank, the Philippine government

Is executing the restoration works in the

following levels: Level I ; urgent restoration for

recovering the function ( temporary

bridges, detour construction).

Level II ; restoration for recovering the

pre-earthquake functions ( repair and

retrofit), to be completed In two years. Level 111 ; restoration Including

strengthening against future possible big

earthquakes, to be completed In more than

two years.

3. Soil liquefaction

Damage of a wide variety of structures due to soil liquefaction was observed over

a wide area, especially In the coast area along the Llngayen gulf.

Failure of river levees, collapse of bridge

substructures and resulting falling off of

girders, settlement of highway embankments,

tilting and settlement of buildings, lifting up of burled water supply pipes, and

floating up of burled tanks at gasolene

stations were seen. Much of the typical damage was observed

In Dagupan city. Dagupan city with a population of 110 thousand people runs along the both sides of the Pantal river. Many reinforced concrete buildings were

damaged by tilting and sinking without

having major structural failures. Road

embankments, even though which were as low

as 1.5-2 m, settled at several sites and

failures of pavements were observed near

Dagupan. The Magsaysay bridge, which Is a key means of connecting the parts of the

city on either side of the Pantal river,

failed. The girders fell down because the

abutment moved laterally when liquefaction

induced lateral spreading of the ground.

In southern part of the city, where many

low rise dwelling houses are located, 1.5-2 m of settlement of the ground surface took

place which caused difficulty with

drainage.

FIgures-3.1 and 3.2 show the damaged area

of Dagupan city. The area shown consists of

thick alluvium deposits having loose sand

layers and sUty sand layers up to 7-10 m below the ground surface. The ancient

113
river channels are shown In FIgure-3.2. The

area where damage to structures was heavy

Is located on the old river channel, and

Inside the area surrounded by winding course of existing river.

The Department of Public Works and

Highways (DPWH) of the Philippine

government conducted standard penetration

tests at many sites around Dagupan city in order to Investigate the soli conditions of

the damaged area after the earthquake. An

example of that SPT results Is shown on

Flgure-5.5. The SPT method conducted In the

Philippines seems to be slightly different

from that In Japan as they do not use bentonlte suspension to protect the bore

hole, the rope to pull up the drop weight Is controlled by drum. The measured blow

count number needs to be modified to account for the energy transmitted.

The maximum ground acceleration was

estimated to be around 200 gals from the

empirical attenuation formula In Japan. The

liquefaction potential assessed by the

method in Japan Highway Specification

utilizing the estimated ground motion and

the measured blow count is also shown In Flgure-5.5. From this Figure. It Is considered that the layers up to around

7-10 meters below the ground surface may be

liquefied.

At the time of the lecturing seminar In March, 1991, they were Intensely conducting

restoring work of the damaged drainage

system and streets In Dagupan city, and of

the Magsaysay bridge.

The damaged Magsaysay bridge was a seven

span simply supported bridge resting on

pile-bent foundations. The foundation type

of the new bridge was a cast-ln-place large

diameter pile foundation. The existing old

piles are obstacles for constructing the new cast-ln-place plies, so the old piles for

piers p-4 and p-5 were pulled out from the

ground. The piles for pier p-5 were cracked

and failed In bending at around 10 meter below the ground surface, and the upper

portion of the plies were bent 'towards the

river side. From the depth of the cracks on

these piles, the thickness of the liquefied

layer was estimated to be around 10 meter below the ground surface. The piles for p-5 were constructed in single row whereas the

piles for p-4 were constructed In three

rows, all of which were undamaged.

4. Damage to Road-Side Slopes

Status of Damage to the Japan Philippine

Friendship Highway

Damage to the Japan Philippine Friendship

Highway by the earthquake was serious,

covering a 76 km section which runs through

mountainous section between San Jose (kms

160) and Aritao (kms 236).

Fig. 4.1 shows an outline of the damaged

section. In the Bayombong district (km 265)

north of Aritao a few cracks occurred on

the road shoulders only, while on the road

between San Jose and Aritao, the traffic was shut down due to slope failure of the road.

Since large-sized fault displacements were produced along the Highway In the

neighborhood of Dlgdlg (kms 183) it Is thought that the strong shaking of the

earthquake motion was centered around the

mountainous areas. According to investigation by a helicopter, in addition to slope failure

along the Japan Philippine Friendship

Highway, slope failure on surrounding

mountainsides was observed; the scope of

this slope failure spreads to Bagulo and Its surroundings.

The most prominent feature of the

earthquake damage along the Highway was

that a great number of continuous slope

failures occurred not only on the

artificially cut slopes but also on the

natural slopes. However, the slope failures

all had shallow surface sliding.

These districts have the regional

characteristics that there are few trees on

the surface of the mountains and no slope

protection measures are used. Surface

layer of slope which is weathered due to the

heavy rains and the high tropical

temperatures and barely balance against

the static force of rainwater erosion was

easily slid when subjected to strong

vibrations.

It may be noted that the annual rainfall

in Santa Fe (kms 215) on the north side of

Dalton Pass (kms 208, 1000 meters above sea

level) and San Jose to the south of the Pass from 1976 to 1984 is that, 1500 to 3400 mm In Santa Fe and 1700 to 3200 mm in San Jose. Sometimes these districts are hit by heavy

rains accompanyed by the progress of

typhoons.

Since the Highway runs along the fault

topography with a NW-SE direction

branching off from the Philippine fault, the

geological features there are disturbed and

changed in quality by its influence.

The geology is composed of the following:

the basement is granite and diorlte of the

Mesozolc era. Cretaceous period; diabase

and andeslte of the Paleogene period to Neogene period; and sandstone, mudstone

and limestone of the Neogene. The basement

rock is covered by terrace gravel of

Dlllvlum period In Quarternary, alluvium

sand, gravel and clay of Quarternary

114
period; then the surface geology Is composed of talus sediment and the present

riverbed sand and gravel.

The basement rock is in an unstable state having clayey seams caused by fault

activity. Furthermore, since the granite

mainly distributed in the Santa Fe neighborhood Is decomposed and so unstable

that It can easily cause the slope failures.

Slope failure hardly occurred where there were slopes composed of hard rock, low

slopes (less than 10 m) and gentle grade

natural slopes. Furthermore, compared with

the many slope failures, there was little damage to the structures, such as stone masonry retaining walls and gabions which

reinforce the lower parts of slope faces. One of the features of the disaster is that

damage was hardly observed on the structures with a base rigidly connected

with the sound natural ground.

Some pavements were damaged. Cracks were

caused by settlement of the embankment

shoulders, there was damage from falling stones caused by the earthquake motion and

thrusting up of pavement slabs due to fault

displacement. At the places where faults cross the road, the whole road surface slid

laterally (about 2 m to the axis line) and

the pavement slabs were badly damaged.

However, it was not enough to stop the

traffic.

In addition to the damage to the road

Itself due to the afore-mentioned collapse

of slops, there were outflows of huge

amounts of muddy sands which caused a large

amount of damage from flowing mud and

drifting wood in the low-lying areas

(rivers) of the region.

Table 4.1 shows the types and distribution

of damage along tiie Highway from visual

observation. The number of the places with

slope failure reached more than 200 and

totalled about 12 km of the Highway. Most

of the slope failures occurred continuously

along the direction of the road.

Status of Damage to The Three Main Routes

to Baguio

Though the Naguillan Road could be passed

through, some sections still had only one

passable lane because of the large quantity

of collapsed earth and sand from the failed

slopes and the settlement of road surfaces on banked sections after the earthquake.

In the areas of KM267+700 and KM261+000,

collapsed earth and rocks of more than

5,000m^ remained. The total amount of

collapseOd earth and rock was estimaed to exceed SO.OOOm^.

It was difficult to remove this large

quantity of collapsed earth, rocks and huge

fallen rocks, and addlning to this difficulty

rehabilitation work was prevented by

rainfall. The section between KM272+000 and

KM277+500 of the Marcos Highway was closed

until September 1990 because It took a long

time to remove the collapsed soil. Although

the features of the damage due to slope

collapse on the Marcos Highway were similar

to those on the NaugUIan Road, the scale of

the slope failures was much larger than

that on the Naguillan Road. The obstacles

which cover the inaccessible sections

exceeded 164,000m^.

The Kennon Road, one of the three main routes leading to Baguio city, was the most extremely damaged (Fig. 4. 2). The damage to the Kennon Road differs

from the Marcos Highway, in the point that

the large quantity of debris from the slopes

along the road Included huge fallen rocks.

The classification of slope failures In Kennon Road is shown in Table 4.2.

The mountain slopes on both sides of the

Buedo river failed during the shaking and

supplied a large quantity of debris which

raised the bed of the Buedo river. When the

rehabilitation plan for Kennon Road Is formulated, the partial realignment of the

road must be considered in sections

beginning with the locations where the

former road is susceptible to be flooded

due to the raised river bed.

On every route, even when the accumulated

earth was removed was susceptible to newly-fallen debris and further erosion of

the shoulders which may further damage to the functions of the roads due to the heavy

rain. Therefore, measures to provide

sufficient drainage facilities are necessary.

5. Damage to bridges

1) Status of Damage to Bridges and Damage

Analysis

Damaged bridges ranged so widely from

Manila to the northern part of Luzon Island

that a field survey on all of them in a limited period was hardly possible.

Therefore, the team decided on a survey

policy which would cover as many sites as possible and check on bridges heavily

damaged.

In fact, field surveys on a total of 30 bridges were made. The team also checked

nine damaged bridges on the Japan

Philippine Friendship Highway by hearings

from the local engineer as well as by

looking at photographs which had been

taken in previous surveys.

115
According to the DPWH damage survey

results as of Aug. 1990, the surveys had not been conducted yet on 25 bridges. However,

these apparently did not Include bridges

with serious damage, such as fallen bridge.

Fig. 5.1 shows the locations of damaged

bridges, 79 bridges. Including the

additional results of subsequent Inspection.

Bridge damages extended to large area of

Luzon Island, including many bridges

particularly on the Japan Philippine

Friendship Highway and the Manila-North

Road. These are trunl< roads linking the

northern part of Luzon Island to Manila.

On the Manila-North Road linking Manila

and San Fernando via Tarlac and on roads in Its vicinities, the Carmen, Calvo and Tabora

Bridges fell and the Arlngay Bridge was

seriously damaged.

In Dagupan City and Its suburbs where

heavy damage was caused by liquefaction,

the Magsaysay Bridge fell and the

Embarcadero and Rabon Bridges were heavily

damaged.

On the Japan Phllipine Friendship Highway,

the Manicla Bridge fell and the Slcslcan

Bridge was In danger of falling. As for the

Puncan II Bridge, no abnormalltly was discovered In its structure but it seemed

that it could have been damaged by the

oncoming large quantity of driftwood.

Table 5.1 shows data, status of damage and

damage sketches concerning major damaged

bridges . Damage characteristics include falling off

of the superstructure due to pier

Inclination caused by the liquefaction of

ground, damage of the girder ends, shoes,

and bridge seats caused by Inertlal force,

damage of the abutment piles, footings, and

girder ends caused by earth pressure and

inertlal force, damages of the pile heads in pile bent piers, settlement of piers, and

settlement of approaches.

Of these damage types, severe damage by

liquefaction, damage to pile bent abutments

and piers and damage to shoes and bridge

seats were remarkable.

Mention must first be made of the fact that

the horizontal bearing force of damaged

bridges In general was small because their

foundations generally had insufficient

setting depth (presumed to be about 10 m, at

most). The single-row pile form (one row in the direction at right angles to the bridge

axis) was common (pile bent) and the

commonly used piles were made of wood or

RC. Their shoes were relatively simple structures and did not have a sufficient

number of anchor bolts. The bridge seat on

which the shoes rested was narrow so that

there was a danger of bridge fall or easy

slip-out of the shoe in the event of a large

displacement of the substructure. Also,

shear fracture was liable to occur in the

bearing edge at the pier top and in the

girder end.

The degree of damage had to be judged in order to carry out emergency

rehabilitation. The team decided upon

assessing the of damage degree for

Individual bridges In accordance with

Manual for Repair Methods of Civil

Engineering Structures Damaged by

Earthquakes (Ministry of Construction,

March 1986). The bridges of major damage

levels, D and E were shown In Fig. 5.1. The

degrees of damage concerning bearing

capacity were decided according to the

following five ranks:

A: No damage There is no known

abnormality concerning

bearing capacity.

B: Minor damage There is no possibility

of immediate decline of

bearing capacity.

C: Medium damage .. The damage may lower

the bearing capacity,

but If it does not worsen from

aftershocks, live load,

etc., the bridge can be

used tentatively for

the traffic.

D: Major damage ... The damage seriously

affects the bearing

capacity and may

result critical

consequences, such as a bridge fall.

E: Bridge fall The case of a fallen

bridge

In deciding upon degree of damage

concerning bearing capacity, It Is most Important to determine whether or not a bridge fall will eventually occur. Here,

bridges with the possibilities of a fall due

to live loads, aftershocks, etc. are

classified as being in the D rank. As a whole, these bridges must be totally closed

to traffic.

2) Seismic Design for Bridges

According to the recent "Design Guidelines

Criteria and Standards by DPWH 1989 (hereafter referred to as the Guideline),

aselsmlc design Is to be applied to civil

engineering structures such as dams, tanks,

and bridges. Detailed procedures and

methods of seismic design are not shown In this Guidllne.

Detailed information on bridges is given in the "National Structural Code of the

Philippines, Vol.11, Bridges by the

Association of Structural Engineers of the

116
Philippines, 1987" (hereafter referred to as the Code). This paragraph describes the

seismic design of bridges,

History of Seismic Design for Bridges

The seismic design of bridges before 1972 was not clear.

Seismic design was performed on the basis

of the guldenne issued In 1972. It was

specified that 10% of the dead load must be

considered as the minimum seismic effect. According to the guideline (1972), it was

specified that the actual seismic design Is to be performed on the basis of "Earthquake

Engineering for the ILZGAN-BUTAN ROAD In the Island of Mindanao-Philippines by the

Asian Development Bank" Issued In 1972.

From 1989, It was specified that 10% (dead

load + 1/2 live load) Is to be considered as the design earthquake force In the

Guideline above which was revised from the

1972 guideline.

Actual seismic design was done on the

basis of the aforementioned Code Issued in 1987.

Seismic Design Method for Bridges (as per

the Code In 1987)

Seismic design for those areas expecting

earthquake occurrence Is to be done by

considering the relation to active faults of

the construction sites, response

characteristics of the ground and the

bridges, etc.

Both the equivalent static load method

(seismic intensity method) and the response

spectrum method are shown in the Code as the practical calculation methods. The

former is specified as follows:

For almost equal underground layer

rigidity, design earthquake force (EQ) Is calculated using the following equation.

EQ = C • F • W

Where, F = Compensation factor for each

structure (1.0: single column type

piers, 0.8: Rigid-frame piers) W = Dead load C = A combined response factor which

is greater than 0.1

Value C Is obtained using the maximum

acceleration (A) of the foundation,

thickness of alluvlnm layer to baserock,

and the natural period (T) of structures.

If the aforementioned data (maximum

acceleration of foundation, etc.) are unknown, the following equation can be used

to calculate EQ.

EQ = 0.10 (W + L/2)

Where, W = Dead load, and L = Live load,

The calculation for complicated

structures is to be made by the response

spectrum method.

Load for the stopper of girder ends at movable shoes. Is as follows:

EQ = 0.25 X DL - (shearing force generated

on a pier due to earthquake force).

Where, DL = Reaction force on the

superstructure to be considered for design

However, there seems to be some

uncertainty for the practical application of

the above guideline.

3) Calculation of Lateral Resistance of Pile

Foundations and The Seismic Force for

Magsaysay Bridge

Fig. 5.3 and 5.4 show the fallen old

Magsaysay Bridge and the newly planned

Magsaysay Bridge respectively.

Calculations were executed in order to estimate earthquake force which actually

acted, the level of liquefaction and lateral

resistances of pile foundations.

Since there Is a shortage of precise data on

pile foundations, pile cross section, and

soil properties, estimated data were used

in the calculations.

Conditions and Methods of Calculations.

® The soil conditions and models shown in Fig. 5.5 were used for the calculation. The

soil parameters were obtained at the site

away from the bridge, because the soil condition was unknown at the bridge site. Both models for piers P4 and P5 are assumed to have the same soil condition . The calculation models shown in Fig. 5.

5

used the survey results obtained after the

earthquake.

® The horizontal ground reaction

coefficients are assumed to be from KH=1.0

to 10.0 Kgf/cm3( Kh = 10 kgf/cm3 is equivalent to SPT N value of 10 ). Calculations were also carried out assuming

outstanding piles saused by liquefaction occurrence.

(D The ultimate bending moment capacities

of piles were obtained assuming RC plies

with 40 cm X 40 cm cross section and 3 types of reinforcing steels.

(3) Maximum bending moments (Mm) of a pile

and the displacements ( 5 T) at the pier tops were obtained and Illustrated when

earthquake force (Ho) was applied. The

seismic Intensity Kh was defined as the ratio

of horizontal force to dead load.

Results of Calculation

117
® Regarding to the pier P5, results of

calculation are Illustrated In Fig. 5.6. It can be concluded that the piles would be

broken If seismic force equivalent to more

than Kh = 0.1 acted even without the occurrence of liquefaction. The actual piles were failed In bending due to liquefaction.

(2) Piles for the pier P4 would be broken If deep range of ground were liquefied. There

was little damage to the actual piles as mentioned In the previous chapter.

6. Damage to rivers and dams

1) Damage to Rivers

At the Agno river system in Fig. 6.1, the team observed damaged river facilities on a wide scale. Cracks at the crests of the

levees and settlement and sliding of the

levees were marked and some of them had

collapsed. Furthermore, cracks and

collapsing were observed at the parapet and

revetment structurs. There were few damaged sites at the

Pampanga river system, however, a collapse

of the levee near Mt. Arayat was particularly notable.

No significant damage was seen at the

gates, which seemed to be operated

smoothly.

• The situation of the damage at the Agno

river site was as follows.

The Tarlac River. A Branch of The Agno

River (Panigui Site)

Large cracks occurred at the crowns of

levee of 5 - 6m in height, and sliding

occurred as shown In Fig. 6. 2. The material

of the levee was mainly composed of fine

sands. Some places were temporarily

rehabilitated with sand-bag piling.

The revetments, made of plain concrete

using boulders, were dameged by sliding

with the collapse of levees in such a way

that the foundations were laid directly on

the natural ground. Although these

revetments were badly damaged, the revetments. In which foot protection on the

river frontage had been conducted by the

placing of boulders, suffered little or no

damage.

The Agno River (Vlllasis to Bayambang

Sites)

Many cracks were observed at the crests of levees of 5 to Gm height. Some levees

sank about Im as a whole, or, some of the

levees slid and collapsed. Sliding occurred

towards the direction in which the ground

height, Inside or outside of the dike, was

relatively lower. The main material of the

levees seemed to be silt.

The Agno River (Bayambang to Urbiztondo

Sites)

Several oblique cracks on the crowns of

levees of about 5m in height, and big

collapses caused by sliding were observed.

According to a housewife who saw the

collapse at Its actual location, serious

horizontal shaking occurred, cracks

opened, and then the earth dikes slid

(Flg.6.3b).

Both sides of the earth dikes were marsh

and the ground conditions were very bad.

2) Damage to Dams

The team investigated the following dams;

the Ambuklao Dam and the Binga Dam of the

Agno river system, and the Pantabangan Dam

complex (composed of the Pantabangan Dam,

the Aya Dam and the Masiway Dam) and the

Angat Dam of the Panpanga river system.

All the dams except for the Masiway Dam

were embankment dams. Cracking and settlement of the crests of

the dams and shallow sliding on the slopes

of the dams were observed mostly in the

Ambukulao Dam (Fig. 6.4), the Binga Dam (Fig.

G.5) and the Masiway Dam. Since all the dams had become old and

maintenance on them seemed to be

Insufficient, a lot of damage was also

observed which was unrelated to the

earthquake.

7. Conclusion

As described previously in Chapter 2, the

characteristic features of the seismic

damage this time are: 1) that the damage was spread over a wide range, 2) that the types

of damage suffered were diverse and 3) that

about 5m lateral dislocation along the

Philippine fault appeared on the ground

surface over a long distance (confirmed at

least approx. 120km). The main affected area was roughly within a trianglar area

with one side approx. 50-60km connecting

Bagulo, Dagupan and San Jose, with damage

found scattered to Cabanatuan City and

colonies along the Japan Philippine

Friendship Highway.

The types of damage are roughly divided

into 2 types i.e. 1) collapse of structures by shaking and 2) ground failure. The

ground failure Included: a) sinking and

tilting of buildings and damage to Infrastructures such as bridges, due to soli

liquefaction as well as damage to pavements

and b) damage to roads by land slides

centered around In the mountainous area.

In the " Report of Expert Team of Japan

Disaster Relief Team on the Earthquake in Philippines of July 16 ,1990 " the technical

recommendations, such as items desired to

118
be urgently acted on. Items for which quick

action Is desired to be taken and Items for

which actions should desirably be taken over the long time period have been made

based on field reconnaissance.

Among these Items, Items to be acted on urgently for civil engineering structures

are briefly cited here. Analysis of damaged

pile foundation for Magsaysay Bridge Is also

mentioned here.

Items Desired to Be Urgently Acted on

(1) DIagnosls, repair and reinforcement of

bridges suffering from the disaster

Trunk roads connecting the northern part

of Luzon to Manila Include 2 routes I.e., the

Japan Philippine Friendship Highway and the

Manila North Road. In the former, some

Impassable zones still remain due to the

existence of a large quantity of debris,

In the latter, because the Carmen Bridge

collapsed, a bypass through LIngayen has

been provided.

Many bridges suffered from the disaster

along the Manila North Road, which Is left as only one road. Including this bypass

road, connecting the northern region of

Luzon to Manila and Its peripheral roads.

There are even some of these bridges which

are susceptible to the progressive status of

damage under the load of traffic of heavy

vehicles. Considering this road's

Importance as a trunk road. It goes without

saying that the diagnosis on the degree of

damage and repair and reinforcement of

damaged bridges are necessary for the rapid

recovery of transportation means. As a result of field surveys on 39 bridges

at this time, 12 bridges were found to have

suffered high degrees of damage. Including

6 bridges which fell down. Practical

proposals on Immediate countermeasures and

a full-scale recovery plan were offered In the tentative report for 10 of these

bridges.

Of the remaining bridges suffering from

the disaster which were not diagnosed this

time because of time restrictions, diagnosis

should be conducted soon.

(2) SettIng of urgency bypasses and erection

of emergency temporarily erected bridges

The diagnosis, repair and reinforcement

described In the previous Item are one set of countermeasures to be urgently taken,

but substantial time will be required before

their completion. As a stop-gap measure to be taken In the meantime, setting urgency

bypasses and erecting emergency temporary

bridges are necessary.

Such actions have already been taken to some degree, and It looks as If emergency

recovery has become efficient to some

extent, but further treatment of the whole

range of damaged roads and of all damaged

bridges will be necessary.

(3) Measures for prevention of drifting of

bridges,, etc., by flowing sediment and logs

At the Puncan n Bridge on the Japan

Philippine Friendship Highway, danger of

the bridge being set adrift by running

sediment and logs was observed.

At the Carmen, Calvo, Manlcla and

Magsaysay Bridges which fell down, girders

which fell down have been left as they are, Impeding the passage of flowing water

during floods. Some measure to remove them

quickly needs to be taken.

(4) Early opening of 3 routes to Bagulo

Among the 3 main routes to Bagulo, only

the NaguUlan Road has been opened.

However, along this route, substantial

lengths of road are left where only one-way

traffic Is available because of large scale

collapses and potential disasters resulting

from rain fall. It should be an urgent

priority to hasten the opening of the

Marcos Highway.

The remaining 2 routes, which have

relatively short Impassable sections, can be

quickly repaired providing at least 2 routes which lead to Bagulo.

For this purpose, larger numbers of heavy

duty machines for removal of collapsed

sediment should be provided.

In the opened road. It Is desired to keep

watching and to control traffic volume to prevent a secondary disaster from

occurring.

(5) Emergency recovery of river embankments

and dams

Early execution of emergency recovery of

the damaged parts of levees and dams which

suffered from the disaster Is necessary.

This Is because the rainy season has set In. Furthermore, after the season Is over, full-scale permanent recovery work needs

to be done.

(6) Study for acquiring basic data for future

recovery plans

In proceeding with the above-described

countermeasures, records on the planned

processes and methods of their execution,

etc., should be left, such that they will serve not only as mere records of recovery

of this time's disaster but as data for

reference In planning countermeasures for

the future.

Analysis of The Damaged Bridge of

Magsaysay

From analysis of the pile foundations of

fallen Magsaysay Bridge, It Is clearly seen

that there Is no consideration of

liquefaction In design and that the seismic

Intensity Is considered to be at most 10%. In

119
conclusion the horizontal bearing capacity

of a pile foundation Is Insufficient to cope

with liquefaction. Furthermore the pile bent

foundation Itself Is very flexible,

therefore. It can be said the bridges are susceptible to fall during a strong

earthquake.

8. Acknowledgement

This paper Is prepared mainly based on

"Report of Expert Team of Japan Disaster

Relief Team on the Earthquake In Philippines

of July 16, 1990", which was authored by the

Technical Expert team of seven members

headed by Prof. T. Okada of Tokyo

University with lots of assistance and

support provided by the DPWH, DOTC. NHA,

Armed Forces of the Philippines. PHILVOCS,

local authorities of the Philippines

Government, the Japanese consulting firms,

and JICA experts. The manuscript of this

paper was typed by Mr. S.Takagl of the

Foundation Division, PWRI and English was

reviewed by Dr. G.McRay of Earthquake

Engineering Division, PWRI. The authors greatly appreciate these

contributions given by all Individuals

above.

Reference

1) Allen, C.R.: Geological criteria for

evaluating selsmlclty. Geol.Soc.Amer.Bull.

vol.86, PP1041 ~ 1057, 1975

2) DPWH : Designing Materlals(Magsaysay

Br. etc)

3) JICA : Report of Expert Team of Japan

Disaster Relief Team (JDR) on the Earthquake

In Philippines of July 16, 1990.

4) A.Shlmazu, Y.Sasaki and M.Okahara : Reconnaissance Report on the Philippines

Earthquake of July 16, 1990. Civil

Engineering Journal, VOL.32-10, 1990.

5) T.Nakata, et al. : Earthquake Fault of

1990 Philippines. Journal of Geology, 99-5,

1990.

6) Y.Sasaki : Recrinalsance Report on the

Philippines Earthquake of July 16, 1990,

Tsuchl-to-klso, 38-11, Nov. 1990.

7) M.Okahara and M.Mltanl : Study Report on

Damages by the Philippines' Earthquakes,

Journal of Foundation Eng., 18-12, Dec. 1990.

8) T. Ikeda : Earthquake of Luzon Island and

Aids of Restoration, Monthly Journal of

Construction, Jan. 1991.

9) Y.Sasaki : Characteristics of Disaster by

Recent Overseas Earthquakes, Technical

Memorandum of PWRI. VOL. 2928. Feb. 1991.

10) Nippon Koel : Kennon Road Disaster

Prevention Project Final Report, Feb. 1991.
Table 2.1 Destructive Earthquakes and their impacts in the

Philippines

DATE EPICENTER INTENSITY MAGNITUDE CASUALTY/ INJURED

01 Jul.

1954

Bacon,

Sorsogon

VII 8.3 13 101

01 Apr.

1955

Lanao,

Mindanao

VII 7.5 291 713

02 Aug.

1968

Casiguran,

Aurora

VII 7.3 • 270 600

07 Apr.

1970

Baler,

Quezon

VII 7.3 15 200

17 Aug.

1976

Moro Gulf,

Mindanao

VII 7.9 3,739 8,000

121
ig. 2. 1 Trenches and Major Active Faults in the Philippines

and its Surroundings

122
0 300km

1 I

Fig. 2. 2 Major Earthquakes along the Philippine Fault

and Seismic Gaps along the Fault (After Nakata et.al)

123
116E 117E 118E 119E 120E 121E 122E 123E 124E 125E 126E 127E 128E

Fig. 2. 3 Seismic Intensity Distribution

124
Vertical Dlsplacemehl Horizontal Displaement

Fig. 2. 4 Displacements along the Fault Caused

by the 1991 July Earthquake

125
Fig. 3. 1 Damaged Area of Dagupan City

due to Liquefaction

I

126
LEGEND

0 TOTALLY DAMAGED BUILDING

PARTIALLY DAMAGED BUILDING 1 UNDAMAGED BUILDING //

I

', ANCIENT RIVER CHANNEL

« x ^ RECLAIMED RIVERINE AREA

Source: Philippine Insritule of Volconology and Seismology

7^^ ROAD

Fig. 3. 2 Affected Buildings in Central Area of Dagupan City

127
Table 4.1 Outline of Damage (Number of Damaged Places)

1

1 Section

1 Cut slope1 Bank slope 1 Pavement

1 Others | Larg e Med

.

Small 1 Large Small i Thrust

1 «P Faults

|Kms 160 (San Jose)

jto Kms 183 (Digdlg) 36 10 10 1 3 6 1 12 9 Hud flow 1 Accumulation of | drifting wood j

|Kds 183 (Digdlg)

jto Kms 200 51 9 2 1 3 5 1 9 2 Mud flow I Drifting wood j

IKns 200

|to Kms 208 (Dalton) 30 3« 15 1 5 21 1 3 2

IKms 208 (Dalton)

jto Kms 215 (Sta. Fe) 18 7 37 1 3 6 1 2

|Kms 215 (Sta. Fe)

jto Kns 236 (Aritao) 11 11 1 1 12

1 Total 1 U6 60 75 1 15 50 1 26 13

1

1 281 places j

1 65 pi aces

1 J Note: Cut slope ; Large: more than 30 m of slope height

Medium: 15 to 30 m

Small: less than 15 m

Bank slope; Large: collapse from the boundary of the cut-bank

Small: shoulder settlement

128
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IsAii iuAaaii BIHC8 "n~ |SA1 JUAQaH BlltCI I

SA» Amelia Biitat m | SA> AIIIELIO BIIDat 0

\ fUHCAII 8IIDCE n

TVUAIIA BSIDCE ~

\ TU»Tail» n BIIDOE TUIH8III 1 BIICaE

HAIICLA Bllaci

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LABI IIIDGE

ttGA aiAKCB BiHGI

haiaciac stiaaB BOHGABOna BUDGE

SAHGAIIt BIIDat LCBinaAi BiiDai

SICSICAII BUDGE

CE». lUIA BUDGE

milATDLA BIIDGI

CAIAI Biuai

Major Damage

Bridge Fall

Fig. 5.1 Location of Damaged Bridges

134
254 559

13 165 165

5£s- ' --ess

89

i

-I •

I

241 241

6 SO

13

I--

89

(a) F i xd Bear i ng

11.

1

229 229

4E>7

-i

25 64 241

1—

r

241 64

660

25

(b) Movab 1 e Bearing

Fig. 5. 2 Bearing Details

135
136


Mm

(tf-m)A 20-

10-

<D Mru=18. Otf-m

D Mru=15. 8tf-m

<D Mru=13. 9tf-in

5 T (cm) n-20

- 10

Vo =SWd+ (Buoyancy)

=273-30=243tf

Ho =kh-SWd

400

0.10 kh

(27.3) (Ho(tf))

- Mm

-5t

kh ;Seisniic Intensity Equivalent

to Horizontal Force Due to Earthquake

Mm ;Calculated Maximum Bending

Moment of a Pile

Mru;Ultimate Bending Moment Capacity

of a Pile Determined by Pile Section

(5 TiCalculated Displacement at the Pier Top

KH :Assumed Lateral Ground Constant

Assumed Pile Section

Calculation Cases of Reinforcing Steels

® 8*D19 =22.9cm2

(E) 8+D22 =31.0cm2

d) 8*D25 =40. 5cm2

Fig. 5.6 Relationship between the Lateral Resistance of a Pile

and Effect of Seismic Force in Case of Pier 5

139
Mm

(tf-m) P i e r 4

Mru=12. 8tf -m

Mru=10. 5tf -m

10-

Mru=8. 5tf -m -CD

0.3 kh

(100.5) [Ho(tf)3

Mm

5t

kh ;Seismic Intensity Equivalent

to Horizontal Force Due to Earthquake

Mm ;Calculated Maximum Bending

Moment of a Pile

Mru;Ultimate Bending Moment Capacity

of a Pile Determined by Pile Section

(5 T;Calculated Displacement at the Pier Top

KH ;Assumed Lateral Ground Constant

Vo =Sffd+ (Buoyancy)

=335-55=280tf

Ho =kh-5:Wd

4QQ_

Assumed Pile Section

Calculation Cases of Reinforcing Steels

. ® 8*D19 =22. 9cm2

(2) 8*D22 =31.0cm2

(D 8*D25 =40.5cm2

Fig. 5.7 Relationship between the Lateral Resistance of a Pile

and Effect of Seismic Force in Case of Pier 4

140
One span of the

bridge fell

fcalvo Cal \

I Bridge Abutment wa cracked

Embankment protected

by parapet was cracked

and had settled at this location.

1.

Agno River

( a )

1 J

Agno River (low flow channel located far away)

Marshy area

Marshy area

( b )

Fig. 6.3 Damaged Sites of Agno River

141

Recorded Structural Response During the M7.1

Loma Prieta Earthquake

A. Gerald

ABSTRACT

The recording of structural response

during strong-motion earthquakes has

been a goal of the structural engineering profession since the 30' s, in

the profession's quest to understand

how buildings behave and to design

against collapse. The Loma Prieta

earthquake provided an unmatched collection of records, although few instrumented structures were even close

to collapse. Near-collapses or structurally damaging motion will be

recorded at a much slower rate than

the engineering profession desires

and funding programs will need to address this question.

KEYWORDS: collapse, instrumentation,

response, strong-motion, structures.

1 . INTRODUCTION

The Loma Prieta earthquake provided

strong-motion accelerograms from 137

ground and structural stations

throughout the epicentral area, the

San Francisco Bay Area and surrounding areas. The largest networks are those of the California Strong-Motion

Instrumentation Program, CSMIP, (of

the Division of Mines and Geology,

CDMG, of the Department of Conservation) , and the US Geological Survey,

USGS . Several structures instrumented by private organizations add to

the total. The data from CDMG and

USGS are made readily available as soon as processing is completed

(CSMIP staff, 1989; Brady and Mork,

1990) , and much research on structural behavior has already been

initiated. A seminar was held by

CSMIP on 30 May, 1991, where short

presentations were made by researchers having already spent a year on funded research on Loma Prieta data

(CSMIP staff, 1991) . A further year

of funding, with grants of up to

$25,000, was launched in May, 1991.

Brady*

As of this writing, May-June, 1991,

data are available from 47 structural

stations: 37 multi-story buildings,

four dams, three bridges, a gymnasium, an elevated rail and a wharf.

Some instrumentation dates from times

when only a limited amount of data

was satisfactory (three of the multistory buildings and one dam) , but 94%

of the structural stations recording

Loma Prieta have extensive instrumentation, with between 10 and 30 channels designed and placed to help

solve specific problems related to

the structure, in all or part.

Structural stations cover a wide

range of epicentral distance, from a four-story Watsonville building at 18

km and Anderson Dam at 27 km, to a 13-story Berkeley building at 99 km

and a 3-story Richmond building at

112 km, although beyond an epicentral

distance of 100 km, ground level ac- celerations at structural sites do

not exceed O.lOg. Azimuthal coverage

is restricted to approximately 180

degrees, since the San Andreas fault

lies close to the Pacific Ocean. The

prominent centers for these structures are the urban areas of San

Francisco, Oakland and San Jose, and

nearby locations. However, the re- mainder of the northeastern side of

the fault is uniformly covered.

2. INSTRUMENTATION GOALS.

Structural instrumentation reflects a longheld desire to understand how

structures behave during earthquakes,

particularly during damaging motion.

Recorded weak motion, defined for the

purposes of this paper as a level of

motion for which the structure re- mains primarily elastic and undamaged, is useful for the interpretation of a specific structure's

design and the relationship between

* U.S. Geological Survey

Menlo Park, CA94025

143
design concepts and actual elastic

dynamic behavior. Recorded strong

motion, on the other hand, concentrates attention on damage and poorlyunderstood dynamics and can, if called for, provide evidence that building codes need improving. It is the

intelligent analysis of recordings

from strong damaging motion that will

decrease the seismic hazard in the

years ahead, as well as the much more

frequently occurring visual inspection and subsequent interpretation of

damaged structures.

As structures have become more complex, architecturally and geometrically, so has our interest heightened

in their behavior during strong motion. More sensors have clearly

become necessary as structure size

increases. However, until every

joint is monitored for displacement

and rotation, a clearly impossible

task, missing structural dynamic

properties will be needed from estimates, calculations and assumptions, in order to provide the motion

at non-instrumented points. Graphical representations of structural dynamic response use joint motions that are recorded, together with estimated

values for all other points

.

The more obvious structural parameters that affect dynamic behavior

during strong damaging motion, and

would have to be incorporated in the

analysis of strong-motion recordings,

include

:

a. The force-displacement relationship, perhaps as simple as elastoplastic

.

b. Joint detailing, the reinforcing

steel's containment of concrete,

and the ability to maintain integrity.

c. The ductility ratio able to be

achieved

.

d. Behavior under repeated excursions

into the nonlinear regime (see, for

example, Perez and Brady, 1984) . These considerations complicate

enormously the analysis of structural

strong-motion recordings, although

not from the Loma Prieta earthquake

because none of the instrumented

buildings were damaged.

3. SYSTEM IDENTIFICATION.

The goals of system identification

routines include the identification

of basic vibrational patterns in the

response of a structure to an earthquake input, from which the total

vibrational behavior can be built up.

The extent of the participation of

the basic patterns, or modes, can change with time as the earthquake

input progresses (see, for example,

Safak, 1988) . Taking into account

the incomplete instrumentation that

is lamented on earlier, any mode

shapes associated with system identification might well be reliable at

instrumented points, but unreliable

at all others. In spite of this,

much useful information is obtained

on frequencies, damping, participation factors, mode shapes of basic

vibrational patterns, and the variations of each of these with time,

from system identification investigations. It is clear that the use- fulness of the information is limited

by the same concerns expounded above,

namely, incomplete instrumentation,

unknown nonlinearities, and assumptions, however subtle, on the way

structures behave during strong

damaging earthquake motion. A few

records from earthquakes other than

Loma Prieta have been recovered from

structures suffering structural damage from earthquake shaking, and fewer still from structures with extensive instrumentation. These records

offer the real opportunity of corre- lating earthquake input, nonlinear

structural motion, and documented

damage. The structural engineering

community and the building code writers need more of these research

results

.

4. VERTICAL MOTION

A specific area where an understanding of earthquakes, structures, and

their relationship appears to be less

than adequate is in the analysis of

vertical structural motion. Although

designing for vertical loads is basic, the dynamics associated with

measured vertical motion adds a dimension of difficulty. Differing

vertical motion at two different locations at the same level (for ex144
ample, basement level) , implies either the rocking of the structure on

its foundation as in the rolling of a ship at sea, or the vertical motion

of columns that remain vertical as in

the stepping of stilts. Without the

measurement of joint rotations, we cannot distinguish between these two

possibilities, because there can be an arbitrary amount of the second,

the stilt-stepping, without any

addition to the horizontal structural

motion. How much, then, of the re- corded vertical motion is rocking?

5. PERIOD LENGTHENING DURING STRONG

MOTION

Records from an extensively instrumented structure generally permit the

calculation of natural periods and

lowest modes of vibration even from

original analog records, should this

be the recording medium. As the

excitation continues, or as it builds

in intensity, the stiffness of individual components decreases, due to

infinitesimal cracks, and, the mass remaining constant, the modal periods

consequently lengthen. Under stronger

excitation still, structural members

weaken from cracks, initially infinitesimal, but becoming damaging given

sufficiently high excitation levels.

This is accompanied by further period

lengthening. This behavior is not

yet sufficiently documented nor, consequently, able to be estimated. Its

presence in the strong-motion records

of a structure provides a quantitative measure of the decrease in

stiffness, but no details of its

occurrence

.

6. DESIGN CONSIDERATIONS

The successful design of structures

for seismic effects, apart from following the relevant building codes,

is dependent on the ability to understand how a completed structure will

respond to a future earthquake, how

this response builds, and how the behavior changes from the expected norm

with the lowering strength of critical sections. This task will slowly

become easier as the response of specific structures to weak motion is

measured. All instrumented structures in the Loma Prieta area reached

at least this stage. Further success

will follow the recording of response

of structures to strong motion at

non-linear levels (a small fraction

of Loma Prieta structures) . Less common still, we await correlations

made between damage to a structure

and its recorded motion during the

damaging experience. Post-earthquake

reconnaissance of damaged structures,

an endeavor that precedes recorded

motion by many hundreds of years, re- tains its position as a fundamental

provider of experience, namely, the

partially documented detailed dynamic

response of structures. Documentation can only be considered partial

because there need not be a one-toone correspondence between the timevarying dynamics of a structure and

the observable damage after an earthquake . The goal of reducing the earthquake

hazard in structures yet to be designed and built relies in part on understanding the dynamics of existing structures. Even though a structure's behavior during a specific

strong-motion earthquake might be

well documented, there is very little

data to suggest that the same dynamic

behavior will be evident during the

next . To the author' s knowledge

there is in fact no such example. Insofar as the next earthquake to cause structural damage to a specific

structure is not going to be a duplicate of one that already has done so, it is likely that specific instances

of dynamics-damage pairs will remain

part of a very slowly growing and

consequently little-used database.

This statement might well be used as an argument for increasing the number

of thoroughly instrumented structures in high seismicity areas, but

the best that can be hoped is that

the time scale for useful data be

reduced from hundreds of years to

perhaps scores. In any case, the

time is much further into the future

than funding programs can usually

envisage

.

The structural engineering community

and particularly that segment of it

that is responsible for changes to

building codes cannot wait for a database of pairings of measured dynamics and inspected damage. There exists a risk in attempting to consider

changes to a code on the basis of a specific dynamics-damage trend in

145
that no such trend can yet be con- sidered reliable. Changes to codes

are slow to appear for this reason.

It is clear that the tightening or loosening of some seismic code re- quirements, at the expense of those

remaining unchanged, might possibly

be associated with not fully understanding the complete structural

dynamics phenomenon. The detailed

faulting mechanism and wave propagation characteristics of all future

damaging earthquakes are surely not

yet sufficiently understood to be of

specific use to the structural engineering profession in its efforts to

reduce seismic hazard.

7. BAD GROUND, AMPLIFICATION, AND

RESONANT FREQUENCIES

One of the interesting research efforts in the San Francisco Bay Area

after the M7 . 1 1989 Loma Prieta

earthquake has been the identification of areas of soft soil able to

amplify vertically-travelling shear

waves on their passage up from underlying bedrock. Horizontal ground os- cillations from these waves are generally considered the principal cause

of damage to residential areas in San

Francisco, to the Bay Bridge, and to

the Cypress Street Viaduct (Hanks and

Brady, 1991) . In addition to identifying the existence of soft soil of

sufficient depth, questions are raised as to why some of these areas did result in significant amplification, with resulting damage to structures thereon, while others, even much closer to the epicenter, did

not . One of the obvious contributing

factors is that each soft soil area

has its own set of resonant frequencies appropriate to large, nonlinear,

damaging motion, at which amplification is certain. Outside these frequencies, and how far outside is dependent on the damping in the soil at

large, nonlinear, damaging motions,

there is little amplification. At

the rock/soil interface, or more ac- curately, at all interfaces between

high shear wave velocity media, and

low, an amplification must occur,

independent of frequency, because of

energy considerations. If the resonant amplifications are dominant at a specific site, then this site will

suffer large, damaging motions only

if the resonant frequencies and the

input frequencies of the earthquake

coincide, and if the input amplitudes

are sufficiently high.

Although surface waves will have

played a part in the ground motion

throughout the Bay Area, they did not

apparently contribute to any of the

damaging motion in the San Francisco

area

.

Shear waves emanate from a rupturing

fault as slip proceeds. One prominent propagation path is in the direction transverse to the rupture,

with particle motion parallel to the

rupture, driven by the slip on the

fault. Along the direction of rupture, both ahead and behind, trans- verse shear waves travel, with trans- verse particle motion, in sympathy

with the relaxation of the transverse

restoring forces. The frequencies of

these waves, and all other associated

body waves, are entirely dependent on the detailed travel history of the

rupture. This detail is not known,

and cannot be measured, but efforts

at estimating it are made from

strong-motion records written at the

surface in the epicentral and adjacent areas. There exists no way of

estimating the predominant frequencies or their possible range for

large damaging earthquakes. Representations of estimated rupture

detail indicate, in general, non- uniform rupturing and variable rupture speeds. For any large earthquake, even a repeat earthquake on the same section of fault, generated

frequencies are clearly arbitrary.

The highest and lowest frequencies

measured on strong-motion instruments

on rock in the epicentral area of

large earthquakes are each affected

by signal-to-noise limitations, but

are of the order of 50 and 0.05Hz,

respectively. Fourier spectra or response spectra of such records

confirm the apparent arbitrariness of

prominent frequencies between these

bounds

.

8. CONCLUSIONS

The structural engineering profession

is wise to limit its treatment of the

seismic component of codes to a general one. For example, in light of

the above paragraphs, there is no justification at present, nor in the

146
forseeable future, for treating any- soft soil in high seismicity regions

differently from other soft soil.

Although the understanding of the

seismological and geophysical aspects

of the most recent multiple-recorded

earthquake (in the present case, Loma

Prieta) has grown considerably in

recent years, the ability to predict

the effects of the next remain in

question. There are not enough data,

and there have not yet been enough

earthquakes, for the situation to be

any different.

9 . REFERENCES

1. CSMIP staff, 1989, "Plots of the

processed data for the interim set of

14 records from the Santa Cruz Mountains (Loma Prieta) earthquake of October 17, 1989." Office of StrongMotion Studies Report, CDMG, Sacramento, Calif., and subsequent reports

and tapes.

2. Brady, A.G., and P.N. Mork, 1990,

"Loma Prieta, California, earthquake,

October 18 (GMT), 1989; processed

strong-motion records, vol 1," U.S.

Geological Survey Open File Report

90-247, and subsequent tapes.

3. CSMIP staff, 1991, "SMIP91 semi- nar on seismological and engineering

applications of recent strong-motion

data; preprints," Calif. Dept. of

Conservation, CDMG; Sacramento,

Calif.

4. Perez, V. and A. G. Brady, 1984,

"Reversing cyclic elastic demands on structures during earthquakes and

application to ductility requirements," EERI Earthquake Spectra, 1,

1, p. 7-32.

5. Safak, E., 1988, "Analysis of

recordings in structural engineering:

adaptive filtering, prediction and

control, " US Geological Survey Open

File Report 88-647.

6. Hanks, T. C. and A. G. Brady,

1991, "The Loma Prieta earthquake,

ground motion, and damage in Oakland,

Treasure Island, and San Francisco,

Bull. Seis . Soc. Am., in press.

Monotonic and Cyclic Loading Tests of Rockfill Materials

Norihisa MATSUMOTO*'- Nario YASUDA*^. Masahiko OHKUBO*^, Ryo-ichi YOSH IOKA=*''*

ABSTRACT

In this paper the static and dynamic

strengths of rock materials of rockfill

dam body are investigated through

monotonic and cyclic loading tests.

Keyword Rockfill materials,

large-scale triaxial test. cyclic

loading, monotonic loading

1 . Introduction

In an undrained cyclic loading test

on saturated loose sands under

laboratory conditions, the sand tend

to flow losing its original structure,

resulting in liquefaction. In contrast,

dense sands don't tend to behave this

way and shows the condition of cyclic

mobility. Although this happens, due to

the cyclic loading the pore water

pressure increases causing cumulative

strain magnification. Therefore, the

dynamic strength can be established

from the number of cycles and

resulting strain. The static strength

can be simply obtained from a monotonic loading test.

From this, it seems that monotonic

and cyclic loading tests are not

mutually exclusive .but actually

rel ate to each other.

This report is concerned with the

monotonic and cyclic test result for

well compacted rockfill materials. The

paper aims at illustrating the

definition of shear strength by the

laboratory testing, and presents the

effect of cyclic numbers, prescribed

strain, relative density, initial

confining stress, initial shear stress

and the testing method for dynamic

strength, and the overall relationship

between static and dynamic strength.

2. Specimen Preparation ajid lasi

Procedure

The materials used in this study were

transported from a quarry site for the

Sagurigawa Dam. The rock material

consists of very hard and high quality

gabbro.

A rigid specimen of 30-cm in diameter

and 60-cm in height was utilized for

the cyclic triaxial test. A hollow

cylindrical specimen was an 80-cm in

outer diameter. a 40- cm in inner

diameter and a height of 80-cm was used

for the cyclic torsional shear test.

These specimens are shown in Fig.l. The

physical properties of the test

materials are given in Table 1. The grain size distribution curve for

the test materials is illustrated in

Fig. 2. The maximum particle size of

specimen is 63.5 mm. which is less than

1/5 of the diameter of the specimen

used for the monotonic loading tests.

The uniformity coefficient of the rock

material was set 30 in order to minimize

the effect of the system compliance. In this figure, the gradation curve for

the rockfill materials of the Oroville

Dam (with which Banerjee et al.'^

conducted cyclic loading tests with

undrained condition) is also plotted.

The gradation curve for the Oroville

Dam materials had the maximum particle

size of 50.8 mm (which is slightly small

when compared with 63.5 mm for the

maximum particle size of our specimens)

is almost the same as that for this

study Banerjee etal. used a 10%

reduction in the cyclic strength ratio

as correction factor for the membrane

compliance of the undrained cyclic

triaxial test, for materials with a diameter of 9.5 mm at Deo (50% diameter

of soil particle). Therefore, for the

purpose of this study, the 10%

reduction in the cyclic strength ratio

is to be repeated.

The materials were placed in six

layers, with each layer compacted by

33kgf weight electric vibratory

rammer. Finally, the specimens with 85%

and 95% of relative density were

prepared.

The procedures for cyclic testing is given as foil ows.

1) A confining pressure of 0.2 kgf/cm^

was applied to the specimen (30-cm in

diameter and 60-cm high) and

subsequently the vacuum pressure

inside the specimens was released.

2) Carbon dioxide (CO2) was supplied

under pressure of 0.05 kgf/cm^ to the

t] Dr. Eng., Chief. Filldam Division.

P.W.R.I., Ministry of Construction,

Tsukuba City. Japan (Presently Japan

Dam engineering Center)

t2 Senior Research Eng.. ditto

t3 Eng.. ditto

tA Visiting Research Eng.. ditto

(Presently Shimz Corp.)

149
specimen for a 24 hours period, in

order to fill any air void completely

with CO2. CO2 was used because it is much easier to dissolve in water than

air and a higher degree of saturation

for the specimens can be accomplished.

3) After replacing the air with CO2.

70 liters of de-aired water was supplied to the sample for a 24 hours

period.

4) The degree of saturation of the

specimen was then investigated by

Skempton's B-value method and testing was carried out if the specimen had a B-value of more than 0.96.

5) A 1.000 cc burette was used for

measuring the volume change in the

sample during an average 45 minute

consolidation. Cyclic loading tests

were started once the rate of volume

change in specimen became less than

0.05% per mi nute.

A cyclic undrained test was carried

out under load-controlled condition of

0.1 Hz. The specimen was subjected to a back pressure of 2.0 kgf/cm^.

The digitized test results were

recorded on a personal computer using

a low-pass filter of 1.0 Hz and were

subsequently plotted.

3. Dynamic strength

Tdf is defined as the cyclic strength

achieved during cyclic loading test , where Td is the cyclic stress when the

shear strain reaches £pf at failure

after N^- cycles. Generally. T df is

effected by the vertical effective

stress CT'vc . horizontal effective

stress cr'hc of the field, cyclic number

N and the initial shear stress Ts. The

shear strength is expressed as merely

Tdf or (Ts+Tdf) which is the sum of

Ts and Td-p. and the shear strength

ratio is obtained by dividing both

shear strength values by cr'v. These

points are used in the following

d i scussi 0 n

.

In the case of the triaxial tests,

shear stress ratio is expressed by the

following equations;

T d -f T d -f

crr

cr •(1)

( cr 3 m a X -crr)/2

cr cr where

cr'no - initial normal stress on 45°

plane after consolidation,

cramax = maxImum axial stress.

cr r - confining stress.

The dynamic strength ratio gained

from the triaxial test has to be

corrected since tests in the

laboratory are not representative of

the in-situ considering the the

rotation of the axes of principal

stresses. horizontal stress and

plane-strain state etc.

=C., -(2)

cr'^ IN-SITU or'v TRX

Several values of Cri have been

reported by previous researchers

according to Ko(= cr 30/ cr 1 ol the

principal stress ratio during a consolidation) as shown in Table 2. On the other hand, the cyclic shear

stress Td is added with the initial

shear stress Ts for torsional simple

shear tests. T dr is defined as the

shear strength achieved during cyclic

torsional shear test for a Td of a cyclic stress when the shear strain

reaches a certain value, which is established as the axial strain t pt or shear strain T p-p at failure, after U-p loading cycles. Hereinafter. the

strength is expressed by a

d i mensi 0 n 1 ess stress ratio Td-f/CT'v or

( Tdi-+ Ts)/ cr'v as the case of the

triaxial test. A correction factor Crs

shown in the following equation is

required to compare the in-situ stress

ratio with the laboratory stress ratio.

This is because the plane-strain

condition is not completely satisfied

unless the radial confining stress is

controlled so as to maintain the areal

strai n at zero. T df T df

= Cr2 -'(3) a\ IN-SlTU cr'v TS

It can be said that Crs is approximately 1.0.

4. Test results

In this report, the strain regulating

dynamic strength in the cyclic loading

test is defined as single amplitude,

which is half of the double amplitude

for isotropic consolidation, and the

peak strain which remains in

compressed direction for anisotropic

consol idation.

The dynamic Strength ratio is expressed by the following equation.

T d+ T s

...(4)

where. Td = cyclic shear stress.

Ts = initial shear stress.

Pc=initial effective mean

principal st r ess. Fig. 3 illustrates the relationship

150
between the dynamic strength ratio and

the number of cycles. Each curve shows

the dynamic strength ratio when the

axial strain reaches 0.5. 1.0. 2.5 and

5.0% respectively. The dynamic

strength ratio for the anisotropical 1 y consolidated condition is 2 to 3 times

as large as that of the isotropical 1 y consolidated one. The dynamic strength

decreases according to the increase in

the number of cycles or the decrease

of prescribed strain. The curve of the

dynamic strength ratio for the

anisotropical 1 y consolidated condition

has an inclination to close at the

boundary line of stress reversal

(( 'rd+Ts)/Po=0.75) . Fig. 4 shows the

relationship between Po and the

dynamic strength ratio. which is obtained from the curves for 20 cycles

with 1% of axial strain for the

specimens at 85% and 95% of relative

density. The dynamic strength ratio for

the specimen at 95% of relative density

under both isotropical I y and

anisotropical 1 y consolidated condition

is 1.5~2.0 times larger than that at

85% of relative density. The dynamic

strength ratio depends upon the

confining stress and increases

according to the decrease in Pc.

Yunoki, et al®^ reported the same tendency as in the results for dense

sands.

Fig. 5 indicates the relationship

between the dynamic strength ratio and

the relative density. The increase in

the dynamic strength ratio at each

level of shear strain is equal for the

isotropical 1 y consolidated specimens.

In the case of unisotropical I y consolidated samples, the increase in

dynamic strength ratio is remarkably

reaching 2.5% of the axial strain.

The relationship between the dynamic

strength ratio and principal stress

ratio during consolidation is indicated

in Fig. 6. The dynamic strength ratio

has a tendency to increase

accompanying the increase in the

principal stress ratio as described by

Yasuda. et al ."'^ Fig. 7 shows curves of the dynamic

strength ratio for the torsional shear

test under isotropic consolidation,

which coincides with that of the

triaxial test. This corresponds to

about a value of 1.0 for Cri for a value of 1.0 for Ko in Table 2. The

shear strain y in the cyclic torsional

test is compared with axial strain £a

in the cyclic triaxial test and

Poisson's ratio P which is 0.5 for

undrainedtests.

Yamashita. etal.®' reported that the

result of the torsional simple shear

test on dense sands is greater than

that for the triaxial test on the

sample prepared using the

ai r-pl uviation method. On the other

hand. Ochi. et al.^' claimed that the

relationship between the result of the

triaxial test to that of the torsional

shear test. is dependant upon

differences in the sample preparation

method and relative density.

Fig. 8 shows the comparison between

the undrained dynamic strength, and

the drained and undrained static

strengths for rockfill materials. The

dynamic strength is the value

recruited at 20 cycles and 2.5% of the

axial strain. Incidentally, the static

strength is defined as the peak stress

within 15% of the axial strain. It is recognized that the undrained dynamic

strength with an anisotropic

consolidation is larger than the

drained static strength at a normal

stress of less than 5.0 kgf/cm^.

However. 2.5% of the peak strain and 20

cycles have been adopted, and these

values are not appropriate for

application as the dynamic failure for

filldams. These have been adopted in

this report because of the limitation in

the loading capacity of the triaxial

testing equipment used here.

Therefore, the dynamic strength at

larger axial strain is expected to

i ncrease.

5. Conclusion

The following points were concluded

after carrying out a series of

monotonic and cyclic loading tests,

using well compacted rockfill

materials.

® In the undrained cyclic loading

test, the pore water pressure rises and

the strain increases according to the

increase in cyclic frequency. The

dynamic strength obtained depends on the cyclic frequency, reference strain,

relative density, initial confining

stress, initial shear stress and testing

method. The overall tendency in this

study coincides with that expected for

a dense sand.

(D The dynamic strength in cyclic

torsional shear is almost the same as that for the cyclic triaxial test with

the isotropic consolidation.

(3) The dynamic strength for the

anisotropic consolidation was greater

than the drained static strength at a

151
normal stress of less than 5.0 kgf/cm^.

References

1) N.G.Baner Jee. H.B.Seed, C.K.Chan.

"Cyclic Behavior of Dense CoarseGrained Materials in Relation to the

Seismic Stability of Damsj. UCB/EERC79/13. 1979.

2) W.D.L. Finn. J.J.Emery. Y.P.Gupta.

•"A Shaking Table Study of the

Liquefaction of Saturated Sands during

Ear thquakes J . Proc. Third European

Symposium on Earthquake Engineering.

1970.

3) H.B.Seed and W.H. Peacock, rjest

Procedures for Measuring Soil

Liquefaction C h a r acte r i st i cs J . Proc. of

ASCE, SMS. Vol.97. 1971.

4) G.Castro, f" L i q u ef act i o n and Cyclic

Mobility of saturated Sandsj. Proc. of

ASCE. Vol.101. N0.GT6. 1975.

5) Ishihara.K. and Li, S-|..

^Liquefaction of Saturated Sand in

Triaxial Torsion Shear TestJ, Soils and

Foundations, Vol.12. No. 3, pp.l9~39.

1972,

6) Yunoki, Y. et al.. rEffect of

Confining Pressure on Cyclic Triaxial

Test Results of Dense Sandj, The 17th

Japan National Conference on Soil

Mechanics and Foundation Engineering,

pp. 1649-1652. 1982. (in Japanese)

7) Yasuda.M. et al.. ''Cyclic Undrained

Deformat ion-Strength Character istics

of Akashi Undisturbed Gravity Samples

in Large-Scale Triaxial Testsj, The

24th Japan National Conference on Soil

Mechanics and Foundation Engineering,

pp. 817-820. 1989. (in Japanese)

8) Yamashita. S.. et al.. ""Cyclic

Undrained Triaxial and Torsional

Shear Strength of Sand with

Consideration of Membrane

Penetration J. The 22th Japan National

Conference on Soil Mechanics and

Foundation Engineering. pp. 777-780,

1 987, ( i n Japanese)

9) Ochi. K. et al., rComparison of Shear

Strength Between Triaxial and

Torsional Shear Testsj. The 22th Japan

National Conference on Soil Mechanics

and F 0 u ndat i on E n g i neer i ng. pp. 583-586.

1 984, ( i n Japanese)

152
Table-1 Physical Properties of Materials

Spefific gravity Gs 2. 933

Water content W 0. 54(%) A i r -dry

Maximum void rat i o e m a X 0. 569

Minimum void ratio e m I n 0. 142

Maximum grain size d m a X 63. 5 mm

Coefficient of uniformity Uc 30

Dry density of specimen r d 2.46. 2 . 53 g f /c m

^

Table-2 Cn of Previous Researchers

E quations Cr 1 for Ko = 0.Cr 1 for Ko=l.

Finn, et al. 2> Crl=(l+Ko)/2 0. 70 1. 0

Seed and Peacock^' Varies 0. 55~ 0. 72 1. 0

Castro^' Cri=2(l + 2Ko)/3>A 3 0. 69 1. 15

Ishiiiara and Li^' Cri=(l+2ko)/3 0. 60 1. 0

153
® GAP SENSER

(D AXIAL LOAD AND TORGUE PICKUP

(D DISPLACEMENT TRANSDUCER

0 ROTARY DEGREE PICKUP

© CAP

(D MEASUREMENT FRAME

@ HALLOW CYLINDRICAI SPECIMEN (08O-4Ocm x H80cm)

(D PEDESTAL

(9) CELL BOTTOM

@ OUTER VOLUME CHENGE PICKUP

(D) CELL PRESSURE EOUIMENT

@ TORQUE LOADING EQUIPMENT

O TORQUE ARM

(Q) TORQUE REACTION

® LOADING AND REACTION ROD

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Fig. 1 Large-Scale Cyclic Torsional

Shear Apparatus

154
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155
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Initial mean plicipal stress pc (I(gf/cn2)

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ratio and initial mean principal stress

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Q.

3.0

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156
1.

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Cycl ic loading test for rock

(Dr=85X, pc=4. 0kjf/ctii2, n=20)

1.0 1.5 2.0

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torsional shear tests

157
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158
Full-Scale Structural Applications of Active Control Systems

by

T.T. Soong*, A.M. Reinhom* and M. Shinozuka*

ABSTRACT

An active bracing system has been designed, fabricated, and installed in a full-scale dedicated

test structure for structural response control under seismic loads. Presented in this paper are the design of the active system, control algorithm

development, and simulated as well as preliminary test results. Detailed design and analysis of

the active system are carried out with respect to hardware development, control force constraints,

and power and energy requirements. It is shown

that a full-scale efficient active structural control system can be developed within limits of

current technology. Simulation and test results provide information on performance bounds that can be expected of active systems in structural control under seismic loads and under constraints

imposed by practical considerations.

KEYWORDS: structural safety; active control; earthquake engineering.

1. INTRODUCTION

The possible use of active control systems as a means of structural protection against environmental loads, such as wind and earthquakes, has

received considerable attention in recent years.

It has now reached the stage where active sys- tems have been installed in fuU-scale structures (Kobori, 1990; Soong, 1990). As an example, an active mass driver system was recently in- stalled atop of the eleven-story Kyobashi Seiwa

Building in Tokyo. This system, designed primarily for comfort control of building occupants,

has performed well under several hurricanes and

moderate earthquakes encountered since installation.

Another active mass damper system was recentiy

installed in a dedicated full-scale 600-ton test structure on the grounds ofthe Takenaka Research

Laboratory. It has also performed well during a recent moderate earthquake. In addition, a full- scale active bracing system has been installed

in the same structure for performance evaluation

purposes. The advantage of having the perfor- mance of two active systems evaluated using the same structure is obvious. In addition to pro- viding the same base parameters for performance

comparisons, this arrangement aUows the calibration of both systems by using one ofthe systems as motion inducer and the other as motion controller. Even without actual environmental disturbances,

much of the performance characteristics of these

systems can be assessed.

To provide a focus, this paper addresses design,

fabrication, and performance issues related to the

development of the active bracing system. These

discussions provide a realistic assessment of the

potential benefit which can be derived from an

active system on the one hand, and system design

and capability requirements on the other.

2. TEST STRUCTURE AND

ACTIVE BRACING SYSTEM

2.1 Test Structure

As mentioned above, a dedicated full-scale struc- ture was erected for performance verification of

two active structural control systems, i.e., an active mass damper and an active bracing system.

Located in Tokyo, Japan, the structure is a symmetric two-bay six-story building as shown in Fig.

1. It was constructed of rigidly connected steel frames of box columns and W-shape beams with

reinforced concrete slabs at each of the floors. Weighing 600 metric tons, the structure was designed as a relatively flexible structure with a fundamental period of 1.0 sec in the strong di- rection and 1.5 sec in the weak direction in order

to simulate a typical high-rise building. It is as- sumed that the floor slabs are rigid in their planes

and the structure does not twist; accordingly, the

* National Center for Earthquake Engineering

Research and Department of Civil Engineering, State University of New York at Buffalo,

Buffalo, NY 14260

159
space structure was modeled as a fixed-base plane

frame with three degrees of freedom (horizontal,

vertical and rotation) at each joint in the eigenvalue analysis. The stifftiess of the concrete slabs was taken into account by considering rigid zones

at ends of the members in analytical modeling.

The spectral properties of the structure obtained

analytically as well as experimentally are compared in Table 1 , showing that their differences

are within 3% for all modes.

2.2 Active Bracing System (ABS)

As shown in Fig. 1, the biaxial active bracing

system consists of a pair of hydraulic actuators

in each of the principal directions with the di- agonal braces connected to the first floor. The

actuators, monitored by two hydraulic servovalve

controllers with auxiliary hydraulic accumulators,

are capable of generating 685 kN of control force

in each direction.

Velocity sensors are attached to the base, the first, the third and the sixth floors and an accelerometer

is placed at the base. Measurements derived from

these sensing devices serve as the feedback state variables in either of the two proposed control

algorithms. The PC-Limited 286 microcomputer

facilitated with A/D and D/A boards is used for

on-line computation. The control logic is implemented on an 80286-24 MHz processor and an 80287-10 MHz co-processor.

2.3 Design Earthquakes

For design purposes, the peak velocity of the

design earthqu^es was taken to be 10 cm/sec

based on local seismic records of the past seven

years (maximum = 9.5 cm/sec). Accordingly,

the scaled (32%) El Centro earthquake with 98

cm/sec^ (0.1 g) peak acceleration was determined

as the design earthquake which corresponds to the criterion of 10 cm/sec maximum velocity.

Response analyses were also carried out using

a series of recorded earthquake time histories to verify the adequacy of the design specifications.

2.4 Control Algorithms

The real-time computations in the microcomputer

were based on a reduced order model (ROM)

active control algorithm. Two basic algorithms

based on the classical closed-loop linear optimal

control law (Reinhom, Soong et al, 1989) were

designed for this investigation: (a) three-velocity

feedback which uses only velocities measured

at three critical floors in the structure (Soong,

Reinhom, et al, 1991) and (b) velocity feedback

with observer which accounts for full dimensional

state feedback with aid ofa state estimator (Soong,

Reinhom et al, 1991).

Both algorithms were modified to include the

compensation of delays in the force application

with respect to the measured signal. The time

delays are a result of phase shifts in instruments,

conditioners, time required for on-line computation and time required for the actuator to respond.

A phase type compensation was employed in connection with compensation of a parallel time delay

in computations and actuators. AU compensations were based on field measured performances

of the control components (Reinhom, Soong, et

al, 1989).

2.5 Simulated Results

Simulations using several design earthquakes produced satisfactory results as shown in Table 2. The maximum power required during an earth- quake reaches 20 kw for most earthquakes except

for Hachinohe earthquake where 50 kw are re- quired. The total energy consumption during an

earthquake of one minute will be ofthe order of0.3

kwh to 0.9 kwh which can be delivered from passive electrical power sources. Additional power

should be considered for the continuous operation

of changing electrical and hydraulic equipment.

A typical structural response of the full-scale sys- tem is shown in Figures 2. The performance

is shown for the top floor of the structure where

both displacements and accelerations are reduced.

As a result, the base shear is also substantially

reduced. The required resources to produce such

performance are shown in Figure 3 (for the strong

direction of the structure for both actuators operating in this direction). The large forces are required for short period of times associated with

several power peaks and a rapid initial energy

demand.

3. DESIGN OF ACTIVE BRACING

SYSTEM (ABS)

The primary parameters upon which the detail

system design is based are the control force, the

actuator displacement and the actuator velocity,

160
which are related to the determination of the ac- tuator capacity, the cylinder stroke, and the flow

rate requirement of the hydraulic servovalve. In

addition, the total flow of the hydraulic fluid re- quired during a seismic event is the basis of sizing

the hydraulic supply system. During the operating period of die active system, the hydraulic

servo-controller supplies a constant flow of oil regardless of the actual requirement. When the

demand of oil is more than what is supplied, the

accumulator will pressure a subsidiary flow to the

actuator cylinder; inversely, when supply is more

than demand, excess oil will be discharged into

the accimiulator, thus no additional pumping is necessary during the short period of earthquake

excitation. For economic reasons, the pumping

rate of the controller is determined to be the av- erage flow rate estimated over the time history

of the cumulative flow such that the accumulator

volume is minimized. These design parameters were established in accordance with the simulated results of resource demands required for a

desired structural performance under the design

earthquake as discussed below.

3.1 Preliminaries

The maximum control force and the specification of the hydraulic actuators were determined

based upon the experimental study of a 1 :4 scaled

model. The fundamental frequency of the 1:4 scaled model was about 2.5 Hz and the time scale

of the ground motions used in the tests was a re- duction by a factor of 1/2 of the prototype, which

made the model dynamically similar to the full- scale structure having a fundamental frequency of

approximately 1 Hz and subjected to the design

earthquake. Therefore, the expected maximum

control force of the full-scale system can be esti- mated to be proportional to the dimensions of the

structure and the intensity of the ground motion.

The maximum control force normalized with re- spect to the weight of the structure was 9.5%

for the 1:4 scaled model associated with a 40%

reduction in relative displacement under the 25%

El Centre earthquake with an appropriate change

in time scale. It was assumed in the scaling of

the active system that, for the same degree of

reduction in response, the force to weight ratio remained the same for the full-scale structure if it is subjected to earthquakes of the same intensity as used in the scaled model tests. Accordingly, the

expected maximum control force for the full-scale

model weighing 600 metric tons was estimated to be about 7 12 kN under 32% El Centro earthquake.

Based on these estimates and the available manufactured imits, two 35-ton hydraulic actuators were chosen in each direction. In other words,

the active system can provide 685 kN maximum

control force uniaxiaUy.

3.2 Analysis and Design

Having determined the specification of the hydraulic actuators, further analysis of the control

system can be performed based on the associated

parameters. A stiffness of the diagonal brace (including the actuator) of 6 1 .3 kN/mm was assumed

in the simulation in order to estimate the actuator displacement and velocity. Changes in the

structural eigenproperties due to implementation

of additional braces are insignificant, therefore

the influence of bracing on the structural dynamics was neglected in the analysis. The flow rate requirement of the actuator or the hydraulic ser- vovalve was in turn calculated from the piston

area of the cylinder and the actuator velocity

based on a first-order approximation. Analytical

simulations were performed only in the strong

direction for design purposes and the same design

specifications were used in the other direction.

3.2.1 Design of Passive Povi^er Resource

The required flow rate of the hydraulic cylinders can be determined approximately in terms of the

piston area and the actuator velocity. Figure 4 il- lustrates the cumulative flow accumulated during an earthquake, which is obtained by integrating

the time history of the flow rate. The slope of

this curve represents the instantaneous flow rate required to achieve the control goal. It is observed from interpreting the slope of the curve

that system demand is the highest between 2 and

5 seconds and less so over the rest of the time

history, a property apparently resulted from the

nonstationary nature of the earthquake motion

and the control effect contributed in the previ- ous time period. The linear curve represents the

cumulative volume of a constant flow which is obtained by minimizing the difference between the

demand and supply of oil using the least-square

criterion. The largest difference between the cu- mulative flow and the average flow indicates the

minimum volume of the hydraulic accimiulator to be considered in design.

161
3.3 Specification

3.3.1 Braces

The design of the braces is based on the maximum control force and the anticipated stiffness with the assurance that buckling will not occur under actuator actions. Circular steel tubes were

used as bracing members with the following specifications: length = 360.5 cm, diameter = 165.2

mm, thickness = 4.5 mm and strength = 564 kN.

3.3.2 Hydraulic Actuators

Four units of PARKER, heavy duty hydraulic

cylinder series 2H - style TC (NFPA style Mx2)

were selected as controllers with the following

specifications: length = 73.5 cm, piston diameter = 152.4 mm, rod diameter = 63.5 mm, stroke = ±5 cm, and average capacity = 344 kN.

The average capacity of the actuator is calculated

from the working pressure (3,000 psi) of the

hydraulic oil and the average piston area, i.e., average of the piston area on one side and the same area minus the rod area on the opposite side

of the piston. The capacity can be improved by

increasing working pressure of the hydraulic oil.

3.3.3 Hydraulic Servovalve Manifold

Apair ofhydraulic actuators is monitored by a unit

of servovalve controllers in each direction. Hydraulic accumulators are provided to regulate oil supply without additional pumping during earth- quake motions (within 1 min. in general). Only

1/3 of the accumulator volume contains hydraulic

oil, the rest is filled with nitrogen. The mechanical system gains substantial pressure with

only a slight volume change of the nitrogen. A

margin was added in sizing the accumulator, in

addition to the calculated quantity shown in Table

2, to account for energy loss due to pressure drop

and error due to the first-order approximation,

respectively.

Two units of MTS model 252.2x servovalves

with flow rating of 15 gpm (55 1pm) at 1000 psi

pressure drop were used accompanied with two

five-gal. pressure accumulators and two two-gal.

return accumulators.

3.4 Verification

From the analytical data shown in Table 2, it was found that the design was appropriate for

almost all earthquake records used except for the case of Hachinohe earthquake. In that case, it requires a maximum control force of 696.5 kN

which is far beyond the design capacity of the

system if the same control strategy is to be used.

Investigation was made by restricting the output

control force within the design capacity of the

actuator while using the same feedback gains.

Results show less reduction of structural response

(about 30% in displacement and 47% otherwise)

when the control force is restricted to 333 kN

per actuator whereas stability of the mechanical

system is preserved.

This situation should also be taken into account in

the on-line control practice due to erratic nature of earthquake ground motions.

4. EXPERIMENTAL PERFORMANCE OF

CONTROL SYSTEMS

Recently, the construction was completed and

initial tests were performed on the site. The tests included three stages: (a) control of microtremor

excitation due to traffic and wind; (b) control

of free vibration of the structure after excitation

produced by the use of the active mass damper as a force exciter; and (c) control of forced vibration

using the active mass damper exciter. Topical test results are presented herein as an illustration of

the system performance.

Figure 5(a) shows the uncontrolled structural re- sponse in both directions. The decayed rate of the

uncontrolled response is very slow because the

system is only slightly damped (1% in the strong

direction and 0.5% in the weak direction). Figure

5(b) shows the build-up of the structural vibration

and then the influence of control applied at the

peak of excitation. The increase in damping is evident, showing, for example, an increase from

0.5% to 3.5% in the weak direction. The influence

of control during forced vibration is shown in Fig.

5(c). The response builds up in the transient stage

and reaches its steady state, which is evidendy

smaller than the previous tendency as shown in

Fig. 5(b).

162
5. CONCLUSIONS

Presented in this paper are design, simulation and

performance of a full-scale active bracing system

which was recently installed in a full-scale test structure. All components of the system were

selected from currently available hardware and

the power and energy resources required for this structural application were found to be within

limits of current technology.

Analytical simulation and initial structural tests indicate that the system can produce signifi- cant structural response reduction. While more

widespread applications of active structural control systems must await more extensive analytical

and experimental verification, results obtained

to date have been very encouraging and more

full-scale demonstrations are expected to confirm

these findings.

6. ACKNOWLEDGEMENT

This woric was supported in part by the National

Center for Earthquake Engineering Research imder Grant Nos. NCEER-89-2201 andNCEER-90-

2201. The authors are indebted to MTS Systems

Corporation, Takenaka Corporation and Kayaba

Industry Co., Ltd. for their technical assistance,

consultation and financial support.

7. REFERENCES

1. Kobori, T. (1990), "State-of-the-art of Active

Control Research in Japan," Proc. National

Workshop on Structural Control, Los Angeles. CA.

2. Reinhom, A.M.andSoongT.T.,etal. (1989),

"1:4 Scale Model Studies of Active Tendon

Systems and Active Mass Dampers for Aseismic Protection," Report NCEER-89-0026,

National Center for Earthquake Engineering

Research, Buffalo, NY.

3. Soong, T.T. (1990), "State-of-the-art of Active Control Research in the U.S.," Proc. National Workshop on Structural Control,

Los Angeles, CA.

4. Soong, T.T. (1990), Active Structural Control: Theory andPractice, Longman, London

and Wiley, NY.

5. Soong, T.T. and Reinhom, A.M., et al. (1991), "Full-scale Implementation of Active

Control, I. Design and Simulation," ASCE J. Structural Engineering (in press).

O

in eo

d

o

3 e

Actuators

1 n

A (

Hydraulic

tzX 3 E

& Control

Equipment

z 5.0

J. 5.Q

IQ.Q m

-7

(a) Top View

10.0 m

(b) Front View

Figure 1. Test Structure and Active Bracing System
Table 1. Modal Frequencies of Test Structure

Morle

X-direction Y-direction

Rpcord

(Hz) (Hz) (Hz)

/-viicuy bis (Hz)

1 0.900 0.922 0.650 0.643

2 2.700 2.703 1.800 1.750

3 4.650 4.737 2.975 2.918

4 6.850 6.973 4.125 4.117

5 9.400 9.481 5.425 5.511

6 12.75 13.03 7.250 7.623

Table 2. Summary of Response Analysis Under Design Earthquakes

Performance indices Resource requirement / Actuator

Eartliquakes Scale

factor Top fl. disp. (cm)

Top fl. acc.

(g)

Base

siiear (kN)

Control

force (kN)

Actuator

disp. (cm)

Max.

flow

rate (Rpm)

Ave.

flow

rate (gpm)

Vol.ofa

ccuni.

(gal.)

Total volume

of oil (gal.)

Power

(kW)

EI Centre 0.32

Unctrl

Ctrl Red.(%)

7.98 4.66 41.60.37

0.23

36.5

1018.3 618.0

39.5 313.9 0.51 19.72 2.7 0.11 0.74

in 20 s 20:02

Miyagioki 0.68

Unctrl

Ctrl Red.(%)

5.30

3.27

36.4

0.43

0.21 50.7

847.6

547.4

35.8 245.3 0.40 15.24 2.0 0.12 0.71

in 30 s 16.08

Mexico

(N90W)

0.65

Unctrl

Ctrl Red.(%)

9.70

5.80 40.2

0.32

0.18 44.2

1347.9 877.0

35.0 255.1 0.41 6.21 1.2 0.18 1.33

in 100 s 5.65

Mexico

(SOOE)

1.115

Unctrl

Ctrl Red.(%)

9.95 6.99

29.8

0.34 0.24

30.4

1389.1 1000.6

28.2 323.7 0.53 9.36 1.5 0.34 2.13

in 100 s 10.77

Pacomia Dam

(S16E)

0.095

Unctrl

Ctrl Red.(%)

5.58 3.90

30.1

0.35

0.17

50.7

741.6

500.3

32.7 255.1 0,42 10.94 2.0 0.06 0.47

in 20 s 10.06

Pacomia Dam

(S74W)

0.104

Unctrl

Ctrl Red.(%)

4.14

2.68

35.1

0.30

0.14

52.6

606.3

406.1 33.0

176.6 0.29 9.35 2.0 0.055 0.44

in 20 s 4.88

Taft (N21E)

0.715

Unctrl

Ctrl Red.(%)

7.64 4.36

42.9

0.63 0.30

53.3

1083.0 535.6

50.5 333.5 0.54 18.29 3.0 0.12 0.96

in 20 s 21.48

Tokyo 1.48 Unctrl

Ctrl Red.(%)

4.85

3.73 23.2

0.33 0.26

21.1

623.9

406.1

34.6 333.5 0.54 12.63 2.7 0.07 0.37

in 10 s 21.60

Hachinohe 0.60

Unctrl

Ctrl Red.(%)

17.93 9.48

47.1

0.79

0.41 48.6

2731.1

1471.5 46.0 696.5 1.13 26.74 6.0 0.17 1.51

in 20 s 59.02

Hacliinoiie (limited Ctrl force) 0.60

Unctrl

Ctrl Red.(%)

17.93 12.65 29.4

0.79

0.69

12.5

2731.1 1660.8

39.2 333.5 0.54 31.04 4.0 0.17 1.21

in 20 s 55.54

164
0 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 TIME (sec) UncontroQed Controlled

0 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15

25.0

0.00 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15

8.00-

0.00

0 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 TIME (sec)

Figure 2. Structural Response Under 32% Figure 3. Control Requirements Under 32%

El Centro Earthquake - El Centro Earthquake

165
0 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15

TIME (sec)

Figure 4. Cumulative Oil Flow During 32% El Centro Earthquake

166
(a) Uncontrolled Free Vibration (b) Controlled Free Vibration

TIME B LIN

X; 12.734 SEC Y: -.7831 V BOSEC

(c) Controlled Forced Vibration

Figure 5. Preliminary Test Results of Full-scale Structure

167

Base-Isolated Buildings and Seismic Safety Evaluation in Japan

By

Yoshikazu KITAGAWA i and Mitsumasa MIDORIKAWA2

SUMMARY

A total number of 59 design plans on

base-isolated building structures have obtained

the required special permissions in Japan by

the end of March, 1991. This paper presents

the characteristics, present state and safety

evaluation guideline of base-isolated building

structures. KEY WORDS: Base-isolated Building

Structure, Characteristics of Base-isolated

Building Structure, Seismic Safety Evaluation.

1. INTRODUCTION

Japan is characteristically subjected to seismic activity, therefore, it is a matter of concern how building structures behave under

earthquake conditions. Currently the research

focus for a seismic design method has been

changed from static analysis to dynamic

analysis.

Many studies and proposals concerned

with the base isolation system have been

carried out. Some of those studies and

proposal were realized by actual structures in

New Zealand, France, U.S.A., Japan and

other countries. Recently, base-isolated

building structures have been proposed and

enthusiastically put into practice in Japan,

although the ball bearing system was proposed

by Dr. K. Kito in 1924.

In this paper, characteristics, present state and safety evaluation guideUne of base-isolated

building structures are briefly presented.

2. CHARACTERISTICS OF BASEISOLATED BUILDING STRUCTURES

There is a recent technical trend that

introduces the base isolation method to diminish the seismic load. The ultimate

objective of base-isolated structures is that the

structure not be destroyed. In that sense, baseisolation may be said to be method for making

aseismic structures. However, its aim of

obtaining earthquake resistance by means of

reducing the seismic load on the structure

makes the base isolation system slightly

different from other types of earthquake

resistant design methods in that the increased

strength and ductility of the structure actually

improve its earthquake resistance.

If we assume that ball-bearings can be

easily placed between perfectly flat plywood

slabs, the resulting structure will possibly have

a natural period of unlimited length. This

implies a static point so that the structure will

not move due to horizontal earthquake ground

motions. However, since the structure is actually supported at the points where it touches the ball-bearings, the response force

concentrates there. This leads to a supporting

point, not allowing a static point to be formed.

Since there is no restoring force, there is no

guarantee that the structure, after the

earthquake ends, will remain its original

position. There are several possible means to secure a restoring force, such as the use of

both bearings and springs, or roughening the

steel surface. With an accompanying restoring

force, the resulting structure will possess a

longer, but not infinite, natural period.

The base isolation system thus tries to reduce input earthquake ground motion by

means of lengthening the overall natural

period. Nowadays, laminated rubber bearings

composed of thin layers of steel plates and

rubber are primarily used. This type of

systems is firm against vertical loading and is horizontally flexible (see Fig. 1). Energy

absorption mechanisms, in use today, include

combinations of laminated rubber and various

kinds of dampers, (e.g., elasto-plastic, viscoelastic, viscous, friction), laminated rubber

containing lead, high-damping laminated

rubber, as well as a combination of elastic

sliding bearings that use low-friction Teflon

and horizontal dampers (see Table 1). In this

last case, the overall restoring force

characteristics are determined by the friction of

the sliding bearings and the restoring force of

1) Dr. Engrg., Director of Production Dept.,

Building Research Institute, Ministry of

Construction.

2) Dr. Engrg., Head of Earthquake

Engineering Div. H, USEE, BRI, MOC.

169
the dampers. The natural period of the overall

system is able to be adjusted.

Base-isolated building structures have

two goals: first, to reduce transmitted

earthquake motion by means of lengthening the

overall natural period; second, to absorb the

earthquake ground motion energy during

severe earthquakes by the use of various types

of dampers (see Fig. 2). As a result, the baseisolated building structure itself behaves as a

rigid body at time of an earthquake. Also,

since it forms a so-called "non-vibration

space", it does not suffer any story drift as do

other structures which are directly lined to the

ground (see Fig. 3). This has the effect not

only of reducing input earthquake ground

motion to the structure during an earthquake,

but also of increasing added effect to the

structure. The base isolation system is beneficial to buildings that require a more variety of

possible construction sites and/or improved

safety and so on. Such building might include

museums, art galleries, socially and culturally

important buildings containing motion sensitive

equipment, such as built-in sophisticated

computer network systems, hospitals with

expensive medical facilities and equipment,

medical-care centers, hotels and condominiums

needing a variety of facades, low- and

medium-rise apartment buildings containing

personal property as well as energy-related

buildings such as nuclear power stations.

3. PRESENT STATE OF BASE-ISOLATED

BUILDING STRUCTURES

Currently, construction of base-isolated

building structures in Japan requires the special

permission from the Minister of Construction

to fulfill Article 38 of the Building Standard

Law. The reason is that the base isolation

system is a new technology where the design

standards and specifications are not yet fully

documented.

The first step in the building process

requires that the design be examined by the

Base Isolation Research Committee (Chairman:

Prof. M. Izumi, Univ. of Tohoku ) in the

Building Center of Japan. This committee

specializes in examining design criteria and

specific base isolation system mechanisms to determine whether or not the base-isolated

building structure has attained a safety level as a total system. The committee then states the

necessary recommendations for compliance.

The next step in the process is to have the

planned structure examined by the Baseisolated Building Structure Appraisal

Committee (Chairman: Prof. Y. Inoue, Univ.

of Osaka). The planned structure is inspected

from technical point of view as to its structural

strength, safety, the durability and maintenance

of the base isolation device. In the case where

the planned structure is equipped with a similar

base isolation device previously examined by

the Base Isolation Research Committee, the

Base-isolated Building Structure Appraisal

Committee examines only the base isolation

system in question.

By the end of March, 1991, a total

number of 59 structure design plans have

obtained the required special permissions.

Construction companies have rapidly been

putting these plans into practice. Experiments,

earthquake observations and various types of

analysis have been carried out on such baseisolated building structures. The results have

been presented at a conference sponsored by

the Architectural Institute of Japan, at the

World Conference on Earthquake Engineering

as well as other presentations.

Fig. 4 and Table 2 show an example of

design procedures of base-isolated buildings,

and an example of design criteria on earthquake

response of a base-isolated building structure,

respectively.

4. SAFETY EVALUATION GUIDELINE FOR

BASE-ISOLATED BUILDING

STRUCTURES

The Base Isolation Safety Evaluation

Committee (Chairman: Prof. H. Aoyama,

Univ. of Tokyo) established by the Building

Center of Japan proposed the safety evaluation

guideline for base-isolated building structures

3) in 1989.

This guideline applies to a base-isolated

building structure that is constructed on either

hard or intermediate soil and with a height less

than 45 meters. However, application of part

of this guideline may be waived for a baseisolated building structure which was designed

in accordance with special investigations. This

guideline requires that target performances in

six areas be carried out in the following areas: (1) overall seismic safety; (2) overall wind

safety; (3) overall safety in architectural details

and equipment design; (4) overall durability of

isolation devices; (5) overall fire resistance of

isolation devices; and (6) overall construction

and maintenance of the isolation devices.

170
Base-isolated building structures are

classified based on their response to level-2

earthquake ground motions 2) (see Table 3). A

level-2 earthquake ground motion corresponds

to a probable maximum earthquake event

currently defined as earthquake ground motion

at a ground velocity of 50 cm/sec in the Tokyo

area. Buildings in each category are reviewed

with accordance to safety consideration. The

notation A"*", A', B"^, B" and C in Table 2

correspond to the safety evaluation levels: a

level A"*" design is considered to be the most

conservative design, hence inherently safe,

whereas a level C design is the least

conservative. Differences between these safety

evaluation levels result in requiring which

safety evaluation items be addressed. The

required items may include, but are not limited

to, the following: architectural design, soil

condition, structural design, isolation devices,

earthquake response analysis, analytical model,

natural period, earthquake ground motions,

evaluation for level-2 earthquake ground

motions, and/or evaluation of safety margin. A

level A"*" design will require the least number of

items, while a level C design will require the

most.

Article 83 of the Building Standard Law

specifies the combination loads used in

determining a design load acting on a baseisolated building structure with the exception of an earthquake load. This guideline specifies

the calculation method for the earthquake load

case.

First, an earthquake load acdng on the

superstructure is to be calculated using the

following relationship.

Q = a-Z-BAi-BCo (1)

where,

Q = design story shear coefficient of the

superstructure;

a = isolation characteristics factor

reflecting the base isolation effect; Z = seismic zone factor;

sAi = distribution factor of story shear

force coefficient along the superstructure (bAj = 1 at the lowest story of the superstructure);

and

bCo = standard shear coefficient to be

taken equal to 0.2.

Second, an earthquake load acting on the

substructure is the sum of earthquake load

acting on the superstructure as determined by

Eq. (1) and the earthquake load acting on each

part of the substructure. The latter is determined by the procedure with the exception

when isolation devices are installed in an

intermediate story.

Lastly, for the case when isolating

devices are installed in an intermediate story an earthquake load is determined by an

appropriate analysis that accounts for the effect

of response amplification of the substructure to both the isolation devices and superstructure as well as the transfer of the earthquake load

from the superstructure to substructure.

The effect of base isolation is best

reflected by an appropriate response analysis of

the structure to earthquake-type base excitation.

The coefficients a and gAj are, thus, to be

determined by the preliminary response

analysis to level- 1 earthquake ground motions.

This level corresponds to an once-in-a-lifetime

earthquake event defined as a maximum

ground velocity of 25 cm/sec in the Tokyo

area. In general, a base-isolated building

structure must be shown to respond to two

levels of earthquake ground motions ^) for

target performance in earthquake resistance.

First, during a level- 1 earthquake, the

superstructure must remain in the elastic range,

and the isolation devices must remain within

stable deformation (see Fig. 5). Also, during a

level-2 earthquake, the superstructure remains

wiUiin a general yield of the stories (or within a permissible ductility factor), and isolation

devices must remain within a performanceinsured deformation.

The degree of safety allowance must be

investigated for all base-isolated building

structures except for a level A-i- design in either case where the superstructure is expected to absorb energy by inelastic deformation, or where the isolating devices undergo

deformations larger than stable deformation.

Confirmation of the safety allowance is necessary as the level-2 earthquake response

analysis does not necessarily cover all the

possible unfavorable conditions. These

include uncertainty in the input earthquake

waveform, the analysis method and the

analytical model. Particularly important is the

uncertainty in stiffness evaluation of laminated

rubber bearings, variation due to

manufacturing conditions, hardening or

softening in the large deformation range, and

eccentricity. Also important is the evaluation

of load variation due to earthquake ground

motion. Laminated rubber bearings are currently designed in such a way that no

tension would be allowed even in case of a level-2 earthquake. However, in the

171
confirmation of safety allowance, performance

of isolating devices beyond this limit is investigated.

5. CONCLUDING REMARKS

This paper has presented a brief oudine

of: (1) characterisdcs of base-isolated building

structures; (2) present state of base-isolated

building structures; and (3) safety evaluation

guideline for base-isolated building structures. When the guideline is finally approved

and in effect, a relatively conventional and

conservative base-isolation design can be

reviewed directly by the Safety Evaluation

Committee. The state of a base-isolation

system is sdll in its development stage, and a general social consensus for seismic safety of

this kind of structure as a new technical trend is yet to be developed.

REFERENCES

1) "Report on Durability and Reliability of

Laminated Rubber Bearings for Base-isolated

Structures", the Society of Rubber Industry,

Japan, 1988 (in Japanese).

2) High-rise Building Structure Appraisal

Committee, "Earthquake Motions of Highrise

Building for Dynamic Analysis", the Building

Letter, No. 203, the Building Center of Japan,

June, 1986, pp. 49-50 (in Japanese).

3) "Report on New Building Construction

Technology-No. 2, Safety Evaluation

Guideline for Base-isolated Building

Structures", the Building Center of Japan,

1989 (in Japanese).
Table 1 Classification of Isolating Devices"

Isolating

Bearing

I— Laminated

Rubber Bearing

Slide

Bearing

Damper —

Hysteretic

Damper

Velocity-Type

Damper

— Natural Rubber Bearing

Lead Rubber Bearing

^

High Damping Rubber Bearing

Elastic Slide Bearing

Rigid Slide Bearing

Bar Type — Loop Type

Portal Type — Plate Type

Ring Type

— Steel Damper —

— Lead Damper — Friction Damper — Oi 1 Damper — Viscous Damper

Table 2 Example of Seismic Response Criteria

Earthquake

Level

Seismic Performance

Superstructure Laminated Rubber Bearing

1

(25cm/s)

Not exceeding

Al lowable Stress

Less than 100% of Shear

Strain

2

(50cm/s)

Not exceeding

Initial Yield

Strength

Less than 200% of Shear

Strain

Earthquake for

Confirmation

of Safety

Al lowance

(50^75cm/s)

Not exceeding

General Yield

Strength

Less than 300% of Shear

Strain

173
Table 3 Classification of Base- Isolated Building under Level 2 Earthquake

Charcteristics of

^^^~---,^s0 1 a t i n g Char ac t er i s t i c s^~^^.^^ Dev i ce of Superstructure

B- I Within Stable

Deformat ion

B- n

Within Performanceinsured Deformation

S - I In Elastic Range

s - n

Within General Yield of

Stories

A" B"

S -ffl Within Story Mechanism C C

174
Steel Plate

Rubber —

Verticad Loading

Horizontal Loading |

(Original form) /Under the horizontal.)

[ Vertical Load / Fig. 1 Laminated Rubber

h = 0. 0 5

Increase in Damping

h = 0.

O Structure set up on Ground

# Base-Isolated Structure

•T (sec.)

to

h = 0. 0 5

h-Lengthening

I Natural Period i Increasing in tlie damping

h = 0. 2 0

To T,-^T (sec.)

Fig. 2 Schematic Characteristics of Base-Isolated Structure

Non Vibration Space No Storey Drift

1^

1 1 /1/ u-i; 1212

G.L. .-J

Fig. 3 Dynamic Behavior of Base-Isolated Structure

175
START

Grasp of General Concept

Selection of Design Criteria

Structural Plan

Design of isolating Device

Static Design

Nk Seismic Response Analysis

YES

Overall Construction and

Maintanance of isolating

Devices

END

Fig. 4 Flowchart of Design Procedure

176
Fig. 5 Schematic Restoring Force Characteristics of Laminated Rubber Bearing''

177

Dynamic Modeling of Soil-Structure Systems for

Control Applications

by

H. A. Smith* and R. I. Borja*

ABSTRACT . Much attention has been focused recently on the use

of actively controlled structural systems in seismic

zones. However, the majority of the models utiUzed

in this research are highly simplified representations

of the structural and foundation system, and usually

do not consider model nonlinearities or complexities

such as soil-structure interaction.

This paper presents and discusses a modeling

technique for studying the effects of soil- structure

interaction (SSI) on the active control of seismically

excited structures. Instantaneous optimal control

algorithms are used to formulate a system of

equations applicable to both linear and nonlinear SSI

systems. The resulting control equations of motion

are solved in the time domain using a direct

integration scheme which can be combined with

Newton-Raphson iteration for nonlinear systems

This soil-structure interaction (SSI) model developed

for use in this study is represented using a coupled

finite element (FE) - boundary element (BE)

formulation which has the capability to resolve

fundamental difficulties such as radiation boundary

conditions, nonproportional damping, and hysteretic

behavior effectively. The foundation of the building

is assumed to behave as a rigid body with both

translational and rotational degrees-of-freedom.

KEYWORDS: active control; nonlinear; soil- structure interaction; earthquake; time-domain

1. INTRODUCTION

In recent years, considerable research efforts have been

placed on the appUcation of active control systems to structures in seismic zones. Many of these structures exhibit inelastic or nonlinear response which aids in the dissipation of energy but causes damage and

unserviceability when excessive. Active control is used to ensure that these structures do not undergo

extensive and prolonged deformations which could

result in destruction.

Various control algorithms and systems, such as the active tendon control systems and active tuned mass

dampers, have been investigated for tall structures undergoing earthquake excitations. However, most

of the methods studied thus far have involved rigidbased structural models and rarely are the issues of

nonlinear behavior and soil-structure interaction

effects addressed.

Accurate representation of soil-structure interaction (SSI) effects is a crucial part of earthquake

engineering analyses. To date, little work has been

done to study the influence of SSI effects on the

effectiveness of control algorithms. The properties of

the underlying soil can significantly influence

structural behavior and consequently alter the performance of a control algorithm formulated with

the assumption of a rigid structural base. In this study, a nonlinear SSI model is developed for use

with an instantaneous optimal control algorithm.

The purpose of the research is to determine how soil- structure interaction effects influence the active

control of buildings subjected to strong ground

motion. In Section 2, an overview of the previous

work on nonlinear optimal control theory is given.

Section 3 formulates the equations governing optimal

control of the linear and nonlinear systems, and

Section 4 describes the SSI model developed for use with the control algorithm. A summary of the work

is given in Section 5,

2. OVERVIEW OF ACTIVE CONTROL

FOR STRUCTURES WITH RIGID

FOUNDATIONS

2.1 General Theory

For a linear, actively controlled one-dimensional

structure represented using an N degree-of-freedom

finite element model, the equation of motion takes

the form:

My(t) + Cy(t) + Ky(t) = Hu(t) + FeU) (1)

where M, C, and K, are the system mass, damping,

and stiffness matrices respectively; y(t) is the vector

* Department of Civil Engineering, Stanford

University, Stanford, California USA 94305-4020

179
of system response; H is the matrix relating the locations of the control actuators to the system

degrees-of-freedom; u(t) is the vector of control

forces; and Fg(t) is the vector representing the

external excitation. Equation 1 can be transformed

from a set of N equations of motions into a set of 2N

state equations in the form:

z(t) = Az(t) + Bu{t) + WFE(t) (2)

where z(t) is the state vector; A is the system matrix;

B is the matrix specifying the locations of the control actuators; and W is the matrix specifying the

locations of the applied external forces.

Control algorithms for linear systems are well

developed and extensively studied. Optimal control algorithms are based on the minimization of a

quadratic performance index whose objective is to maintain the desired system state while minimizing

the control effort. For a linear system, this performance index takes the form:

J = [z(t/Qz(t) + u(t)'^Ru(t)] dt (3) Jo

where Q and R are the weighting matrices dictating the relative importance between the response state of

the system, z(t), and the control forces, u(t), and tj is the duration of the earthquake.

In 1987, Yang et al. [Ref. 14] showed that the optimal control theory based on Eq. 3 is not

applicable to seismically excited systems because the time-history of the excitation is not known a priori.

As an alternative, they introduced instantaneous

optimal control algorithms which are derived by

minimizing a time-dependent quadratic performance

index at every time step. This instantaneous

performance function takes the form:

J = [z(t)'^Qz(t) + u(t)'^Rii(t)] (4)

These instantaneous optimal control algorithms were

derived for open-loop, closed-loop and closed-openloop control systems. Yang concluded that all three instantaneous control algorithms yield identical

results in an ideal control environment.

The equation of motion governing the response of an

actively controlled system undergoing inelastic or

nonlinear deformations takes the form:

My(t) + Fi(t) = Hu(t) + WFE(t) (5)

The internal force, Fi, is a function of y(t) and y(t)

[Ref. 2] and is found from the expression:

Fi(t) = B a dV (6)

where B is the matrix relating strains to displacements and a is the vector of internal stresses. Methodologies for the control of nonlinear systems

have been studied less extensively than active control of linear systems. The remainder of this section

presents the present state-of-the-art in active control research for systems behaving nonUnearly.

2.2 Review of Control Algorithms for

Nonlinear Structures

In 1981, Masri et al. [Ref. 13] presented a method

which utilized portable gas pulse generators to actively control structural response. This control algorithm continuously monitors the system state and

applies a series of force pulses when a specified

threshold value is exceeded. The determination of the pulse magnitude is based on a performance criteria which is linearly dependent on the deterministic component of the structure's response. Both linear and

nonlinear MDoF systems subject to deterministic or stochastic response can be analyzed with this optimum pulse control method. Masri et al. [Ref. 8]

later developed an on-line control method also using

pulse generators. This methodology, applicable to linear and nonlinear systems, determines the pulse

magnitude based on the instantaneous magnitude of

the structural velocity.

A non-optimal active control algorithm for inelastic (bilinear) SDoF systems was developed by Reinhom

et al. [Ref. 10, 11]. This closed-loop algorithm is based on a reformulated version of the Newmark-beta

time integration scheme in conjunction with the Newton-Raphson method. Here the active control mechanism is triggered when nonlinear response

exceeds a prescribed limit which would render the system unserviceable. Like the previous work of

Masri et al., this methodology also utilizes a pulse

control system.

Abdel-Rohman and Nayfeh [Ref. 1] used a combination active/passive control system to control nonlinear oscillations of a single-span bridge. A

truss tendon is used to provide both the passive

control (increasing stiffness and shifting frequencies

away from resonance) and active control (increasing

damping to limit peak amplitudes). The magnitude

of the active control force is proportional to the bridge velocity at midspan.

180
In a follow-up to his earlier work, Yang et al. [Ref.

15] proposed the use of the instantaneous control algorithms for nonlinear systems with bilinear

elastic-plastic characteristics. The time-dependent

stiffness and damping terms of the equation of motion

are approximated using influence coefficients

representing the tangent stiffness and damping

matrices. A solution to the resulting set of nonlinear

equations is found using the Wilson-9 numerical

integration scheme. This methodology is successfully proven using examples with both active and hybrid (active/passive) control systems. Yang

proposes combining an adaptive control algorithm

with an instantaneous control algorithm to account

for system degradation in the control process for nonUnear systems.

In 1988, Kamat [Ref. 6] developed a control

algorithm for flexible spacecraft with both geometric

and material nonlinearities. This method is based on

an extension of the Davidon-Fletcher-Powell

algorithm for multi-variable function minimization

previously studied by Tripathi and Narendra [Ref. 13]. The Newmark-beta average acceleration scheme is used to integrate the nonlinear equations and a

globally convergent Newton-Raphson method is used

to solve the resulting nonlinear algebraic equations.

Nagurka and Yen [Ref. 9] convert the optimal control problem to an algebraic nonlinear programming

problem using a Fourier series approximation of each

generalized coordinate. This method generates near optimal trajectories of nonlinear dynamic systems

represented by deterministic, lumped-parameter

models.

Nonlinear optimal control for Duffing oscillators was studied by Suhardjo, et al. [Ref. 12]. Duffing

systems primarily occur in systems with large

deflections exhibiting geometric nonlinearities. Here

the nonlinear optimal regulator problem is examined

and represented using indicial tensor notation and the resulting control methodology is used to study the Duffing systems.

Though all the previously discussed research addresses

active control algorithms for structures behaving

nonlinearly or inelastically, none of the work

specifically discusses the issue of soil-structure

interaction. For this reason, the following section

formulates the instantaneous optimal control

algorithm [Ref. 23] developed specifically for application with nonlinear soil-structure systems.

3. AN OPTIMAL CONTROL THEORY

FOR NONLINEAR SOIL-STRUCTURE

SYSTEMS

3.1 Control Theory for Nonlinear Systems

In this study we formulate a control theory for nonlinear soil-structure systems as a constrained

minimization problem. As previously presented, a time-dependent quadratic objective function J(t) for instantaneous optimal control in the context of Yang

et al. [Ref. 14] is assumed.

We seek to minimize the performance index at t =

tn+l>

J (tn+l) = z(tn+l)^Qz(tn+l) + u(tn+l)'^Ru(tn+l) (7)

subject to the constraint (i.e. the equation of motion

at t = tn+i as stated by Eq. 5)

My(tn+l) + Fi(tn+l) = Hu(tn+l) + FE(tn+l) (8)

For convenience we write the performance index of

Eq. 7 expUcitly in terms of the individual elements of

the state vector z(tn+i). Since J(t) is a quadratic

function, the performance index can also be written in the following form:

J M= y(tn+i)^Qiy(tn+i) + 2y(tn+lfQ2y(tn+l)

^g-^ + y(tn+lfQ3y(tn+l) + U(tn+lfRu(tn+l)

where the weighting matrix Q is written in partitioned form as

. Q2 Q3 .

Note that the minimization of J(t) is performed

instantaneously at each time station, as discussed by

Yang [Ref. 14]. Equations 7 through 9 represent a general instantaneous optimal control algorithm

applicable to both linear and nonlinear soil-structure systems.

Equation 8 represents a numerical integration of the hyperbolic dynamic equation of motion in time. In

practice, time-stepping of a second-order ODE may

be performed effectively using either the Newmark

method or any of its variants.

One can establish the equivalence of the control algorithm given by Eqs. 7 and 8 with the nonlinear

control algorithms proposed by Yang et al. [Ref.

15] by applying the Taylor series expansion of the nonlinear constraint (Eq. 8) about the current time

181
station tn. Upon ignoring tiie second-order terms and

setting FE(tnf i) = - M vxg(t)^ the expansion takes the form

M[y(tn) + Ay] + Fi(tn) + C*{QAy

+ K*(tn)Ay = -M vxg(tnvi) + Hu(trHi)

(11)

where C (tn) and K (tn) represent the tangent

damping and stiffness matrices at t^. These matrices can be represented as

C (tn) = aFi(tn) / ay(tn)

K*(tn)=aFl(tn)/ay(t^)

(12)

and are the respective gradients of Fi evaluated at time station t^. By letting

-Mvx"g(tm.l) + Hu(tn+l) = -M v[xg(tn) + AxJ + H[u(tn) + Au]

(13)

and substituting into Eq. 11, the incremental timestepping algorithm of Yang et al. is obtained where

M[y(tn) + Ay] + Fi(tn)

+ C*(t^)Ay + K*{QAy (U)

= -Mv[xg(tn) + Ax'J + H[u(tn) + Au]

Clearly, Eqs. 8 and 14 are equivalent for the linear case. For the nonhnear case, the equivalence of the two algorithms can be established in the limit as the time step vanishes. Owing to the explicit form of

Eq. 14, however, numerical stability problems could

seriously afflict the Yang et al. algorithm in the nonUnear regime, particularly when the time step is large. Furthermore, without a sound iterative

strategy, the Yang et al. algorithm could fail to detect the onset of initial yielding or initial elastic unloading in elasto-plastic materials, which could be

crucial to the accuracy of the control algorithm.

Equation 8 is a generalized equation of motion and

includes effects of soil-structure interaction provided

that the equation of motion is imposed over a region

which includes not only the structure but also the foundation and the supporting medium. Thus, energy

dissipation due to radiation damping, which has no

counterpart in a rigidly supported structure, becomes

part of the constraint equation.

3.2 General Solution Algorithm

Let X(tn+i) represent an A^-dimensional vector of

Lagrange multipliers for optimal control at time

instant tn+i, and let the Hamiltonian function H be

defined by

(15)

H = y(tn.i)'^Qiy(t,^i) + 2y(t„.ifQ3?(t:Hi) + y(t.»fl)^Q3y(tm-l) + u(tnvl)'^Ru(t„,i)

+ M.tn,\)^ [My(ti^l) + Fi(trHl)

-Hu(t„^l)- FE(t„.l)]

Together with the equations that characterize a particular time-stepping algorithm we view the Hamiltonian as a function of the control forces

u(tn+i), the Lagrange multipliers M^+i) and, the displacement vector y(tn+i) (note that y(tn+i) and

y(tn+i) can be determined once y(tn+i) is computed).

The necessary conditions for minimizing the

performance index as defined in Eq. 9 subject to the

constraint imposed by the equation of motion are

3" ,0;^=0;^=0 (16)

3y(Wi) 3u(t„.i) dXwi)

Introducing the following (N xN ) gradient arrays where D represents the total derivative:

r(t„.0 = ^^; A(w,) = ^^

^y(ui)

and expanding Eq. 15, we obtain the following three madix equations:

dH

dyitnn) = 2Qiy(t„,i) + 2Q2y(t„,i)

-I- 2 A(t^i)^Q:y(WO + 2 A(t„,i)^Q3y(trK-i)

+ [m r(wi) + 3((t„.i)]\tn.i) = 0

(18a)

= 2Ru(tr,.i) - ifTJitr^l) = 0 (18b)

3u(tm-l) — = MyXtxHi) + Fi(tr^i) - Hu(t„.i)

dMlrwl) (18c)

-FE(tn.l) = 0

By solving Eqs. 18a, 18b, and 18c simultaneously, we obtain the instantaneous values of the vectors

u(tn+i), y(t„+i), and X(tn+i) •

182
Note that r(tn+i) and A(tn+i) are constant gradient

arrays, i.e., their elements are constant scalar

functions of the integration parameters for a particular

time-stepping algorithm. For the linear case ^t^+i)

is also a constant gradient array; however, !Kitn+i) is a

time-dependent function for the nonlinear systems. It -

is observed from Eq. 17 that for linear elastic undamped systems, f?C(tn+i) = K and for linearly

elastic damped systems, ^tn+i) = K + CA(tn+i).

3.2 Linear Optimal Control

Equations 18a, 18b, and 18c can be simplified for the

linear case by first writing Eq. 18a in a form suitable for evaluation of the Lagrange multipliers X(tn+i)

where

Xtr^i) = 2Uy(t:Hi) + 2Vy(tr>.i) (19)

with (assuming symmetric M and K)

U = - [r(tn.l)^M-h Jdtr^li \Q2^ A(Wi)^Q3]

V = - [r(tn.l)'^M + !Kitr^l)]~lQl+ A(t:^l/Q2]

Substituting Eq. 19 in Eq. 18b and solving

for u(tn+i),

u(tn.i) = R-'H'T;Uy(t:^i) + Vy(t,Hi)] (20)

The final recurrence equation is obtained by

substituting Eq. 20 into Eq. 18c where

My(tn.O + Cy(t:^i) + Ky(t„.i) - FE(t„.0

^^j) - HR-'H'^Uy(t^i) + Vy(t^i)] = 0

Together with a given time-stepping algorithm (e.g., the Newmark method), Eq. 21 is a linear system of

equations that can be solved in a straight forward manner in time.

3.3 Nonlinear Optimal Control

Nonlinearities require that the optimal control

problem be solved by iteration. Here, Eq. 21 is a

nonlinear function of y(tn+i). An efficient algorithm

for solving a system of nonlinear equations is by

Newton iteration. Equation 21 is put in the residual

form

My(tr».i) + Fi(tn+i) - FE(t,Hi)

^22) - HR-'H'T;Uy(trHi) + Vy(tr^i)] = r

where r is an A^-dimensional vector function of

y(tn+i). A vector y(tn+i) is sought such that r = 0. The purpose of Newton's method is to iterate over the function

r'{y(tr^i)ik}Ay = r{y(t,Hi)jfc}

1 ^'^^ where y(tnfi)*+7 «- Ly(tm.i)*- AyJ

(where the subscript k in Eq. 23 represents an

iteration counter) until becomes negligible. In

practice, the error tolerance represented by the symbol

||r|| is taken as the Euclidean norm. A major impediment to an efficient solution of Eq.

22 by Newton iteration is the dependence of the matrices on the (as yet unknown) tangent stiffness matrix !?^tn+i). An alternative iteration strategy that does not use this unknown tangent stiffness is provided by the so-called modified Newton iteration,

which requires the evaluation of an initial tangent

operator only; however, the convergence of this iteration is generally very slow. Another possible

approach is to take an approximate linearization of

the left-hand side of Eq. 23 and define the tangent

operator as

r'{y(t.^O*} = Mr(t:,.o + !?a:tn.i) ^^4)

-HR"^Hl^UA(tn.i) + V]

where the matrices r(t„+i) and A(tn+i) are assumed

fixed in the linearization. The convergence

performance of a Newton-like algorithm based on the approximate tangent operator given by Eq. 24 is presently under investigation.

4. APPLICATION OF CONTROL

ALGORITHM TO SOIL-STRUCTURE

SYSTEM

4.1 Overview of SSI Model

The equations formulated and discussed in the previous section are applicable to both linear and

nonlinear SSI systems. In this section the importance of incorporating SSI effects in earthquake

engineering models is discussed. The SSI model

developed specifically for use with the preceding

control algorithm is then presented.

When a structure is subjected to earthquake forces, the

structure's response is modified considerably by the dynamic deformations of the underlying soil. Accurate representation of soil-structure interaction is a crucial part of earthquake engineering analyses. A

183
large number of soil-structure models exist of varying

degrees of complexity. The choice of the most

appropriate model for a particular analysis is based on

the building-foundation system of interest, the desired

level of accuracy, and experienced "engineering

judgment."

A conventional first step in analyzing the dynamic

response of actively controlled MDoF structures subjected to seismic excitation is to input a onedimensional earthquake ground acceleration at the base

of the structure. While this approach may be

appropriate for structures supported on a rigid base

(such as bedrock), it is not realistic for structures that

rest on soft soils where interaction between the soil and structure could dominate the building response. For these cases, it is particularly important that the influence of SSI on actively controlled systems be

carefully evaluated.

In the present study, the actively controlled structure

is considered as part of a larger system which includes

the foundation and the supporting medium. Analysis

of a larger system allows modeling of radiation

damping and the rocking motion at the base which

could otherwise dominate the building response and

consequently alter the effectiveness of the control

algorithm.

The SSI model proposed for use with the previously

outlined control algorithm is shown in Figure 1. The total domain, Q is divided into a finite interior part, Q.K and an unbounded exterior part, Q^.

Nonlinearities are allowed to take place in Q}' but not

in Q.^ where the response is assumed to be strictly linear. Between the two subdomains is a nonreflecting boundary Xb that serves to absorb all scattered waves emanating from Q} during seismic

excitation. The non-reflecting boundary is not

necessarily regular; the only requirement is that it defines the boundary between the linear and the nonlinear regions. The finite element method is used

to solve the numerical problem in Q.^ while the boundary element technique is employed in Q^.

Thus, a typical mesh consists of mixed finite and

boundary elements as shown in Figure 1.

Seismic loads are input directly on the non-reflecting

boundary in the form of free-field excitation,

following Gutierrez and Chopra [Ref. 5] and Bielak

and Christiano [Ref. 3]. Here, the free-field response

is substracted from the total response resulting in a composite problem which is solved for the total field

in Q.^ and a scatter field in Q^. Material

nonlinearities are treated in the finite element

formulation using a bilinear constitutive model for the structure and classical elasto-plastic and elasto- viscoplastic constitutive models with isotropic and

kinematic hardening for soils [Ref. 4]. Timeintegration in is done using Newmark's fully

implicit scheme, while in time-marching is done

by numerical evaluation of the convolution integrals

in the time domain. The composite FE-BE system is then solved simultaneously by Newton-Raphson

iteration.

4.2 Some Notes on the Application of

Control Theory to SSI Problems

In analyzing soil-structiu'e interaction problems for control applications, we need to redefine the state vector z(t). Restricting our structural model to the case where the base of the foundation is assumed

rigid, the state vector can be defined in terms of the nodal velocities y(t) and displacements y(t) of the

structure relative to the motion of the rigid base. If rocking of the base is considered, then the relative motion may be taken with respect to the rigid-body

configuration of the structure compatible with the rotational motion of the rigid base. Hence, in principle, the performance index can be uniquely

defined.

Note that the translational and rocking movements of

the base are themselves unknown in a typical soil- structure problem. Thus, the forcing function F^it)

generally differs from the source-point inertia force, - Mvxg(t). However, using the method of Bielak

and Christiano [Ref. 3] for prescribing the seismic

input through a soil-structure interface, it is always

possible to write the equation of motion for the soil- structure system in such a way that the external force

is driven by the free-field motion. In that case, an

instantaneous control, theory can phrased such that the performance index is minimized at each time step

depending on the instantaneous motion of the free

field.

The control algorithms and SSI model proposed

herein are currently being implemented. Parametric

studies are being made which quantify die influence of nonlinear soil-structure behavior on control theory.

Consideration of system nonhnearities and analysis of

the combined FE-BE soil-structiu'e model in the timedomain allows accurate representation of fundamental

difficulties associated with SSI analysis.

5. SUMMARY

Despite the extensive work done recently in active control of seismically excited structures, little research has directly addressed the issue of the effects of soil-structure interaction. The structural response can be significantly affected by the properties of the underlying soil, which typically are not considered in

184
conventional control applications based on rigidfoundation systems. This paper formulates and

presents an optimal control algorithm for actively controlled soil-structure systems. The purpose of this study is to determine the effects of SSI models on the effectiveness of control algorithms.

An instantaneous control algorithm is formulated

which is applicable to both linear and nonlinear soil- structure interaction systems. The equation

governing the control of a nonlinear soil-structure system subjected to strong ground motion are formulated as a constrained minimization problem,

and a time-dependent performance index is presented.

A solution to the resulting equation is achieved by

incorporating a direct integration scheme with the Newton-Raphson technique.

The SSI model proposed for use with this control

algorithm is based on a boundary element (BE) - finite element (FE) formulation, and is capable of

overcoming the fundamental difficulties typically

associated with soil-structure interaction analysis such

as radiation boundary conditions. The foundation of

the structiu-e is assume to behave as a rigid body with

the ability to translate and rotate. Hence, the state vector z(t) in the control algorithm and performance

index represents the relative motion taken with

respect to the rigid-body configuration considering the rocking motion of the foundation.

Studies are underway which fully implement the

control model and algorithms presented in this paper. Results from these studies are expected to quantify the influence of soil-structure behavior on the

effectiveness of active control algorithms for

seismically excited systems.

6. ACKNOWLEDGMENT

Funding for this research was provided by a grant

from the National Science Foundation under Contract No. BCS-9058316, Presidential Young Investigator

Award.

7. REFERENCES

1. Abdel-Rohman, M. and Nayfeh, A.H. (1987).

"Active Control of Nonlinear Oscillations in Bridges." Journal of Engineering Mechanics. Vol. 113, pp. 335-348.

2. Bathe, K..J. (1982). Finite Element Procedures

in Engineering Analvsis. Prentice-Hall, New

Jersey.

3. Bielak, J. and Christiano, P. (1984). "On the Effective Seismic Input for Nonlinear Soil- Structure Interaction Systems." Earthquake

Engineering and Structural Dynamics. Fol. 12, pp. 107-119.

4. Borja, R.I. (1991). "Modeling the Monotonic

anc Cyclic Viscoplastic Soil Behavior."

Proceedings: Second International Conference on Recent Advances in Geotechnical Earthquake

Engineering and Soil Dynamics . University of

Missouri-Rolla, Vol.1, pp. 37-40.

5. Gutierrez, J.A. and Chopra, A.K. (1978). "A

Substructure Method for Earthquake Analysis of

Structures Including Structure-Soil Interaction." Earthquake Engineering and Structural Dynamics. Vol. 6, pp. 51-69.

6. Kamat, M.P. (1988). "Active Control of Structures in Nonlinear Response." Journal of Aerospace Engineering. Vol. 1, pp. 52-62.

7. Masri, S. F., Bekey, G. A., and Caughey, T. K.

(1981). "Optimum Pulse Control of Flexible." Journal of Applied Mechanics. Vol. 48, pp. 619-

626.

8. Masri, S.F., Bekey, G.A., and Caughey, T.K.

(1982) . "On-line Control of Nonlinear Flexible

Structures." Journal of Applied Mechanics. Vol. 49, pp. 877-884

9. Nagurka. M.L. and Yen V. (1990). "FourierBased Optimal Control of Nonlinear Dynamic

Systems." Journal of Dynamic Systems. Measurement, and Control. Vol. 112, pp. 17-26.

10. Reinhorn, A.M., Manolis, G.D., and Wen,

C.Y., (1985). "An On-line Control Algorithm

for Inelastic Structures." Second International

Symposium on Control . University of Waterloo,

Ontario, Canada, pp. 564-579.

11. Reinhorn, A.M., Manolis, G.D., and Wen,

C.Y., (1987). "Active Control of Inelastic

Structures." Journal of Engineering Mechanics. Vol. 113, pp. 315-333.

12. Suhardjo, J., Spencer, B.F., and Sain, M.K.,

(1990). "Nonlinear Optimal Control of a Duffing System." submitted to International

Journal of Nonlinear Mechanics.

185
13. Tripathi. S.S. and Narendra, K.S. (1970).

"Optimization Using Conjugate Gradient

Methods." IEEE Trans. Automatic Control.

IEEE, New York, New York, pp. 268-270.

14. Yang, J.N. and Akbarpour, A. (1987). "New

Optimal Control Algorithms for Structural

Control." Journal of Engineering Mechanics . Vol. 113, pp. 1369-1386.

15. Yang, J.N., Long, F.X., and Wong, D. (1988).

"Optimal Control of Nonlinear Structures." Journal of Applied Mechanics. Vol. 55, pp. 931-

938.

186
Current Research Efforts in Japan for Passive and Active

Control of Highway Bridges Against Earthquake

Kazuhiko KAWASIIIMA". Klnjl IIASEGAWA"". Shigekl UNJOH'', Illroyukl NAGASIIIMA''. and Illdeyukl SHINIZU""

ABSTRACT

This paper presents current research

activities on passive, active and hybrid control

for application to bridge structures conducted

at the Public Works Research Institute.

KEY WORDS : Highway Bridges, Menshln Systems,

Active Control, Hybrid Control

1. nrTRODUCTION

Menshln systems (Base-isolation), which is being introduced for application to highway

bridges in Japan" ^, is to reduce earthquake

response of highway bridges by "Passive Means". On the other hand, active control is to control

it by "Active Means." Since external energy

supply required to reduce the bridge response

to satisfactory level is significant during a significant earthquake In the active control, It

is still in a laboratory. Therefore, "Hybrid"

control, which is a combination of advantages

of passive and active control, is promising for

reducing the earthquake response of structures

with small amount of external energy. The

development of hybrid control for improving

the seismic safety of highway bridges, which can

not be achieved by extending current seismic

design method, is expected".

A series of researches, and developments on

passive, active and hybrid control for highway

bridges have been made at the Public Works

Research institute since 1987. This paper

briefly presents current research activities on

passive, active and hybrid control for

application to bridge structures conducted at

the Public Works Research Institute.

2. DEVELOPMENT OF MENSIUN SYSTEMS

2-1 Joint Research Prograa on Menshln Systems

for nighway Bridges

For studying the application of the menshln

systems (passive control) to the seismic design

of highway bridges, a committee chaired by

Professor Tsuneo Katayama, University of Tokyo,

was formulated through 1986 to 1989 at the

Technology Research Center for National Land

Development. Three programs were studied In the committee, i.e., 1) a survey of menshln

devices which can be used for highway bridges,

2) a study on the key points of the design of

menshln highway bridges, and 3) designs of

menshln highway bridges. As the final

accomplishment of the study, "Guidelines for

Design of Menshln Bridges (draft)" was published

In 1989^.

Based on the study above, the three-year

Joint research program on the menshln systems

for highway bridges Is now underway between

Public Works Research Institute and twenty

eight private firms since the year of 1989. The

goal of the program is to develop the menshln

design method and the new menshln devices for

highway bridges in order to improve the seismic

performance of new and existing bridges with

less cost by the menshln systems. FIg.l shows

the research Items of the program, which are outlined in the follows"".

2-2 Development of New Menshln DcYlces A lot of base Isolation devices which are

similar to the menshln device have been ever developed for building structures. However, the

menshln devices for highway bridges have to be more compact and more weather-proof than the

devices for building structures since the

menshln devices would be Installed at narrow

and exposed crests of bridge columns. The new

menshln devices should be developed

exclusively for menshln highway bridges to be

effectively constructed In Japan.

The following ten new devices in the six

types are now being developed under the

research program.

1) High Damping Rubber Bearing 4 devices

2) Sliding Friction Damper 2 devices

3) Steel Damper 1 device

4) Roller Menshln Bearing 1 device

5) Link Bearing 1 device

6) Viscous Damper 1 device

All developed menshln devices but the link

bearing were tested by the loading experiment

system of PWRI under the same loading

conditions to verify their performance. In the

tests, the cyclic lateral force was given to the

devices by the actuator under the constant

axial force, which is equivalent to the dead

load of the superstructure, as shown in Photo

1. Since the performance of some devices could

be affected by the temperature, the tests were conducted in the room shown in Photo 2, where

the temperature was kept at 20 °C.

1) Head of Earthquake Engineering Division,

Earthquake Disaster Prevention Department,

Public Works Research Institute, Ministry of

Construction, Tsukuba Science City, 305 Japan

2) Senior Research Engineer, ditto

3) Research Engineer, ditto

4) Assistant Research Engineer, ditto

187
Flg.2 shows the lateral force vs. displacement hysteresis loops of a high damping

rubber bearing and a sliding friction damper. It

Is said that the damping effect of every tested

devices Is satisfactory for highway bridges and

the dynamic character Is also stable with the

smooth hysteresis loop.

The standard of the test procedure for

menshin devices are being specified under the

research program In order to certify the other

devices which would be developed for highway

bridges In future. There are two main Items in the procedure, one of which Is related to the

performance as a support such as the check of

the vertical stiffness, and the other Is related

to the dynamic horizontal performance as a menshin device such as the check of the

stability under the repeated cyclic lateral

loading, the check of the dependency on the

loading velocity or frequency, the check of the

dependency on the axial force, the check of the

dependency on the lateral loading direction

and the check of the dependency on the

temperature. The test condition for each Item and the evaluation of the test result have been

formulated. The size and shape of the

specimens, the loading test apparatus and the

loading process will be standardized.

2-3 Development of Expansion Joints and

Restratncrs for Menshin Bridges

The knock-off mechanism and the finger

expansion Joint which Is distinguished from the

regular finger Joints by the transverse

movement, are being developed under the

research program. The restralner which consists

of the steel bar Installed In the crest of the

substructure and the steel casing with the high

damping rubber Inside Installed under the

superstructure. Is also being developed.

The knock-off mechanism Is Illustrated In Flg.3°'. If the excessive displacement of the

girder toward the abutment were brought by

earthquakes, the apron of the slab would

collide with the top portion of the abutment.

Since the portion Is not actually fixed to the

abutment Itself, It would be squeezed Into the

backfill behind the abutment to ease the Impact

Induced by the collision.

The mechanism was tested as shown In Fig. 4, In which the top portion was pressed back Into the backfill by the actuator. In order to observe the situation of the squeeze and the

damage of the backfill, and measure the

resistant force of the backfill®'. Photos 3 and 4 show the specimen and the situation after the

test, respectively. Flg.5 shows the lateral

force vs. displacement curve of the mechanism.

It Is seen that the Impact force could be

mitigated and the extensive displacement of the

girder would be controlled by the mechanism.

2-4 Development of Menshin Design Method

Flg.6 Illustrates the flow of the menshin

design method. In which the two levels of the

design earthquake loading are considered as limited states. The lower level. Level 1, Is equivalent to the design seismic force of the

seismic coefficient method regulated In the

Design Specifications of Highway Bridges - Part V Seismic Design. This represents the seismic

force for moderate earthquakes which are supposed to take place with higher probability.

The higher level. Level 2. is equivalent to the

design seismic force of the check of the bearing

capacity of the reinforced concrete columns for

the lateral force regulated In the

specification, for larger earthquakes, such as Kanto earthquake In 1923, which are supposed to take place rarely^®. The design procedure will be formulated, being based on the flow.

2-5 Application of Menshin Design

Since expansion Joints make annoying noise

and are vulnerable to the traffic load,

continuous bridges tend to be constructed

recently with the demand of lowering the

number of Joints. Therefore, one target of the

menshin systems Is to construct continuous

bridges. The multi-span continuous bridge with

the girder length more than 1 km, which Is called In the research program as a super

multi-span continuous bridge. Is highlighted as a crucial research Item. It was concluded that

the super multi-span continuous bridge with 26 spans as shown In Flg.7 could be constructed

with the menshin system.

Existing simple girders would be connected one another In order to also decrease

expansion Joints between adjacent simple

girders. Since the Inertia force of the girder

transmitted to each substructure might be

varied by the connecting, the strengthening of some substructures would be required due to the lack of the bearing capacity. It was deduced

that the distribution of the Inertia force could

be Intentionally adjusted with proportion to the capacity of each substructure, by the

replacement of the existing ordinary fixed or movable bearings with the menshin devices. The

construction method and the structural detail

of the connecting and the replacement will be

examined under the research program.

2-6 Draft Manual for Menshin Design of Highway

Bridges

The program will be accomplished In March,

1992. The draft manual of the menshin design

method, the contents of which are shown In the

follows. Is to be composed as the fruits of the

research program.

Chapter 1 INTRODUCTION

1.1 Purpose of Manual

188
1.2 Application of Manual

1.3 Definitions of Terms

Chapter 2 BASIC CONCEPTS

2.1 Purpose of Mcnshln Design

2.2 Earthquake Levels and Limited

States

2.3 Seismic Design Force of Menshin

Design

2.4 Design Method and Criterion of

Safety Check

2.5 Basic Flow Charts of Menshin

Design

2.6 Application of Menshin Design

Chapter 3 DESIGN AND CHECK OF SUPERSTRUCTURE

Chapter 4 DESIGN AND CHECK OF SUBSTRUCTURE

4.1 General

4.2 Design Acceleration for

Substructure

Chapter 5 DESIGN METHOD OF MENSHIN DEVICES

5.1 Basic Concept of Design

5.2 Design Acceleration for Menshin

Devices

5.3 Rubber Bearing

5.4 Lead Rubber Bearing

5.5 High Damping Rubber Bearing

5.6 Steel Damper

5.7 Friction Damper

5.8 Viscous Damper

Chapter 6 CHECK OF MENSHIN BRIDGES BY DYNAMIC

ANALYSIS

6.1 General

6.2 Input Motion of Dynamic Analysis

6.3 Dynamic Analysis Method

6.4 Safety Check by Dynamic Analysis

Chapter 7 STRUCTURAL DETAILS OF MENSHIN

BRIDGES

7.1 General

7.2 Restralners

7.3 Structural Details of Menshin

Devices

7.4 Structural Details of Expansion

Joints

Chapter 8 CERTIFICATION TESTS OF MENSHIN

DEVICES

8.1 General

8.2 Dynamic Loading Test

8.3 Static Loading Test

8.4 Ultimate State Test

Chapter 9 RETROFITTING BY MENSHIN DESIGN

9.1 General

9.2 Application of Menshin Design for

Retrofitting

9.3 Design Method of Menshin Devices

for Retrofitting

9.4 Construction of Menshin Devices

for Retrofitting

9.5 Structural Details of Retrofitting

Appendix I References for Design

Appendix II Design Examples

3. CONSTRUCTION OF MENSOIW niGHWAY BRIDGES

3-1 Pilot Construction of Menshin Illghway

Bridges

Five pilot menshin highway bridges as shown

In Table 1 are under construction with the

supervision of Ministry of Construction, In order to verify the performance of the menshin

system. One of them. Miyagawa Bridge in Shizuoka-ken, was completed and opened for the

public traffic In March 15, 1991 as the first

menshin highway bridge In Japan.

3-2 Miyagawa Bridge

- First Menshin Highway Bridge - DOutllne of Miyagawa Bridge

The Miyagawa bridge located at Haruno-machl, Shu-tl-gun. Shlzuoka-ken, Is a three-span continuous steel plate girder bridge

with the total length of 105.8 m and the width

of 10.5 m as shown In Flg.8, which spans over the

Ketagawa river as a bridge of the national

highway No. 362^', The rubber bearings with the

lead plug Inserted at the center shown In Photo

5, which were developed In New Zealand and are usually designated as lead rubber bearings were used as the menshin devices.

Although the lead rubber bearing is originally

movable to any lateral direction, the

restralners were Installed to prevent the

bearing from the transverse movement so that

the menshin system works only In longitudinal

direction.

2)MeDshln Design

The bridge was designed In accordance with

the Design Specifications of Highway Bridges.

Table 2 summarizes the natural period, the

design horizontal acceleration and the design

relative displacement of the bearing for both

the seismic coefficient method and the check of

the bearing capacity of the reinforced concrete

columns for the lateral force. The design

acceleration Is 0.2 g with the natural period of

0.76 sec In the seismic coefficient method. It Is 0.54 g with the natural period of 1.01 sec In the

check of the bearing capacity, when energy

dissipation effect of the base-Isolated devices

Is taken Into account. However, since the bridge

Is one of the first menshin bridges In Japan, the

damping effect of the devices was not taken

Into account In the check of the bearing

capacity. Therefore, the design acceleration of

0.7 g which was obtained by assuming that the

energy dissipation of the devices Is zero, was

used In the check of the bearing capacity.

189
since the tnenshln system triggers off the

larger displacement of the superstructure

relative to the substructure than the ordinary

bridges, the sufficient gap should be kept

between the abutment and the girder, and at the

expansion Joint. On the other hand, the larger

gap at the expansion Joint would bring the

vulnerability to the traffic load. Therefore,

the following concept was adopted that the gap

between the abutment and the girder should be

decided as the girder would not collide to the

abutment when It Is subject to earthquakes for

the check of the bearing capacity, and the gap

at the expansion joint should be made

equivalent to the relative displacement

developed at the expansion Joint by

earthquakes of the seismic coefficient method.

It was followed that the gaps between the

abutment and the girder were 150 mm and the

gaps at the expansion Joints were 50 mm.

3) Field Vibration Test

The free vibration tests were conducted

under the Joint research program between the

Earthquake Engineering Division of PWRI and the

Shizuoka-ken government prior the opening In order to verify the effect of the menshin

system. The three hydraulic static Jacks owned

by the Research Institute of Japan Highway

Public Corporation were placed between the

temporary steel frames assembled under the

girder and the A2 abutment as shown in Photo G. The force capacity of each Jack is 150 tf and

the maximum stroke Is 150 mm. The force created

by the Jack can be released more rapidly than

the regular static Jack to develop the free

vibration. The girder was forced by the Jack to move against the abutment in the longitudinal

direction to a certain displacement. The

maximum displacement was decided as 8 cm so

that the total force of the Jacks should not

exceed the capacity of the abutment.

The accelerations on the girder at each

substructure (8 points), the accelerations on

the crests of the substructures (8 points) and

the displacement of the girder relative to each

substructure (8 points) were recorded during

the free vibration tests. FIg.9 shows the relative displacement of the

girder on the A2 abutment when the force of the

Jack was released at 8 cm. It Is observed In the

figure that the free vibration of the girder was developed In 0.75 sec after the release and the

girder came nearly to a stop at 4 sec, remaining

the residual displacement of 5.4 cm by the

plastic resistant force of the lead plug.

Fig. 10(a) shows the general model for the

lateral load vs. displacement relation of the

lead rubber bearing. Since the load consists of

the linear restoring force of the rubber and

the plastic restoring force of the lead plug,

the relation can be expressed as a combination

of the linear spring model and the friction

model shown In Figs. 10(b) and (c), respectively.

The natural period and the damping effect was analyzed with a simple single degree of freedom

oscillation system as shown In FIg.ll, which Is a model for the girder and the lead rubber

bearings.

The spring constant and the friction force

necessary for the analysis were obtained from

the result of the loading tests, which had

measured the lateral load vs. displacement

relation of the lead rubber bearing prior to the construction, as 2,351 tf/m and 173.8 tf, respectively. The result of the analysis Is also

shown with a dotted line In the FIg.9.

The analysis evaluates the test very well In the first 0.75 sec after the release. It can be

derived that the dynamic performance of the

menshin bridge could be assessed with the

parameters obtained from the loading test of

the bearing Itself. The gradual decrease of the

relative displacement after 0.75 sec observed in the test, could have to do with the creep of the

lead plug.

The traffic vibration tests were also

conducted by running the trucks under the

Joint research program between the Structure

Division of PWRI and the Shizuoka-ken

government. No unusual vibration due to the

menshin system was monitored around the

bearing.

The long-term measurement of the residual

relative displacement after the release at 8 cm

will be done since the end of a series of tests. The displacement was recovered to 0.9 cm at about one month later. It Is said that the

residual displacement of the bearing cause no

practical problem.

At the Mlyagawa bridge, the strong motion

observation Is now being made by the

Shizuoka-ken government and Ground Vibration

Division of PWRI.

4. ACTIVE CONTROL FOR SEISMIC RESPONSE OF

HIGOWAY BRIDGES

4-1 Introduction

Applicability of an Active Mass Damper

(refers as AMD In the following) for controlling

earthquake response of highway bridges was

investigated by using a single-degree-of

freedom model and reported In the last Joint

meeting. The results showed that a great deal

of external energy is required In order to control earthquake response In appropriate

level by means of AMD. The mass and/or stroke

required are excessively large, and It seems not

practical to control seismic response of bridge

structures against a significant earthquake

Hence, It Is of great Importance to develop hybrid control by combination of

passive and active means. Development of hybrid control technology

190
was Initiated at the Public Works Research

Institute since 1989. Preliminary study on

control of seismic response of highway bridges

using multiple AMDs, which is basis for hybrid

control is presented here.

The cooperative research program Is being

executed as a U.S. -Japan cooperative research

program through the the Panel on Wind and

Seismic Effects, UJNR. The program Is being

conducted at the Public Works Research

Institute In Japan side, and at the National

Institute for Standards and Technology, and at National Center for Earthquake Engineering

Research'*" with supports of National Science

Foundation In U.S. side. Major subjects of the

program In Japan side Includes Investigations on © passive control method and ® active

control method for application to hybrid

control as well as (3) optimum hybrid control

which combines them appropriately.

4-2 Highway Bridge Analyzed and Anaiytlcal

Conditions

For studying where Is the best point for

control and effect of multiple control by means

of AMD, the transverse response of three-span

continuous girder bridge with length of 150m as shown In Flg.l2 is analyzed. The bridge Is modeled as a two-dimensional spring-mass model.

Fundamental natural frequency Is 1.67IIz, and

this Is a first flexural mode of the deck. The

first mode Is most predominant In the

transverse response of the bridge.

For controlling the deck response by means

of one Ai'ID, the weight of a driving mass Is assumed as 180tf, which corresponds to about

1/10 of the deck weight (1.950tf). Control by

means of two AMDs was also studied. In which two AMDs were assumed to be Installed on the girder

at the Pi and Pa piers, respectively. Three

cases are analyzed, that Is, their weights are assumed to be 180tf each. 90tf each, and 120tf

and 60tf (180tf In total). The spring

coefficients, which supports driving masses, are assumed so that fundamental natural frequency

of the mass of AMD be 1/4 of fundamental

natural frequency of the bridge analyzed.

Structural damping ratio of the bridge Is assumed as 5%. A standard ground acceleration,

which Is specified In "Part V Seismic

Design"^ of the "Design Specifications for

Highway Bridges", is used as an Input motion.

Control algorithm of AMD follows the

classical optimum feedback control method. The

control force driving the mass of AMD is assumed as a linear combination of the values

proportional to displacement and velocity of

the structure and AMD. Proportional

coefficients are called as control gain, and the

optimum gain obtained by solving Rlccattl's

Algebraic Equation"" '°' is used.

4-3 Effect of Controlling Point by AMD

In order to Investigate an optimum control

point, the earthquake response analyses are conducted. AMD Is assumed to be placed at the some points on the girder. Fig. 13 shows peak

response displacement and peak response

acceleration of the girder. It Is more effective

In reducing seismic response of the girder when

the AMD Is Installed near the center of the

girder. Comparing with uncontrolled one. peak

displacement of the girder decreases to 1/4 and

peak acceleration decreases to about half. On

the other hand, effectiveness Is smaller when

AMD Is placed near the abutments. For example,

when AMD Is placed on Abutment Aa. although the

response displacement becomes smaller by 15%

than uncontrolled one, displacement response near the ends of girder becomes larger by about

13 times the uncontrolled displacement.

Therefore. It Is more effective to Install

AMD at the point where the mode value of

predominant mode shape Is largest.

4-4 Effect of Multiple Control

In order to Investigate effectiveness of

multiple control, the earthquake response Is analyzed, In which AMDs are assumed to be

Installed on the Pi and Pa piers. Flg.l4 shows

peak displacement and peak acceleration of the

girder controlled at the two points. Results

controlled by only one AMD Is also shown In Flg.l4 for comparison. Two-point control makes

possible to decrease the seismic response of

girder more significantly than the control by

only one AMD. Effectiveness of control of the

response and control force required by AMDs

with the weight of masses of 180tf Is almost same

as those of 90tf both. Effect of the weight of

driving masses appears only In stroke lengths

of driving masses. The strokes of driving mass when the weight of driving masses is 90tf both

is twice those when the weight is 180tf. Therefore, the weight of driving masses affects

the stroke lengths of driving masses, but does

not affect the effectiveness of control. The

similar results can be obtained when the weight

of driving masses are GOtf and 120tf each.

4-5 Concluding Remarks up to the Present

1) It Is most effective that the AMD Is Installed

at the point where the mode shape value of a predominant mode Is largest for controlling the

seismic response of continuous girder bridges.

2) Two-point control using AMDs is more

effective than one-point control In reducing

the displacement response and acceleration

response of the girder. Since multiple control

makes It possible to control the deck response

against any mode shapes. It Is required to further Investigate the effect of the

multiple-point control in detail.

191
5. DEVELOPMENT OF A VARIABLE DAMPER

5.1 Basic Concept of A Variable Daapor for

nighway Bridges

The variable damper Is basically a viscous

damper and the viscous damping force is variable depending on the response of

structures (highway bridges) as shown In Fig. 15. For Instance, the damping coefficient of the

damper is larger during the vibration within a small amplitude and the damper has the same

function as a fixed bearing support against

braiting load of vehicles. It Is. however,

movable against the load with low velocity,

such as elongation of girder by temperature

change. Once an earthquake occurs and

amplitude of vibration of girder becomes larger

up to a certain level, the damping coefficient

of damper decreases so that energy dissipation

be optimum and inertia force to the

substructure be adjusted appropriately.

Furthermore, when the vibrational amplitude

becomes excessive, the damping coefficient

increases gradually in order to suppress the

amplitude and so that the damper have a function as a stopper. Therefore, the variable

damper has the advantages of an usual viscous

damper-stopper, a passive energy disslpator and

a stopper with a shock absorber. The required

external energy by the variable damper for

altering the damping coefficient is greatly

smaller than usual active control devices

because the width of the orifice of the damper

can be changeable with a small amount of

energy'.

5.2 Earthquake Response Analysis of MultlDegrec-of-FreedoB System with Variable

Dampers

The equations of motion for a linear

multi-degree-of-freedom model with the variable

dampers may be written as

M X MC + Cv) X K x = - M xo (1)

in which M, C and K represent mass, damping and

stiffness matrices of the structure,

respectively, x and Xo denote displacement

vector and ground displacement, respectively.

Cv denotes a damping matrix of the variable

dampers and Is given by a function of relative

displacement and relative velocity between the

nodes where the variable dampers are installed.

Since Cv Is time-varying, Eq.(l) has to be solved

by a direct Integration method. According to the Newmark /9 method, velocity and

displacement are assumed as

Xi = Xi-i+(l-a)AtXi-i + a AtXi (2)

Xi = Xi-i + A t - xi-i + (1/2 - ^ )( A t)^ - xt-i ^(At)^x.-i (3)

In which a and $ are parameters of the

Newmark $ method and if a constant

acceleration method Is assumed a and p have

to be taken 0.5 and 0.25, respectively.

Using the above analytical method, a computer program, named as "VDAM", which can analyse earthquake response of multl-degreeof-freedom system with the variable dampers was

developed.

5-3 nighway Bridge Analyzed and Analytical

Conditions

In order to investigate effectiveness of the

variable damper, a simple span girder bridge, as shown in FIg.lG, with a span length of 30m is analysed In the longitudinal direction. The

model bridge has elastic Isolator and the

variable dampers are assumed to be installed

between the superstructure and the top of

substructure. The spring constant of elastic

isolator Is defined so that a fundamental

natural period of the bridge be 1 second.

Fourteen cases In total are analysed as shown in Figs. 17 and 18. In Case 1 the variable

damper is not Installed, in Cases 2 and 3 the

damping coefficient of the damper is In dependent of displacement and/or velocity. The

damping coefficients of Cases 2 and 3 are given

so that damping ratio of critical be 5* (Case 2) and 100% (Case 3). in Cases 4 and 5, the damping

coefficient is dependent on relative

displacement as shown In Fig. 17. Peak relative

displacement dj is defined as a peak relative

displacement computed in Case 2 and It was assumed as 19.4cm. In Cases 6 and 7, the damping

coefficient Is dependent on relative velocity as shown In Flg.l7, Peak velocity v.j is defined as a peak value (113 cm/s) computed In Case 2. Case 8

is a combination of Cases 4 and 6 and the

damping coefficient of the damper is dependent

on relative displacement and relative velocity.

A ground motion acceleration which was

modified so that the response spectra matches

with the target spectra Is used as an input

ground acceleration. The acceleration response

spectrum, which Is specified In "Part V Seismic

Design" of the "Design Specifications for

Highway Bridges", for the check of bearing

capacity of reinforced concrete piers, Is assumed as the target spectra.

Damping ratio of critical of a model bridge

Is assumed 2* as a modal damping.

5-4 Errectiveness of Variable Dampers on

Earthquake Response of Highway Bridges

Table 3 shows the peak response of girder

and the variable damper, where the peak values

of displacement, velocity and acceleration of

the girder and damping force, relative

displacement and relative velocity of the

damper are presented. Figs. 19 and 20 compare

time histories of displacement and acceleration

192
of the girder and damping force of the variable

damper between Case 1 (without the variable

damper). Case 5 (displacement dependent) and

Case 7 (velocity dependent).

According to these results, the following

remarks may be pointed out:

1) If the damping coefficient Is dependent on

relative displacement of the damper. It Is effective to set larger damping coefficient

around zero displacement (Case 5). The peak

damping force In Case 5 becomes 58.2 x 2tf which

corresponds to 48.2* of the weight of the deck

(241. 5tf). The damping force of 96.4tf makes

reduction of acceleration of girder to 566gal/1300gal = 43.5%, and relative

displacement of the girder to 13.25cm/33.04cra = 40.1% comparing with Case 1 In which the

variable damper Is not Installed.

2) The damping force required to control the

response up to the same level Is smaller In displacement dependent dampers (cases 4 and 5) than In velocity dependent dampers (cases 6 and

7). This Is because pulse-type large damping

forces are required In the velocity dependent

dampers. Large damping coefficient at the time

when large velocity is developed causes large

damping force. On the other hand, since the

damping force Is dispersed as time large

damping forces are not required In the

displacement dependent dampers.

3) The time history displacement shows that the

variable dampers operate satisfactorily as a stopper which also has a function as a shock

absorber without an Impact response.

5-5 DevelopBiGnt of A Pilot Model of the Variable

Damper

Photos 7-9 show a pilot model of the

variable damper developed. The model Is designed so that the damping force be variable

within 20kgf to 200kgf. maximum relative

displacement of damper piston be + 5cm, and the

weight be about 40kgf. The variable damper

consists of an usual viscous damper and a control system. Damping characteristics of the

variable damper Is now being studied. Model

experiments are planed.

5-6 Concluding Remarks up to the Present

1) In the displacement dependent dampers, It Is effective to set larger damping coefficient

around zero displacement.

2) The damping force required to control the

response up to the same level Is smaller In the

displacement dependent dampers than in the

velocity dependent dampers.

3) The time history response shows that the

variable dampers operate satisfactorily as a stopper which also has a function as a shock

absorber without an Impact response.

4) The pilot model of the variable damper was

developed. Experimental study on the

characteristics and control method of the

variable damper Is being studied.

5) The variable damper seems to be near a practical use.

6. CONCLUSIONS

This paper briefly presents current

research and development activities on passive,

active and hybrid control for application to bridges structures conducted at Public Works

Research Institute. It Is thought that these

technologies may become very Important In future for constructing super-multiple

continuous girder bridges, long-span bridges,

bridges with high-rise piers, and bridges In urban area with strict land use restrictions.

RKFERENCES

1) Kawashlma, K. : Development and Future Scope

of Seismic Isolation of Structures, - A Review -, Proc. of Japan Society of Civil

Engineers, Structural Eng. /Earthquake Eng.,

Vol.398/ I -10. 1988

2) Buckle, I. : Development and Application of

Base Isolation and Passive Energy

Dissipation - A World Review -, Proc. of A Seminar and Workshop on Base Isolation and

Passive Energy Dissipation, pp.l53~174, San

Francisco, U.S.A.. March 198G.

3) Guidelines for Design of Base-Isolated

Highway Bridges, Final Report of A Committee

for Study of Application of Base Isolation to Seismic Design of Highway Bridges (Chairman : Professor T. Katayama), Technology Research

for National Land Development, March 1989

4) Public Works Research Institute : Report of

Joint Research between PWRI and 28 Private

Firms on Development of Menshln Systems for

Highway Bridges, Technical Report of

Cooperative Research, No. 44 and No.60, March

1990 and March 1991, Public Works Research

Institute

5) Chapman. H. E. and KIrkcaldle, D. K. : Seismic

Design of Base Isolated Bridges

Incorporating Mechanical Energy Dlsslpators,

Transit New Zealand's 1990 Bridge Design and

Research Seminar, Chrlstchurch, New Zealand.

November 1990

6) Goto, Y., KIkuchI, Y. and M. Kurolwa : Investigation on a Knock-off Device as an Expansion Joint for Menshln Bridges,

Proceedings of the 46th Annual Meeting of

Japan Society of Civil Engineers, October

1991

7) Japan Road Association : Part V Seismic

Design of the Design Specifications for

Highway Bridges, February 1990

8) Iwasakl. T. , Kawashlma. K. and Hasegawa. K. : New Seismic Design Specifications for Road

Bridge In Japan, 22nd Joint Meeting • -••o

193
U.S. -Japan Panel on Wind and Seismic Effects.

U.J.N.R.. NIST. Galthersburg. May 1990

9) Matsuo. Y., 0-lshl. A.. Hara, K. and Yamashlta,

M. : Design and Construction of Mlyagawa

Bridge, - First Menshln Bridge In Japan -, Bridge and Foundation, Vol.91. No,2. ppl5-21,

February 1991

10) Taylor, R. G.. and Roblson. W. H. : Highstrain Test on Lead Rubber Bearings for

Earthquake Loadings, Bull, of the New

Zealand National Society for Earthquake

Engineer, Vol.17, No.2. pp90-I05, June 1984

11) Bulling, I. J. and Klrkcaldle. D. K. : Base

Isolation of Bridges In New Zealand, Proc. of

New Zealand-Japan Workshop on Base

Isolation of Highway Bridges, Wellington,

New Zealand, November 1987

12) Kawashlma, K., Hasegawa, K. and Yoshlda, T. : Active Control of Seismic Response of

Structure by Means of Mass Damper, Civil

Engineering Journal. Vol. 31-5, May 1989

13) Taguchi, J.. Iwasakl. T.. Adachl. Y., Sasaki, Y. and Kawashlma, K. : U.S. -Japan Cooperative

Research Program on Hybrid Control of

Seismic Response of Bridge Structures, 22nd

Joint Meeting, U.S. -Japan Panel on Wind and

Seismic Effects, U.J.N.R., NIST. Galthersburg.

May 1990

14) Feng. Q. and Shinozuka, M. : Use of A

Variable Damper for Hybrid Control of Bridge

Response Under Earthquake, U.S. National

Workshop on Structural Control Research,

University of Southern California. CA.,

U.S.A., October 1990

15) Ito, M. : Automatic Control, Shokodo

Publishing Co.. February 1983 lG)Kalman. R. E. : Contributions to the Theory

of Optimal Control. Proc. 1959 Mexico City

Conference on Differential Equations. Mexico

City, 1960, PP102-199

194
Table 1 Pilot Program for Construction of Menshln Highway Bridges

Owner Name of Bridge Type of Superstructure Total Length

Hokkaido developing Bureau On-netoh Bridge Steel Girder 456 m

Tohoku Regional Construction

Bureau, MOC

Nagakigawa Bridge Steel Girder 97 m

Iwate-ken Maruki Bridge Prestressed Concrete 92 m

Tochigi-ken Daiichi Karasuyama Bridge Prestressed Concrete 250 m

Shizuoka-ken Miyagawa Bridge Steel Girder 110m

Table 2 Natural Period, Design Acceleration

and Design Relative Displacement

Design Method Item Al Pi P2 A2

Seismic

Coefficient

Method

Natural Period (sec) 0.764

Design Acceleration (g) 0.20 Design Relative

Displacement (mni) 28.2 21.8 20.2 28.0

Check of

Bearing

Capacity

Natural Period (sec) 1,01 Design Acceleration (g) 0.54 Design Relative

Displacement (mm)

133 109 105 130

195
Table 3 Peak Response of Deck and Variable Damper

(a) Full Control

Analytical Cases

Deck Variable Damper

Displacement

( cm^

Velocity

(cm/sec)

Aceleration

( cm/sec^^

Damping

Force

(tf)

Relative Displacement ( cm \Relative Velocity

Total Energy

Absorption

(tf-m)

1 No Control 33.04 189.0 1300

2 Constant Damping

(h, = 0.05)

25.63 148.7 1012 8.8 19.41 113.1 65.41

3

Constant Damping

(h2 = 1.0)

7.97 45.1 481 47.1 5.73 30.4 81.51

4

Displacement

Dependent

Damping

23.33 140.7 1043 39.2 17.32 142.2 70.70

5 Displacement

Dependent

Damping

13.25 84.8 566 58.8 10.18 123.6 93.62

6

Velocity

Dependent

Damping

21.10 131.3 1117 140.3 16.05 169.4 78.41

7

Velocity

Dependent

Damping

20.35 126.7 1029 109.8 15.44 147.4 76.38

8

Displacement

and Velocity

Dependent

Damping

20.96 122.7 1136 129.3 15.72 168.1 80.52

(b) Partial Control

Analytical Cases

Deck Variable Damper

Displacement

(cm)

Velocity

(cm/sec)

Aceleration (cm/sec^)Damping

Force

(tf)

Relative Displacement (cm)Relative Velocity

(cm/sec)

Total Energy

Absorption

(tf-m)

5 DisplacementDependent

Damping

13.25 84.8 566 58.8 10.18 123.6 93.62

9

Control When

v>0 19.10 113.5 769 71.2 14.87 181.1 94.16

10 Control When

v<0 18.87 112.7 757 70.0 14.57 176.8 92.81

11 Control When

dv>0 13.60 87.2 575 60.0 10.54 136.5 92.83

12 Control When

d-v<0 14.14 86.9 591 60.1 11.41 147.1 94.28

13 Control When

d>0 13.69 84.8 592 58.5 11.48 140.3 92.80

14 Control When

d<0 14.02 89.4 585 61.7 10.78 141.9 93.59

196
Development of Menshin Devices

for High way Bi idges

Development of Expansion

Joints and Restrainers for Menshin Bridges

Development of Menshin Design

Method

"— Application of Menshin Design

- Higli Damping Rubber Bearing

Sliding Friction Dumper

Steel Damper - Roller Type Menshin Bearing - Link Bearing - Viscous Damper - Certification Test of Menshin

Devices

Expansion Joints for Menshin Bridges

Restrainers for Menshin Bridges

Basic Concept of Menshin Design

Dynamic Analysis Method of Menshin

Bridges

Design Method of Menshin Devices

Simplified Design Method of Menshin

Bridges

Design Method of Restrainers and

Expansion Joints for Menshin Bridges

Evaluation of Model Tests by Dynamic

Analysis

Application for Prestressed Concrete

Bridges

Application for Steel Bridges

Application for Super Multi-span

Continuous Bridges

Application for Seismic Retrofitting

Flg.l Research Items of Joint Research Program

24

168

0

8

16 24

-20 -16 -12 -8 -4 0 4 8 12 16 20

LATERAL DISPLACEMENT (cm)

(a) High Damping Rubber Bearing

-10 -8 -6 -4 -2 0 2 4 6 8 10 LATERAL DISPUCEMENT Icm)

(b) Sliding Friction Damper

Fig. 2 Lateral Load vs. Displacement Hysteresis Loop

197
APRON OF SLAB

KNOCK-OFF

GIRDER^>

ABUTMENT

Fig. 3 Knock-off Mechanism

ASPHALT PAVEMENT

RESISTING WALL

SHAKING

TABLE

O

—7 / KNOCK-OFF

/ / \ LOAD CELL

E

in

ro

BACKFILL / 6.6m

1 ^

rKNOCK-OFF

Is

3.0 m 0.25 m

7777 ELEVATION

H-SHAPED

STEEL BAR

DETAIL OF KNOCK -OFF

e-EO\o

PLAN oID

250mm

ANCHOR BAR

lOOmm

E

E

500 mm

Et250mm

1&I

r-U

lOOmm

500mm

Flg.4 Experiment of Knock-off Mechanism"'

30

9 20

WIT

WIT

H ANCHOR BAR

HOUT ANCHOR BAR

fx.-.

5 to 15 20

LATERAL DISPLACEMENT (cm)

FIff.5 Lateral Load vs. Displacement Relation

of Knock-off Mechanism®'

198
[Item] [Object] [Analysis Method ] [Design Force Level]

Assumption of

Initial Dimension

Substructure Static Analysis Level 1

Design

Menshin Device

Static Analysis

Level 2 Substructure Level 1

Check

Menshin Device

Substrcture

Static Analysis Level 2

Dynamic Analysis

Level 1 Level 2

Structural Detail Expansion Joint (Level 2)

Restrainer (Level 3)

Flff.6 Basic Flow Chart of Menshin Design

Super Multi-span Continuous Bridge

199
UJ

-J O

|jj<t o

a: -J t

o (

STOP OF MOTION LEGEND - EXPERIMENT - ANALYTICAL 12 3 4 5

TIME (second)

Flgr.9 Relative Displacement after Release

F

0

'

u

(a) Lateral Load vs Lateral (b) Elastic Spring

Displacement Relation of LRB

Flg.lO Load vs. Displacement Relation of LRB

(c) Friction Force

k

^77777777^772^7^

m

X

Fig. 11 Single Degree of Freedom

150 m 50m

1 50m 50m

At7T

Pi'

(a) Highway Bridge Analyzed

ic ^20

al—2 3 4 5 6 7 8 9 10 •—•

W

(b) Analytical Idealization

(c) Fundamental Natural Mode

of Vibration (f.=1.67Hzj

Fig. 12 Highway Bridge Analyzed

and Analytical Idealization

200
123456789 10 123456789 10

' ' I——I 1—I 1 U—l 1 I I 123456789 to 123456789 10 NODAL POINT NUMBER NODAL POINT NUMBER

(a) Peak Response Displacement (b) Peak Response Acceleration

FIff.l3 Effect of Controlling Point by AMD

AMD1 AMD 2 AMD! AM02

I—I—I—I i_i I I I I I I I I I I 123456789 10 123456789 10 NODAL POINT NUMBER NODAL POINT NUMBER

(a) Peak Response Displacement (b) Peak Response Acceleration

Fig. 14 Effect of Multiple Control by AMDs

201
DECK RESPONSE RELATIVE TO PIER CREST

Fig. 15 Basic Concept of Variable Dampers

Fig. 16 Highway Bridge Analyzed and Analytical Idealization

202
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Fig.20 Computed Response of Variable Dampers

206
Photo 1 Loading Experiment System for Menshln Devices
Photo 4 Situation after Test'

Photo 5 Lead Rubber Bearing

I Photo 6 Hydraulic Static Jack

208
Photo 7 Bridge Model with Variable Damper

209

Status of U.S. Research on Structural Control Systems

by

S.C. Liu*, HJ. Lagorio*, K.P. Chong*

ABSTRACT

This paper assesses the current status of control

technology and its potential application to

structural safety against seismic, wind and other

dynamics loads. An intensive effort underway to develop a national research program is described.

Research needs and important problems

remaining from both the technological and

practical standpoints are suggested.

1. INTRODUCTION

Bridges, buildings and other structures have

traditionally relied on their strength and ability to

dissipate energy through engineering design to

resist dynamic forces such as strong seismic

events, extreme winds, ocean waves, and heavy

traffic or machine loads. In recent years, however, the use of passive and/or active systems

to control and reduce structural and architectural

damage caused by earthquakes, winds, and other

environmental loads has become an area of

considerable research interest. New provisions

dealing with seismic isolators have been included

in the 1991 edition of the UBC which will enhance the practice. Historically there have

been a number of successful applications of

control concepts where the objectives, limits and

criteria for the controller are essentially

established by the specific design requirements.

However, the pace of practical application in a

global sense and technological advancements in the field of intelligent materials and control

systems for civil structural design and

construction, have been falling short of

expectations. Control technologies as required for such civil structural applications clearly need

further development and enhancement through

use of advanced sensors, computers, and actuators

to (a) detect changes in the system internal state and environment, (b) monitor quantitative models

of system performance by means of multiple sensor inputs, and (c) generate real time actions

to modify or alter system behavior, performance,

and operation due to earthquake, extreme winds,

and other natural or man-made hazard

uncertainties.

In connection with passive systems, much research

has been accomplished in specific areas related to base isolation techniques and a variety of

mechanical energy dissipators such as structural bracing systems, friction dampers, viscoelastic dampers, and other mechanical devices. In the area of active systems, active mass dampers,

variable stiffness devices and active bracing or tendon systems represent some of the devices

being developed, tested, or installed both in the

laboratory and, in certain cases, in prototype full- scale structures. Recently, the combined use of

active and passive systems, which are often

referred to as hybrid systems, have been suggested

for some specific structural applications.

2. DYNAMIC CONTROL AND

INTELLIGENCE

An active or hybrid controlled structure can be

viewed under the general context of "intelligent structures" or "intelligent materials". The

intelligence can be programmed or built into the

structure by material composition and processing,

real-time measurement and monitoring of dynamic

response, detection of microfracture, yielding, or other failure, or conditioning through

computerized servo system to adapt in a controlled manner to external forces or other

stimulus. In a general sense, a "controlled" structure usually contains, through ingenious

engineering design, its own sensors, data

acquisition and reduction, actuators,

computational/control, and feedback capabilities.

It appears that such technologies, when

sufficiently developed and broadly implemented,

possess an excellent potential to impact the safety

design and construction of new structures in the

future. Furthermore, the dynamic control and

intelligence technology also offers a viable,

alternative solution to the nation's enormous

* Engineering Directorate

National Science Foundation

Washington, D.C. 20550

211
problems in the rehabilitation of existing structures which are deficient in their seismic,

wind, or other resistant capacities.

It is generally agreed that dynamic control and

intelligence technology holds considerable

promises in the design applications of civil engineering structures, new and existing. By its interdisciplinary nature, such technology also has

the potential impact to the future development

of a diverse field of scientific or technical subjects,

such as material science, vibration, biotechnology,

neural network, and new devices.

To forge ahead in the development and practical

implementation of structural control technology,

a coordinated research program covering these

diverse but interconnected subjects is needed and

should be developed. This paper describes the

development process of such a program by the

National Science Foundation (NSF) and presents

the results of the current technology assessment and future research needs in this field of emerging

importance.

3. RESEARCH PROGRAM DEVELOPMENT

Over the last few years, intensive development

efforts have been made for the purpose of

charting a long-term national research agenda on

structural control. An important capstone of such an agenda is to foster, on a coordinated basis,

multidisciplinary research and development of

passive, active and hybrid control technology for application to structures and nonstructural

building systems under dynamic loads. Collaborative efforts among researchers in interdisciplinary areas involving academics,

industrial experts and professional practitioners are considered extremely essential.

The development effort, which is expected to lead

to an official announcement of research initiative by NSF, can be described by the following series of technical events that have taken place to date:

(1) In 1989, a U.S. Panel on Structural

Control Research was established under the

auspices of NSF. The Panel has the responsibility

of: (a) Facilitating the transmission of information

concerning state-of-the-art developments in the

field. (b) Identifying and prioritizing needed research

and development.

(c) Developing preliminary plans for analytical

and experimental advancement in the field. (d) Developing plans for the performance of full- scale testing and demonstration.

The Panel's organization consists of an Executive

Committee with Professor George W. Housner as

its chairman, and seven (7) Working Groups in: (1) Analytical Methods, (2) Experimental

Methods, (3) Building Applications, (4) NonBuilding Applications, (5) Interdisciplinary

Approaches, (6) International Coordination, and

(7) Information Dissemination. In the meantime,

a Japanese counterpart panel on Seismic Control

Research was also established by the Science

Council of Japan and Dr. T. Kobori is the

chairman of the Japanese Panel. Both the U.S. and Japan panels have been working closely in the

development of a joint US-Japan research agenda

in the areas of active and hybrid control.

(2) On July 23 to 26, 1990, an International

Workshop on Intelligent Structures was held in Taipei, Taiwan. The emphasis of the workshop

was on the development, application, and

modification of active/hybrid control theories, and

through innovative materials processing and

system concepts, to achieve designed intelligence

of civil engineering structures. Three major

theme areas for the technical program of the

workshop were established: (a) Sensing and

monitoring techniques, (b) Structural control, and

(c) Intelligent systems (Chong, Liu and Li, 1990).

(3) On October 15 and 16, 1990, a workshop

was held at the Electrical Power Research

Institute (EPRI) under the joint auspices of NSF

and EPRI. The objective of this workshop is to evaluate the status of and research needs in the area of Intelligent Control Systems (Shoureshi

and Wormley, 1990). The technical program was

structured on four technical areas: (a) General

theoretical foundations assessment, (b)

Knowledge-based systems, (c) Application related research, and (d) Software/hardware requirements.

(4) On October 25 and 26, 1990, the U.S. Panel on Structural Control Research organized

and held the "U.S. National Workshop on

212
Structural Control Research" at the University of

Southern California, Los Angeles. The workshop

identified four specific areas related to earthquake

safety for technical discussions. They are: (a) Analytical research, (b) Experimental research, (c) Building applications, and (d) Non-building

applications. Proceedings of the Workshop is in press and will be available by the Summer of 1991

(Housner and Masri, in press).

(5) On February 25 and 26, 1991, NSF held

another Workshop on "Sensors and Signal

Processing for Structural Control" in Washington,

D.C. This Workshop focused on assessing the current state of knowledge and technology

concerning sensors and measurement techniques,

signal processing, actuator mechanics as they

applied to structural control (Roberts and Li, in

press).

4. PROBLEMS AND RESEARCH NEEDS

From the intensive investigations conducted under

the special meetings and workshops as described

above, a clear picture of the current state of

structural control technology is now emerging.

While considerable progress is being made in research and development of passive and active structural control systems both in the U.S. and

abroad, a number of important technological as well as practical issues of cross-disciplinary nature need to be addressed in order to forge close ties between research and practice and to accelerate

the process of implementation. Such issues by

their disciplinary nature include:

• Determination of quantitative parameters of

the input forcing functions as well as structural response parameters that could be measured

and incorporated in the control algorithms;

• Development of high-tech sensors as well as

effective use of conventional sensing devices

currently available for sensing and measuring

the desired forcing and response parameters

which would minimize the time-delay and spill- over problems;

• Investigation of actuator mechanics which

allow the desired real-time actions to be

generated reliably within practical power

supply limitations for safety or for comfort

control according to the established

performance specifications through system

design and integration;

• Use of, or use through modifications of, existing control concepts, theories, and systems

developed or applied in other fields such as aerospace, navigation, mechanical or electric systems to civil engineering structures;

• Development of new materials and sensors which possess their own intelligence;

• Development of effective signal processing

techniques and algorithms for instantaneous,

feedback/feedforward, adaptive, or optimum

on-line control operation;

• Development of advanced computer

technology and systems incorporating artificial intelligence, fuzzy logic, and/or neural

networks for specific structural control

application;

• Reduction and control of chaotic vibrations in mechanical and transportation systems;

• Development of innovative structural,

architectural and nonstructural systems which

are amenable to the application of structural control;

• Modelling and robustness of structural control

under uncertainties; and

• Reconciliation of the conventional concepts of

structural safety and controlled safety.

It is with such a background that a coordinated

national research initiative in this emerging area of significant application potential is being

formulated. This initiative is designed to facilitate and accelerate research and development in structural control. Through this effort a solid technological base is expected to be established

which will make possible for broad, practical

implementation of structural control systems to safeguard structural and nonstructural systems

against unpredictable natural hazard events.

213
The fundamental objectives of the program

initiative under development are identified as:

• encourage innovations in passive, active, and

hybrid control systems.

• advance sensor and actuator technology and

signal processing techniques and their applications.

• develop intelligent robots, other devices and

computer systems for structural safety.

• develop innovative systems for energy

absorption, added damping, and variable

stiffness.

• study the robustness and system reliability of

control systems.

• develop guidelines, standards and practical

approaches and techniques for design,

fabrication and field installation.

• investigate emerging technology subjects in innovative structural systems, and smart

materials, sensors, and devices for detection,

sensing, monitoring, and diagnosis.

In line with the above objectives, the following

key research problems are identified:

• New materials.

• Performance evaluation, analytical and

experimental verification.

• Standardized test procedures and performance

criteria.

• Integrated structural and architectural systems

design techniques and standards.

• Control algorithm development, robustness,

and reliability, adaptivity, and optimization.

• Sensors, actuators and monitoring system

development and optimal deployment.

• Advanced computer technology incorporating

artificial intelligence, neural network, fuzzy

logic, etc. for on-line targeting, detection,

monitoring, diagnosis, and control.

• Intelligent robots and devices.

• Modelling, signal processing, and computer

simulations.

• Practical methods and techniques for design,

manufacturing, fabrication, and

implementation.

• Laboratory and field experiments of systems

and devices under actual or simulated dynamic

excitations.

• Design and demonstration projects of control

systems as applied to new or existing

structures.

• Innovations in base isolation, energy

dissipation devices, and other passive

techniques.

In some of the research areas identified above,

collaboration among researchers, practicing

engineers, and manufacturers is necessary. Results obtained from such a coordinated

research effort are expected to generate new

knowledge, techniques, and a better understanding

of the working principles as well as practical

limitations inherent in these control systems so

that conditions of applicability and design limits of each class of systems can be established and

evaluated. It should be emphasized again that the

ultimate goal of this multi-year research effort under planning is to bring about technological

advances so that control technology can be

realistically implemented in a timely fashion to provide safety resistance to protect structures such

as buildings, bridges, tunnels, and other critical lifeline systems against earthquakes, winds, and

other natural or manmade hazard environments.

5. SUMMARY AND CONCLUDING

REMARKS

Over the last two decades, considerable progress has been made to improve structural safety

through research in structural, earthquake, and

wind engineering. The possibility of applying

control systems to mitigate the structural damage

against earthquakes and winds opens up a wide

range of challenging research opportunities for multidisciplinary investigators in the field. Further, such challenges can be undertaken most

effectively by engineering professionals through

international research cooperation. In this spirit,

the mechanisms which might be provided by the

International Decade for Natural Disaster

Reduction (IDNDR) program can efficiently serve

as a launching pad for the research projects as

identified. The UJNR Panel on Seismic and

Wind Effects, and the US Panel on Structural

Control Research and its counterpart Japanese

panel are all expected to play a pivotal role in the

advancement of control technology and to lead us toward a safer and more economical future for

civil engineering construction.

214
6. REFERENCES

1. Chong, K.P., Liu, S.C., and Li, J.C., Editors,

Intelligent Structures (1990). Elsevier Applied

Science, London and New York, 460 pp.

2. Housner, G.W. and Masri, S., Editors,

Proceedings of U.S. National Workshop on

Structural Control Research (in Press).

3. Roberts, R., and Li, W.P., Editors, Proceedings

of Workshop on Sensors and Signal Processing

for Structural Control (in Press).

4. Shoureshi, R. and Wormley, D., Editors,

Proceedings of Workshop on Research Needs

in Intelligent Control Systems (1990).

NSF/EPRI.

215

Vibration of Large-Scale Shaking Table Foundation

BY

Chikahiro Minowa^, Nobuyuki Ogawa^, Tadashi MikoshlbaS,

Keilchi OhtanP, Norio OyagU

SUMMARY

The large scale shaking table of National

Research Institute for Earth Science and Disaster

Prevention, was improved in 1988. By this

iirprovement work, the shaking table performances

have been enhanced. Just after this mechanical

effort, the reinforcement work of the shaking

table foundation was conducted. During the reinforcements, the change of dynamic characteristics of large-scale shaking table foundation were measured at the various work stages to extract the dynamic Interaction effects. By the reinforcements, the foundation weight was

increased about 3000tons. The old one was of about 8700tons, and the new one has 11700tons approximately. This paper presents the mea- surement data and describes the vibration control system for shaking table foundation which was proposed during the planning of this reinforcements .

KEY WORD : Shaking Table Foundation. Vibration

Test, Soil -Structure Interaction, Vibration

Control

1. INTRODUCTION

Many studies to verify experimental ly the effects

of dynamic interaction between soil and foun- dation have been carried out in recent years. Most of the dynamic tests, however, have been

conducted using comparatively small scale

foundation models. It is preferable to use foundations as large as possible In order to minimize the effects of the local irregularities

of soil properties which Inherently exist in actual soil. Valuable experimental results for both dynamic

soil-structure interaction and dynamic characteristics of the surrounding soil have been

presented by making use of the large scale shaking

table foundation of National Research Institute

for Earth Science and Disaster Prevention. In 1988. the limit performance of this shaking table

was Improved. The maximum displacement has been

enlarged to 44cm p-p from old amplitude 6cm p-p, also the maximum velocity has been magnified to 75cm/sec. Reinforcement works of the foundation

began just after the improvement. The drives of

this shaking table have been possible during the reinforcement works. In course of the reinforcement works, the embedment conditions, the configuration and the weight of the foundation

have been changed. At various stages of the reinforcement works, the excitation tests of the foundation have been performed by driving the shaking table. The change of dynamic characteristics of the foundation and the surrounding

soi 1 at each stages of the work has been measured.

The one of objective of this study is to measure

the change of dynamic characteristics of the foundation and to extract the Interaction effects

from the observed results.

In the planning of the reinforcements, vibration

control methods for the shaking table foundation,

which already exists, were discussed. The methods

are described in this paper.

2. OUTLINES OF SHAKING TABLE FOUNDATION. EXPERIMENT AND MEASUREMENT

The foundation is made of reinforced concrete

with base area of 39m^5m and embedded in the

soil by 8.2m as shown in Fig. 1. Soil profiles

are shown in Table 1. In the reinforcement works, concrete mass of about l,200m3 was cast between

the buttresses of both sides of the foundation

as Indicated by black pointed area of Fig. 1. The weight of foundation itself increased from

8700tons to lUOOtons after the reinforcement

works.

Figs. 2(a) to (f) show the six states of the foundation when the excitation tests were performed. The first stage corresponds to the state before the reinforcement works; the second to the state when the surrounding soil was removed;

the third to the state when the concrete was

cast; the fourth and the fifth to the states when

the sides of the foundation were half and fully

backfilled, respectively; and the sixth to the state when sheet piles were take out. The

backfill soil is loose and the shear wave velocity

of the soil Is estimated to be about 50m/sec. The dynamic response of the foundation during

the excitation tests have been measured by

displacement meters, accelerometers and earth- pressure gages. Fig. 3 shows the location of accelerometers and the displacement meters( TIsec, ^ " 70%) set on the foundation and the surrounding soil.

1 Director. Disaster Prevention Division,

National Research Institute for Earth Science

and Disaster Prevention, Science and Technology

Agency

2 Head. Earthquake Engineering Laboratory, ditto

3 Head, Earthquake and Volcanic Disaster Pre- vention Section, ditto

4 Senior Researcher, ditto

5 Research Member. Earthquake Engineering Laboratory, ditto

217
3. DYNAMIC CHARACTERISTICS OF SHAKING TABLE FOUNDATION

By comparing the experimental results of six stages of the foundation shown in Fig. 2, the effects of various factors on the dynamic

characteristics of foundation can be extracted.

It should be noted that all of the comparison

are made for the results observed at the measurement points of foundation.

3.1 Effect of Constraint of Soil Figs. 4(a) and (b) show the cotiparison of the horizontal and vertical frequency-response

curves for the first and the second experiments.

These results are normalized by unit excitation

force. The difference of these curves can be considered to be the effect of constraint of soil

at the sides of the foundation. It is clear from

these results that the constraint of soil at the sides of the foundation deceases the horizontal and vertical responses which is caused accom- panied by the rocking motion of the foundation.

It is also noticed that the constraint of soil deceases remarkably the resonant frequency of the rocking motion.

3.2 Effect of Added Concrete Mass

Figs. 5(a) and (b) show the comparison of the horizontal and vertical frequency-response

curves for the second and third experiments. The difference of these curves may be interpreted as the effects of the added concrete masses. The foundation weight of the third state is about

30% larger than the second state. It is noticed

that the effects of added concrete masses appear only on vertical response of foundation. As being expected, it is also seen that the added masses decrease remarkably the resonant fre- quency of the rocking motion. The effect of added masses may also be seen by

comparing the first and sixth experiments, whose

results are shown in Figs. 6(a) and (b). A slight

shift of the resonant frequency of the rocking

motion is also noticed from Fig. 6(b). The main reason that the effect of added masses is less than the cases shown in Fig. 5 is that the increase

of mass the sixth state is only 8% larger than

the first state.

3.3 Effect of Embedment Depth

In Figs. 7(a) and (b), the frequency-response

curves for the third, fourth and fifth experi- ments are shown simultaneously. From these results, it can be seen that with increase of

the embedment depth the horizontal and rocking responses of the foundation tend to decrease. It should be noted that the embedment effects appear only around 4 to 6Hz for the horizontal response

but are more pronounced for the rocking response.

3.4 Effect of Sheet Piles

The sixth experiment was performed soon after

the sheet piles were taken out. Figs. 8(a) and

(b) show the comparison of the frequency-response

curves for the fifth and sixth experiments. The

difference of these two curves may be interpreted

as the effects of the sheet piles. Since the difference is very small, it can be said that

the effect of the sheet piles is negligible.

3.5 Deformation of Foundation

It has been generally pointed out from the theoretical studies that the deformation of foundation beconK remarkable with increase of frequencies (Iguchi and Luco 1982). Perhaps, one of most interesting subjects is to observe

the deformation of the foundation experimentally. Figs. 9(a) to (c) show the deformation

curves of the foundation during a half cycle of harmonic vibrations for f"6, 8 and lOHz of the fifth experiments. The deformation curves are drawn by using the observed results of accel- erometers shown in Fig. 3. These results indicate

that for higher frequencies more than f»8Hz, the foundation may not be considered to be rigid as far as this foundation is concerned.

4. DYNAMIC CHARACTERISTICS OF SURROUNDING SOIL

By comparing the experimental results of six stages of the surrounding soil, the effects of various factors on the dynamic characteristics

of the surrounding soil may be extracted.

4.1 Effects of Constraint Soil Two resonance curves on the points of 35m and lOOm are shown in Figs. 10 (a) and (b). The 35m measure point near to the foundation displayed

the distinct differences between first and second

stage. The curve of second stage gave small amplitudes in comparison with the first stage.

However, The lOOm measure point far from the foundation wouldn't include such a remarkable

differences.

4.2 Effect of Embedment Depth

The resonance curves of the 35m measure point in third, fourth and fifth stages are shown in Fig.

11. The embedment effects to surrounding soil responses might be conflicting to foundation

responses. The increase of embedment depth

decreased the foundation responses, but increased the surrounding soil responses.

4.3 Effect of Sheet Piles

Figs, 12(a) and (b) show the comparison of the 35m point frequency-response curves for the fifth and sixth experiments. The difference of these

two curves may be interpreted as the effects of the sheet piles. Since the slight differences

are recognized.

5. VIBRATION CONTROL METHOD FOR A SHAKING TABLE

FOUNDATION

Several methods can be imaged, in order to decrease the uncomfortable vibrations of shaking

table foundations. It is generally said that the weight of shaking table foundation needs 100 times its power, so as to take good performances

and decrease surrounding soil vibrations.

218
Recently, in constructions of new siiaking table

systems, it can be possible to employ the effective floating foundation which is supported

by air springs on rigid foundations. The floating

foundations have been used in many shaking table

laboratories in Japan. For the shaking table in operation, however, it may be impossible to set

a floating foundation so as to decrease the uncomfortable vibration. In the planning of reinforcement works of the large scale shaking

table foundation, vibration control methods for shaking table foundation were discussed in addition to the mass increase methods for foundation. Severe uncomfortable vibrations of foundations take places in sinusoidal shaking

table motions. If dynamic dampers will set on

the existing foundation, the uncomfortable

vibrations would be decreased much more than the mass increase methods. The dynamic dampers for shaking table foundation have to work effectively

in the frequency range between 3Hz and lOHz. Therefore, the dynamic damper makes resonances

to shaking table motions all over its wide

frequency range. In this reason, the spring

rigidity of dynamic dampers have to be variable.

As the springs with variable rigidities, air

springs and hydraul ic actuators wi 1 1 be proposed.

The high pressured air springs give high

frequency resonance. As for the hydraulic

actuators, the stroke end position of piston

gives high frequency resonance, and the neutral position gives low frequency resonance. If the dynamic damper mass of 600tons will be used, 1000 air springs of 30cm diameter would be necessary,

or hydraulic actuator of 3m stroke and section

area lOOOcm^ would be necessary. Figs. 13 and 14 show the images of air spring and hydraulic

dynamic damper installed in large scale shaking

table foundation of NRIESDP. Fig. 15 is calculation example of dynamic damper effects. Unfortunately, this vibration control methods

have not been adopted to reinforcement works.

6. CONCLUSION

(1) The constraint of the side soil of foundation

has effect mainly on rocking response of embedded

foundation.

(2) Added mass of foundation tends to increase

the horizontal displacement in lower frequency

regions and changes the resonance characteristics of the rocking motion.

(3) Deformation of the foundation becomes

remarkable for higher frequencies more than lOHz.

(4) Even when the backfill soil loose, the embedment effects may appear on the dynamic

response of the foundation.

(5) In order to decrease the uncomfortable

vibration much more, the vibration control methods must be used for the existing shaking

table systems.

ACKNOWLEDGMENT

The Authors wish to thank Dr. Iguchi, Professor

of Tokyo Science University, for his precious

helps to our research. And we express the appreciation to Dr. Edward A. Kausel, Professor

of M.I.T for his precious advice to our

experiments.

REFERENCES

Iguchi, M., Minowa, C, Wakui. E. (1991) , Change

of Dynamic Interaction Characteristics Between

Soil and Large-Scale Shaking Table Foundation

During Reinforcement Works, SMiRT 11 (in Sub- mitted)

Minowa, C, Ogawa, N., Otani, K., Oyagi, N., (1991), Responses of Large-scale Shaking Table

Foundation During Improvement Works, Review of Disaster Prevention No. 151, National Research

Institute for Earth Science and Disaster Pre- vention (in Japanese)

Iguchi, M. and Luco, J. E. (1982), Vibration of

Flexible Plate on Visco-elastic Medium, Jngng

Mech. Div., ASCE, Vol. 108, No. EM6, PP 1103-1120.

219
FOUNDATION CONFIGURATION

PUtQ> KOOH SIDE

Fig. 1 Outline of Shaking Table Foundation

Table 1 Soil Profile

10-

20 -

30 -

40 '

aay&

sot

Saod

Saoi

Sikwitfa

Saod

Sand with

1

J u

L

;y.-i''J ADDED CONCRETE MASS

(a) Pint Eiperimeat

(b) iSo'i'mmI ExpoiajoA

(c) lUid Biperinieat

(d)Fo(Btfa Experimeitf

(e) Hfth Experiment

(0 Sixdi EKpoiment

Fig. 2 Stages of Reinforceie Horks

220
DiqiacoDeot Meter • Afioelero Meter

IS! WlOO W75

S 13

W50 W35

100m i 75m L-5to

T

(i)SectiQO

1.0

Fig. 3 Location of Sensor

») soz

• Freq. OK)

(a) Horizontal (b) Vertical

Fig. 4 Effect of Constraint of Side Soil

16

0.5

A

() 80K O-iS

2W EXP. 3M> GXP, O.IS

O.OS

WD EXP. 3KD EXP.

• Frtq. »2 0« »«

(a) Horizontal (b) Vertical

Fig. 5 Effect of Added Concrete Mass (Comparison of 2nd and 3rd Stage)

221
(a) Horizontal (b) Vertical

Fig. 7 Effect of Eabedient Depth

(a) Horizontal (b) Vertical

Fig. 8 Effect of Sheet Piles

222
(a)6Hz (b)8Hz (c) lOHz

Fig. 9 Defonation Curves of Foundation During a Half Cycle of Vibration (5th Stage)

1st 1st

-t- 2nd + 2nd

(a) H35i Horizontal (b) WiOOi Horizontal

Fig. 10 Comparison of 1st and 2nd Stage

rwajUHCT » It rugUENCY IN Bt 3rd. + 4th, X 5th 3rd. + 4th. x 5th

(a) H35i Horizontal (b) H35i Vertical

Fig. 12 Comparison of 2nd, 3rd and 4th

223
A

N H

I r

c . I0 I

N A

T

/ I

t 0

•

1.89

.SB

s u

raoniQcy n it 5th

+ 6th

(a) W35i Horizontal

A

H H

I f

0 I

M A i.za

I

/

1

' ' ' ' —1—1 1 1 u

nEQUEicv in iz

IS

5th

+ 6th

(b) H35B Vertical

Fig. 12 Conparison of 5th and 6th

Dynamic dampers consisted of hydraulic cylinders Dynamic dampers consisted of air springs

Fig. 13 Image of Vibration Control Fig. 14 Image of Vibration Control

. by Hydrau I ic Actuators by Air Springs

8RSE HRSS > 8RSE SPKINC- CKOUNO HASS « CKOUNO SPRING'

15.6 HOOO.O

SOO.O 130000.0

BRSE VISCOUS- SOO.O

GROUND VISCOUS* 2000.0

OTNR. DRHPER HRSS RNO H • .00000 .00000

. MAX VRLUE - .020204 RT Hi 1.4 a. foundation

5.0

FREQUENCY IN Hz

Fig. 15 Calculated Foundation Responses with Vibration Control Systei

224
Safety and Economic Benefits Realized From Upgrading

the Lawrence Berkeley Laboratory

by

1 Donald G. Eagling

2james R. Hill

ABSTRACT

The Lawrence Berkeley Laboratory (LBL)

Initiated a comprehensive seismic safety

program following the 1971 San Fernando

Earthquake. Since then, 34 buildings have

been strengthened, and emergency

recovery facilities improved significantly.

The 7.1 M Loma Prieta Earthquake,

centered 65 miles south of Berkeley,

produced 12%g acceleration at LBL

compared to 65%g near the epicenter. It caused only non-structural damage at LBL,

because those buildings susceptible to structural damage from minor ground

shaking had been strengthened. For

example, two of LBL's major laboratory

buildings would have sustained diagonal

tension cracking in brittle reinforced

concrete bearing wall window piers if they

had not been buttressed to prevent

damaging deflections. After the

earthquake, it would have been mandatory

to evacuate these two buildings and find

130,000 square feet of laboratory space to lease for 350 people. It would have then

taken two years before these buildings

could be strengthened and reoccupied.

Research would have been disrupted over one year and the U.S. Department of

Energy would have suffered a loss over $30 million if these two buildings had not

been upgraded.

Including other facilities that would have

suffered structural damage, upgrading LBL

saved over $50 million at a cost of less than

$4 million expended since 1971.

KEYWORDS: seismic safety; Loma Prieta;

cost savings.

1. INTRODUCTION

The Lawrence Berkeley Laboratory is a

multiprogram research facility operated by

the University of California for the U.S.

Department of Energy (DOE). It is located

in the San Francisco Bay Area on the hills above the University of California at Berkeley (UCB). The LBL site

encompasses 130 acres including 80

buildings and has a population of 4,000.

As can be seen in Figure 1, the seismically

active Hayward Fault passes at the foot of

the hill near the Laboratory's main

entrance. LBL buildings are shown in the

foreground of Figure 2 above the UCB

Campus. The San Andreas Fault lies

about 15 miles west across the bay just

beyond the Golden Gate Bridge on the

horizon.

In 1971, following the destructive San

Fernando Earthquake in Southern

California, LBL initiated a comprehensive

review of its facilities and operations to improve earthquake safety. Since then, as shown in Figure 3, 34 buildings have been

Retired Plant Manager, University of

California, Lawrence Berkeley Laboratory

^fvlanager. Natural Phenomena Mitigation

Programs, U.S. Department of Energy

225
strengthened. Active and potential

landslide areas have been stabilized, life lines have been hardened, non-structural

hazards have been mitigated and

emergency recovery facilities have been

upgraded and augmented in preparation

for a great earthquake.

As shown in Figure 4, the Loma Prieta

Earthquake, magnitude 7.1, occurred on a

25-mile segment of the San Andreas Fault

about 65 miles south of the LBL site.

Ground acceleration at LBL was only 12%g

as compared to 65% near the epicenter

and 25% on soft sediments near the bay a few miles west of LBL where a portion of

the Cypress Street viaduct collapsed. The

Loma Prieta Earthquake caused only nonstructural damage at LBL, because those

buildings which would have been

susceptible to structural damage in 1971

from 15%g accelerations had been

strengthened by 1 989. The cost to upgrade

LBL facilities since 1971 has been less

than $4 million. If susceptible buildings

had not been strengthened, it is estimated

that LBL (and DOE) would have suffered

financial losses in excess of $50 million.

2. BACKGROUND

The Loma Prieta Earthquake triggered a backlog of developing concern about

liability which had grown with each of the

smaller but damaging earthquakes

experienced in California during the last

decade or so.

Public attitudes about earthquake hazards

and the financial risks of liability have

changed considerably during this period.

Engineers have become more outspoken

about building types which are substandard and hazardous to occupants.

Owners and managers have in turn

become more aware of their corporate and

sometimes personal liability for the safety of

occupants, particularly when a building is known to be unsafe.

Allowing people to continue to occupy a

building after earthquake damage has

shown it to be a potential collapse hazard

is untenable to most responsible owners. It is a liability they are unwilling to bear.

People are also becoming aware of the

enormous costs associated with evacuating a damaged building and relocating

occupants elsewhere until the building can be repaired and made safe for

reoccupancy. Of course, the cost of lengthy

disruptions to operations and research is devastating in itself.

Examples from the Loma Prieta Earthquake

are many. A very large department store in Oakland was out of business for one year

following the earthquake. The Oakland

City Hall, a high rise building, sustained

structural cracking and has not been

reoccupied one and a half years later. Several substandard buildings at Stanford

University which remain unoccupied will require enormous costs for rehabilitation

before they can be reoccupied. None of

these buildings collapsed, but the signs of

structural damage are obvious and

concerns for liability and public safety

required their evacuation.

3. THE LBL EXPERIENCE

Earthquake shaking at LBL due to the

Loma Prieta Earthquake would not have

caused the collapse of any substandard

buildings even if they had not been

strengthened. However, certain buildings

would have sustained structural cracking

which would have required the evacuation

of occupants, repair and strengthening

before reoccupation. Prime examples are

the two major laboratory buildings (50A

and 50B) shown side by side in the center

of Figure 5 as they appeared in 1971

before they were strengthened. The two

tower elements are seven stories high on

the downhill end and four stories on the

uphill end. They are reinforced concrete

bearing wall structures of identical design.

The end walls are shear walls which

provide satisfactory resistance in the

transverse direction. Lateral resistance in

226
the longitudinal direction, provided by

reinforced concrete moment connections,

was deficient in that deflections necessary

to resist earthquake forces would permit

structural failure to take place in brittle bearing elements.

Shaking due to the Loma Prieta

Earthquake would have caused diagonal

tension cracking in brittle reinforced

concrete window piers in the sidewalls if these towers had not been buttressed in the longitudinal direction to prevent

damaging deflections. Consequently, it would have been necessary to remove the

350 occupants and find 130,000 square

feet of laboratory and office space for them

to continue research for two years or more

until these buildings could be repaired and

strengthened for reoccupancy. This would

have cost the U.S. Department of Energy

an extra $30 million and disrupted vital

research programs for at least one year.

A brief explanation of the weaknesses

inherent in the original design and the

measures taken to prevent structural

damage follows.

Figure 6 shows one of the towers in 1973

after it was determined that these buildings

would suffer catastrophic collapse in a

major earthquake and structural damage in minor shaking. Lateral resistance in the

longitudinal direction was originally

provided by moment connections between

the 8-foot v/ide vertical elements at the end

of the window rows and the heavy

longitudinal spandrels between horizontal rows of windows. The 14-inch thick side

bearing walls, perforated with windows,

transmit the vertical load down through the

2-foot wide window piers. The 8-foot wide

end piers were heavily reinforced to resist

longitudinal bending, but intermediate 2-

foot wide window piers were only nominally

reinforced. In an earthquake, the 8-foot

wide end piers could resist the longitudinal

load, but would allow the short brittle

window piers to deflect enough to crack in diagonal tension. To compound the

problem the reinforced concrete beams

supporting heavy concrete sunshades

were cantilevered out from the brittle

window piers. In minor shaking, the piers

would develop typical "X" cracking. In heavy shaking they would disintegrate

leading to collapse of the towers. Soon

- after these hazards were identified the

sunshades were removed as shown to reduce the risk until the buildings could be

strengthened. At that time these buildings

were not evacuated. The concern for

liability was not as strong in 1971 as it is today.

Figure 7 shows a diagonal tension failure

in a non-ductile window pier in the Misawa

Commerce High School as a result of the

1968 Tokachi-Oki Earthquake.

As shown in Figure 8 the solution was to

install heavy stiff buttresses on either side

of the uphill end of each building to limit deflection of the window piers during

earthquake shaking. Each buttress was

wrapped around the end wall and in-filled into window openings. Extensive bolting

into the floor diaphragms and within the

periphery of the in-filled windows tied the

buttresses to the buildings. By observing

the pattern of windows above the buttress,

one can tell where windows were in-filled.

Figure 9 shows the two buildings after

strengthening was complete. One of the

buttresses can be seen at the far left end of

the nearest tower. The cost of

strengthening these buildings in 1974 was

about $0.5 million. The cost today would

be about $1.5 million. Replacement value

today is $60 million. This expenditure

saved $30 million in 1989 when the Loma

Prieta Earthquake took place and will save many lives in a great earthquake.

4. CRITERIA FOR UPGRADE

Structural criteria for strengthening these

buildings were based upon a maximum

credible earthquake of 7-1/4 magnitude on

the nearby Hayward Fault with a peak

227
ground acceleration of 75%g for dynamic

analysis. Results were checked by static

analysis using a minimum base shear of

20%g. These criteria are used for all normal LBL buildings, both for new

construction and upgrading existing

facilities. For essential buildings having

emergency recovery facilities or hazardous

materials, a static base shear of 28%g is used.

5. CONCLUSION

Including other facilities that would have

suffered structural damage, upgrading

substandard buildings at LBL saved over $50 million during the Loma Prieta

Earthquake. The total cost to upgrade

these buildings since 1971 is $4 million.

LBL will spend another $10 million over the

next few years to complete its seismic

upgrade program in preparation for the

great earthquake that is sure to come. In the meantime, the work accomplished thus

far has produced large cost savings for the

U.S. Department of Energy.

The LBL experience in the implementation

of its seismic upgrading program served as

a basis for the development of the "Seismic

Safety Guide" (Eagling, 1983) which was

prepared for the Office of Nuclear Safety,

U.S. Department of Energy. The Seismic

Safety Guide provides the managers of

facilities with practical guidelines for

administering a comprehensive earthquake

safety program.

6. REFERENCE

1. Eagling, D.G., et al, (1983) "Seismic

Safety Guide," LBL-9143, Lawrence

Berkeley Laboratory, University of

California, Berkeley, California
Figure 1 . Aerial View of the

Lawrence Berkeley Laboratory (LBL)

229
i

230
231
1-80 Green Valley

Fault

HealdsburgRodgers

Creek Fault

t-5

^ Sacramento

99 16

A 104 Rancho Seco Nuclear Power Plant Antioch Fault

Concord Fault

LLNL

o Stockton

Greenville Fault

O Modesto

San Gregorio Fault

Calaveras

Fault

Sargent Fault

• Epicenter of Oct. 17, 1989

Loma Prieta Earthquake

Segment that slipped in

1989 Loma Prieta earthquake

1-5

Moss Landing

Fossil Fuel Power Plant

99

San Andreas Fault Not to scale

Figure 4. Active Faults

Local to the San Francisco Bay Area

232
233
Figure 6. Removal of Concrete Sunshades from

Building SOB (About 1972)

234
Figure 7. Diagonal Tension Failure in Window Pier

Misawa Commerce High School

Tokachi - Oki Earthquake 1968

235
236


Structural Regulations of Buildings and Development

of Technology in Japan

By

Yuji OHASHI+1

sumRY

This paper presents a outline of struc- tural regulations of the Building Standard

Law. which prescribes the basic building

restriction of Japan, and relation of building restriction and developement of building

technology, showing some examples of new structural technique.

Building regtilations in Japan are based on the Building Standard Law. There are a

lot of standards in the law. There is a

special approval system by the Minister of

Construction for new structures which do

not comply with the standards to prevent

the resririction from becoming too rigid.

The system has provided the way for a lot of new building structure to date. Airsupported membrane structure, base isolation structure and response control systems are the latest successful examples of this

system.

KEY WORDS : Building Restriction, Structural

Regulation. Seismic Design Method, Building

Codes, The Building Standard Law. Approval

System

1. OUTLINE OF JAPANESE BUILDING RESTRICTION

1.1 Feature of the land and the buildings of

Japan

There have been many disartrous earthquakes in the history of Japan, and every

year, the country is attacked by typhoons.

These extreme forces of nature have

inflicted great damage on building structures, and it has long been a major task of

Japanese building technology to find ways of preventing this constant danger.

Traditionary, wood was used to buildings, from the plentiful supplies in the

mountains. Houses were constructed with

wooden columns and beams, using teclmiques

of disaster prevention developed over years

and years of experience. Most of the urban areas in Japan can be

regarded as a continuation of old towns where wooden buildings are densely packed.

The urban areas in Japan have been subject

to the attacks of a lot of conflagrations

in the past.

The cause of conflagrations is not only

Hiroyuki YAMANOUCHI+2

accidental fires but also the simultaneous

outbreak of fires in many places due to

a large scale earthquake. The Great Kanto Earthquake (1923) was typical of the latter,

with approximately 700,000 buildings being

destroyed, and some 140,000 persons dead or

missing.

Thus, it is highly necessary that buildings in Japan be able to safely withstand

attacks of earthquakes, typhoons and fires. It is the purpose of the building restriction of Japan to safeguard against the

danger.

1.2 Purpose and constitution of the Building

Standard Law Building Standard Law was enacted in 1950

considering the abovementioned conditions,

and since then has been subjected to many

revisions in accordance with the advances in technology and the conditions of building

disasters that have occurred.

The building standard statute consists of

the Building Standard Law, the Building

Standard Law Enforcement Order as cabinet

order, the Building Standard Law Enforcement

Regulations as Ministry of Construction

Order, and Notifications of Minister of

Construction based on these regulations.

The purpose of the statute is to 'safeguard the life, health, and properties of

the people by providing minimum standards

concerning the site, structure, equipment,

and use of buildings, and thereby to con- tribute to the furtherance of the public

welfare.

'

The statute comprises the ' general provisions" on procedures for building con- struction, the 'building codes" on standards

for safety and sanitation of all buildings

in the whole country, and the 'zoning codes' on standards for ensuring the safety and

appropriate environments in urban areas

to be applied to city planning areas based on the City Planning Law. as well as penal

provisions.

1 Senior Research Engineer,

Structural Engineering Department, Building

Research Institute, Ministry of Cnstruction

Tatehara-1, Tukuba-City, Ibaraki, Japan

*2 Dr. Eng. , Director of Structural

Dynamics Div. Structural Engineering Dep. BRI
1.3 Outl ine of the regulations of the Building Standard l.aw

1.3.1 General provisions (Procedures for

building construction)

0 Building confirmation

Generally in the case where a building is to be constructed, the building client must

sulimit the construction plan to a building

official of the local govarnment. and re- ceive the confirmation of the official as to whether the plan satisfy the regulations of

the Building Standard law. If the plan does

not meet the regulatons, construction will not be allowed.

There are about 1600 building officials, over 5000 other local officials who concern with building restriction and about 300 local governments throughout Japan . There are about one million building confirmations

per year. 0 Design and supervision by qualified build

ing engineers

The design and the supervision of con- struction works of buildings exceeding a certain scale must be performed by qualified

building engineers. The number of the building engineers is about 700 thousands since

the enforcement of the Building Standard Law. This number includes a lot of dead engineers.

The number of active engineers is uncertain.

A lot of the engineers work as not the

designers of buildings but the supervisers

of construction works.

(S) Inspection after the completion of con- struction

Once a construction work have been

completed, the building client must submit

a notification of this effect, to the building official. The said building must receive an inspection by the building official, as to whether the buildiiig satisfy the regulations. A certificate of inspection is then

granted when it is confirmed.

@ Periodic in.spection and periodic report

system

The owners of certain buildings such as

hospitals, hotels, department stores, theaters, collective dwellings and offices ex- ceeding certain scales, and the owners of

escalators and elevators and other building

equipments must have regular inspections con cerning the conditions of such buildings and

building equipmerils by a qualified engineer,

every 1 to 3 years, and must report the

results of inspections to the local government.

1.3.2 Building codes

Building codes stipulate standards for

structural safety, fire prevention and sanitation.

1) Standards for structural safety

Regarding structural safety, the codes

lay down the standards of detailed struc- ture for each type of structure such as wooden structure, steel structure and reinforced concrete structure, and standard of

structural calculation.

In order to ensure the safety of a building, the Building Standard Law Enforcement

Order provides that structural calculations must be performed to confirm the safety of

large buildings. The range of building for

which structural calculations must be done

is as follows,

type of construction scale

wooden structure n^3 or A^BOOnf

others n^2 or A^200m'

(where n: number of stories of the building

A: total floor area of the building)

The method of the structural calculation was revised in 1980, and has been in force

since June 1, 1981. Adding to the conventional allowable stress design method for

elastic range, this new method of structural

calculation determines design method considering ultimate stress including the stress within a plastic range. In addition, story

drift, rigidity factor and eccentricity

factor should be checked to ensure a balanced structural plan.

The Enforcement Order prescribes the

loads, the external forces, the allowable stress of building materials necessary for

structural calculation, and also the methods

of calculation that are necessary for structural safety in accordance with the height

and structure of buildings.

Regarding the structural calculation, to confirm structural safety, at least five

kinds of loads and external forces must be

checked, nanely. fixed load, imposed load, snow load, wind force and seismic force. Values of allowable stress are provided

for the common materials such as timber,

steel, concrete, etc. Dual allowable stress system is adopted, namely, allowable stress

for permanent loads and allowable stress for

temporary loads are provided.

Values of allowable stress for permanent

loads are half to one third of ultimate stress of materials, values of allowable

stress for temporary loads are equivalent

to yield level or elastic limit.

The stress generated in each of structu

ral members by the combinations of loads

and external forces, does not exceed the

240
allowable stress. The purpose of the current seismic design

code is that buildings should withstand

moderate earthquake motions with almost no

damage, and should not collapse nor harm

human lives during severe earthquake motions

Regarding seismic design, adding to the

allowable design method, limitations on story drift, rigidity factor, eccentricity

factor and ultimate lateral shear strength

of each story are provided.

The lateral seismic shear Qi of i-th

story is determined by

Qi = Ci Wi (1) where Ci is the lateral seismic shear co- efficient of the i-th story and Wi is the

weight of the building above the i-th story.

Lateral seismic shear coefficient of the

i-th story is given by

Ci = Z Rt Ai Co (2) where Z is the seismic hazard zoning co- efficient. Rt is the design spectralcoeffici

ent, and Ai is the lateral shear distribution factor. The standard shear coefficient

Co is 0. 2 for moderate earthquake motions

and 1.0 for severe earthquake motions. For buildings higher than 31 m. or with

irregularities exceeding the specified limits, the calculated ultimate lateral shear

strength of each story must not be less than

the specified ultimate lateral shear Qun : Qun = Ds Fes Q (3) where Ds is the structural coefficient. Fes

is the shape factor, and Q is the shear for severe earthquake motions which is given by

Eqs. (1) and (2) letting Co = 1.0. The structural coefficient Ds is determined by the energy absorbing capacity of

the structure. The shape factor Fes is the

product of the factor Fe, which is determined by the excentricity facte r and the

factor Fs. which is determined by rigidity

of each floor.

2) Approval of the Minister of Construction

for new structure and special material

There is a special provision which makes

allowance for construction using new struc- ture and materials, in response to progress

in constrution technologies and development

of new materials. According to this provision. Article 38 of the Building Standard

Law. such a building plan requires only

special approval by the Minister of Construction, overriding standards of the building codes. The article says as follows.

The provisions of this Chapter or those

of orders or cordinances based thereon shall not apply to buildings using special building materials or methods of construction

unanticicipated thereunder, in cases where

the Minister of Construction determines that

the said building materials or methods of

construction are equal or superior to those

specified in the said provisions.

Before the approval of the Minister, the

safety of the buildings shall be evaluated

by the committee of the Building Center of

Japan under the supervision of the Minister.

The buildings approved by this system in recent years average about 150 per year. Super-highrised buildings higher than 60m

have been approved by this system. To evalu- ate the safety of such buildings, adding the

allowable stress design, dynamic response

analyses are taken using records of the

ground motions of past earthquakes, where some typical ground motions such as' El Centro' and the ground motions which were recorded near the construction site of the

building are used. At this time. The scales

of the ground motions are normalized and 2 levels of the ground motions are adopted.

Maximum velocity of the ground motions of

level 1 is some 25 cm/sec and that of level

2 is some 50 cm/sec. Response from level 1 ground motions shall not exceed elastic

limit of the building materials of structural members and that of level 2 ground

motions shall be within apt value of ductility factor.

3) Other regulations

Outline of other regulations of the

Building Standard Law is as follows. As for fire prevention, the codes stipulate that roofs and external walls of buildings in a specified area be made fire

preventive construction, that department stores and other buildings for special use be made fireproof construction, and that

fire compartments be created in large scale

buildings according to a specified floor

space. Standards for emergency facilities,

including hallways, staircases, smoke-exhaust

equipment, etc. are also stipulated. To secure sanitation, the codes lay down

standards for natural lighting and ventilation of rooms, structure of lavatories, structure of elevators and escalators, etc. The zoning codes prescribe relationship

between sites and roads, land use designated

districts, ratio of total floor area to site

area, height of a building, restrictions on buildings in fire protection districts. etc.

2. OUTLINE OF THE HISTORY OF SEISMIC CODE OF

BUILDINGS OF JAPAN

With the end of the Edo Period and the

241
Meiji Restoration in 1868, Japan set the course for a modern nation. Abolishing the

insulation policy of the Edo Era, the newly

established Meiji government carried out various measures to open the country and

catch up with western advanced nations.

Modem structures such as steel structure and reinfoced concrete structure were introduced from these nations.

With the outbreak of World War I . the

Japanese economy made outstanding progress, causing concentration of the population and

industries in urban areas and making it necessary to establish urban planning and

building regulations in big cities such as Tokyo and Osaka.

In 1919, the City Planning Law and the

Urban Building Law was enacted for six major

cities. The two laws served as the starting

point of modern building regulations in Japan.

The Urban Building Law provided for

such systems as building lines regulations

and zoning systems. It also stipulated

standards for buildings in terms of f ireproofness, safety in structural strength,

and guarantee of sanitation.

Regarding structural safety, the codes

stipulated the standards for detailed struc- ture and structural calculation. As loads,

only fixed load and imposed load were provided in regulations of structural calculation.

In accordance with progress in construc- tion technology, and economic and industrial

growth, the Urban Building Law underwent

several revisions.

In 1923. the Great Kanto Earthquake

attacked the Kanto Area, destroying a huge

number of buildings. Fires caused by the

earthquake burned down wooden houses in the region, killing about 140,000 people.

In reflection on the massive damage, the

Urban Building Law set up seismic force and

made it obligatory to calculate the seismic

strength of building, where seismic coefficient shall be more than 0.1 in 1924. The

article says. ' Horizontal seismic coefficient shall be more than 0.1.' That means each building shall be designed and const- ructed so that it can resist a lateral force

of at least 10 % of its weight.

Regulations on safety of buildings in Japanese legislation have since centered on fireproofness and earthquake-proofness. The lesson of the earthquake demonstrated

the great importance of bearing walls in resistance to earthquakes.

There had been only one allowable stress

for each material which is about one-third

of the yield level of the material.

During the World War II, because of

the shortage of construction materials, a dual allowable stress system was introduced

into Japan. That is. the stress caused by

permanent loads was to be less than the

allowable stress for permanent loads, and

the stress from temporary loads was to be

less than the allowable stresses for temporary loads. During the World War II. building rest- riction based on the Urban Building Law were suspended except for some regulations on fire prevention.

After the World War II . the Urban Building Law was restored. In accordance with the enactment of the Constitution of Japan, the

Building Standard Law was enforced in 1950,

abolishing the Urban Building Law. The ap- plication area of the Building Standard Law

is the whole nation, not limited districts as in the Urban Building Law. Technically speaking, the Building Standard Law was basically the extension of

the Urban Building Law. But. regarding

structural calculations, the standard was

revised, where dual allowable stress system was adopted and seismic coefficient became

0.2 and allowable stress for temporaly load

like seismic load regulated about elastic

limit of building materials.

Nearly 40 years have passed since the

Building Standard Law was enacted in 1950. During the period. Japan has made remarkable

progress in economy and building technology.

And the Building Standard Law has been

revised several times, taking into account progress in building technology and damages

to buildings caused by earthquakes.

The Tokachi Earthquake in 1968 showed

importance of the ductility as well as strength of structural members. As a result,

the structural codes of the Building Standard Law stipulated ductility of buildings

with the revision of reinforcement for

shearing of pillars of reinforced concrete buildings in 1971. During that time, the development of

computer technology also promoted structural analysis knowhow. Through computerized

seismic analysis, construction technology

of super-highrise buildings was developed

around 1960, using the earthquake ground

motions recorded in the United States. The construction of super-highrise buildings requires approval of the Minister of

Construction based on the regulation of

Article 38 of the Building Standard Law and

242
evaluation by the committee of the Building

Center of Japan before the constructin work

of the building. In order to evaluate the

safety of the buildings which use special strucure or special materials, the Building

Center of Japan was established in 1965 under the supervision of the Minister of

Construction.

Against the background of these factors,

the Ministry of Construction led the development project for new seismic design method

from 1972 to 1977 with a view to thoroughgoing revisions of seismic design method in

the Building Standard Law. And in 1981,

the drastic revision took effect with

dynamic considerations on seismic calculation. For instance, the seismic force,

which had been set at more than 0. 2 of

seismic coefficient, was fixed in accordance

with natural periods of buildings. The revision also called for two-stage designing,

namely design against the seismic force

equivalent to self weight of a building,

in addition to the conventional allowable stress design against seismic force of 20%

ofthe self weight of the building. Moreover,

retained ultimate lateral shear strength was to be calculated with consideration for

ductility of a building, not allowable

stress. Revisions and addition have been made to standards for detailed structure in accord- ance with progress in structural materials,

such as steel products, and development of new structuralforms, including wood frame

construction.

3. BUILDING RESTRICTION AND DEVELOPEMENT OF

TECHNOLOGY

3. 1 Building regulation and technological

level

A lot of technlogical standard s are established in the Building Standard Law. Most of the standards are regulated based

on the general technological level and so- cial and economical conditions of the times

when the standards are regulated. Also, they

reflect the average form and structure of

buildings of the time. Thus, in a rapidly changing society with

highly advanced technology, occasionally the

standards of the existing building standards cannot deal with the newly developed structure and materials. Some of these buildings

do not fit in the formal framework of the

existing building regulations.

Sometimes, these existing standards stand

in the way in developing these new buildings. However, there is a special provision,

which allows the construction of these

buildings that do not satisfy the standards

by the approval of the Minister of Construction, as explained in sction 1.3.2 of

this paper. This approval system has covered the

construction of these new buildings and has

contributed the developement of building

technology. Examples of those buildings are super-highrised buildings and pre-fabricated

houses, etc. 3.2 Relation between building restriction

and developement of technolog,y

Here, the writer shows relation between

building restriction and developement of

technology, using air- supported-membrane

structure, base-isolation structure and the structure which adopts response control

system as examples.

3.2.1 Air-supported-membrane structure

Air-supported-membrane structure usually

makes its roof with membrane and makes

internal air pressure slightly higher than

external air pressure to add the ridigity of

the roof. Large-scale air-supported-membrane struc

-ture features a lot of special characteristics when compared with conventional

structures. The following are some of these

characteristics.

DThis type of structure features an ex- tremely large roof with low rigidity. Wind

pressure acting on the roof is different

from that of usual rectangular medium sized

buildings, which have high rigidity.

2) This structure requires a snow-melting

system. Snow load on the roof requires

higher internal air pressure, which dose not good for usage of the building because the

the higher air pressure is, the higher

velocity of the air flow at doorways is. Thus, a snow-melting system is required.

However, in Japan, we do not have the ex- periences of such a system.

3) This structure uses membrane, special

material.

4) The single huge space created by this structure require s a new concept for the

fire prevention and emergency facilities of

the building.

5) This structure requires special facilities which sustains internal air pressure. The facilities must be properly operated

and well maintained for stability of the

structure. The standards of the Building Standard Law do not anticipate these specific charac243
teristics of this structure. These characteristics requires the ap- proval of the Minister of Construction and

evaluation of safety by a committee of the

Building Center of Japan, before the const- ruction of the building. To this effect, a new concept had to be created for design

and evaluation which covers these characteristics.

Around 1980, some major Japanese const- ruction companies began to develop the technology of large-scale air-supported membrane structure like a baseball stadium with roof,

studing such existing stadiums in the United

States. A special committee was set up in the Building Center of Japan in 1983, based on the agreements among the construction

companies, the Building Center of Japan,

the Ministry of Construction and exparts.

The committee, composed of representatives

from these parties, began to work out the

standards for evaluation of the safety of

the large -scale air -supported membrane

structure.

In December 1984, the evaluation standard was made. This work promoted planings of

such buildings. The decision to construct

the Tokyo Dome, first baseball stadium with

roof in Japan, was made. The evaluation

standard was used as a guideline for the

design of the Tokyo Dome. After the dome was

designed, the evaluation committee of the

Building Center of Japan made the safety

evaluation based on the standard. After

that, the Minister of Construction approved

the construction of the dome with Article

38 of the Building Standard Law. Since then, a lot of air-supported mem- brane structure of medium size have been

constructed in Japan.

3.2.3 Base isolation structure and response

control system

Advances in response analysis technology

had given the knowledge that seismic force

could be reduced by lengthing the natural

period of a building, and research was carried out into how this could be done. Laminated rubber bearing was invented around

1980. The bearing is made of several alternating layers of rubbe r sheets and steel

plates and can withstand great forces in a vertical direction, such as the self-weight

of the building, while horizontally it is

flexible in character. When they are install

ed in the foundation of a building, or the

columns of its lowest floor, they can sup- port the self-weight of the building and at same time, because of their horizontal flexibility, can lengthen the building' s natural

period. That reduces seismic force. This

technique is known as base-isolation. In Japan, it is already beyond the experimental

stage and has been introduced gradually into

actual construction. Compared with conven- tional structure it is capable of considerably reducing the effects of seismic force. Research has also done into other methods

of absorbing or controlling the response

from an earthquake. One of them is to install a 'damper*. Dampers can be installed

independently in the walls or structural

braces, or can be used in combination with

laminated rubber bearings as part of a base

isolation structure system.

Another technique involves installing a large mass on the roof or top floor of a

building. The special characteristics of

themass mean that when it is connected

through the superstructure of the building

to springs and dampers, the vibration caused

by external forces such as seismic force is

converted to the vibration energy in the

mass. The superstructure of the building

will then vibrate less violently than it would do without the mass. This kind of

system is called 'tuned mass damper'.

Structures which use such devices in

addition to conventional structural members are known as ' response control structures'

.

There are already several of them in Japan,

and further research is still under way. The most advanced systems of response

control involve sensors and computers. The sensors can measure the intensity of an earthquake motion or a strong wind force,

and response vibration of th e building

itself. Signals from the sensor are fed into

a computer, which calculates and triggers

off a

' control force ' in the building to reduce the response of the building. This

kind of system is known as 'active mass

damper'

.

In other structures, a computer can calculate to what extent the rigidity of structural braces should be modified, and can alter their rigidity accordingly.

These systems which can reduce the

response of structure by using appropriate

energy are known as ' active response control

system '. The research and developement of

the systems are now being carried out. These systems are special ones, and the

regulations of the Building Standard Law do not anticipate. Thus, the construction of

the building which uses such a system re- quires approval of the Minister of Const- ruction and evaluation of the committee of

244
the Building Center of Japan. It is supposed

that it will take long time to evaluate

safety of such systems because we do not have the experiences of planings and const- ruction works of the kinds of the systems.

In order to reduce the time of evaluation

of the actual construction, prior to the

evaluation, special committees which studies

model buildings that adopt such systems are sometimes set up base on the requests of the

companies which are developing these systems.

These committees are studing basic structural problems of the systems. They contrib- ute to reduce the time of the evaluation of

actual construction plans.

4. PRESENT PROBLEMS RELATED TO BUILDING

RESTEICTION AND DEVELOPEMENT OF TECHNOLOGY

A lot of buildings which use newly

developed technology that dose not comply

with the regulations of the Building Standard Law have been constructed by the approval of the Minister of Construction and evaluation of the committee of the Building

Center of Japan. The approval and evaluation

system based on Article 38 of the Building

Standard Law contributes the developement of

building technology. However, there are some problems in this system and its smooth

administration. The following are some of

these problems.

1) Technical evaluation requires a lot of

time and effort

The evaluation must be conducted care- fully to ensure the building safety. Thus,

this system requires much time and manpower. This has been pointed out not only by const- ruction companies and structural consultants

but also by officials and experts in charge

of the evaluation, corresponding to the i ncreasing number of buildings to be evaluated.

2) Intermediate coordination may be required

during the technological developement stage.

In general, official evaluation is to be

conducted after the completion of the building design. However, a building incorporating extremely new technologies may require

coordination between the developer or designer and officials in charge during a technological developement or design stage

to promote smooth building developement. The

Building Center of Japan recently started

' technical guidance ' service or setting up

of above mentioned special committee which

studies new technique to handle such cases, when necessary, prior to the official evaluation based on the request from the developer of the technique.

5. CLOSURE

Building regulations in Japan are based on the Building Standard Law. There are a

lot of standards in the law. There is a special approval system by the Minister of

Construction for new structures which do not comply with the standards to prevent the

resririction from becoming too rigid. The

system has provided the way a lot of new building structure to date. Air-supported

membrane structure, base isolation structure

and response control systems are the latest

successful examples of this system.

This system requires evaluation of the

committee of the Building Center of Japan

prior to the approval of the Minister.

However, there stil 1 remain some problems

to be solved. It is hoped that the system

will be simplified to respond to these

problems, creating a favorable environment

that will promote smooth technological developments while at the same time satisfying

the purpose of the building restriction.

245

Seismic Hazard Assessment of Liquefaction Potential

at Mormon Island Auxiliary Dam, California, USA

by

David W. Sykora'

Joseph P. Koester* Mary E. Hynes^

ABSTRACT

A seismic hazard analysis was conducted

using various combinations of a^^^ and earth- quake magnitude at a dam site where post-earthquake stability is suspect. Results of the limit equilibrium slope stability analysis using residual

strengths and pore pressures were combined with

seismic hazard calculations to estimate the likeli- hood of liquefaction of dredged foundation

gravels and consequent failure of the dam-foundation system. The hazard associated with each

interval of earthquake magnitude was calculated.

KEYWORDS: Earthquake; Seismic Hazard; Post- earthquake Stability

INTRODUCTION

A seismic stability evaluation of the

Mormon Island Auxiliary Dam (MIAD), part of the

Folsom Dam and Reservoir Project, located near Sacramento, California, was performed as a

cooperative effort between the US Army Engineer

Waterways Experiment Station (WES) and the US

Army Engineer District, Sacramento (SPK). A

location map of the project is shown in Figure 1

.

From a Phase I study (Hynes-Griffin et al. 1988), it was determined that the portion of MIAD

with shells founded on dredged tailings will not be

stable against large-scale internal sliding during

and after the design earthquake event. Remedial measures were recommended over this length of the dam. It was concluded from Phase II studies

(Wahl et al. 1988) that the segments of MIAD with

shells founded on rock or on undredged alluvium

will be stable both during and after the design

earthquake event, hence, remedial measures in these sections are not required.

Further engineering analysis in the form of

limit equilibrium slope stability analyses and

seismic hazard assessments were conducted to

assist in the decision making process for remedial measures at MIAD. A re-anaiysis of post-earthquake slope stability, using stratigraphic and

cyclic strength data obtained at the site, was

conducted to evaluate the slope stability of a

representative section of MIAD founded on

dredge tailings (gravels) if subjected to the design

earthquake or less severe shaking. A seismic

hazard assessment (SHA) was performed to determine the risks associated with different combinations of earthquake magnitudes and peak

horizontal accelerations causing liquefaction of dredged foundation gravels.

These additional engineering studies were

aimed at assessing the potential for occurrence of a damaging earthquake at the site, and provide

background analyses from which possible postearthquake embankment configurations could be

envisioned. An exhaustive analysis was not conducted because of the short time frame within

which the results were applied to the construction

of remedial measures. Calculated return periods

are considered to be lower-bound values because

of the conservatism applied throughout the analysis. The results of the seismic hazard analyses

are summarized in this paper.

Project History and Description

The Folsom Project was designed and

built by the Corps of Engineers during the period

1948 to 1956. Upon completion of the project in

^ Research Civil Engineer, Geotechnical Laboratory, US Army Engineer Waterways Experiment Station, Vicksburg, Mississippi, USA

^Supen/isory Civil Engineer, Geotechnical Laboratory, US Army Engineer Waterways Experiment

Station, Vicksburg, Mississippi, USA

247
May 1956, ownership of the Folsom Dam and

Reservoir was transferred to the US Department

of the Interior, Bureau of Reclamation, for operation and maintenance. The Folsom Project is composed of a concrete gravity dam, two adja- cent earth-fill wing dams, the earth-fill MIAD, and

eight earth-fill saddle dikes. Additional aspects of MIAD, documented in detail by U.S. Army Engi- neer District, Sacramento (1953 and 1955), are presented below.

MIAD is a zoned embankment dam

4,820 ft (1,469.1 m) long and 165 ft (50.3 m) high

from core trench to crest at the maximum section.

The shells were constructed of dredged tailings

gravels. The narrow, central impervious core is a well-compacted clayey mixture founded directly on rock over the entire length of the dam to provide a positive seepage cutoff. Two transition zones, each 12 ft (3.7 m) wide, flank both the

upstream and downstream sides of the core. The

transition zones in contact with the core are composed of well-compacted decomposed

granite which classifies as a silty sand according

to the Unified Soils Classification System (USCS).

The second (outer) transition zones are construct- ed of the fraction of the dredged tailings passing

the 2-in. (5.08 cm) sieve. Typical sections of the

dam are shown in Figure 2. MIAD was constructed across the Blue

Ravine, an ancient channel of the American River

that is about 1 mile (1.61 km) wide at the dam

site. For about 1 ,650 ft (502.9 m) of its width, the

Blue Ravine is filled with auriferous, gravelly

alluvium of Pleistocene age. The maximum

thickness of the channel gravels is about 65 ft (19.8 m). The gravels have been dredged for

their gold content in the deepest portion of the

channel, and the tailings were placed back into the partially water-filled channel. The replacement

process tended to deposit the tailings in a very

loose and random condition. The remaining

undisturbed alluvium is crudely stratified and

slightly cemented.

POST-EARTHQUAKE SLOPE STABILITY

Background

Slope stability studies were performed

using the computer program UTEXAS2 (Edris and

Wright, 1987). This program may be used to determine limit equilibrium factors of safety

against sliding along surfaces prescribed by the user within the soil profile of interest, using one or more of several computational schemes available

to the geotechnical engineering community. The

scheme selected for this re-analysis calculated

factors of safety along circular slip surfaces using

Spencer's Method (Spencer, 1967).

Material properties used In UTEXAS2 were

derived from the various sources of field and

laboratory measurements including pre-construction investigations, construction testing, and

recent studies including crosshole and downhole

seismic measurements, in-situ density measurements, cyclic triaxial testing, and Becker Hammer

Penetration Testing (BPT).

Earthquake-induced shear stresses were computed in the Phase I study using the computer program FLUSH (Lysmer et ai. 1973). Zones

in the foundation where low factors of safety

against liquefaction (high excess pore pressures) were predicted were enlarged slightly to account

for lower cyclic strengths indicated by lower (N,

values. Strengths were assigned corresponding

to residual strength of liquefied soil, S^,, for the

purposes of post-earthquake stability analysis.

The relationship of residual strength to (Ni)6o

values used in this re-analysis is shown in Fig- ure 3. Residual excess pore pressures (expressed by means of a residual excess pore

pressure ratio, F\) induced by earthquake shaking

in saturated zones with factors of safety against

liquefaction (FS^) greater than unity were determined and represented in the manner described

by Hynes-Griffin et al. (1988) and Wahl et al. (1988). Average F\, values of 0.25 and 0.50 were

used in the non-liquefied portions of saturated

embankment and foundation gravels, respectively.

These values were selected on the basis of upward adjustments to values reported by HynesGriffin et al. (1988) to account- for tower cyclic

strengths indicated by the lower (Ni)6o values

discussed previously. The effect of these earth- quake-induced excess pore pressures was modeled as causing a reduction in available frictional strength (by means of a reduced effective angle

of internal friction, <t>') for input to UTEXAS2. An

artificially reduced <{>' was calculated for nonliquefied materials using the following expression

(after Hynes-Griffin, 1987):

sin(j)'F^^, = (1 -F\,)sin<|>'p.o (1)

The zonation of materials and strengths input to UTEXAS2 for slope stability analyses under design

pool and earthquake conditions is shown in Figure 4.
Determination of Foundation Zones Critical to Post-Earthquake Stability

The earthquake acceleration necessary to

trigger liquefaction in the free field gravels (average ((N, )eo = 7.1 ) was estimated for several earth- quake magnitudes for level ground conditions

using the simplified procedure (Seed and Idriss,

1982). Accelerations estimated by this procedure

range from 0.08 g to 0.10 g for magnitudes

ranging from 6.5 to 5.25, respectively, but it was

found that liquefaction in the free field {{t^)eo = 7.1) may not control threshold stability. A

residual excess pore pressure ratio, F\, of 0.40 was estimated for zones of foundation gravel with

average (N^)6o = 13.7. Vasquez-Herrera and

Dobry (1989) found that F\ < 0.40 generally

defines stable excess pore pressures in loose, fine sands; this value corresponds to FSj^ of about 1 .2

in the foundation gravels at f^lAD. Residual excess pore pressure ratios in the remaining

saturated, non-liquefied gravels were assigned

values between 0.10 and 0.40 for the purpose of evaluating stability with liquefaction restricted to

free-field gravels. Excess pore pressure ratios were estimated in the remaining saturated, nonliquefied zones. Two of the several upstream potential

post-earthquake failure surfaces evaluated for the case of liquefaction only in free-field gravels (average ((N^ )6o =7.1) were found to produce a factor

of safety against post-earthquake sliding of 1 .05. The factor of safety calculated using UTEXAS2

and material properties associated with the less severe shaking for this particular surface is 1.73. The results of these efforts indicate that

liquefaction of the foundation materials with

average (N^)6o - 13.7 controls stability with

respect to damage to the core of the dam.

Determination of Unsafe Combinations of Acceleration and Magnitude

The principal finding of this re-analysis of post-earthquake slope stability was that deep

failure surfaces would not likely develop to the

extent of intersecting the core of MIAD unless the zones of dredge tailings in the upstream foundation having an average (N, )go = 13.7 were caused

to develop residual excess pore pressures about

equal to the total overburden stress in situ (that is, and i\ =1.0, or expressed as 100 percent). It then remained to determine the level of earth- quake shaking necessary to induce 100 percent

pore pressure response in the (N^)6o = 13.7 zones. One-dimensional dynamic response

analyses were performed for an idealized column

through the upstream slope at the location of the

thickest part of these materials (i.e., through the

break point between 3:1 and 4.5:1 slopes) using

the computer program SHAKE (Schnabel, Lysmer,

and Seed 1972) to determine approximate rela- tionships between rock outcrop acceleration and

induced shear stresses. The use of SHAKE for

this case Is expected to yield fairly good, slightly

conservative estimates of cyclic shear stresses in the foundation soils. These estimates are somewhat higher than cyclic shear stresses that would

result using FLUSH. Stresses calculated using a

one-dimensional analysis in the embankment are

less accurate for the two-dimensional condition.

Cyclic shear strength profiles were developed as a function of (N, and multiplied by factors to correct for the effects of initial static shear stress on horizontal planes, , initial effective confining

stress, (both of these stresses were obtained

using FEADAM (Wahl et al. 1988)), and the number of significant cycles based on earthquake

magnitude K^ using the following relationship

(after Seed and Idriss, 1982):

KoM = (N.)eo/90x 1S.X K^x 1^ (2)

Figure 5 shows profiles of cyclic strength

developed for earthquakes of magnitude 5.25 to 6.5 for the SHAKE idealized column of materials,

along with profiles of average cyclic shear stress induced by the design earthquake acceleration

history scaled to various (rock outcrop) levels. A

shear modulus proportionality constant (K2)^a^ = 30 was assigned to foundation- gravels (dredge

tailings) to generate the stress profiles. Cyclic

shear stresses induced by the design earthquake

(predicted using SHAKE) exceed the available

cyclic shear strength throughout the foundation

gravels in Figure 5 by about 20 percent. It is thus

estimated from SHAKE stresses that an increase

in cyclic shear strength of about 20 percent would

be necessary to avert extensive liquefaction in the

gravels with (Ni)6o = 13.7. Figure 6 gives similar relationships for development of 40 percent residual excess pore pressure in the dredge tailings zone with (N^ )go = 13.7. A relationship between peak horizontal

acceleration, a^^^ , and earthquake magnitude was derived to differentiate between potentially

liquefiable and non-liquefiable (unsafe and safe,

respectively) conditions for dredge tailings foun249
dation gravels at MIAD. This relationship, which

is specific to a value of (Nilgo = 13.7 for foundation dredge tailings gravels, is shown in Figure 7. Correspondence between earthquake magnitude

and number of equivalent cycles as depicted in the figure is adapted form Seed and Idriss (1982).

For the stresses and strength conditions examined in this study, it was determined that rock

outcrop accelerations near MIAD of 0.25 g from a magnitude 6.5 event or 0.37 g from a magnitude

5.25 event would be required to initiate liquefaction of sufficient extent in the dredge tailings

foundation gravels to develop critical failure surfaces passing through the core and potentially

leading to loss of the reservoir.

SEISMIC HAZARD ANALYSIS

The objective of a seismic hazard analysis

is to estimate the return period (inverse of hazard)

associated with the exceedance of a particular

ground motion, usually a^^^. The probability that a certain level of a site ground motion parameter

is exceeded in a specified time interval may be

calculated from the analysis results. In addition,

for this study, the hazard was estimated for combinations of a^^^ and magnitude causing

liquefaction of foundation gravels at MIAD.

California and western Nevada comprise

the most seismically-active region of the continental United States. In these areas, much emphasis

has been placed on recording earthquake-induced ground motions, locating epicenters of

earthquakes, deriving distinct source zones, and

deriving characteristic parameters, such as those

defining attenuation, for various source zones. These efforts greatly benefitted the study of seismic hazard assessment at MIAD.

Several capable faults exist in central

California and are summarized in Table 1. The

closest capable fault to the Folsom Project is the East Branch of Bear Mountains fault zone, which

has been found to be capable of generating a maximum magnitude 6.5 earthquake. The minimum distance between this fault zone and MIAD

is 13 km.

Earthquake Source Zones

Algermissen et al. (1982) proposed the use of numerous seismic source zones in the U.S.

to facilitate probabilistic assessment of maximum

acceleration in rock. Seismic source zones are

selected on the basis of several criteria, including

the existence of capable faults, the rate of previous seismic activity, and the controlling type of seismotectonic regime.

The number of seismic zones in California

(including near off-shore zones), northern Baja

California (Mexico), and western Nevada suggested by Algermissen et al. (1982) is about 60.

Initially, about 50 of these original source zones were considered for this study to determine which zones would significantly affect the seismic hazard assessment. On the basis of this sensitivity

analysis, nine of Algermissen's source zones were

found to significantly contribute to seismic risk at the site. Only these nine zones were used in the

remainder of the study. These nine seismic zones

are described below. The locations are shown in Figure 8, and the recurrence parameters are summarized in Table 2. A zone delineating the East Branch of the

Bear Mountains fault zone (Zone 10), not considered by Algermissen et al. (1982), was included as an option for this study because of its close

proximity to MIAD. The computer program RISK

(McGuire, 1976) was used for this study and does

not allow for true line sources so a thin area source (about 0.7 km wide) was used to repre- sent this fault zone. The location of this fault zone with respect to source zones, in plan, is shown in Figure 8.

Recurrence Parameters

Recurrence parameters used for this study were obtained from a national study by

Algermissen et al. (1982) and a site-specific study

by Tierra Engineering Consultants, Inc. (1983).

Algermissen et al. (1982) also presented contour

maps of seismic risk for the U.S. based on computer analysis using general, -regional - source

parameters. These contour maps were later updated to include the results of more sophisticated analysis and were published by the Federal

Emergency Management Agency (Building Seismic Safety Council, 1988). The value of lower

bound magnitude, rT\ , is chosen to represent a magnitude cutoff point below which earthquakes

are too small to be of engineering interest. A

value of n\ = 5.0 is generally recognized to be

reasonable for engineering studies and was used

for this study.

The annual rate of earthquake occurrence,

also called the annual activity rate, is a function of earthquake magnitude and may be determined for each source zone from the historic data by

dividing the number of earthquakes that exceed a

given magnitude by the length of the observation

250
period, n, in years. A complete recurrence rela- tion may be compiled by considering increments

of magnitude for the full range of magnitudes

possible for the source zone.

Historic reports of earthquakes in the U.S. date back to the fourteenth century (Chiburis,

1977). However, consistent reporting and

recording did not occur until the late 1800's and

instrument measurements did not begin until the

early 1900's, and then, were only sparsely locat- ed. In most cases of reported earthquake activity

in central California prior to 1911, only intensities were recorded. Therefore, earthquake magnitudes, M, must be estimated from values of Modified Mercalli (epicentral) Intensity, MMI, for a

majority of the records. The annual activity rates corresponding to a MIVll V event for ail seismic zones used in this study (except Zone 10) were

derived from Algermissen et al. (1982). The rate

of occurrence for Zone 10 (East Branch of the

Bear Mountains fault zone) was reported by Tierra

Engineering Consultants, Inc. (1983). The rates of occurrence for all seismic zones are listed in Table 2. The rate of return corresponding to n\ = 5.0

is a necessary input for RISK. This rate of return may be obtained by relating magnitude and

intensity scales of earthquake motions and using

published rates of return based on MMI. A correlation between MMI and magnitude, M, was

determined for this study using seismic data

(6809 sets of MMI and M) measured before 1 March 1985 from the western U.S. (west of 10^

W longitude), as contained in the Stanford University data base (Nielsen, 1982). The magnitude at the average plus one sample deviation for Intensi- ty V data is 4.1. Algermissen et al. (1982) used a similar

relation suggested by Gutenberg and Richter

(1942) to relate M and MMI:

M = 0.6 MMI + 1.3 (3)

A magnitude of 4.3 is obtained for an intensity of V using equation 3. The seismic risks calculated

for this study are lower than those calculated by

Algermissen et al. (1982) because a lower value of n\ was used, if all other input were equal.

The recurrence rate for earthquake magnitude

is assumed to be of the form (Gutenberg and

Richter, 1954):

log [N (M)] = a - b M , M < M„^, (4)

or

In [N (M)] = a - p M , M < rv^ax (5)

where

N (M) = average annual number of earthquakes with magnitudes

greater than or equal to M;

a, b, a = parameters determined from

activity rates for respective source zones.

The parameters p and b are related by:

P = b In 10 (6)

The values of p used in the seismic hazard

assessment for this study, obtained from

Algermissen et al. (1982) and Tierra Engineering

Consultants, Inc. (1983), are listed in Table 2. The

magnitude-recurrence relations for these 10 source zones are plotted in Figure 9.

Attenuation Relations

Relationships proposed by Joyner and Boore

(1 981 ) to calculate mean values of peak horizontal

acceleration were used for this study. An upper

bound value of acceleration of 0.60 g was applied

to these relationships for mean acceleration as suggested by Bolt (1989) for seismic hazard

studies. These relationships have been derived

from measured data with magnitudes greater than

5.0 in western North America, predominantly in the state of California. The equation proposed to

calculate the mean peak horizontal acceleration

is:

-1.02 + 0.249 M - log^o r - 0.00255 r (7)

where

a^3^ = peak horizontal acceleration

(g's) M = magnitude

r = (d" + 7.3")'" 5.0< M< 7.7

d = (surface) distance from source

to site.

251
The standard deviation of loao a^ax reported by

Joyner and Boore (1981) is 0.220 (0.507 for

Recent studies by seismologists indicate that thrust faults produce accelerations that are about

20 percent greater than accelerations produced

by other types of faulting at a given distance

(Bolt, 1989). Reviews of seismologic data for each of the source zones described in Table 2

indicate that thrust faults exist in Zones 2 and 3. In Zone 2, the thrust faults are considered to be

inactive and, therefore, were not considered

separately for purposes of this study. In Zone 3, however, active thrust faults exist. The modified

attenuation relation used for Zone 3 to account

for the influence of the type of faulting is:

-0.941 + 0.249 M - log,o r - 0.00255 r (8)

A standard deviation for \oq^q a^^^ of 0.220 was

assumed. The same cap of a^^^ = 0.60 g was

applied to Equation 8. Mean values of a^^^ from

thrust fault zones are not expected to exceed

0.60 g since the closest point of Zone 3 to MIAD

is about 40 km.

Seismic Hazard Calculations and Results

The computer program RISK was used to calculate the annual probabilities of exceedance

given the aforementioned input parameters. The

program RISK is based on the theory described

by Cornell (1968). A Poisson distribution of events is assumed with the assigned recurrence parameters over each seismic source. The

following assumptions are inherent with a Poisson

model:

a. The earthquakes are spatially independent.

b. The earthquakes are temporally independent.

c. The probability that two seismic events will take place at the same location at the same

time approaches zero.

The probabilities are computed using the combined contributions of sources and a numerical

integration method.

The overall study included the calibration of the input data with the results of nation-wide

seismic hazard calculations made by others and

calculations that followed the profession-wide

procedures to calculate the seismic hazard at any

given site. The set of calculations presented

herein were developed specifically for this study

and are reported herein. These hybrid calculations were developed to incorporate information

available on critical combinations of earthquake

magnitude and peak horizontal acceleration that cause liquefaction for the dredged foundation

gravels at f^lAD. This latter set of calculations were thought to provide the best values for decision making.

Results of traditional seismic hazard calculations indicate that not all seismic zones used for

this study contribute a significant amount to the

seismic hazard at WAD. Rather, only three

seismic source zones were determined to make a

contribution of 5 percent or greater to the total

risk for a^^^ = 0.25 g and two seismic source zones contribute 5 percent or greater for s^^^ =

0.37 g. These sources zones are: 7, 9, and 10 for = 0.25 g and 9 and 10 for = 0.37 g. The magnitude and percentage of contributions

for each zone are listed in Table 3. Zones 9 and

10 produce 87 and 94 percent, respectively, of the risk at these two levels of acceleration. The

East Branch of the Bear Mountains fault zone is the largest contributor to the total seismic risk. The results presented in Table 3 suggest that some of the source zones used for analysis did not contribute any hazard for these two levels of acceleration, including Zone 3 which utilized modified attenuation relationships specific to

thrust faulting.

The contributions of different magnitude

events were also examined for purposes of this study. The contributions of magnitude intervals of 0.5 were calculated for peak horizontal acceleration of 0.25 and 0.37 g and are presented in Table

4. Eighty-seven and 94 percent of the contributions for 0.245 and 0.37 g, respectively, come

from magnitudes between 5.0 and 6.5. Notice

that these numbers correspond to the contributions of Zones 9 and 10 (which have M^^^ of 6.1 and 6.5, respectively). Therefore, Zones 9 and 10

essentially make the significant contributions to seismic risk between magnitudes of 5.0 and 6.5; the remaining zones make contributions at greater

magnitudes. There is relatively little to no contri- bution for the magnitude range of 6.5 to 7.0 or from Zone 3. The probability of exceeding a given acceleration alone does not signify a specific risk to

liquefaction of the dredged foundation gravels.

The results of the seismic hazard and post-earth252
quake slope stability analyses were combined to estimate the risks for various conditions that will cause initiation of liquefaction of dredged foundation gravels and consequent failure of MIAD. The

results presented in Figures 7 and 10 were combined and are presented In Figure 1 1 and tabulat- ed in Table 5. The annual (total) risk associated

with critical combinations of acceleration and

magnitude leading to liquefaction of the dredged

foundation gravels is approximately 0.0009 (an

annual return period of about 1100 years) for magnitudes that could have an impact on the site (greater than 5.0). The greatest contributions to the seismic

hazard causing liquefaction at MIAD come from

critical combinations of peak horizontal acceleration for earthquake magnitude intervals greater

than 7.0 (corresponding peak accelerations less than 0.21 g). The sum of the probabilities of occurrence for magnitudes greater than 7.0 Is 6.5 X 10"". The sum for magnitudes less or equal to

6.5 is 2.5 X 10"". The seismic hazard for the

magnitude interval 7.0 to 7.5 comes from Zones 1 (San Andreas Fault zone), 4 (Hayward and

Calaveras Faults zone), and 7 (Genoa-Jack Valley

and West Walker River Faults zone). The seismic

hazard for the magnitude interval from 7.5 to 8.0 comes from Zones 1 and 4. The seismic hazard

for the magnitude interval from 8.0 to 8.5 comes

from Zone 1 . This finding differs from the conclusion drawn from the traditional seismic hazard

analysis.

The Poisson distribution may be used to

calculate the probability of exceedance, Pg, by:

Pe = 1 - e"^'" (9)

where:

t = time period of interest

T = return period (yrs)

Using Equation 9 and considering a return period

of 1100 years, the annual probability of liquefaction occurring is 0.09 percent.

SUMMARY AND CONCLUSIONS

Analyses of slope stability for post-earthquake

conditions and seismic hazard have been con- ducted to assist in making decisions regarding

remedial measures necessary over portions of MIAD. For the post-earthquake slope stability

analysis, accelerations from both the design

earthquake and lower magnitude events were

used. The seismic hazard analysis used various

combinations of a^^^ and earthquake magnitude

to assess the seismic hazard. Results of the two

analyses were combined to estimate the likelihood of liquefaction of the dredged foundation gravels

and consequent failure of MIAD.

Acceleration levels required to cause liquefaction only In free-field gravels with (N^)eo = 7.1 were estimated using the Seed-ldriss simplified

procedure (Seed and Idriss, 1982, and elsewhere)

to range from 0.08 to 0.10 g for magnitudes

ranging from 6.5 to 5.25, respectively. Onedimensional dynamic response analyses were

conducted to determine approximate levels of acceleration required to produce the critical conditions. A range of induced average shear stress- es and strength conditions were examined to determine critical combinations of peak acceleration and magnitude. For example, these studies

indicated that rock outcrop accelerations near MIAD of 0.25 g produced by a magnitude 6.5 event (8 cycles) or 0.37 g produced by a magnitude 5.25 event (2.5 cycles) would be required to

initiate liquefaction of sufficient extent in the

dredge tailings foundation gravels to develop

critical failure surfaces passing through the core and potentially leading to loss of the reservoir. The input data used for seismic hazard

calculations are primarily from Algermissen et al. (1982) and Joyner and Boore (1981). The computer program RISK by McGuire (1976) was also

used. Best estimates of seismic hazard at MIAD

from earthquakes with magnitude greater than 5.0 are shown in Figure 10. The annual return periods for the best-estimate relation, including the

contribution of all magnitudes, at peak horizontal

accelerations of 0.25 and 0.37 _g are 1,500 and

10,500 years, respectively. It was determined, in general, that over three-fourths of the total seismic

risk is generated from the Sierran block and the East Branch of the Bear Mountains fault zone and

that these same percentages of contribution apply

to magnitudes between 5.0 and 6.5. The overall risk of experiencing a critical acceleration and magnitude level causing liquefaction of dredged foundation gravels at the site was estimated to be 0.0009 or an annual return period of about 1,100 years. It was determined

that the seismic hazard for critical combinations

of peak horizontal acceleration and earthquake

magnitude originates from three source zones: San Andreas Fault zone, Hayward/Calaveras

Fault zone, and Genoa-Jack Valley/West Walker

River Fault zone.

253
The return periods calculated for this study

are considered to be lower bound values. Many

conservative estimates were made in the course

of this seismic hazard assessment due to the

uncertainties associated with many of the input

parameters. More sophisticated analyses would

require greater effort and time. However, it was

known at the Initiation of this study that the results would be applicable for only a short duration due

to the construction of remedial measures. (The

first phase of remedial construction was complete

in November 1990, nine months following completion of the seismic hazard study.)

ACKNOWLEDGEMENTS

The data presented and analyses described,

unless otherwise noted, were obtained from re- search conducted for the US Army Engineer District, Sacramento (SPK), by the US Army Engineer

Waterways Experiment Station and reported by

Sykora et al. (1990). Permission was granted by

the Chief of Engineers to publish this information.

The overall Project Manager was Mr. John

Nickell, of the US Army Engineer District, Sacramento. Messrs. Frank K. Chang, Ronald E. Wahl,

and Daniel Habeeb, and Dr. Ellis Krinitzsky, WES,

provided assistance throughout the study.

REFERENCES

Algermissen, S. T., Perkins, D. M., Thenhaus, P. C, Hanson, S. L, and Bender, B. L, 1982.

"Probabilistic Estimates of Maximum Acceleration

and Velocity in Rock in the Contiguous United

States," Open File Report 82-1033, US Department of the Interior, Geological Survey, Denver,

CO.

Bolt, B. A. 1989. "Draft Seismological Report on

Seismic Hazard Assessment for Mormon Island Study," Report to US Army Engineer District,

Sacramento, Dec. 15.

Building Seismic Safety Council 1988. "Recommended Provisions for the Development of Seismic Regulations for New Buildings," FEMA Reports 95, 96, and 97, (3 Vols.), Federal Emergency Management Agency, National Earthquake

Hazards Reduction Program, Washington, D.C.

Chiburis, F. 1978. "Earthquakes of the Northeastern U.S. - 1534 through 1977," Preliminary un- published report, (cited in Nielsen, 1982).

Cornell. C. A. 1968. "Engineering Seismic Risk

Analysis," Bulletin of the Seismological Society of America. Vol 58, No. 5, pp 1503-1606.

Edris, E. V., Jr. and Wright, S. G. 1987. "User's Guide: UTEXAS2 Slope Stability Package," Technical Report GL-87-1. Vol. 1, US Army Engineer

Waterways Experiment Station, Vicksburg, MS.

Gutenberg, B., and Richter, C. F. 1942. "Earth- quake Magnitude, Intensity, Energy and Acceleration," Bulletin of the Seismological Society of America. Vol 32, pp 163-191.

Gutenberg, B., and Richter, C. F. 1954. Seismicity

of the Earth. 2nd ed., Princeton University Press,

Princeton, NJ.

Hynes-Griffin, M. E. 1987. "Seismic Stability

Evaluation of Folsom Dam and Reservoir Project,

Report 1: Summary Report," Technical Report

GL-87-1 4, US Army Engineer Waterways Experiment Station, Vicksburg, MS.

Hynes-Griffin, M. E., Wahl, R. E., Donaghe, R. T., and Tsuchida, T. 1988. "Seismic Stability Evaluation of Folsom Dam and Reservoir Project, Report

4: Mormon Island Auxiliary Dam - Phase I," Technical Report GL-87-1 4, US Army Engineer Waterways Experiment Station, Vicksburg, MS.

Joyner, W. B. and Boore, D. M. 1981. "Peak

Horizontal Acceleration and Velocity from StrongMotion Records Including Records from the 1979

Imperial Valley, California, Earthquake," Bulletin of the Seismological Society of America. Vol 71 , No.

6, pp 2011-2038.

Lysmer, J., Udaka, T., Tsai, C. F., and Seed, H. B. 1973. "FLUSH: A Computer Program for Approximate 3-D Analysis of Soil-Structure Interaction

Problems," Report No. EERC 75-30, Earthquake

Engineering Research Center, University of California, Berkeley, CA.

McGuire, R. K. 1976. "FORTRAN Computer

Program for Seismic Risk Analysis," Open File Report 76-67, US Department of the Interior,

Geological Survey, Denver, CO.

Nielsen, R. J. 1982. "Stanford Data Base for Earthquakes in the United States," Report 58,

Blume Earthquake Engineering Center, Stanford

University, Palo Alto, CA.

254
Schnabel, P. B., Lysmer, J., and Seed, H. B. 1972.

"SHAKE, A Computer Program for Earthquake Re- sponse Analysis of Horizontally Layered Sites,"

Report EERC 72-12, Earthquake Engineering Research Center, University of California, Berkeley,

CA.

Seed, H. B., and Idriss, I. M. 1982. Ground Motions and Soil Liquefaction During Earthquakes. Volume 5 of a series titled: Engineering Monographs on Earthquake Criteria, Structural Design,

and Strong Motion Records, Earthquake Engineering Research Institute, Berkeley, California.

Spencer, E. 1967. "A Method of Analysis of the

Stability of Embankments Assuming Parallel Inter- Slice Forces," Geotechnique, Vol 17, No. 1, pp

11-26.

Vasquez-Herrera, A. and Dobry, R. 1989. "Re- evaluation of the Lower San Fernando Dam,

Report 3: The Behavior of Undrained Contractive

Sand and Its Effect on Seismic Liquefaction Flow

Failures of Earth Structures," Contract Report GL89-2, US Army Engineer Waterways Experiment

Station, Vicksburg, MS.

Wahl. R. E., Crawforth. S. G.. Hynes, M. E., Comes. G. D., and Yule, D. E. 1988. "Seismic

Stability Evaluation of Folsom Dam and Reservoir

Project, Report 8: Mormon Island Auxiliary Dam - Phase II." Technical Report GL-87-14, US Army

Engineer Waterways Experiment Station,

Vicksburg, MS.

Sykora, D., Koester, J., Hynes, M., Fea, T., Gewerth, J., Jorgensen, K., McCune, R., and Treat,

R. 1990. "Estimation of Seismic Hazard and

Slope Stability Assessment of Mormon Island

Auxiliary Dam of Folsom Dam and Reservoir

Project" U.S. Army Engineer Waterways Experiment Station, Draft report to U.S. Army Engineer,

Sacramento District, Vicksburg, MS, February

1990.

Tierra Engineering Consultants, Inc. 1983.

"Geologic and Seismologic Investigations of the

Folsom, California Area," Report to US Army

Engineer District, Sacramento, Sacramento, CA.

US Army Engineer District, Sacramento 1953.

"Foundation Report, American River, California,

Mormon Island Auxiliary Dam, Folsom Project."

Sacramento, CA.

1955. "Foundation Report, American

River, California, Part A: General," Sacramento,

CA.

US Department of Interior, Bureau of Reclamation

1977a. "Earthquake Evaluation Studies of the

Auburn Dam Area," prepared by Woodward-Clyde

Consultants, 8 Vols.

1977b. "Seismic Evaluation of Auburn

Damsite: Supplement to Project Geology Report,"

Vol 20F6.

1978. "New Melones Dam Project,

California: Geologic and Seismologic Investigations."

255
Table 1 Estimated Seismic Characteristics of Capable Faults' (modified from Tierra Engineering Consultants. Inc.. 1983)

Name of Fault Zone

Minimum

Distance

To Site (km)

Type of Faulting

Maximum

Earthquake

Magni tude^

Approximate

Slip Rate

(cm/vr)^

Most Recent

D i splacement

Known^ San Andreas 170 Strike-sl ip 8.5 1 - 2 Historic

Hayward 140 Strike-sl ip 7.9 0.5 Historic

Calaveras 130 Strike-sl ip 7 0.25 Historic

Genoa Jack Valley 117+ Normal -si ip 7.25 0.01-0.02 Holocene

West Walker

River 142 Normal -slip 7.25 0.01 Historic

Me I ones 27.5 Normal -slip 6.5 0.0006-

0.0001 Pleistocene

+100,000

East Branch

Bear Mtns. 13 Normal -si ip 6.5 0.0006-

0.0001 Pleistocene

+100,000

1 A capable fault was defined as one that exhibited displacement at or near the ground surface within the past

35,000 years, recurrent movement within the past 500,000 years, exhibits creep movement, and/or exhibits aligned

macro-seismicity (M < 3.5) determined from instrunents.

2 Estimated by Tierra Engineering Consultants, Inc. (1983) based on various studies including: US Department

of the Interior, Bureau of Reclamation (1977a, 1977b, and 1978), and others. Modified slightly using data from Algermissen et al. (1982).

3 Slip rates estimated from historic, geomorphic, or geologic evidence.

4 Late Pleistocene displacement may be as old as 500,000 years ago or as young as 10,000 years ago.

Table 2 Description of Seismic Source Zones Used for this Study

Recurrence Parameters Source Code Algermissen et al. (1982)

This

Study

Algermissen

et al. (1982) Name Rate

at MMI V 5

1 C24 San Andreas Fault zone 1.97 0.99 8.3*

2 C39 Northern Coast Ranges 0.358 1.0 ~7.9

3 025 Western Great Valley 0.168 1.5 6.1

A C38 Hayward & Calaveras Faults 0.827 1.2 7.9

5 024 Elk & Kettleman Hills 0.247 1.5 6.1

6 030 Northern Sierra Nevada 0.588 1.5 7.3

7 029 Genoa- Jack Valley and West Walker River Faults

1.32 1.5 7.3

8 027 Southern Sierra Nevada 0.111 1.5 5.5

9 026 Sierran block 0.477 1.5 6.1

10 East Branch, Bear Mtns. fault zone 0.851** 1.96 6.5

* Adjusted from 8.5 as per Bolt (1989).

** Rates reported in terms of magnitude.

256
Table 3 Contributions of Seismic Zones to Seismic H3Z3rd

at Mormon Island Auxi I iary Dam

0.25 9 0 .37 q Contribution Contribution

Annual. Rate to Tn1"Al Ha7arH Annua

Rate iUtal naZarQ

Source (events/yr) (percent) (events/yr) (percent)

1 1.12 X 10-5"' 2 5.45 X

10"'

1 5.78 X 10"* 1 3.14 X

10"'

1.02 X

10-* 4.64 X io-« 2.15 X

10-^ 3 1.33 X

10"*

1 5 4.0 X 10-'

6 2.84 X

10"* 1.53 X

10"'

7 4.50 X

10"* 7 3.62 X

10"* 4 8 6.89 X 10"' 2.76 X 10"*

9 2.77 X 10"* 42 4.77 X

10"=^ 50

10 2.91 X

10"* 45 4.14 X

10-^ 44

100 100

Table 4 Distribution of Seismic Hazard Among Earthquake Magnitudes

Range in Contribution to Total Hazard (percent)

Magnitude 0.25 g 0.37 g

5.0 - 5.5 12 4

5.5 - 6.0 44 44

6.0 - 6.5 31 46

6.5 - 7.0 0 , 0

7.0 - 7.5 6 3

7.5 - 8.0 4 . 2

8.0 - 8.25 3 . . 1

100 100

257
Table 5 Risks Calculated for Various Combinations of Magnitude artd Peak Acceleration

Causing Liquefaction of Dredged Foundation Gravels. (N ,)^^

Magni tude Range Mean MagnitudeNumber of Equiv. Cycles Hazard Return

Period(yrs)

5.0 - 5.5 5.25 2.5 0.37 0.000003 333,000

5.5 - 6.0 5.75 A.

5

0.31 0.000094 10,600

6.0 - 6.5 6.25 6.75 0.26 0.00015 6,670

6.5 - 7.0 6.75 10.0 0.22

7.0 - 7.5 7.25 13.5 0.20 0.00012 8,300

7.5 - 8.0 7.75 17.5 0.17 0.00021 4,760

8.0 - 8.25 23 0.16 0.00031 3,020

Combined: 0.0009 1,100

258
Figure 1 . Location of Folsom Dam and Reservoir Project

259

261
O1

<«-( <«-( o o o o o

D. c 1 O

O o ^£) vO O. CO o 1 .—<

»-< o 1 <J 1 1 1 1 1 n o 1 u o «

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3: c c c c

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4) <v <u o 4J

a a;

u

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-o0

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inu

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uua

4J

O

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CO

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f-< 0 3 X 1 B <u

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4) -r-(I 0

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cr i-( JZ «

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u

^1

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<0 00c u

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i-i 3 c 0 CO

-r-( 3 03 •0 41

4) -e CX HO

fn 1 I I I I ac

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D)waD

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C«c

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-a-

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262
500 1000 1500

20

40

60

80

100

1 rDynomic sheor stresses computed

using SHAKE85, upstreom

mid-slope column. Mormon Islond Dom

Volues stiown ore 0-65(mox.)

trni^onkment

Foundotion

Cyclic Strength

(K2)„„ = 30

in foundotion grovels

0.15q 0.25g 0.35g ATM.=0.1g 0.2g o,3g (1 pgf - 47.88 Ps)

J L J I ] L.

20

40

- 60

80

500 1000 1500 OrNAMIC SHEAR STRESS, psf

100

Figure 5. Profiles of average shear stress for upstream, mid-slope column with

(K2)max 30 in foundation gravels, showing cyclic strength profiles for earthquakes ranging in magnitude from 5.25 to 6.5

1000 1500 200

20

40

60

80

100

120

Dynomic sheor stresses computed

using SHAKE85, upslreom

mid-slope column. Mormon Islond Dom

Volues shown ore 0 65(mox.)

Embonkment —

Foundotion

i R„ = 405e in foundotion grovels

with (t^Oeo = 13.7

(K2U. = 30

in foundotion grovels

0.1 5q 0.25g 0:35g ATM.=0.1g 0.2g o.3g (1

_i I I _i L

psf - 47.88 Pa)

J I 1 L.

20

40

80

100

500 1 000 1 500

DYNAMIC SHEAR STRESS, psf 200WO

Figure 6. Profiles of average shear stress for upstream, mid-slope column with

Climax = 30 in foundation gravels, showing cyclic stress profiles

required to produce f\ = 40 percent with (N^)eo = 13.7 for earthquakes

ranging in magnitude from 5.25 to 6.5

263
0.50

0.00

Number of Equivalent Cycles

2.5 4 8 20

5 6 7 8

Earthquake Magnitude, M

Figure 7. Estimated combinations of a„3, and earthquake magnitude causing

liquefaction of dredged foundation gravels for (N,)go = 13.7

264
265
Magnitude (^)

Figure 9. Recurrence relations for seismic source zones used in this study

Peak Ground Accelerotion (q's)

Figure 10. Best estimate of seismic hazard versus peak horizontal acceleration

recommended for use in selecting remedial measures 266
0.50

0.45

0.40

c

o

o

c

o

N

0.20

0.00

Number of

2.5 4

En ijivQ lent

8

Cyc 1 es

20

- 1 1

' (J) '

' C\l

1 ^ 1 ^ 1

-

' O ' (7) 1

1 O 1 O 1 1 C\] 1 1

1 o 1 O 1 o 1 1 o 1 o 1 O 1

, O , O o , o 1 o ' O ' O ' o ' o ' o ' o ' O

'

' O ' O

' o ' o ' o 1 o 1 ^ 1

0.35 -

o 0.30

u 0.25 h

<

o 0.15 -

S 0.10

Q_

0.05

SAFE

(N,) 60 13.7

5 6 7 8

Earthquake Magnitude, M

Figure 11. Probabilities of occurrence for different ranges in magnitude at combinations of a^^^ and earthiquake magnitude causing liquefaction

of dredged foundation gravel for (N,)go = 13.7

267

System Identification of Spirally Reinforced Circular Bridge

Columns Subjected to Cyclic Lateral Loading

William C. Stone^

and

Andrew W. Taylor^

Abstract

Problems relating to present procedures for the design of bridge columns to withstand seismic loads are discussed and a more rational design procedure is proposed. Central to the proposed analysis method

is the requirement for an analytical capacity to predict the hysteretic behavior of reinforced concrete

bridge columns when subjected to inelastic lateral loads up to and including failure. A three-parameter

model with a nonsymmetric trilinear envelope was used in an initial effort to identify relationships between

physical column properties and hysteretic model parameters. A digital database of load-displacement

histories was established for circular, spirally reinforced bridge column tests. A system identification analysis was performed on each specimen in which the error in predicted cyclic absorbed energy was

minimized. Three parameters, comprising a stiffness degrading coefficient, a strength degrading

coefficient, and a pinching coefficient, were employed in this study. Subsequent multivariable least squares regression analyses were conducted to correlate the variation of the three failure model

parameters with the specimens physical variables as well as cummulative absorbed energy. Conclusions

are presented regarding the sensitivity of hysteretic behaviour to certain physical characteristics of

circular, spirally reinforced bridge colums.

Introduction

Research on the design of bridge structures to resist seismic loadings has been driven by the inadequate

performance of constructed facilities under actual earthquakes. The catastrophic collapse of the 1-880 Cypress Freeway, and the significant damage sustained by other elevated highway structures in the

Oakland-San Francisco region during the Loma Prieta earthquake of 1 989 (Lew, 1 990) has given rise to a research project at the National Institute of Standards and Technology (NIST) to develop a rational

procedure for the design of ductile bridge columns. This paper describes the fundamental tenets comprising this concept and the results of preliminary work on one phase of the project, that of developing a suitable hysteretic model for the prediction of inelastic dynamic response of ductile bridge

columns.

Background

Before discussing what constitutes a "rational" design procedure, it may prove useful to summarize the

current generally accepted design procedure for single, cantilevered bridge columns employed by the

California Department of Transportation (Caltrans, 1984), as this represents a more stringent version of. the AASHTO (AASHTO, 1978) procedures presently in effect throughout the United States for the design

of bridge structures. The Caltrans procedure, presented in Table 1 , requires the selection of a maximum

expected bedrock acceleration at the bridge site from a seismicity contour map; the selection of a

Research Structural Engineer, National Instrtute of Standards and Technology, Gaithersburg, Maryland

Research Structural Engineer, National Institute of Standards and Technology, Gaithersburg, Maryland

269
response spectrum based on generic soil conditions; and the calculation of the anticipated fundamental

period of the structure. The designer then reads a static equivalent lateral design force from the response

spectrum and proceeds along the design steps outlined in Table 1 . This procedure, while lending Itself to a codified standard, suffers from several drawbacks:

• A single peak acceleration value for a regional area ignores important localized phenomena

associated with distances from active faults, variations in earthquake magnitudes and their effects on frequency content, and attentuation and frequency shifting with distance from the active faults.

• A sample of representative response spectra for homogeneous soil conditions for a few depth ranges

will not necessarily yield a realistic measure of the lateral surface excitation force as experienced by

the structure, particularly if the frequency content of the earthquake rock motion is other than

assumed in the generation of the generic design charts and if there are multi-layered sedimentary

deposits.

• There is no basis for the prediction of the post-earthquake damage status for the structure. Thus,

the design codes for lifeline structures presently include arbitrary modifiers to the static equivalent

lateral design load based on the level of 'importance' of a particular structure. Because of this,

there is no ability to perform an effective optimization study which would result in the least-cost

design solution to meet a particular survival requirement.

A Rational Design Procedure for Bridge Columns in Seismic Regions

A more rational approach, described graphically as a computer program flow chart in Fig. 1 , to the design

of bridge columns is proposed. This approach is based on recent advances in synthetic earthquake

generation techniques and the development of reliable models for the prediction of hysteretic behavior

in reinforced concrete structures. Because the concept requires the integration of both seismological and

structural analysis components, the input data upon which the analyses are based are site specific.

Presently there Is no national data base from which one might, for example, obtain a set of representative

acceleration-time histories for a particular site which might then be used for design. For this reason, it was decided that a demonstration project, using the San Francisco-Oakland Bay area as the example,

would be used to limit the scope of the required data base generation. Similarly, because there are many

possible variations to superstructure architecture, we have limited the initial scope of this project to the

performance and design of circular, spirally reinforced, single cantilevered columns which are

monolithically cast with continuous longitudinal reinforcement through the potential plastic hinge regions.

These criteria are applicable to a broad number of elevated highway systems presently in use in the State

of California.

From a fundamental standpoint there are seven unique, and interdependent, aspects relating to the

design of a bridge column that is required to perform in a certain manner when subjected to certain levels

of seismic excitation. These are as follows:

• Superstructure and Performance Definition • Subsurface (Soil Column) Definition • Bedrock Motion Definition • Site (Surface) Motion Calculation • Hysteretic Failure Model Selection • Inelastic Dynamic Analysis • Cost Optimization

In considering the first of these, Superstructure and Performance Definition, the engineer would prefer to enter as few pieces of information into a computer program as necessary. In the case of geometry and

270
loads, the minimum data required are the height of the column, L, and the effective axial load, P^, from

the roadway. In addition the engineer would also like to specify the level of acx^eptable damage following an earthquake of a given magnitude. It may be appreciated that a bridge column which is required to be totally functional, with no damage at all, following a Magnitude 6.5 earthquake, may in fact demand

a more robust design than the same column which would be required to remain marginally standing (and

require subsequent demolition) following a Magnitude 7.5 earthquake. Since these criteria will depend

on site-specific characteristics, the engineer would also want to provide the computer program with a

geographical location (longitude and latitude) of the construction site, from which, ultimately, design

loadings would be automaticaliy generated. Obviously, this design approach represents a departure from

current practices. Its implementation is predicated upon a number of developments taking place and on

the integration of what are presently disparate research software into an automated system.

The Issue of structural input loadings is addressed in the second through fourth aspects listed above. The

seismic excitation function (acceleration-time history) is dependent upon the following parameters, among

others:

1 . The magnitude of the earthquake.

2. The local tectonics which affect the generation and propagation of seismic waves.

3. The distance from the epicenter to the structure.

4. The underlying soil conditions which exist between the foundation of the structure and bedrock.

Given these parameters the development of an appropriate 'design' earthquake loading is a probabalistic

process which is presently the subject of considerable research. We envision, ultimately, a national

database which would permit the user, by specifying a geographic location for the construction site, to automatically develop a best estimate for the soil column profile from surface to bedrock level; and to automatically generate a suite of design bedrock acceleration-time histories, at varying magnitudes and

frequency contents which would be representative of that site. These two pieces of data, combined with

existing wave propagation software (e.g. SHAKE, 1990), can be further used to automatically generate

surface-level acceleration-time histories which would serve as the input loading for the structural design.

Initially, it is clear that the sparsity of available boring logs to bedrock will relegate such automated soil selection procedures to preliminary analyses, pending availability of specific construction site logs.

However, with required filing of boring logs to bedrock at all construction sites in seismic regions with a

national organization, such as the United States Geological Survey, the completeness of such a soils database would gradually improve to the point where a refined analysis would become unecessary.

Procedures are currently available (e.g. Joyner and Moore, 1988) for the development of synthetic

earthquake records at a given site, provided information on local geology and tectonics are available. For

any given site, and any given structure, it may be appreciated that a single acceleration-time history, just as with a single equivalent static lateral load, is insufficient for proper analysis. Because some structures may respond more dramatically to a particular segment of a response spectrum, it is important to develop,

for each site, a suite of acceleration-time histories which represent, in addition to the entire response

spectrum, segments of it. Likewise, because different magnitude earthquakes may initiate from different sources, a separate suite of design acceleration-time histories for each magnitude under consideration

must be generated. Given these data, one may then develop a suite of surface-level time histories

suitable for subsequent structural analysis. An interactive graphics-driven system which employs the

above principals is presently being developed for the Oakland-San Francisco Bay area and will be the

subject of a forthcoming paper.

Hysteretic Failure Models

The remainder of the proposed design procedure rests upon the ability to analytically predict the dynamic

response of the target structure, not just to its elastic limits, but to complete failure. Several computer

271
programs presently exist including SAKE (Otani, 1974), SARCF-II (Rodriguez-Gomez, Chung, and Meyer,

1990), and IDARC (Park, Reinhorn, and Kunnath, 1987), among others, which have, to varying degrees,

achieved some measure of success in achieving this goal. All are 'research codes" and to date are

limited to the analysis of two dimensional structural systems. At the heart of these programs are analytical

models to idealize the inelastic behavior of reinforced concrete under dynamic or quasi-static loading (El- Borgi, White, and Gergely, 1991). These are generally accomplished with the use of macromodeling

schemes in which inelastic action is represented through hysteretic models describing force-deformation

characteristics at a cross-section. Such models attempt to capture overall behavior of the structural element based on load-displacement predictive algorithms which are calibrated to experimental data.

These latter models, generally referred to as hysteretic failure models, were selected for the project

described in this paper because of their ability to be calibrated to experimental data and because the

output directly lends itself to the computation of absorbed hysteretic energy and displacement ductility,

both of which have been shown to be correlated to a quantifiable measure of the state of damage

sustained by the structure following a seismic event (Park, Ang, and Wen, 1984).

Since the mid-1 960s research has progressed towards the development of two general hysteretic failure models which can be classified as either "smooth" or "multi-linear," depending on the mathematics of the

algorithm employed. State-of-the-art summaries of the various methods can be found in El-Borgi, White

and Gergely (1991) and in Park, Reinhorn and Kunnath (1987). In the present study a multi-linear

hysteretic model, known as a "Three Parameter Model," based on improvements to the Clough-Takeda

Model (Takeda, 1970) by Takayanagi (Takayanagi and Schnobrich, 1977) was employed. This same

algorithm is presently used in the program IDARC (Park, Reinhorn, and Kunnath, 1987).

It is common practice to describe the envelope curve of the force-deformation relation of reinforced

concrete components by a multi-linear function with three turning points, e.g., cracking, yield, and ultimate

strength. The tri-linear approximation to the envelope curve is known as the "skeleton" curve. Procedures

have been developed to extract this curve from an equivalent monotonic load-displacement envelope

curve that has been fitted to experimental data for columns tested under reversed cyclic loading (Park,

Reinhorn, and Kunnath, 1987). A variety of hysteretic properties can be obtained through the combination

of the tri-linear skeleton curve and three characteristic parameters, a, (3 and 7, as will be subsequently

described.

The parameter a controls the degradation (softening) in the unloading stiffness that is generally observed

in reinforced concrete members as they degrade under reversed cyclic loading. In the three-parameter

model, stiffness degradation is introduced by setting a common point on the extrapolated skeleton curve.

It is assumed that the unloading lines aim at this point until they reach the displacement-axis, as shown

in Fig. 2. The parameter a directly affects the slope of the unloading branches of the load-deflection

history and, as can be seen in Fig. 2, indirectly affects the area bounded by the hysteresis loops, which

is a measure of the seismic energy absorbed by the structure during any given cycle. As a decreases so does the unloading stiffness.

The parameter 0 determines the rate of strength degradation, as shown in Fig. 3., and is defined as the

ratio of the incremental damage sustained by the structural element caused by the increase of the

maximum response, dS^S^, to the normalized incremental hysteretic energy, dE/(8^jPy), as:

m

= ^

(1)

dE dE

y

272
where:

6y = Ultimate deformation under monotonic loading; generally defined as the displacement on the

descending monotonic failure curve at which point the lateral load has decreased to 80% of the

maximum lateral load. P = Yield strength, calculated from the envelope curve. dt = Incremental hysteretic absorbed energy during the current cycle as shown in Fig. 3. 6^ = Maximum observed displacement during the present cyclic loading history.

As can be seen from Fig. 3, a value of /S equal to zero will produce hysteresis loops with no strength

degradation, no matter how many subsequent cycles are carried out, provided all other parameters

remain the same. Increasing, non-zero values of 0 produce increasingly rapid decay in the maximum load on subsequent cycles.

Pinching behavior, such as illustrated in Fig. 4, may be introduced by lowering the target maximum point

(labeled as the Initial Target Point in Fig. 4) to a level of ^Py (labeled as the New Target Point in Fig. 4)

along the previous unloading line. A new stiffness is calculated such that the reloading path aims at this new target point until the crack closing point is reached. The stiffness of the reloading path is changed

again at this point such that subsequent increases in deformation produce increases in load along a

steeper (stiffer) slope which ultimately passes through the initial target point. The introduction of such

a pinching parameter leads to a reduction in the cyclic energy absorbed by the structural element.

These three parameters, together with the skeleton curve information described above, comprise the

variables required for implemention of the hysteretic rule described in Park, Reinhorn, and Kunnath (1 987).

El-Borgi, White, and Gergely (1991) warn that the proper implementation of this type of model requires

calibration of the variables to a specific type of structural element if reliable results are to be obtained.

It was for this reason that the initial project scope was restricted to circular, spirally reinforced, single

bridge columns. Our first task was thus directed to assembly of a suitable digital database of tests of

spirally reinforced, circular concrete bridge columns subjected to reverse cyclic loading with combined

axial load. The process of determining the relationship between the hysteretic parameters a, 0, and y

and the column geometry and materials is known as system identification. The ability to implement the

optimized inelastic dynamic structural analysis described in Fig. 1 is predicated on the ability to reduce

such a relationship to a set of closed form equations.

A Proposed Standard for Digital Test Records

An extensive literature search identified several sources of bridge column test data. Digitized records for 8 column tests were available from Stone and Check (1989) and Check and Stone (1990); 12 additional

records were obtained from Ang, Priestley, and Paulay (1989, 1985, 1981); Mander (1984); Munro, Park,

and Priestley (1976); Ng, Priestley, and Park (1978); Priestley and Park (1984); and Zahn (1986). As

most of these latter tests were originally conducted using analog recorders a digitizing program was

developed at NIST to produce digital load-displacement records from high resolution photographic

enlargements from the technical papers. Additional usable data was identified in Lim, McLean, and

Henley (1990); McLean and Lim (1990); Petrovski and Ristic (1984); and Wong, Paulay, and Priestley

(1990) but this latter data became available too late for inclusion in the analyses reported in this paper.

A summary of the available tests and their properties are reported in Table 2.

273
One aspect concerning experimental data merits discussion at this point, and that is the matter of 'noise',

Introduced into the load-deflection record either by spurious electronic signals or by mechanical sources such as leaking servo-hydraulic loading apparatus. Inevitably all experimental data is laced with noise

to some degree. Furthermore, there exists the issue of how many data points are necessary to effectively

capture a single hysteresis loop. For the purposes of analysis, it is desireable to: a) reduce the

experimental record to as few digitized points as necessary to accurately capture the load-displacement

record; b) eliminate to the best possible degree the influence of noise; c) represent quarter and half cycle

hysteresis curves as smooth, continuously differentiable mathematical functions and d) have within each

cycle a one data point with precisely zero displacement at the start of each loading cycle. All of these

have one ultimate purpose: to reduce computer processing time when performing a system identification analysis, in which several tens of thousands of iterations through the entire data set may be required to

arrive at a best fit for the particular parametric model being used. These criteria were achieved in the

present study by fitting the initial quarter load cycle, and all subsequent half load cycles with n**^ order

polynomial curves. The t>est fit for the first quarter cycle in almost all cases was achieved through an

equation of the form:

Pi(5) = CqB^'^ + 0^5=^^ -I- CgS^^ + CgS^^ + 0^5^^ + CgS^^^ (2)

The best fit for all subsequent half cycles was generally achieved through an equation of the form:

= Cq + C^d + C^^ + Cgfi^ + C^b^ + Cgfi^ -I- Cg6® (3)

The atx)ve fitting was done using a special purpose interactive graphics program, NI DENT 3.0, which was

developed on a high performance workstation as part of the present project at NIST. Using NIDENT 3.0,

the researcher may experiment with different polynomial fits and visually assess the accuracy of the fit. An Initial density of 25 points per half cycle was selected for subsequent regeneration of the loaddeflection hysteresis record using the polynomial regression equation appropriate to each half cycle.

Records processed in this manner were then used for system identification analyses as will be described

below.

In the process of assembling the digital records of the 'rav/" data for each column test it became apparent

that a specific format would have to be established so that each test could be handled by automated

system identification routines being developed at NIST. As there is presently no standard protocol for such data records we propose the format described in Table 4. The substantial problems encountered

in developing the digital database used for this paper indicate that it is particularly important that a

standard digital record format be defined soon and used throughout the international structural research

community in the future. Such a standard, combined with increasingly effective inter-computer

communications (e.g. FTP, InterNet, ArpaNet, etc.), would make a much broader range of experimental

data available for the refinement of hysteretic failure models for all types of structural components. We

feel that such analysis methods represent the vanguard of future seismic-resistant design and as such

merit the cooperation of international research laboratories.

System Identification

The term "System Identification' is generally associated with the experimental acquisition of such dynamic

characteristics of a structure as its mode shapes, frequencies, and damping coefficients. In this paper

we use the term to refer to the determination of the three parameters a, P, and y, which best characterize

the hysteretic behavior of a given test specimen. The system identification procedure used in this study

consists of a three dimensional trial and error search with bounded limits on a, 0, and y subject to the

constraint of minimizing the error between predicted and experimentally observed differential energy

absorption. In equation form, we seek to minimize the function F, where F is defined as follows:

274
E 0.5{S,., - d,.,Xi'^(0 - P^(Of " (4)

where:

displacement of next point in load-displacement record

displacement of previous point in load-displacement record

experimentally observed load at current point

predicted load at current displacement, calculated using the multi-linear hysteretic rule and

the currently selected values of a, 0, and y

number of data points in the experimental record

In mathematical terms, the above analysis constitutes an optimization problem. Where smooth,

continuously differentiable (to second order) equations can be used to describe the hysteretic rule, such

as are presently employed in the so-called "smooth" hysteresis models (see El-Borgi, White, and Gergeley,

1991, for example). It is possible to make use of procedures such as the Levenberg-Marquardt method

(More, 1977), among others. Unfortunately, attempts to integrate a Levenberg-Marquardt solver into NIDENT 3.0 (the NIST graphics-based system identification package) failed with the Three-Parameter

model. The reasons for this stem from the fact that the hysteretic rule (see Park, Reinhorn, and Kunnath,

1 987 for a detailed presentation) for the Three Parameter model is presently comprised of 1 5 discrete cases which govern such situations as loading, unloading, re-loading, and pinching. This results in the

presence of step-like facets to the function describing the relationship between a, 0, and 7. Optimization

routines such as the Levenberg-Marquardt method base their "next best guess" at the optimum solution

by taking exceedingly small steps to obtain estimates of the local first and second derivatives of the

function to be optimized. For the Three-Parameter model, this generally results in several iterations until a particular facet is reached. There, considerable variation in a given parameter may not produce any

significant change in the predicted load. Once such a situation is encountered most optimization solvers

terminate, indicating that they have found the 'optimum' solution, even though the fit, as visible in a

graphics display, is poor. A solution to this problem was developed for NIDENT 3.0. This consisted of employing the coarse-mesh three dimensional trial and error search for an initial estimate of the

parameters a, 13, and 7. These values were then used to calculate the predicted load-displacement record

which was subsequently superimposed graphically, in a different color, on top of the experimental record.

Using a high performance graphics workstation it was then possible to interactively adjust any of the three

model parameters by means of a dial-box while viewing the results in real time on the graphics screen. This visual-feedback approach proved surprisingly effective in arriving at a "best fit". As a final check, the

absorbed energy was calculated for the experimental data model estimates for each cycle and compared

in a histogram plot. When the error between the theoretical and experimental sums of the cyclic absorbed

energies was reduced to a level of a few percent, the final values of a, /3, and 7 were recorded.

Examples of the fit between the analytical model and experimental data are shown in Figs 5-9. Several

details of these figures bear discussion. The first, as evidenced, for example, in Fig. 8, is that as the

column begins to fail, there is a tendency for the predicted loads to overshoot the experimentally observed

load at the maximum observed displacements for each cycle. This is an expected consequence of using a tri-linear skeleton curve where the post-yielding stiffness for such a model is defined to be positive for numeric stability. Secondly, as evidenced in the associated cyclic absorbed energy histograms (see Figs.

5-9), there is a tendency for the predicted absorbed energy to overestimate the experimental data for low

values of lateral displacement and to underestimate the experimental data for high displacement ductility.

Despite these shortcomings, the Three Parameter model is able to generally predict both overall absorbed

Pexp(0

p (i) =

275
cyclic energy and displacement ductility to within a few percent error, which was considered acceptable

for the puposes of the present project.

Multi-variable Stepwise Regression Analysis

Given best-fit values of the parameters a, /3, and 7 for many different column tests, the next objective is to ascertain if there is any correlation between these parameters and the physical properties and

dimensions of the test specimens which were modelled. For this objective a stepwise linear regression

analysis was conducted. Of the 20 digital test records listed in Table 2 all but three were able to be

estimated to within 2% error in total absorbed energy using the techniques described above. Two

specimens represented re-tests of previously loaded specimens and were therefore considered suspect

and the third, for reasons yet unknown, proved intractable to accurate solution using the Three-Parameter

model. These three specimens were therefore eliminated from the analyses. In addition to the basic

specimen properties listed in Table 2, a large series of variable permutations, shown in Table 3, were

considered in the regression analyses. The regression analyses were conducted using the commercial

software package SAS/STAT (SAS/STAT, 1987) and the results are presented in Table 4. (square of the regression correlation coefficient) values of 91%, 86%, and 97% were obtained for the estimates of the parameters a, /?, and 7, respectively. The regression equations associated with these parameters

are given by:

a = 0.832x10^ + 0.116x10-V^ - O.SOSxIO^A

y (5)

+ 0.430x1 0-«E^ - 0.665x1 0-'ifij, + 0.572^-^jo.6''*

P

= 0.189 + --^(13.14/^ - 4.457) + ^[-^][4.051 - 30.826--^

- 0.001715^ + O.IOIy^-^j + (0.356p, - i.soejco.g)"'

(6)

Y

= 0.845 + 0.00309- + 0.0704-^f-?

- 1.879-^ + (0.0826 - 0.748x10%)/?;

(7)

The above relations should not be construed as a reliable basis for seismic design in their present form,

given the limited size of the database upon which they were developed. However, from a research

standpoint, several useful observations can be drawn which may guide further development of hysteretic

failure models.

Stiffness degradation, as defined in Fig. 2, is heightened by lower values of a. It is therefore retarded by

high values of spiral reinforcement yield stress, f ; by high concrete strength to spiral reinforcement yield

strength, fyfyg! by increased initial column flexibility, as inferred by the column aspect ratio, L/D; and by

Increased amounts of confining reinforcement, p^. It Is of particular note that the total absorbed energy,

E^Q^, enters In this equation. All of the other variables can be derived from material properties and

276
geometry. In a computational analysis it is possible to calculate the cummulative absorbed energy at each point in the time-history. This opens some rather interesting opportunities which may solve the

problem of overestimation of cyclic absorbed energy at low displacement ductilities and the

underestimation of cyclic absorbed energy at high displacement ductilities. It may be possible, with

further research on an expanded test database, to develop a correlation between cummulative cyclic

absorbed energy and all the parameters defining a hysteretic model. If such a correlation exists, rt may

then be possible to answer one of the fundamental questions faced by structural researchers: the

quantification of the effects of cyclic load history on the behavior of structures.

Of the three parameters defined above, /3 (strength degradation) appears the least subject to precise

definition. Although a relatively high correlation coefficient was achieved, this was at the expense of a

significantly higher number of participating terms in the regression equation. This was not altogether

unexpected, for Park, Ang, and Wen (1984) had already reported in the derivation of an energy-based

failure criterion for reinforced concrete that there existed significant scatter in the data used to define when

failure" actually occurred. The term 0 is the most closely correlated to the state of damage, as previously

described, and we therefore anticipated significant scatter. Still, a number of useful observations can be

made. Since increased values of 0 correspond to more rapid failure of the column, it may be concluded

from a sensitivity analysis of the above equation that higher axial stress levels lead to more rapid failure of a column, as does a lower aspect ratio. This would serve to reiterate the conventional wisdom that

highly stressed short columns are at greatest risk in earthquakes. It is also interesting to note that for longitudinal reinforcement ratios of less than about 4%, increasing the confining steel reinforcement serves

to improve column performance (see final term in 0 equation).

The third parameter, y, influences the pinching behavior in the hysteresis loops. A smaller value of 7

leads to a more pronounced pinch and, therefore, less energy absorption during any given cycle.

Desirable performance would dictate the maintainence of fat' hysteresis loops. Although a regression

equation is presented for 7, we caution against drawing any behavioral information from it. Only one of the specimens presented in Table 2 exhibited any observable pinching behavior (Ng No. 3, shown in Fig.

7); all others had 7 values greater than or equal to 1.0. It may be appreciated from Fig. 4 that values of 7 in excess of 1 .0 result in a re-targeting of the initial target point on the unloading curve; that the results should therefore be insensitive to changes in 7 beyond 1.0 were verified by means of interactive variation

of the parameter using the program NIDENT. It is of some note, however, that all but one specimen

exhibited a lack of pinching behavior through failure. Since all of these test specimens were designed

in accordance with current Caltrans or New Zealand 3101 criteria, it would appear that these provisions

provide for adequate confining reinforcement to prevent significant pinching behavior.

Conclusions

Within the bounds of the available test data for circular, spirally reinforced concrete bridge columns

subjected to cyclic lateral loading, it was found that the Three Parameter model was able to generally

predict the observed experimental load-deflection histories, and to produce an estimate of the

cummulative absorbed cyclic energy within an error bound of several percent. Preliminary regression

equations were developed for the parameters a, 0, and 7 and explanations were offered to explain the

relationship between these parameters and the material and geometric properties of the columns. With

only one exception, all columns were able to be accurately modeled with a 7 value of 1.0, indicating that

the design guidelines utilized for the design of the individual specimens were sufficient to prevent

significant pinching hysteretic behavior through failure of the columns. The accuracy of the regression

equations can be substantially enhanced through expansion of the available database. To this end a

standard format for reporting digital records for tests of reinforced concrete columns subjected to cyclic

lateral loading was proposed.

277
Table 1: Caltrans Single Column Design Procedure

1 . Design the main longitudinal reinforcement for biaxial bending based on AASHTO group loads l-VI and Caltrans seismic loads for group load VII.

2. Determine the probable plastic moment of the column.

3. Design lateral reinforcement for the controlling group load or the lesser of the Caltrans group load VII unreduced earthquake force or the plastic moment, whichever controls.

4. Confinement and shear reinforcement are designed using the AASHTO Code and modifications as

specified by Caltrans:

a) Confinement is covered in (8.18.2.2) as modified by Caltrans Code in the plastic hinge zones. Two

formulas are used based on the size of the column. In all other locations the confinement equation

is (8-62) as defined in the AASHTO Code.

b) Shear strength (8.16.6.1.1) equations (8-46) and (8-47) as defined in the AASHTO Code. Concrete

shear strength (8.16.6.11.4(a) is modified by Caltrans in the area of plastic hinging.

c) Reinforcement shear strength (8.16.6.1 1.4(b)) conforms to AASHTO Code equations (8-53) and (8- 63).

5. Footing design equations conform to the AASHTO Code. Footings are designed for AASHTO group

loads l-VII and Caltrans seismic loads for group load VII.

278
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Regression

Variable OS vm hnl J

111 U Vll C 1 1111 llUll

X22 Ps_

fya

X23 Ps

f f lya I c X24 Etot

Measured total absorbed energy

X25 Measured lateral yield load

X26 Measured lateral yield disp.

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X28 "maxDuctility ratio

X29

Etot

Py5y

Normalized absorbed

energy

X30 Vf^Ag —

X31 Vf^AglLj

X32

Pe /Dn

f'c Ag lI

X33 Inverse of aspect ratio

X34 —

X35

X36 Ps /D\ —

X37 Ps —

X38 lys c X39 - X48 Where K = 0.1, 0.2, 0.3,...

0.7, 0.8, 0.84, 0.9

X49 - X58 Where K = 0.1, 0.2,

0.3,... 0.7, 0.8, 0.84, 0.9

X59 - X68 Where K = 0.1, 0.2, O.3.... 0.7, 0.8, 0.84, 0.9

Table 3 (Part II). Regression variables

280
Variable Symbol Definition

XI L^oncreie coiripressive suengui

X2 fya LXjngiiuaiiiai reiniorceiiieni

yield stress X3 fys Confining reinforcement yield

stress X4 D Column diameter

X5 L

Column height or

effective length

X6 Ps

Confining reinforcement ratio

X7 Pa

Longitudinal reinforcement ratio

X8 Pe Axial load

X9

A Gross cross section area XIO

Pe Normalized axial load

Xll Ac Effective concrete area

X12 Ac —

X13

£c

^ya

X14

Aci fyaj

X15 Ag/fc^ Pe > —

Ac (fya^ [fc AgJ

X16 L

D

Column aspect ratio

X17

fys

X18 Ps

Pa —

X19 —

X20

fya L

X21

fys

^ya

Table 3 (Part I). Regression variables.

281
Table 4: (Page 1 of 2)

Proposed Standard for Digital Test Records of Reinforced Concrete

Columns Subjected to Cyclic Loading with combined Axial Load

Field FORTRAN Description

Number Formal

Administrative:

1 1X,A10 Year of Test

2 1X,A80 Test Specimen Name

3 1X,A80 Laboratory Name

4 1X,A80 Laboratory Street Address

5 1X,A80 Laboratory City.State, Country, Zip

6 1X,A80 Laboratory FTP Digital Communications Address (if available)

7 1X,A80 Researchers Name(s)

8 1 X, A80 Title of Most Appropriate Report/Paper

9 1X,A80 Reserved for future data/aux. notes

10 1X,A80 Reserved for future data/aux. notes

Specimen Properties:

11 1X,E12.6 f'g Concrete Compressive Strength (MPa)

12 1X,E12.6 f Longitudinal reinforcement yield strength (MPa)

13 1X,E12.6 Confining reinforcement yield strength (MPa)

14 1X,E12.6 D Column diameter, if Circular (mm)

15 1X,E12.6 H Column depth, in loading direction, if Rectangular (mm)

16 1X,E12.6 W Column width, if Rectangular (mm)

17 1X,E12.6 c Concrete cover to centerline of confining steel (mm)

18 1X,E12.6 L Effective column length (mm)

19 1X,E12.6 Confining steel reinforcement ratio (%)

20 1X,E12.6 s Confining steel pitch or spacing (mm)

21 1X,E12.6 ds Confining steel bar diameter (mm)

22 1X,E12.6 T Confining steel type: Spiral; Hoop; or Rectangular Ties.

23 1X,E12.6 Longitudinal steel reinforcement ratio (%)

24 1X,E12.6 dl Longitudinal steel reinforcement bar diameter (mm)

25 1X,E12.6 N Number of longitudinal bars

26 1 X,A80 Q Longitudinal splice description (if any)

27 1X,E12.6 Pg Axial load (kN)

28 1X,E12.6 Experimentally measured cracking displacement (mm)

29 1X,E12.6 Pg^ Experimentally measured cracking load (kN)

30 1X,E12.6 Experimentally measured yield displacement (mm)

31 1X,E12.6 Py Experimentally measured yield load (kN)

32 1X,A80 Rate of Loading (include units)

33 1X,A80 P-Delta effects included in load-displacement data?

(yes/no)

34 1X,E12.6 . reserved for future data -

' 35 1X,E12.6 reserved for future data

36 1X,E12.6 reserved for future data

37 1X,E12.6 reserved for future data

38 1X,E12.6 reserved for future data

39 1X,E12.6 reserved for future data

40 1X,E12.6 reserved for future data

282
Table 4: (Page 2 of 2)

Proposed Standard for Digital Test Records of Reinforced Concrete

Columns Subjected to Cyclic Loading with combined Axial Load

Field FOR"mAN

Number Format

Description

Loading History:

41 1X,I5

42

43

44

45

1X,E12.6

1X,E12.6

1X,E12.6

1X,E12.6

no

max,1

^min.l

s;''max,2

^111111,2

Number of cycles in test loading history, including any

used in the initial determination of the yield displacement

Maximum (positive) displacement for cycle 1 Minimum (negative) displacement for cycle 1 Maximum (positive) displacement for cycle 2

Minimum (negative) displacement for cycle 2

41+nc

42+ nc

1X,E12.6

1X.E12.6

Load-Displacement Record:

43+nc 15 44+nc 1X,2(E12.6,3X)

45+ nc 1X,2(E12.6,3X)

'max.nc

'min.nc

npts

Maximum (positive) displacement for cycle "nc" Minimum (negative) displacement for cycle "nc"

Number of data points in the load-displacement record

Lateral displacement (mm); Lateral Load (kN) pair Lateral displacement (mm); Lateral Load (kN) pair

43+nc 1X,2(E12.6,3X) ^npts'^npts Lateral displacement (mm); Lateral Load (kN)

+npts pair at final sequential point in the record.

283
Table 5: Regression Analysis Coefficients

Regression Analysis for Variable Alpha

R-square = 0.91085245

Parameter Standard

Variable Estimate Error

Intercept: 8.31876014 4.89794429

X3 0.01160500 0.00834998

XI 7 -50.93576455 24.16746952

X24 0.00000043 0.00000020

X26 -0.06646607 0.02255637

X54 0.57236815 0.32174906

Regression Analysis for Variable Beta

R-square = 0.86416021

Parameter Standard

Variable Estimate Error

Intercept: 0.18964653 0.35159550

XI

0

-4.45792174 2.02667807

X14 4.05133212 2.06712680

XI

5

-30.82632478 12.65111662

X26 -0.00170923 0.00062060

X30 13.14743204 5.32542453

X34 0.10056988 0.02926639

X48 -1.30608709 0.57406484

X68 0.35640403 0.11675052

Regression Analysis for Variable Gamma

R-square = 0.97634278

Parameter Standard

Variable Estimate Error

Intercept: 0.84494717 0.01515021

X16 0.00309135 0.00072887

X17 -1.87851685 0.11806923

X19 0.08262993 0.00615182

X36 0.07035537 0.01795630

X38 -0.00007482 0.00000591

284
QrapNci Prt-n-oOMtor

AnalyaliTyp*

1 : Cdwnn h

2: Aim Uad. P

1:CalunDH«<^h

2:AxWUiad. P

3: Coneret* Str., fc

4: SM yMd, Fy

5: Column Ola., 0

6: MmI r«., rhoa

7. oarl. rit.rtw*

I Sduton -~

Conrtvotlen mi bvdreekT No

SatSell ProdtoUmjH CMnllon

Soi

n*cordi Datttm* KM)

Contov

(Sin Pnndtco)

1.QrvMe«Plci<o<

Sl» kxation ttam Map

1.1 Liy«f 0^ (1) 1.2Layw(1)PnipM«M

N.I Layw (N) D«pth

N.2 Layw (N) PrupwHat

Daptii (N) Propartat (N) Oapt» (N) PropanlM (N) ZT

Mrock Motion

Dtraa

Auto

1: SiM bcason pick tom dlgrtzad Sar Francisco map OalarmlnM daaign auKa

01 1(1) hMori** tar a ipadfc

magrrtKda aarthqiMtta

• w

11 • (1) tn1

».

In «

mn m « o( MagnMixlaa n - ntrrAar o( hi*t canp/lting siita

a. Updated afrpiricil modal! ralotng ratponaa •p*c*a Id Magrituda &

datoxa tx>n aourca

(Caiilomia tactonics)

b. Updatad amplflcal modalt

tar duration at a Kncton ol Magnitjda and diatnaca torn aouroa (Calitomia

tactonica) c Whara avaUbIa, Mlad S

•calaba<»Dd(

racorda mstcNi'ig crttarla a) and b) tor apacifc San Frxdaco aitaa.

d Mwa raccrdt ara not availabla atachaatcally ganarata arHdal aarTvquafcaa aa

raqtirad.

Uaartuppliad accalaraHon tma Natory. a(l) m • 1 n • 1

Yaa

Soi Column ftopagaton

Coda: QanarMa auiaca

•or*. 8HAKE2.0

FEWCodaa

Hyataratlo Fallura lledai

UaarData

3-Parant«taf

Polynomial

Srangth Oagradaton

Saflnau Dagradation Hnehing

Baaad on

ragrataion tt to aiparlmartal bddga colifm data. Cydlc RUa only.

I^PolynomiilJ Y Is-Paramatar

Salact: 1:a^

2:bata

3: gamma

Beaad on ra^aaalon

Ittoaxparimanlal

bridga column data

a: Riad Paramalara

b: CycHcRula

NIDENT1.0

hilaracfva

Uannncalon Coda

SoMkm

Flag -1,2

SoluHon Flag.l

Solution Rag .2 3:

Initid Ectimatat: fe, D, ly. ftioa, rtwa, h, P, m, n

Magriluda

©

I SuWa NuTfcar [Do J.I.N I y^optlmlzad

Stora: Coat, fe, fy, D, itioa, rhoa. m. n

|

IDARC 2.2 Matllc Dynamic

SVucturU Analytii Coda

Naw Eiln iataa: rc, D, ty, rtioa, rhoa

Dana(I) E(m.n,l<) Etot (m.n)

Dl (m,n) Daltamax (m.n) Mmo (m.n)

Optimization

Oplmlzalion Coda:

Minlmlza: Conatuetlon Coat C • F(rc.lyAD.rhaa.moa^

Subject to tha eonttaint: Dl> >1.i2. >3.>4 • ifiowabla damaga

indcaator Magnltuda(U)

Cani»airc

01 » F(rcfy, 0. rhoa. rtna) c

•5

9

(3

Li. Oa

2

Sto>. X

OUTPUT

Figure 1. Proposed rational design procedure for bridge substructures in seismic regions.

285
286
287
288
NIST Full Scale Flexure Specimen

2000

1500

1000

500

o 0

-500

-lUUU

-1500

-2000

Experimenial

Thewetical

No. Cycles 20

Alpha 7.5 Beta 0.062

Gamma 1.00 Energy Error -0.8 %

-600

^

-200 0 200 400

Lateral Displacement (mm)

NIST Full Scale Flexure Specimen

600

Experimental

0 Theoretical

N cycles p p

111

0.00 10° Ijl

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20

Cycle Number

Figure 5. Experimental and analytical results for NIST full scale flexure specimen.

289
NIST Model Specimen Nl

90 I I r

!

:

: :

I

: : i

-70 -60 -50 -40 -30 -20 -10 0 10 20 30 40 50 60 70

Lateral Displacement (mm)

NIST Model Specimen Nl

^

8.00 10^

1 2 3 4 5 6 7

Cycle Number

Figure 6. Experimental and analytical results for NIST model specimen NL
Ng '78 Specimen 3

Ng *78 Specimen 3

^

5.00 10^

I

1 2 3 4 5 6 7 8

Cycle Number

Figure 7. Experimental and analytical results for specimen 3 of Ng.

1
Ang *81 Specimen 1

200

150

100

50

U

00

-100

-150

-200

Experimental

ThcOTCtical

No. Cycles

Gamma

Energy Error

1.00 1.4 %

-60 -40 -20 0 20 40

Lateral Displacement (mm)

Ang *81 Specimen 1

60

1.60 10"

1.40 lO"

1.20 lO''

O' 1.00 10"

8.00 10^

o

ftn o An ia3

bp

g 6.00 10^

4.00 10^ o

^ 2.00 10^

0.00 10°

Experimental

Theoretical N cycles g _ g

Cumulative Error = ]^ (

" "P )

xlOO = -1.4% i= 1 ^exp

II

II

III

3 4 5 6

Cycle Number

Figure 8. Experimental and analytical results for specimen 1 of Ang.

292
Potangaroa *79 Specimen 1

1000

800

600

400

200

o 0

ca -200

Lai •400

-600

-800

-1000

Expaimental

TheOTetical

No. Cycles 8 Alpha 8.75

Beta 0.10

Gamma 1.00 Energy Error -0.04 %

-50 -40 -30 -20 -10 0 10 20 30 40 50

Lateral Displacement (mm)

Potangaroa '79 Specimen 1

0.00 10°

2 3 4 5 Cycle Number

Figure 9. Experimental and analytical results for specimen 1 of Potangaroa.

293

Simplified Procedure for Inspecting Earthquake Resistant

Potential of Highway Bridges Against Soil Liquefaction

by

Ken-lchl Tokida', JIro FukuP and Hideo Matsumoto^

SUMMARY

Reasonable and simplified procedures are necessary to Inspect the earthquake resistant

potential of existing highway bridges. Effects

due to soil liquefaction during earthquakes

are very important problem on the earthquake

resistance of bridge foundations and should

be considered for the inspection of bridges

adequately. Inspections on the earthquake

resistant potential of highway bridges have

been carried out In Japan since 1971, and the

Inspection procedure has been Improved step

by step.

The next inspection of earthquake resistant

potential for highway bridges will be con- ducted in Japan, in 1991. For this inspection,

the procedure to Inspect the liquefaction-induced damage potential of highway

bridges has been investigated to improve the

former procedure of 1986-inspection. The

procedure for the next Inspection will be

drafted In the near future, referring to the

results of the detail investigation shown in this paper.

In this paper, newly Investigated procedure

for inspecting the llquefactlon-lnduced damage potential of highway bridges Is presented.

KEY WORDS : Earthquake resistance of bridge

foundation. Site liquefaction potential, Liquefaction-Induced damage potential. Procedure for inspection

1. FOREWORD

Two kinds of procedures to reduce the

earthquake-Induced damage of highway bridges

can be considered; (1) the earthquake design

method for newly-constructing bridges; and

(2) the retrofit method for existing ones. The

earthquake design methods have been In- vestigated for a long time and applied to the

actual design of highway bridges in Japan [ 1, 2 ] . On the other hand, the inspections to check the earthquake resistance of existing

bridges also have been carried out since 1971

for retrofit of highway bridges.

Table-1 shows the history of procedures for

Inspecting the earthquake resistant potential

of highway bridges in Japan. As shown in this

table, soil liquefaction was not inspected directly at the first stage of Inspection in 1971

and 1976. However, liquefaction potential

of a ground was noted concretely for the first

time in the 1979-lnspection. Furthermore, in

the 1986-InspectIon, both liquefaction potential of a ground and its effects on

bridge foundations were considered. The

next inspection for the earthquake resistant

potential of highway bridges will be con- ducted in 1991. Based on results of the 1986-inspection and

lessons from the past earthquakes, the

following is necessary to be investigated for

establishing the reasonable procedure for the

coming Inspection.

1) Lack of data for Inspection:

For Inspecting the liquefaction potential of

a ground and the liquefaction-induced damage

potential of a bridge suitably, the data

related to both ground conditions and

structural conditions of highway bridges,

especially, foundation conditions should be

surveyed. However, because It is generally

difficult to get the data, especially on

borings and/or structures related to the

bridges constructed long ago, It should be

considered how to deal with the bridges

without suitable data for inspection. From

this point of view, simplified procedures are proposed for the coming inspection with use of

data which can be obtained easily in this

paper.

2) Reasonable estimation of both liquefaction

and its effects on structures:

Not only the occurrence of soil liquefaction but also the effects induced by

liquefaction should be considered. From this

point of view, two kinds of Indexes; (1) the

site liquefaction potential; (2) and the liquefaction-induced damage potential are proposed

to estimate the scale of liquefaction and

damage of highway bridges more quantitatively

than the 1986-inspectlon in this paper.

2. GROUND LIQUEFACTION POTENTIAL

2.1 Indexes of Ground Liquefaction Potential

As a first stage for Inspecting the earthquake resistant potential of highway bridges,

a ground liquefaction potential defined as the

probability and/or scale of occurrence of

IHead, Ground Vibration Division, Public Works

Research Institute, Ministry of Construction,

Japan.

2Senior Researcher, Foundation Division, ditto.

3Researcher. Ground Vibration Division, ditto.

295
soil liquefaction at the bridge site should be

estimated roughly and/or in detail according

to the data obtained. In this paper, indexes

for estimating the ground liquefaction potential are proposed as follows :

1) Ground Group and Topographical Group:

Because these indexes can be got easily

comparing with the boring data, they can be

applied for inspecting the ground liquefaction potential roughly [see Section 2.3] .

2) Liquefaction Resistance Factor (Fl), Modified Thickness of Liquefiable Layer (H*) and

Site Liquefaction Potential (Ipl):

However It needs to get boring data and

test results of soli properties to calculate

these Indexes, they can be applied for in- specting the ground liquefaction potential in detail [see Section 2.2] . Furthermore, the relation of site liquefaction potential between the objective

bridge site and the neighboring site of the

bridge is investigated in Section 2.4.

2.2 Procedure for Detail Inspection of Ground

Liquefaction Potential

Liquefaction potential as for sandy soil

layers can be examined based on the liquefaction resistance factor Fl defined by the

following equation applied In the

"Specification for Highway Bridges" [1, 2] .

Fu = R / L (1) where

Fu : Liquefaction resistance factor

R : Resistance of soil elements to djmamic

loads, and R = Ri +R2 +R3

Ri : Resistance based on relative density

R2 : Resistance based on Dco-values

R3 : Resistance based on fine contents

R, = 0.0882^/ N/(ct v'+0.7 ) r 0.19 (0.02mm gDso^ 0.05mm)

Rs =<. 0.

R3 =

_ r

.2251ogiD(0.35/DBo)

(0.05mm < Doo^ 0.6 mm)

0.05 (0.6 mm < Dsn ^ 2.0 mm)

0.0 ( 0 % g FCg 40 %)

Cz

Co

Ci

1 0.004FC-0.16 (40% < FCgl00%)

; Dynamic load to soli elements

L = r<i • ks • ( <T v/ (7 v'

)

rd = 1.0 - 0.015X

ks = CzCQCiko

: Blow count by standard penetration test

: Mean grain size (mm)

: Fine contents (%) ;Depth from the actual ground surface (m) ka: Seismic coefficient for evaluating liquefaction

Seismic zone factor

Ground condition factor

Importance factor

N

D50

FC

X

Kd : Standard seismic coefficient (= 0.15) a V : Total overburden pressure(kgf/cm^)

a v': Effective overburden pressure at the

static condition (kgf/cm^)

Soil layers with the liquefaction resistance

factor Fl smaller than 1.0 are judged to liquefy during earthquakes. However Flvalue is very suitable to estimate the liquefaction potential of a certain layer at the

objective depth, another index is necessary

to estimate the liquefaction potential of the

whole subsurface ground in stead of FL-value.

In this paper, the following Index Is proposed as the simplified parameter to represent

the ground liquefaction potential.

H*

Ipu = (2) 20 where

Ifl : Site liquefaction potential [^1.25]

H* : Modified thickness of liquefiable layers, and H* = H, * + Ha*

Hi* = 1.5Hfli + 1.0Hfl2 + 0.5Hfl3

Ha* = I.OHfli + 0.5Hfl2

Hi* : Modified thickness of liquefiable lay- er at the depth of 0 to 10 m

H2* : Modified thickness of liquefiable lay- er at the depth of 10 to 20 m

Hfli: Total thickness of layers

with Fl ^ 0.6 (m) Hfl2: " with 0.6 < Fl ^ 0.8 (m) Hfls: " with 0.8 < Fl g 1.0 (m)

The coefficients in the parameters Hi* and

H2* are decided for convenience sake, by re- ferring the relation In Table-2 [ 1 ] where

bearing capacities of liquefied ground are

either neglected or reduced by reduction

factors De determined In accordance with

FL-values.

The damages of bridges Induced In the past

earthquakes In Japan are Investigated In view

point of the effects of liquefaction and so on. Table-3 shows the typical examples of bridges

which can be estimated to be damaged mainly

by liquefaction. Site liquefaction potential

(Ifl) is calculated with use of boring data at the bridges as shown In Table-3. Furthermore, the damages of bridges related to liquefaction during earthquakes are classified

into five levels as shown in Table-4 in this

paper. For the bridges in Table-3, the

relation between the damage level of bridges

and the site llque- faction potential (Ifl) can

be drawn In Fig. 1. The standards on site

liquefaction potential can be classified Into three groups as follows, considering the

damage level of bridges as shown In Table-3.

C Ipl < 0.2] There were not any case of

damage related to girder fall Induced by

296
liquefaction. The scale of liquefaction and

Its effects on bridges can be assumed to be

small. (Estimation-1)

[ 0.2 ^ Ipu < 0.4] Tiiere were cases of damage

related to liquefaction. The effects by liquefaction on bridges can be assumed to exist.

(Estimation-2)

[ 0.4 g Ipu ] There were cases of girder

fall Induced by liquefaction. The scale of

liquefaction and its effects on bridges can be

assumed to be very severe. ••• (Estimation-3)

From the above results, it can be indicated

that the index of Ipu can be used as one of

indexes to estimate the ground liquefaction

potential at the bridge site. The Ifl values

of 0.2 and 0.4 can be applied as the boundary

to judge the level of ground liquefaction

potential, considering the damage level of

bridges.

2.3 Procedure for Macro-inspection of Ground

Liquefaction Potential

Both ground group and topographical group

are selected as the indexes for macroinspection of the ground liquefaction potential, and their effectiveness are Investigated

in this section.

2.3.1 Ground Group and Site Liquefaction Potential

The ground condition can be classified

roughly according to the stiffness and/or

amplification characteristics of ground at the

bridge site. In this paper, the ground conditions are assumed to be classified Into three

groups; (1) Ground Group I ; (2) Ground Group

n ; and (3) Ground Group m as shown In Table-5 which shows the classification of

ground groups defined in the "Specification

for Highway Bridge" [1, 2] . Table-6 shows the results investigated on

the relation between the ground group and

the site liquefaction potential (Ipl) at 217

sites of bridges. As shown in Table-6, the

sites included in Ground Group I can be

assumed as the ones with a low liquefaction

potential related to the damage of bridges,

considering the Estimation-1 of Section 2.2. In other words, the sites classified as Ground

Group n or m are in danger to have high

liquefaction potential related to damage of

bridges.

2.3.2 Topographical Group and Site Liquefaction Potential

It Is well Itnown that the topographical

condition is related to the probability of liquefaction occurrence. In this paper, topographical conditions are classified roughly

Into three groups, i.e.. A, B and C which

are proposed as Table-7 for inspecting the

earthquake resistant potential of highway

bridges.

Table-8 shows the results investigated on

the relation between the topographical group

and the site liquefaction potential (Ifl) at 332 sites of bridges. As shown In Table-8, the

site classified as Topographical Group A can

be assumed to have high liquefaction potential

related to the damage of bridges. In other

words, the sites classified as Ground Group B or C can be assumed to have low liquefaction

potential related to damage of bridges.

2.4 Liquefaction Potential at Neighboring

Sites

For dealing with the problem on the lactc of

boring data at the site of the objective

bridge, the relation of liquefaction potential

between the bridge site and Its neighboring

site Is Investigated with use of the index

defined by Eq. (3).

HT = Hi ** + Ha** (3) where

HT : Modified thickness of llquefiable layer

(m) [gl6(2/3)m]

H,** = I.OHfli + (2/3)Hfl2 + (1/3)Hfl3

Hs** = (2/3)Hfci + (1/3)Hfu2

Hi**: Modified thickness of llquefiable lay- er at the depth of 0 to 10 m

Hs**: Modified thickness of llquefiable lay- er at the depth of 10 to 20 m

Hfui, Hfl,2, Hfu3 : See Eq.(2)

The coefficients In the parameters Hi** and

Ha** are decided In the same way with the

parameters Hi* and Ha* shown In Eq. (2). For the above Investigation, 187 boring

data with llquefiable layers are selected

within a radius of 750 m around 10 bridges.

The modified thickness of llquefiable layer

(HT) of selected sites distributes as shown in Fig. 2. Five items which seem related to liquefaction and their categories are selected

to Investigate the relations between the

boring data of a bridge site and the ones at

its neighboring sites as follows.

(ltem-1) Difference of above the sea level.

Category: ~ -3m~ -Im— lm~ 3m~

(ltem-2) Distance.

Category: 0~100m~ 300m'- 500m~ 750m

(Item-3) Difference of ground water level.

Category: —3m~-lm~lm~3m

(ltem-4) Difference of ground group.

Category: -1, 0, 1, 2 (Item-5) Difference of topographical group.

Category: 0. 1. 2

297
Fig. 3 shows the distribution of data

according to the categories of each items. The

regression coefficient and correlation

between the difference of modified thickness ( A HT) and the above Items with each category

are calculated with use of type I quantification analysis, and the results are summarized In Tablc-9. Based on the results

of Table-9, however all of correlations are very low, the three correlations on the

differences of ground water level, topographical group and ground group are relatively high comparing with other items.

Furthermore, as for two items of difference of

ground water level and topographical group,

the difference of modified thickness ( A HT) is inclined to increase according to the increase

of them.

From the above Investigation, the difference of topographical group, difference

of ground group and difference of ground

water level seem to relate to the liquefaction

potential between two sites, relatively. The

guidelines to adopt the boring data at the

neighboring site for estimating site liquefaction potential (Ipl) of the objective

bridge can be proposed as shown in TabIe-10

for convenience sake.

3. LIQUEFACTION-INDUCED DAMAGE POTENTIAL OF

BRIDGE

3. 1 Indexes Related to Damage of Bridge by

Liquefaction

For surveying the indexes related to the

liquefaction induced damage of bridge foundations, the following five indexes are se- lected considering the results of Table-3.

Ij : Supported condition on pier-top.

If : Foundation type.

Ifs : Bearing condition of foundation.

Ifl : Site liquefaction potential.

Igc : Ground group.

Table-11 shows the categories of each index

mentioned above. With use of type I quantification analysis, five cases for

analyzing the effects of these Indexes to the

damage of bridges are conducted by combining

them as shown in Table-12. In this analysis,

the damage level (see Table-4) of bridge foundations is defined as a dependent variable and

the independent variables are combined with

the indexes as shown In Table-12.

Table-13 summarizes the results of cat- egory-scores and correlation-coefficients for

each case. From these results, it can be seen

that the three indexes, i.e.. If, Ifs- and Ihu can explain the damage level of bridge

foundations relatively well.

3.2 Simplified Method for Estimating Liquefaction-induced Damage Potential of Highway

Bridge

It should be noted that the damage patterns

of bridge foundations induced by soil

liquefaction is different according to the

foundation type as shown in Table-14. In other words, it is desirable to select the

suitable indexes to explain the cause of

damage induced by liquefaction. Based on

results in Section 3.1 and lessons from the

past earthquakes, three indexes, i.e., PI, Pn

and Pin are proposed in Eqs. (4-1), (4-2) and

(4-3), respectively for convenience sake to estimate the liquefaction-induced damage

potential of bridge foundations according to the foundation type (see Table-15). For es- timating these indexes quantitatively, the

weighted points are supposed roughly,

considering the scale of effects to damage of

bridges.

Abutment: P 1 = ( ic + Ij + If + Ifs ) X Ipl X Ihe (4-1)

Movable Pier: P n = ( Ic + If + Ifs ) X Ipl X Igc (4-2)

Fixed Pier: Pffl = ( If t Ifs ) X Ifl • (4-3)

where

Ic : Effect of device against fall of a girder

' Continued girder, Fixed girder : 0.0

• Both insurance of Se and

installation of stopper/connection : 0.5

• Insurance of Se or installation of

stopper/connection : 1.0

• Not Insured : 1.5 (Note)SE: Distance between the end of gird- er and the crest of a substructure.

Ij :Effect of girder-support condition on the

top of pier/abutment

• Movable bearing : 0.0

• Fixed bearing : 1.0

If : Effect of foundation-type

• Pile. Caisson : 1.0

• Wooden pile. Pile bent : 2.0 lFs:Effect of bearing condition at the bottom

of foundation

• Perfect support : 1.0

• Imperfect support : 5.0

• Shallower than liquefied layer : 6.0 lHB:Effect by flow of abutment-backfill

(Represented with Height of backfill:HB)

• Hb < 5.0 m : 1.0

• Hb ^ 5.0 ra : 1.2 lFL:Effect by liquefaction (Site liquefaction

potential : Ipl)

• 0.2 ^ Ifl < 0.4 : 1.0

• 0.4 g Ifl : 1.5 (Note)The site with Ipl < 0.2 is omitted

because of low liquefaction potential.

locrEffect by deformation of surrounding

ground (Represented with Ground group)

298
• Ground group n : 1.0

• Ground group m : 1.2

3.3 Liquefaction-Induced Damage Potential of

Highway Bridge In the Past Earthquakes

The simplified procedure for estimating liquefaction-induced damage potential of highway

bridge proposed in Section 3.2 are applied to the bridge foundations damaged in the past

earthquakes as shown in Table-3. Figs. 4(1),

4(2) and 4(3) show the relations between

liquefaction-induced damage potential and

damage level for abutment, movable type of

pier and fixed type of pier, respectively.

Based on these results, the boundary values

of indexes related to the severe damage of

bridges, i.e., with the damage level more than

4 can be estimated as follows.

Abutment: P I < 6.0 :Damage induced by liquefaction can be estimated to be small.

P I ^ 6.0 :Be in danger of severe damage.

Detail Investigation should be

necessary.

(Relation-4)

Movable Pier: P n < 6.0 : Damage Induced by liquefaction can be estimated to be small. P n ^ 6.0 : Be In danger of severe damage.

Detail Investigation should be

necessary.

(Relatlon-5)

Fixed Pier: P in < 4.5 :Damage Induced by liquefaction can be estimated to be small. P in ^ 4.5 :Be In danger of severe damage.

Detail investigation should be

necessary.

(Relation-6)

4. SIMPLIFIED PROCEDURE FOR ESTIMATING

EARTHQUAKE RESISTANT POTENTIAL OF HIGHWAY

BRIDGE AGAINST SOIL LIQUEFACTION

From the above Investigation and other

lessons In the past earthquakes, the simplified procedure for estimating earthquake

resistant potential of highway bridge against

soil liquefaction can be proposed such as

Fig. 5. As shown in this flow-chart, the

Inspection is classified Into four stages as

follows.

1) First-stage of Fundamental Inspection:

In this stage, before inspecting the site

liquefaction potential and the liquefactioninduced damage potential, the bridges in the

following conditions can be excluded.

(1) Bridges designed with use of the specifications after 1971 In which the effects of

liquefaction are considered for earthquake

resistant design of bridges.

(2) Bridges whose superstructure can be

considered as the types not effected by soil

liquefaction severely, for example. Deck

Arch Bridge, Rigid-frame Bridge and Culvert

Bridge.

2) Second-stage of Site Liquefaction Inspection:

In this stage, the site liquefaction potential is Inspected before Inspecting the

effects of liquefaction on bridges. The

inspection flow Is divided into two flows ac- cording to the existence of boring data at the

bridge site. When the boring data Is possible

to get, the site liquefaction potential can be

calculated with use of the index of Ipl proposed in Eq. (2). On the other hand, when the

data can not be got, the ground liquefaction

potential is estimated roughly with use of

either ground group or topographical group

of the bridge site based on the results In Section 2.3. Furthermore, for the bridge site judged as both Ground Groups n and EI and Topographical Groups A , the boring data at the

neighboring site of the bridge site which can

be selected by the standard proposed In Table-10 is possible to use to calculate the

site liquefaction potential (Ifl). When the

site liquefaction potential (Ipl.) Is less than

0.2, the effects of liquefaction is judged to be small by the RelatIon-1 shown in Section

2.1 and further inspection is not necessary.

When the soil data for the Index of Ifl can not be got at both bridge site and neighboring site, the Inspection should progress In the next stage.

(3) Third-stage of Damage Potential Inspection: In this stage, the liquefaction-induced

damage potential of P I , Pn and P m should

be calculated by Eqs. (4-1), (4-2) and (4-3) for

the abutment, movable pier and fixed pier,

respectively. When the values of P I , P D and

P m are less than the boundary shown In RelatIons-4, -5 and -6 in Section 3.3, the

damage Induced by soil liquefaction Is assumed

to be small and further Inspection Is not

necessary.

(4) Four-stage of Detail Stability Inspection:

In this stage, the detail Inspection should

be conducted to calculate the stability and

resistance of foundations by the earthquake

resistant design of highway bridge [1, 2] . For the unstable bridges for soil

liquefaction, the countermeasures for

reducing the effects of liquefaction should be

selected and designed for planning the

retrofit of bridges in the future.

299
5. FUTURE REMARKS

The results Introduced In this paper will be

referred in the next Inspection of earthquake

resistant potential for highway bridges con- ducted In 1991. The further Investigation

win be carried out after obtaining the

results of next inspection to improve the

procedure more reasonably In the future.

6. ACKNOWLEDGEMENT

The authors wish to express their sincere

gratitude to Dr. T. Iwasakl, Dr. Y. Sasaki and

Dr. K. Kawashlma of Public Works Research in- stitute (PWRI) for their invaluable comments

and to Road Disaster Prevention Section of

Road Bureau, Kanto Regional Construction

Bureau, Metropolitan Government, Kanagawa

Prefectural Office. Chlba Prefectural Office,

Yokohama City Office, Kawasaki City Office.

Japan Highway Public Corporation and

Metropolitan Expressway Public Corporation

for their cooperation of pilot study. The

authors also wishes to express their gratitude

to Dr. S. Uehara and Dr. T. MorohoshI of

National Research Institute for Earth Science

and Disaster Prevention for their allowance

to use the boring data-base and to Mr. T. Azuma of Ground Vibration Division for his

assistance to prepare the manuscript of this

paper.

REFERENCES

1) Japan Road Association. "Specification for

Highway Bridges, Part V. Earthquake Re- sistant Design", February. 1990 (in Japanese).

2) iwasakl T. , Kawashlma K. and Hasegawa K., "New Seismic Design Specifications for Road

Bridge In Japan". 22nd Joint Meeting U.S.- Japan Panel on Wind and Seismic Effects.

U.J.N.R.. Galthersburg, Maryland. U.S.A., May.

1990.

3) Toklda K, Fukul J. and Matsumoto H.. "Simplified Examination Method on the Stability of Bridges Considering the Effects of

Soil Liquefaction". International Workshop on

Remedial Treatment of Potentially Llquefiable

Soils. Task Committee "H" of Panel on Wind and

Seismic Effects. U.J.N.R.. Public Works Re- search Institute. Tsukuba Science City. Japan,

January. 1991.

300
Table-1 History of Inspection for Earthquake Resistant Potential of

Highway Bridge Considering Liquefaction Effects in Japan.

Stage Inspection Procedure for Liquefaction

1 1971 • Views by Site Observation.

Liquefaction is not considered directly.

2 1976 • Check List for Present Condition and Damage.

• Damage Potential (A, B, C) for Each Check-index.

• Liquefaction is not considered directly.

3 1979 • First- and Second-stage Inspection.

• Ground /Superstructure/Substructure.

• Calculation of Bearing Capacity of Pile (2nd-stage).

• Estimation of Cost for Countermeasure.

• Liquefaction potential of ground is considered,

ist-stage : Existence of Saturated Sandy Layers (Yes/No).

2nd-stage : Thickness of Saturated Semdy Layers

(~ 1.0 m ~ 10.0 m ~).

4 1986 • First-, Second- and Third-stage Inspection.

• Ground /Superstructure/Substructure.

• Calculation of Bearing Capacity of Pile (2nd-stage).

• Estimation of Cost for Countermeasure.

• Liquefaction potential of ground and Its effect to foundation

are considered.

Ist-stage : Existence of Saturated Sandy Layers (Yes/No).

2nd-stage : Calculate Ft. and Thickness (Hi_) of Layer with Fl. less than 0.6. Compare Hu with Depth of Foundation.

5 1991

• Inspection Procedure is investigated at present.

Table-2 Fu~De Relation"'

Depth. Z (m) Reduction Factor, De

Fl g 0.6 Z g 10

10 < Z S 20

01/3

0.6 < Fl S 0.8 Z g 10

10 < Z g 20

1/3 2/3

0.8 < Fl S 1.0 Z S 10

10 < Z S 20

2/3

1

1.0 < Fl 1

301
Table-3 Damage of Highway Bridge and Liquefaction In Japan

Earth- quake

(EQ)

Name of Highway

Bridge

Typical Diunage of Bridge Structures

Ob

No.

Jectlve Substructure i Damage L€

£ind Its Damage ' vel Liquefaction Phenomena

Site

Llq. Pot.

Ipl. NUgata

EQ

1964.6

M= 7.5

SyowaBrldge

• 5 simple girders

(P2--7) fell down.

• 2 piers subsided and

others Inclined.

• Pile deformed, etc.

1 PI Settled(3cm), Pier-top Disp.(34cm) 4 Severe

liquefaction occured.

0.57

2 P6 Subsided In the River. 5 9.47

3 P7 SettledOcm), Pier-top Dlsp.(7cm) 4 0.47

Hlgashl- Kosen

Bridge

• 1 simple girder

(P8~9) fell down.

• Pier subsided and In- clined.

• Piles deformed.

• Retaining wall abut- ment subsided, etc.

4 AL Slight damage (Unknown In detail) 2 Severe

liquefaction occured.

9.90

5 P7 2 0.06

6 P9 Settled(2.5cm), Pier-top DIsp.(15cin) 5 0.43

7 AR Settled(32cm), Pier-top DIsp.(13cm) 4 0.23

YactUyo

Bridge

• Substructure broke.

• 1 glrder(P2--3) was

In danger of fall.

• Retaining wall abut- ment subsided, etc.

8 AL Settled(4ecm), Inclined with 19.5° 4 Movement of ground by

liquefaction occured.

0.34

9 P2 Pier-body broke. 5 9.34

19 P10 Up-llft(7cm). Pier-top DIsp.(7cm) 3 9.13

Talhel Bridge

• Pier Inclined.

• Slightly damaged.

11 PI (Not damaged) 1 A few meters

of surface

liquefied.

0.42

12 P8 Pier- top Disp.(6cm) 3 0.17

Shln- Matsuhana

Bridge

• 1 girder (Pie- 11) fell down.

• Under-construction.

13 PI (Not damaged) 1 9.23

14 P7 Settled(2cm), Pier-top Dlsp.(2ecm) 4 19 m more llq. 9.78

Kosudo

Bridge

• Abutment and pier subsided.

IS PI Settled(75cm) 3 (Unknown)

9.17

16 AR Sett led( 7cm) 3 9.99

Mly«i«l- Ken-Okl EQ

1978.6

M« 7.4

AbukuDA

Bridge

• Crack/peel-off of

pier. Bearing broke. 17 P8

Pier cracked. Bearing broke. 3 Slightly sand boiled. 9.93

Yuriage

Bridge

• Crack of pier/girder

/parapet.

• Joint damaged.

18 PI Peel-off of pier. Bearing broke. 3 Slight sand boll.

9.18

19 P6 Movement of glrder(6cm) 3 9.99

Kaihoku

Bridge

Bearing broke.

• Slightly damaged.

20 P2 Bearing broke. 3

Not liquefied.

9.99

21 P4 Slightly damaged. 2 0.06

22 AR Slightly damaged. 2 0.06

Eal Bridge

• Crack/peel-off of

pier 23 P5

Pier cracked. Movement of girder(6cm) 3 (Unknown) 0.43

Klmazuka

Bridge

• Base of bearing

broke.

• in danger of fall.

24 P3 Base of bearing broke. 3 Sand boll. 9.98

NlhonKal- Cbubu EQ

1983.5

M= 7.7

Oga

Bridge

• Pier subsided and In- clined.

• Approach settled.

25 P4 Settled(13cm), Pier-top Dlsp.(23cm) 3 (Sand boll. Crack)

9.90

26 P5 Settled(15cm). Pier-top Disp.(26cm) 3 0.00

Ryuma

Bridge

• Abutment cracked.

• Bearing broke.

• Slightly damaged.

27 AL Abutment-body cracked. 4 Sand bolL 9.99

28 P2 (Not deunaged) 1 (Sand boil) 9.99

Gomyokoh

Bridgel

• Backfill of abutment

subsided severely.

• Bridge-body was damaged sllgtly.

29 AL Plle-bent cracked. 3 Sand boll. Crack.0.93

36 AR Slightly damaged. 2 9.99

Narusawa

Bridge

• Backfill of abutment

subsided.

• Bridge-body was damaged slightly.

31 AL Slightly damaged. 2 Sand boll, Crack.9.90

32 PI Not damaged. 1 0.00

Note: ( ) shows the unknown or estimated results without actual data.

302
Table-4 Classification of Damage Level of Highway Bridge

Related to Liquefaction

Damage

Level

Liquefaction Related

to Damage of Bridge

Standard to Estimate Damage Level

of Bridge and Countermeasure

5 Liquefaction is main factor and causes severe damag of

bridge.

• Girder is fell down.

• Substructure is broken.

• Countermeasure for liquefaction is needed preferentially.

4

Damage of bridge is relatively

severe, and liquefaction Is considered as one of causes.

• Pier-top displacement by subsidence

and/or inclination is more than 19 cm.

• Width of crack is more than 1 mm.

• Countermeasure for both liquefaction and other cause of dtimage is needed.

3

Liquefaction is not directly related to damage of bridge.

• Pier-top displacement by subsidence

and/or inclination is less than 19 cm.

• Width of crack is less than 1 mm.

• Genaral retrofit and/or repair for damage is needed.

2 Slightly damaged.

• Slight repair for deimage is needed.

1 Not damaged.

Note: Damage level for severe damage which was not Induced by liquefactln is considered to be 3.

Table-5 Classification of Ground Group"

Group Characteristic of Ground TG (s) •

I

• Rock, Diluvial Ground

(Firm) TG<9.2

n

• Alluvial Ground

(Medium-stiff ) 9.2STG< 9.6

m

• Soft and Thick Alluvial Ground

(Soft) e.esTG

• : Characteristic Value of Ground

Table-6 Relation between Ground Group and

Site Liquefaction Potential (Thl)

Ground

Group

Without Saturated Sandy Layer Ipi.<0.2

0.2g

Ipu< 0.4 0.4SIpu Total

I 32 7 0 0 39

n 14 75 13 2 104

in 0 34 27 13 74

Total 46 116 40 15 217

Estimate Suitably Est. Dangerously

Est. Safely Estimate Suitably

303
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305
Table-14 Cleissiflcatlon of Damage of Foundation Induced by Liquefaction

Abu tnicn t Movable Pier r ixea r ler

r

^iipiiiiiiii • ' ' ' ^'"li -u -

• Fall-down Induced by large displacement of pier-top.

• Foundation-collapse Induced by loss of bearing of ground.

• Settlement Induced by loss of friction of ground.

• Fall-down Induced by large displacement of pier-top.

• Settlement Induced by loss of friction of ground.

• Foundation-collapse Induced by loss of bearing of ground.

• Settlement Induced by loss of friction of ground.

Table-15 Indexes to Estimate LiquefactionInduced Damage Potential of

Highway Bridge

Yschiyo Bri. AL~P2

f 4

s 1 o

Hiqasi-Kosen Bri. P9 (Fjll«n Part) Showa Bri. P6 (Fillen Part)

Index Substructure Type

Abutment Movable Pier Fixed Pier

Device Against Fall O 0

Supported Co. on

Pier-Top Ij O

Foundation Type O O O

Bearing Co. of Foundation O O O

Height of Backfill O

Site Liquefaction

Potential Ipl. 0 O o

Ground Group loo O

r-*T—I\—I—

r

Ryuma Bri, A

I

Eai Bri. PI Taihei Bri. P5 Shin-Matuhaia Bri. PI

Note: • Is added Index on Llquefactlon-lnduced Damage Potential.

Si n-Matuhama Bri. P7

0.5 Site LIqnefaction Potential i'''-

1.0

Fig. 1 Relation between Site Liquefaction Potential

(fhu) and Damage Level of Substructure

10 12 14 16 18 20

HT (I) Thickn«ss of Lifuatiikii LiMr

Fig. 2 Distribution of Modified Thickness

of Liqueflable Layer (HT)

306
ISO

140

130

120

no

100

90

ao

70

eo

50

40

30

30

10

0

Totil : 187 Siiti

-I I

Difftrinet of Abflvi tkt S<> Lt>tl (>|

-I I 3

Oilf«r«Mt of trttii

I«t«r Livtl M

-10 12

Difftrtnet ol Ctoii4 imp

0 I

Dilfirtnet if

Fig. 3 Distribution of Categories

In Each Item

1 ) P I for Abutment

0

0

-

0

-

0

-

o ^

c ^

15.0

14.0

12.0

10.0

8.0 -

6.0

4.0

2.0

2 3 4 Damage Level

2 ) P n for Movable Pier { 3 ) P m for Fixed Pier

Fig. 4 Relation between Damage Level of Substructure

and LIquefactlon-lnduced Damage Potential

307
START

YES DESIGNED BY

DESIGN CODE

SINCE 1971 ?

YES

NO

BRIDGE TYPE NOT AFFECTED

SEVERELY BY LIQUEFACTION ?

FUNDAMENTAL

INSPECTION :

No

YES

N9

POSSIPLE TO GCT

BORING DATA AT OBJECTIVE

BRIDGE SITE ?

NO

SATURATED

SANDY LAYER

EXIST ?

GROUP I

YES

WHAT IS GROUND GROUP ?

GROUP

B OR C

GROUP n OR m

WHAT IS TOPOGRAPHICAL GROUP ?

SITE liquefaction

inspection:

'

GROUP A

POSSIBLE TO GET

BORING DATA AT

NEIGHBORING SITE

NO

YES

SATURATED SANDY LAYER EXIST ?

YES

YES SITE LIQUEFACTION POTENTL\L

Ipl < 9.2 ?

NO

CALCULATE

LIQUEFACTION DAMAGE POTENTIAL

P I . P a, P m

ABUTMENT

YES MOVABLE PIER YES

NO NO

FIXED PIER YES

P ni<4.5

DAMAGE

POTENTIAL

; INSPECTION

NO

DETAIL INVESTIGATION AND/OR CHECK OF STABILITY OF FOUNDATION

NO NECESSARY TO RETROFIT FOR LIQUEFACTION ?

YES

DETAIL SATBILnT ; INSPECTION

SELECT COUNTERMEASURE

END

Fig. 5 Example of Flow Chart for Inspecting Earthquake

Resistant Potential of Highway Bridge Foundation

308
Evaluation and Repair of M^jor Water Resources Structures

for Improved Safety and Seismic Stability

by

Dr. WilliEim Roper*

ABSTRACT: This paper addresses recently

developed technology for evaluation and repair of water resoiirces structirres for improved safety and

seismic stabihty. It covers selected examples of

research conducted imder the Repair, Evaluation,

Maintenance, and RehabUitation (REMR) Research

Program. The REMR program was a comprehensive six-year effort in which 54 individual

studies were conducted. Those studies which can be

related to wind and seismic events are discussed

under three categories: Monitoring and Inspection

of Existing Structures, Remedial Treatments to Improve Safety, and Repair of Damage.

KEYWORDS: Breakwaters; dissimilar armor;

foundation assessments; foimdation grouting;

geophysical methods; jetties; instrumentation

automation; Uquefiable foundations; Uquefiable soUs; monitoring structwes; nondestructive testing; rehef

wells; rubble-mound structure toes; underwater

inspections; ultrasonic pulse-echo; wave runup and

overtopping

1. INTRODUCTION

The percentage of the US Army Corps of

Engineers' budget devoted to maintenance, repcdr

and rehabihtation of its civil works projects

increased throughout the 1980's as the average age

of Corps structiu-es increased. About 70 percent of

the Corps' hydrauhc structures are over 20 years of

age; 49 percent are more than 30 years old; and 26

percent were constructed prior to 1940. Since it generally cost less to rehabilitate a structure than to replace it and budget constraints continued to increase, a major need was for technology to extend

the life of existing structures. The Repair,

Evaluation, Maintenance and RehabUitation

(REMR) Research Program was initiated in 1984 to meet this need. The overall objective of the

program was to identify and develop effective and

affordable technology for maintaining and extending

the service life of existing Corps civil works

structures.

The Corps was one of the first Federal agencies to recognize the need for research to address the

nation's deteriorating infrastructure. The REMR

Research Program has clearly demonstrated the

benefits of research in getting more value for the

dollars spent on maintenance and repair activities.

Technology from the REMR Research Program is credited with producing savings in excess of

$70 miUion dollars with estunated savings over the next 5 years exceeding $200 miUion for the Corps

alone. In addition to doUiu- savings, the program

has resulted in other benefits such as improved

safety and reliabihty, reduced manpower

requirements, and improved operational capabihties.

These figures represent a significant return on a $35 milhon dollar investment.

REMR was a broad based research program which was broken down into seven problem areas: (1)

Concrete and Steel Structures, (2) Geotechnical, (3)

HydrauUcs, (4) Coastal, (5) Electrical and

Mechanical, (6) Environmental Impacts, and (7)

Operations Management. A total of 54 individual

research studies were conducted over the 6 year life of the program.

Although most of the evaluation, maintenance,

repair and rehabihtation efforts at Corps projects are due to normal wear and tear, aggressive

environments, and the need for increased capacity,

damage from wind and seismic events does pose a

significant threat. Several REMR research studies

produced technology for evaluation, remedial

treatment, and repair efforts related to wind and

seismic events. This paper summarizes those

studies.

* US Army Corps of Engineers, 20 Massachusetts

Avenue, NW., Washmgton, DC 20314-1000

309
2. MONITORING AND INSPECTION OF

EXISTING STRUCTURES

The monitoring of existing structures during various

loading events c£m provide advanced warning of

impending failure or allow timely actions to be

taken which will prevent failure. Once damage has

occurred, it is important to thoroughly inspect the

damage prior to any repair or rehabilitation effort. The cause and extent of damage must be considered

for proper selection of repair or rehabiUtation

procedures. A summary of the results of REMR

studies on inspection and monitoring of existing structures is provided below. The results are appUcable to the evaluation of structures subjected

to wind and seismic events.

2.1 Nondestructive Testing

Nondestructive testing of concrete allows the inspection of larger areas at less costs than coring

and provides more information than a visual inspection. A comprehensive REMR technical

report describes the available techniques, theu- applications, advantages, and disadvantages.

A new, improved prototype ultrasonic pulse-echo

system was developed for the evaluation of concrete

structures where only one surface is accessible. The

system uses piezoelectric crystals for both signal

generation and detection. The 200-kilohertz

two-transducer system has a signal-to-noise ration of

18. The weight and dimensions of the improved

system, which has the shortest pulse length on

record, works well in making thickness

measurements of portland-cement concrete (<12

in.) and can indicate the presence of reinforcing

steel, voids, and inferior quaUty concrete both above

and below water.

To keep civil works metal structures fully functional

and safe, a thorough inspection procedure must be

implemented. A guide for the selection of an

appropriate test method for metal structures was

published. The guide includes information on the

theory, applications, advantages, and disadvantages

of each nondestructive testing method. Case

histories are included which describe how

nondestructive testing procedures can enhance

inspection routines.

2.2 Geophysical Techniques

Existing structures compUcate and limit routine

application of standard engineering geophysical

techniques for foundation assessments. New

geophysical methods were developed and standard

methods adapted that allow definitive and

cost-effective application to existing structures.

Field procedures, data processing methods, and

interpretation methods were developed or adapted

for high-resolution seismic reflection, groundpenetration radar, microgravity, and self-potential

methods. Two field test sites were selected and the

methods applied to the assessment of various

structure and foundation problems. At the

Lockport Approach Dike on the Illinois Waterway,

near Joliet, Illinois, a ground-penetrating-radar

siuA'ey detected possible seepage paths through the

dike, verified the location of the concrete core wall,

and assessed its condition. At Dike 1, Beaver Dam,

Arkansas, a comprehensive foundation assessment was conducted. The geophysical methods applied at Dike 1 succeeded in mapping seepage paths and

delineating the complex geology (Figure 1). Based on the results of the geophysical surveys and other

geotechnical investigation the rational design of a concrete cutoff wall to eliminate anomalous seepage was possible.

2.3 Evaluation of Damage to Underwater Portions

of Coastal Structures

Effectiveness of techniques for inspection and

evaluation of underwater damage is Umited by the severe conditions encoimtered in the zone

surrounding coastal structiu^es, including high wave

energy, strong currents, and low underwater

visibility. Current technology £md methods for performing inspections were examined, including

diver and crane surveys, imderwater video, and

commercially available side scan and other sonars. Guidance was developed, through testing and

experience in actual field conditions, on the

selection and use of these techniques. Training was

provided to Corps personnel, and the guidance was

used in evaluation of underwater parts of Humboldt

and Crescent City jetties on the California coast.

Since none of the available technologies is entirely

satisfactory, an effort was undertaken to develop an

underwater survey system that would be less

310
dependent on favorable conditions and provide less subjective results. A prototype system called the

Coastal Structure Acoustic Raster Scanner

(CSARS) (Figure 2) has been designed, constructed,

and improved through field trials. Results, thus far, have been promising.

2.4 Instrumentation Automation for Concrete

Structtires

Automation of instriunentation at existing concrete

structures can increase the timeliness and accuracy

of data and can reduce man-hour requirements. It can also permit the instruments to be monitored

from a remote location or from an off-site office. Information was accumulated on commercially

available hardware and software, and guidance was

developed on the use of these products for instrumentation automation.

This technology was demonstrated at Beaver Dam

near Eureka Springs, Arkansas, where pressure

cells, weirs, joint meters, and piezometers were

automated. Prior to the automation of the

84 piezometers, a team of field personnel took four

hours to read the piezometers which can now be

read in less than a minute.

2.5 Continuous Monitoring System for Structural

Safety of Large Dams

The normal behavior of a dam needs to be

imderstood so that abnormal behavior can be

detected. Precise daily studies of dam motion were

not previously feasible because the measurements

could not be automated. Under this study, a continuous monitoring system was developed that uses the Navstar Global Positioning System to determine three-dimensional differential movements

at up to eight points on a structure. Each position

is then compared to an original position to determine the total amoimt of movement.

Accuracies of + /-5 millimeters in the

three-dimensions are possible. The monitoring and

data recording equipment are linked together with

telephone lines, R-232 cable, or fiber optic cable. No system operator is required, and real-time

structural deformations may be monitored at remote locations such as an off-site office. A

prototype system was installed and demonstrated at Dworshak Dam, Idaho from July to September

1989.

3. REMEDIAL TREATMENTS TO IMPROVE

SAFETY

Improved inspection and analysis techniques now

available often point out the need for remedial

treatments to existing structures to improve their

safety. REMR studies to develop remedial

treatments for structures subjected to wind and

seismic events are described below.

3.1 Reducing Wave Runup £md Overtopping

Problems associated with runup and overtopping

have occurred on about 20 percent of the Corps'

coastal structiu^es. Problems from rimup and

overtopping range from (1) excessive wave action on

the lee side of breakwaters and jetties, (2) flooding

and erosion on the backside of seawalls, sea dikes,

and bulkheads, to (3) backside subsidence, erosion,

and sometimes coUapse of revetments.

Mathematical models and flume studies were used

to evaluate and rank the hydrauhc performance of

a ntmiber of seawall/revetment configurations. For

existing revetments, it was foimd that an attached

low berm fronting a rubble revetment significantly

decreases wave rtmup and overtopping. This

technology was used at the Temple Emanuel project on the Chicago waterfront (Figure 3) and resulted

in a 12-percent savings over the costs of the

traditional high berm revetment.

3.2 Remedial Treatment of Liquefiable Soils

Liquefaction of foundation soils can cause

catastrophic failure or limit the usefulness of a

project. It is now known, through new and

improved testing procedures, that many Corps

structures are located on liquefiable foundations.

Techniques appUcable for remedial treatment of the

foimdations of existing structures were identified,

and guidance was provided for their selection and

use. Two of the more promising techniques are sand compaction piles and stone colimins. The

relative costs of these techniques are moderate and

they provide (a) vertical support, (b) drains to reUeve pore water pressure, and (c) shear resistance

in horizontal and inclined directions. Stone coliomns were used to improve foundations conditions at Steel Creek Dam, Georgia (Figure 4).

311
3.3 Restoration of Relief Wells

Relief wells, a safety feature when installed at a

structure, can become clogged by chemical or

biological incrustation allowing uplift pressures to increase beneath the structiu-e. A procedure was

developed for restoring the efficiency of these wells using a process called "Blended Chemical High

Temperature," or BCHT. The Corps has

approximately 5,000 relief wells that must be treated

periodically or replaced. The BCHT process will extend the maintenance interval for these wells by

at least one year and will result in annual savings of

$500,000.

An example of additional savings is the relief wells

at the New Alton Pumping Station Project in the St. Louis District. Standard cleaning methods failed to bring these wells up to the required 80 percent of

original capacity. The BCHT process was

demonstrated to be capable of successfully cleaning

the wells, thus eliminating the expense of

replacement wells. The cost difference between

restoration and possible replacement of as many as 60 wells at this site was estimated to be $1.5 million.

4. REPAIR OF DAMAGE

Several repair techniques and materials were

evaluated under REMR and undoubtly many of

them can be used to repair damage from wind and

seismic events. However, those which would seem

to relate more specifically to repair of damage from

seismic and wind events are foundation grouting and

repair of coastal structures damaged from wave

action.

4.1 Grouting for Foundation Repair and

Rehabihtation

Foundation grouting has long been an accepted

procedure for improving foundation conditions.

The objective of this effort was to evcduate and

develop modern methods of foundation grouting to improve quahty control and reduce costs. Several

products were developed under this effort including a comprehensive technical report on consolidation

grouting of rock masses. A computerized grout control and monitoring

system was developed under a joint effort with the Bureau of Reclamation. The use of this system is

expected to substantially reduce the number of

changed condition claims by grouting contractors. A method to evaluate rock properties prior and

subsequent to grout injection was developed. A

drill rig instrumented with a drilling parameter

recorder (Figure 5) is used to predict groutabiUty,

and an acoustic emissions monitoring system is used

for mapping grout location. Test grouting on this method was conducted at Little Dell Dam, Utah.

4.2 Rehabilitation of Rubble-Mound Structiu^e Toes

Many breakwater failures have been attributed to

failure of the toes of these structures. No firm

guidance existed to aid Corps personnel in designing

the berms and most design work was carried out

using limited local field experience with past successes and failures. A series of two-dimensional and three-dimensional

breakwater stability physical model experiments was

developed and conducted to address the sizing of

toe berm and toe buttressing stone in breaking wave

environments. The two-dimensional tests focused on sizing of toe stone on rubble-mound structiu"e trunks exposed to 90-degree wave attack, i.e. wave

orthogonals perpendicular to structure crest. Toe

berm armor stone sizing for obhque wave attack on

rubble-mound structure heads and trunks was

examined in the three-dimensional model tests. Guidance for sizing toe berm armor stone was

developed for a range of wave and still-water level conditions. Guidance for sizing of toe buttressing stone was addressed for a limited set of incident wave conditions on structure trunks. Use of this new guidance should eliminate wave induced toe berm failures and result in significant dollju" savings

for the Corps.

4.3 Use of DissimilcU" Armor for Repair and

Rehabilitation

Model test have shown that overlays of concrete

units such as dolosse and tribars over stone require

larger sizes than if the structure were originally

designed for those particular units. Thus, the design

guidance for new structiu-es is inappropriate for use

in the design of repairs and rehabihtations involving

dissrmilcU" armor.

312
Stability tests were conducted of dolos overlays on

stone-armored, dolos-armored, and tribar-armored

breakwater and jetty trunks subjected to breaking

waves. Tests were edso conducted on tribar overlays

of stone-armored breakwater and jetty trunks and

dolos overlays of stone-armored breakwater heads

subjected to breaking waves. The results from these

tests can be used to design appropriately sized armor units for repair and rehabilitation of

breakwaters and jetties.

4.4 Sealing of Permeable Breakwaters and Jetties

Rubble-mound breakwaters and jetties are intended

to protect harbors and navigation channels from

excessive wave energy and to prevent shoaling of

chaimels and boat mooring areas by littoral material. The cores of many such Corps coastal

structures have deteriorated and become permeable

to such an extent that they do not serve, or only

partially serve, their original intended function.

A cost-effective alternative to traditional methods of

rubble-moimd structure sealing (i.e. dismantling to rebuild core sections, chinking layers along surface,

additional armoring layers, etc.) was determined to be drilling and grouting (sealing) a vertical barrier

curtain along the center line of the structure from

the bottom to approximately mean higher high water (Figure 6).

Sealing permeable brecikwaters or jetties should be

approached from the stamdpoint of preventing wave

or sand movement through the structure and not from the requirement of impau-ting structural

stability or strength. In planning a sealing

operation, a quantitative determination must be

made of the wave energy or sand passing through

the structure to ascertain economic benefits.

5. INTRODUCTION TO REMR-II PROGRAM

The successes of the REMR Program, the high

demand for its products, and the opportimity for similar successes on other REMR type problems were the justification for getting approval for a foUow-on effort (REMR-II), REMR-II was

initiated this fiscal year (FY91) and is proposed as a 7-year, $35-million effort. It wiU address new and

different needs which have been identified by Corps

field personnel. REMR-II will concentrate on

problems and areas which have the potential for

large payoffs and widespread applications. Much of

the technology that wUl be produced will have

appUcations outside the Corps to other types of

infrastructures. A significcmt effort will be devoted

to the coordination and sharing of this technology

with other federal agencies, state and local governments, and the private sector.

6. CONCLUSIONS

The U.S. Army Corps of Engineers was one of the

first Federal agencies to recognize the need for research to address the nation's deteriorating

infrastructure. The REMR Research Program was

implemented to address that need and has clearly

demonstrated the benefits of research in getting more value for the dollars spent on maintenance

and repair activities.

The REMR Research Program was a very broad

based program with 54 individual research studies. Several of these studies are applicable to the

evaluation, remedial treatment, and repair of

structures subjected to wind and seismic events, as summarized in the above text.

7. REFERENCES

1. O'Neil, E. 1989 (Apr). "Instrumentation

Automation for Concrete Structures, Report 4, Demonstration of Instrumentation Automation

Techniques at Beaver Dam, Eureka Springs,

Arkansas," Technical Report REMR-CS-5,

U.S. Army Engineer Waterways Experiment

Station, Vicksburg, MS.

2. Thornton, H.T. Jr., and Alexander, A.M. 1987

(Dec). "Development of Nondestructive Testing

Systems for In Situ Evaluation of Concrete

Structures," Technical Report REMR-CS-10,

U.S. Army Engineer Waterways Experiment

Station, Vicksburg, MS.

3. Kisters, F. and Kearney, F. 1991 (Jan).

"Evaluation of Civil Works Metal Structures,"

Technical Report REMR-CS-31, U.S. Army

Construction Engineering Research Laboratory,

Champaign, IL.

4. Barksdale, R.D. 1987 (Nov). "State of the Art

for Design and Construction of Sand Compaction Piles," Technical Report REMR-GT-4, U.S.

313
Army Engineer Waterways Experiment Station,

Vicksburg, MS.

5. Wilt, M.J. and Butler, D.K. 1990 (Mar).

"Geotechnical Applications of the Self Potentieil (SP) Method Report 4, Numerical Modeling of

Spanomahus: Documentation of Program SPPC

and AppUcations," Technical Report

REMR-GT-6, U.S. Army Engineer Waterways

Experiment Station, Vicksburg, MS.

6. Barksdale, R.D. 1987 (Dec). "Applications of

the State of the Art of Stone ColumnsLiquefaction, Local Bearing Failure, and

Example Calculations," Technical Report

REMR-GT-7, U.S. Army Engineer Waterways

Experiment Station, Vicksburg, MS.

7. Dickinson, R.M. 1988 (Jul). "Review of

Consolidation Grouting of Rock Masses and

Methods for Evaluation," Technical Report

REMR-GT-8, U.S. Army Engineer Waterways

Experiment Station, Vicksburg, MS.

8. Dobecki, T.L., Mueller, T.L., and Savage, M.B.

1989 (Jul). "High Resolution Seismic Reflection

Investigations at Beaver Dam, Arkansas,"

Technical Report REMR-GT-10, U.S. Army

Engineer Waterways Experiment Station,

Vicksburg, MS.

9. Carver, R.D. 1989 (Jan). "Prototype Experience

with the Use of Dissimilar Armor for Repair and

Rehabihtation of Rubble-Mound Coastal

Structures," Technical Report REMR-CO-2, U.S. Army Engineer Waterways Experiment Station,

Vicksburg, MS.

10. Carver, R.D. and Wright, B.J. 1988 (Feb).

"Stability of Dolos and Tribar Overlays for Rehabilitation of Stone-Armored RubbleMound Breakwater and Jetty Trunks Subjected

to Breaking Waves," Technical Report

REMR-CO-4, U.S. Army Engmeer Waterways

Experiment Station, Vicksburg, MS.

11. Carver, R.D. and Wright, B.J. 1988 (Jun).

"Stability of Dolos Overlays for Rehabilitation

of Dolos-Armored Rubble-Mound Breakwater

and Jetty Trunks Subjected to Breaking Waves,"

Technical Report REMR-CO-5, U.S. Army

Engineer Waterways Experiment Station,

Vicksburg, MS.

12. Carver, R.D. and Wright, B.J. 1988 (Aug).

"Stabihty of Dolos Overlays for Rehabilitation

of Tribar-Armored Rubble-Mound Breakwater

and Jetty Trunks Subjected to Breaking Waves,"

Technical Report REMR-CO-6, U.S. Army

Engineer Waterways Experiment Station,

Vicksburg, MS.

13. Ahrens, J.P. 1988 (Oct). "Methods to Reduce

Wave Runup and Overtopping of Existing

Structures," Technical Report REMR-CO-7,

U.S. Army Engineer Waterways Experiment

Station, Vicksburg, MS.

14. Simpson, D.P. 1989 (Apr). "State-of-the-Art Procedures for Sealing Coastal Structures with

Grouts and Concretes," Technical Report

REMR-CO-8, U.S. Army Engineer Waterways

Experiment Station, Vicksburg, MS.

15. Carver, R.D. 1989 (May). "Stabihty of Dolos

Overlays for Rehabilitation of Stone-Armored

Rubble-Mound Breakwater Heads Subjected to Breaking Waves," Technical Report

REMR-CO-9, U.S. Army Engineer Waterways

Experiment Station, Vicksburg, MS.

16. Kucharski, W. and Clausner, J.E. 1990 (Feb).

"Underwater Inspection of Coastal Structures Using Commercially Available Sonars, Techniccd

Report REMR-CO-11, U.S. Army Engineer

Waterways Experiment Station, Vicksburg, MS.

17. Markle, D.G. 1989 (Sep). "Stabihty of Toe

Berm Armor Stone and Toe Buttressing Stone on Rubble-Mound Breakwaters and Jetties,"

Technical Report REMR-CO-12, U.S. Army

Engineer Waterways Experiment Station,

Vicksburg, MS.

18. Rosati, J.D. and Denes, TA. 1990 (Oct).

"Field Evaluation of Port Everglades, Florida,

Rehabilitation of South Jetty by Void Sealing,"

Technical Report REMR-CO-15, U.S. Army

Engineer Waterways Experiment Station,

Vicksburg, MS.

314
Beaver Dam, Arkansas

Major Rehabilitation

Concrete Cutoff Wall

Figure 2. The Coastal Structure Acoustic Raster Scanner

is raised from its underwater surveying position.

315
316


Shaking Table Tests for Hydrodynamic Pressure Acting

on Double Cylindrical Caisson Type Breakwater in the Deep Sea

By

Tatsuo UWABE-'-, Makoto OSADA^ and Setsuo NODA"^

SUMMARY

Shaking table tests were conducted to investigate characteristics of hydrodynamic

pressure acting on a double cylindrical

caisson. This caisson is composed of a permeable outer cylinder whose upper part has

many openings and an impermeable inner

cylinder

.

The size of caisson model made of acrylic

boards was 50cm in height and 50cm in

diameter. The openings of outer cylinder

were made within the 20cm limits below the

top of the model. Opening ratio of 10%, 25%

and 100% were used. These models were called

as a permeable type. The model, that has a permeable half circle in front of cylinder

and an impermeable half circle in rear side, was made and called as a wave dissipating

type. In the experiments, hydrodynamic

pressures were measured under the conditions

of single cylinder type and cylindric wall

type. In case without opening, the resultant force

of hydrodynamic pressure acting on the

cylindric wall measured here was almost equal

to the value of Westergaard ' s formula for the

wall whose width was the same as the diameter

of outer cylinder.

In case with openings, the resultant force of

hydrodynamic pressure on the permeable type

model with 25% opening was about 0.7 times of

that on the model without opening. The ratio

of resultant force on the wave dissipating

type model with 25% opening to that on the

model without opening was above 0.9.

KEY WORDS : Double Cylindrical Caisson,

Hydrodynamic Pressure, Opening, Shaking Table

Test.

1. INTRODUCTION

As offshore and waterfront development is

strongly required recently, many large

structures have been constructed in deep sea area for port and airport facilities.

Because in this area, most of the grounds are very weak and large hydrodynamic pressure

acts on these structures, they are supposed

to be easily damaged by an earthquake. So we have to investigate the dynamic stability of

these structures during earthquake.

Since the breakwater is constructed on deeper

seabed ground as compare to other port

facilities, it is necessary to go ahead with

development of a deep-sea breakwater. In Japan, a composite type of breakwater that

consists of a rubble mound and concrete

caissons is constructed in general. The cost

of the rubble mound of breakwater in deep sea strongly increases. To reduce this cost, it

is needed to introduce a large size type of

caisson. The double cylindrical caisson

discussed in this study is one of them. This

caisson is composed of a permeable outer

cylinder whose upper part has many openings

and impermeable inner cylinder. Wave that

permeate through outer cylinder is dissipated

in the wave chamber that is formed by outer

and inner cylinders.

Generally a design section of the breakwater

is determined by wave forces. But the

hydrodynamic pressure during earthquake that

become large with the increase of the water

depth is a main external force for design of

the deep-sea breakwater. In the current

design standard of port and harbour

facilities, the hydrodynamic pressure is

given by the Westergaard ' s formula for a vertical wall, and is given by Zangar's

formula for an inclined wall . Uwabe, Noda,

Chiba and Higaki reported that the

hydrodynamic pressure on the rectangular

caisson is calculated by the Westergaard '

s

formula when the breakwater doesn't show

large response. On the other hand, many

studies for hydrodynamic pressure on a cylinder have been carried out to design of a footing of bridge. Sakurai presented

general solution of hydrodynamic pressure on

a long cylinder, which is derived from the

equations of motion considering the flexural

oscillation of the cylinder. Goto and

Toki reported the solution of three

dimensional hydrodynamic pressure using

Euler's equation and velocity potential that

is derived from wave equation. But these

studies were carried out for the single

cylinder, and characteristics of hydrodynamic

pressure on a cylindrical caisson with

openings were not analyzed in the past

studies. Also it is necessary to calculate

the hydrodynamic pressure acting on a cylindric wall when this cylindrical caisson

is adapted to a composite type breakwater.

But there is no description about calculation

1 Dr. Eng., Chief of Earthquake Disaster

Prevention Laboratory, Structural Engineering

Division, Port and Harbour Research Institute,

Ministry of Transport, 3-1-1 Nagase,

Yokosuka, Japan.

2 Member of Earthquake Disaster Prevention

Laboratory, ditto

3 Dr. Eng., Director, Structural Engineering

Division, ditto

319
method of this pressure in the current design

standard. Therefore, shaking table tests were conducted to investigate the

characteristics of hydrodynamic pressure on

the double cylindrical caisson.

2. MODEL OF DOUBLE CYLINDRICAL CAISSON

2 . 1 Outline of Double Cylindrical Caisson

Figure 1 shows an outline of double cylindrical caisson type breakwater. This type of

caisson has many advantages as follows;

(1) Necessary volume for stability is effectively given.

(2) Costs of materials are cut down.

(3) Surface area of the structure is small.

(4) Various permeable ratio is chosen.

This structure is composed of double vertical

cylinders. The upper part of outer cylinder

above an intermediate floor slab has many

openings. Waves that permeate through these

openings are dissipated in the wave chamber

that is formed by the intermediate floor

slab, and the inner and outer cylinders. An

opening ratio of outer cylinder is defined as the ratio of the total area of openings to the upper part area of outer cylinder. On

the other hand, the inner cylinder and the

lower part of the outer cylinder are impermeable. Inside them, concrete and sand

are filled.

2.2 Model

Caisson models for shaking table tests were

made of acrylic boards. For these experiments, we used a measuring model and two dummy models. Figure 2 shows the outline of

these models. The size of the model was 50cm

in height and 50cm in diameter. The diameter

of the inner cylinder is 30cm. On the top of

the model there is a cover to link both

cylinders and at the bottom of the model

there is an acrylic square board to connect

to load cells. Sixty four openings whose

shape was rectangular were made around the

outer cylinder within 20cm limit below the

top of the model. The opening ratios of

model were 0,10,25 and 100%. The opening

ratio of 0% means that there is no opening on

the outer cylinder. The opening ratio of

100% means that there is no outer cylinder

above the intermediate floor slab. In this

study, these models were called as a permeable type. Besides these, the model

that had a permeable half circle in front of

cylinder and an impermeable half circle in rear side was made and called as a wave dissipating type. Therefore, we used four

permeable type models ( opening ratio: 0, 10,

25, 100% ) and three wave dissipating type

models ( opening ratio: 10, 25, 100% ).

3. MODEL TESTS

3.1 Outline of Model Tests

The contents of model tests are as follows;

(1) In case without opening

a) Single cylinder type model

b) Cylindric wall type model

c) Dotted line set type model

(Space between models: 0.5, 4.0cm ) (2) In case with openings

a) Single cylinder type model

(opening ratio: 10, 25, 100%)

b) Cylindric wall type model

(opening ratio: 10, 25, 100%)

The tests of single cylinder type model mean

that only the measuring model was set in a container on the shaking table, while the

tests of cylindric wall type model mean that

the dummy models were set in contact with the

measuring model. The tests of dotted line

set type model were only conducted using 0%

opening models. In this case, the dummy

models were set leaving a space from the

measuring model, then relation between the

characteristic of hydrodynamic pressure and

the distance of space was investigated.

3 . 2 Measuring items

Response accelerations and hydrodynamic

pressures on the model were measured. The

hydrodynamic pressure was measured by water

pressure gauges that were set on the surface

of the model. Distribution and resultant

force of the hydrodynamic pressure on the

model were given by these measured values.

However in case with openings, it was difficult to measure the hydrodynamic pressure on openings. Therefore the resultant

force of the hydrodynamic pressure was measured by four load cells set under the

model

.

Figure 3(a) shows an array configuration of

accelerometers , water pressure gauges and

load cells in case without opening. Because

the model was symmetrical with the direction

of excitation, the water pressure gauges were deployed in six rows and three columns on a quarter area of the outer cylinder. Angles

from the direction of excitation of each

column were 0, 34 and 67.5 degree. The

response acceleration of the model was measured by six accelerometers. Five

accelerometers were set in a row inside the

outer cylinder and another accelerometer was

put on the top of the model.

Figure 3(b) shows an array configuration of

instruments on the model with openings.

Seven water pressure gauges were used to measure distribution of the hydrodynamic

pressure on the inner cylinder. Three of them were set in a column on 0 degree from the

direction of excitation and others were set

in a row around the half part of the inner

cylinder. Positions of the water pressure

gauges and the accelerometers set on lower

part of the outer cylinder were same to those

of the case without opening. In case of the wave dissipating type, a permeable half

320
circle was set on a side on which the water

pressure gauges were set..

3.3 Setting of Model

Two kinds of shaking table that were of

electromagnetic type and an underwater

shaking table were used for the present

study. The underwater shaking table was developed for large structures constructed in

deep sea area^ . Since the shaking table is set on the bottom of the large water tank, it can reproduce the seabed ground during

earthquake

.

In model tests using electromagnetic type

shaking table, a container whose size was 3m

long, 1.5m wide and 1.2m high was set on the

shaking table. Figure 4 shows the location

of the models. The load cells were set in a steel box and the measuring model was put on

them. Crushed stones were poured around the

steel box so that the level of the bottom of

the model was same to that of the surface of

them. The dummy models were set on rectangular steel plates that were put on the

crushed stones. In order to fix the model to the container during shaking, the dummy

models were connected to a cross beam at the

top of the container. In case using the load

cells, it was necessary that the measuring

model didn't touch the dummy models and the

steel box during shaking. Therefore, little

spaces were inserted between them. As the

effect of this, it was expected that the

hydrodynamic pressure on the model decreases

in compare with a case without a space.

However, we judged that it was possible to estimate the correlation between the

hydrodynamic pressure and the opening ratio

from the results of test using load cells.

Besides the tests described above, the model

tests in which the models were directly set on the bottom of the container were conducted

to get rid of effects of the load cells and

the crushed stones.

The underwater shaking table was used for the

dotted line type model tests. Figure 5 shows

the outline of these tests. The single

cylinder type model tests and the cylindric

wall type model tests were carried out using

the underwater shaking table to compare with

the results of the tests using container.

Input motion used for the experiments was of

sinusoidal wave form that was of 5Hz and of

lOcycles. Maximum accelerations of the input

motion were 50, 100 and 200Gal. Table 1 shows the whole test conditions mentioned

above

.

4. RESULT OF MODEL TEST

4 . 1 Distributions of Accelerations

Distributions of peak values of response

accelerations obtained from time histories

are shown in Fig. 6. Figure 6(a) shows

results of cases that the models were

directly set on the bottom of the container.

It is easy to recognize that the response

acceleration of the top of the model(Al) is almost same to that of the bottom of the

model(A5). Figure 6(b) shows the results of

the case using load cells. In this case, the

response acceleration of Al is a little

larger than that of A5. This tendency was more obvious in single cylinder type than in

cylindric wall type, and was more obvious in

impermeable type than in permeable type. But the ratio of acceleration at Al to that at A5

is no more than 1.1 in maximum. Therefore,

it was also judged that the models were

almost shaken as a rigid body in case with

load cells.

4 . 2 Distributions of Hydrodynamic Pressure

Distribution of the hydrodynamic pressure on

the model, which was set on the bottom of the

container, is described here. Peak values of

second and third half waves were chosen from

time histories for analysis (called as p). As mentioned earlier, the model was shaken as

a rigid body so that a mean value of

accelerations measured each accelerometer was used here as a response acceleration of the

model. A ratio of this response acceleration

to the gravity acceleration was named a seismic intensity (called as k). Since a relation between p and k was almost liner, we used a mean value of p/k for following

discussion.

4.2.1 Distribution of Hydrodynamic Pressure

( without opening ) Figures 7(a)and(b) show vertical distributions of the hydrodynamic pressure on the

model in case of the single cylinder type and

of the cylindric wall type respectively. The

hydrodynamic pressure on the cylindric wall

is larger than that on the single cylinder.

Figure 8 shows horizontal distributions of

the hydrodynamic pressure on the model at the

height of 17cm from the bottom of the model.

It is easy to recognize in this figure that

though the hydrodynamic pressure become small

with increase of the angle from the direction

of excitation in the single cylinder type,

that become large in the cylindric wall type.

The hydrodynamic pressures measured in the

dotted line type are between them. These

facts suggest that the hydrodynamic pressure

concentrates near contact point of cylinders

in case of the cylindric wall type, and when

a little space is set between them, the

hydrodynamic pressure is rapidly decreased.

4.2.2 Resultant Force of Hydrodynamic

Pressure (without opening)

Figure 9 shows vertical distributions of

resultant force of the hydrodynamic pressure

on the model that are approximately

calculated using the horizontal distributions

at each level. Distributions of the Goto&

321
Toki's formula and the Wes tergaard ' s formula

are also shown in Fig. 9. ^he distribution of

the Westergaard ' s formula is obtained for

the wall whose width is the same as the

diameter of the outer cylinder. The

distribution in case of the single cylinder

type agrees with that of the Goto&Toki's

formula, and the distribution in case of the

cylindric wall type is almost same to that of

the Wes tergaard ' s formula. Therefore, it is

considered that the resultant force of the

hydrodynamic pressure on the cylindric wall

is given by the Westergaard ' s formula.

Distributions in case of the dotted line type

lie between those in case of the cylindric

wall type and in case of the single cylinder

type.

Whole resultant force of the hydrodynamic

pressure was approximately obtained by calculating the areas formed by each distribution.

The results of calculations are shown in

Table 2. The resultant force in case of

dotted line type with 4.0cm space, which is

equal to 8% of the diameter, is 0.67 times of

that in case of the cylindric wall type. As described earlier, this result suggests the

hydrodynamic pressure is decreased by

inserting a little space between caissons.

4.2.3 Distribution of Hydrodynamic Pressure

( with openings ) Figure 10(a) shows vertical distributions of

hydrodynamic pressure in case of the single

cylinder type with openings. The values above

31.5cm were measured by the water pressure

gauges on the inner cylinder. Though the

hydrodynamic pressures measured at Wla-W3a

show large value with increase of the opening

ratio, the values at W4-W6 decrease. In particular, the value at W4 considerably

decrease. From these results, it seems that

the hydrodynamic pressure on the lower part

of the outer cylinder was decreased with

decrease of hydrodynamic pressure on the

upper part effected by the openings. On the

other hand, it is estimated that the

hydrodynamic pressure on the inner cylinder

shows large value with increase of the

opening ratio. Figure 10(b) shows vertical

distributions in case of the wave dissipating

type. Differences between the hydrodynamic

pressures of each opening ratio are small as compare with those of the permeable type.

Figures ll(a)and(b) show horizontal distributions of the hydrodynamic pressure on the

lower part of the outer cylinder in case of

the permeable, and the wave dissipating type.

It is obvious that the hydrodynamic pressure

decrease with increase of the opening ratio.

Figure 12 shows the comparison of the

distributions of hydrodynamic pressure

between the permeable and the wave dissipating type. Though the hydrodynamic

pressure in the wave dissipating type is

larger than that in the permeable type as

expected, the differences of the values at W4-W6 are very small. These characteristics

of the single cylinder type as described

above were observed similarly in the

cylindric wall type.

Figures 13(a)and(b) show horizontal distributions of the hydrodynamic pressure on the

inner cylinder in case of the permeable type

and the wave dissipating type. In case of

the permeable type, values at W4a and W6a,

and at W3a and W7a are symmetrical about W5a

respectively. Absolute values of each water

pressure gauge increase according to the

opening ratio. In case of the wave dissipating type, however, signs of the

hydrodynamic pressures are same at each point

and the absolute values of them increase as increase of the angle. Also the hydrodynamic

pressures measured at W5a, W6a and W7a

increase with decrease of the opening ratio.

In the cylindric wall type, though the

hydrodynamic pressure was larger than that in

the single cylinder type, same characteristics were also obtained.

4 . 3 Relation between Resultant Force and

Opening Ratio

The resultant force of the hydrodynamic

pressure was estimated from measured values

of load cells. They contained inertial

forces of the model, load cells and water

inside the model. In the experiments,

therefore, the model was shaken in case without water and in case with water, and the

inertial forces of the model and the load

cells given by subtracting the measured

values without water from those with water,

and the inertial force of water inside the

model was removed from measured value of load

cells to obtain the resultant force of the

hydrodynamic pressure. In this process,

however, the water in the wave chamber was regarded as hydrodynamic pressure in case with opening, and was regarded as inertial

force in case without opening. Therefore,

the inertial force inside the wave chamber

was added to the hydrodynamic pressure on the

model without opening when the resultant

force on the model without opening was compared with those on the models with

openings

.

Peak values of the second and third waves of

time histories measured by load cells were chosen as measured values of the load cells

in the same way as the hydrodynamic pressure

(called as f). Figure 14 shows relation

between the accelerations and the values of

the load cells. Straight lines in Fig. 14 shows mean values of f/k used for this

analysis

.

Figure 15 shows relations of the resultant

forces of the hydrodynamic pressure and the

opening ratio. The vertical axis shows the

ratio of resultant force in case with

openings to that in case without opening. As

322
mentioned above, the resultant force in case without opening contains the inertial force

of the water inside the wave chamber. In

case of the permeable type of single

cylinder, the ratio of resultant force with

10% opening is 72% and that with 100% opening

is 62%. In case of the permeable type of

cylindric wall, as similar to these results,

the ratio of resultant force with 10% opening

is 78% and that with 100% opening is 66%. On

the other hand, the ratio of resultant force

in case of the wave dissipating type shows

little decrease with increase of the opening

ratio.

5. CONCLUSION

Shaking table tests were conducted to investigate characteristics of the hydrodynamic

pressure acting on the double cylindrical

caisson. The following conclusions are obtained.

(1) The resultant force of hydrodynamic

pressure on the cylindric wall was almost

equal to the value of the Wes t ergaar d'

s

formula for the wall whose width is same as the diameter of the cylinder.

(2) The hydrodynamic pressure on the

cylindric wall concentrates near the contact

points of cylinders. This pressure concentration is relaxed by slightly separating

them each other.

(3) In case of the permeable type, the

hydrodynamic pressure on the lower part of

the outer cylinder tends to decrease with

increase of the opening ratio. In particular,

the pressure near the openings considerably

decreases. In contrast with this, the

hydrodynamic pressure on the inner cylinder

tends to increase with increase of the

opening ratio. These characteristics were

also observed in case of the wave dissipating

type.

(4) In case of the permeable type, the horizontal distribution of the hydrodynamic

pressures on the inner cylinder is almost

symmetrical with zero. But in case of the wave dissipating type, the sign of the

hydrodynamic pressure around inner cylinder

is same all at once and the absolute value of

it increases as the angle increases. Also,

the absolute value of it increases as decrease of opening ratio for the angle above

45 degree.

(5) The resultant force of hydrodynamic

pressure on the permeable type model with

25% opening is 0.7 times of that on the

model without opening. The ratio of

resultant force on the wave dissipating

type model with 25% opening to that on the

model without opening is above 0.9.

REFERENCE

1) Japan port and harbour association:

Technical standards for port and harbour

facilities, March 1989.

2) UWABE, T., NODA, S., CHIBA, S. and HIGAKI,

S. : Coupled Hydrodynamic Response Characteristics and Water Pressures of Large

Composite Breakwaters, Report of the Port

and Harbour Research Institute, Vol. 20, No. 4, December 1981.

3) SAKURAI, A.: Vibration of Pier constructed

in the Water, Civil Engineering, Vol.16,

No. 6, 1961.

4) GOTO, H., TOKI, K. : Fundamental Studies

on Vibration Characteristics and

Aseismic Design of Submerged Bridge Piers,

Proc. of JSCE, No. 100, 1963

5) NODA, S., KURATA, E. and OSADA, M.

:

Development of Underwater Shaking Table,

Proc. of the 20th Earthquake Engineering

Conference, 1989.

6) Westergaard, H.M. : Water Pressures on Dams

during Earthquakes, Trans. ASCE98, 1933,

pp. 418-434.

Table 1 Test conditions

Type of Model

Single Cylinder Type

Cylindric Wall Type

Dotted Line Type

(Space between Models : 0.5,4.0cm)

Opening Ratio 0 , 10 , 25 . 100 %

Type of Caisson Permeable , Wave Dissipating

Measuring Item Hydrodynamic Pres. , Acceleration

Measuring Method of Hydrodynamic Pres.

Water Pres. Gauge Water Pres. Gauge + Load Cell

Input Motion Sinusoidal Motion (5Hz, lOWaves)

Max. Acceleration 50 , 100 , 200 Gal

323
Table 2 Measured and theoretical resultant forces of hydrodynamic pressure

Measured Values Theoretical Values

Wall Type

Dotted Line Type

Single Type West ergaard Goto&Toki

0. 5 cm 4. 0 cm

Resultant

Force (kgf)

133. 1 113. 6 89. 5 57. 3 138. 6 55. 8

Ratio 1 0. 853 0. 672 0. 431 1. 04 0. 419

Fig.l Outline of double cylindrical caisson

type breakwater

, 50

30

Unit : cm

Fig. 2 Model of double cylindrical caisson

324
51.5

Acryl ic

Board

\

.A-6

Al oWI3 oW7 .owi

A2 owi4 owe

OWI5 OW9 • A3

W2

<PW3

• pV

A4 1 oWI7 oWII

OWI8 O^'^AS*?^^

(12.5 1(12.5)

I Load Cel

V- -1-

Cn3

Outer Cyl inder ; Inner Cylinder

Cross

Section

T-r-r-r-2

Shaking Table

(a) Without opening

.30 0—^'^-

,A6

Ala A 2*0(^W2c •6w3o oW4a

OWI7 own

A3a

)W4

M

W5

JiW6

OW5o

I 500_L

Load Cel

LO

LO

LO

Unit : cm

(b) With openings

Fig, 3 Locations of pickups

Model of Double Cylindrical Caisson

2

I 50 0

50.0

1 Inner Cyl inder (d=30)

"T 7

-1 ? > 1

-i—

CS]

Dummy Models

Outer Cyl inder (d=50)

"T 1 1 T-

"T 1 T r

_! I I 1 Steel Box Cover of Load Cells

Crushed Stone

_j I )t_ 299 0

Direction of Vibration Unit : cm

Fig. 4 Setup of model

325
0*

Shaking Tabic

D: Space between models

V

EoO

inI

Fig,5 Setup of dotted line type test

Second half wave

T i me Third half wave[-60

Thi rd ha 1 f wave

AiA2

A3

A4

A5AoL

-40

-20

Single type

Wal 1 type

Second half wave

]

-250 -200 - 100 0 100 Acceleration (Gal)

J 200 250

(a) Direct setting on container

Th i rd half wa ve

-250 -2

—•- -80

-60

Single type

Wall type

-40

20

' I Second ha If wave

-100 0 100 Acceleration (Gal)

200 250

(b) With load cell

s

Fig. 6 Vertical distributions of acceleration (0% type)

326
50.0 r40.0

4-> 30.0 -

20.0

10.0

0.0

0 SI

00 A

Locations of

Water Pres. Gauge (deg)

0 : 0

: 34 ^ : 67.5

t3> A

0 0

0

0.0 10.0 20.0 30.0 40.0 50.0

P / k (gf/cm2)

50.0

40.0

30.0

CI) EC 20.0

10.0

0.0

A 00

A DO

A DO

Locations of

Water Pres. Gauge (deg)

0 : 0 Q : 34 ^ : 67.

O 0

a 0

o 0

0.0 10.0 20.0 30.0 40.0 50.0

P / k (gf/cm2)

(a) Wall type (b) Single type

Fig. 7 Vertical distributions of hydrodynanii c pressure (0% opening)

-A-0—

Wall Type

Line Type 0. 5 cm Line Type 4. 0 cm Single Type

/

/

30 20 10

P / k ( gf/cm^) Direction of

Vibration

Fig. 8 Horizontal distributions of

hydrodynamic pressure (17cm height)

327
P/ k (kgf/cm )

F i g. 9 Measured and theoretical values

10.0 -

Opening ratio

0%

10% 25%

100%

0.0 10.0 20.0 30.0 40.0

P / k (gf/cm^)

51.5 5O0

40.0

3O0

= 200

10.0 -

0 0

Opening ratio —o— : 0% —(h- : 10% -<y- : 25% —^ : 100%

0.0 10.0 20 0 30 0 40.0

P / k (gf/cm^)

(a) Permeable type (b) Wave dissipating type

Fig. 10 Vertical distributions of hydrodynami c pressure

(s i ngl e type)

328
Wall

type

Single Opening

type ratio —o

: 0%

: 10% —-- : 25%

: 100%

30 20 10 P/k (gf/cm^) Direction of

V i 1) r a I i 0 ii

20 E

30 20 10 P/k (gf/cm^)

0

Direction of vibration

(a) Permeable type (b) Wave dissipating type

Fig, 11 Horizontal distributions of hydrodynamic pressure

(26cni height)

51.5 50.0

400

30.0

20.0

10,0

0.0

Opening ratio 10% — : Permeable type —0— : Wave d i s s i p . type

0.0 10.0 20.0 30.0

P/k (gf/cm^)

400

Fig. 12 Hydrodynamic pres. of permeable and wave dissipating type (10% opening)

329
Fig. 13 Horizontal distributions of hy d r od y nam i c pres.

on inner cylinder (single type)

330
—o— : Permeable , Single Type —•— : Permeable , Wall Type

—-0- : Wave Dissipating, Single Type

: Wave Dissipating, Wall Type

50 1 00

Opening Ratio (%)

Fig. 15 Hydrodynamic pres. versus opening ratio

331
>1

i
Theme III

Storm Sm*ge and Tsmiamis
I

1

i

t

I

I
Subaerial and Submarine Landslide Generated Tsunamis

by

James F. Lander', Patricia A. Lockridge^, and Herbert Meyers^

ABSTRACT

Subaerial and submarine landslide-generated

tsunamis constitute a class of tsunamis which,

although they have caused at least 58,000

fatalities in the last four hundred years, are inadequately studied and generally fall outside current warning and mitigation

systems. This is because their effects,

although potentially disastrous, are local with

little available response time once they are generated. They are most common in areas of glaciation, volcanic activity, submarine

canyons, and reservoirs. While warning and

mitigation activities are difficult for this class

of tsunamis, there are monitoring and other

actions that can be taken on a site by site basis. The definition of a tsunami is expanded to include all impulsively-generated

traveling gravitational water waves whose

velocities are a function of the water depth.

KEYWORDS: Landslides, Natural Disasters,

Natural Hazards, Tsiuiamis, Water Waves.

1. INTRODUCTION

Most tsunamis, including almost all of those

which cause damage beyond their source

regions, are caused by tectonic displacements

associated with earthqu£ikes. The effect of

these can be mitigated to a degree by

warning systems, particularly in regions

beyond the immediate source area. However,

there are also tsunamis caused by a variety of

mechanisms associated with volcanoes,

landslides and other phenomena which are

significant hazards and, in general, are inadequately treated by research and

mitigation efforts. This paper wiU examine

subaerial and submarine landslide-generated

tsunamis including those triggered by

earthquakes or volcanic activity and other

indirect causes.

1.1 Frequency and Distribution of Occurrence

Landslide-generated tsunamis causing

fatalities have occurred more than 25 times

since 1596 and occur in many places

worldwide. They have caused about 58,000

fatalities including about 1500 in the 20th

Century.

Figure 1 shows the location of the known

fatal landsUde tsunamis. It also shows the

four types of environments most common for

landslide tsunamis: areas of glaciation, areas of recent volcanic activity, submarine

canyons, and reservoirs.

Table 1 lists these tsunamis. This Ust understates the problem since it is not always

clear when landslides are the cause of a tsunami during complex earthquake or volcanic sequences.

1.2 Definition

Since some of the waves discussed here may

seem to be different from what is usually

meant by the term "tsunami" the following

definition for "tsunami" is used:

A tsunami is an impulsively generated,

traveling gravitational water wave which

has a velocity (v) proportional to the

square root of the depth (d) and where

if the depth is great in comparison to the wave amplitudes.

This definition uniquely identifies a single

geophysical phenomena eliminating seiches

'University of Colorado CIRES, Boulder, CO

80309

^NOAA National Geophysical Data Center,

Boulder, CO 80302

335
which are standing waves, storm surges, and

various other waves. As a practical matter

seiches generated by tsunamis are usually

included as part of the tsunami phenomena.

The original Japanese use of the term and

earlier European use included storm waves

and surges, but these two are two distinct

phenomena. Also in much current western

literature many events here called tsunamis

are called "water waves," an unnecessary and

vague term.

1.3 Causes

The basic cause of landslide tsunamis is the

failure of over-steepened deposits or slopes

above or below the water surface. Once

unstable conditions have been created the

failure can be triggered by any of several

agents such as earthquakes, volcanic activity,

rain (which increases the mass and reduces

the friction), and continued deposition or progressive failure from creep. Once the

slide material has been released, the resulting wave depends on the kinetic energy in the

sUde, the volume and geometry of the body

of water, and possibly other factors. On

impact of subaerial landslides on water, both a surge or splash and a tsunami are generated. The surge such as the famous 524

m runup associated with the Lituya Bay event

is not part of the tsunami as defined here

since its velocity is kinetically-generated

rather than gravitation in nature. Its velocity

does not depend on the depth of the body of

water. The displacement of the water surface

does create a true tsunami.

Submarine landslides are less well

understood. They often generate turbidity currents but may also be rotational.

Turbidity currents as inferred by broken

submarine cables may or may not be

sufficient evidence of the generation of small

tsunamis not otherwise observed. Submarine

landslides associated with submarine canyons

may be due to the purging of sediments from

coastal sources or sediments transported along

the coasts by currents or they may be part of

the process of creating and expanding the

canyons laterally or headword. Surprisingly,

perhaps, large river deltas do not seem to be

sources of submarine slump tsunamis. None

have been reported from the great deltas of

the Mississippi, NUe, Amazon or other large

river deltas. This may be due to the better

sorting and finer grain size of these deposits.

1.4 Environments

Landslide tsunamis are associated with four

principal environmental regimes as follows:

I. Glacial

a) over-steepened fjord walls, i.e., Lituya Bay, Alaska 1958; Tafjord,

Norway, 1934

b) glacial outwash deposits, i.e.. Gulf

of Alaska 1964

c) glacial ice, hanging glaciers and

calving, i.e.. Disenchantment Bay,

Alaska, 1905.

II. Volcanic

a) Failure of pyroclastic deposits on

volcanic slopes, i.e., Shimabara,

Japan, 1792.

b) Collapse of lava flows upon reaching

the sea, i.e., Hawaii, 1919

c) CoUapses accompanying eruptions,

i.e., St. Augustine, Alaska 1883.

d) Collapses of the volcanic edifice

produced by earthquakes at the site,

i.e., Hokkaido, Japan, 1741.

in. Submarine Canyons, i.e.. Grand Banks,

Newfoundland, 1929; French Riveria

1979.

rV. Reservoirs, i.e., Vaiont Dam, Italy 1963;

Roosevelt Lake, Washington 1949-51.

Reservoirs are a particular problem. These

are typically built in steep valleys which may

show signs of landsliding before the reservoir was constructed. In addition, the reservoir,

may raise the water table, create microclimates of enhanced rainfall or snow,

and create the possibility of inducing

earthquakes due to crustal loading. Their

construction also creates the possibility for

landslide-induced surges and tsunamis which

can cause damage upstream, transverse to the

sUde, and downstream overtopping of the

dam. Downstream of the dam, flooding may

occur from the breaking of the dam or its overtopping but these are floods and do not

meet the tsunami definition criteria.

336
1.5. Warnings and Mitigation

The problems associated witii the warnings

and mitigation of the landslide tsunami

hazard are more difficult than for those due

to earthquakes. As can be seen in the

followdng examples, the effects are generally

limited to a few kilometers or a few tens of

kilometers which means that the time for

evasive action is limited to a few minutes.

The hazard for each potential landslide

tsunami must be evaluated. Depending on

the population at risk and the potential time

available once a sUde has been initiated,

various schemes involving monitoring, slope

stabilization, and land use may be employed.

Most mass failures are detectable weU before

the failure, and sometimes years earlier. A

history of such failures in the area or in

similar areas indicates a need for evaluation

of the risk. Research can be done to delineate the risk through modeling, both

physical and numerical. Surveys can be

made of slope stability and rates of

deposition, particularly for offshore areas.

2.0 CASE HISTORIES

The rest of this study will illustrate the

problem with several case histories selected

to show some of the features of this hazard.

2.1 Glacier Related Sources

2.1.1 Oversteepened Fjord Walls

2.1.1.1 Lituya Bay. Liniya Bay is a Tshaped glacially carved fjord approximately

13 km long, the top being two inlets with

glaciers lying in the Fairweather fault plane.

On July 10, 1958 a magnitude 7.9 earthquake

set in motion at least six landshde-induced

tsunamis including one in Lituya Bay. A 30

million m^ sUde was released from 200-1000

m above Gilbert Inlet down slopes of about

40° on the east side of the mlet (Miller,

1960). The impact of the slide on the water

sent a surge across the inlet to the opposite

point of land, clearing the forest to a height

of 524 m. The resulting tsunami swept down

Lituya Bay destroying two of three fishing

boats, killing two people about three minutes

after the wave was first seen. It was 30 to

50 meters high near the middle of the bay.

Figure 2 shows the configuration of Lituya

Bay and the maximum inundation level. Lituya Bay has been the location of repeated

landsUde tsimamis including one in 1936

which appears to have been induced by

excessive rainfall, and others documented by

tree ring dates of new growth areas, although

the exact site and nature of these earher

tsunamis have not be exactiy determined.

2.1.1.2 Norwegian Fjords. About 210

people have been killed by tsunamis

generated by rockfalls and landslides in

Norwegian fjords and glacial lakes of

Langfjord, Loen Lake, and Tafjord. The

tsimamis have had heights of 5 to 15 m with

surges to 70 m on opposite shores.

Earthquakes are not a factor. (Miller, 1960)

At Tafjord, Norway, on April 7, 1934, about one million m^ of rock feU from a maximum

height of 730 m from a 45° slope along a 229 m front. Tafjord has a maximum depth

of 213 m. The tsunami was about 12 m high

at opposite ends of the fjord and 1 m at Nordal about 50 km away. Figure 3 shows

the location of the sUde and nmup heights.

The tsvinami is beUeved to have been

triggered by melting and freezing of water in

the fracture (as such rockfalls are most

common in the spring and fall). The

beginning of such failure may be detectable

several years in advance. (Peck, 1978)

2.1.2. Glaciation Outwash Deposits

2.1.2.1 Prince William Sound. A magnitude

8.6 earthquake on March 28, 1964, generated

a major tectonic tsimami observed around the

Pacific Basin and triggered many local

tsunamis mosdy from submarine landshdes

from glacial outwash sediments. Almost 80%

of the tsunami-caused fatalities in Alaska and

most of the damage was due to these local

landsUde tsunamis. In Seward, Alaska, a

section of the waterfront 1070 m long and 91

meters wide slid into the bay. The material was water-soaked alluvium with an offshore

slope of 30° to 35°. A nine meter wave

returned in about two minutes. Eight

separate local tsunamis caused a total of 71-

337
82 fatalities with waves up to 30 m. The

fatalities could have been much higher but

this is a sparsely inhabited region. (Cox and

Pararas-Carayannis, 1976)

2.1.3 Glacial Ice

2.1.3.1 Fallen Glacier. A glacier in a hanging valley in Disenchantment Bay,

Alaska, apparentiy suddenly surged out of its valley and fell 305 m down a 16° slope into

Disenchantment Bay without an obvious

trigger. The ice fall generated waves 37 m

high as far as four km away. Indian legends

report that several hanging glaciers had

surged in a similar maimer in earlier times.

Collapse of glacier tongues can also generate

large local tsunamis. (Tarr, 1912)

2.2 Volcano-Related Forms

2.2.1 Pyroclastic Deposit Failure

2.2.1.1 Shimabara, Japan. In 1792 of a 535

million m' portion of Unzen Volcano

collapsed. TTie collapsed area was located some 4 km from vents which had been active a month or two earlier. The collapse and

resulting landslide was probably caused by a

strong swarm of volcanic earthquakes or an

explosion. The landsUde-generated waves

reached a height of about 10 m and caused

devastation for a distance of 77 km along the

Shimabara peninsula. Fifteen thousand

people were killed or disappeared without a

trace and 6,200 houses were destroyed.

(Soloviev and Go, 1974)

2.2.2 Lava Hows into the Sea

2.2.2.1 Mauna Loa, Hawaii. On October 2,

1919, an eruption of Mauna Loa produced a stream of lava which reached the sea near Alika on September 30 creating unusual waves and boiling water and creating a delta. Many tourists were attracted to tihie sight.

The submarine extension of the flow became

unstable in the marine environment and

collapsed creating a 4.3 m wave which

caused minor damage and swept several

people out to sea. They were all rescued.

(Cox and Morgan, 1977)

2.2.3 Collapse Accompanying Eruption

2.2.3.1 St. Augustine I., Alaska. An

explosion on October 6, 1883, of the summit

of the St. Augustine volcano generated a

debris avalanche and tsunami which nearly

destroyed the community of EngUsh Bay, 85

km to the east. A wave 7.6 to 9.1 m high

reached the community about 25 minutes

after the explosion. Fortunately, the wave

arrived at ebb tide. The tidal range here is 4.3 m, so only minor damage occurred.

Figure 4 shows the location of the slide area on the volcano and the location of English

Bay relative to the volcano. (Kienle et al., 1987)

2.3. Submarine Canyons

2.3.1 Grand Banks Earthquake. A

magnitude 7.2 earthquake on November 18,

1929, was associated with a massive sUde

down the Laurentian Channel at the Cabot

Trench. Hasegawa and Kanamori (1987)

believe the volume of the material, 550 x 10* m\ is sufficient to account for the observed

earthquake. The resulting tsunami may have

been 15 m on the Newfoundland coast where

26 lives were lost. The earthquake and

tsunami occurred at the time an abnormally

high tide was expected and while a gale

storm was in progress. Twenty-eight cable

breaks occurred from the turbidity currents. The landslide could have been triggered by a small earthquake or perhaps was related to the storm. (Keith, 1930, Smitii, 1968)

2.3.2 French Rivera. Two waves

approximately 3 m high struck a 30 km

stretch of the French Riviera Coast from the

Italian border to Antibes, France including

the shore at Monaco and Nice on October 16,

1979. The waves caused considerable

damage and 10 deaths. Evidence suggests

that the waves were caused by the failure of

a ground fiU intended to extend an airport

runway or by two much larger submarine

slides that apparentiy occurred on the

continental slope, perhaps triggered by the

landfill slide. Turbidity currents continued

beyond the slides and cut submarine cables at 80 and 100 km offshore of Nice. Figure 5

shows the location of the canyon, sUde and

affected community.

338
2.4. Reservoirs

2.4.1 Vaiont, Italy. The Vaiont Dam in

Italy, the second highest dam in the world at 265.5 m, was completed in the fall of 1960.

On October 9, 1963 the worst dam disaster in

history occurred when 240 miUion m' of

Mount Toe slid into the reservoir completely

filling it over a length of 1.8 km and a depth

150 m above the original reservoir level. The surge wave overtopped the dam by 100

m and was still 70 m high downstream at the

entrance to the canyon. The engineers had

been aware of the slope instability since June

and several days prior to the disaster reaUzed

that the affected area was larger than

originally thought. They had begun to try to lower the water level in the reservoir and had

thought about evacuating the dovmstream

communities. It had been raining heavily in

August and September hampering the effort

to lower the reservoir, and was raining hard

when the landsUde occurred. This, the

largest flood disaster ever, IdUed about 2000

people. (Miiller, 1964: Kiersch, 1965)

Figure 6 shows the location of the sMde and

the affected communities.

Less well reported was a tsunami which

traveled upstream demolishing homes and

struck the town of San Martino with fuU

force. There were 158 fatahties in Casso and

Erto. The toe of the slide was under the

level of the reservoir; the ground water level

had risen due to the impoundment.

Geologically the area was unstable with steep

slopes due to earUer glacial action and recent

river erosion. The bedding planes dipped

steeply toward the river valley. (QuarentrUi,

1979)

2.4.2 Franklin D. Roosevelt Lake.

Between April 8, 1944, and August 19, 1953,

eleven landsUdes occurred in terrace scarps

underlain by bedded unconsoUdated deposits

in the Reed Terrace area near Kettle Falls,

Columbia River VaUey, Washington. Narrow

segments of the scarp on slopes averaging

about 23° suddenly gave way and sMd into

Franklin D. Roosevelt Lake. Debris came

down from a maximum height of 63.6 m

above water. Waves were generated by at least 11 different shdes; the largest wave rose

to a maximum height of 19.7 m on the

opposite shore and was observed 10 km up

the lake where boats and barges broke loose

from the dock. Observed velocity of one

series of waves was about 75 km per hour.

(Miller, 1960)

3. CONCLUSION

1. The term "tsunami" should be broad

enough to include waves which meet the

following criteria: spontaneous generation,

traveling gravitational water waves with

velocities defined as a fimction of the water

depth. This definition includes landshdegenerated "water waves" not previously

classified as tsunamis although physically

there is no difference once the waves have

been generated.

2. Landslide-generated waves are most

common in four environments: glacial and

glaciated area, areas of volcanic activity,

submarine canyons, and reservoirs. Locally

they can be disastrous and collectively they

are a significant hazard.

3. Since the hazard is limited in area and in

available response time, special warning and

mitigation efforts tailored to the individual

site are needed.

4. REFERENCES

1. Beminghausen, W.H. (1968). "Tsunamis

and Seismic Seiches Reported from the

Western North and South Atlantic and the

Coastal Waters of Northwestern Europe,"

Informal Report, Naval Oceanographic Office,

Washington, D.C. 20390, 48 p.

2. Cox, Doak C. and Joseph Morgan. (1977).

Local Tsunamis and Possible Local Tsunamis

in Hawaiian Islands. HIG-77-14, Hawaiian

Institute of Geophysics, 118 p.

3. Cox, Doak C. and George Pararas- Carayannis. (1976) Catalog of Tsunamis in

Alaska. SE-1, United States Department of

Commerce, National Oceanic and Atmosspheric Administration, 43 p.

4. DaU, W. H. (1883). United States Coast

and Geodetic Survey Pacific Coast Pilot. Part L p. 202-204.

339
5. Hasegawa, H.S. and H. Kanamori. (1987).

"Source Mechanism of the Magnitude 7.2

Grand Banks Earthquake of November 1929:

Double Couple or Submarine Landslide?"

Bulletin of the Seismological Society of

America. Vol. 77, No. 6, p. 1984-2004.

6. Heck, N.H. (1947). "List of Seismic Sea

Waves," Bulletin of the Seismological Society

of America. Vol. 37, No. 4, p. 269-286.

7. Johnston, J.H.L. (1930). The AcadianNewfoundland Earthquake of November 18. 1929. Proceedings of the Transactions, Nova

Scotia Institute of Science, Halifax, Nova

Scotia, Vol. 17, p. 223-237.

8. Kiersch, George A. (1965). "The Vaiont

Reservoir Disaster," Geologic Hazards.

Mineral Information Services. State of

California, Sacramento, p. 129-138.

9. Keith, Arthur. (1930). "The Grand Banks

Earthquake," Earthquake Notes. Eastern

Section of the Seismological Society of

America, Vol. 2, No. 2, p. 1-5.

10. Kienle, Jurgen, Zygmunt Kowalik, and

T.S. Murty. (1987). "Tsunamis Generated

by Eruptions from Mount St. Augustine

Volcano, Alaska," Science. Vol. 236, p. 1442-1447.

11. Miller, Don J. (1960). Giant Waves in

Lituya Bay. Alaska. United States Geological

Survey Professional Paper 354-C, United

States Government Printing Office,

Washington, D.C., 83 p.

12. Muller, Leopold. (1964). The Rock SUde

in the Vaiont Valley," Felsmechanik und

Ingenieur Geologie. (Rock Mechanics and

Engineering Geology) Springer Verial Wien,

Vol. 2, p. 148-212.

13. Peck, Ralph B. (1978). "Laurits Bjerrum:

Contributions to Mechanics of Rockslides,"

Chapter 3 in Voight, Barry (ed.) RocksUdes

and Avalanches. Elsevier Scientific

Publishing Co, Amsterdam, Vol. 1, 833 p.

14. Quarentilli, E.L. (1979). "The Vaiont

Dam Overflow: A Case Study of Extra

Community Responses in Massive Disasters,"

Disasters. Vol. 3, No. 2., p. 199-212.

15. Smith, W.E.T. (1968). Earthquakes of

Eastern Canada and Adjacent Areas.

1928-1959, PubUcations of the Dominion

Observatory, Ottawa, 121 p.

16. Soloviev, S.L. and Ch.N. Go. (1975). A

Catalogue of Tsunamis of the Eastern Shore

of the Pacific Ocean. Academy of Sciences of

the USSR, "Nauka" Publishing House,

Moscow, 204 p. (Translation by Canada

Institute for Scientific and Technical

Information, National Research Council,

Ottawa, 285 p.

17. Soloviev, S.L. and Ch.N. Go. (1974). A

Catalogue of Tsimamis on the Western Shore

of the Pacific Ocean. Academy of Sciences of

the USSR, "Nauka" Pubhshing House,

Moscow, 310 p. Translation by Canada

Institute for Scientific and Technical

Information, National Research Council,

Ottawa, 439 p.

18. Spaeth, M.G. and S.C. Berkman. (1972).

"Tsunami of March 28, 1964, as Recorded at Tide Stations and the Seismic Sea Wave

Warning System," in The Great Alaska

Earthquake of 1964. Oceanography and

Coastal Engineering, National Academy of

Sciences, p. 38-1 10.

19. Tarr, R.S. (1909). The Yakutat Bay

Region. Alaska: Physiography and Glacial

Geology. United States Geological Survey

Professional Paper 64, pt. 1, 144 p.

20. Tarr, R.S. and L. Martin, (1912). The

Earthquakes at Yakutat Bay. Alaska, in

September 1899. United States Geological

Survey Professional Paper 69, 135 p.

340
Figure 1. Locations of landslide-generated tsunamis.

Table 1. -IDENTIFIED LANDSLIDE-GENERATED TSUNAMIS

CAUSING FATALITIES

DATE LOCATION MAX RUNUP SETTING* DAMAGE FATALITIES

(SURGE)

(m)

1596 09 04 SEIKAIDO, JAPAN 5.0 V? ISLAND SUN 708

1597 09 10 BEPPU BAY, JAPAN 2.0 V? SOME 40

1700 04 01 KYUSHU, JAPAN 2.0 7 300 HOUSE 1000

1741 08 29 HOKKAIDO, JAPAN 90.0 V 800 HOUSES 1600

1756 02 22 LANGFJORD, NORWAY 40.0 O BUILDING & BOATS DESTROYED 32

1792 05 21 SHIMABARA, JAPAN 11.0 V 6200 HOUSES 15030

1815 11 22 JAWA (JAVA), INDONESIA 3.6 V SOME 1200

1853 00 00 LITUYA BAY, AK 37.0 o SOME MANY

1871 03 02 N. MOLUCCA IS. 25.0 V MUCH DAMAGE 277

1883 08 27 JAVA SEA 35.0 V TOWNS DESTROYED 36000

1888 03 13 BISMARCK SEA 60.0 V VILLAGES DESTROYED MANY

1899 09 10 S.E. ALASKA (60.0 G VILLAGE DESTROYED SOME

1905 01 15 LOEN LAKE, NORWAY (40.0) O BUILDINGS & BOATS DESTROYED 61 1928 08 04 FLORES SEA 10.0 V SHIPS SUNK 128 1929 11 18 GRAND BANKS, NFLND. 4.6 SC CABLES BROKEN 28

1934 04 07 TAFJORD, NORWAY (62.0) G EXTENSIVE 44

1936 09 21 LOEN LAKE, NORWAY (70.0) G EXTENSIVE 73

1946 06 23 B. C, CANADA 30.0 SC? CABLES BROKEN 1 1958 07 10 S. ALASKA (525.0) G 2 BOATS SUNK 2

1963 10 09 VAIONT, ITALY (240.0) R SOME 159' 1964 03 28 GULF OF ALASKA 67.0 G MUCH DESTRUCTION 71-82

1971 03 18 CHUNOAR, PERU (30.0) G CAMP DESTROYED 500

1979 07 18 LOMBLEN I., INDONESL\ 10.0 SC? 4 VILLAGES 540

1979 10 16 FRENCH RIVIERA 3.0 SC CONSIDERABLE 10 DEATH TOTAL APPROXIMATELY 58,000 •KEY: G = GLACIAL AREA, V = VOLCANIC AREA, SC = SUBMARINE CANYON, R = RESERVOIR

'Number due to tsunamis of total of about 2000 killed in the flood.

341
Figure 2. Map of Lituya Bay showing setting, location of boats, and trimline caused by the 1958

tsunami.

Figure 3. Tafjord showing wave heights in meters produced by the rock slide of 1934.

342
0 5 10 km

Figure 4. Map showing the location of the sUde area on the St. Augustine volcano and the

location of English Bay relative to the volcano. .

343
Figure 5. Location of the Grand Banks slide, cable breaks, and areas affected by the tsunami.

Figure 6. Vaiont Reservoir sUde showing slide area and communities affected by the wave.

344
A Study on the Effect of Tsunami Breakwater

by Numerical Simulation

Chiakl goto" and Katsutoshi Tanimoto*'

Abstract

In order to estinate the tsunami reduction

effects of bay-mouth breakwater, the numerical

simulation assuming the 1960 Chilean Earthquake

Tsunami las carried out upon the Ofunato bay in

Sanriku Coast. Through the comparison of the numerical

results Kith and without bay-mouth breakwater,

it was concluded that the bay-mouth breakwater

was able to reduce the tsunami height to half

order.

|(ey Words : Tsunami. Numerical Simulation. Baymouth Breakwater. Tsunami Reduction Effect.

1. Introduct ion

The Sanriku coast had been attacked by

tsunami many times. in the recent 100 years,

Meiji Great Sanriku Tsunami occured in 1896.

Syouwa great Sanriku Tsunami in 1933 and Chilean

Earthquake Tsunami in 1960 caused serious damage

to Sanriku Coast. The 1896 and 1933 Great

Sanriku Tsunami were the huge tsunami and run-up

on the land more than 30 meters hight. The 1960

Chilean Earthquake tsunami propagated on the

Pacific ocean from South America.

As a tsunami defence work, bay-mouth breakwater with 736 meters long was constructed in

Ofunato Bay in 1968. When it was planed and

examinated. the hydraulic model experiments or

theoretical analysis were carried out. However,

the problems of scale effect and initial tsunami

wave profile remained to be unsolved. It was

caused by that the ability of computer was not

enough and the Investigation on tsunami was not

fully developed.

Then, In this paper, the tsunami reduction

effects of bay-mouth breakwater was examined by

use of recent developed technique of tsunami

\tr i iM* E 110* E lie iw I ivr t iw I tor t m" t U X1 r

k

^.^

V

y /(

w $

ur s

Figure 1 Computational region of trans-oceanic propagation

Figure I Compuatlonal region near Sanriku Coast

1) Chelf, Ocean Energy Utilization Lab. PHRl

2) Directer, Hydraulic Engineering Dlv.,PHRl

345
nunerical sioulation.

2. Tsunami Nunerical Simulation

2. 1 CoBPUtat ional Region

The numerical simulation are carried out for

the Chilean Earthquake tsunami occured in 1960. The simulation is divided into tvo part; trans- oceanic propagation and near Sanriku Coast pro- pagation of tsunami.

The computational region of trans-oceanic

propagation is shown in Fig. 1. The region

includes all over the Pacific Ocean surrounded

by Far Eastern Asia. Australia, North America

and South America, which extends from 60 degrees

south to 60 degrees north in latitude and from

120 degrees east to 170 degrees west in longitude. The computational grid size was 10 minutes in latitude and 10 minutes longitude

(about 111 kilometers near the equator).

The computational region near Sanriku Coast

is shown in Fig. 2. The region is extended from

the Nemuro Peninsula in Hokkaido to the coast of Miyagi Prefecture and include the area from 36 degrees and 20 minutes to 43 degrees in north

latitude and from 139 degrees and 40 minutes to

145 degrees and 20 minutes in east longitude.

This area is further divided into six subregions. The grid size in the deepest area is

set 5. 400o, and is made small in shallow sea

area up to 1,800m, 600b, 200m, 100m. and SOm.

2.2 Governing Equations of Tsunami in Deep Sea

for the governing equations of nunerical

Simula t ion. The earth is considered as the sphere with

the radius R. the latitude and longitude

coordinates are denoted by (X, ij)) and discharge

flux in each direction by (M,N). The governing

equations'' of tsunami on trans-oceanic

propagation are expressed as follows:

_L

(U RcosX RcosX ai|> 3 [-^F.]

(1)

(2)

(3)

F3 RcosX h\.Bx [ ;i.;iJ"cosX) + 1,

2. 3 Governing Equations near Sanriku Coast

In the computation near Sanriku Coast, two types of tsunami theories in the plane

coordinates are used. The following nonlinear long wave theory

(shallow water theory) with the nonlinear term

and the friction tern are used for the computation near the coastal region which contains

the runup on a dry land.

^t ^x dy

(4)

JUL

at dx^h + n^ ('y^ h + n

] + g(h+ri)

ClSl

znL.

MQ (5)

In the problem of tsunami generated in Japanese coast, the computational area considered is about 100km width and 100km length, thus the Cartesian coordinates system is used usually.

However, in this problem, the tsunami is generated in Chile, propagates over 10,000kn and attacks Japan, thus the spherical coordinates

system must be employed. Moreover, because the Coriolis' force and dispersion effect could not

be neglected in the tsunami propagation in deep

sea. The linear Boussinesq theory must.be used

rJi * ayfh + r,J *

f)x^ h + r, ^ ^ ^(h+r,).^^

+ ^

NQ = 0 (6)

where (x,y) denotes the special coordinates, t

the tine coordinate, n the elevation of tsunami

surface from the still water level, h the still

water depth, g the gravitational acceleration,

n the Manning's roughness parameter, (M.N) the discharge flux in the (x.y) direction, and

Q = (M^ +N2 .

346
The shallow water theory stated above is used

for the tsunami simulations near the coastal

line. On the other hand, in the sea area with

large depth, the easier equations can be employed. Because the assumption of small amplitude holds, nonlinear tern in the shallow

water theory can be neglected. Moreover, the bottom friction say not have a large influence

on wave nation. Thus, in the coastal area except in the vicinity of coastal line, the following the linear long wave theory can be

used.

at

'

at ^ay

(7)

(8)

(9)

2. 4 The Initial and Boundary Conditions

As the initial condition of tsunami numerical

simulation, the vertical component of deformation of sea bottom estimated by the earthquake dislocation model is given to the wave surface prof i 1 e. When the strain accumulated in the earth

reaches the limit, the dislocation is generated

and the crust of both side slip suddenly. This

phenomenon is the dislocation motion which

generates the earthquake. When the dislocation

motion happens below the sea bottom, the upheaval and subsidence of sea bottom occurs and tsunami is generated.

The dislocation motion can be described by

the easy fault model. The distribution of vertical displacement of the sea bottom near the fault can be calculated by the method of Mansinha and Smylie'' using the given fault parameters. The vertical displacement obtained _

is added to the water surface and used for the initial condition of tsunami.

For the boundary condition of runup, the method of Iwasak! and Mano^' is used, in which

the topography near the wave tip is considered

to be like stairs. When the water elevation at the sea side grid point is higher than the ground elevation at the land side grid point,

the difference of both elevations is used for the total depth in the computation.

3. Tsunami Reduction Effect

3. 1 Reproducibility of Numerical Simulation

The numerical simulation results of Chilean

Earthquake Tsunami in Ofunato bay are shown in Fig. 3. Figure shows the distribution of maximum

water elevation by contour lines and the values

at some grid points. The values denotes the computed tsunami height in meters and the values

in ( ) denote the measured tsunami heights.

Figure 4 shows the comparisons between the computed and measured tsunami height. Aida'"

proposed the parameters K and k defined by the following equations.

logK = (1/n) 1 logK i (11)

log/c = [(1/n) I (logKi)^ - (logK)^]

(12)

where Ki denotes the ratio of the computed value

and the measured one at point i, K the mean ratio of the computed and the measured one, and

K the scattering (corresponds to the standard

deviation in logarithmic values).

In this computation. K=0. 94 and K=1.19. Thus

it is concluded that the computed results agree with the measured values well. However, the computation seems underestimating where the runup height is large and overestimating where the runup height is small. The area where the runup

height is large is the inner bay. therefore, it

is thought that the effect of resonance does not appear sufficiently because the characteristic

of period of incident tsunami is small different

from the actual one. As regards the large

measured values in the bay center and bay mouth,

there are the problems of the accuracy of measurement and that the highest tsunami trace

347
km

3.S0(4.3)

288(3 0)

2.79(3 7) 270(2 6)

306(2.4 )

333(4.1 )

3.92(3.9) 389(3 I ) 388(4.2)

4 08(4.4 ) 393(5.0)

399(4 6) 365(4.3)

388(4.7 )

2.95(3 9)

I 55

0 I 2 km

I—I I I I I

'

Figure 3 Maxinun tsnami heights

( without breakwater)

Figure 5 Maximum tsnani heights

( with breakva ter)

Computed

tsunam i height ^

0 1 2 3 4 5

Measured tsunami height

Figure 4 Comparison between numerical

and measured data

6 (m)

tend to be measured as the tsunami runup height.

Moreover, because the computed value is the averaged one in 50 meters grid, it is possible

that the local movement of tsunami can not be

reproduced.

3. 2 gffects of Bay-mouth Breakwater

Figure 5 shows the computation results of the

case in which the contemporary topography with

the bay-mouth breakwater is used. When the prediction and the reproduction

computations are compared to estimate the effect

of bay-mouth breakwater, the parameters A and a

are introduced, which are similar to Eqs. (11),

348
(12) and are calculated froa the ratio of tsunaai height vith and without breakwater.

The bay-DOutb breakwater reduced the tsunaai

heights in the bay to half order ( A=0.47.

a=l. 12 ), the BaxiDun tsunaoi height becones 2 meters. Inundation area are restricted to the

low ground area, for exanple. the Chayanae down

town, reclaiaed land and Uosoura in Ofunato

Fishing Port.

4. Conclus ions

Froa the tsunani nuoerlcal siaulations, the tsunaai heights are analyzed, and the reproducibility of computation and the effect of bay-aouth breakwater are considered. The aajor

conclusions are as follows:

(1) The tsunaai nuoerical siaulation gives good

results with the acceptable error 10%.

(2) The bay-aouth breakwater is able to reduce

the tsunaai heights by half in the Chilian

Earthequake Tsunaai.

References

1) Goto, C. : A study on numerical siaulation

for trans-oceanic propagation of tsunami, PHRl

rept.. (To be published).

2) Mansinha.L. and D.E.Soylie: The displacement

fields of inclined faults ; Bull. Seisn. Soc.

Am., vol.61. No. 5, pp. 14 3 3-1 440. , 1 97 1.

3) Iwasaki, T and A. Mano : Nuaerical siaulation

of two dimensional runup on land in the Eulerian

description, Conf. of Coastal Eng. in Japan, Vol

26. pp70-7 4, 1 97 9.

4) Aida. i. : Reliability of a Tsunami Source

Model derived froa Fault Parameters; J. Physics

of Earth, vol. 26. pp. 57-7 3. , 1 978.
I

1

(

i

'I

!
Theme IV

Joint Cooperative Research Program

The U.S.-PRESSS Program

M.J. Nigel Priestley

General

Precast Seismic Structural Systems have been chosen by the U.S. and Japan as the topic for

the 4th phase of the U.J.N.R. Panel on Wind and Seismic Effects Large Scale Testing Program.

U.S. funding for an initial three year program was approved by the National Science Foundation

in May 1990, and the project has proceeded very rapidly from that time.

Research Emphasis

The U.S.-PRESSS project aims (1) to develop comprehensive and rational design

recommendations based on fundamental and basic research data which will emphasize the

viability of precast concrete construction in seismic zones, and (2) to develop new materials,

concepts and technologies for precast concrete construction in seismic zones.

This second aim also indicates the emphasis and direction of the U.S.-PRESSS research

project. It is felt that the greatest challenge and promise for precast structural systems in seismic zones is to develop new and innovative technology appropriate to the construction constraints.

The approach of designing precast structures to emulate monolithic reinforced concrete

structures has already been established by earlier experiments and design studies, and is felt to require less research effort. Further, it is doubtful if emulation of monolithic reinforced concrete

structures is economically viable for precast systems in the United States.

Research Projects

The following projects are currently funded in Phase I of U.S.-PRESSS. It will be seen that

the emphasis is on design studies and analytical development. Experimental research is currently

being carried out in an associated study at National Institute for Standards and Technology, but

will also be the focus of a second phase of the U.S.-PRESSS project, expected to start in 1992.

Project 1.1 - Concept Development: Development of new concepts for precast concrete

construction for seismic zones focus on innovative structural solutions and the incorporation of

advanced technology to ensure relevance for the needs of the 21st century. A series of seminars

involving practicing designers, precasters and contractors has already been held to promote

discussion and development of new concepts. Conclusions from these seminars will be available

shortly. From these preliminary studies, a selection of systems which have been established as economically and technically viable will be chosen for further experimental and analytical study.

353
Project 1.3 - PRESSS Connection-Classification and Modeling: An important task for future

analytical and design studies is the development of a classification scheme for generic PRESSS

buildings based on connection type and function. This involves a thorough review of existing

research and design information, and original work to predict the characteristics of innovative

and currently untested connections. There is considerable cooperation with the

Japanese-PRESSS group in this area, and also with Project 1.1, since it has been found that

strong interaction between Projects 1.1 and 1.3 is essential to the success of both studies.

Project 1.4 - PRESSS Analytical Platform DeveSopment: The special needs of dynamic

inelastic analyses of PRESSS are being met by developing advanced analytical programs based on the DRAIN-2D architecture. Both 2-D and 3-D versions, based on fiber constitutive models,

and capable of reproducing the deformation characteristics of precast connections are being

developed. These programs will be used in analytical studies to predict structural response in the

different seismic zones of the U.S.

Project 1.6 - PRESSS Preliminary Design Recommendations; Current seismic design codes

for precast concrete structures are being reviewed, and design philosophy for the final design

recommendations formulated. Prescriptive design requirements for 'strong-joint' (i.e., monolithic

reinforced concrete emulation) are being based on existing information, and those for ductile

joints are being based on the results of Projects 1.1, 1.3 and 1.4. The format of the

recommendations is expected to be strongly influenced by structural reliability theory.

Project 1.7 - PRESSS Coordination: This project provides the umbrella under which the

separate projects listed above are interrelated. Regular coordination meetings involving the

entire PRESSS research team are held twice a year, and are attended by researchers working in

allied fields not specifically funded in the PRESSS project. The coordination between U.S. and

Japanese research teams is also an extremely important part of this project. The U.S. team is desirous of having closer interaction between the U.S. and Japanese teams than has been possible

in previous U.J.N.R. joint studies. As a means toward closer cooperation, both teams have have

agreed on a standard 'test-bed' building, to enable different design philosophies and details to be

compared. Commitments have been made toward producing joint documents on design

recommendations and connection classification.

The first U.S.-Japan Joint PRESSS meeting, held in San Diego in November/December

1991, was extremely productive and effective in terms of cooperative effort.

354
Summary

Although the U.S.-PRESSS project has only been in operation for 12 months, it is already

producing useful information in the form of preliminary reports on structural concepts,

connection classification, and a draft version of the analytical platform. Cooperation with the

Japanese team has been effective and rewarding.

Although the scope of Phase I of the U.S.-PRESSS program is comparatively modest, it is expanding by association with the NIST experimental program and with the ATLSS building

systems program. It is expected that further experimental projects will be added to U.S. PRESSS

early in 1992.

355

U.S.-Japan Cooperative Research Program Precast

Seismic Structural System

by

Shin Okamoto.Shinsuke Nakata.Takashi Kaminosono

Hitoshi Shiohara and Mizuo Inukai

Part 1 Design Trial of 15 story Precast Reinforced Concrete Buildings

1. Introduction

In 1990, Japan side did three cases of trial design

of 15-story precast concrete structures using

current Japanese building codes and design re- quirements. Here one of three design trials is introduced as case 1

Case I consists of moment-resisting concrete

frames without structural walls. The dimensions

are 30 m X 60 m in plan (6 m x 10 m in bay size),

57 m in height (3.8 m in story height).

Common designed conditions are as follows:

(1) no-basement, but with sufficiently rigid footing beam

(2) seismic zone factor =1.0

(3) medium grade ground (soil period Tc= 0.6

second).

The precast elements in the designed buildings

are used for columns, beams, semi-finished slabs

and spandrels.

This building is designed in a manner the same as a monolithic structure, because the precast ele- ments are connected rigidly with in-situ concrete. Therefore, the design is based on "AIJ

Standards for structural calculation ofreinforced

concrete structures" and related standards. The

strength of concrete is 360 kg/cm2, the uppermost value allowed in the standards.

High-rise precast concrete buildings in Japan

have a short history. The precast concrete struc- ture was introduced to Japan more than 30 years

ago; only a few precast concrete buildings were

constructed as compared with other types of

construction. A major reason may be that ductile

frame buildings such as steel frame or steel- framed reinforced concrete are preferred in Ja- pan attacked frequently by earthquakes; and the ductile performance of precast concrete construction has not been clearly understood.

Another factor is that the construction cost of

precast concrete structures is higher than cast-insitu reinforced concrete structures. Accordingly,

precast concrete structures which are assembled

by their units have not been used positively up to

this time.

With a rapid development of experimental re- search on joint performance and joint construction methods to assure acceptable performance,

not only high-rise reinforced concrete structures but also high-rise precast concrete structures have been constructed in the last several years. Furthermore, the lack of labor power caused the

development of construction of precast concrete

structure; Through the study of the three buildings on

precast concrete, we believe that the designed

buildings satisfy the requirements in serviceability and aseismic performance, and contribute to reduction of construction term, building cost. Outline of designed building, material design

concept and precast joint system are here mentioned.

Fig. 1 shows the joint concept of precast members integration . In the beam-column joint re- gion , longitudinal reinforcements of columns

and beams are exposed from these members and

are casted in rigid body by post casted concrete.

Detailed section of column and beam are shown

in Fig.2. Hatched parts show the precast concrete in this figure. Fig.3 shows the detailed ar- rangement of reinforcements in comer column.

The longitudinal reinforcement of beams are anchored in the beam column joint with 90

degree hooks.

Fig.4 illustrates the procedure of construction set up of precast members. In the step 1 , precast col- umns are jointed at column base through sleeve

joints. Step 2 shows the setting beam members

from both directions. In step 3, half-precast slabs are installed and top reinforcements of slab are arranged. Step 4 indicates pouring cast-in-situ

Building Research Institute ,MOC

357
concrete at slab surface, beam top and beamcolumn joint. Size of designed column section

is I m by I m at lower stories, and 85 cm by 90

cm in lower beams.

Table 1 shows the defined condition of criteria for "ductile" members. In columns, the limitation of axial force ratio, shear span ratio and

longitudinal reinforcement ratio are defined. In beams, applied shear stress ratio is limited. According to the a check of this criteria to this design example, this building is estimated as ductile structure.

3. Results of Structural Analysis

Fig. 6 shows calculated skeleton curves of horizontal force - story drift relationship in each

story. ,Using this mathematical model, elastoplastic frame analysis showed around 2.0 cm

story drift in each story due to design shear force

(CB=0.168). And due to 60 percent higher

design shear force, responsed story drift reached

to hinge mechanism in each story. Fig. 7 is the

results of dynamic response analysis to four

input earthquakes. Upper two graphs are in the case of 25 cm/sec max. velocity input. Lowers

are in case of 50 cm/sec input. The results of

responsed story drift due to 25 cm/sec input,

show 1.53 cm max. story drift. Incaseof 50kine

input, the max. response show due to Hachinohe

Earthquake and its responsed shear force is almost same as that of failure mechanism of structure. This means that the designed building is enough

safe to big earthquake.

4. Conclusion

In this report, an example of three design trials are introduced and discussed its seismic safety

based on the existing Japanese seismic design

criteria. In order to develop more grade ofdesign

freedom, it is recommended that prestressed

concrete system should be much more used.

Through this project results from US side and

Japan side, Japanese seismic design standard and

design manual of precast joints are to be established and next year this designed building will

be discussed its seismic safety based on these

new research products.

358
Building

1-1 Outline of Building

1) Plan

2) Elevation

3) Number of Stories

4) Type of Structure

External wall

Internal wall

Logitudinal Direction:

1 0 @ 6 . 00m= 60 .00m

Transverse Direction:

3 @ 1 0 , 00m= 30 . 00m

Total Building Height 57m

Story Height 3.8m

1 5 (No-basement)

Precast reinforced concrete frame

Precast concrete panel cladding

Board (fireproof board) partition walls

5) Usage Office

1-2 Materials

1 ) Concrete

• Precast Concrete Normal Concrete

For Column, Beam and Slab

• In situ concrete Normal Concrete

• Precast wall panel Light Weight Concrete

Fc= 360 kg/cm"

Fc= 360 kg/cm2

Fc= 300 kg/cm^

2) Reinforcement

• Column and Beam

For Hoop, stirrup

• Secondary beam and Slab

For stirrup

Deformed bar SD 40

fy=40,000 kg/cm«

High strength deformed bar

f y ^ 13 ,000 kg/cm'

Deformed bar SD 35 SD 30

Deformed bar SD 30

359
?ra:ni3 DoUib

PrQcast Ua^n unit Procaat beain unit

/

- \. / —

y Precaal column \

^—In-aitu concrete I 1

1

/-SpUca sleeve joint

/—Filled ^ith /—Beani-Beoxa joint

y awrtar / Un-situ concrete)

/ 1 i ^2F.L.~15F.I

1

1 Precast beam I ^!

Precast beam '• Precast beam

\

^ Precast column

1

r

1

i ^

^ \—Beam-Column joint ^

(In-situ concrete)

Duct for reinforcement

(Filled with ortar)

71F.L,

Filled with BortAT ^—Filled with nortar

X DIRECTIOII FRAHE

-Beam-Column joint

(In-aitu concrete)

•In-situ concrete

Precast column

Splice sleeve joint

Filled with mortar V2F. L.-15F.L

\

Precast beam Precast beam

-Conection part of

Secondary beam

-Precast column

LJ

II I

for reinforcement

(Filled with iwrtar)

•Precast column

VlF.L.

lied with mortar -Filled with mortar

Fig. 1 Frame Details

360
Column DoUila

(High strength

deformed bar)

Mala bar-^

B

•Spiral hoop

0 P

'c

a

c

c a Q C c Duct for rolnforcemont

Duct through type Splice sleeve type

1 story column 2~15 story column

Beara Details

X direction beam Y direction beara

In-situ ^ concrete L

.Precast beaa

Conatniction jointq u u u

0

•3>

•;Half precaat slab —

a

-a

•Spiral stirrup

(High strength

deformed bar)

In-situ ^ concretelv^

Precast beaa

Construction joint

—

f

. slab Q

9

Spiral stirrup

(High strength

deformed bar)

Anchora;^e of main reinforcements of beam

In-situ concrete

-O 0 4 O :7/ ecast

'

Precast Column

Beam-Beam joint

Enclosure welding

Precast beaa

In-aitu concrete

Fig.2 Details of Columns and Beams

361
Detail of Cornor boara-co liimn joint

115 'flO CcT)

S5Q

M UO Ifl 150

I

' ilil I'M,

I Vi 111'''

^1' !i !

,11 II jj

115 Precast beoin

beam-column joint

In-situ concrete

O

VI 11 1-1 11 i

i J: ' " UJJU

m: B

—•<- in

I0 O

90 110 fOO liiOpQ

I.O'OO

Precast beara

PLAN

Bo|loj 630 raw:

1.050 e

Precast colunin

< > 1 1 >

' Ct Cl Cl

1

i Cl

> Cl Cl

s Cj C| C| C| Cl Cl ,C| C| C| C|

CD

en od C) Ci Cl ,Cl ,C| Cl Cl ,C| Cl

cn

Cl ,Ct Cl Cl Cl ,Cl Cl Cl

<

Precast beam KEY PLAH

Precast column

Fig.3 Detail of Corner Beam Joint

362
Construction Procedure

® Setting precast concrete Column

® Setting precast concrete beam and half precast concrete slab

(D Arrangement of beam reinforcement, floor slab reinforcement,

beam-column joint reinforcement

® Placing concrete into beam-column joint, on the half precast

beam, and on the half precast slab

Fig. 4 Construction Procedure

363
Table 1 Required Condition for Ductile Members

For Column

1 1 emi

Limltillon of

RAIIK A Designed value Eva I ua 1 1 on

Coll apse node

Should be

flexural yielding

All colunns are

flexural yielding

A

Axial stress at 35

A 9847*

X-directlon B 494*

C 474*

( 91X)

(4.GX) . A

Co 1 1 apse nechan i sn A 7609*

Y-direciion D 2720*

C

*

( 74X)

- ( 26^)

( »

A

Shear spaa ratio of

1st story coluans

' formiD; hloies

2H/Q-D^2.

5

nore than 3.97 A

Relnf orceDent ratio

for bendlnc noneot

0. 545( to 0.80X A

Heaa shear stress

-at col lapse

DechaDlso

X u/Fc^O. 1 less than 0.099 A

For Beam

Items

Lini tat Ion of

RAHK A Designed value Evaloat Ion

Collapse node

Should be

flexural yielding

All beam are

f lexura] yielding

A

Mean shear stress

at col lapse

nechaal sn

X u/Fc^O. 15 less than 0. 091 A

'RANK A' can be Judged to this frame from synthetic evaluation above nenlloned.

364
(r 41 ir ti 1'.. . 1, ..

ijn. h„ 1 a 1 ^1 1 l^, 1 hi, 1 h„ 1 h.. 1 hi, 1 h„ 1 h„ 1 1 I,. . 1 u

° 1 91 1 hJ h„ 1 h„ 1 hi. 1 h„ 1 h„ 1 h„ 1 1 . II . .1 u

IKT.

1 CI

I

1 uJ [ .:i 1 hJ h„-

1

h„ i hi, 1 h„ 1 h„ 1 h„ 1

1 " 1

1 .11 .1 u

llfL a 1 ^ 1 I 11 i h 1

I 11

J

h„ 1 h„ 1 h„ 1 h„ 1 h„ 1 h„i h„ 1

I CI1 I .1l_l

c< m 1 " 1 1 ti 1 1 91 1 h„ 1 h., 1 hi, 1 hi, 1 hi, 1 h„ 1 h., 1

1 ° 1 1 . 11 1

ct

iin.

1 " 1 1 11.

1

h„ 1 hi. 1 h„ 1 h., 1 h. 1 h., 1 h„ 1 h„ 1

1 1 1 II 1

CI

. ^ a 1 " 1 1 . 11 ..1 h,. 1 hi, 1 hi, 1 h„ 1 hi, 1 hi, 1 h„ 1

1 ^ 1 L 11 .1

1 <^ 1 1 t, 1

C4

1 " 1 1 41 1 h 1

1 . 11 ..1 hi, 1 h„ 1 hi, 1 h„ 1 hi. 1 h„ 1

1 ^ 1 L 1). 1

1 ^ 1 1 .. 91 1

IM.

1 " 1 1 li 1 h 1 1 11 1 h„ 1 h„ 1 h„ 1 h„ 1 hi, 1 h„ 1 h„ 1

1 ^ 1 1. II .1 ux 1 ^ i 1 II 1 h„ 1 h„ 1 h„ 1 h„ t hi, 1 h„ 1 hi, 1 h„ 1

1 <^ 1 1 II 1

. K\ h„ 1 h„ 1 h. 1 h„ 1 h. 1 h„ 1 h„ 1

1 ^ 1

1 91 1

..fa h.. 1 hn 1 h„ 1 hi, 1 h„ 1 h„ 1 h„ 1 hi, 1 h., 1

1 a 1 1 .. 11 1 r\ h.. 1 h„ 1 h„ 1 hi, 1 h,. 1 h„ 1 hi. 1 hJ h„ 1

1 <3 1

1 11 ,

!

C4

h.. 1 1 h„ 1 h., 1 hi. 1 h., 1 l^, 1 h., 1 h„ 1

1 1

1 . 9, 1

C4

5

1 "niA I 1 "rii 1 1 '^ni 1 h,„ 1 hn, 1 h., 1 hn, 1 hm 1 I'^rai. 1

1 ^ 1 1 raiA 1 U

Ull Ull un Ull VIM

<5)

K di

J) <^

r Q c

J

tior

J <^

1 fr amir

i

ig Q

J) <5

lev atic n

i) i

Fig.5 Plan and Elevation of Designed Building
Story niinbor Slorr nunl.xir

0 1.0 2.0 3.0 4.0 0 1000 2000 3000 4000 5000 0000

Inter »tory drlfl(ca) Shear forcfl(t)

Haximura Response Inter Story Drift Maximum Response Sliear Force

X Direction (25cm/sec)

Story nujabor Story nuiabor

0 1.0 2.0 3.0 4.0 5.0 0 1000 2000 3000 4000 5000 0000 7000 8000

Intor story drlfl(ai) Slionr forced)

Maximum nesponso Inter Story Drift Maximum Rosponso Slioar Force

X Direction (GOcm/sec)

Fig.6 Siccleloii Curve of Eacli S1017 and Results of Frame Analysis

366
Slonr 3hear(l)

8.0 10,0 Inter story driftCca)

Story shear-Inter story drift reration

X Direction

Story shear (()

10. 000

8.0 10.0 InUjr slory dx-iftCcn)

Story sliear-Inter story drift reration

y Diroction

Fig.7 Results of Dynamic Response Analysis

367
U.S.-Japan Cooperative Research Program PRESSS

Part 2-1

Tensile Tests of Lap-Spliced Joint

for

Precast Concrete Panel Connections

1.1 Introduction

Tensile force occurring in connections of preccLst concrete frame-wall structure, such as (1) wall panel to boundary column connections, (2) wall panel to boundary beam connections, and (3) slab- to-slab connections are often designed to be trans- mitted at joint by lapped deformed bar wliich are embedded in site-cast concrete. Many researchers have reported the behavior of these joints under

shear stress, although reports on the tensile be- havior are limited. Thus, the provisions in cast- in-place concrete building codes are used to design

of lapped-spliced joint to tensile force. It is aimed

to develop lapped-spliced joint in which the site as- sembly may be simplified and necessary site work

may reduced and in which satisfactory performance

may be assured.

In this paper, results of 17 panels tensile tests are reported and tensile behavior for tlie details which

include (l) the type of splice (2) lap length of splice, (3) space between splice reinforcement, (4) shape of

splice reinforcement, and (5) reinforcement ratio, are discussed.

1.2 Test Program

1.2.1 Specimens

The specimens tested were seventeen full-scale pre- cast concrete flat panels. Their dimensions are 500 mm in width X 1,000 mm in length and 100 or 120 mm in thickness. Figure 1 shows the loading apparatus and the shape of the specimens. The specimens are con- nected at the center. The four different types of lap joint tested are shown in Fig. 2. They include

• straight lap splice joint (straight lap joint),

• loop shaped lap splice joint [loop splice joint),

• straight lap splice which laid on precast concrete panel which embedded in topping concrete (topping concrete lap joint), and a

• splice reinforcement in between full-precast

3egment3(preca5^ segment lap joint).

The test variables are

• panel thickness i (100 mm or 120 mm),

• lap length ( (10 d, 20 d, 30 d, where d the nominal cross sectional area of deformed bar),

• spacing of lapped bar s (60 mm or 30 mm),

• the presence or absence of supplemental rein- forcement, which is laid transverse to reinforce- ment in loop in case of a loop shaped splice

joint only,

dus well as four different lap details. Seventeen specimens and the material properties

are listed in Table 1. SD30 grade steel were used for the splice reinforcing bars. Normal weight concrete was used and the compressive strength of concrete

of site-cast concrete was 384 kgf/cm^.

All of the specimens were placed horizontally and

concrete was cast on it to form an integrated unit. The concrete was left for 15 to 22 days before casting. No preparation on concrete joint was carried out except topping concrete lap joint. The inter- face of precast concrete and topping concrete was

brushed, veneer panel was used for the form for side edges of segment lap splice joint.

1.3 Loading and Measurement

The specimens were loaded with increasing monotonic tensile force until failure occurred. The de- formations were meeisured between two targets on

specimen. The targets were anchor bolts embedded

ill concrete and the distance between them was 600

mm and the distance is nearly equal to the 6-times

of the panel thickness. Average tensile strain was

defined here as the ratio of the deformation to the

distance between target3(= 600 mm). Average tensile stress was defined as the ratio of the total tensile force to the total cross sectional area of splice

reinforcement.

1.4 Test results and Discussions

Figure 3 show the relation of average tensile strain and average tensile stress. The Ultimate strength

368
and the Ultimate strain at strength are listed in Table 3. Observed failure modes are shown in Fig.

5. -

1.4.1 Straight lap splice joint

During the tests of the specimens, the lap length

of which exceeded 20 d ( d; nominal diameter of rebar ), yield of the steel and the failure in the zone

with no lap joint were observed(specimens #1, #2,

#3, #4). Almost of all deformation are able to be

attributed to opening of the crack at the concrete

construction joint. The specimens with thickness of 100 mm and 120 mm had almost the same strength

and ductility. In the case of the specimens which

had a lap length of 10 d, tensile yield of the steel was observed. The average strain was almost 1-

2 %, before sudden rupture at the splice occurred

and the strength suddenly dropped off (specimen

#5'. #6).

The tensile stiffness was almost equal to the stiff- ness of the reinforcing steel by themselves. The

opening of the crack at concrete interface may re- duce the stiffness. The different spacing of lapped

reinforcing bars seemed to have little effect on their strength and ductility.

1.4.2 Loop splice joint

The specimen #7 and #8 had a splice length of

10 d and had transverse reinforcement placed in- side the loop. The splice reinforcement yield ob- served and attained to maximum strength at av- erage strain of 1.0 - 1.3 % . It sustained strength

and a gradual failure observed. In the case of the specimens which had a shorter splice length of 5 <i showed gradual failure at cast-in- place concrete before the reinforcement yielded, while it had a trans- verse reinforcement inside of the loop bar (specimen

#10). The specimens without transverse reinforce- ment showed brittle rupture at average strain of

0.05 %, before yielding of the reinforcement (specimens #9, #11).

1.4.3 Topping concrete splice joint

In the case of specimen wiiich hcuJ a splice length

of more than 20 d showed yielding of the reinforce- ment and a gradual failure at the section occurred

at the point where the truss reinforcement a was

terminated (specimen #12 #13). In the case of a 10 d splice length, just after the yielding of the reinforcement, a sudden failure and reduction of strength was observed at the splice joint. The stifT- ness before yield was generally higher than for the

other types of splice joint due to the tensile force transmitted by concrete of cast-in-place panel.

1.4.4 Precast segment lap joint

When the stress in reinforcement exceeded 51 - 71 % of the yield strength, shear slip at the construction joint between precast concrete panel and cast- in-place concrete occurred, and a sudden reduction

of strength was observed. Bond failure along the

surface of splice reinforcement wets not observed,

(specimen #15 #16 #17).

1.5 Concluding Remark

Tensile tests of the spliced joints with embedded

cast-in-place concrete were carried out. Monotonically increEising tensile force were applied. Splice

reinforcement was SD30 grade steel and the compressive strength of concrete was 384 kgf/cm^.

From the test result, the following conclusions are mode.

1. Straight lap longer than 20 d ; where d the nominal diameter of the bar, show excellent behavior. The maximum average strain exceed

3 % . When the lap joint length was 10 d, yield strength of the steel attained but after that sudden failure occurred.

2. Loop-shape splice joint showed better per- formance than straight lap splice joint with

lap length 10 d, if transverse reinforcement

is added. Without transverse reinforcement,

brittle failure occurred at average strain of 0.05 %.

3. Straight lap splice joint in topping concrete show excellent behavior similar to straight lap

splice joint. In addition, higher stiffnesses were

obtained compared with straight lap splice

joint.

4. The lap splice joint in precast segments showed

brittle failure before the yield strength was

reached. Failure at the construction joint be- tween the precast concrete and cast-in-place concrete were observed.

369
Table 1: List of Specimens (cont'd)

No. Type o{ Splice Tliicknes) Lap Length Spacing

( (mm) ( (mm) i (mm)

# 1 100 30 d 60

TT * 1 20 30 d 60 nn SLrAirht I An Snlice 100 20 d 60 #4 100 20 d 30

100 10 d 60

100 10 d 30

rr • 100 10 d' 120 10 d' #9 DIO Loop-Sh&pe 100 10 d #10 100 5 d' #11 100 5 d #12 120 30 d #13 DIO Topping-Concrete 120 20 d #H 120 10 d #15 100 30 d #16 D13 Ptecajt Segment 100 20 d #17 100 10 d

* : nilh IcaniveKC reinforcement in loop

Table 2: List of Specimens

Concrete Compressive Steel Tensile Strength Strength Splice Reinforcement

No. Pre-Cast Casl-in-Place Yield Ultimate Ratio kgf/cm' kgf/cm' kgt/cm' kgl/cm' %

#1 ibb 1.01 #2 455 0.84 #3 474 384 3489 5082 1.01 #4 455 1.01 #5 459 1.01 #6 474 1.01 #7 457 0.86 #8 474 0.71 #9 480 384 3890 5787 0.86 #10 474 0.86 #11 459 0.86 #12 459 #13 457 384 3890 5787 0.71 #u 480 #15 480 #16 480 384 3897 5704 0.40 #17 480

Table 3: Test Results

No.

Ultimate

Tensile Load

tonf

Average Strain at Ultimate Load mm/mm

Ultimate Stress Yield Stress Failure Mode'*

#1 21.4 0.0255 1.21 A

#2 > 23,6 • > 0.0343 1.33 A

#3 > 22.2 • > 0.0178 1.25 A

#4 > 24.7 • > 0.0329 1.40 A

#5 20.6 0.0198 1.16 C

#6 18.4 0.00G9 1.04 C #7 17.3 0.0096 1.04 B

#8 19.1 0.0130 1.15 B #9 10.8 0 0011 0.65 D

#10 9.5 0.0178 0.57 C

#11 6 3 0.0006 0 38 D

#12 23.1 0.0215 1.38 A

#13 19.1 0 0040 1.14 A #H 15 6 0.0013 0.94 C

#15 5.5 0.0010 0.71 B #16 4 9 0.0008 0.63 E #17 4.0 0.0006 0,51 E

* Ultimate Strength vrere not attained during tests see Fig, 5

370
Load transducer

Maximum Lo*d : IStoaf

1000

(ft

i!1! (Preca

Sp

i ii 0

St concrete)

ecimen

(Precast co

0 Ii

ncrete)

i li

o

#

o

]

Specimen extension observation length (600)

J

. 500- y//////////////////

Figure 1: Specimen and Loading Apparatus

o-joo-

(a) Siriight Lip Spliced Joint o

(d) Precist Se(m«nt Joint

-}00-

r

1000 120

• ISO

o

(c) Topping Concicle Lap Spliced Joint (b) Loop Spliced Joint

Figure 2: Details of Splice Joint

371
HM

SMI

Mi VI0

iiiestr Ittt

1 im

u

CO 1

f-»n(LW)8rtm,(i>iiiire.

* t-i i.« I a. ts 0. It • 0. li Average tensile strain. (%)

(d) Piecast Segment Joint

(Note : • SpUce ralnforcelng bar yield. X Failure. ^ S(rcss-S(iaJn (clailon of idnrordng bar only.)

1 1 1

i i 1

i ! I -

iKlOd) ill. 11 I

- ! y % " 1 -•"1 "'^

l-:' :tli^.\C:.A. .. . i iS:

: i 1 1 ; i

• t.i I. • \.i l.t l.i 1. Average tensile strain. (%)

(b) Loop Spliced Joint

(Note : • Splice ralnforcelng bar yield. X Failure. #1 Slres-Slraln relallon of reinforcing bar only.)

Figure 3: Average strain vs. Average Stress Relations (cont'd)

Figure 4: Average Strain vs. Average Stress Relations (cont'd)

372
o o

o

A : Gradual failure after yield due to opening at construction Joint. (#1.#2.#3.#4,#12.#13)

C&D

o

o

o

o

o

B : Gradual failure after yield due to concrete fracture near spliced bar. (#7,#8)

C : Brittle failure after yield due to fracture near spliced joint. (#5.#6.#10.#14) D : Brittle Failure befor yield due to fracture near spliced joint. (#9.#11)

E : Brittle failure due to slide along construction joint between segments.

(#15.#I6.#17)

Figure 5: Failure Modes

373
U.S.- Japan Cooperative Research Program PRESSS

Part 2-2

Strength of Concrete Interface

Subjected to Compression and Shear

2.1 Introduction

In precast concrete buildings, site-ca-st concrete connections require cast-in-place concrete for their completion. In the site-cast connection, the in- terface between precast concrete and cast-in-place concrete have an important role in force transfer. The majority of existing data about shear transfer has been obtained from tests of push-off speci- mens which have splice reinforcement across inter- face, but without a clamping force on the interface. Building code requirements for vertical joint or horizontal joint of a wall-panel construction re- flects these test results and study.

At the connection of line element such as beam

and column, the shear should be transferred by ag- gregate interlock under high bending moment, such

as in flexure hinge section. Paticulary, in the member with a small shear-to-span-ratio, shear force is smaller than the normal compressive force. In this condition of stress, the interface does not cause any

slip and behaves as a monolithic constructed concrete with a crack. This kind of concrete interface should be regarded to be equal to monolithic with

respect to the shear transfer by concrete. But, less data is available on slip behavior of an interface and

the failure mode with high clamping force. This paper reports tests of concrete interfaces. The purpose is to find conditions, at which inter- face behaves monolithically. Deformation at the

interface and the failure mode are also discussed.

2.2 Test Program

2,2.1 Variables

This is a simple direct shear test to examine concrete interface between the precast and cast-in- place concrete. Table 1 lists the specimens. Number of specimens is 46 in total. Tlie two variables of the tests are listed as follow:

1. Five different interfaces are tested. The inter- faces are chosen considering the possibility of usage in practice, which does not necessary re- flect the current practice in precast concrete member manufacturing.

2. The inclination of the loading direction to in- terface 9. This is chosen to find the angle at

which the specimen behave equal to monolithic

one. The five different surface conditions of precast concrete are shown in Fig. 1. The characteristics of these surface conditions are briefly explained as follows;

t Type A specimens have no interface because

they are monolithically constructed and in- tended to be used as control specimens.

• Type B specimens have concrete interface without any intentional preparation. Thus the

surface was as smooth as the steel form surface.

• Type C specimens have a concrete interface in- tentionally roughened by brush before precast concrete hardened. The depth of roughened

surface was estimated to be 1 - 2 mm.

• Type D specimens have concrete interface shot-brusted with small steel balls. This operation was intended to be done until the aggre- gate become visible. But due to the limitation

of ability of machine, the obtained roughness was estimated to be 1 mm.

• Type E specimens have a concrete interface with small shear keys. The size of the shear

keys were 2.8 mm height with section of tri- angular section and which were set by 9 mm

spacing.

• Type F specimens had a castelerated concrete

interface. The dimensions of casteleration were

10.0 mm height and the width was about 35 mm. The spacing was 60 mm.

2.2.2 Sliape of specimen, materials

Figure 2 shows the shape of specimens. All 46

specimens have the same shape and dimensions,

20 X 20 X 10 cm. They have no internal reinforce- ment. On four sides, shear keys with spacing of 20 mm were made, in order to prevent non-uniform

stress causing irregular failure of specimen. Nor- mal weight concrete with compressive strength of 213-294 kgf/cm^ was used. The maximum aggre- gate size was specified to be 20 mm. Table 2 lists

374
the concrete compressive strength. Ca3t-in-plax:e concrete was placed to make the direction of the

interface vertical to horizontal plane.

2.2.3 Loading

Figure 3 shows a loading rig. The loading rig was

set between the parallel loading plates of the hy- draulic universal testing machine, the capacity of which is 200 tonf. Monotonically static increas- ing compressive load was applied to the specimen,

maintaining the constant angle between the concrete interface and loading direction. The value of tan d of the angle was 0.4 to 2.5. In order to prevent from irregular failure due to uneven stress by bending moment, the rotational freedom of the loading plates was fixed. To relea-se confinement of interface after large slip at interface, a set of slide bearing was inserted between the load- ing rig and the base plate.

White gypsum capping was used to get uniform

stress transfer between specimen and 10 mm lliich steel Lrshape loading plates, the surface of wliich had the same shape as the shear key on specimen

sides. The gypsum capping was also intended to keep the opposite sides of specimen parallel. Tests were carried out without initial cracks at interface.

2.2.4 Measurement

Applied load and the deformation near the inter- faces were measured. Assuming the shear stress along the interface r and clamping stress normal

to interface (7„ is uniformly distributed, the follow- ing equations are obtained from the equilibrium of internal stress and applied force.

T = ^sin^ (1) A

P

a„z=-cos9 (2)

where, P : the applied load, 6 : the angle from

loading direction to the normal line to interface, r

: shear stress along the interface, a„ normal stress acting on interface, A : area of interface. By increasing the compressive strength of the concrete, the ultimate shear stress of the specimen

also increases. Thus, for the purpose of comparing the ultimate shear stress of specimen, the ratio of the shear stress to compressive strength of concrete referred to as the normalized shear stress is used. In the case that the strength of the precast concrete is different from that of the casl-in-p!ace concrete, the mean value was used to normalize the shear stress.

Translational displacement along the interface and at 90 " to the interface were measured with

four crack displacement transducers, which were at- tached to the both surfaces of the concrete with adhesive. The distance between measurement targets of each crack displacement transducer was 60 mm

in common.

2.2.5 Test procedure

More than two specimens were tested with the same

test variables. Some specimens show irregular fciil- ure patterns, for example, the value of strain mea^

sured on opposite side of specimen were signifi- cantly different, and the failure modes differ on two

sides. The reason of irregular failure may be at- tributed to errors in the capping procedure. In this case, the irregular data was discarded and another

specimen was tested. The actual number of spec- imens loaded was 63 and the 17 specimens failed irregularly, (27 % of total number of specimens)

the data of which were discarded. Thus, the tests of only 46 specimens are reported here. The specimens with small tan 6 value were tested

first. By increasing the tan ^, the dominant failure pattern changed from crushing of concrete to translational displacement along interface. Once change

of failure mode occurred, further tests with speci- men with larger tan Q values was not conducted.

2.3 Test Result and Discussion

Here the (1) ultimate Shear Stress, (2) shear stress- translational displacement relation, and (3) fail- ure criterion of the interfaces are reported and dis- cussed.

2.3.1 Ultimate shear stress The ultimate shear stress of different types of inter- face are compared with that of monolithic one in Fig. 4. The horizontal axis represents the normalized shear stress defined in equation (1) and the

vertical axis represents the tan The values of tan 0 at which the specimens showed almost equal

or greater strength were 0.6, 0.8, 1.00, 1.00 and 1.25

for Type B, C, D, E, and F, respectively.

The provision for shear-friction in the ACI code

[l] specify an eflfective friction coefficient 0.6 to concrete to concrete and 1.0 for intentionally roughened concrete interfaces respectively. If the effec- tive shear friction coefficient in the ACI code is used, the interfaces are expected to have higher

strength than monolithic concrete interface.

375
2.3.2 Translational displacement - shear

stress relation

The shear stress vs. translational displacement

along interfaces are shown in Figures 5 to 7. The

translational displacement is the mean value of four displacements, obtained with four crack displacement transducers. The elastic stiffness of the inter- faces showed good correlation with that of monolithic concrete, except that the Type B specimen

with unroughened interfaces show low stiffness. As

the value of tan d decreased, the elastic stiffness de- ceased. The ultimate shear stress was attained at the translational displacement of 0.1 - 0.25 mm.

The stiffness of interface and the translational dis- placement at maximum shear stress of Type F in- terface was variable but wets generally larger than

that of monolithic specimen and specimen with

Type E interface.

2.3.3 Failure criterion of intorfacns

The ultimate shear stress divided by the concrete strength r//g, and normal stress (^dS'c are plotted

in Figures 8 to 12, and compared with the yield surface of monolithic concrete. Type A Monolithically constructed specimens of Type A construction, showed compressive fail- ure, when tan $ was in the range of 0.4 - 1.67. In this CELse, diagonal sliding was occurred after the ultimate loaui was reached. The specimen

with a value of tan 0 of 2.5 showed failure along

the plane corresponding to interface. The plot marks show good correlation to the arc of Mohr's stress circle, which had principal

stresses of zero and 1.25 (where, is the compressive strength of a concrete cylinder (10

cm dia. X 20 cm))

Type B In the case of the interface Type B, at which no roughening was carried out, no fail- ure along interface was observed, with the val- ues of tan S of 0.4 and 0.6. The spocimen witii a value tan of 0.8, showed suddcMi .slip and

brittle failure at a shear stress of 2.') % - .3.') % of ultimate shear stress of monolithic speci- men. Type C The interface of Type C, which had a brushed interface, showed no failure along the interface when value of tan from 0.4 to 1.0. The specimen with tan 6 = 1.25, showed a sud- den slip and reduction of restoring force just af- ter attaining to maximum load. Tiie inaxininni load exceeded that carried by a monolilhic

specimen. From the failure surface observed,

the cast-in-place concrete was scratched off by

precaist concrete. Type D The shotbrust roughened interface of about 1 mm depth, was used in Type D. The

specimen of with tan ^ values of 0.4 - 0.6 showed no failure along the interfaces. The

specimen with tan 6 val ues of 0.8, showed sud- den failure after it attained its maximum load. The maximum load was larger than that of the Type A specimen. The failure plane seemed to have been made by scratching of the cast-in- place concrete by precast concrete like that of Type C. Type E The interface Type E had small shear

keys. In this type, specimens with tan 0 from

0.4 to 0.8 showed no failure along interface. The maximum strength of specimens in which

tan d equaled 1.0 exceeded the strength of the monolithic specimen. One of them showed a sudden failure along the interface after maximum strength was reached, the other specimen

showed different failure mode and the^trength

rapidly decreased. The maximum strength of the specimens with tan 6 of 1.25 exceeded the

strength of the monolithic one. After attained

to the maximum strength, sudden failure along

interface occurred. The cast-in-place concrete seemed to be scratched off by the shear key of the precast concrete. Type F The Type F interface has a casteleration. No failure along the interface were observed

in the specimens with tan d value of 0.4 - 1.25. During the loading of the specimens with tan 0 values of 1.67, sudden reduction of load oc- curred at a strength of 70 % of the monolithic specimen strength. Failure of cast-in- place concrete along the lines which connect

the peaks of the casteleration of precast concrete was observed.

2.3.4 Discussion

The values of tan 9 at which the failure mode

change from crushing of concrete to failure along

interface are listed in Table 4. From these test re- sults, it was found that by increasing the roughness of the interface, the ability to transmit skew

compressive force increases. It is also seen that

the specimen with higher shear strength than the monolithic specimens did not necessary showed the same failure mode as monolithic specimens.

376
2.4 Conclusion

Forty three specimens were subjected to compression and shear. Testing was were carried out in order to examine the performance of tiie concrete r interface between the cast-in-place concrete and tlie precast concrete. The conclusions are given below.

1. When the angel d from the direction of loading

to the normal line to interface was small, No

failure along the interface occurred and both

sides of the concrete acted as one unit. The

maximum strength was greater or equal to that of monolithic specimens. By increasing the angle 0, the failure mode changed to the failure along the interface which was aJways accompanied by sudden reduction of restoring force. Therefore, if the angle d is smaller than a par- ticular value (which depends on the type of

interface) the interface of precast concrete and

the cast-in-place concrete behave as one.

2. The angle at which the failure mode changed

from failure other than interface to the failure at interface, varied according to type of inter- face. The tan for interface without roughening was 0.6 - 0.8. As increasing the depth of roughing, The tan ^ increase up to 0.8 - 1.67. In case of monolithic concrete, the value tan 0 was more than 1.67.

3. The translational displacement at the maximum shear stress wa.s estimated to be 0.1 -

0.25 mm. Whether or not an intcrffice was

used, did not affect the elastic stiffnass before the maximum strength. The stiffness of an in- terface without any roughening was lower than

other interfaces.

References

[l] ACI Committee 318. Building Code Require- ments for Reinforced Concrete (ACI 31S-S9)

and Commentary (ACI 318R-89). American

Concrete Insitute, Detroit, Michigan, 1989.

377
Table 1: List of Specimens

Type Interface tan ( —fTotal No.

Condition 0.4 0.6 0.8 1.0 1.25 1.67 2.5 A Monolithic 2 1 12 B Concrete to Concrete

1 2 2 5

C Concrete to Roughened

Concrete with brush

2 2 1 2 7

D Concrete to Shotbrusted Concrete

2 2 2 6

E Concrete to Concrete

with Small Shear Keys

2 2 2 2 8

F Concrete to Concrete

with Caateleration

2 2 2 2 8

where, 9: angle (see Fig. 3)

Table 2: Concrete Compressive Strength

Concrete Compressive Strength in kgf/cm'*

Type Precast Caat-in- Place Average

Concrete Concrete Strength

A 213-248 B 270 245 258 C 226 251 239 D 248 242 245 E 224 204 214 F 213 291 252

Table 3: deserved Failure Modes

Type Interface tan^

Condition 0.4 0.6 0.8 1.0 1.25 1.67 2.5 A Monolithic CC C CC CC c CC SS B Concrete to Concrete C CC SS

C Concrete to Roughened

Concrete with brush CC CC c SS

D Concrete to Shotbrusted Concrete CC CC SS

E Concrete to Concrete

with Small Shear Keys

CC CC cs SS

F Concrete to Concrete

with Casteleration CC CC CC SS

where, C : Crushing of Concrete, S : Sliding along interface and

8: angle (see Figure 3)

Table 4: tan 6 at which the failure mode change from crushing of concrete to slide failure along interface

Type Interface Condition tan^

A Monolithic 1.67 - 2.5 B Concrete to Concrete 0.6 - 0.8 C Concrete to Brushed Concrete 1.0 - 1.25 D Concrete to Shotbrusted Concrete 0.6 - 0.8 E Concrete to Concrete with small shear keys 1.0 F Concrete to Concrete with Casteleration 1.25 - 1.67

378
Cast-ln-PliM 200

Precast

(a) Type B Concrete to Unroughened Concrete

(d) Type E Concrete to Concrete with Small Shear Keys

Casl-ln-P!ace 200 -40 ^

_4a

_fiJL r_41L

k i Prtcasl

(e) Type F Concrete to Concrete with Casleleration

unit : nnm

Figure 1: Configuration of Interfaces of Specimens

100

unit : mm

Figure 2: Geometry of Specimens

Cajt-ln-Place 200 113

• .. -180

(b) Type C Concrete

Precast

to Roughened Concrete with brush

SO.100

Caist-ln-Place 200 —=iafl 3,

• f

Precast X

30 ^

Toilni MacMia

UlKV Slldtc

Figure 3: Loading Method

unit : mm

Ult(n4ic shear stress ^snoele compressive stienglh Uliimale shear stress / Concrete compiessive slienglh

0.4

1 0 H9 ^—jUl Type B Concrete to Unroughened Concrete

Ulitraiie shear stress "onciele compiessive strength

1.23

a49i ^^^^^^

0.663

, 10SI7

04

at

1.0

1.23

(d) Type E Concrete to Concrete with Small Shear Keys

Ultimate shear stress / Concrete compressive itienglh

'^Mi-y.-:-:-:

10 )n

:-:-:-:jx3ii^:-:><>::-:->>:':::^

10)0

•:-:-:-iuij-:':':-:':':':':'>il o.t is

.730 o«

1.23

(b) Type C Concrete to Roughened Concrete with brush 1.67

|0 -

i

1 i OSIT :':v;5^1ui%';-:'x'l)374

1 410

UI:iTjie shear stress Concrete compressive strength Concrete with Casteleration

04 f4.IO

oi.54

t

^06^

(c) Type D Concrete to Sholbrusted Concrete

I I nrPt * UonoJllhM

TTfPE B Cmnu k> Unnwtl»i>d Cnnia

j rrPC-C Cencrtv to RAi|hmd

TVPC-O Coccm* 10 Sholtmlcd Onnru

nrPC e Concnw u Concnu

MID Snill Shiv Ktri

I* ' ' w wlACMirkndoa

I unl \ «»ilo to >tnn|lli of BonoinNc Ipcdan

Figure 4: Comparison of Ultimate Sliear Stress

379
0.1S

t.it

t.li

e.io

0.2S

, —,:Fai;r /^CO.I-J

fO. 4-1

EO. 4-1

M.4-J..

CO. 4-1

00.4-]

AO. 4^1.

tan (9 =0.4

M. l-l U.(-l

tan (9 =0.6

FO.J-l

EO. l-l K.lrl..

co.»-:

BO. 1-2

AO.tri.. DO.I-ZAIIura iloni tnlerfaca

X BO.t-]AIIunilon| Inlcrfs tan (9 =0.8

-«.]> 0

° t.2i o.io o.ii 1.00 i.n i.io ttanslational ditplacement along intefface, mm

Figure 5: Translational Displacement along Concrete Interface vs. Shear Stress Relation (Cont'd)

S

s.Eo

II. 0-1

I Cl.O-l Ar.4-r

tan (9 = 1.0

E1.2}-ZFUIin iloDl kllstan CLZS-lAllunalonilnlcrfi

A1. 25-}

ri:

fi.ji-i

El. 2S-t Cr.25-I

Al. 2S-J

tan (9 = 1.25

ri. iT-i

Al.lJ-2

F1.67-l.F>llun iloni InltrfK

j

tan (9 = 1.67

-«.2S 0 0.2S O.SO O.li 1.00 1.2S

translational displacement along interface, mm

I.JO

Figure 6: Translational Displacement along Concrete Interface vs. Shear Stress Relation (Cont'd)

&

E

8

V O.SO

tan (9 -2.5 *2. i-2

£ -0.2S « 0.2S O.IO 0.71 1.00 I.2S I. Si ttantlational displacement along interface, mm

Figure 7: Translational Displacement along Concrete Interface vs. Shear Stress Relation

380
0. 2S 0. SO 0.7S 1.00 1.2S Ultimate nocmil stiesi / Concrete compressive strength

I. SO

Figure 8: Failure Criterion of Type B Concrete to Concrete unroughened

^ 0 25 0. 50 0.75 1.00 1. H 1.50 Ultimata normal stress / Concrete compressive strength

Figure 11: Failure Criterion of Type E Concrete to Concrete with small Shear Keys

(b) ^J^^ C C«icifli 10 Roinhfii«J Coociilt wllli bink • A : Compnulon ftlkire

1 Cwt-to-Pau 1 j O ^ I Futun ajong bilcWsct

1 rnua FiOn alool bHrfaca ^

I

' i - 1 i

.

s i i /

/

/o . ~j y

1 //

/

• / i / \

\

1

1

!

i

N j

1 s

i

i

i "b^rj-* TYPE-A

0

1.00

0.:5 0.50 0. 75 1.00 1. 25 Ultimate normal stress / Concrete compressive strength

1.50

Figure 9: Failure Criterion of Type C Concrete to Roughened Concrete with Brush

0.2S 0.50 0. 75 1.00 t.25 Ultimate normal stress / Concrete compressive strength

1. SO

Figure 12: Failure Criterion of Type F Concrete to Concrete with Czisteleration

Ultirntti notmal streii / Concrete compressive strength

Figure 10: Failure Criterion of Type D Concrete to Shotbrusted Concrete

381
Japan-U.S. Cooperative Research Program PRESSS, Part 3

Experimental Study on Shear Walls of High-Rise

Precast Reinforced-Concrete Frame Buildings

.

by

Shinsuke Nakata, Takashi Kaminosono, Hitoshi Shlohara, and Mizuo Inukai

1. ABSTRACT

We make an experimental study on two lower three-storied shear walls

of fifteen-storied precast reinforced concrete frame building. We have

not yet made criteria about high-rise precast reinforced concrete frame

building. This experimental study will be a fundamental data which may

be useful for making the criteria. In this experiment, we put an attention on the vertical joints of the shear walls, in order to discuss why

and how the vertical Joints are destroyed and what effects on the

shear walls this destroying happens. So we get the results of destroying and not destroying of vertical joints on first floor.

At the results, it is clear that at the shear wall which vertical

joints are destroyed its strength easily decrease and at the shear wall

which vertical joints are not destroyed its strength keep itself until

joint translation angle is 1/100. When the horizontal load marks the

maximum load, the displacement of both shear walls concentrates on the

first floor, and after that, the rotation angle and story drift angle

decreases on over second floor.

Therefore we recognize that it is useless for seismic system to design

the vertical joints which is destroyed under earthquake.

2. KEY WORDS

1) precast. 2) frame. 3)shear wall. 4)cotter-bar in vertical joint.

382
. . Japan-U.S. Cooperative Research Program PRESSS, Part 3 Experimental Study on Shear Walls

of Hlgh-Rlse Precast Relnforced-Concrete Frame Buildings.

by

Shinsuke Nakata*^ Takashl Kamlnosono*^, Hitoshl Shlohara*"^, and Mlzuo Inukal*

1

.

ABSTRACT

We made an experimental study on two lower three-storied shear walls of fifteenstoried precast reinforced concrete frame

buildings. We put an attention especially on vertical Joints between column and

shear wall which Is supposed to effect on

strengths of buildings.

2. DESIGN of TEST SPECIMENS

The specimens are two shear walls, that Is three-storied, one-third scale, 3 meters

span, 1.17 meters height on each floor. The specimens are made of columns, beams,

shear walls, slabs, a foundation, and a loading-beam (Fig. 1). The precast members

of the specimen are columns and shear

walls with beam (Flg.l, white parts), and

the cast-ln-place members of It are slabs, a part of beams, vertical joints, and so on (Flg.l, gray parts). The size of members

are as follows; column section: 300x300mm,

beam sectlon:300x230mm, wall thlckness:100

mm, slab thlckness:100mm. The specimen has

vertical Joints and horizontal Joints. The

vertical Joints are reinforced concrete

150mm wide, 230 or 100mm thick, between

cast-ln-place column and beam, or precast

column and shear wall, respectively. They

Include shear-cotters(100xl00mm section,

20mm thick, 45 degrees) and maybe cotterbars. The horizontal joints have two

types; one Is spllce-sleeves-and- mortar- type which Is 10mm thick between shear

wall and slab, or column and slab, and the

other Is cotter-bars-type between beam

and slab. In order to discuss why vertical

Joints fall, we decide the parameters of

test specimens as the existence of cotter- bars and vertical-bars area of shear wall

In the horizontal Joint on first floor. The specimen PCWALLl has no cotter-bars

and all vertical-bars of shear wall In the

horizontal Joints on first floor, and the

specimen PCWALL2 has cotter-bars and

three vertical-bars In each side of shear

wall on first floor. We show the data of

test specimen In Table 1, the bar arrangement of test specimen In Fig.2, the section

detail of column In Fig. 3, the detail of

vertical Joint In Fig. 4, and the material

properties In Table 2.

3. TEST SETUP and LOADING

HISTORY

We show test set-up equipment In Fig. 5. We put the specimen on reaction floor,

four actuators upon the specimen for

axial force loading, two actuators on

reaction wall for horizontal force loading. This test setup Is called a cantilever

loading. The specimens are subjected to cyclic load history.

During cyclic loading, the specimen

PCWALLl Is subject to the axial force

(210tonf) and the top moment which keeps

the Inflection point 7-meters high. The

specimen PCWALL2 is subject to the axial

force (lOOtonf). The inflection point of

PCWALL2 Is 4.2-meters high.

The cyclic load history Is as follows; one positive-negative cycle at lateral

force (=20 tonf which Is below bending

crack strength), one cycle at Joint translation angle (R= 1/2000, 1/1600 rad.), two cycles at R=l/800, 1/400, 1/200, 1/100

(rad.), and then, for specimen PCWALLl, two cycles at R=l/50, or, for specimen PCWALL2,

three cycles at .R=l/67 (rad.)and the deflection for failure.

During loading, we sample the data of

bending deformation In column and shear

wall. Joint slip and opening displacement

at horizontal and vertical joints, and

steel strain of column-maln-bar, wallvertlcal-bar, beam-main-bar, and cotterbar.

• 1 Head Research Associate, Building Re- search Institute (BRl). 1 Tatehara, Tsukuba City, Ibarakl-Pref., Japan

•2 Chief Research Associate, BRI. •3 Research Associate, BRl. •4 Research Associate, BRl.

383
4 . RESULTS

The crack configuration In Fig. 6 and

lateral load - displacement relationship In Fig. 7 are shown. About specimen PCWALLl,

bendlng-crack Is found at R= l/1600(rad.),

bending yield happens at about R=l/250,

and specimen strength decrease largely at R=l/50 because of shear-slip- failure at precast shear wall on first floor. The

bending yield Is caused by tensUe-yleldIng of column-maln-bars In tensUe-sIdecolumn on first floor and wall- vertical - bars In quarter-tensUe-sIde - wall from

first to third floor. It Is unexpected that

there Is no failure In vertical Joints In spite of no cotter bars. About specimen PCWALL2, bendlng-crack Is found at R=l/1600(rad.). bending yield

happens at about R=l/270, and specimen

strength decrease largely at R=l/67 be- cause of vertlcal-jolnts-fallure on first

floor. The bending yield Is caused by

tensUe-yleldlng of column-maln-bars In tensUe-sIde-column on first floor. The

result is that the vertlcal-jolnts-fallure

Is cause by the decrease of axial force,

the inflection point height, and the wallhorizontal-bar area on first floor.

4.1 COMPARISON of TEST RESULTS

With CALCULATED RESULTS about

STRENGTHS

.

Table 3 Is the comparison of test with

calculated results about bendlng-crack,

bendli\g-yield, and maximum strength of

specimens. The equations for calculations are below Table 3. About bendlng-yleld

and maximum strength, calculated results

follow test results perfectly, but about

bendlng-crack strength, calculated result

Is greater than test. It Is because we

decide the bendlng-crack-happening as the

time when we can see a crack at precast

members Ignoring horizontal Joints.

4.2 OVERAI.L DISPLACEMENT

FIg.8 shows the overall displacement of

specimens at R=l/400, 1/200, 1/100, 1/50 or

l/67(rad.), and maximum R (joint translation

angle). We describe the displacements by

ten times multiplying. In Flg.8, we can find

that the overall displacements follow the

bendlng-dlsplacement-type from first to third floor until smaller displacement, and

the horizontal displacement concentrates

on first floor at maximum R while the

story drift displacements on 2nd and 3rd

floor decrease.

FIg.lO shows rotation angle - story drift

angle relationship. It is clear that the

relationship of each floor is linear until

R= l/50(rad.) In PCWALLl. or 1/67 in PCWALL2, and the displacement concentration on first floor is caused by the rapid Increase of story drift angle on first

floor and decrease of rotation angle.

4.3 DISCUSSION about DISPLACEMENT CONCENTRATION

Flg.ll shows the lateral load - opening

or slip of vertical and horizontal Joints

relationship. Fig.9 shows the sampling

place of these data. Both of Joint opening

displacement and slip is tend to increase

at positive loading (column tensile). In specimen PCWALLl, the Joint slip is about

7-milllmeters at R= l/50(rad.), and In specimen PCWALL2. It is about 14 millime- ters at R=l/67(rad.). After that, fn PCWALLl

the precast wall on first floor is mainly

deformed while negative loading (column

compressive), and in PCWALL2 the vertical

joints failure Is caused by rupture of

wall-vertlcal-bars, Increase of joint-slip

displacement at first floor wall, and

compressive failure of columns while positive loading (column tensile).

5. CONCLUSIONS

1. The specimen PCWALLl (without cotter

bar): failure mode Is shear-slip of precast

shear wall and hysteresis characteristic is bending type Including lager hysteresis

energy within about R=l/100. There Is no

failure at joints.

2. The specimen PCWALL2 (with cotter bar):

failure mode Is vertical-joints when Joint

translation angle Is 1/67. The displacements concentrate on first floor.

3. The horizontal displacement concentrates on first floor because of shear

failure or horizontal joint slip displace- ment of first floor wall while the rotation angle and story drift angle from

second to third floor decrease.

4. The vertical joint failure on first floor is caused by the increase of horizontal joint slip of shear wall.

384
5. The equations for calculated results

properly follow the test results even If precast concrete shear wall.

6

.

REFERENCES

1. Y.Suenaga, T.Goto. Y.Sonobe, "Guide for

DeslgTi and Prefabricatlon of Precast Reinforced Concrete Structures", 1986,

Architectural Institute of Japan.

2. S.Mochlzukl. E.MakltanI, T.Shlnto,"Behavlors and Design Method on Vertical Joint

of Wall Panel Precast Concrete Structures", 1989, Architectural Institute of

Japan.

3. K.Ogura, H.Aoyama, H.Tomosawa,"Standard

for Structural Calculation of Reinforced

Concrete Structures", 1988, Architectural

Institute of Japan.

385
Table 1 Section Size of Speciuen

M e m b e r Sec t i o n Size

Section

1 300x300 (mni) Co I u m n Main bar

| 12-D13 Main bar ratio 1.69 %

(PCa) Hoop i D6840 Hoop ratio 1.07 X

Section

1 230x300 (mil) Beam Main bar

j 12-D13 Main bar ratio 2.21 X

(PCa) Stirrup j mM Stirrup ratio 0. 348 %

Length \ 2400 (mill) Wa 1 1 Height i 856 (mm) Wall bar D109100

(PCa)

1 Thickness! 100 (mm) Wall bar ratio 0.71 X

Length

j 4000 (mm) Slab Width 1 1000 (mm) Slab bar 6^@50 mesh

(Cast-in-pl ace) Thickness] 100 (ram) Slab bar ratio 0. 560 X

Table 2

a) Tensil Strength of Bar

Results of Material Test unit:kg/cm'

Type Name

Strength

Veild Rupture

PCa

D 6 3560 5060

DIG 3950 5810

D13 4030 5950

Cast ill-place D 6 3690 5260

D13 3650 5140

b) Compressive Strength of Concrete

Type

1st

Story

2nd

Story

3rd

Story

PCa PCWALLl 505 515 507

PCWALL2 526 522 526

Cast-in-place 555 505 617

Mortar 646 680 691

Grout 691 691 691

m

Elevation 1 Elevation 2

notes DiPCa . Si:Cast-in-place

Fig. 1 Specimen

2MFig. 3 Section Detail of C:oluinn

ParALL2 PC*ALL1

(ilth cotter bars) (without cotter bars)

Fig. 2 Bar Arrangement

Fig. 4 Detail of Vertical Joint

386
Ilei»ll»e Lotdlif Poiltl»e Lotdln

i:<:>:ww:t:KKia>JH:<:i:i6iia;<ffi:iiiH:t>»i-ti:KnrS|-

reacl'lo'n fToor^

IMil \ 3Q00 I 7850

a) Specimen PCffALLl

Hesatlve UidUg PoiiUve Lojdiii

e a c 1 1 o d

' f I ooV ^////'^^

11500115001

i Tqqq I

Fig. 5 Set-up Equipment

(-1—• —()

-1

ri

d

1 1 b) Specimen PCWALL2

Fig. 6 Crack Configuration

Table 3 Comparison of the Test Results lith Calculated Results

Specinen PCWALL

I

PCWALL2

Strength Bending Crack Yelid Mail nun Bending Crack YelId Max! nun

Test results (tonf) 4 0.2 84. 8 10 0.8 5 3. 5 9 3.4 10 4.3

Calculated

Results

(tonf)

Moaent (tonf-n) 2 12.3 6 5 8.9 6 17.4 16 1.5 3 5 8.7 3 6 2.5

Shear (tonf) 3 0.3 7 9.8 8 8.2 3 8.0 8 4.5 85. 4

Test Resul ts/Shear 1.33 1.06 1.14 1.4 1 1.11 1.2 2

1 . Tension crack louent : Mc=(l. 8/"FC+N/Ae)»Ze 2. Yelld Boaent : M»r=0. 8-at-sayDtO. 2-ai-s<7 lyDtO. 5-N-D(l-N/b-D-Fc) 3. Ullliate «oaent : Mu^at -sff y If -40. 5-af -sa wy li»0. 5-N- If Values fith caluculatlon :s<7y='4O00 kg/ca' •safy'JSOO kg/ca'

PCirALLl:Fc=502 kg/ca' • N=226.0 tonf • h=700. 0 ca

PCI(ALL2:Fc»533 kg/ca' • N'115.8 tonf • h=424. 7 ca

Qc=Mc/h Qty=M»y/h

Qu=Hu/h

120 10080

6040

20

0 -20

-40

-60

bending ye

bending crack

ld^^^_^ •

aax. strength

-80 -60 -40 -20 0 20 40 60 80 Horizontal Disp. (4 F) (mm)

T:fallure of shear slip

a) Specimen PCWALL

1

T3o

120

100

80

6040

20

0 -20

-40

-60

-80

-100

-120

bending yelld

bending crack

—'—1—'—1— 1

' 1—'—I aai. strength

-80 -60 -40 -20 0 20 40 60 80 100 120 Horizontal Disp. (4 F) (ram) A:railure of vertical Joint

b) Specimen PCWALL2

Fig. 7 Lateral Load - Displacement Relationship

387
1/400 1/200 «»l R-^/h

'1/100 *•

1/30

direction

ai. disp

1/400 11

1/67

(1) PCWALLl

^ (2) PCWALL2

Fig. 8 Overall Displacement

O).

20

10

rt -10

•*-» o

-20-40

(measured point)

disp.

slip of vertical joint

IF horizontal slip of column base

notes DcPCa , ^:Cast-in-place

Fig. 9 Measured Points of Opening and Slip

PCHfALLl 1

i \/m

a 4 -o 3F

-1/50 i

A'

D.V:..v.D-2F

-

A-j—A IF

1 -20 0 20 Story Drift Angle (xlO'*rad.

)

a.::7::r70 3F.

I 2F A i-A IP

-25 0 25 50 75 100 Story Drift Angle (xlO"'rad.

)

Fig. 10 Rotation Angle - Story Drift Angle Relationship

IZO

-10 12

IF Opening of

Vertical Joint

^ 120 c

o

60

-o° 0

-60

120

60

0

-120

1/200 "

1 -i/?oo •

-i/;so^

-l/(0»

-1/200 : -1/100 -4 0 4 IF Slip of Vertical Joint

120

60

0

-60

-120

U/lOO 1

-|/S0

i -1/100

f-1/100:

'-1/(00 : -1/100 i

i -4 -2 0 2 4 IF Hori2ontal Slip

of Wall Base

7.0 120

60

0

-60

-120

! 1 i/ioo ! : l(10ttjY<>:l/SO

A!

\ W... mrum. l

-i/5o<r

/5W-I/400 i / ftS-l/jOO : -i/iool ; . -5.0 -2.5 0 2. 5 5.0 IF Horizontal Slip

of Column Base

1/100

: 1/200

: 1/(00

;...l/«QO V \

-i/ibb

:

-1/(00 : -I/20C : -I/IOO ; 60 -30 0 30 60 IF Opening of

Vertical Joint

120

60

0

-60

-120

1/200

1/400

....V.'.oo

/ \-|/«7 i

f-l/lOO -1/100

-1/200

I/IOO

a) Specimen PCWALLl

120

-10 0 10 20 IF Slip of

Vertical Joint

30 -40 0 40 80 120 IF Horizontal Slip

of Wall Base

9.0 -4.5 0 4.5 9.0 IF Horizontal Slip

of Column Base

b) Specimen PCWALL2 (tir max. disp.) (nun)

Fig. 11 Lateral Load - Joint Opening-Slip Relationship (IF)

388
Theme V

Joint Performance of

Nonstructural Systems
'1I
Current Efforts to Improve the Performance of Nonstructural

Systems Based on Experience From the Loma Prieta Earthquake

by

Robert Reitherman*

ABSTRACT

The October 17, 1989 Loma Prieta

earthquake in Northern Cahfomia produced a few new findings concerning the behavior of

nonstructural components, while in most

respects the earthquake confirmed past

conclusions. The psychological impact of the

earthquake, however, was more historic than

the technical nonstructural lessons learned:

Expectations have been raised, at least in

California, as to the level of seismic

performance, including the performance of

nonstructural items. These higher

expectations on the part of the public must be

balanced against the increased cost of higher

levels of protection. Some nonstructural

features, such as elevators and various

hazardous materials components, are likely

candidates for improvements, while past

practice will proceed unchanged for most

items.

KEYWORDS: contents; equipment;

nonstructural components.

1. INTRODUCTION

A variety of nonstructural earthquake

damage, as well as good performance or lack

of damage, occurred in the Loma Prieta

earthquake. None of the 63 earthquake

fatalities was caused by nonstructural

damage, as shown in Table 1. However, this

optimistic fact should be cited in conjunction

with another fact: the earthquake's ground

motion only peripherally affected the heart of

the urbanized portion of the region, causing

only a moderate level of shaking. Even many

deficient or vulnerable nonstructural items

escaped damage. Judgement must be used in

interpreting earthquake casualty statistics:

There is sometimes a tendency for the latest earthquake's figures to be quickly seized and

used to prove a particular theory or to

support a certain policy prescription. Some

kinds of straightforward engineering

solutions, such as anchoring relatively rigid

pieces of equipment or cabinets, appear to need no revision. Future changes in practice

will be more focused on more complicated

nonstructural components that did not fare as well as most designers and clients would

like, such as elevators and hazardous

materials components.

2. LOMA PRIETA NONSTRUCTURAL

DAMAGE

A summary of nonstructural damage in this

earthquake is provided in Ding and Arnold et

al. (1990) and discussed by Soong (1990)

and by many of the authors of papers in the

Applied Technology Council ATC-29

proceedings (Bruce, 1991 in progress). One

convenient means of pulling together the

variety of effects is to group them by building

occupancy. Using this approach, some of

the more interesting and significant kinds of

damage are seen to have occurred in medical,

manufacturing, and laboratory settings, rather

than residential, school, or office buildings.

Ubiquitous types of nonsti'uctural items such

as suspended ceilings or light fixtures

suffered damage where not designed in

accordance with then current (1985 or 1988

edition) Uniform Building Code provisions,

and in the case of some ceilings there was

perhaps damage even where those provisions were met. However, damage to the more

specialized kinds of nonstructural features in

medical, manufacturing, and laboratory

buildings was more notable.

The prime case of medical facility damage

was the WatsonviUe Community Hospital.

* The Reitherman Company, Half Moon

Bay, CA 94019

391
Though no strong motion instrumentation

was at the site, it probably experienced

shaking of about the intensity measured at the

Corralitos station (0.47 g PGA vertical, 0.64

g horizontal) or the Watsonville Pacific Bell

building (0.66 PGA vertical, 0.39 g

horizontal). These peaks are at least double

the typical values for the more urbanized

portion of the region around San Francisco

Bay. Damage to nonstructural items at the

Hospital comprised a long list, as reported by

Angell (1990): fallen and spilled

pharmaceuticals and chemicals, (including a dangerous hazardous material incident);

malfunctioning radiology processor (no Xray service); spilled files and leaning,

overturned, or jammed file shelving (only ad

hoc patient tracking, billing, reference to medical histories, etc.); lab equipment thrown

to floor (no "high tech" support services

available); broken water pipe (3 inches of

water in the key Central Supply area within

10 minutes); broken gas line (one of the

danger signs that almost led to complete

evacuation); large exterior oxygen tank tipped

20 degrees (adjacent to 30,000 gallon

propane tank, causing another situation that

almost triggered complete evacuation);

toppled refrigerators (no food service); main

doors jammed (usual access had to be

bypassed); damaged elevators (manual

transportation of patients from the area of the

building that was evacuated--the Critical Care

Unit on the top or fourth floor); one of two

diesel motor-generators displaced off its mounts, disabling itself and causing a fuel

outage to the other unit (concurrently with a 26-hour utility company service outage);

paramedic base station radio damaged and not

repaired until two days later (while telephone

service was sporadic after an initial complete

outage). Some of the above, such as the

generator damage, should have been

prevented by the use of currently required

California Hospital Act building code

requirements. Other examples, however,

such as spilled chemicals and perhaps

elevator damage, seem to be problems that

current standards do not completely solve.

The case of this hospital-one of the smaller

of the 112 acute care hospitals in the region

but the most severely shaken (California

Association of Hospitals and Health

Systems, 1990)—illustrates the range of

effects possible, and also indicates the

distinction between damage that can cause

injuries vis-a-vis damage that can interrupt an

essential service. The near collapse of the

oxygen tank represents both types of risk, the

damage to the generators was only a

functional loss problem, while the extensive

amount of glass (window and contents)

damage in the building was primarily a safety

threat.

At one large high-technology research and

manufacturing facility with numerous labs, a

similar long list of nonstructural damage

pertains, though at a lesser scale. This

complex of buildings was near to sites where

the recorded ground motion reached a peak of

0.3 to 0.4 g acceleration. Damage included:

toppled compressed gas cylinders, spills of

hazardous materials in labs, broken

glassware, overturned file cabinets; asbestos

ceiling material was released, small

computers fell, equipment on wheels rolled

and hindered exiting, some hght fixtures fell,

and several chilled water lines broke and

caused water damage.

A commonality between the public hospital

and the private research and manufacturing

facility is that loss of function proved to bein retrospect, from the point of view of

tenants or owners—intolerable. In both the

hospital and lab facilities, extensive remedial measures have been implemented after the

earthquake, totalling between one and two

million dollars for nonstructural work in the case of the company and $42 million in the case of the hospital to demolish the existing

facility and construct a new one The drastic

solution in the case of the hospital was

necessitated by structural and soils failure

problems with the building, but the result wiU

be a new facility with greatly enhanced

nonstructural protection since it will be built

under 1990s California Hospital Act criteria

which exceed ordinary UBC requirements.

3. CURRENT EFFORTS

Some of the current efforts to deal with the

observed performance of nonstructural

components in the Loma Prieta earthquake

are educational in nature, rather than only a

392
matter of engineering. Many, perhaps most

building owners and users assumed the

building code was providing them with

extensive nonstructural protection, while in

fact the aims of the Uniform Building Code's

seismic regulations, as stated by the

Structural Engineers Association of

California (1990), are rather modest.

Protection against functional losses, such as flooding from an occasional broken pipe that causes a building to be shut down for repaks,

isn't promised. There are even many

nonstructural items that may cause serious

injury—tall file cabinets or chemicals on

shelves for example~that are not within the

scope of the building code, and are outside

the process by which drawings are produced

to guide the construction of a building under

the regulations of a building department. The

efforts to better educate clients so that they can either knowingly accept the risk of codeminimum design or select a higher (and

costlier) level, are not coordinated but are

rather the efforts of individual consultants or building code personnel. A standardized and

quotable explanation, with specific mention

of nonstructural components that are included

within the scope of the building code and

their intended performance levels, would be a useful addition to chapter one of the UBC.

More sophisticated clients, such as some

facilities managers of larger companies or agencies, knowledgeably discuss the subject

with their designers, but this is still unusual.

Short of a definitive public opinion survey

there is no way of determining what the

general public or typical building

owner/occupant expectation of nonstructural

seismic performance is, but among

earthquake engineers it is commonly stated

that the public expects to receive more than

the code intends to deliver. The following

three facts add up to a communication

problem:

1. The public commonly uses the term

"earthquake proof;"

2. Earthquake engineers know that codeminimum designs will experience extensive

nonstructural (and structural) damage in large

earthquakes;

3. Most buildings are designed to a codeminimum level (and many fall beneath that

level with respect to nonstructural

components—water heaters have often been

installed without seismic restraint in recent

construction in California, for example).

The solution, if there were to be excellent

communication between earthquake engineers

and the public over this issue, would not be

simply to universally raise minimum

requirements to ensure very high levels of

performance. Once the public found out the

cost of that extra protection, they would not

be willing to pay for the extra benefit. The

solution, if communication were better,

would probably be a matter of selective

voluntary use of higher standards for projects

where the client needs and can afford higher

protection (e.g., the labs of the corporation in

the above example), and a slight

improvement in the mandatory minimum

requirements for all buildings based on better

technical seismic protection techniques.

The State of California is now pursuing

improvements in nonstructural performance

with respect to schools and hospitals, the two

types of buildings for which the State

preempts local jurisdictions in the

enforcement of building regulations.

The Office of the State Architect, which

controls the building permit process for

schools in California, has been continually

updating its nonstructural regulations since

the 1950s for new construction and is now

also looking into the issue of the retrofit of

schools up to a current level of seismic

protection. The California Seismic Safety

Commission is currently investigating the

cost and implications of a statewide survey of

nonstructural hazards in schools. The

Berkeley Unified School District

commissioned a nonstructural seismic survey

of its schools after the Loma Prieta

earthquake (Reitherman and Kustu, 1990),

including prioritized hazard rankings, cost

estimates, and other necessary guidance to decide how to proceed. The same District

accomplished a structural seismic safety

study of its schools by SOH & Associates

along with a disaster planning effort by

Emergency Management Group.

If the structural, nonstructural, and

emergency preparedness needs found in

Berkeley are projected statewide, including

393
desirable improvements as well as bare

minimum estimated needs, the potential cost

would be enormous. With regard to nonstructural retrofitting, the only feasible

approach is to narrow down the scope of

evaluations to potential damage that is ti^uly

unacceptable, beginning (and perhaps

ending, for most schools) with items that

would cause serious injury. In the Berkeley

Unified School District nonstructural study,

approximately half the estimated

nonstructural life safety retrofit cost was

attributable to items in the highest of three

hazard categories, which is a common

finding: Half the cost can be eliminated if only the most serious life safety threats are

tackled. In many buildings, naiTowing down

the scope results in a nonstructural life safety

retrofit cost of fifty cents or a dollar per

square foot, instead of a figure several times

higher if more conservative life safety criteria are used. Extending the objectives to include

protection of property value and essential

functions has an even greater associated cost

increase factor than extra conservatism in the

life safety evaluation process.

With regard to hospitals, the Office of

Statewide Health Plannmg and Development

has commissioned a study of elevator

performance at medical facilities in the Loma

Prieta earthquake . Like suspended ceiling

systems, elevators ai"e a nonstructural system

that, even if constructed up to cuiTent

California requirements, don't seem to yet

provide acceptable performance. In both

cases, the systems have many elements, thus

complicating earthquake solutions, and the

construction industry developed the basic

product designs prior to the evolution of

seismic regulations. Some of the more

reliable or easier engineering solutions

interfere with the usual way these systems are installed or built, or increase cost

significantly. Solutions must involve

improvements in the products manufactured

by the construction industry, not just

individually engineered retrofit or new

installation designs.

Staehlin (1990) reported that fifty elevators in

hospitals required repairs as a result of the

Loma Prieta earthquake, and Schiff (1990)

catalogued several areas of concern with

regard to elevator safety and post-earthquake

functionality. Many tenants or owners in

buildings other than hospitals also concluded

after the earthquake that the performance of

their elevators was unsatisfactory, and it is likely that improved elevator design

requirements will be one result of the

earthquake for both market and regulatory

reasons.

The Association of Bay Area Governments

(1990) has completed a National Science

Foundation sponsored study of hazardous

materials incidents caused by the earthquake,

(there were over 1 50) and this is another area where some specific changes are likely to be

soon implemented. The topic of hazardous

materials and earthquakes includes a wide

vaiiety of nonstructural items-compressed

gas cylinders, fume hoods and ventilation

systems, piping, lab chemical bottles, etc. Some of these items, such as compressed gas

cylinders, are most likely to be regulated by a building operation safety code, such as the

Uniform Fire Code in the case of California

jurisdictions. There has been interest on the

part of the fire service and fire protection

engineering community responsible for

developing the UFC in incorporating

eaithquake engineering rules, and so it is a matter of developing workable

(predominantly prescriptive rather than

analytical) provisions. Hazardous material

piping issues have been studied by Kircher

and Loshkari (1990), and the applicable

avenue for incorporation of improved

performance may lie in the revision of piping

standards such as those of the American

Society of Mechanical Engineering. Unlike

some of the other nonstructural problem areas

like ceilings or light fixtures, the hazardous

material piping problem will get worse rather

than better unless future practice considerably

improves: CuiTcnt practice often neglects

earthquakes, and combined with the fact that

large amounts of such nonstructural

components are being installed every year,

the level of vulnerability will increase.

4. CONCLUSIONS

The Loma Prieta earthquake was of greater

stmctural or geological significance than for

its nonsti^uctural lessons, but there were still

394
some noteworthy effects within the scope of

this paper's topic:

1. Psychological effects: Building

tenants and owners, on the one hand, and

earthquake engineers and building code

personnel on the other, are dealing with the

disparity between perceived protection and

actual protection, and between the

performance expected by the earthquake

engineering community and by the general

public.

2. Technical developments: Elevators

and hazardous materials are two areas where

the development of improved design

solutions will probably be quickly reflected in revisions to models codes and standards.

With regard to most nonstructural

components, the earthquake validates current

approaches and emphasizes the need for

effective implementation of existing criteria.

5. REFERENCES

1. Ding, Day, Christopher Arnold, et al. (1990). "Architecture, Building

Contents, and Building Systems," in Lee

Benuska, ed., Loma Prieta Earthquake

Reconnaissance Report. Earthquake

Engineering Research Institute.

2. Soong, T.T., 1990. "Seismic

Performance of Nonstructural Elements

During The Loma Prieta Earthquake," in

Noel Raufaste, ed.. Wind and Seismic

Effects: Proceedings of the 22nd Joint

Meeting of the U.S.-Japan Cooperative

Program in Natural Resources Panel on Wind and Seismic Effects . National

Institute of Standards and Technology.

3. Robert Bruce, ed., 1991 in progress.

Seismic Design and Performance of

Equipment and Nonstructural Elements in Buildings and Industrial Structures . AppHed Technology Council.

4. Angell, Lisa, 1990. "Earthquake '89

Summary Report: Facilities Problems."

Hospital Council of Central and Northem

California course on Earthquake Hazard

Reduction and Emergency Planning for

Hospitals.

5. California Association of Hospitals and

Health Systems, 1990. "Hospital

Earthquake Preparedness: Issues for

Action: A Report on the Loma Prieta

r Earthquake."

6. Structural Engineers Association of

California, 1990. Recommended Lateral

Force Requirements and Commentary.

7. Reitherman, Robert and Onder Kustu,

1990. Nonstructural Seismic Hazard

Sui'vey for the Berkeley Unified School

District. The Reitherman Company, for

the Berkeley Unified School District.

8. Staehlin, William, 1990. Personal

communication to Anshel Schiff in Lee

Benuska, ed., Loma Prieta Earthquake

Reconnaissance Report. Earthquake

Engineering Research Institute.

9. Schiff, Anshel, 1990. "Performance of

Elevators," in Lee Benuska, ed., Loma

Prieta Earthquake Reconnaissance

Report. Earthquake Engineering Research

Institute.

10. Association of Bay Area Governments,

1990. Hazardous Materials Problems in

Earthquakes: A Guide to Their

Mitigation.

395
Table 1. Loma Prieta Earthquake Fatalities

CAUSE OF DEATH / COUNTY NUMBER

Collapse of Cypress "doubledecker" freeway / Alameda

Unreinforced masonry building damage/ San Francisco

Unreinforced masonry building damage/ Santa Cruz & Monterey

uweiiing coiiapse / cjan rrancisco

Fall on stairway / San Francisco

Fall from tower / Santa Clara & Monterey

Landslide on coastal highway / Santa Cruz

Smoke inhalation from natural gas caused fire / San Francisco

Gunshot wound to civilian while directing traffic / San Francisco

Motor vehicle accident, collapsed span of Bay Bridge / San Francisco

Motor vehicle accident with horse on highway / Santa Cruz

Carbon monoxide inhalation from emergency generator / Alameda

41

5

4

O

2

2

TOTAL 63

source: "Earthquake-Associated Deaths - California," Morbidity and Mortahty Weekly Report.

Vol. 38, No. 45, November 17, 1989, Centers for Disease Control.

396
State-of-the-Art Report on Seismic Design of Building Equipment

and Nonstructural Components in Japan

By

Masaya HiROSAWA ^, Hatsukazu MIZUNO \ and Mitsumasa MIDORIKAWA 3

SUMMARY

Recently in Japan, in the event of an earthquake, more investigations have been made into damage to building equipment, furniture and nonstructural

components such as interior and exterior finishing and

nonstructural walls rather than structural members, and

there have been not a few reports on the analysis of such

damage. Accompanied by this trend, seismic design

guidelines for nonstructural components have been

prepared under the supervision of the administrative

organizations concerned and some of the guidelines have

been used for actual construction. In this paper, we will survey damage to these

nonstructural components and members used for construction and introduce various related guidelines

which have recenUy been prepared to reveal their goals

and other principal contents.

KEY WORDS: Earthquake Resistance, Building

Equipment, Nonstructural Component, Seismic

Guideline, Confirmation, State-of-the-Art

1. INTRODUCTION

Recently, earthquake hitting cities with many

medium- and high-rise buildings have inflicted great

damage to nonstructural components. For example,

windows and water storage tanks at the roof are broken,

interior and exterior finishing come off, and indoor

furniture is tumbled. Also, as for reinforced concrete

buildings, what is called nonstructural walls have often caused a brittle failure of structural columns, and have

displaced doorway sashes to stop the inhabitants from

going in and out. Like collapsing buildings, such

breakage is in most cases very dangerous because it causes building inhabitants and passers->by to be injured or killed as well as significandy decreasing die functions

of the buildings which are expected to work in case of a

disaster.Table 1.1 shows damage to equipment machines

and so on in the recent severe earthquakes in Japan.

Also, Table 1.2 shows the history of various seismic

guidelines related to equipment machines and so on. As

seen in these tables, in the 1978 Miyagiken-oki

Earthquake which inflicted great damage mainly to Sendai, the most populous city in the Tohoku region,

major damage was done to large computers, many water

storage tanks at the roof, and a lot of other interior and

exterior finishing and equipment machines. Based on

this experience, the Ministry of Construction and other organizations concerned started to prepare seismic

regulations concerning main nonstructural components

for construcdon and equipment machines, while as seen

in Table 1.3, some related industries tried to meet the challenge by having a private organization (for example,

the Reinforced Plastic Association) prepare related

technical guidelines. That was the time when work was

under way to sweepingly revise the existing seismic

regulations concerning buildings based on the results of

the five-year national research & development program.

Therefore, preparation of new regulations on equipment

machines followed the example of this revisional work.

In 1980, laws and orders concerning the seismic

design methods for building structures saw a sweeping

revision for the first time in thirty years. In the revised

version, the goal of seismic safety was to allow

buildings to survive without collapse such a severe earthquake motion that they might possibly encounter

only once in their lives. For that purpose, it was

stipulated that ultimate strength should be designed by

using 1.0 as an elastic response shear force coefficient. The revised laws and orders (Enforcement Order

of die Building Standard Law) stipulated that in light of

the damage to nonstructural components in 1978,

chimneys and water tanks extruding from the roof

(Article 39-2) and water-supply and drainage and other

piping facilities (Article 129-2) should be safe in die event of earthquakes. After diis performance provisions were set, die Building Center of Japan published in 1982 under the supervision of the Ministry of

Construction "Guideline for Seismic Design and

Construction of Building Equipment" ^\ provisions

covering concrete seismic design mediods for the overall building equipment. Although these provisions cover

earthquake forces and displacements to be considered,

structural design methods, and recommended details,

which are noted in Section 2.2, the targeted seismic

safety is a little smaller than that for structures. Moreover, the 1983 revision of "Technical Guideline

and Commentary for Water Supply and Drainage

Equipment" ^ ^ involved the addition of seismic

countermeasures, which followed the above guidelines.

The Ministry of Construction notified each building

officials that these guidelines should be used for

reference in designing.

On the odier hand, immediately after the 1978

Miyagiken-oki Eardiquake, the Architectural Institute of

Japan established a committee, which then began

1) Dr. Engrg., Deputy Director General, Building

Research Institute, Ministry of Construction.

2) Ms. Engrg., Head of Large Scale Structure Testing

Div., Production Dept., BRI, MOC.

3) Dr. Engrg., Head of Earthquake Engineering. Div. II, USEE, BRI. MOC.

397
investigating the seismic safety of nonstructural

components for construction such as various interior and

exterior finishing. In 1985, the results of the

investigation were published as "Guideline and

Commentary for Seismic Design and Guideline for Seismic Design and Construction" ^\ As previously

stated, these guidelines were mainly intended for nonstructural components such as interior and exterior

finishing but are not often used for actual designs now

partly because the 1975 version of laws and orders do

not contain any provisions for nonstructural

components.

For government buildings which are often used

as disaster prevention bases in the event of a disaster, the Government Building Division, the Ministry of

Construction had had its own regulations based on the buildings important factors, but in 1987, it prepared a

seismic design guideline chiefly intended for interior and

exterior finishing ^\ This technical guideline, whose

details are noted in Section 2.3, is now used to design

many government buildings.

On top of the above examples, another building

element which is likely to suffer from earthquakes is reinforced concrete nonstructural walls in reinforced

concrete buildings. Despite repeated experiments since

the 1968 Tokachi-oki Earthquake where many related

disasters were observed, there has been no coordinated

technical guideline available yet. Also, there is something to be modified in the particular sections of

the Japariese Industrial Standard which provide for design

seismic forces and other matters for various vending

machines, which are often installed in the middle floors of the building these days. In addition, dedicated

investigations have been made into the access flooring

system which is used to install computers.

Seismic guidelines on these nonstructural walls,

vending machines, and free access flooring systems are covered in Section 2.5. In addition to the several nonstructural structures stated above, chimneys and reinforced concrete block

fences are examples of building belongings which have

often sustained damage from earthquakes and are provided with well-coordinated technical guidelines, and

equipment piping for nuclear power plant facilities is a

nonstructural element which is scrupulously designed to resist great seismic forces. On the contrary,

nonstructural components for buildings with a base

isolation system or with an active seismic response

control system are a typical example for which no

seismic technical guidelines have ever been considered.

This report does not refer to these matters for want of

space. Chapter 2 of this report gives the outline of the

typical guidelines which are used for actual design.

Moreover, Chapter 3 describes the seismic design for several equipment machines covered by these guidelines.

This chapter is divided into several sections each of

which refers to design seismic forces or other major

items of those guidelines.

2. OUTIJNE OF TYPICAI TFCHNICAL GUIDELINES

2.1 Outline

Each of the following sections covers the guidelines listed below as typical seismic guidelines for several building nonstructural components and outlines

their contents.

a) Guideline for Seismic Design and Construction of

Building Equipment (Building Center of Japan, revised

in 1984) 3). 3) b) Guideline for Seismic Design of Building

Nonstructural Components (Public Buildings

Association, 1987) ^) c) Technical Guideline and Commentary for Water

Supply and Drainage Equipment -1983 version

(Building Center of Japan, 1983) 1') Similarly, Section 2.5 outlines seismic

guidelines for reinforced concrete nonstructural walls,

and vending machines and free access flooring systems

installed inside and outside the building as examples for which several guidelines have been made but still have

much to be modified.

2.2 Guideline for Seismic Design of Building

Equipment (Building Center of Janan. 1984^ 3). 3')

2.2.1 Background

The Building Center of Japan, a juridical

foundation supervised by the Ministry of Construction,

compiled this guideline in 1982. Damage to nonstructural components and building equipment

during the 1978 Miyagiken-oki Earthquake motivated

intensive studies on seismic performance of them. The

Enforcement Order, Revised Building Standard Law ^) executed in 1981 includes specifications on seismic

safety of water tanks, chimneys, etc. extruding from

roofs, piping equipment and elevators. Notification No. 1101 of the Ministry of

Construction, 1981 '^^ specifies lateral design seismic

forces of water tanks, chimneys, etc. extruding from

roofs. The lateral shear force, P, for water tanks, etc. extruding from roofs shall be determined from the following formula:

P = k«W (2.1)

where,

P = lateral design seismic force acting on water

tanks, etc.; k = design seismic force coefficient of water

tanks, etc. which shall be determined by multiplying a

value more than 1.0 by a value of seismic zone factor Z; Z = seismic zone factor specified in article 88,

Enforcement Order, Building Standard Law. The

Minister of Construction specifies a value within a range between 1.0 and 0.7 according to the extent of

earthquake damage, seismic activity and other seismic

characteristics based on the record of earthquake in the region concerned; and,

W = sum of dead loads of water tanks, etc. and

supporting parts of structure, and live loads of water

tanks, etc. (unit: kgf).

The lateral design seismic force on water tanks,

etc. can be reduced to a half of that mentioned above, if some countermeasures of preventing damage and harm

398
to human lives due to lateral movement and toppling of

water tanks, etc. during earthquakes are effectively done.

The guideline conforms to the Enforcement

Order, Building Standard Law and Notification No. 1101

of the Ministry of Construction, 1981 mentioned above,

and includes knowledge of some studies, new

countermeasures from practical viewpoint and some

design examples. The scope of the guideline is: (1) Installation of building equipment in buildings of more than three stories and their heights not taller than

60 meters. Seismic performance of equipment itself is supposed to be confirmed by equipment manufacturers;

and

(2) Installation of equipment lighter than 100 kgf is excluded, if equipment is installed on carefully treated surface of structure in a way specified by equipment

manufacturers. The guideline adopted two types of design

seismic force on equipment; One is the force specified in

the local seismic coefficient method, another the force

specified in the modified seismic coefficient method.

Both includes acceleration amplification effects with

stories of building, and the latter explicitly considers

acceleration amplification effects of equipment.

2.2.2 Seismic Force 1 Lateral design seismic forces to equipment

installation are the followings and act on the center of

the gravity in principle:

Fh = Kh-W (2.2) Kh = Z.Ks (2.3) where,

Fh = lateral design seismic force; Ks = standard design seismic force coefficient

(Table 2.1); and

Z = seismic zone factor.

If vertical seismic forces are considered, the vertical design seismic force is given as follows: Fv = Kv«W (2.4) Kv = Kh/2 (2.5) where,

Fv = vertical design seismic force; and

Kv = vertical design seismic force coefficient

If a tank with free water surface is considered, W of the above expression is substituted for Wq.

Wq = ttT^W (2.6)

OCqp is prepared in figures for cylinder and ball tanks, and

in formula for rectangular, cylinder and ball tanks. Namely the lateral design seismic force to a water tank

is given as follows: Fh = Kh'Wq = Z»Ks«aT»W (2.7) The force acts at the following height, hog, from the bottom:

hog = PT*h (2.8) where,

h = equivalent height of a water tank (unit: cm);

(1) full height for steel tanks and the like; and (2) height

of the center of overflow pipe for a fiber-reinforced

plastics tank. In a case of hopper-type overflow, the height shall be the sum of height of water surface and

radius of overflow pipe; and

Px = ratio of force-acting height hog and

equivalent height h. Figures and formula are prepared.

Liquid tanks with free surface shall be designed in a similar way as water tanks.

2.2.3 Seismic Force II Lateral seismic forces to equipment installation

are: Fh = Kh-W (2.9) KH = Z«I»Ki«K2»ko (2.10)

where,

Kh = lateral design seismic force coefficient; W = weight of a nonstructural component or equipment (unit: kgf);

Z = seismic zone factor (0.7-1.0);

I = importance reduction factor (1.0, 2/3); Ki - floor response amplification factor of a

building (1-10/3);

K2 = response amplification factor of a nonstructural component or equipment (1.0-1.5 -2.0);

and

ko = standard design seismic force coefficient

(0.3).

If vertical seismic forces are considered, the

vertical design seismic force coefficient Ky is given as:

(1) supported on a vibration-reduction apparatus;

Kv = Kh/2 (2.11)

(2) fixed with base, floor, etc. by anchor bolts,

etc.* Kv = Kh/(2K2) (2.12)

Importance reduction factor I is given as shown

in Table 2.2:

(1) important equipment I = 1.0; and

(2) general equipment I = 2/3. Building amplification factor Ki is given as shown in Fig. 2.1: Ki = Ag ; penthouse (2.13) = l+(AB-l)»h/H ; general floors = 1 ; basement and first floor where,

H = height fi-om ground level; and

h = height of equipment-installed floor (unit m).

Ab means response amplification at the roof, and

is given as follows: Ab =10/3 ;Tt<0.6 (2.14) = 10/3-2/3(Tt-1) ;0.6<Tt<1.2 = 3.2A't ; 1-2<Tt

Fundamental natural period Tj (sec) is evaluated

as follows: Tt = (0.02+0.0 la)«H (2.15)

The coefficient a is the ratio of the total height of steel construction stories to the entire height of a building. If all stories over the ground level are of steel construction, a = 1. If all stories are of reinforced concrete or of steel and reinforced concrete composite

construction, a = 0. Equipment amplification factor K2 is given as follows:

(1) fundamental natural frequency fm (Hz) is unknown, K2 = 2; and

(2) fundamental natural frequency f^ is known.

399
1) in a case of fm^l5, K2 is given for the ratio fb/fm Fig. 2.2. In a case of important equipment, a value of I«Ki«K2«ko shall not be less than 0.6, and in a case of general equipment, shall not be less than 0.4. 2) in a case of fm>15, K2 = 1.

2.3 Guideline for Seismic Design of Building

Nonstructural Components (Public Buildings

Association. ]9&7) ^)

The contents of the guideline are the following

:

Chapter 1. General

1.1 Purpose

1.2 Scope

1.3 Target of Safety

Chapter 2. Seismic Design

2.1 Inspection for Seismic Safety of

Nonstructural Components

2.2 Deformation Capacity against Story Drifts and Nonstructural Components

2.3 Inspection for Strength Capacity of

Nonstructural Components

2.4 Inspection for Overturning Possibility of

Nonstructural Components

Chapter 3. Design of Details

3.1 Principles of Detail Design

3.2 Exterior walls and their Finishing

3.4 Ceiling and Floor Materials

3.5 Opening Members such as Windows

and Doors

3.6 Furniture and Equipment

3.7 Computers and Associated Equipment

2.3.1 General

The purpose of this guideline is to provide

minimum standards for use in nonstructural component

design and inspection for the seismic resistant capacity

of nonstructural components mainly in government

buildings to maintain the seismic safety of total building systems in a major earthquake.

The guideline applies to all nonstructural

components in buildings not taller than 60 meters. Nonstructural components include all members except

for structural members, equipment, chimneys, parapets,

exterior stairs, and miscellaneous structures, such as: a) exterior walls and their finishing; b) exterior fixture; c)

partition walls and interior materials including interior doors; d) ceiling and floor materials; e) furniture and

equipment; and f) computers and associated equipment.

The target of safety is as follows:

(1) In inspecting the safety of the earthquake resistant capacity of a government building, the target of the seismic performance in a major earthquake is decided by

taking into consideration the functions to be performed

and the location conditions;

(2) Activity section rooms in a government building to be used as disaster prevention bases in a major

earthquake shall possess the seismic resistant capacity to perform their functions after a disaster; (3) The target of seismic resistant capacity of

nonstructural components, as shown in Table 2.3, is the following:

a) Ordinary government buildings shall be guaranteed to be safeguard against loss of life and secondary disasters; and

b) Activity section rooms in a government building to be used as disaster prevention bases in a major

earthquake shall be secured their functions in addition to the condition mentioned above; and

(4) In the seismic design of nonstructural components,

the target of seismic resistant capacity shall be classified

into two groups, exterior and interior, in a building:

a) All exterior parts of a building shall be guaranteed not

to fall down due to break; and

b) The interior parts of a building shall be designed by

distinguishing the activity section rooms from the

others.

2.3.2 Seismic Design

Nonstructural components shall be designed to possess the seismic resistant capacity against expected

story drifts and seismic forces in a major earthquake.

The following inspections are required in the seismic design of nonstructm-al components:

a) Inspection for the deformation capacity against story

drifts; b) Inspection for the strength capacity against seismic

forces; and

c) Inspection for the possibility of overturning and

sliding.All of above inspections are not always required

depending on conditions such as characteristics of

nonstructural components, fixing positions, and fixing

types. Fig. 2.3 shows the general flow of the seismic

design of nonstructural components.

In the inspection for the strength capacity of

nonstructural components, the procedures are in the following:

(1) Nonstructural components and their joints, if necessary, are designed to resist the design seismic

forces determined from the calculations;

(2) The design seismic forces for a nonstructural

component shall be calculated considering the floor response amplification of the building concerned, the response amplification of the nonstructural component

itself, the seismic zone factor, and the importance factor; (3) The design seismic forces may be assumed to act noncurrently in the following direction: a)horizontal (in- plane); b)horizontal (out-of-plane); and c)vertical

directions; and

(4) In the design of nonstructural component joints, the maximum working stresses shall not exceed the allowable stresses of material under temporary loads. The design seismic forces for a nonstructural

component is determined according to the "Overall

Design Standard for Earthquake Resistant Capacity of

Government Buildings" prepared by the Minisu^y of

Construction, which includes seismic: a) forces based on

the modified seismic force coefficient; b) forces based on

the local seismic force coefficient; and c) forces

determined from the rigorous analysis procedure.

According to the modified seismic force

coefficient, the lateral design seismic force, F^, shall be

determined from the following formula:

400
Fh = Kh»W (2.16)

Kh = Z.I.Ki«K2*ko (2.17)

where,

Kh = lateral design seismic force coefficient; W = weight of a nonstructural component or equipment (unit: kgf);

Z = seismic zone factor (0.7-1.0);

I = importance reduction factor (1.0, 2/3); Ki = floor response amplification factor of a buUding (1-10/3);

K2 = response amplification factor of a

nonstructural component or equipment (1.0-1.5-2.0);

and

ko = standard design seismic force coefficient

(0.3). The vertical design seismic force, Fy, shall be

determined by the following formula: Fv = KvW (2.18)

where,

Kv = vertical design seismic force coefficient

(=Kh/2).

In the formula (2.17), the value of I«Ki«K2»ko shall be

not less than 0.6 and 0.3 for the important and ordinary

nonstructural components, respectively.

2.3.3 Design of Details

In the design of nonstructural components, the

details shall be examined by properly grasping their individual characteristics and their fixing conditions on

the building.

The detail design of nonstructural components

depends on their fixing conditions which can be

classified into three groups such as surface, line, and

point joints.

The seismic effects on nonstructural components

depend on the fixing conditions mentioned above. The

nonstructural components such as exterior tiles and

mortar with surface joint are dominated by the effects of

story drifts. The inspection for the strength capacity

against seismic forces is required for the nonstructural components such as ceiling with point joint. The

inspection for both deformation and strength capacity is required under certain circumstances for nonstructural components with line joint.

2.4 Technical Guideline and Commentary for

Water Sur)nlv and Drainage Eauipment (1983

version) i')

2.4.1 OLftline

Based on the 1970 revision of the laws and

orders, technical guidelines for building water supply

and drainage equipment were published as "Technical

Guideline and Commentary for Water Supply and

Drainage Equipment", but this first edition had no

regulations on seismic design. Later, the 1980's revised

version stipulated that safety in the event of earthquakes

should be considered for storage tanks, piping, and

equipment, resulting in the 1982 seismic design

guideline -^^ intended for the whole equipment. The

guideline covers water supply and drainage equipment

and piping.

This guideline is extensively used for actual design partly thanks to the administrative guidance ^\

Thus, when the guideline ^

) was revised in 1983, the

related part of the above equipment guideline was added

to it as a new section. Since then, the seismic

regulations of the guideline have been almost the same

as the technical guideline ) shown in Section 2.2. Here, we focus on piping and outline its related

guidelines, which are not covered in Section 2.2.

2.4.2 Scope

This guideline mainly covers equipment such as water storage and supply tanks, and water supply and

drainage pipes. Also, as for seismic design, it refers to

a) design seismic force, b) seismic measures at the base

of equipment, and c) seismic measures for piping.

2.4.3 Seismic Measures for Piping

Principles

Piping sometimes suffers damage due to the deformation of buildings forced by earthquakes and

excessive vibrations of the piping itself. To avoid such

damage, the seismic force working on the middle floor piping is considered as a horizontal force equal to 60%

of the dead weight of the piping and the allowable stress design method is applied for the working stress generated at each element by this horizontal force. In addition, if the story drift angle Rd caused by the seismic force is not to be particularly considered, use Rp = 1/200 and appropriate expansion joints. To be

specific, seismic support intervals for the piping, the cross section of the support members, and required

erection anchors are determined to meet the above

conditions.

Piping through Expansion Joint of Buildings

For piping though the expansion joint of

buildings, the relative displacement 6d of both

buildings shall be defined as follows. Flexible joints to be provided (expressly illustrated) must resist the following displacement:

6D = 2RD»h (2.19)

Here, h is the height from the ground to where the piping runs.

Connections between Eauipment and Piping

Basically, the equipment shall be fixed, and the piping shall be supported by sliding supports and the

like so that it will not be subjected to excessive

displacement. Then, flexible joints are used to connect them (illustrated).

Horizontal Piping

Horizontal piping is given seismic supports at an

interval appropriate to impede excessive displacement in the axially perpendicular direction on the occasion of

earthquakes. On the assumption that fixed-type or supporting-type seismic supports are used, a convenient

table shows various kinds of piping, how to actually

support them and to select the cross section of the

401
supporting members and required anchors according to the intervals of the seismic supports.

Vertical Piping

Vertical piping is given seismic supports to hamper excessive displacement in the axially

perpendicular direction in lime of earthquakes and to flexibly resist the story drift of the building. Like the case of horizontal piping, a convenient table is given to be utilized for design.

2.5 Seismic Guidelines for Other Nonstructural

Components

2.5.1 Guidelines for Reinforced Concrete

Nonstructural Walls

(1) Outline

In the 1968 Tokachi-oki Earthquake, two to four

story reinforced concrete school buildings particularly

sustained heavy damage. Most damage was caused by

the nonstructural and hanging walls (These are known

generically as secondary walls. Refer to Fig. 2.4)

installed in the reinforced concrete frames without

making any calculations for these elements. These

walls shortened the clear height of the structural columns, increasing stiffness to concentrate stresses, and

in turn causing brittle shear failures in many columns.

Since then, secondary wall-caused damage to reinforced concrete buildings has been observed in many

earthquakes at home and overseas, amounting to one of

the important specific challenges which must be overcome to minimize seismic damage. Though many

experimental investigations have been made since the 1968 Tokachi-oki Earthquake to produce certain results, there have not yet been any measures commonly

applicable to a variety of reinforced concrete structures. The current specific measures for such secondary walls

are as follows:

a) as for ultimate strength design methods where ideal seismic behaviors are pursued, secondary walls are used

in such ideal forms as perfect slits b) as for general reinforced concrete structures, a method

of making the thickness of secondary walls, t, small

enough for the column width, b, (t/b<l/6 and t<10 cm)

is illustrated in the seismic technical guideline and

c) as for thick-walled frame structures which have

substantial stiffness and strength, recommended standard

secondary walls are detailed ^\

The above third problem will be detailed below.

(2) Secondary Walls in Guideline for Structural

Design of High-rise Walled Frame Structures

Outline

Intended for six to eleven story apartment houses

(now applicable to as fifteen story buildings), these

structures contain a frame structure composed of flat rectangular-sectional wall columns in the longitudinal

direction and a shear wall structure with surrounding

wall columns in the span direction (refer to Fig. 2.5). It

is necessary to specify how to treat secondary structures because they are mainly built in the longitudinal

direction but are essential as partitioning walls or

exterior walls which arc needed for apartment houses to work properly. The superstructure in the longitudinal

direction is a frame structure and a failure mechanism is assumed to be a total failure mechanism which is ductile enough to develop yielding hinges at the end of the beams and the wall base in time of severe earthquakes.

However, since the structure has extremely high elastic rigidity and a short fundamental natural period of

approximately 0.012h (h: height of a building), the required horizontal bearing capacity is defined as 0.35 or more in terms of yielding base shear coefficient to obtain a response deflection of 1/100 or less in terms of

drift angle.

Recommended Details of Secondary Walls

It is recommended that anything but wall

columns used for structural members should be

considered as secondary walls, and that secondary walls should have a thickness of 12 cm or so, and out of four

sides in contact with the structure, at least two should

be notched into slits because they do not have any

adverse effects on superstructure behaviors on the occasion of earthquakes (Fig. 2.6). Incidentally, it was

revealed that in the failure experiments on semi-full and

full-scaled structural frames with secondary walls

complying with these details, the secondary walls

ruptured before developing major displacement and the

subject structure showed the expected inelastic deformation capacity.

2.5.2 Seismic Measures for Vending Machines

(1) Outline

At present in Japan, there are various vending

machines installed on the street or in the building and

the variety is on the increase. Some of them are for bottled drinks and so heavy that they can probably

damage as if they should be tumbled. Thus, by listing

design forces, the Japan Industrial Standard (JISB-8562,

1985) recommended that before installing vending

machines, not only impact by human bodies and wind

forces but also seismic forces should be investigated to obtain required strength capacity.

(2) Design Seismic Forces and Their Problems

It is stipulated that for a vending machine of

weight W (its height of gravitational center is h), a 0.4W horizontal force and a 0.4W»h overturning

moment on the installation surface should be considered

to enable the machine to resist seismic forces. There is no provision, however, on the location of installations or the allowable strength capacity of anchor bolts. Therefore, this JIS can be applied even to the case where the machines are installed in the middle

floors of the building. In that case, what is called floor responses should be considered to increase design

stresses, which has not yet been covered by this JIS.

2.5.3 Free Access Floors in Computer Room

The design and construction of free access floors

in a computer room are specified in guidelines of Refs.

402
5 and 4. The lateral design seismic forces are determined

basically by the formula (2.16). The design target of each guideline is the following:

(1) Guideline of Ref. 5 (Public Building

Association, 1987)

Free access floors in a computer room shall resist the lateral seismic design forces, due to a major

earthquake, without brittle failure or collapse to secure

the ways of refuge. This guideline applies to the computer rooms of ordinary use except for those to be

used in a major earthquake.

In the design of supports for free access floors, the maximum working stresses shall not exceed the allowable stresses of material against lateral design

seismic forces, and principally the strength capacity

shall be confirmed by an experiment.

(2) Guideline of Ref. 4 (Architectural Institute of

Japan, 1985)

The extent of damage in nonstructural

components is divided into five ranks according to the amount of damage, the necessity for repair and

exchange, the fall down, and/or the deterioration of

important functions. The allowable limit of damage due to a moderate

or major earthquake, as the design target of

nonstructural components against story drifts, shall be

provided by one of five ranks according to the importance of a building, the type of nonstructural

components, the effects on refuge, and so on. In the design, the maximum working stresses

shall not exceed the allowable stresses of material

against lateral design seismic forces.

3. COMPARISON OF SEISMIC DESIGN

METHODS IN RELATED TECHNICAL GUIDELINES

3. 1 Outline

In this chapter, we refer to the design seismic

forces, design methods for each element, and other matters covered by the above guidelines for several nonstructural components, while making a comparison

with design methods for building structures. In

addition, we outline the seismic safety levels aimed at by each guideline, the actual conditions pf nonstructural component design, and the confirmation methods.

3.2 Principles of Seismic Design

The lateral design seismic force coefficient, gk,

of equipment and nonstructural components is generjdly

given by the following formula:

gk = Z«gK»gko (3.1) where,

gk = lateral design seismic force coefficient; Z = seismic zone factor (0.7-1.0);

gK = factor due to floor response amplification of

a building, response amplification of a nonstructural

component, importance, and so on; and

gko = standard design seismic force coefficient

The value of gko is 0.2 in Rcfs. 3, 3' and 5. The

value of gK'gko for important equipment and

nonstructural components is increased by 1.5 times that for general ones. These values are determined by

expecting a major earthquake of maximum ground

acceleration of 0.2-0.3 g (g: acceleration of gravity),

which is a smaller level of maximum ground

acceleration for the seismic design of building

structures. The value of gK is 2-10 in Refs. 3 and 3', and

1.5-10 in Ref. 5. The allowable stress design is basically required

for equipment and nonstructural components against the

lateral design seismic forces determined by the formula

(3.1), while the ultimate strength design is principally

required for building structures. The formula (3.1) is compared in detail with each

guideline in the following section.

3.3 Design Seisnnic Forces and Deformation

Copacit y

Table 3.1 summarizes two typical guidelines for the seismic design of equipment and nonstructural

components. The table includes the seismic structural design of example buildings (31 m < height < 60 m) for

reference.Design seismic forces to building equipment and

nonstructural components are given based on a seismic

force coefficient. This is derived from specification of

local seismic coefficient to water tanks, chimneys, etc. extruding from roofs specified in Notification No. 1101

of the Ministry of Construction, 1981, Building

Standard Law. The design seismic force coefficient of

building equipment and nonstructural components is determined by the local seismic coefficient method or by

the modified seismic coefficient method. Both includes

acceleration amplification effects with stories of a

building, and the latter includes explicitly those effects of equipment. An importance factor is also a parameter

of seismic force coefficient. Building owners and

structural designers can determine whether specific

building equipment and nonstructural components are general or important.

Representative seismic coefficient values of

general building equipment and nonstructural

components at the first floor level are compared in Table

3.1 with a representative base shear coefficient value of

buildings higher than 3 1 meters and not taller than 60

meters. The seismic zone factor Z is assumed 1.0, and

the structural characteristics factor Rt is also assumed

1.0 in building base shear determination. The design

seismic force coefficient is 0.4 in the local seismic

coefficient method, and 0.2 in the modified seismic

coefficient method. These values are equivalent to a design base shear coefficient of 0.2 for the first phase

design or for possible several moderate earthquake

ground motions during building lifetime. The design of

members are based on allowable stresses for temporary

loads in both building equipment and nonstructural

components, and building structures. In addition, the

structural design requires a check of ultimate strength of

403
an overall structure against a design base shear

coefficient of 1.0. This check is on the basis of

possible one major earthquake ground motion during

building lifetime. The seismic design of building equipment and

nonstructural components requires a deformation

capacity check in addition to a stress check. Buildings

are designed for moderate earthquake ground motions so

that story drift angles are less than 1/200. Vertical

pipes of buildings are designed so that the pipes can deform to 1/200 story drift angles without damage.

Pipes through expansion joints are designed so that the pipes can deform to maximum relative displacements

among two structures in possible reverse phases. The

guidelines include several detail examples to secure enough deformation capacity.

The seismic design of building equipment and

nonstructural components for major earthquake ground

motions (Ref. 5) assumes the following story drift angles of three types of buildings: a) Steel Structure - more than 1/100 story drift angles; b) Reinforced

Concrete Structure, and Steel/Reinforced Concrete

Composite Structure with comparatively a few walls - less than 1/100 story drift angles; and c) Low-rise and

Medium-rise Reinforced Concrete Structure, and

Steel/Reinforced Concrete Composite Structure with

comparatively many walls - less than 1/200 story drift angles. Story drift angles in building structures during a major earthquake are evaluated from those for moderate

earthquake ground motions in the presented formulae of

Ref. 5. As the first step of the seismic design of

building equipment and nonstructural components, the maximum story drift angle of a structure shall be

estimated and evaluated. Then, the prospected damage of

building equipment and nonstructural components shall be compared with that of a building structure, and

suitable ones shall be selected.

3.4 Seismic Design Methods for Equipment

Machine. Piping Bodies, and Their Supporting

Members

(1) Outline

Metallic materials are often used for various water tanks, boilers, equipment machines, storage boxes

and piping bodies. Lately, FRP (fiber reinforced

plastics) has sometimes been used for storage tanks. Also, steel angle members are used as material for supporting frames used to support piping against the

structure. In designing each member such as a machine

body by one of these materials, the allowable stress design is often used to control working stresses generated in each member by the design force noted in Section 3.3.

(2) Stress Analysis of Each Member

A working stress generated in each member is approximately in elastic calculated on the basis of the

elastic stiffness of each material.

(3) Allowable Stress of Each Member

The Enforcement Order, Building Standard Law

stipulates that the dead weight of each machine and the weight of its contents should be defined as permanent

loads and these weights plus seismic forces as temporary

loads. In case of the materials covered by JIS, it lists allowable stresses for separate loads. As seen in Table

3.2, the temporary allowable stress of metallic material

is the standardized yield point strength of that material. Incidentally, in normal cases, the standardized yield

point strength is approximately 1/1.1 to 1/1.3 and never below the actual yield point strength of the material

although it depends on the type of steel material. The

0.2% offset point strength is used for steel materials

which have opaque yield points. Moreover, the order

stipulates tJiat allowable stresses of adhesives and the

like which do not have any standards should be

determined by making experiments on it and then

setting appropriate safety factors on the results.

(4) Confirmation of Strength Capacity of Each

Member

In terms of the strength capacity, the safety of

various equipment machine bodies is usually confirmed

by checking to see if the working stress of each member

obtained as stated above is not over the temporary

allowable stress. Never available are those design

methods which use the ultimate strength and expect the

effects of inelastic deformation capacity as in building

superstructures.

3.5 Allowable Stress of Anchors

(1) Outline

What is called post-installed anchors are often used to attach equipment to reinforced concrete buildings or floor slabs. In the seismic actions for equipment and

piping covered in this report, such anchors are used in the following cases and play an important roll: a) installing equipment machines or vending machines on the floor slabs; b) attaching seismic supporting member of piping to walls; and

c) attaching prefabricated precast exterior panels.

Therefore, some related technical guidelines

provide for the allowable strength capacity of the postinstalled anchors. But there are still many problems

unsolved because of the variety of the products on the market

(2) Design Methods of Anchors

Like the equipment machine bodies noted in Section 3.4, the allowable stress design method is used

to design anchors, so it is necessary to set the allowable

strength capacity for anchors.

(3) Temporary Strength Capacity of Anchors

Since anchors are ordinarily subjected to tensile forces, shear forces, and/or their combined forces when

used to attach equipment machines, temporary strength

capacities are set for those working stresses as follows: Tensile bearing capacity. Ta

Ta = Oi«Min(Tai,Ta2, Ta3) (3.2)

404
where,

Tai = strength capacity which is caused by the

tensile yield of anchor and depends on their sectional

area, aA, and yield point stress, ay;

Ta2 = strength capacity which is caused by the cone-type failure of concrete and depends on the anchors' embedded length, 1a and concrete strength, Fc; Ta3 = strength capacity caused by the bond

failure of the connecting plane between adhesive and

concrete when adhesive anchors are used; and

0\ = strength reduction coefficient which

represents the safety factor of the material. Shear capacity. Oa

Qa = <D2»Min(Qai , Qai) S (3.3)

where,

Qai = bearing strength which is caused by the shear failure of anchor bodies and depends on aA and aA

and ay; and Qa2 = strength capacity which is caused by the

failure of concrete and is usually represented by a

function of Fc, Ec (elastic modulus for concrete) and aA- Safety with anchors subjected to combined

stresses(Q, T) shall be considered by using charts based on the following expression or other similar

alternatives. (Q/Qa)" + (T/Ta)" < 1 (3.4) Ordinarily, the value of 1 or 2 is used for n.

(4) Problems with Anchors

As stated above, anchors play an important role

in the seismic design of nonstructural components but

there are many problems unsolved, including those

under investigation:

a) There is no standard for industrial products

manufactures, while many products are on the market;

b) Few research materials are available in terms of the

effects of dynamic cyclic loading and creep phenomena

on strength capacities or load-deflection relationships;

c) There are few data on time-dependent durability or fire proof abilities; d) Few date are available in terms of how base concrete conditions and construction preciseness work on the

anchors; and

e) There are few date on limit conditions for combined

stresses.

3.6 Structural Details for Seismic

Countermeasures

(1) Outline

Since actions based on structural calculations

cannot cover the whole sphere of aseismicity

improvement for building equipment, structural details

to be adopted are often illustrated and recommended

specifically. Among such structural details, many have

been developed to cope with horizontal deformations

occurring in time of earthquakes, while some are related

to the strength capacity. These structural details are

divided in several major categories below.

(2) Structural Details for Seismic Deformations

The structural details are mainly related to piping

joints, the formation and joining of interior and exterior

finishing, structural strength capacity, and other matters.

a) Many of the structural details for piping joints are to control deformations occurring in time of earthquakes.

For example, as for piping through expansion joint

members as shown in Section 2.4, the detail describes

how to install flexible pipe joints or flexible joints in a manner that they will not be damaged by deformations

of possible intensity. Also it describes how to use

flexible joints where the piping enters the building from

outside so as to control uncalculated deformations which

are possible to occur vertically and horizontally.

b) The structural details related to the formation and

joining of various interior and exterior finishing 1) contain typical details about fasteners for precast curtain walls (fixed or sliding type), dry-type partition walls in steel buildings (flexible type), glazing compounds of

glass windows (clearance with frame members), and

seismic doors and ceilings (flexible enough to resist in- plane deformations).

c) The structural details for strength capacity include

details about reinforced concrete nonstructural walls

stated in Section 2.4 (1) and convenient tables for selecting appropriate seismic support members for piping (see Section 2.3), as well as the detailed figures

of structural joints for the supporting frames of

equipment machines (showing the shape of gusset plates for the end of braces and the position of rivet holes).

They are used to omit calculations for the strength

capacity except the cases of secondary walls.

3.7 Equipment Machine Design and

Confirmation

Today in Japan, licensed building engineers are responsible for the seismic designing of building

structures based on the related laws and orders

whereas building officials at the supervising authorities are in charge of structural calculations and structural drawings based on those calculations. On the other

hand, although building equipment and piping under

control of laws are similarly confirmed, there is no large control or building confirmation for various interior and

exterior finishing. Partly due to this, except the cases of official buildings, aseismicity has not necessarily

inspected for nonstructural components. Also, building

confirmation for building equipment is not as scrupulous as that for strength capacity.

4. CONCLUDING REMARKS

We reported above on the guidelines, and related actual design and confirmation regarding the seismic

design for building equipment and other nonstructural components or members. As stated, there are not a few

problems as some guidelines have much to be

coordinated or improved. These guidelines are established by several committees of learned and

experienced people, affiliated with the Ministry of Construction, and the Ministry of International Trade

405
and Industry. Researchers at the Building Research

Institute commit themselves to almost all these committees and offer them technical information so as

to maintain consistency among the guidelines.

Table 1.1 Examples of Recent Earthquake Danage to Building Equlpaents, Nonstructural Blenents. etc.

Vear Earthquake Contents of Typical Danage

1964 Nllgata Erq. Danage to sever pipe and Tloatlng-up ot septic tanks due to liquid-faction of sand layer

1965 Shizuoka Erq. Danage to vindov glass

1968 Tokachl-okl Erq. Breakage of roof chloney and independent chlnney

1978 Izuohshina-klnkal Danage to »ater and gass supply, electric facilities

Erq. due to landslide, danage to nedlcai facilities O 1978 Mlragiken-okl Erq. Danage to big computers, roof sater tanks, piping,

ducts, gass facilities, fire facilities and nany other nonstructural elenents

1983 NIhonkai-chubu Erq. Outdoor underground piping, septic tanks and oil tanks

1987 Chlbaken-Tohookl Erq. Sliding anf falling dovn of nany roof tiles

Table 1.2 Protress to EslabI Ishient of Aselsitc Design Code and Regulations

Regarding Nonstructural Building Eleaents

Tear-Monlh Hens. Code and Guideline *' Mark In ( ) 8ho«s the refference

nunber In this report

Iteas related tith Equipaent etc.

Erthquake DaaageCode or Order

Technical

Guldel Ine Structural Regulation

1976.3 End of the national project on nei aselsile design nethod (a) o

1976. T. G. for vater supply and drainage

equlpient

(1): Vlthout aselsnic regulations *' o1978.6 Miyaglken-okl Earthquake

1980~ Many technical guidelines by private associations (2. 1~2. 7) o

1980 large revlalon of the Building

Standard La> <b) o

1981. 1 T.C. for structural calculation basing the revised code (2.8) o

1981.6 Notification No. 1101 by MOC on the revlsend code 'b* (c) o

1982. 1 T.G. for building equlpients basing

on 'c' (3) o

1982.2 Director's Nontlflcatlon No. 32 by MOC recoiendin» to apply T.G. *3'(d) o

1988.2 Revliion of (1) (1'): Addition of aselsnle regulations o

1983. S Nihonkal-chubu Earthquake o

1984. 9 Revision of (3) - (3") o

198S. 11 T.C. for iselsalc design of nonstructural eleaents by AU (4) o

1987.2 T.G. for aselsalc design of nonstructural eleaents by G. B. S (S) o

1987. 10 T.G. for acdiua-high rised «all fraae construction by 6CJ (6) o

1988. 10 T.G. for reinforced concrete

buildings based on ultlaate strength

concept by AIJ (7) o

Note : •!) See note of table 1.3 concerning on the aark AIJ. G. B. S. . and AIJ.

•3) Mark O aean the Itea has priaary laporlance.

406
Table 1.3 (i) Trplcil Aieisalc Guidelines for Arcli! teetural !(onstructuril Eteaenti

.Equipaent etc.. and the Main Contents

No. In Table

1.2

Publisher *' (Supervision)

Year of Edition

(First Edition)

Title Main Subjetcts

Main Content! "

Design

Load

Design

Defftec.

Al lotable CapacltrDetails

I

( 1 )

B C J

(H. B. of mc)

t3t) (1978)

Technical standard and coaaentarj'

for «ater suppir and drainage

quipaent

Water suppir and drainage

equipaent

(J-

)

(3'

)

No Yes

3(3)

B C J (H. B. of XOC)

1934 (1992) Gidellne for aselsiic design and construction of building equipaent

Water tank, boiler, cooling to»er. air conditioner

piping, generator, etc.

Yes No Yes Yes

4 A I J 198S Recoaaendat ions for aseisaic design

and construction of nonstructural

eleaents

Curtain lall of RC. . PC. ALC. glass «lndo*. door,

eel 1 Ing. furni ture etc.

Yes Yes Yes Yes

5 P. B. A

(G. B. D. of MOO

1987 Guideline for aseisaic design for architectural nonstructural

eleaents

Interior and exterior

tai 1. eel 1 Ing and floor. <indo« and door, furniture

and equipaent, computer, access flooring srstea

Yes Yes Yes Yes

6 B C J (R.B. of MOC)

19S7 Guideline for structural calculation

and construction of aedlui-highrlsed tailed fraae reinforced concrete construction

(Structural guideline) Secondary »all with selsalc slitNo Yes No Yes

7 A I } 1999 Design guideline for earthquake

resistant reinforced 'concrete buildings based on ultlaate strength

concept

(Structural guideline) Nonstructural eleants(Exclusion of nonstructural eleaents

Tith unclear structural behavior)

Note *1 BCJ : Building Center of Japan. AU : Architectural Institute of Japan. P. B. A. : Pubric Building Association

H. B. or G.B.D. of MOC : Bousing Bureau or Goveraaent Building Division of Ministry of Construction

*2 Yes or No lean the guideline contains or does not contain regulation related (ith the ilea, respectively (3) aeans

the guideline applies the guideline No. 9° correspondingly related tIth the itea.

Table 1.3 (b) Technical Guidelines related vlth Building Equlpnents etc. Prepaired

by Private Sectors and Reffercd at Conipllatlon of the Guideline (3) (Except 2. 8)

No. In Table 1.2 Title of Guideline Year of EditionPubl Isher

2 . 1 Guideline for aseisaic design and construction of building equlpaents

1980 Building Center of Japan

2. 2 Guideline for aseisaic design of engine generator equlpaents

1981 Nippon Engine Generator

Association

2. 3 Guideline for aseisaic design of building equlpaents (Draft)1981 The Society of Heating. Air Conditioning and Sanitary

Engineering of Japan

2. 4 Data book for aseisaic design of building equlpaents1980 Building Bureau of Japan Telephon and Telegraph Public Corporation

2. 5 Aseisaic design standard for FRP latcr

tank (Draft) Structural calculation acthod for FRP Tater tank

1980

1981

The Japan Reinforced Plastics Association

2. 6 Manual for aseisaic design and construction of building electrical equlpaents

1983 Japan Electrical Construction

Assoc iat ton

2. 7 Guideline for aseisaic design and construction of copper piping for

bul Iding equlpaents

1384 Japan Copper Developaent Association

2. 8 Guideline for structural calculation

and coaaentarr basing on net aseisaic

code

1981 Building Center of Japan

407
Table 2.1 Standard Seismic Coefficient K,

Important equipment General equipment

The highest floor, roof and penthouse

1 . 5 (2. 0)

I . 0

The second floor and floors higher than the second floor

1 . 0 (1. 3)

0. 6

Basement floor and the first floor 0. 6 0. 4

NOTE Values in parentheses means seismic lateral coefficients

of Important equipment with vibration Isolator apparatus.

Table 2.2 Importance Reduction Factor I

Important

Building Equipment

General Building Equipment

importance Reduction

Factor 1

1 . 0 2/3

Table 2.3 Targots of Seismic Safety of Nonstructural Components

Target of seismic safety of nonstructural components

Required capacity

Class A

Buildings aimed to function as disaster

prevention bases

Safeguard of the human life

Guarantee of the functions required

In disaster pre- vention bases

Exterior of the building

(all parts) Interior of the bui Iding

{all spaces besides the activity sections)

Interior of the bui Iding

(activity sections and

the like)

- No fal 1 due to break

- No break or col lapse due to brittleness

- Damages. If any. should be local ones

1 Imi ted to nembcrs.

Joints, etc.

Class B

Government buildings

required for the sake of refuge and

first aid

Safeguard of the human life Guarantee of the functions required

In disaster pre- vention bases

Exterior of the building

(all parts) Interior of the bui Iding

(practically the totality)

- No fall due to break

- Damages. If any. should be local ones

1 imI ted to members, joints and the like

Class C

Ordinary government

but idings

Safeguard of the human life

Exterior of the bui Iding

(all parts) Interior of the bui Iding

(all parts)

- No fal 1 due to break

- No break or col lapse due to brl ttleness

Class D

Bui Idings storing or handl Ing dangerous

aaterials

Ditto Exterior of the bui Iding Interior of the bui iding

(all parts)

- Ditto

408
Roof

V Ki=Ai.

H

GL

V '

h /

Basement Floor

.Ki = 1

Fig. 2.1 Building Amplification Factor Ki

0 '—-J— uj 1 , 0.3 0.6 0.9 1.5 2 fb/f.

Fig. 2.2 Equipment Amplification Factor K

( f - g 1 5)

START

Inspection lalnljr for deforaatlon capacity

STEP 1

Classification of rooa functions against * severe earthquake

Estlnatlon of storx drift angle

In structures

STEP 2

Selection of nonstructural conponents

Inspection aalnly for overturning posslbllltr

Classification of rooii functions against a severe earthquake

Is 11 necessary

to check strength under >-

seisilc force?

inspection lainiy for strength capacity

YES

STEP4 Estliatlon of selsnic force

Classification of rooa

functions against a severe earthquake

Design of nonstructural coiponents and joints

Design of detal Is

END

Fig. 2. 3 General Flof of Selsaic Design of Nonstructural Coaponents

409
Bens (ith hree strength

and dcfonibilllr

Wall cohin tnd boi

with saae thickness

7.500 10,000

Fig. 2.4 Cross Section in Span Direction of Typical Japanese School Building

(Usually, columns In the outer frames are »ith nonstructural spandrel »all like this figure)

Shear vail vithout

inner beats

Broad vail cohio *ilh lane

strength and deforiabililr

Fig. 2.5 Outline of Structural System of

the R/C. Frame Structure with Wall Column

200mm n zr 1 \

/

\

n 200mm Tv7nvi

JAI

a) Arrangement of Structural Slit b) The Same

(Case A : one side and top of n.s. walls) (Case B : top and bottom of n.s. walls)

may be made

at one side

7- t-R11_

Anchar bar^^"^ ^ 120m;ii D10@400

c) Horizontal Section of Structural Slit

DI0«»400

Beam

d) Vertical Section

Fig. 2.6 Treatment of Non-structural Walls in HFW Construction*'

410
TABLE 3. J SEISMIC DESIGN METHODS OF BUILDING EQUIPMENT. NONSTRUCTURAL EQUIPMENT AND BUILDING STRUCTURES

NAME OF GUIDELINE or CODE

Iteas

GUIDELINE ON SEISMIC DESIGN OF BUILDING EQUIPMENT (1982) GUIDELINE OF NONSTRUCTURAL

COMPONENTS OF BUILDING (1987) BUILDING STRUCTURE

( BUILDING STANDARD LAf ) (for Reference)

Objects Selsale Design on Installation of

Bui idlng Equlpaent

Selsale Design of Nonstructural Coaponents In Bui Idlng Selsale Design of Buildings

Scope (1) Equlpaent Installation In buildings

on aore than three stories, and lith building height lover than 80 acters.

(2) Installation of equlpaent lighter

' than 100 kgf Is excluded.

(1) Goveraental and public buildings

lover than (0 neters

(2) All nonstructural components

other than building structure, building equlpaent. ehianeirs, pa raphe t s,

(an eiaaple]

Buildings higher than 31 aeters and lover than 60 aeters

Design Method Local Seisale

1.061 lie leni mcmou

Modified Selsale

i«oeiiicieni mcinoQ

One of the foiloving the aethods

(1) Modified selsnic coefficient

aethod

(2) Local selsale coefficient aethod

(3) Rigorous callculatlon

(Earthquake response analysis) Modified selsale coefficient aethod

seeas to be reconaended.

Equivalent Lateral Force Method (storr shear coefficient aethod) (Local selsale coefficient aethod for vater tanks, ehlaners. etc. protruding

froa roofs, and underground

(baseaent) structures }

Design Selsale Force

r N R H • t u • z u

there

Z • Selsale Zone Coefficient

Ki 'Local Selsale Coefficient

CO. 4,0. 6. 1.0

(I.J), 1.5. (2.0))

Hh " Z • 1 • Ki • !(« • Ko there

Z : Selsale Zone Coefficient

1 : Equlpaent laportance Reduction Factor

Ki: Building Aapllfleatlon Factor

K>: Equlpaent Aapliflcatlon Factor

ko: Standard Selsale Coefficient (0.3)

Modified selsale coefficient aethod

Is the saae as left.

C, • Z • R, • A, • Co vhcre

Ci : Storr Shear Coefflent

Z * Selsale ZonA Cnpfflrlpnt

Ri : Factor of Soil Profile Characteristics

Ai : Factor of Vertical Distribution of Selsale Forces

Co : Standard Shear Coefficient

Se i sa i c Zone Coeff iclenl

Z • 0. 7, 0.8, 0.9. 1.0 The saae as left. The saae as left. The saae as left.

Selsale Coefficient of General Equlpacnt at First Floor and

B;)senent or Base Shear Coef. of Building

0 . 4 (Lateral Selsale

Coef. of General Equlpaent at First

Floor and Baseaent

) tz-i ]

0 . 2 (Lateral Selsale Coef. of General Equlpaent at First Floor and Baseaent

) [l-li'U'\. 1-2/3. ko'

0.3 ]

The saae as left. Base Shear Coef. C

i

Ct ^ 0. 2 for Moderate Eq. Ground Motions

Ci 2 1.0 for Major Eq. Ground Motions

(Cheek of Ultiaate Strength) tZ-A,-R,-l ]

Building

laportance

Factor

Governnental and public buildings

are classMfled into three categories on the basis of required selsale perforaance: The lovest correspondes

the level of Bldg. Stan. Lav. The

helghest Is selsale perforaance of disaster relief and rehabilitation

bases after aajor earthquakes. Specific perforaance is not shovn.

Bldg. Stan. Lav. declares providing ainiaua standards concerning the site, structure

. equlpaent. and use of bulldl

ngs*. Building ovners and designers can deteralne target

level of selsale perforaance

higher than ainlaun standards

specified in Bldg. Stan. Lav.

Bui Idlng

Aapl If Icatlon Factor

K.

Not eiplicitlr con

sidered. But local selsale coefs. are

given for three location trpes of

stories.

K, r •- 10/3 The saae as left.

A factor At of the above selsale force includes acceleration aaplif Icatlon effects

vlth estlaated fundaaental period of building.

Equlpaent

laportance Reduction

Factor

I

Local selsale coef

s. are given for general eqipaent

and iaportant

equlpaent. Bldg. ovners and str. de signers can select

the value.

I • 2/3 - 1.0

Bui Idlng ovners and str. designers can dcteralne

laportance and select

the value.

The saae as left.

411
TABLE 3. I SEISMIC DESIGN METHODS OF BUILDING EQUIPMENT. NONSTRUCTURAL EQUIPMENT AND BUILDING STRUCTURES

(Continued)

Equlpient /

Conponenls

Aiipllf Icatlon FactorK.

Not considered. K, • 1 - 2 The sane as left In principle. Specific values of the factor are given socc nonstructural coaponents

(Joints of PC panel, etc.. celling,

furni ture).

Range of Design

Selsale Coefficient KnKh • 0 . 2 8 - 2 Kh 0. 1 4 - 2 The saae as left. Design selsalc forces are auch affected by the above cocfs. and factors.

Selsalc Design

of Equlpnent /

Coaponents

Seismic perforaance of equlpacnt Itself is supposed to be conflraed bjr aanufacturers. Selsalc design procedures of soae nonstructural coaponents (exterior

tall, partion tall, celling, floor, free access floor, furniture, coaputer, etc.) are presented.

Notification. Notification

Order and regulation spedify

lateral selsalc force of tatcr

tanks, chlaneys. etc. protru- ding froa roofs.

Dcforaabllltr of Equlpscnt /

Components

No connents for equlpaent In general. Vertical pipes shall bo subjected to storr

drift of I/70Q raHlsn of stnr-r h(>lffht in principle. Pipes through expansion Joints

shall be subjected to possible aailaun

relative displaceaent aaong two structures. Several datall exaaples of piping at expanti

on Joints and service connections to bldg.

to absorb possible displaccaents.

Possible aaxlaua story deflection

angles are estlaated as follots;

Steel str. - aorc than 1/100 radian.

Relnf. con. str.. and steel and rein

f. coaposlte str. tlth conparat Iveir

fct tails - aore than 1/100 rad. Mldlua-rlse and lot-rise relnf. con.

str. and steel and relnf. eon. conp.

sir. tlth aanr talis - aore than

1/200 rad. Evaluation foraulae arc shotn. Max. story def. ang. arc evaluated froa

str. def. ang. during aoderate eqs.

.

Maxiaua str. def. during aajor eqs. are evaluated at first, and then nonstructural coaponents are selected on the basis of deforaabi 1 1 ty. Nonstructural coaponents are classified Into three categories on the basis of deforaabi 1 1 ty.

Story deflection angles for aoderate earthquakes

: not aore than 1/200 rad.

: not aore than 1/120 rad. In cases there such deforaatlons

of structure of a specific

building is not likely to lead considerable daaage of any part of the said building.

Cross Section

Design of Henbers of Equlpnent /

Conponenls /BIdg

Allotable Unit Stress for Tent)orarlly Load The sane as left. For Moderate Eqs. : Allowable

Unit Stress for leap. Load

For Major Eqs. (Checking of Ultiaate Capacity basing on

the ductility of the str.)

Table 3.2 Exaaple of Allotable Stress for Steel" \^ Allotable

^vjinlt stress

Kind of^\

steel \^

Allotable unit stress for sustained

loads (unit: kg/sq ca)

Allotable unit stress for teaporary loads (unit: kg/sq ca)

Coapresslon

Tension Bending Shear Coapres- Tension Bending Shear

slon

General structural

steel

Ifeldlng structural

steel

F F F F

1. 5 tiaes the values of allotable unit stress for coapresslon, tension, bending or shear for sustained loads, respectively.

1.5 1.5 1.5 1.5 [3

Bolts

Black

F

1.5

Finished

F f

1.5 2

Rivet steel

F F

1.5 2

Cast steel

F F F F

I.S 1. 5 1. 5 1.5 JT

Cast iron

F

I.S

In this table. F shall represent a value of standard strength (unit: kg/sq ca) to bo specified

by the Minister of Construction according to the kinds and quality of steel, etc.

412
Theme VI

International Decade for Natural

Disaster Reduction
I
Damages and Lessons From the Loma Prieta Earthquake

Stephen A. Mahin*

ABSTRACT

The Loma Prieta earthquake provides a wealth of

information on the seismic response of a wide

variety of structures located in a large metropolitan

area. Soil amplification of motions at sites distant

from the epicenter contributed significantly to the

damage developed during the earthquake. Because

of the large shaken area, the earthquake provides

much useful information on earthquake resistant design and preparedness planning. Structural

damages resulting from the earthquake are reviewed herein with emphasis on buildings and

bridges. Implications for modern design and

retrofit methods are highlighted. Emphasis is placed on the need to carefully consider soil conditions, to treat a structure as a system rather

than as an assemblage of independent elements, to

explicitly define performance expectations, and to increase efforts to retrofit older seismically hazard- ous structures. KEYWORDS: buildings; bridges; damage;

earthquakes; retrofit; seismic resistant design.

1. INTRODUCTION

The Loma Prieta earthquake of October 17, 1989

has proven to be one of the most costly natural

disasters in American history. Probably the most

vivid examples of structural damages from the

period immediately following the earthquake were

the catastrophic collapse of the Cypress Street viaduct on Interstate 880 in Oakland, the partial

collapse of a section of the San Francisco - Oakland Bay Bridge, and the distress to numerous

buildings in San Francisco's Marina district. As

additional data was gathered during the days and

weeks following the earthquake, it became

abundantly clear that damages were far more

extensive than originally suspected.

In total more than 18,000 dwellings and several thousand other structures were significantly damaged by the earthquake. Nearly a thousand of

these were destroyed by the earthquake itself and

500 more have been demolished since the earthquake. Numerous buildings and bridges stand

empty today as studies continue to determine the

technical and economic feasibility of restoring their structural integrity.

Particularly severe damages were observed the

epicentral region. The older downtown regions of

Santa Cruz, Watsonville and Los Gatos (Fig. 1) were hard hit as were individual homes in the

Santa Cruz mountain. Today, several square

blocks in Watsonville and Santa Cruz have been

demolished and stand vacant as silent reminders of

the devastation caused by this earthquake. Highway and building structures were also damaged in localized regions throughout the greater San

Francisco Bay Area. Significant damages were also observed in lifeline facilities (water pipelines and

treatment plants, in particular), telecommunication

facilities, and in the architectural and mechanical

components and contents of buildings.

The total cost of the physical damages to structures is estimated by the Federal Emergency

Management Agency (FEMA) to be in excess of

$ 6.7 billion. However, the total cost of the

earthquake, considering the loss in immediate and

long term business revenue, damage to contents and inventories, medical and workman's compensation claims, and reallocation of resources, is likely several times this amount.

The human toll of the earthquake was also substantial. Sixty seven fatalities resulted from the

earthquake and more than 3,700 injuries were

reported. While these numbers are large, they are no where as large as might have been expected

considering severity of structural damage. The

low number of casualties is attributable in part to the fact that on the day of the earthquake many

people had gone home early to watch the baseball

World Series.

* Professor of Civil Engineering, University of

California, Berkeley, CA 94720

415
For example, based on the number of vehicles

expected on the Cypress Viaduct on a normal

workday, estimates of fatalities made just after the

earthquake exceeded 700, substantially greater than

the 42 deaths that actually occurred. This is indeed fortunate, but it is clear that had the

earthquake occurred under more normal circumstances, casualties in this and other structures would likely have been far greater.

In addition, the number of damaged and destroyed

dwellings resulted in more than 10,000 displaced

persons. Finding adequate food and shelter for these individuals contributed significantly to the

recovery efforts. Fortuitously, the weather was good and, with a few important exceptions, the

transportation infrastructure remained intact permitting relief supplies to be delivered to the

needy.

The Loma Prieta earthquake is the first major

seismic event to strike a major metropolitan area

in the U.S. in nearly 20 years. As such, it provides

the earthquake engineering profession with a unique opportunity to assess structural design and

retrofit methods. As the earthquake affected an

area where even larger earthquakes are expected, an examination of the damages will provide particularly valuable indications of future damage

trends. Because the levels of motion was not

unusually severe, and because many different

common structural types and soil conditions were

excited, the earthquake also provides planners and

engineers in other parts of the world with many

valuable lessons.

In this paper, the overall nature of structural damages observed during the Loma Prieta

earthquake are reviewed. After making some

overall comments on the apparent severity and

distribution of the damages, information regarding

the specific types of structural damages observed is presented. Due to space limitations, emphasis is placed on building structures and bridges. Finally,

the implications of these damages to the design of

structures in seismically hazardous areas are

offered.

Additional detailed information on the damages

can be found in the References.

2. SEVERITY AND DISTRIBUTION OF

DAMAGES

Ground motions were recorded at 131 sites and in 46 buildings during the earthquake. Correlation of

this information with observed damage will provide

a focus for research for many years to come. However, several important observations regarding

the distribution and severity of structural damage

can already be made.

First, ground motions recorded in the epicentral

region were quite severe, ranging up to 64% of

gravity. Motions greater that 30 to 40 percent of

gravity were detected over a very wide area. Since

these levels of motion approach those considered

in the design of new structures, one might expect

to see in these areas some significant damage to engineered structures, especially to older ones

designed to lower force levels. However, damage

in these areas with few exceptions concentrated in older unreinforced masonry and wooden structures.

Older types of engineered structures with known

seismic vulnerabilities (such as non-ductile

reinforced concrete buildings, tilt-up structures,

precast buildings, etc.) generally survived the

motions in the epicentral region without serious

damage. While the precise reasons for this unexpectedly good behavior are under study by

many at the moment, this anomaly reconfirms the

limitations of peak ground acceleration as sole

reliable index of earthquake damage potential. In

this case, the recorded motions in the epicentral

region ranged from 6 to 15 seconds, depending on

the record and the method used to determine

duration. This duration is short for this magnitude

event. In addition, surface displacements induced

by the shaking were also relatively small in comparison with those developed during other more typical and damaging earthquakes of this magnitude.

Comparison of the motion and damage data

obtained for this and other earthquakes will provide very important insight into the factors that influence damage in structures. Until the out- comes of such studies are known, it is prudent to acknowledge that the Loma Prieta earthquake did not generate particularly damaging motions in the

epicentral region and that it may not be conservative to extrapolate damages occurring in this region

to other parts of the Bay Area or the world where

earthquakes with greater damage potential might

occur.

416
Damage was quite serious to a wide variety of

structural types in some areas outside the epi- central region. In particular, regions of San

Francisco, Oakland and other areas as far as 80 to 100 km. from the epicenter suffered significant

damages. Peak ground accelerations recorded on

firm soil at these distances were generally around

10% of gravity or less. However, most of the

structures damaged in these areas were concentrated around the San Francisco Bay on man made

land overlying bay mud or on soft, deep soil deposits. In these areas motions were recorded

with peak accelerations ranging between 20 and

30% of gravity. Nonetheless, duration of shaking was relatively short, generally less than 6 sec, at these sites.

Attenuation of peak ground acceleration at distance was substantially less severe than might be

expected on the basis previous west coast earthquakes. Whether this discrepancy relates to special features of the fault mechanism and resulting

directivity for the Loma Prieta earthquake or

unusual, high amplification of the earthquake waves by the soft bay mud requires additional

study. Regardless, the large inventories of older,

seismically vulnerable, structures located over

these soils resulted in substantial numbers of

damaged structures. Soil effects clearly had a dominant influence on the unusual severity and

distribution of damages throughout the Bay Area.

Nonetheless, serious damage still concentrated in only the most vulnerable of structures since motions still were generally significantly below current design levels. On the other hand, the damage

observed possibly provides a reasonable indication

of the damages that might develop over a wider

area if a large local earthquake occurred in the Bay

Area. In this case, peak accelerations would

approach those that developed only on soft soil during the Loma Prieta earthquake. These damages also may provide a useful indication of the nature of damages that might occur in other areas of the world where similar soft soil conditions

exist. The 1989 Loma Prieta and 1985 Mexico

earthquakes clearly point out the potential vulnerability of urban areas overlying soft deep soil deposit, even for distant earthquakes. This should

have major implications for the Pacific Northwest,

the central mid-west and many locations on the

eastern seaboard of the U.S. and elsewhere in the

world where similar conditions exist.

Significant damages also occurred to new and older

buildings in the Palo Alto area. This region was

relatively close to the epicenter and had large

stocks of buildings. Ground motions ranging

between 30 and 40 percent of gravity were recorded. Several new concrete and steel buildings

suffered damaged as did older reinforced concrete,

wood and masonry structures.

Thus, it appears that structural damage is highly

dependent on the nature of the building inventories present in the shaken area, the types of soil conditions and the specific characteristics of the

motions generated by the fault rupture and that

arrive at the site. While efforts aimed at microzonation must continue, this sensitivity makes it from

a practical perspective all the more important to design structures in accordance with the basic

lessons learned from past earthquakes.

These past earthquakes have suggested the

desirability of designing simple structures that are

inherently insensitive to the uncertainties

associated the input motions and soil conditions.

Thus, emphasis should be placed on avoiding

systems with limited redundancy or that tend to concentrate damage in a few locations, on

providing details capable of large inelastic deformations, on selecting structural systems that are able to limit deformations to reasonable levels, and on attaching nonstructural components in such

a manner that they do not adversely influence

structural response and are not extensively

damaged by the structural deformations.

The damages produced by the Loma Prieta earthquake must be reviewed to determine the continued soundness of these past lessons. In the following sections, the performance of a variety of struc- tures will be reviewed with this in mind.

3. DAMAGE TO BRIDGE STRUCTURES

More than 1500 bridges exist in the area affected by the Loma Prieta earthquake. Eighty of these

suffered some relatively minor damage during the

earthquake, ten required shoring (though traffic continued to flow over them during repair) and

another ten were closed due to the severity of the

damage. Three bridges suffered collapse of one or more spans. These were the double deck Cypress

Street Viaduct, the truss section of the San

Francisco - Oakland Bay Bridge and the Struve

Slough bridge west of Watsonville. The collapsed

span on the Bay Bridge was replaced within about

417
a month. The Cypress Street Viaduct has been

demolished, and six other double deck viaducts in and near San Francisco remain closed a year and

a half after the earthquake awaiting the results of

engineering studies to determine whether they can be economically retrofit. The Struve Slough bridge was removed and rebuilt within a few months using

modern design practices.

Virtually all of the bridges in the area had

undergone the Phase I retrofit program, initiated by the California Department of Transportation

(Caltrans) following the 1971 San Fernando earthquake. These retrofits consisted of installing cable

restrainers at bridge expansion joints which are intended to prevent bridge deck members from

sliding from their seats. These retrofits proved to be generally quite effective during the earthquake.

In addition, at the time of the earthquake several bridges in the Bay Area, including the double deck

Interstate 480 Embarcadero viaduct, were under

design review for Phase 11 retrofits. The focus of

this new retrofit effort was on tall single column

bents that failed catastrophically during the 1971

San Fernando event.

4. DAMAGE TO STEEL BRIDGES

A number of long span steel bridges are used in the Bay Area to cross the bay and to span over the

Sacramento River. Damage was reported to three

of these: the San Francisco - Oakland Bay Bridge on Interstate 80, the San Mateo - Hayward Bridge on Highway 92 and the Carquinez Bridge where

Interstate 80 crosses the Sacramento River at Carquinez Straits.

The portion of the San Francisco - Oakland Bay

Bridge east of Yerba Buena Island consists of a double level freeway supported on truss spans. A

large steel tower (numbered E9) is provided near

the center of this segment of the bridge. This

tower resists longitudinal loads from a single

506-ft. span to the west and two 290-ft. spans to the east. The two concrete roadways extend across the tower on sets of 50-ft. long, simply supported

steel beams running parallel to the bridge's

longitudinal axis. Each of these beams was bolted

to a seat angle on the east, but the seat was

allowed to slide freely at the west end. During the

earthquake the beams slid from the free end with

the west end of the upper deck resting on the

lower one (Figs. 3 and 4). The lower deck was

also unseated and came to rest several feet lower

on an electrical transformer housing and the tower

braces.

In addition, the bottom chords of the trusses were

attached to the tower by means of twenty 1-in. diameter bolts. During the earthquake the bolts in the connection on the east side of the tower

sheared completely and the bearing plate moved

approximately 5-10 inches to the east. This was

the proximate cause of the roadway beams slipping

from their seats.

Additional damage to the eastern portions of the

bridge were also identified. These damages ranged

from spalling of concrete in support piers to

shifting of bearing plates and breakage of anchor

bolts.

The bridge was repaired by jacking the eastern segments back into place and providing new roadway beams and supports. The support seats were

constructed about 50% longer than in he original

design. A more extensive study is currently under

way by Prof. A. Astaneh and his colleagues at the

University of California, Berkeley and by Caltrans

to assess more reliable long term retrofits.

The San Mateo - Hayward Bridge is a steel orthotopic box girder bridge with a main span of 750-ft. During the earthquake one of the bearing

assembles on an approach pier shifted. The cap

screw used to retain the 5-1/2-inch diameter

bearing pin stripped off. The upper portion of the

bearing assembly shifted with respect to the lower

potion by about 2-3/4 inches. The bridge was

jacked back into position and repaired without

interruption of traffic.

The Carquinez Bridge consists of a pair of steel cantilever truss spans. Bearings supporting steel plate girder on the approach to the bridge re- portedly tipped over. These were repaired by

jacking the girders up and re-setting the girders. In addition, the piers supporting the truss spans

shifted in some locations as much at 2-inches.

The closure of the San Francisco - Oakland Bay

Bridge and the damage to the other bridges cross- ing the bay clearly indicated how vital these

vehicular arteries were to transport within the Bay

Area. The Governor of the State of California has

since signed a proclamation mandating that such

critical bridge structures be designed to remain

operable following a major earthquake. This is a

significant departure from the emphasis usually

418
placed on life safety in the design of such

struaures.

5. DAMAGE TO REINFORCED CONCRETE

BRIDGES

The most spectacular damage to reinforced concrete bridges was the catastrophic collapse of

nearly a mile long portion of the upper deck of the

Cypress Street viaduct located in west Oakland on

poor soil reclaimed from the bay (Fig. 5). This

structure was designed in the early 1950's and

construction was completed in 1957. The death

toll from this collapse was 42, though as mentioned earlier a thousand casualties or more

might have been expected had the earthquake

occurred during congested rush hour traffic. A variety of factors contributed to the failure of

this structure. These include the amplification of

the ground motions due to the soft soil conditions

in the vicinity, traveling wave effects, and variable

soil conditions. The structure was also built at a time when earthquake resistant design procedures were not highly developed. Thus, while the bridge

appears to have been conservatively designed

relative to the structural requirements in force at the time of its construction, it has a number of

major deficiencies with respect to current construction practices.

This bridge represented in many respects a substantial advance in the state of the art at the time

it was built. Not only did it have two levels, it was one of the first major bridge structures in the U.S.

to employ post-tensioning and large diameter

reinforcing bars. In most cases the structural system incorporated a significant number of

flexural hinges in order simplify the future

addition of access ramps, to reduce secondary

forces resulting from deformations associated with

post-tensioning and foundation settlement, to reduce computational effort and to control

moments that could be transferred to the

foundations. This resulted in a structure that was

nearly statically determinant. In addition, detailing

did not incorporate features needed to impart

ductility to the elements (shear reinforcement

sufficient to develop the flexural capacity of

members, confinement of potential plastic hinge

regions and inadequate development length on

reinforcement), and joints were neither confined nor designed to resist the shear forces associated

with realistic lateral loadings. Details of a typical transverse bent are shown in (Fig. 6). In addition.

in the longitudinal direction, frame action

depended on the bending forces developed in the

decks being transferred to the columns by torsional forces developed by shear friction. A relatively

massive and brittle structure resulted.

Detailed studies by Nims et al (1989) and others

have indicated that any of a number of structural

deficiencies could have triggered the collapse. The

lack of redundancy (needed to redistribute forces

in the event of the failure of an element) and the

brittle details utilized made the final catastrophic

outcome a certainty, regardless of the aaual trigger

of the failure and details of the seismic input.

The dominant features of the typical collapse

mechanism are shown in (Fig. 7). Shear cracks in the pedestal supporting the hinged upper level columns were able to extend downward into the

joint region following the projection of downward

bending hooks on the end of the transverse girder's

(cap's) top level reinforcement. The lack of shear

reinforcement in the joint allowed this inclined

crack to act like a slide, with the resultant pancake

collapse of the upper level.

The sudden and catastrophic nature of this failure mechanism was demonstrated convincingly during

the demolition of the viaduct. A wrecking ball was

used to initiate failure in a single column. Almost

immediately progressive collapse occurred

throughout the upper level not only extending

transversely across the roadway, but also longitudinally for six bays to the end of the

remaining standing portion of the viaduct (Fig. 8). Clearly, such brittle structures are not compatible

with earthquake resistant design.

Upon construction of the Cypress Street viaduct a

series of six other double deck viaducts were

designed and constructed in the San Francisco Bay

Area. All of these suffered damages during the

earthquake. Each suffered characteristic distress in the lower joint regions similar to that developed in the Cypress Street structure. An example of this

is the Southern Freeway portion of Interstate 280

located near the southern border of San Francisco

(Fig. 9). A bent in a curved connector structure between

Interstates 880 and 980 near the Cypress viaduct

also suffered damage. A knee joint extending

transversely from the side of the deck developed

severe shear cracks and a #18 bar fractured at the

hook (Fig. 10). The apparently inadequate shear

419
capacity of such joints is of concern, since the

structure was only about two years old at the time

of the earthquake. Similar damages were seen on

several older single level viaducts (e.g., where

Interstate 280 crosses Mission Creek). Improved

design criteria and retrofit procedures for such

knee joints are under investigation by Caltrans.

Two short bridges crossed over Struve Slough

near Watsonville. These consisted of multiple,

skewed T-beam spans supported on monolithic pile bents. During the earthquake the supporting piles failed in shear, resulting in the deck of one of the

bridges collapsing and the piles punching up

through the deck.

6. RETROFIT TESTS OF CYPRESS VIADUCT

The need for rapid retrofit of the damaged double

deck freeways in the San Francisco area led to a

series of field tests on undamaged portions of the

Cypress Street viaduct. These tests consisted of

forced vibration studies to determine the dynamic

characteristics of the structure and static lateral load tests in the transverse direction to determine

force and deformation capacities. Tests were

performed on a relatively unmodified segment of

the structure, as well as on the same segment after

it was retrofit. The three bents used in the tests are shown in (Fig. 11). The results of these tests have been reported by Moehle and Mahin (1990).

Tests on the original structure indicate that the

upper column pedestal and the adjacent lower

joint developed cracking characteristic of the

damage leading to the failure of the structure

(Fig. 7) at a displacement of about 3/4-in.

(191 mm)at the upper deck level. The total lateral load of 1400 kips (635t) applied to the three bents

corresponds to an equivalent base shear coefficient

of about 0.32 assuming a 2:1 distribution of loads

to the upper and lower deck levels. A simple

elastic time history analysis of the test structure has shown that it would have required a base shear

coefficient at least 61% larger to have remained

elastic for a ground motion recorded nearby. The

working stress base shear coefficient utilized in the

original design was 0.06.

Three different types of retrofits for the lower

joints were investigated. In addition, upper and

lower bent caps (girders) were externally posttensioned to help compensate for inadequate bar

development lengths, and to provide increased

strength and confinement in the joints. Most of

the columns were reinforced in shear by adding

exterior post-tensioning.

The retrofits were able to maintain the vertical load integrity of the test structure. However,

significant damage was observed. In particular, the

critical failure zone moved away from the lower

joint to the upper joints, which began to fail and

disintegrate during the tests. Diagonal cracks in the upper joints began at displacements at the

upper deck level of less than 1-in (254 mm). In

addition,the non-retrofit columns began to fail in shear at an upper deck level displacement of about

1.5-in. (381 mm) and an equivalent base shear

coefficient of about 0.61.

After strengthening the shear damaged columns,

the structure was again cycled and lateral dis- placements of 9.8-in. (0.25 m) were achieved at its top along with equivalent base shear coefficients of

0.91. These values are about 13 and 3 times

greater, respectively, than the values that could

have been developed by the original structure

(Fig. 12).

One of the bents had heavy steel wide flange

sections post-tensioned vertically along the face of

the columns like splints. During the tests composite action could not be maintained and the

relatively long lengths of unbounded post-tensioning steel used resulted in large cracks and

spalling in the lower joints and in gaps between

the steel and the concrete (loss of confinement)

over large portions of the columns and joints.

Another bent had inclined rock anchors grouted in the lower joints. This bonded reinforcement

proved effective in resisting perpendicularly

oriented shear cracks. However, towards the end

of the tests spalling and flexural cracking and joint

shear cracking had begun. The continued integrity

of this retrofit scheme under these conditions was

not fully addressed.

The third retrofit consisted of a heavy steel collar placed around the pedestal and the base of the

upper level column. This collar was post-tensioned

to the sides of the lower joint. During the test the

collar was observed to rotate about the lower joint

and, upon its removal following the tests, fully

developed sets of inclined cracks were observed in the joint along with much pulverized concrete.

420
These tests indicate the effectiveness of retrofitting

existing bridge structures. However, they point out

the need to consider a structure as a system and

not to simply fix the damaged or over-stressed

portions. TTie limitations of the tests must also be

recognized. Loading of the test structure was

halted when the test site had to be vacated.

However, the structure was not believed safe to continue testing at that time. The deformation

history imposed may not be conservative for the

most severe ground motions expected in the Bay

Area. In addition, no attempt was made to retrofit or test the structure in the longitudinal direction

and this mode of behavior must be carefully

considered.

7. DAMAGE TO WOODEN BUILDINGS

As indicated previously nearly 20,000 dwelling

units were damaged during the Loma Prieta earthquake and 10,000 individuals were forced to relo- cate as a result. A majority of these displaced

people lived in wood frame houses and apartments.

Many homes located in the Santa Cruz mountains,

in the immediate epicentral region, were damaged.

Damages ranged from fallen chimneys and porches

to partial or complete collapse. Many wood frame

buildings in the hardest hit areas were over- whelmed by the seismically induced inertia forces. Sadly, the majority of the damage occurred in older wood frame buildings where the structures were simply not connected to their foundations.

During the earthquake, as they have in innumerable past earthquakes, the structures simply shifted

off their foundation with resultant vertical dislocation (Fig. 13).

Another common type of failure that was observed

related wood frame dwellings built upon short

pony or cripple walls. These short walls provided

a two to four foot (0.6 to 1.2 m) tall access space

that was used for storage, ventilation and to accommodate sloped sites. In many older

buildings the framing between the ground floor and the foundation consisted of vertical studs and

horizontal siding. Even though diagonal braces were provided in the upper levels to resist lateral loads, braces were typically omitted from the pony

walls. In addition, the horizontal siding covering

these walls in most cases had badly deteriorated

and effectively provided little or no lateral resistance (Fig. 14). The failure of these short

support walls resulted in large vertical and lateral displacements of the supported structures (Fig. 15).

The unfortunate aspect of both of these forms of

damage is that simple and economical retrofit procedures, if implemented prior to the earthquake, could have prevented or significantly

reduced the severity of the observed damages.

Repair following the earthquake was generally

extremely expensive not only as a result of the

required structural repairs, but because of wide

spread damages to architectural, electrical and

mechanical features. Pre-earthquake retrofits to these walls and foundations were seen to be effective in several instances in Santa Cruz and else- where.

Another form of damage to wooden structures was

observed in the Marina District in San Francisco.

This area was constructed on loose, fine sandy

hydraulic fill. During the earthquake the area

exhibited evidence of significant liquefaction with

localized sand boils, lateral spreading, slumping

and heaving of sidewalks and roadways.

Many wood buildings were damaged or collapsed

in this area. However, a detailed inspection of the

damages indicates that the collapses were

concentrated in three or four story apartment

buildings (Fig. 16) located at the comers of blocks

(or where a structure was not sandwiched between

two other adjacent buildings with little or no separation). These older apartment buildings were

generally constructed with relatively massive

apartment floors supported over a level of garages

(Fig. 17). The bottom level thus was largely open

to allow for multiple garage door openings.

Typically, parking stalls were several cars deep and

often extended out though the back of the building

to provide for even more parking. This situation

resulted in a soft first story. Lateral load resis- tance was provided by horizontal wood sheathing

(often covered by brick veneer) placed over only a few walls. This sheathing was observed to have decayed severely in many cases during the life of the

structure.

The amplified seismic motions in the Marina

District resulted in the collapse of the lower level of these corner apartment buildings in many cases.

It is significant to note that these types of apartment complexes are quit common throughout San

Francisco. Some apparently identical buildings

located only a few blocks away from the heavily

damaged area, but on firmer ground, remained

virtually undamaged following the earthquake.

421
Nonetheless, the potential hazard posed by these

structures during future more severe earthquakes

should be carefully investigated.

Initial suppositions regarding to the causes of the

severe structural damage in the Marina District focused on the observed soil liquefaction.

However, structures did not collapse or suffer substantial damage even in the most heavily

afflicted area, if the lateral load resisting system was continuous over the height of the building.

Another remarkable feature was the limited

damage to residential homes located along the

middle of a block. These structures typically were

two or three stories tall, with adjacent structures having the same floor elevations and virtually no

separation. As noted in during other recent earthquakes, such structures seem (at least

initially) to buttress one another and suffer only

limited damage. More extensive damage occurred

in buildings located at the corners of blocks. Where liquefaction occurred near or under a

structure, vertical differential settlements of a few

inches were observed with resultant structural and

especially architectural damage.

Similar damages were also observed in other areas of San Francisco, notably in the Mission Creek

area located south of Market Street. The

distribution of this damage is similar to that following the great 1906 San Francisco earthquake.

8. DAMAGE TO UNREINFORCED

MASONRY BUILDINGS

The older downtown areas of many well

established cities in California were built of

unreinforced masonry. These commercial areas

suffered significant damages in Watsonville, Santa

Cruz, Oakland and San Francisco. However,

damage appears to be strongly influenced by local

soil conditions and the intensity of ground

motions. This can be seen in Table 1 where the

approximate numbers of unreinforced masonry

(URM) buildings vacated following the earthquake

are compared with the total stocks in various

cities. Cities are listed in order of decreasing

epicentral distance.

Table 1, Damage Statistics for URM Buildings*

City Approx. Number Number

of URM Buildings Vacated

Berkeley 400 5 San Francisco 2000 252

uaKiana zOUO 40U

South San

Francisco 42 2

San Mateo 28 1 Salinas 96 16

Hollister 17 14

Palo Alto 49 0

Mountain View 20 6

Gilroy 41 2

San Jose 230 3

Los Gatos 29 11 Santa Cruz 46 36

Adapted from Benuska (1990)

Only limited damage was detected in Gilroy, for instance, even though motions in excess of 50

percent of gravity were recorded in the vicinity of

several unreinforced masonry structures. In other

cities damage only concentrated in certain areas, with severe damage in some areas and virtually none in others. While soil conditions play an

important role in this distribution, the precise

interrelationships between the various factors influencing damage in this class of hazardous

structure need careful study.

The most prevalent form of damage to URM

structures was the collapse or dislocation of parapets. In most of these cases the masonry was tied to the floor diaphragm by means of steel anchors.

The short extension of the wall above the roof fell from many buildings, even where only moderate

ground motions occurred far from the epicentral

region.

A particularly devastating form of damage resulted

when the walls were not tied to the floor diaphragms. In many cases the walls failed outof-plane, falling into the street (Fig. 18). In several notable cases the walls fell from upper

floors of a building onto the roof of a lower

adjacent structure (Fig. 19). The damage to the

lower building resulted in two deaths in two

separate buildings in Santa Cruz. In another

instance, five casualties resulted when a wall

collapsed onto the sidewalk area in front of an

422
unreinforced masonry structure. For the level of

shaking encountered, moderate amounts of tie reinforcement appeared effective in limiting this type of damage.

In some cases in-plane shear failures of brick walls was observed (Fig. 20). In some cases this distress resulted in local collapse of wall panels. In other

buildings wide spread, inclined cracks developed,

stepping along weak mortar joints. In many cases buildings with moderate cracking have been left un-repaired with continued occupancy or the

cracks have been simply repaired with epoxy

injection to insure water tightness.

Brick and stone architectural veneers fell from

many buildings. Typical metal inserts embedded in the mortar joints proved insufficient when the

attachment nails pulled from the wood framing.

The framing under the veneer had deteriorated

badly in many of these cases. In other cases the veneers were dislodged as a result of hammering

against adjacent structures.

9. DAMAGE TO REINFORCED CONCRETE

STRUCTURES

While non-ductile concrete frames have performed

poorly in past earthquakes, few of these types of

buildings were located in areas subject to severe ground shakings during the Lxjma Prieta earthquake. In some cases severe shear cracks were

observed in columns. However, the buildings in question (for example, along Mission Street near Third Street in San Francisco) were located midblock with no space separating them from the

adjacent structures. It is expected that the

adjacent structures provided the required lateral resistance in these cases.

In Watsonville, where peak accelerations of 39% g were recorded, an old five story reinforced concrete frame structure appeared to suffer mainly

architectural damage. The building had a number

of broad shear walls which apparently accounted

for its good behavior in spite of the severe motions

recorded in the area.

Damage was observed in a number of newer rein- forced concrete structures. For example, a mechanical penthouse, containing a water tank, fell from the top of a reinforced concrete multistory

hotel in Burlingame. Across the street another

hotel, build less than 16 months prior to the

earthquake, suffered significant damages to its shear walls in the lower levels and to its dia- phragms near the shear walls. Both of these hotels were constructed on fill over bay mud. Motions at the nearby San Francisco International Airport were up to 33%g.

A ten story moment frame constructed in San Jose

of lightweight concrete suffered severe spalling

damages in several locations and architectural

damages in spite of ground motions in the area

less than 11 % g. lightweight concrete shear walls

in a 15 story telecommunications building (Fig. 21)

in Oakland suffered severe cracking, spalling and

splitting in the lower story. This building was

designed as a dual system with steel frames

proportioned to carry 25% of the lateral forces along with the gravity loads.

Damage also occurred in reinforced concrete

buildings used as hospitals and schools. The seven

story tower of the Peralta Hospital, built in Oakland in 1927, was severely damaged and has

been demolished following the earthquake. The

Palo Alto Veterans Administration Hospital suf- fered serious damage, including shear cracking to the columns in two of six buildings. The four

buildings without damage had been previously

retrofit. A number of reinforced concrete buildings on the Stanford University campus, including

the School of Business Administration, the Library

and some residence halls suffered moderate

cracking. The John O'Connell High School in San

Francisco suffered serious structural cracking to its frame-wall system.

10. DAMAGE TO STEEL BUILDINGS

Damage to steel buildings was typically less obvious than that to masonry or concrete structures. However, there were several reports of

buckled braces in several buildings along the San

Francisco Peninsula near San Mateo and Palo

Alto. In most cases these could be fixed within a few days. In other cases substantial effort was

needed to find damage in buildings where large

amounts of architectural distress suggested the

presence of structural damage. Evidence of panel zone buckling, gusset plate yielding and buckling,

and column buckling were found in several

buildings.

423
11. DAMAGE TO RETROFIT STRUCTURES

The engineering profession has known for many

years of the seismic vulnerability of many types of

structural systems and has attempted to retrofit many of these structures. The Loma Prieta

earthquake provides an opportunity to assess the

efficacy of the procedures used.

A good example of the effectiveness of retrofits is the performance of a four story reinforced concrete telecommunications building in Watsonville. This

building had been be upgraded by infilling windows

and adding shear walls. During the earthquake

accelerations were recorded up to 1.24g at the roof

without the building suffering any significant structural damage. Another example is the retrofit reinforced concrete structures at the Veterans

Administration Hospital in Palo Alto. As

mentioned previously, these suffered little damage

in comparison to the non-retrofit units.

On the other hand, there is ample evidence that some retrofit schemes did not work as well. This

is attributable to the lack of any performance

standards for retrofit work, any code to set load

and detailing requirements, and virtually no

research on the effectiveness of many types of

retrofit procedures used in practice.

One example of this situation is a 6 story rein- forced concrete building constructed in San

Francisco at 150 Fourth Street during the early 1980's (Fig. 22). During change of ownership a few years after it was built a number of seismic

deficiencies were identified. These were remedied on the basis of protection of the life safety of the

occupants by adding steel braces and shear walls. Following the earthquake, connections in the steel braces were observed to buckle (Fig. 23), the

attachments of the braces to the concrete frames

had slipped (in some cases due to faulty instal- lation of anchor bolts), extensive cracking occurred

in the shear walls, and initiation of punching shear

failures in a few column to flat plate connections.

The building has been since repaired (Fig. 24) to restore the capacity that existed prior to the

earthquake.

Another example, is the Hotel Oakland in Oakland. This is a steel frame building with extensive

masonry infill panels. In retrofitting this structure new lateral load resisting elements were added and

the exterior masonry elements were positively

attached to the building to prevent large panels

from falling into the street. As result of the

earthquake nearly all of the masonry piers in the

building developed distinctive x-shaped shear

cracks. In addition, at a few locations masonry

elements were dislodged from the building and fell to the street (Fig. 25). This building apparently

preserved life safety as intended, but extensive and

expensive repairs were required to restore the

building fully to service.

Most of the focus of retrofit work has logically

been on unreinforced masonry buildings. In a

recent study by Conrad (1990) 400 of San Francisco's more than 2000 URM buildings were

inspected and 69 were identified as having been

retrofit to some degree. This would indicate that only about 3% of the city's URM buildings have

been retrofit. Most of the retrofits had been

implemented through the introduction of steel braces. Retrofit structures tended to be found in clusters where a local community was being

redeveloped. The report found that 54% of the

retrofit buildings suffered no damage, but no

comparison with the performance with adjacent

non-retrofit structures was offered. Light damage

was observed in 22% of the retrofit structures and

moderate damage was seen in another 20%.

Heavy damage was seen in three buildings (3%).

One of these structures (259 Front Street) was

located at the corner of a block and it was heavily

braced along two street sides (Fig. 26). The added

braces at the street level buckled out of plane

significantly (Fig. 27), but no damage was actually

found in the masonry portions of this four story

building. The second structure (1051-1075 Battery

Street) was also braced but the mortar in the

existing masonry walls was apparently so poor that random sections of the exterior wall fell from the

building. Another building (located at the corner

of Sixth and Bluxome Streets) had apparently been

partially retrofit by addition of wall anchors. Five

people were killed as the upper level wall on one

side of the structure fell onto the sidewalk and

street. This damage may have been initiated by

pounding with an adjacent building. In all of these

cases it is clear that the lack of a consistent,

systems approach to the design of the retrofits had

a detrimental effect and that a new class of

potentially hazardous buildings (i.e., the

inadequately retrofit structure) needs to be

investigated.

424
12. DAMAGE DUE TO HAMMERING OF

ADJACENT BUILDINGS

Another important observation from the Loma

Prieta earthquake is that many buildings suffered

damage due to pounding. The separation of many

buildings was insufficient to avoid collision of the

upper stories. A wide range of damage resulted,

from local spalling of veneers near the impact

point through cracking of vertical supporting

members to failure (though not collapse) of

columns. Field inspections have identified several hundred cases where pounding contributed to the

observed damage. In many instances it was

apparent that a few more cycles of impact could

have caused the failure of the structures involved

(Fig. 28). In many of the severely damaged

unreinforced masonry structures evidence suggests

that pounding was a contributor to the damage.

In one instrumented building in Oakland (at the

intersection of 17th and Harrison Streets) the

records show clear evidence of pounding at one

corner of the building. The steel moment frame

building (with reinforced masonry walls along the

property lines) was separated from the adjacent

unreinforced masonry storefront by about an inch

(25 mm). However, no significant damage was observed in either building as a result of the

pounding.

Damage due during the 1985 Mexico and other

earthquakes suggests that pounding is a serious

problem that needs to be addressed by the engineering profession. A number of investigators are

currently studying this problem as it relates to the

Loma Prieta earthquake.

13. LOSS OF FUNCTIONALITY

An important observation regarding damage to

structures during the Loma Prieta earthquake has

been the public's reaction. Most structures performed generally in conformance with the

design professions expectations. New structures and most engineered structures did not collapse,

and life safety was protected, even in regions of

severe shaking. However, these motions were not

the most severe or damaging that might be

expected in the Bay Area. Nonetheless, the public

reaction to these damages was generally that they were excessive.

There was much nonstructural damage to partitions, ceilings, cladding and contents (Fig. 29) in

structures with little or no structural damage. This

damage not only contributed a significant life safety threat in many cases, it disrupted the

functionality of the structures and added

substantially to repair costs. One ten story

building in Oakland reportedly had $6 million in water damages due to the breakage of a fire sprinkler line.

Structural damages were repaired quickly in many

cases, but in other cases considerable delays were

encountered as owners await financing or as technical difficulties are encountered. In many

cases the resulting repairs were cosmetic or

intended to simply restore a structure to its pre-quake condition.

Large institutional owners or ones with significant

capital investments have increasingly developed

heightened awareness of the need to control

damage. In addition to consideration of life safety for a structure's occupants, they have become

concerned over institutional life safety. For

example, Stanford University has developed much

more stringent criteria since the earthquake for

retrofitting their seismically hazardous buildings.

Having had to relocate several departments and

instructional and research facilities following the

earthquake, they now stipulate that buildings

should be designed or retrofit to permit repairs to be made within a few days or weeks of the

earthquake (depending on the occupancy of the

building). Other institutions that cannot similarly

just pick up and move have come to similar

conclusions. Similarly, the Governor's

proclamation also indicates that California state buildings should be built or retrofit so that they

would remain operable following a major

earthquake. This is a major departure relative to the conventional design of buildings.

14. IMPLICATIONS FOR EARTHQUAKE

RESISTANT DESIGN

Few new lessons have been developed as a result of the Loma Prieta earthquake. However, as the

first major urban earthquake in the U.S. in nearly

two decades, the earthquake stands as a compelling

reminder of many important past lessons. We

have again seen the vulnerability of several major

classes of buildings designed in earlier eras when

our knowledge of earthquake resistance was not

fully developed. The damages to numerous

unreinforced concrete buildings, wood dwellings

with inadequate foundation attachments, soft story

425
structures (e.g., Marina District apartment houses)

and non-ductile concrete frames (e.g., double deck

viaducts) are reminders of well-known past lessons.

Similarly, the important effects of soil liquefaction

and site amplification have been seen in many past

earthquakes, such as the 1985 Mexico, 1967

Venezuela and 1964 Nigatta, Japan earthquakes.

The Loma Prieta earthquake indicates that these

soft soil sites should be carefully considered in design as they will be sensitive not only to close

earthquakes, but also to any number of

earthquakes generated on other relatively distant

faults.

The special need to have a safe and dependable

transportation system has been again demonstrated. It has been seen that certain structures may be more critical than others, and greater levels of

conservatism should be considered to assure their

operability even following a major earthquake.

At the same time some past phenomena were not

seen. However, the Loma Prieta earthquake

appears to be an unusual earthquake in many

respects and its motion in the epicentral region

appears not to be as damaging as other typical

earthquakes in the western U.S. Thus, the good

behavior of some types of structures during this earthquake should not lead to complacency in implementing their retrofit.

The damages provide a clear warning to the Bay

Area and to other seismically active areas of the

world to increase preparedness activities. In

particular, the substantial hazard posed by our

older seismically vulnerable structures must be

remedied. This can only be done through reliable evaluation and retrofit criteria developed and

validated on the basis of laboratory research and

quantitative investigations of the seismic

performance of buildings during earthquakes.

The earthquake has shown a greater value being

placed by the public, government and corporate

entities on reducing economic as well as life hazards. Control of damage during earthquakes

need not involve great expense, but additional

attention to the selection of the structural system

and to detailing of nonstructural components.

Additional research is needed to provide a reliable design basis for such considerations.

Finally, it must be recalled that the timing and

location of the Loma Prieta earthquake were

fortuitous. The epicenter was in a sparsely

populated region relatively far from urban centers.

It also occurred at a time when many people were away from the most vulnerable structures. This set of favorable circumstances can not be expected to occur for all future earthquakes.

15. ACKNOWLEDGMENTS

Many of the research observations made in this paper were made possible by the generous research

support provided by the National Science Foundation and the California Department of Transportation. In addition, the author would like to thank

his many colleagues, especially Jack Moehle, Hasan

Astaneh and Roy Stephen, for sharing their observations with him. The findings and conclusions of this paper, however, must be blamed

solely on the author alone.

16. REFERENCES

Astaneh, H. et al (1989), "Preliminary Report on

the Seismological and Engineering Aspects of the

October 17, 1989 Santa Cruz Loma Prieta) Earthquake," Report No. UCB/EERC-89/14, Earthquake

Engineering Research Center, Univ. of California,

Berkeley, October.

Benuska, Lee, Editor (1990), "Loma Prieta Earthquake Reconnaissance Report," Earthquake

Spectra, EERI, Supplement to Vol. 6, May.

Conrad, J. (1990), "Performance Evaluation of

Strengthened Unreinforced Masonry Structures Following the October 17, 198 Loma Prieta

Earthquake," Graduate Student Report, Dept. of

Civil Engineering, Univ. of California, Berkeley,

May.

Dames and Moore (1990), A Special Report on the

Oct 17, 1989 Loma Prieta Earthquake, Los Angeles, CA, January.

EQE Engineering (1989), Oct. 17, 1989 Loma

Prieta Earthquake: A Quick Look Report, San

Francisco, CA, November.

426
The Governor's Board of Inquiry on the 1989 loma

Prieta Earthquake (1990), "Competing Against

Time," Report to Governor George Deukmejian,

Housner, G.W., Board Chairman, Thiel, C.C.,

Editor, May.

Nims, D.K. et al (1989), " Collapse of the Cypress

Street Viaduct as a Result of the Loma Prieta

Earthquake," Report No. UCB/EERC- 89/16, Earthquake Engineering Research Center, University of

California, Berkeley, November.

Seed, R.B. et al (1990), "Preliminary report on the

Geotechnical Aspects of the October 17, 1989

Loma Prieta Earthquake," Report No.

UCB/EERC90/05, Earthquake Engineering

Research Center, Univ. of Calif., Berkeley, April.
428
429
Fig. 7 Postulated Failure Mode Fig. 8 Failure Mode Observed during Demolition

Fig. 9 Damage to Southern Freeway Fig. 10 Damaged Knee Joint in New Freeway

Fig. 11 Cypress Street Test Structure

430
(30 •H

Fx4

431
Fig. 16 Collapse of Apartment in Marina District Fig. 17 Typical Marina District Apartment

Fig. 18 Out-of-Plane Failure

Of Masonry Wall

.ig. 19 Collapse of Masonry Wall

Onto Adjacent Roof

Fig, 20 In-Plane Failure of Masonry Wall

432
Fig. 22 Retrofit of New Reinforced Concrete

Building

Fig. 21 Lightweight Concrete Shear

Walls Failed At Base of Building
i

[I

Fig. 25 Damage to Retrofit Steel Frame Building

with Masonry Infilled Walls

Fig. 26 Retrofit URM Building with Steel Braces

Fig. 27 Buckled Brace in URM Building Fig. 28 Pounding Damage

Fig. 29 Damage to Building Contents

434
Damages and Lessons Learned From Hurricane Hugo

by

Peter R. Sparks*

ABSTRACT _ Wind and surge conditions in

Hurricane Hugo are reviewed and

their relationship to the extent and

nature of damage is discussed. It is

shown that in areas where the gust

wind speed exceeded 50 m/s,

10%-20% of the buildings were

destroyed or seriously damaged. At

40 m/s the figure was about 6%

and at 30 m/s only 1% were

seriously damaged. Surge damage,

although locally severe, only

accounted for about 10% of the

overall damage.

In general damage was related to

the local conditions not to the

relative severity. Thus the wind

damage in Charleston, where the

wind conditions had a recurrence

interval of 50 years, was similar to

that in Sumter where the same

wind speed had a recurrence

interval of more than 200 years. It is suggested that this was due to

the fact that more than 95% of

structural systems and 99% of

building enclosures had been

selected without direct reference

to design wind conditions which

had been based on appropriate

recurrence intervals. For recently

constructed buildings design flood

levels did have a bearing on the

structural form.

Certain types of buildings and

components performed very

poorly, primarily mobile homes,

unreinforced masonry buildings,

older single family dwellings close

to the ocean and some newer ones

with masonry foundations, most

roofing materials and some wall

cladding components.

The severe disruption of the

electricity supply system,

beginning when the wind reached

about 30 m/s, and its effect on the

community is discussed and

contrasted with the much better

performance of the telephone

system.

It is concluded that the extensive

damage resulted in general not

from unprecidented wind and

surge conditions, but from the

failure to use available knowledge

on wind and surge effects in the

selection of building and utility

systems.

KEYWORDS: buildings; damage;

Hurricane Hugo; storm tides;

tropical cyclones; utilities.

1. INTRODUCTION

By the time Hurricane Hugo

crossed the coast of South Carolina

just after midnight on September

22, 1989, it had already caused

over $2 billion worth of damage in

the Caribbean. In its passage

through the states of North and

South Carolina it caused further

damage estimated to be nearly $7

billion, making it the most

expensive storm in the United

States history.

* Department of Civil Engineering,

Clemson University, Clemson, SC

29634-0911

435
At landfall Hugo had a central

pressure of 934mb and an eye

about 50 km in diameter. Winds

circulated in a counterclockwise

direction about the eye, but the

high forward speed produced a

highly asymmetric wind field,

increasing speeds to the right of

the track and reducing them to the

left. The highest winds and worst

storm surge occurred where the

northeastern part of the eye-wall

crossed the coast about 35 km

northeast of the city of Charleston

at Bulls Bay. The high forward

speed (13 m/s-18 m/s) and a track

almost normal to the coast resulted

in very unusual wind conditions in

the interior of South Carolina and

the western part of North Carolina.

2. WIND CONDITIONS

A number of anemometers

measured wind conditions during

the passage of Hugo. Unfortunately

only one of these did so under

standard meteorological conditions

and no anemometers survived in

the area of highest winds.

Figure 1 shows the probable

maximum gust wind speeds at a

height of 10 m in the most exposed

locations, based on corrected

observations in Charleston, Myrtle

Beach, Beaufort, Sumter, Camden,

Florence, Charlotte, and Hickory

plus the use of a theoretical wind

field model.

3. STORM TIDES

Continuous records of water level

were obtained near the Customs

House in Charleston and at Winyah

Bay near Georgetown. Additional

still water levels were obtained by

surveys of tide marks along the

coast.

In the Charleston area tides ranged

from 3 m above mean sea level in

Charleston Harbor to 4.5 m at the

north end of the Isle of Palms. At

the south end of Bulls Bay the tide

reached 6 m and at the north end

of the bay, in the fishing village of

McLellanville, it was 4.2 m to

4.8 m. In the resort areas along

the northern coast, including

Myrtle Beach, tides were 3 m to

4 m above mean sea level. The

normal tidal range is less than 2 m

and much of the occupied land is

only 2.5 m to 4 m above mean sea

level.

In the areas that experienced the

eye of the storm, Charleston, Folly

Beach, Sullivans Island and the Isle

of Palms, the offshore winds before

the eye held the surge back until

the winds dropped as the eye

passed. The water level then rose

rapidly, remained up while the

trailing eyewall passed and then

dropped as the winds subsided.

To the northeast of the eye the

surge rose more steadily, reaching

its peak at about the time of the

highest winds. The surge began to

drop as the winds diminished and

moved around to the south.

To the southwest of the eye the

offshore winds suppressed the

surge and little flooding was

experienced.

4. SEVERITY OF THE STORM

Based on its central pressure, Hugo

was the 10th most intense storm to

strike the United States this

century, being about the same

strength as the 1926 storm

436
in Miami and Hurricane Hazel

which struck the Carolinas in 1954.

The hurricane produced the

highest storm tide recorded on the

Atlantic coast this century, but a

similar tide was recorded at

Savannah Beach near the South

Carolina-Georgia border in 1893

and higher tides might have been

experienced in the Charleston area

in the storm of 1752. However, in

the southern part of Bulls Bay the

tide was a very rare event which

would be expected to be exceeded

on average only once in over 500

years. In McLellanville the mean

recurrence interval was about 300

years and at the northern end of

the Isle of Palms about 200 years.

On SuUivans Island, in Charleston,

and along most of the northern

coast the mean recurrence interval

was approximately 100 years and

about 80 years at Folly Beach.

In the Charleston area the winds

had a mean recurrence interval of

about 50 years, but similar

conditions had not been

experienced since 1911. They had

however also been experienced in

1893, since records began in 1879.

Based on the description of

damage, the wind conditions in

Hugo were almost certainly

exceeded in 1752 and possibly in a

number of other storms in the 300

year history of the city.

In the Bulls Bay area the wind

conditions probably had a mean

recurrence interval of about 100

years, possibly as high as 200 years.

It was about 60 years in

McLellanville and less than 20

years in the Myrtle Beach area. The combination of an intense

storm and high forward motion

produced rare wind conditions in

the interior of South Carolina and

in the Charlotte area of North

Carolina. At Sumter, 150 km from

the coast, the mean recurrence

interval was probably between 200

and 300 years. The very gusty wind

conditions in Charlotte, 300 km

from the point of landfall, may have

had a recurrence interval more

than 100 years.

5. BUILDING PERFORMANCE

Since the early 1960s engineers

have recommended that buildings

and other structures be designed

for conditions with a mean

recurrence interval of 50 to 100

years. Well designed structures

usually incorporate factors of safety,

so serious structural damage should

not be expected in conditions with

a mean recurrence interval less

than 500-1000 years. Such

conditions existed only for surge

action in Bulls Bay and probably did

not exist anywhere for wind action.

Nevertheless the damage was

widespread. The extent of damage

appeared to be related to the local

conditions, not the relative severity

of those conditions, nor was

damage confined to buildings

constructed before appropriate

design criteria were available. In

retrospect this was not surprising.

Most of the damage occurred in

South Carolina where the building

stock consists of approximately

75% wood-framed single-family

dwellings, 15% mobile homes, and

about 10% other types of buildings.

These other types, however

comprise about 35% of the value of

construction.

437
Single-family dwellings are usually

constructed on traditional lines,

sometimes influenced by

prescriptive building code

requirements. These requirements

generally do not adequately reflect

the need to resist wind and water

effects.

Since 1976 mobile homes have

been constructed to Federal

Government standards. The wind

loading requirements in these are

much lower than those generally

set for conventional construction

and reflect a very poor

understanding of wind effects on

buildings.

In South Carolina buildings over

450 m2 or three or more stories in

height must be designed by a

registered architect or engineer.

Even these buildings often contain

components selected on the basis

of empirical design procedures and

product data which do not

adequately consider the effects of

wind (or surge).

Under these circumstances it is

unlikely that, of all of the buildings

affected by Hurricane Hugo more

than 5% had structural systems

designed to resist appropriate wind

and surge loads. Probably less than

1% of the buildings had roofing

and cladding systems specifically

designed for appropriate wind

conditions. Design wind speeds

based on recurrence intervals

therefore had very little bearing on

the wind resistance of most

buildings.

Design flood levels had more of a

bearing on coastal construction

because the availability of insurance

through the government-run

National Flood Insurance Program

was tied to satisfying floor elevation

requirements based on the 100

year flood level. Unfortunately the

flood maps were not very accurate

and buildings were often

constructed without consideration

of the forces generated by the

storm surge and the associated

wind loads.

Table 1 gives the insured losses for

buildings and their contents. The

number of insurance claims for

each type of building is in almost

exactly the same proportion as its

relative contribution to the building

stock and the amount of damage is

in the same proportion as its

relative value. Although the flood

damage was widely reported and

many buildings were destroyed,

flood losses were only about 10% of

the total property losses and

accounted for only 4% of the

insurance claims.

Table 2 gives the Red Cross

estimate of the number of housing

units damaged. This was made

shortly after the storm,

subsequently insurance claims were

made on nearly 3 times as many

properties as were initially

observed to be damaged.

Ultimately claims were made on

30% of buildings in South Carolina

but only 3% were seriously

damaged or destroyed. The vast

majority of damage was due to

relatively minor wind damage

followed by extensive rain damage.

The distribution of damage was

closely related to gust wind speeds,

the proportion of mobile homes,

and, for surge damage, proximity to

the ocean. In some rural areas

where 20%~30% of the population

438
lives in mobile homes, 70%-80% of

the buildings completely destroyed

were of that type.

On the islands near Charleston and

on the shores of Bulls Bay nearly all

buildings experienced some wind

or surge damage and in some

coastal communities nearly 50% of

the ocean-front buildings were

completely destroyed.

In counties where the wind gusted

to more than 50 m/s at least 80%

of the buildings experienced some

damage and 10%-20% were

destroyed or experienced extensive

damage. Damage rates diminished

as the wind speed dropped. At

40 m/s about 6% of buildings were

destroyed or seriously damaged. At

30 m/s less than 1% received

significant damage.

5.1 Detailed Survey of Wind Damage

Wind damage to buildings and

other structures was observed

along the South Carolina coast from

Edisto Beach 55 km southwest of

Charleston, to the North Carolina

border and inland to beyond

Charlotte. Damage to roof

coverings, wall cladding, and signs

occurred where the local gust wind

speed exceeded 30 m/s. Major

structural damage, including loss of

roof structure, collapse of singlestory masonry buildings, complete

destruction of mobile homes, and

extensive damage to older preengineered metal buildings and

wood-framed construction, took

place where the gust wind speed

exceeded 50 m/s. However,

damage was dependent on the

upwind terrain and shelter. In

sheltered locations even old woodframe houses remained completely

undamaged where the gust wind

speed in open country exceeded

55 m/s.

South Carolina is a heavily forested

state. The trees provided

significant protection from the

wind and undoubtedly reduced the

amount of damage to buildings.

However pine trees apparently

began to fall as the wind speed

approached 30 m/s and many

buildings, although protected from

the wind by the trees, subsequently

suffered damage from them.

In the Bulls Bay area, where few

trees were left standing, the

exposure of buildings would have

changed significantly during the

course of the storm.

Two other factors greatly reduced

the wind damage. Firstly, the

densely populated islands of Folly

Beach, Sullivans Island and the Isle

of Palms experienced the eye of the

storm. This meant that the

strongest winds would have come

from the north to northeasterly

direction and south to

southwesterly direction i.e. along or obliquely to the shore. Many

ocean-front buildings would

therefore have received shelter

from neighboring buildings.

Secondly, the strongest winds in

the storm affected the most lightly

populated area of the coast,

occupied primarily by a wildlife

refuge. In this area all settlement

is landward of the Intracoastal

Waterway in forested areas. Had

the storm hit 30 km to the

southeast, Charleston and the

adjacent islands would have

experienced onshore winds

producing wind loads on oceanfront properties perhaps double

439
those actually experienced. Had

the storm hit 80 km to the

northwest all of the buildings on

the densely populated northern

coast would also have experienced

wind loads more than twice those

actually experienced.

5,1.1 Single Family Dwellings

In general these performed

reasonably well, probably more as a

result of shelter than any other

factor. Major structural failures

resulted from loss of roofs, shear

failures and foundation failures.

Failures of roof structures were

relatively rare, probably due to the

use of metal anchors to attach roofs

to walls in exposed coastal

locations. However, in the highest

wind areas, failures occurred even

where anchors had been used,

particularly after window damage.

A typical roof failure which led to a

general collapse is shown in Figure

2.

Failure to provide bracing led to

some shear failures in two story

buildings. The traditional bracing

techniques proved to be

insufficient, especially in buildings

with few interior cross-walls and

large openings in exterior walls.

In most coastal flood-prone areas

buildings are elevated on wood or

prestressed concrete piles. On the

South Carolina coast many buildings

had been constructed on concrete

block piers. The local building

code had permitted unreinforced

masonry piers to have a height to

width ratio of 10, if filled with

concrete. Piers meeting this

requirement failed, as did those

that were lightly reinforced (Figure

3). In meeting the elevation

requirements for flood insurance,

the very large forces applied to the

piers by the wind acting on the

building had apparently been

overlooked.

Many single family dwellings lost

shingles. These apparently minor

wind failures resulted in significant

rain damage, often magnifying the

initial damage by a factor or 10.

Repeated on an enormous scale,

this contributed significantly to the

total cost of the storm.

In some areas wind and surge

damage to single-family dwellings

was difficult to differentiate,

particularly in older buildings built

on grade or newer buildings

meeting the flood insurance

elevation requirements, but

subjected to surge in excess of the

design level, as in the northern

part of the Isle of Palms and in

Bulls Bay. In some instances wind

debris was transported by the surge

making the nature of the wind

failure difficult to determine.

5.1.2 Mobile Homes

As mentioned earlier these

buildings had a failure rate far

higher than other classes of

structures. Their vulnerability was

recognized and the Governor of

South Carolina ordered the

evacuation of all mobile homes in

the coastal region prior to the

arrival of Hugo. This probably saved

many lives.

Structural collapse of mobile homes

was often sudden and complete,

either by overturning or by the

separation of the superstructure

from the chasis. A typical example

440
is shown in Figure 4. In general

the more stable "double-wide"

homes, formed by joining two

"single-wide" units together,

performed better than "singlewide" homes.

5.1.3 Multi-Family Wood-Framed

Construction

Several of these professionally

designed buildings in exposed

locations suffered extensive roof

failures, often precipitated by

window damage. In some of these

buildings metal anchors had been

used to secure the roof. Whether

or not these had been selected

after a structural analysis is not

known, but consideration of

internal pressure would not have

been required by the building code

and the uplift loads could have

exceeded the design value after the

failure of windward windows,

5.1.4 Masonry Walled Buildings

Failures of unreinforced masonry

buildings were observed over a

wind area (Figure 5). The

instability of these buildings has

been documented in several past

hurricane investigations, but the

form of construction is still

permitted by many local building

codes.

5.1.5 Pre-Engineered Metal

Buildings

A number of these buildings located

in exposed locations suffered

serious damage (Figure 6). Being

light and very efficiently designed,

these structures were very

sensitive to the under-specification

of wind loads in earlier building

codes. Buildings designed to

increased requirements in recent

years appeared to have performed

much better.

5.1.6 Steel and Concrete Framed

Buildings

As observed in other hurricanes,

the frames of buildings designed by

structural engineers to meet the

wind loading requirements in the

building code preformed well. The

wall cladding, roof system, and roof

covering, often selected without

reference to the wind-load

requirements, performed very

poorly.

An exterior wall system which had

been employed extensively in the

area used gypsum board attached to

metal studs. The gypsum board

was covered with insulation and

stucco. Based on published data on

the ability of such systems to resist

wind suction, these systems should

not have been used on any building

over 10 m high located on the

South Carolina coast. Nevertheless,

this system had been used

extensively on medical

facilities, educational buildings,

condominiums, office buildings,

and hotels. Failures were found on

the coast as far north as the North

Carolina border where the actual

wind speeds were less than half of

the design value. Several buildings

in the Charleston area, including

hospitals and other critical

facilities, were seriously damaged

(Figure 7). Window systems also

failed due to positive pressure and

suction, giving rise to extensive

internal rain damage.

Failures of flat-roof coverings, both

build-up-roofs and single-ply

systems, sometimes coupled with

441
the roof-deck failures, were also to

be found along the South Carolina

coast as far as the North Carolina

border and inland as far as

Charlotte. These failures, which

included schools used as

emergency shelters, hospitals, and

other critical facilities, resulted in

extensive rain damage to the

interior structure and contents of

the buildings. A limited survey of

insurance data indicated that the

rain damage in multi-story

buildings was often 20 to 30 times

the actual wind damage to the

structure and multi-million dollar

claims occurred in areas where the

storm had a recurrence interval of

less than 20 years.

The local building code, in use

when most of these buildings were

designed, underestimated the

uplift loads on the edges of roofs.

The design loads were almost

certainly exceeded in many places

in Hugo. In some cases the

shortcomings in the building code

may have led to the selection of an

adequate system, but in many cases

the system was probably chosen on

the basis of cost rather than winduplift resistance.

5.2 Detailed Survey of Surge

Damage

Storm surge damage was observed

along 240 km of coastline from

Folly Beach near the entrance to

Charleston Harbor to Cape Fear

50 km beyond the North CarolinaSouth Carolina border. Shallow still

water flooding of sheltered lowlying buildings occurred throughout

the area, including the older parts

of the city of Charleston. The worst

structural damage resulted from

wave and current action along the

beachfront and the unprotected

shoreline of Bulls Bay.

5.2.1 Single Family Dwellings

Many older buildings built on grade

or on low piers floated off their

foundations. Those that were

better anchored or heavier

experienced damage from rising

water or, if near the ocean,

extensive damage by wave action

(Figure 8).

As mentioned in the survey of wind

damage, many oceanfront houses

had been elevated on slender

unreinforced or lightly reinforced

masonry piers. These were

supported on shallow concrete

footings, some less than 0.5 m

below grade. In addition to lacking

resistance to wind forces

transferred to them by the

supported building, many piers

became unstable when erosion

exceeded the embedment depth.

Several hundred oceanfront

buildings of this type collapsed and

many more were damaged. Other

buildings had been supported on

cast-in-place concrete columns.

Many of these were also underdesigned, poorly constructed and

lacked proper embedment. They

also performed poorly (Figure 9) as

did wood piles that had very short

embedments, some as little as 1 m.

Buildings constructed on wood or

prestressed concrete piles,

embedded at least 3 m, generally

performed well, although in some

cases buildings were damaged

when the wave height exceeded

the lowest floor elevation.

442
Wave and current action extended

farther inland than predicted by

the flood insurance maps. Some

foundations, particularly solid

masonry walls, had therefore been

selected on the assumption that

they would be subjected only to still

water flooding. These foundations

performed poorly when subjected

to a combination of wind, wave, and

current action.

5.2.2 Other Buildings

Ocean-front condominiums, hotels

and other commercial buildings

had usually been supported on

piled foundations, but the first

floor of many of these buildings had

not been elevated above the 100

year design flood level and wave

damage was quite common.

When the bearing walls were

perpendicular to the shoreline the

first floor rooms were gutted by the

waves, but the upper floors

remained stable and undamaged.

In the few cases where the bearing

walls were parallel to the shoreline

the first story bearing wall failed

and parts of the two or three floors

above collapsed.

Some buildings had used slab-ongrade floors at the lowest level

which were independent of the

piled foundation supporting the

rest of the building. Even small

amounts of erosion under these

slabs caused their collapse.

Many sea walls, bulk heads and

retaining walls failed, leading to the

destruction of almost all oceanfront

swimming pools in the Myrtle

Beach area.

6. PERFORMANCE OF UTILITIES

The electric utility companies had

constructed transmission and

distribution systems which were

very vulnerable to wind damage.

The decision to use wood poles for

high voltage transmission systems

and overhead local distribution

systems was probably made on

economic grounds. Although

interruption of power supply is

common in the region in severe

thunderstorms and had been

experienced in earlier hurricanes,

the extreme vulnerability of the

system had not been appreciated by

the utility companies nor their

customers. It had certainly not

been appreciated that loss of

electric power for an extended

period of time would have such an

effect on the basic essentials of

life - the ability to obtain food and

water and the provision of

wastewater facilities.

Approximately 1.5 million people

living in areas where the wind

speed exceeded 30 m/s lost

electric power. The utility

companies made tremendous

efforts to restore power, but the

extent of the damage was so great

that in many cases the system

could not be repaired, but had to be

rebuilt, usually to the previously

existing standards. Eight days after

the storm only 23% of the

customers in the Charleston area

had power and in some rural areas

it was two to three weeks before

power was restored.

The telephone companies had

attempted to reduce the

vulnerability of their systems by

burying cables. In most areas there

was very little interruption of

443
telephone service. Other forms of

communication, including

television and radio broadcasts and

even two way radio

communications of emergency

services, were seriously affected by

loss of electric power and the

collapse of transmission towers.

7. CONCLUSIONS AND LESSONS

LEARNED

Hurricane Hugo taught South

Carolina and the nation some very

expensive lessons. The potential

for disaster had been recognized

but the political and economic

systems made it difficult to remedy

the situation. It is worth examining

what went wrong.

The state had chosen a laissez-faire

system of building control. The

power to adopt and the

responsibility for enforcement of

building codes had been delegated

to local jurisdictions. This resulted

in considerable variation in the

quality of control, from some of the

best in the country to no control at

all. The model building code adopted

showed a very poor understanding

of the wind and surge effects on

structures and permitted forms of

construction in wood and masonry

which were inconsistent with the

loading requirements. Lacking

proper deemed- to-comply

documents, the performance

requirements of the code were

unenforceable for non-engineered

structures. Using the permitted

forms of construction in the code

led to structural systems unsuitable

for a hurricane, tornado, and

earthquake prone area such as

South Carolina.

In poor rural areas mobile homes

were the only affordable form of

housing for many people. These

homes were built to inadequate

standards and when delivered to

site were often not properly

anchored.

Despite the fact that cladding

failures can pose a threat to public

safety and even compromise the

structural integrity of a building,

cladding systems had been

specified by those without adequate

training in structural design. The

professional licensing laws had

permitted this.

In recent years owners had often

been unwilling to pay for adequate

professional inspection during

construction. The quality of

construction had suffered.

Compounded by poor building code

requirements, the use of

inappropriate design professionals

and inadequate material testing

standards, this had led to cladding

systems and, in some cases,

structural systems with very low

wind resistance.

The insurance industry had been

willing to insure property with

little regard for its wind resistance.

Premiums were often based on the

building's fire resistance. Thus

there was little or no incentive for

the owner to improve wind

resistance in order to meet

insurance requirements.

The electric utility companies had

been permitted to construct

transmission and distribution

systems which were very vulnerable

to damage by a hurricane. The

public was unaware of how fragile

444
the system was or how dependent

they were on a continuous supply of

electricity.

Finally, there was the lack of

experience. For 30 years prior to

Hugo, during which time the

hurricane-prone areas of the state

expanded rapidly, no major

hurricanes hit the South Carolina

coast.

From an engineering standpoint

the dominating lesson is clear and

the solution is straightforward.

Design and construction practices

must reflect the loads which are

known to occur in conditions with

reasonable recurrence intervals.

The prospects for achieving this

and correcting past mistakes are

not good.

The building stock changes slowly

and the fragmented building

control system in the United States

is ill-equipped to impose improved

construction standards. Based on

the success of the National Flood

Insurance Program, the best

prospect for improved construction

is by insurance industry action.

That industry certainly learned its

lesson in Hugo when it paid out the

equivalent of all property insurance

premiums collected in South

Carolina in the last 5 years.

Not all lessons learned were

concerned with what went wrong.

Properly designed buildings

performed well and the telephone

service was a fine example of what

can be achieved with good

planning. Perhaps the most

valuable lesson learned was that

modern technology can be used

very effectively to monitor storms

and warn coastal residents of

impending danger.

In 1893 a storm of similar strength

to Hugo drowned 1000-2000

people in South Carolina, Prior to

Hugo's landfall over 250,000

people were evacuated from the

coastal area. Only a handful of

people were killed, none by

drowning in their homes. It is a

poor reflection on the construction

industry that many evacuees

returned to buildings that were in

considerably worse condition than

their counterparts after the 1893

storm.

8. ACKNOWLEDGEMENTS

This study of the effects of

Hurricane Hugo was made possible

by grants from the National Science

Foundation, South Carolina Sea

Grant Consortium, South

Commission on Higher Education,

the Federal Emergency

Management Agency and Clemson

University. This support is

gratefully acknowledged.

445
Table 1. Insured Damage to Buildings and Contents in South Carolina

Number of Claims Total Amount

($ million)

Wind

Conventional dwellings 287456 (74%) 1423 (55%)

Mobile homes 57808 (15%) 172 (7%)

Commercial property 41282 (11%) 997 (38%)

Total wind damage 386546 2592

Flood

All types 15739 320

Table 2. Number of Dwelling Units Destroyed or Damaged in South Carolina

Destroyed Major Minor * Total

Damage Damage

Single family dwellings 3783 18146 57698 79627

Apartments 313 2647 8948 11908

Mobile homes 5200 5976 9063 20239

Total 9296 26769 75709 111774

Externally observable damage only, excludes damage to outbuildings,

fences, decks, porches etc.

446

-i-. ^^^^ Figure 2. Wood-Framed

House,

Folly Beach

Figure 3. Masonry Pier Failure

by Wind Action,

Isle of Palms

Figure 4. Mobile Home

Near Sumter

448
449
450
United States Federal Agency Programs in Disaster Countermeasures

By

Paul F. Krumpe

Office of U.S. Foreign Disaster Assistance

Agency for International Development

Department of State Washington, D.C. 20523 ^ and

Kenneth F. Sullivan

Earthquakes and Natural Hazards Division

Federal Emergency Management Agency

Washington, D.C. 20472

INTRODUCTION

The purpose of this paper is to present an

overview of the roles and responsibilities of U.S. Federal Agencies and the major programs they are conducting which involve mitigating the impacts of

natural and man-made or technological hazards on

vulnerable populations, economic investment and

critical facilities both in the United States and

abroad. The United States through its many

Federal Agencies, state and local governments is committed to protecting its people, businesses,

communities and economic resource base in the

face of disaster.

According to the U.S. National Academy of

Sciences, Advisory Committee on the International

Decade for Natural Disaster Reduction, as many as 70 million people in 39 states face significant risk from earthquakes and secondary hazards such as

landslides. It is estimated that more than $100

billion in total economic losses could occur from a

single major California earthquake. Billions of

dollars in damage could occur from a devastating

hurricane such as Agnes in 1972 or Hugo in 1989

which we know resulted in $9 billion in losses,

including 8,800 square miles of downed timber. The

U.S. experiences some 900 tornadoes annually;

wildfires destroy hundreds of thousands of acreage

each year and volcanoes are a constant threat. The

eruption of Mount St. Helens in 1980 caused over $1 billion in damage. Recent estimates indicate

landslides cause $1 to $2 billion in economic losses each yeai with flood disasters ($1 billion annual

loss) and drought impacts ($40 billion estimated loss

in 1988 alone) causing recurring intermittent losses.

The Office of U.S. Foreign Disaster

Assistance, Agency for International Development,

estimates that more than 3 miUion people have died

and more than 900 million have been seriously affected worldwide by disaster during the past 22

years. Property damage and economic losses are

believed to exceed $75 billion during the same

period, although data collected over the years remains imprecise.

The magnitude of destruction and human

loss both physical, social, and economic is dramatic

when viewed on a globed scale and at the national

level. Governments have, in several instances, failed following significant disasters and their impacts.

We, in the United States, recognize that we live in a disaster-prone world and nation and have

determined to take concerted action to mitigate

these negative impacts now and in the future.

DISASTER REDUCTION PROGRAMS OF THE

U.S. GOVERNMENT

Within the 14 cabinet level departments of

the federal government and more than 70

independent agencies, there are thousands of

bureaus and offices dealing with various aspects of

hazard reduction either at the personal, local, state, federal or global scale. BiUions of dollars are expended annually to monitor, regulate, control,

document and respond to the many facets of naturcd and technological hazards as they affect the American people both in the pubUc and private

451
sectors. The nature of our government is to protect

its citizens, as well as to provide humanitarian

assistance to those in need abroad. The lead U.S. domestic agency responsible for disaster

management and civil defense is the FederzJ

Emergency Management Agency (FEMA). The

Agency for International Development's (A.I.D.)

Office of U.S. Foreign Disaster Assistance (OFDA)

coordinates U.S. international disaster assistzmce to host countries abroad. Both FEMA and OFDA

work closely with other federal agencies to ensure

effective and timely implementation of disaster countermeasures as mandated by legislated laws and

regulations at aU levels of government. The federal departments and their many agencies have specific

missions and progrjuns designed to prevent,

mitigate, prepare for, respond to, rehabilitate or otherwise ameliorate the impacts of disaster.

The Lead Role of the Federal Emergency

Management Agency (FEMA) in Earthquake

Hazards Reduction

Under the 1980 amendments of the U.S. Earthquake Hazards Reduction Act, the Federal

Emergency Management Agency (FEMA) was

assigned the leadership role for the National

Earthquake Hazards Reduction Program (NEHRP).

FEMA's leadership responsibilities include a range

of activities that focus on planning and coordinating

the work of the NEHRP, and broadening its scope

of participation and effect.

The purpose of the National Earthquake

Hazards Reduction Pro-am is to reduce the risk to

lives and property. This is accomplished through a comprehensive, multi-agency program of scientific research, mitigation, preparedness and response

planning and public education. FEMA, as the lead agency, has the statutory responsibihty to plan,

coordinate and recommend goals, priorities and

budgets for earthquake activities among the

principal agencies authorized under the Earthquake

Hazards Reduction Act of 1977, as amended. The

agencies include the United States Geological

Survey, the National Science Foundation and the

National Institute for Stzmdards and Testing.

The primary activities of the program are

to: (1) develop improved seismic design

and construction practices for adoption

by Federal agencies. State and local

governments and the private sector;

(2) provide financial and technical

assistance to State cuid local goverimients

to implement comprehensive earthquake

haizard reduction programs;

(3) develop pubhc education and

awareness progr2uns and

(4) plan for and coordinate an

adequate Federal capability to respond to a catastrophic

earthquake.

FEMA's Role in Managing Other Natural and

Technological Hazards and Emergency Response

FEMA's natural and technological hazards

programs include the following elements: (1) National Earthquake Hazards Reduction, (2) Hurriceme Preparedness, (3) Dam Safety, (4)

Radiological Emergency Preparedness, (5)

Hazardous Matericils and (5) Chemical Stockpile

Emergency Preparedness Program.

The goal of the Hurricane Preparedness

Program is to reduce the loss of life and property

damage from hurricanes in high-risk populations

and to develop State Hurricane Evacuation Plans. FEMA, as the chair of the Interagency Coordinating

Committee on Hurricanes, coordinates ongoing

hurricane-related planning and mitigation activities of the U.S. Army Corps of Engineers, the National

Weather Service, the National Hurricane Center

and the Office of Ocean and Coastal Resource

Management.

The objective of the Dam Safety Program is to enhance the safety of the Nation's dams, thereby

protecting lives and property. FEMA exercises dual

responsibilities through its Dam Safety Program to

(1) coordinate Federal dam safety activities and (2) coordinate and implement activities designed to encourage States to implement strong dam safety

programs.

The mission of the Hazardous Materials

Program is to provide technical and financial assistance to State and local governments. In

addition, FEMA coordinates £md cooperates with

the private sector in developing, implementing and

evaluating hazardous materials emergency

452
preparedness programs for State and local governments. The mission is accomplished through

five separate functional elements—planning, training,

exercising, information exchange and

intergovernmental coordination/cooperation,

FEMA develops and distributes planning and

preparedness guideince to State and local governments in cooperation with the 13 member

agencies of the National Response Team.

Hazardous materials training courses and course

materials are developed and financial assistance is provided to State and local governments.

As a result of a Presidential Directive in 1979, FEMA was assigned the lead Federal role for radiological emergency planning and response. Under FEMA's Radiological Emergency

Preparedness Program, the goal is to enhance

integrated emergency planning and response for all types of peacetime radiological emergencies by the

State, local and Federal governments. The primary

emphasis is directed to planning and preparedness

for commercijil nuclear power plants, nuclear fuel cycle and material Ucense holders. Department of

Defense 2md Department of Energy facilities and

transportation accidents. Key activities pertaining to

offsite radiological emergency planning and

preparedness include evaluation of emergency

response and utility plans, review of public

emergency information materials, review and testing

of utility alert and notification systems, periodic

exercises to test emergency response plans and

periodic program activities such as drills, plan

updates and public meetings.

FEMA, more than any other Federal

agency, provides the financial and technical

assistance to State and local governments to prepare

for, respond to, and recover from natural disasters. This is done through the Civil Defense Program,

which provides a system of survival capabilities to protect life and property from natural hazards

through the construction of emergency operation

centers, warning systems, and training of State and

local emergency managers. Other relevant FEMA

programs assisting natural hazard reduction efforts of State and local governments are (1) The national

Flood Insurance Program, which makes flood

insurance backed by the U.S. Government available

to communities that have agreed to adopt and

enforce floodplain management ordinances to reduce future flood losses through wise utilization of

their floodplains; and (2) The Fire Prevention and

Control Programs of the U.S. Fire Administration,

which provides training £ind grants that support the

efforts of State and local governments to reduce the number of fires nationwide and the loss of life and

property from fire.

The Role of the Department of the Interior's U.S. Geological Survey (U.S.G.S.) in Geophysical

Hazards Countermeasures In the United States A principal goal of the National

Earthquake Hazards Reduction Program is to

identify, evaluate, and characterize accurately the seismic hazard in areas subject to high or moderate

seismic risk. The U.S. Geological Survey (USGS)

is the principal agency responsible for addressing

this goal. There are two primary program

objectives: to determine the tectonic and geologic

framework emd earthquake potential of the various

earthquake-prone (or seismogenic) zones in the

United States through detailed seismological and

geological studies; and, as these zones are defined

and their seismic potential quantified, the hazards

and risks to the population, structures, and lifelines must be identified, evaluated, and documented.

Decisions made by various governmental and

private sector entities regarding construction

practices, land use planning, and disaster preparedness should be based on accurate and

complete assessments from work under this prograim element.

The USGS also conducts earthquake

engineering research focused on strong groundmotion data collection, processing and analysis,

maintains centers for seismological data collection amd dissemination and leads the nation's earthquake

prediction research programs in prediction

methodology and experiments; theoretical,

laboratory and fault zone studies; and induced

seismology research.

The USGS currently is managing several projects focusing on earthquake hazards in the Eastern United States. These include: studies of

the structure of the crust in earthquake prone areas, examination of faults, calculation of horizontal

stresses over large areas, modeUng of the local ampUfication of shaking due to earthquakes, and

earthquake monitoring networks. The Parkfield

Earthquake Prediction experiment on the San

Andreas fault in central Cahfornia is now fully

operational and being monitored in real time. The

use of high-precision geodetic surveying across large

453
distances of the Earth's surface now allows scientists to monitor minute changes in the dimensions of the

Earth's crust in earthquake-prone regions. This

information provides a more accurate assessment of

where cind how stresses are accumulating in the

crust.

In earthquake engineering research the

USGS in cooperation with other agencies is conducting laboratory simulation of earthquake

strong ground motions which promises to increase

dramaticcdly our ability to lecirn about the effect of

damaging earthquakes on engineered structures. Research on spatial variation of strong ground

motion is helping to illuminate the patterns of local damage during earthquakes. Understanding this variabiUty is prerequisite to controlling it by means

of protective measures or appropriately conservative

seismic design.

The USGS is at the forefront of producing

and deploying airrays of new digital seismographs in the U.S. and elsewhere in the world. The southern

Quebec Earthquake of November 25, 1988, was a

real-time test of the Earthquake Early Alert Service.

Understanding earthquake risk is fundcimental to disaster planning and mitigation

policy and implementation of countermeasures. Despite the Loma Prieta Earthquake, the risk of a damaging earthquake in the San Francisco Bay

region still confronts nearly 6 million people. A

new study, released in July 1990 by the USGS, says

that there is approximately a 70 percent chance of

another earthquake equal to or larger than Loma

Prieta int he next 30 years, meaning that the

earthquake is twice as likely to happen as not. The

same report examined iadividuid segments of Bay

Area faults and gave probabUities of major

earthquakes on each of those segments during the next 30 years. It is important to recognize that there are vcirious potential sources for the next big

earthquake, and that these sources are much closer

to heavily populated areas than the Loma Prieta

event, hence the potential for loss is greater.

USGS Technical Support For International

Activities in Geophysical Hazards Reduction

Since 1983 the U.S. Geological Survey

(USGS) has worked closely with the Agency for International Development's Office of US Foreign

Disaster Assistance (AID/OFDA) in support of

geophysical hazards reduction programs in the less

developed countries (LDCs).

These programs have produced results that have placed AID/OFDA and the USGS in a highly

visible role as leaders in reducing the impact of

natural disasters. In addition to cooperating with

OFDA in crisis responses such as the catastrophic

eruption of Ruiz Volcano (Colombia), the toxic gas emissions of Lake Nyos (Cameroon), and the earthquake devastation in Armenia (USSR), the USGS provides ongoing technical assistance in OFDA's prevention, mitigation and preparedness

activities.

In 1984 the USGS and OFDA sponsored a Geologic and Hiizards Training Program in Denver

which has proven to be a milestone in the

technology transfer of earthquake countermeasures

information to the LDCs. The purpose was to develop hazard mitigation expertise in order to save

lives and reduce economic losses in countries where

geologic and hydrologic hazards are prevalent.

Forty-two international participants from 28

disaster-prone countries attended. The proceedings

volume (1,112 pp.), has been used extensively by

other organizations as a text for subsequent hazards

training courses.

In 1988 the USGS completed a series of

detailed earthquake conditional probability maps

and reports which cover the Circum-Pacific region

and show comparative earthquake and tsunami

potential for zones in the U.S., Central and South

America, Southeast Asia, Japan and Alaska.

The Volcano Monitoring and Research

Project in Indonesia has been a cooperative

AID/Jakarta-funded five-year volcanology program

between the Volcanological Survey of Indonesia and

the USGS. At the conclusion the agreement was

renewed by AID/OFDA. The purpose of the

program was to provide the on-site consultant

services of an eminent advisory volcanologist, who,

with a number of USGS scientists and technicians,

prepared a national volcano hazards assessment of

Indonesia and trmned the Indonesians so that they

could respond to volcanic crises. This work

included upgrading the volcano monitoring networks

and providing transfer of technology and expertise

in volcanology.

In 1986, as an outgrowth of the USGS's

Volcano Crisis Assistance Team (VCAT) and as a

result of the Ruiz Volcano disaster in Colombia, a

454
5-year program was set up between the USGS and

AID/OFDA. This cooperative program, equally

funded by the USGS and OFDA, has estabUshed

the U.S. capability of providing rapid, efficient and

cost-effective volcano emergency response, initially in Latin America and also worldwide. VDAP

provides on-site monitoring by a technicallyqualified volcano crisis assistance team, ongoing

mapping and analyses of volcano hazards,

preparation of data on the threat information to

facilitate evacuation planning, an equipment cache,

and training. To date VDAP has responded to volcanic threats in Colombia, Costa Rica,

Guatemala, Ecuador, and Peru.

The U.S. Geological Siuvey is currently

implementing the Worldwide Earthquake Risk

Management Program (WWERM) as a one-year

interagency cooperative pilot program (1990-91)

with OFDA. Subsequent activities will depend on

co-funding commitments from other donors and

organizations. The objective of WWERM is to develop an action plan and conduct the one-year

pilot program in each of three earthquake-prone

target regions to demonstrate the utility, importance, feasibihty and need for uniform global

earthquake hazard mapping and risk

assessment/management in preparation for a program linked to the International Decade for Natural Disaster Reduction. In each of the pilot

project's four target countries (Morocco, Peru,

Chile and Indonesia) the USGS, in cooperation with

host country experts, will prepare probability

earthquake ground motion maps and preliminary

estimates of expected physical losses within the

"buUt environment" as a basis for risk management,

life saving preparedness planning, and the

development of mitigation strategies.

The Department of Commerce's National Institute of Standards and Technology (NIST) Improving

Seismic Design And Construction Guidelines

The role of the National Institute of

Standards and Technology (NIST) in the NEHRP

is to conduct research and provide technical support

for the development and appUcation of improved

seismic design and construction practices. The

research conducted by NIST is driven by two

general objectives: to make new construction safe and economical with respect to eairthquake hazards

and to mitigate potential earthquake hazards in

existing facilities. Thus, laboratory and analytical

studies are conducted to provide technical

information needed for improving building codes

and standards pertaining to new construction, and to develop new criteria regarding the repair and

strengthening of existing structures.

Several years ago, NIST initiated a multi- year research program to develop technical data on

the behavior of connections between beams and

columns to provide a basis for a rational seismic

design method for precast beam-column

connections. Inelastic deformations are concentrated in the connections, which are smaller

and weaker than the precast elements being

connected. Thus, the connections are often the weak links in the structure, and their strengthening

is essential as an earthquake countermeasure. NIST

is also developing other strengthening techniques

which are aiimed at improving lateral strength and

ductility and are being apphed to deficient

structures. A set of design gmdeUnes for assessing

the overall capacity of strengthened reinforced concrete frame structures is being developed. A

prerequisite to the estabUshment of such guidelines

is knowledge of the interaction between existing structural members and strengthening elements and

of the behavior of the anchoring elements used to join the structural members.

The National Science Foundation's (NSF)

Responsibility For Upgrading Experimental

Research Laboratories in Earthquake Mitigation

Experimental reseairch in earthquake

engineering is conducted at many laboratories

located at universities across the United States. These laboratories vary from very large multipurpose facilities to small specialized facilities. NSF

is continually providing resources to upgrade and

expand the capabilities and instrumentation at these

faciUties. At the present time, several major

facilities exist that conduct research sponsored by

the NSF. These include The University of

California (Berkeley and San Diego), University of

Colorado, Cornell University, University of Illinois, State University of New York at Buffalo, University

of Texas at Austin, Lehigh ' Jniversity and The

University of Michigan.

Additionally the NSF has been at the

forefront of supporting information dissemination

centers and clearing houses in natural hazards. A

number of these include: the National Information

Service for Earthquake Engineering (NISEE) at the

University of California, Berkeley, and the

455
California Institute of Technology, which conducts

comprehensive earthquake engineering

dissemination programs; the National Information

Service of the National Center for Earthquake

Engrneering Research (NCEER) at the State University of New York at Buffalo; and the Natural

Hazards Information Center at the University of

Colorado, a major clearinghouse for social science

and poUcy information in the hazards field.

The information the NSF disseminates

reaches a diverse audience that includes engineers,

architects, building officials, planners, government

officials, and researchers. Numerous channels of

communication are employed to inform users about

the results of countermeasures research from

projects supported by the NSF in disciplines such as engineering, the social sciences, architecture, and

planning.

The Role of the Department of Commerce's

National Oceanic and Atmospheric Administration

(NOAA) In Mitigating the Impacts of Floods,

Severe Storms and Tsunamis In the United States and Abroad

The primary mission of the National

Weather Service (NWS) is the protection of life and

property and the enhancement of the national

economy. Hence, the basic functions of NWS are

the provision of forecasts and warnings of severe weather, flooding, hurricanes, and tsunami events; the collection, exchange, and distribution of

meteorological, hydrologic, climatic, and

oceanographic data and information; and the

preparation of hydrometeorological guidance and

core forecast information. The NWS is the single

"official" voice when issuing warnings for life- threatening situations and is the source of a common national hydrometeorological information

base. The national information base forms an

infrastructure on which the private sector can build

and grow.

Accurate and timely weather and river forecast and warning systems are vital to the safety

and well-being of the Nation's population. Weather

and water resources forecasting harnesses modern

advances in information to increase the productivity

of American industry, thereby contributing to economic growth.

The National Weather Service (1) collects and exchanges hydrometeorological data and

information on a national and international basis;

(2) issues warnings and forecasts of severe weather,

floods, hurricanes and tsunami events which

adversely affect life and property; (3) issues weather, river and water resources forecasts and

related guidance materials used to form a common

national hydrometeorological information base for the general public, private sector, aviation, marine,

forestry, agricultural, navigation, power interests,

land and water resources management agencies and

emergency managers at all levels of government; (4)

provides climatological simimaries, quality controls

for observed and collected data; and (5) other

services such as global, national or general regional

atmospheric models.

The National Weather Service (NWS) is charged with responsibility for the Tsunami

Warning Service and for the preparation and

distribution of Watches and Warnings of tsunamis

that may threaten lives and/or property in and

around the Pacific Ocean. The NWS operates and

administers the Pacific Tsunami Warning Center

(PTWC) and the Alaska Tsunami Warning Center

(ATWC). The PTWC functions as the National

Tsunzuni Warning Center responsible for providing

tsunami warning services for all coastal States and

other U.S. interests throughout the Pacific. The

PTWC also provides regional warning services for Hawaii; the ATWC provides regional warning

services for Alaska, British Columbia, Washington,

Oregon, and California. The PTWC is the International Operational Center for the Tsunami

Warning System in the Pacific.

The relationship between earthquake size and historic tsunami generation is being studied by

the PTWC on an area-by-area basis to determine

the influence of regional and local tectonics and

submarine geology on the generation of tsunamis.

These studies are providing the basis for a better evaluation of the present warning thresholds and

operational procedures.

The PTWC continues to work closely with

Hawaii Civil Defense in coordinating and supporting

State legislation for enhancement of tsunami

mitigation measures for the State of Hawaii. The

State Legislature approved funding for the

installation of nine additional tide gauges within the Hawaiian Islands, equipped with digital data access

via a computer telephone query. Each of the

County Civil Defense Agencies, as well as the State Emergency Operating Center, will have

456
microcomputers capable of automatically

interrogating and plotting sea level data from any

station using software developed at the PTWC.

Telefax units are also being installed to provide a

graphics commimication link between the PTWC

and all Civil Defense Agencies.

The National Geophysical Data Center

(NGDC) located in Colorado acquires, processes, and analyzes technical data on earthquake hazards,

and disseminates the data in many usable formats to Federal agencies, private industry, academia, and

the pubUc.

The NGDC Pacific Tsunami Data Base

includes more that 1,600 events since 49 B.C. and more than 5,100 locations where tsunamis were

observed. Times of generating earthquakes,

tsunami arrival times, travel times, first motion of

the wave, and periods have been added to the data

base. The data base has been designed for use on

IBM-compatible personal computers.

The Department of Enei^ (DOE) Program in Environment, Safety and Health For Mitigating

Technological, Earthquake and Other Hazards

The Department of Energy (DOE) plays a major role in the National Earthquake Hazard

Reduction Program (NEHRP) and is committed to supporting the program through the Department's

environmental, safety, and health obhgations. The

support has been forthcoming from all levels of the Department, especially from the operations

offices and power administrations that serve as NEHRP contacts. The Department's

representatives have participated in NEHRP

activities of the Interagency Committee on Seismic

Safety in Construction and the Subcommittee on

Federal Earthquake Response Planning.

In April 1989, the Department issued DOE

Order 6430.lA, General Design Criteria, which

specifies the use of design basis earthquake levels for nuclear processing facilities and other hazardous

operations based on probabilistic hazard curves. The DOE Order requires the use of the Design and

Evaluation Guidelines for DOE Facilities Subjected

to Natural Phenomena Hazards. This was re-issued

as an interim report in 1989, and its methodology is based on the approach used in the Uniform

Buildmg Code UBC (1988 Edition) and the U.S. Army's Seismic Design Guidelines for Essential

Building. DOE's Office of Safety Appraisals is sponsoring workshops on the use of the guidelines.

The Department of Energy (DOE)

continues to strengthen the awareness and broaden

the involvement of its program offices, field orgcinizations, and energy contractors in the

National Earthquake Hazards Reduction Program

(NEHRP). The following major organizational

units of DOE are involved in natural phenomena

hazard mitigation programs: 1) Management and

Administration; 2) Defense Programs, 3) Energy

Research, 4) Environment, Safety and Health, 5) Fossil Energy, 6) International Affairs and Energy

Emergencies, 7) Nuclear Energy, 8) Civilian Radioactive Waste Management, 9) Power

Marketing Administrations, 1) Operations Offices

and 11) Energy Technology Centers throughout the U.S.

These organizations are responsible for implementing the requirements of Departmental

Orders, including specified consensus standards

(e.g., seismic requirements of the Uniform Building

Code (UBC) cind the Seismic Design Guidelines for Federal Agencies).

Department of Defense (DOD) Activities in Civil Works, Research and Education for Disaster

Reduction and International Humanitarian

Response

The work of the U.S. Army Corps of

Engineers (CE) military and civil works programs

and activities in earthquake, tsunami and flooding

countermeasures conducted by the U.S. Navy are important contributions to the protection of critical coastal facilities worldwide.

The Loma Prieta Earthquake demonstrated

the need for an increased effort in the area of

structural evaluation of existing structures and

methods of retrofitting. This has set the stage for research work on base isolation and elastic damping

systems. A large percentage of essential miUtary

installations are located in areas of high seismic

activity, and therefore should be ^ble to function

during and after a seismic event. Many of these

facilities are constructed of masonry units and are vulnerable to seismic damage.

The Corps of Engineers continues to develop and improve its seismic design and

457
construction requirements, using the latest seismic

design provisions and innovative design concepts for water resource type structures. A new geotechnical

seismic design manual applying the latest findings of

research for the practical design of shallow and

deep foundations, retaining walls, buried structures,

slopes, and embankment dams is in prepeu^ation.

The manual, developed with the U.S. Naval Civil Engineering Laboratory, provides geotechnical

design guidance for waterfront structures. In

addition, structural engineering guidance is being

prepared for the seismic design of concrete towers

required for the intake of outlet works for dams.

The U.S. Navy Facilities Engineering

Command continues to investigate naval shore

facihties located in areas of high seismic risk. The

program is approximately 85 percent completed and

will ultimately involve 95 Navy activities. Areas of

vulnerability to potential liquefaction, tsunami

flooding, and landslides have been identified, and

widespread deficiencies are being addressed.

Approximately 14,500 structures have been

screened, approximately 1,600 structures analyzed,

and 650 found to be seriously vulnerable. For the

structures identified as hazardous, nonstructural

components as well as the lateral force resisting

system were evaluated. In areas where site hazards

are identified, they are shown on planning

documents, and new construction restricted or adapted to provide protection.

Bureau of Reclamation Earthquake and Flood

Hazards Reduction Applications Research And

Implementation

The Bureau of Reclamation (Reclamation)

earthquake hazards reduction activities basically

involve seismic analyses and the designs or modifications of embankment and concrete dams

and large structures. These seismic analyses require a variety of state-of-the-art technical studies,

including regional and local seismotectonic

investigations, microseismic monitoring, subsurface

drilling and trenching explorations, materials

sampling and testing programs, and various types

and levels of dynamic analyses, including risk-based

analyses. The studies include the installation of

field instrumentation and the development of a wide

latitude of seismically related programs.

Reclamation also conducts research related to earthquake effects on materials and structures. Projects underway in Fiscal Year 1990 include

centrifugal modeling of embankment dam cracking

and soil strength and deformation following

liquefaction.

The Bureau of Reclamation is the lead

technic£il agency in the Department of the Interior's Safety of Dams Program and is involved in seismic

issues throughout the contiguous 48 States and

Alaska. There are, at present, 385 dams in Interior's Safety of Dams Program. Of these, 269

are Reclamation, 52 are Bureau of Indian Affairs

(BIA), 29 are Fish and Wildlife Service (FWS), 26

are National Park Service (NPS), and 9 are Bureau

of Land Management (BLM). This Bureau has

identified about 70 dams that may require

modifications for safety reasons; however, most of

these safety issues are not associated with

earthquakes.

Within the Bureau of Reclamation, a major

program is continuing to evaluate the safety of

existing dams and foundations of new dams in highly seismic areas. Dams or dam foundations that are suspected of being liquefiable are first evaluated

using in situ field testing and simplified empirical

methods. Techniques for determining in situ void

ratio and the steady state of deformation are used

in the evaluation of liquefaction potential on

projects for which the simplified empirical methods

do not give a definitive result.

The Role of the Department of Transportation In Protecting Critical Facilities and Lifelines In the

U.S.

While the Department of Transportation

(DOT) is emphasizing seismic safety in new

construction, its long term objectives are to develop

prioritized courses of action involving mitigation or

retrofit programs for existing buildings and hfeUnes

in high risk areas, and to promote long term

research into ways to increase seismic resistance for both new and existing structures. DOT also expects

that this seismic safety program will have the added

benefit of reducing loss and disruption risk from

other natural and man-made hazards affecting

transportation systems.

A goal of DOT'S hazards reduction

program is to achieve earthquake resistance

commensurate with the seismic risk for that region

for new structures and lifelines. All agencies of the Department of Transportation (DOT) are encouraged to identify cost-effective, low expense

activities for quick enhancement of their present

458
programs. In this regard, procedures are being

established for a future review of the vulnerability of

existing DOT facilities in areas of higher seismic

activity.

Another goal calls for the review of DOT

Orders that govern leasing of space, with adherence

to local building codes on seismic safety as an

eveduation factor in renting building space. Agencies are also encouraged to continue building a seismic knowledge base and sponsoring low cost research. Underlying these objectives is the Departmental pohcy to develop awareness and

information programs for all its constituencies and

employees on the need for increasing seismic

resistance for the Nation's transportation system.

The 1971 San Fernando earthquake was a major turning point in the development of seismic

design criteria for bridges in the United States. The

earthquake started a chain of events which led to the American Association of State Highway and

Transportation Officials (AASHTO) adopting, in 1983, a set of guidelines prepzu-ed by Federal

Highway Administration (FHWA) to establish design and construction provisions for bridges to minimize their susceptibility to damage from

earthquakes. DOT is also concerned with the

additional effects of dynamic earthquake-induced

ground shaking on the behavior of structures and on

soils underlying ports and harbors, as well as tsunamic threat.

The Department of Veterans Affairs (VA) Hospitals

and Medical Facilities Protection Against Disasters

The Department of Veterans Affairs (VA)

develops plans and strategies for the provision of

emergency health care services to veteran

beneficiaries in Veterans Administration medical

facihties, to active duty personnel and where

possible to civilians in communities affected by

national security emergencies. The VA works

closely with other USG agencies to apply applicable

building codes and construction techniques to protect its facilities and equipment against natural

disasters. The VA also assists The Department of

Health £md Human Services (HHS) in developing

national plans to mobilize the health care industry

and medical resources during times of national

emergency.

The Department of Veterans Affairs (VA)

seismic strengthening program strives to ensure the

safety of patients and staff through the seismic

strengthening or replacement of all essential substandard buildings in seismic-risk areas. Since

the inception of the program in 1971, approximately

120 buildings have been strengthened, are being

strengthened, or are presently in the design stage of a construction project for strengthening. The

seismic structural strengthening program has made

significEint progress toward achieving its objective,

although much more remeiins to be done.

The Loma Frieta Earthquake prompted the

VA to reexamine its seismic correction program. A

comprehensive study of VA requirements in this areas has been completed. There is an increased

priority for projects that address seismic

deficiencies. The seismic correction program has

been instrumental in proposing funds for seismic

projects; $5.9 miUion for design of seismic

correction at Long Beach Medical Center was

budgeted for Fiscal Year 1990, with construction

funding of $67.2 million planned for Fiscal Year

1991. Design funding of $15.3 million for seismic

corrections at the Martinez Medical Center is planned for Fiscal Year 1991, with construction

funding of $157.8 million planned for future fiscal years.

Commensurate with a recent VA study, the

development of future budgets will continue to give

high priority to projects at medical centers requiring

seismic correction, including future projects at Memphis, Albuquerque, American Lake, and Manlo

Park. The Loma Frieta Earthquake significantly

damaged the VA medical center in Palo Alto,

California, and necessitated an emergency response

by the VA. The VA has been provided $54 milUon

in emergency funds for medical supplies and staff equipment, temporary bed and support space, seismic upgrading, and the design and site preparation for a replacement facility. The total project cost of the replacement facihty is $252

milUon.

Also included in VA earthquake hazard

mitigation efforts is a wide range of programs

initiated to: (1) Anchor major mechanical and

electrical equipment to prevent dislodging or

disruption during an earthquake. (2) Install an emergency radio network to provide direct communication between any of the VA facihties in the United States; (3) Provide emergency utihty

services, especially water and electrical power, in all VA medical centers; and (4) Install special

459
earthquake provisions for equipment, furniture, and

supplies to assure the operation of VA medical

centers in the postearthquake period.

Nuclear Regulatory Commission (NRC) Seismic

Countermeasures for Civilian Nuclear Electric

Power Plants

Under its congressional mandate to regulate commercial production and use of nuclear

materials, the Nuclear Regulatory Commission

(NRC) conducts a variety of regulatory activities and manages a strong research program to support

its regulatory operations. NRC-supported research

encompasses the areas of seismicity, tectonics, and

stress/strain measurements. These studies provide

a scientific basis for regulating the activities of the

nuclear industry and improving the safety of nuclear

plants and waste disposal facilities. NRC-supported

reseairch programs in these areas contribute to the

efforts of the National Earthquake Hazards

Reduction Program (NEHRP).

Since 1980, the NRC has sponsored

analytical and experimental research on low-rise

reinforced concrete shear walls, buildings, and

building segments. The research is focused on

amswering certain structural concerns related to (1)

the adequacy of criteria currently used in design and

zmalysis, and (2) the abiUty of existing nuclear

power plants to withstand earthquakes greater than

considered in the original design.

While the adequacy of nuclear plant

equipment to withstand earthquakes has been a continuing consideration, only the more recent

plants have used a rigorous testing program to qualify their equipment for design level earthquakes.

The resistance to earthquakes larger than design

levels has not been documented. An extensive NRC

and industry effort nearing completion has used

both the collection of earthquake experience data

from non-nuclear facilities and the synthesis of

existing nucleair equipment qucilification test data to form a program to reevaJuate the seismic activity.

The studies performed have shown that, on

a generic basis, much of the equipment used in nuclear plants is inherently seismically-rugged if properly anchored and not subjected to damaging

interactions (e.g., the collapse of un-reinforced

masonry walls onto equipment). In general, it has

been found that the seismic strength evaluation of

electrical equipment is the most critical for

operating nuclear plants.

In recent years, both industry and the NRC

have been concerned with the appropriateness of

NRC's piping design rules for seismic and other

dynamic loads. An over-design of piping for dynamic inertial loadings can produce negative

effects from thermzd expansion £md the relative displacement of piping and points and can reduce

its overall reUabihty.

Both the NRC and the Electric Power

Research Institute (EPRI) have sponsored programs

to evcduate the ec^thquake experience of piping in heavy industrial facilities. The basic conclusion of

these programs is that welded above-ground piping,

designed according to the more relaxed criteria of

the non-nuclear Industry, performs very well during

large earthquakes.

Environmental Protection Agency (EPA) Natural

and Technological Hazards Abatement

The Environmental Protection Agency

(EPA) works to prevent and respond to hazardous

materials releases and related incidents that are caused by natural disasters. Building on EPA's

existing statutory response activities under the National Oil and Hazardous Substances Pollution

Contingency Plan, EPA provides managerial and

technical leadership to State and local goverimients

requesting assistance with hazardous materials

releases, damage to hazardous materials containers,

fires involving hazardous materials, stabiUzing leaks and spills to protect public safety and the

environment, and related incidents caused by

natural disasters. EPA also conducts environmental

research contributing to our understanding of

natural hazards. The Environmental Monitoring

and Assessment Program (EMAP) provides

statistically determined baselines on ecosystem

status from which to measure and understand

impacts of disasters. Research is also conducted to determine the environmental and ecological impacts

of genetically engineered biological agents and to assess the cumulative loss of wetlands on the frequency and severity of floods.

The Role of the U.S. Department of Agriculture

(USDA) in Mitigating Drought, Flood, Fire and

Landslide Impacts

The United States Department of

Agriculture's Office of International Cooperation

460
and Development (USDA/OICD) is coordinating a Famine Mitigation Project for Africa under a Resource Support Service Agreement (RSSA) with

OFDA for an initial two year period. It is widely

acknowledged by emergency experts that much can be done to mitigate the effects of a famine situation prior to its onset. The intent of this activity is to

identify, test under actual field conditions, and

develop standard implementation models for a

limited number of interventions that can

dramatically reduce the impact of famines prior to

their onset, and to reduce the recovery period after the conclusion of a famine. The major areas that

will be investigated under this activity include Early

Warning Systems and Rapid Assessment, Seeds and

Tools, Livestock Preservation, Water Resources, Cash

for Work, and Conflict Modifications. Expert teams

will be formed to design and field test these

interventions in Sudan, Ethiopia, Somalia, Angola,

and MoZfUnbique. Following implementation amd

evaluation of the selected interventions, the teemis

will develop generic guidelines so that these

activities may be implemented wherever they are necessary and appropriate. Finally, workshops will be conducted for famine response personnel in the

United States and in East and Southern Africa to provide an overview of the tested activities and to develop practical plans for their use during future famine situations.

The Early Warning and Rapid Assessment

team will examine current Early Warning Systems

and options for OFDA, develop methods and survey instruments for establishing deviations from norms, and identify indicators for determining appropriate

intervention measures for at-risk populations. The

Seeds and Tools team will identify agricultural

baselines zmd identify appropriate seed varieties, tools and methods that could be employed prior to famine onset. This team will also examine

agronomic and horticultural packages and seed bank

activities as well as traditional crops and practices.

The Livestock team will identify and test activities to maintaiin livestock through famine situations. The Water Resource team will identify interventions

that can be implemented at regional and local levels to maintain a reliable, safe water supply. Although

Food for Work activities have been extensively

employed and examined in the past, the Cash for Work team will identify direct cash transfer opportunities and how they may be employed to support and strengthen the agricultural, livestock and water resources activities to be tested by the

other teams. The Conflict Modifications Team will identify the barriers to working in and applying the above interventions under conflict situations.

Additionally, the USDA, U.S. Forest

Service conducts both domestic and international

drought, flood, fire and landslide impacts programs

which focus on disaster preparedness (primarily

emergency planning, wildfire prevention, and

wildland fuel modification) and wildfire suppression.

The USDA adso conducts research examining

ecosystem processes; people/structure/hazard

interactions; vegetation management and biological

controls: climate/weather/hazard linkages, and

post-hazard recovery.

The Soil Conservation Service of the USDA

aids local communities sponsoring projects to reduce flood and drought hazards; provides

assistance for emergency intervention and recovery;

education and awareness; drought prediction and

warning; community mitigation planning; and

research on the relationships between soil moisture

regimes and agricultural productivity.

The U.S. Forest Service's Disaster

Assistance Support Program (DASP) was

established in 1985 to increase cooperative efforts between OFDA and the U.S. Department of

Agriculture (USDA).

The primary goal of DASP is to assist OFDA in disaster prevention, preparedness, and

emergency response, mainly to developing nations

of Africa, Latin America and the Caribbean, and

Asia and the Pacific regions. The Forest Service is thus able to:

(1) Provide technical support in disaster prevention;

(2) Develop and implement programs

for training disaster management personnel from

OFDA, United States embassies and USAID

missions, and from host countries;

(3) Develop and use a computerized

roster system for identifying technical experts and

disaster management speciaUsts;

(4) Plan and coordinate workshops,

conferences, reports, and publications that promote

effective disaster prevention, preparedness, and

management; and

461
(5) Mobilize disaster management

specialists, equipment, and supplies to disasters. DASP works with OFDA to strengthen the

ability of countries to cope with disasters by helping

them improve their own disaster preparedness and

response capabilities. DASP has provided the

following assistance to countries around the world

through its support to OFDA:

(1) Landslide hazard training and

assessments in Asia, Latin America, and the

Caribbean;

(2) Development and implementation

of disaster management training courses and

simulation exercises;

(3) Sponsorship of workshops and

conferences;

(4) Locust control operations in Africa;

(5) Wildfire assessments and technical

assistance in Latin America;

(6) Operation of water purification

systems at disaster sites;

(7) Production of operations manuals

and guidebooks for disaster management;

(8) Shipment of fire suppression

supplies and equipment; and

(9) Supply of communications

personnel and equipment.

U.S. Public Health Service (PHS), Centers for Disease Control (CDC), Epidemiological

Surveillance and Emergency Management

The Centers for Disease Control,

International Health Program Office in Atlanta,

Georgia provides epidemiological technical

assistance internationally in natural and man-made

emergency situations. The CDC organizes and

maintains a resource center (including a library and

data base of consultants) to support CDC-wide

organizational activities and objectives. CDC assists international organizations and their respective

constituents in preparedness activities leading to more effective provision of assistance in emergency

situations, and provides technical assistance

internationally to U.S. government agencies and

multilateral organizations involved in refugee health

care.

In 1989, eighty-sbc members of CDC staff participated in forty-three emergency response

missions and developed for the World Health

Organization (WHO), a distribution network for rapid assessments for nine types of emergencies. In

1990, CDC, WHO and AID are designing and

implementing a project to develop the surveillance

and response capabilities of African countries lying

within the meningitis belt. CDC has also developed

a briefmg package for use by consultants in emergency and non-emergency situations. This

package includes such information as communications procedures, cultural sensitivity

guidelines, and instructions for shipping supplies

and samples to the U.S. They are also conducting

retrospective evaluations of recent emergency

responses to determine most appropriate and

effective response procedures.

Agency for International Development (A.I.D.),

Office of U.S. Foreign Disaster Assistance (OFDA)

Program in Disaster Prevention, Mitigation and

Preparedness In The Developing World

A.I.D., through the Office of U.S. Foreign

Disaster Assistance (OFDA), is responsible for providing emergency rehef assistance to foreign

nations affected by natural or man-made disasters and for providing other assistance in disaster prevention, mitigation, and preparedness. This

program has been structured to (1) concentrate OFDA relief resources on those activities which

save the greatest number of lives; (2) concentrate OFDA mitigation and preparedness resources on

those activities which save the greatest number of

lives and protect the economic infrastructure of

developing countries; (3) reorient OFDA's approach

to disasters by integrating development concepts

into relief, rehabilitation, prevention, mitigation and

preparedness activities; and (4) encourage other U.S. Government departments, private industry,

international institutions and A.I.D. regional

bureaus to integrate disaster preparedness and rehef

principles into their projects.

The ultimate goal of this program is to save

lives and reduce human suffering in disaster-prone

countries. A.I.D. recognizes that property loss and

subsequent economic and social disruptions

associated with disasters are major deterrents to

462
A.I.D.'s and the developing countries' goals of

fostering economic growth and sustaining the

viabihty of development assistance programs.

OFDA currently works with the Agency's regional

bureaus to reduce the threat posed by natural

hazards to development.

A.I.D.'s strategy addresses the principal

goal of saving the greatest number of lives by (1)

responding with reUef goods and services during

disasters in foreign countries, and (2) preparing for and reducing the likelihood of economic impacts

and loss of life prior to the occurrence of disasters. The two main components of this strategy are as

follows: (1) U.S. Government Emergency ReUef

Coordination. Through the use of all appropriate U.S. Government resources and cooperation with

private sector resources, A.I.D. seeks to alleviate suffering and reduce the number of deaths in the

aftermath of disaster by providing efficient and

rapid responses to requests for emergency relief and

to enhance recovery through rehabilitation

programs, and (2) Prevention, Mitigation, and

Preparedness. A.I.D. seeks to apply

multidisciplinary approaches in the fields of disaster prevention, mitigation and preparedness to reduce

the impacts of disasters.

The program in disaster prevention,

mitigation and preparedness (PMP) stimulates new

and dynzmiic approaches to reduce the impacts of

disasters on potential victims in target countries.

The goal is to save hves. Through pre-disaster

action, OFDA assists vulnerable countries in reducing the economic and social impacts caused by

disasters. The PMP program provides support to appUcants from the public and private sectors,

institutions, imiversities and agencies. Groups

implementing activities are encouraged to provide

matching support to ensure the transfer of findings

cmd results. The following illustrate areas of

concentration which may constitute the multi-year

disaster PMP program: (1) private sector activities;

(2) famine; (3) hydrometeorological events; (4)

geophysical conditions; (5) medical and health concerns; (6) environmental and technological

disasters; (7) A.I.D. mission program support; and

(8) project design, analysis, and evaluation. The

A.I.D. PMP program will support the U.S. and

international goals established for the U.N.

International Decade for Natural Disaster

Reduction.

The program will provide the analytic.

technical, economic and social expertise necessary to support A.I.D. poUcy and program interventions to reduce the impacts of disasters on development.

SustainabiUty wiU be assured by not funding projects

where the host government commitment is inadequate. Activities will contribute to sustaining

economic development by reducing the vulnerabihty

of the work force and economic assets to disasters.

This progrjun supports A.I.D. efforts to provide more effective humanitarian assistance to those who potentially would suffer from natural and

man-made disasters cmd to intervene on their behalf

prior to the occurrence of disaster. The program

will also help to protect private sector investment

and economic development worldwide.

Principal beneficiaries will be the potential

victims of disasters in the developing countries and

officials in public and private sector organizations

who have disaster PMP, reUef, rehabiUtation and

reconstruction responsibilities.

Major outputs of this program are: (1)

hazard-specific technical assistance and training in selected countries based on strategic planning; (2) building host country Institutional capacity to manage disasters and monitor hazards; and (3) reduction of disaster potential through protection of

infrastructure, mitigation of impacts, and helping

host governments plan for, prepare and, where

possible, prevent disasters.

U.S. FEDERAL PROGRAM STRATEGY FORTHE

DECADE OF THE 1990'S

The U.S. Federal Program for reducing the impacts of natural disasters in the decade of the

1990's focuses on reducing fatalities, human

suffering, environmental damage and economic

losses by coordinating programmatic resources to resolve priority hazard-related issues and integrate

and enhance planning and response capabilities

from the Federal to the local levels. Integration of

extensive interdisciplinary expertise, various missions

of the Federal agencies and State and local governments, and existing components of private sector disaster reduction efforts is key to the

successful reduction of hazards during this decade.

The federal program strategy seeks to integrate

mitigation measures across hazards by applying

common tools for pre-disaster interventions,

463
examining causal relationships, geographic and

spatial similarities and other factors leading to successful forecasting and implementation of

appropriate risk reduction measures prior to event occurrence. The strategy for the 1990's will better define acceptable risks by hazard type and location

and will increase public awareness as to those

practical countermeasures which can be taken to reduce personal and community economic, social and physical impacts. The Federal Subcommittee on Natural Disaster Reduction of the Committee on

Earth and Environmental Sciences of The

President's Office of Science and Technology Policy

(OSTP) has recently completed a draft report. This

report will help establish the U.S. national strategy

for disaster reduction in the 1990's. At the same

lime, these activities are complemented by current program planning by the National Research

Council's U.S. National Committee on the Decade.

Their report wdth a recommended plan of action,

will be available in late June, 1991.

Sources

Advisory Committee on the International Decade

for Natural Hazard Reduction, 1987. Confronting Natural Disasters: An

International Decade for Natural

Hazard Reduction. National Academy

Press, Washington, D.C., 60 p.

Advisory Committee on the International Decade

for Natural Hazard Reduction, 1989. Reducing Disasters' Toll: The United

States Decade for Natural Disaster

Reduction. National Academy Press,

Washington, D.C.

Committee on Earth and Environmental

Sciences, 1991. Hazard-Proofing the

Nation: Strategy to Reduce Natural

Disasters (In Draft) Federal Subcommittee on Natural Disaster

Reduction. Washington, D.C.

48 pp.

Federcd Emergency Management Agency, 1987. Plan for Federal Response to a Catastrophic Earthquake. FEMA,

Washington, D.C.

Federal Emergency Management Agency

(FEMA) 1990. Report to the U.S.

Congress - National Earthquake

Hazards Reduction Program:

Fiscal Year 1989 Activities. Washington, D.C. FEMA.

Federal Emergency Management Agency, 1991. , Principal Threats Facing Communities

and hocdl Emergency Management

Coordinator. A Report to the U.S. Senate Committee on Appropriations.

115 pp.

Federal Emergency Management Agency (FEMA)

1991. Report to the U.S. Congress - National Earthquake Hazards Reduction

Program: Fisc£j Year 1990 Activities. Washington, D.C. FEMA.

Hennig, Leaura M. 1990. The Natural Hazards

Data Resources Directory. Natural

Hazards Research and Applications

Information Center and U.S. Dept of

Commerce, National Geophysical Data

Center, Boulder, Colorado. 247 pp.

Krumpe, Paul F. 1990. Office of U.S. Foreign

Disaster Assistance (OFDA) 1990

Program in Disaster Prevention,

Mitigation and Preparedness. A.I.D. Washington, D.C. 9 pp.

National Oceanic and Atmospheric

Administration. 1991. Policy

Statement on the Weather Service/

Private Sector Roles. DOC/NOAA.

Federal Register Vol. 56 No. 13. Washington, D.C.

U.S. Public Health Service. 1987. Disaster

Response Guides. PHS Office of

Emergency Preparedness. Rockville,

Md.

464
Manuscripts Authored for

Panel Meeting but Not

Presented Orally
I.
Boundary Layer Wind Tunnel Study on Full Aeroelastic

Long Span Bridge Model

by

Kolchi YOKOYAMA", Tomoki KANAZAKI^^ and Hasahlko YASUDA^'

Abstract

In order to carry out final

verification of aeroelastic safety of the Akashi Kaikyo(strait) Bridge, which has a main span of 1,990m and two side spans of 960m and a truss stiffening girder, its full aeroelastic bridge model wind tunnel testings were planned and are now in progress. A geometrical model scale of 1/100 was selected and primary dimensions

to be required for the test section of wind

tunnel were determined; width 41m, height

4m and length 30m. This paper presents some discussions related to this large boundary

layer wind tunnel facility and ideas of modeling for the bridge.

Keywords: wind, design, full model,

suspension bridge, wind tunnel

1. Introduction

The Akashi Kaikyo Bridge, which will connect the Honshu(main island) to the Awaji Island, is now under construction. As the bridge is a very long span suspension

bridge whose main span length is far beyond

that of the current world longest

suspension bridge, it is one of the most important themes to secure the wind

resistant safety of the bridge. Therefore,

it is necessary to verify the safety more accurately not only by evaluating from the ordinary wind tunnel tests using a sectional model but also by carrying out wind tunnel tests using an aeroelastic full bridge model whose shape and elasticity

should be as similar as possible. Fig.l

shows the general view of the Akashi Kaikyo

Bridge.

This paper summarizes conclusions of the study concerning the large boundary

layer wind tunnel facility and a full aeroelastic model of the Akashi Kaikyo

Bridge.

2. Purpose of the Full Aeroelastic

Bridge Model Wind Tunnel Test

The purposes of the full aeroelastic

bridge model wind tunnel test are classified as follows;

1) Verification of Aerodynamic

Stability

As the fundamental natural frequency

of long span suspension bridges such as the Akashi Kaikyo Bridge is very low, flutter

characteristics of those bridges are very sensitive to slight change of member

arrangement and to the difference in test conditions. Therefore, it is necessary to verify the aerodynamic safety more precisely by full model wind tunnel tests.

2) Estimation of Wind Load

Main members of the stiffening truss

of the Akashi Kaikyo Bridge are decided by

static wind load. In the static design with

wind load, gust response of drag, lift and moment are considered but these are all estimated by theoretical formulas. If the accuracy of those formulas is verified

quantitatively by means of observing three

dimensional behavior of the full model in a simulated turbulent wind, more rational

design for wind load will be possible.

3) Verification of the Ordinary Wind

Tunnel Testing Method

Not only in Japan but also in other

countries, wind-induced vibration of a long

span bridge has mainly been predicted by

wind tunnel tests in smooth flow using a sectional model. However, in truth, wind is changing in time and in space. The structure also has three dimensional spread

that can not be expressed by the sectional

model. Therefore, it should be quite

important to study three dimensional

effects of wind and the structure by

carrying out large scale full model wind

tunnel test in order to understand accuracy

1) Head, Structure Division, Structure and Bridge Department,

Public Works Research Institute,

Ministry of Construction

2) Senior Researcher, ditto

3) Head, First Design Division.

Design Department,

Honshu-Shlkoku Bridge Authority

467
and limit of the current ordinary wind

tunnel testing method quantitatively.

As described above, the main purposes

of full aeroelastlc bridge model wind

tunnel tests are to verify the safety of the Akashi Kalkyo Bridge to strong wind, to establish rational wind resistant design

method and to improve current wind tunnel

testing method.

3. Construction of the Large

Boundary Layer Wind Tunnel Facility

3.1 Required Characteristics

of the Wind Tunnel

At first, the model scale of 1/100 was decided considering not to fail to simulate

the characteristics of the bridge and to make it as small as possible. Then the width 41m and height 4m of the test section were decided to put whole the model in it. The length of the test section 30m were selected considering to generate boundary

layer turbulent flow. The maximum wind speed 12 m/s was aimed at considering the Froude number

similarity and loss of energy due to generating turbulence.

The properties of smooth flow were aimed at as follows;

a) Turbulence Intensity

; less than 1 %

b) Mean Wind Speed Distribution

- ; less than 3 %

3.2 Pilot Wind Tunnel Test

In order to determine the structural

configuration of the wind tunnel and the building which satisfy the required

properties economically, a pilot wind

tunnel test was carried out. Its scale was

1/16. The pilot wind tunnel is shown in Photo 1. The facts understood by this test were as follows:

1) Specifications of the contraction

nozzle ; a) Contraction ratio 2 can satisfy

the required flow properties.

b) Two steel screens are enough for

the required properties.

c) One of steel screens is to be installed at the entrance

of the bell-mouth.

2) Building : a) The minimum length of 75 d is required.

b) The minimum height of 19.5 m is required.

c) The ceiling can be made 2 m lower

at the side of the lower reach.

3) Others ; a) Corner-vanes at the entrance are

very efficient and those at the exit

are not.

b) The flow is little affected by the Inside trusses supporting the celling of building.

c) The flow Is little muddled by

the streamline-section posts

of corner-vanes.

3.3 Final Plan of the Wind Tunnel

The general layout of the wind tunnel

facility selected as a final plan is shown

in Fig. 2. Improved points after the pilot

wind tunnel test considering the usage are

as follows;

a) Three test sections are prepared:

smooth flow test at upstream side,

turbulent flow test at downstream

side and different wind direction

test at the center.

b) Sensor rooms are arranged at both

sides of the test section.

c) Size and number of fans are selected as 1.8m In diameter

and 36 respectively.

d) Wind speeds are controlled by

rotation cycles of fans.

This wind tunnel facility had been

completed on the end of Feb. 1991 at the site of the Public Works Research Institute

of Ministry of Construction. Flow

properties were measured Just after

completion of the wind tunnel and the results are quite satisfactory.

4. Modeling of Full Aeroelastlc

Long-span Suspension Bridge

4.1 Required Characteristics

of Aeroelastlc Model

In time of carrying out the full model

wind tunnel tests for the Akashi Kalkyo

Bridge to verify aerodynamic stability and

gust response considering three dimensional

effects of wind and the structure, it is considered that the most important subject

Is to realize a good similarity of the model. In modeling of a full suspension

bridge, following attention should be paid

generally;

a) to satisfy the similarity of shape, mass distribution and stiffness

distribution of the structure at the same time

b) to reduce the damping of model

as small as possible

On the other hand, larger scale will cause higher costs of the model as well as the wind tunnel facility. Through a lot of studies on modeling, it was found possible

to make a full aeroelastlc model of the

468
Akashi Kalkyo Bridge with a geometrical

scale of 1/100, having a pretty

satisfactory performance except minor

points in geometrical and mechanical

similarity. For example, mass and stiffness of the girder, which actually are

not uniform, are almost uniformly

distributed to the bridge axis. However, it

is confirmed by soaie analytical studies

that those relaxation of similarity

requirement would give little effects on aerodynamic characteristics of the bridge.

4.2 Modeling of Stiffening Truss Girder

In order to realize three dimensional

aeroelastic behavior of the truss model, the similarity of geometrical shape as well

as stiffness of vertical, lateral and torsional directions Is really Important. From a lot of sectional model wind tunnel studies repeated so far. It is quite clear

that the cross section shape of the stiffening girder causes a lot of effects

on flutter characteristics. Therefore, in modeling of the stiffening truss girder, at first, attention was paid to realizing the similarity of the stiffness without

providing a kind of stiffening spine at the center of cross section. Two types of the truss model were

studied. One is the rigid block model

connected by springs, and the other is the wholly elastic model in which four stiffening bars are used as main cords. As for the latter, it was found difficult to support deck girders separated at each panel In a row. It Is needless to say that misalignment In the deck girders affects

flutter characteristics of the bridge very

much. Consequently, the rigid block model

was adopted. In the model, the truss girder

Is composed of rigid blocks of every four

truss panels to simulate the aerodynamic

characteristics of the actual truss girder, and springs installed between blocks

representing the mechanical

characteristics. In this modeling, it was important and also really hard to design

the spring. Based on the preliminary study

on the characteristics of those springs,

V-shape springs were finally selected. The

outline of the model is shown in Fig. 3 and Photo 2. Actually, the problems were still to be worked out in how to mlnialze effects of springs and clearances between the blocks

on the flutter characteristics, and how to design such rigid blocks as expected. As for the former problem. It Is confirmed

that there are few effects of V-shape

springs by sectional model wind tunnel

test. And for the latter, it is solved by

using light new material of carbon fiber

plate for the truss members.

The detailed modeling of the truss girder is one of reaso..^ / a geometrical

scale was determined as 1/100.

4 . 3 Modeling of Tower

The model of tower is composed of a frame of stiffening bars Installed inside

and exterior wooden forms similar to the geometrical shape of actual structure. According to the analysis on the actual

bridge, the effect of towers on total

behavior in static as well as dynamic

responses can not be ignored. However, it

is really hard to realize all the similar

stiffness only by scaling down actual

structure. So It was preliminarily studied

to check the effect of stiffness of tower

columns, diagonal and lateral members on static deformation as well as oscillation

behavior of the tower, and also the effect

of stiffness of towers on the behavior of girder.From the results of the study, It was found that the stiffness of bridge axial

and torsional directions would be easily

simulated by choosing an appropriate

section of stiffening bars. Simulation of stiffness transversal to the bridge axis

was not easy, but was confirmed by

replacing the effect of actual diagonal

members to those of lateral members by

cutting at their cross points.

However, it was also found that small error in torsional deflection of the girder

would be Inevitable because It is very hard

to simulate the axial stiffness of tower

columns.

Fig. 4 shows a frame of stiffening bars

of the tower model. The level of similarity

was checked by analyzing natural

frequencies and static deflection on the total bridge structure. The results were

almost satisfactory.

4.4 Modeling of Cable

As for the modeling of the cable, the

mass, drag force on it and axial stiffness

should be simulated, especially considering

the effect of difference In Reynolds number

on drag force similarity of circular cable

section. In this sense, it is out of the question to model them by using only one kind of uniform wire. Therefore, the design

was made by using a uniform wire similar

to only axial stiffness and weight pieces

similar to mass and drag force arranged at

a certain interval. The dimension and

arranging pitch of weight pieces were determined after the decision of the mass similarity requirement. The preliminary

study was finally checked by measuring the drag force with wind tunnel test and the

469
damping for the extended cable. Aerodynamic interference between a stiffening girder and cables close to it near the center of main span is another

problem to be considered. A preliminary

wind tunnel test was carried out with a sectional model of two cables and a girder

combined. Judging from the results, weight

pieces, which are similar to only shape and

not to drag force, are to be attached in a certain position near the center so that there is no wrong effect of cables on aerodynamic characteristics of the girder.

Fig. 4 and 5 show the modeling of cables.

4.5 Condition of Connection

Connections of each element such as end links, tower links, cable anchors, tower bases etc. were carefully designed so

as to realize the same functions as those

of the actual bridge and a small structural

damping on the total model system.

The completed full model of the Akashl

Kaikyo Bridge is shown in Photo 3.

5. Concluding Remarks

This paper describes the circumstances

of studying on a large wind tunnel facility

and a full aeroelastic model of the Akashl

Kaikyo Bridge.

Main subjects of this study are summarized as follows;

1) Construction of the large boundary

wind tunnel

2) Modeling of a stiffening truss without

a stiffening spine

3) Consideration on aerodynamic

Interaction between stiffening truss and cables

After finishing the tests for the Akashl Kaikyo Bridge, this facility will

also be used for full model wind tunnel

tests of the Kurushlma Bridges, three

suspension bridges with center spans of

600m, 1,020m and 1,030m of stiffening

boxes, and the Tatara Bridge, the world

longest cable-stayed bridge with a center

span of 890m. Through these tests, it is expected that wind resistant design of long

span bridges will make further progress.

470
471
00561

472
V-shape Spring

Fig. 3 Idea of Stiffening Truss Model
Ik -

Photo 3 Full Model

474
Comparison of Responses of a Select Number of Buildings

to the 10/17/1989 Loma Prieta (California) Earthquake

and Low-Level Amphtude Test Results

by

M. gelebi\ L. T. Phan^ and R. D. MarshaU^

ABSTRACT

This paper summarizes dynamic characteristics of five buildings within the San Francisco

Bay area. The dynamic characteristics are extracted from the October 17, 1989 Loma

Prieta earthquake response records and from

ambient tests conducted in November 1990.

Dynamic characteristics determined for two

of the five buildings prior to the Loma Prieta earthquake are also included for comparison. The results are to be used to draw conclusions related to the validity of small amplitude tests in assessing the dynamic characteristics of buildings and other structures.

The preliminary results show, in some cases, that low-level tests are useful and, in other

cases, they show considerable differences

with those determined from strong-motion

response records.

a select number of buildings in the San Francisco Bay area. The identified characteristics

from the recorded responses of the subject

buildings to the Loma Prieta earthquake of

October 17, 1989 (Ms = 7.1) and those from

recent ambient vibration tests are compared.

None of the buildings studied in this project

suffered visible damage during the Loma Prieta earthquake (LPE). For some buildings,

available results from small amplitude tests performed prior to the LPE are also used

for comparison. The main objective of the

project is to study the validity of small amplitude tests in assessing the dynamic characteristics of buildings and other structures.

2. APPROACH

The steps to be taken during the course of

the project are as follows:

KEYWORDS: vibration; testing; strongmotion; earthquakes; buildings: damping;

frequency; period.

1. INTRODUCTION

This paper provides a summary, progress to

date and preliminary results of a project that

aims to study the dynamic characteristics of

^U.S. Geological Survey, 345 Middlefield Road

Menlo Park, CA 94025 USA

2 Building and Fire Research Laboratory

National Institute of Standards and Technology

Gaithersburg, Maryland 20899 USA

1. Select a feasible number of buildings in the San Fransisco Bay area from

which recorded response data are/will

be available. The buildings selected are representative of different designs and

construction materials and yet form a basis to provide a representative inventory of structures and data, both from

strong-motions such as the LPE, and

small amplitude data.

2. Perform ambient vibration testing. - This step has been carried out using a PC-based data acquisition system. The

475
system accommodates 16 data channels,

has an input range of +/ — 10 volts and

uses a 12-bit A/D converter capable of

performing up to 250,000 conversions

per second. For this study, each channel

was sampled 50 times per second. In

execution of this step, in a majority of

the cases, permanently-installed forcebalance accelerometers were used as the

sensors. This facilitates easy access to

various floors of the building, without

having to provide temporary cables or sensors and makes it possible to directly

compare the results with those from

strong-motion excitation. Nominal ac- celerometer sensitivity is 2.5 volts/g.

3. Analyses of Data. Both the recorded

strong-motion responses and the ambient responses are analyzed to identify

relevant dynamic characteristics and to

provide a basis for comparison of these

characteristics. Whenever applicable,

studies performed by other investigators

will also be used. This step will include

detailed time-history analyses.

4. Derivation of Conclusions and Recommendations: This is the last step and

ultimate goal of the project which seeks

to determine the validity of smallamplitude testing as a means of assess- ing the dynamic characteristics of buildings and other structures.

3. SELECTED BUILDINGS

After considering several candidate structures, five buildings were selected for this

study. Significant characteristics of the se- lected buildings (structural type, foundation

type and height), distance from the Loma

Prieta earthquake epicenter, number of channels of strong-motion sensors within the su- perstructure and the peak accelerations at

the ground (or foundation level) and at roof

level for the selected buildings axe summarized in Table 1. Also indicated in the table

is the agency (USGS or CDMG) responsible for the instrumentation in each building

and the nominal north-south building orientation. The orientation is important in identifying the direction of recorded motions and

resulting evaluations.

4. SUMMARY OF ANALYSES TO DATE

To date, processed strong-motion data is available only for three of the five buildings.

The analyses and investigations performed

on these three structures are summmarized

below. General information is provided in

those cases where detailed records are not

yet available. For some buildings, small

amplitude test data and results were available prior to the LPE and these are summarized for each building in Table 2. Also

summarized in Table 2 are the results of

data-analyses to date.

4.1. Administration Building (CSUH)

The instrumentation scheme of the Administration Building of the California State University, Hayward, is seen in Figure 1. This

CDMG-instrumented building is square in

plan and sits on an IS" thick (45.7 cm) mat

foundation supported by bearing piles. The

13-story building has a steel moment-frame

core and exterior reinforced concrete frame

(Shakal and others, 1989). There were no

small-amplitude tests conducted prior to

LPE.

476
4.1.1. Strong-Motion Data

Copies of the unprocessed strong-motion ac- celeration records obtained from this building during LPE are seen in Figure 2. Detailed analyses of these records will be performed when data is available. At present

we estimate the period to be approximately

1.1-1.2 seconds determined from the average

of several cycles of the top floor records in

Figure 2.

4.1.2. Ambient-Vibration Data

Figure 3 shows time-histories and amplitude

spectra of ambient vibration response of the

roof of the building recorded after the LPE.

The identified frequencies (periods) are 0.93

Hz (1.07 sec) in the N-S direction and 0.86

Hz (1.16 sec) in the EW direction. Note that

the ambient vibration data obtained in this

study is expressed in microvolts rather than

as a percentage of g.

4.2. Santa Clara County Office Building

The instrumentation scheme and general dimensions of this building are seen in Fig- ure 4. The 13-story, moment-resisting steelframe building experienced strong shaking

due to other earthquakes that occurred prior

to the LPE. Records of the response of this

building to the Morgan Hill (1984) earthquake are also available (CDMG, 1985). The

building is founded on a concrete mat foundation (Shakal and others, 1989).

4.2.1. Strong-Motion Data

The unprocessed copies of the strong-motion

acceleration records obtained from this building during LPE are seen in Figure 5. Detailed analyses of these records will be performed when data is available. At present we

estimate the period (average of several cycles

of top floor records in Figure 5) to be approximately 2.4 seconds in both directions.

4.2.2. Ambient-Vibration Data

Figure 6 shows time-histories and amplitude

spectra of ambient vibrations of the roof of

the building recorded after the LPE. The

identified periods (frequencies) are 1.92 sec- onds for both NS and EW directions. Other

identified frequencies are shown in the figure.

4.3. San Bruno Office Building

This 6-story, reinforced-concrete momentresisting framed building is founded on individual spread footings. Overall dimensions

and the general instrumentation scheme of

the building are shown in Figure 7 (Shakal

and others, 1989).

4.3.1. Strong-Motion Data

The processed strong-motion accelerations

and displacements are shown in Figure 8. For the sake of brevity, we show only preliminary results obtained by system identification procedures whereby a model is estimated based on observed input-output data

(Ljung, 1987 and Mathworks, 1988). The

model used in this paper is the ARX model

based on the Least Squares method (for single input-single output) and as seen in Fig- ure 9, provides an excellent match for the

accelerations at the roof (output) with the

ground fioor motion as input. From these

analyses, we extract damping values of 7.2%

477
and 4.1 %, and periods of 0.85 and 1.02 sec- onds for the N-S and Fi-W directions, re- spectively. As seen in Figure 9, the timehistories of the recorded and calculated motions of the roof and corresponding amplitude spectra are in excellent agreement.

4.3.2. Ambient-Vibration Data

Figure 10 shows the time-histories and amplitude spectra of ambient vibration of the

roof of the building recorded after the LPE.

The periods identified from these records are

0.59 (NS) and 0.71 (EW) seconds. Another

significant peak at 1.93 Hz (0.52 seconds) is seen in the EW direction (north end). Damping values are 2.9% (NS) and 2.3% (EW).

4.4 Transamerica Building

4.4.1 The Building and Strong-Motion Data

Figure 1 1 shows the instrumentation scheme

of the Transmerica Building as well as the

orientation (in degrees) of the building. Also

shown in the figure are the processed accelerations and displacements at different levels. Detailed studies of these records were performed by Qelebi and §afak (1991) and

§afak and Qelebi (1991). According to these

studies, the building experiences rocking

motion at approximately 0.5 seconds (2 Hz)

and the fundamental translational period

(frequency) is 3.57 seconds (0.28 Hz).

Figure 12 provides results determined from

applying system identification techniques as described previously (Ljung, 1987 and Mathworks, 1988). The figure shows recorded motions and spectra and calculated motions and

spectra for the 49th floor corresponding to

basemat input motion. In addition to identifying the fundamental frequency at 0.28 Hz,

also identified is the 2 Hz rocking motion

frequency apparent for both the output and

input motions. This rocking frequency also

appears in phase with the spectral analyses

using the 49th level horizontal and basemat

vertical motions.

4.4.2 Pre-LPE Tests and Analyses

Dynamic response characteristics determined

from small amplitude (forced and ambient)

vibration tests and analyses performed on

the Transamerica Building in 1972-73 are summarized by Stephen, Hollings and Bouwkamp (1974) and Kinemetrics (1979). Analysis performed on the structure did not include soil-structure interaction. The fundamental periods (frequencies) determined from

these investigations are 2.94 seconds (0.34

Hz) for both the NS and EW directions. The

modal damping percentages are 0.9 and 1.4

respectively. These characteristics are summarized in Table 2.

4.4.3 Post- LPE Ambient Test

A 40-second window of recorded ambient vibration data of both the NS and EW direction at 48th floor of the Transamerica Building and corresponding amplitude spectra are shown in Figure 13. The data has been fil- tered with a Butterworth filter (order 4, at

0.1 Hz). The frequencies identified from

this analyses are shown in the figure. The

0.34 Hz frequency is identical with the preLPE identified frequencies from tests and

analyses. Also identified is the 2 Hz frequency

but whether it corresponds to the rocking

motion is still under investigation. Damping

values based on combined mode response are 1.5% (NS) and 1.9% (EW).

478
4.5 Pacific Park Plaza Building

4.5.1 The Building and Strong-Motion Data

The instrumentation scheme and recorded

acceleration responses at different levels of

the 30-story three-winged reinforced concrete

moment-resisting frame building as well as the two free-field sites in proximity to the

building are shown in Figure 14. Peak val- ues of accelerations at roof and ground level are also provided in Table 1. System identification techniques were applied and the

recorded and calculated output acceleration

responses at the 30th floor level and cor- responding amplitude spectra are shown in

Figure 15. Also shown are the input ground

level motion records and their amplitude spectra. The identified fundamental frequency

(period) is 0.38 Hz (2.63 seconds) for both

NS and EW directions. However, this mode

is a torsional-translational mode. One of the

striking results of this set of records is that

the ground level and free-field records show

frequencies that are clearly structural; thus,

there is feedback of structural frequencies.

Further detailed analyses of these records

showing this feature are in Qelebi and §afak

(1990) and §afak and Qelebi (1990).

4.5.2. Pre-LPE Tests and Analyses

Non-destructive testing and analyses of the

Pacific Park Plaza Building was performed in

1985 by Stephen, Wilson and Stander (1985).

They determined the first mode frequencies

(in both directions) and the first torsional

frequency to be 0.59 Hz. The difference between this frequency and the 0.38 Hz frequency determined from strong-motion re- sponses can be attributed to several factors

including significant soil-structure interaction

during the strong-motion event which possibly did not occur during small amplitude

tests. Another possibility is the action of

the piles and non-linear behavior of the foundation and sub-strata of the soil during the

strong motion event. Yet another possibility is the micro-cracking of the concrete al- though no visible damage was experienced

by the structure.

4.5.3. Post-LPE Ambient Vibration Tests

In Figure 16, we show the recorded ambient vibration responses at the center core and the wings of the 30th floor. Also shown

are the amplitude spectra with identified frequencies. These spectra show the first mode

(translational-torsional) frequency to be 0.49

Hz. Again, as with the Transamerica Building, we observe that the post-LPE first mode

frequency is closer to the pre- LPE frequency

than it is to the frequency identified from

strong-motions.

5. DISCUSSION AND CONCLUSIONS

The recorded strong-motion and ambientmotion data of the five buildings provides a unique opportunity to study the dynamic

characteristics of a variety of building structures. In addition, it provides an opportunity to assess the validity of the low-level

non-destructive testing procedures of structures—one of the stated objectives of this

project. The preliminary results of the postLPE ambient vibration investigation (this

study) and those available from strongmotion data and pre-LPE tests and analyses

are summarized in Table 2 for easy comparison and are discussed below. The study to

date covers preliminary assessments and has

not advanced to a stage whereby the significant differences between the characteristics

479
determined from low-level amplitude tests and those from strong-motions can be sat- isfactorily explained. Also a clear statement

about the usefulness of low-level tests in as- sessing vibrational characteristics of structures can not be made at this stage.

CSUH Administration Building periods from the strong-motion data and

post-LPE ambient study are similar.

Santa Clara County Office Building

periods from the strong-motion records

and this study differ considerably. The

periods identified from ambient test data

acquired in the two orthogonal directions of the roof are also quite different

raising the question as to why the building was not excited in a strong translational-torsional mode considering its significant ecccentricity.

The San Bruno Office Building re- sults from strong-motion and ambient

data are not similar. Periods in both

directions obtained from strong-motion

data are longer than those obtained from

ambient tests.

The pre- and post-LPE Transamerica Building first mode periods are identical but differ from the period determined using the strong-motion records.

Furthermore, using the strong-motion

records, a rocking mode at 2 Hz (0.5 sec- onds) has been identified. The investigation of the 2 Hz frequency from ambient data is being carried out to see if that is associated with a rocking mode.

As in the case of the Transamerica

Building, there are substantial differ- ences between the periods identified from

pre-and post-LPE low level ampHtude

tests and the strong-motion records of

the Pacific Park Plaza Building.

Final conclusions will be made in the future

regarding the testing techniques employed

in this study and their accuracy in predicting dynamic characteristics of building structures during strong-motion events which is the basis for earthquake resistant design procedures. During the course of this work, we

have found system identification techniques

to be a very useful tool in assessing the re- sponses from both the strong-motion and ambient tests.

6. ACKNOWLEDGMENTS

The authors acknowledge the data provided

by CDMG and USGS, as well as the rel- evant information related to the buildings

tested. Dr. M. Huang and Mr. Robert Land

of CDMG and Mr. Marion Salsman of USGS

provided help in carrying out the testing of

the buildings. Carol Sullivan of USGS typed

the manuscript.

7. REFERENCES

1. gelebi, M., and §afak, E. (1991). "Recorded Seismic Response of Pacific

Park Plaza (Part I-Data and Prehminary Analysis)." Submitted, ASCE

Journal of Structural Division.

2. gelebi, M., and §afak, E. (1990). "Recorded Seismic Response of Transamerica. Building (Part I-Data and

Preliminary Analysis)." Accepted,

ASCE Journal of Structural Division

(August 1991).

480
3. Kinemetrics, Inc. (1979). "Application

Note No 5: Transamerica Building Ambient Vibration Survey." September 1979.

4. Ljung, L. (1987). "System Identification:

Theory for the User." Prentice-Hall.

5. The MathWorks, Inc. (1988). "User's

Guide: System Identification Toolbox

for use with Mathlab." South Natick,

MA.

6. Maley, R. et. al. (1989). "U.S. Gelogical Survey Strong-motion Records from

the Northern California (Loma Prieta)

earthquake of October 17, 1989." U.S.

Geologicad Survey Open-file Report 89-

568, October 1989.

7. Marshall, R. D., Phan, L. T., and Celebi,

M. (1990). "Measurement of Structural

Response Characteristics of Full-Scale

Buildings: Selection of Structures."

U.S. Geological Survey Open-file Report

90-667." March 1991. (Also pubhshed as NISTIR 4511 , February 1991, Gaithersburg, MD.).

8. CDMG (1985). "Processsed Data from

the Strong-Motion Records of the Morgan Hill Earthquake of 24 April, 1984."

Report OSMS 85-05 . California Department of Conservation, Division of Mines

and Geology, Office of Strong Motion

Studies, December 1985.

9. §afak, E., and Qelebi, M. (1991). "Recorded Seismic Response of Pacific Park

Plaza (Part Il-System Identification)."

Submitted, ASCE Journal of Structural

Division.

10. §afak, E., and Qelebi, M. (1990). "Recorded Seismic Response of Transamerica Building (Part Il-System Identification)." Accepted, ASCE Journal of

Structural Division (August 1991).

11. Shakal, A. P., Huang, M. J., Ventura,

C. E., Reichle, M., Ventura, C, Cao, T.,

Sherburne, R. W., Savage, M., Darragh,

R., and Petersen, C. (1989). "CSMIP

Strong-Motion Records from the Santa

Cruz Mountains (Loma Prieta), California Earthquake of October 17, 1989." Report No. OSM 89-06. November 19897"

12. Stephen, R. M., Hollings, J. P.,

and Bouwkamp, J. G. (1974). "Dynamic Behavior of A Multi-Story Pyramid Shaped Building." EERC Report 73-

17. University of California, Berkeley,

September 1974.

13. Stephen, R. M., Wilson, E. L.,

and Stander, N. (1985). "Dynamic

Properties of A Thirty-Story Condominium Tower Building." EERC Report 85-03. University of California,

Berkeley. April 1985.

481
TABLE 1. Characteristics and Loma Prieta Peak Acceleration of Buildings.

Building Na/Nb- D" n" LPE Peak Comments

km Accel, (g)

m Grnd. ILOOI

(1) Administration 13/0 70 16 (NS) O.OY 0.15 Steel moment-frame core. Building (CSUH) 61 (EW) 0.09 0.24 Exterior reinforced concrete Hayward, CA (Vert) 0.05 moment-frame.

(CT>MG) l!rm rrpt p q n p^ r wall q a mnn

n

|VTc_o9no(**») ^Ipva toT c n a if c fn corr»n A fl rtr\r lO 4—.^ TTi ^laH on oraHp

and bearing piles)

19/1 "^5 (NS) 0.11

Office Bldg. 57 (EW) 0.10 0.36 (Concrete mat)

(Vert) 0.10 (CDMG)

NS=337°

(3) Office 6/0 81 13 (NS) 0.14 0.25 Reinforced concrete momentBuilding 24 (EW) 0.12 0.32 resisting frame. San Bruno, CA (Vert) 0.12 (Individual spread footings)

(CDMG)

NS=335°

(4) Transamerica Bldg. 60/3 97 22 (NS) 0.11 0.24 Steel frame. San Francisco, CA 257 (EW) 0.12 0.31 48tli floor is the top used floor (USGS) (Vert) 0.07 (2.75 m thick concrete mat)

NS=351° (No piles)

(5) Pacific Park Plaza 30/1 97 21 (NS) 0.17 0.29 Reinforced concrete momentEmeryville, CA 95 (EW) 0.21 0.38 resisting frame. (USGS) (Vert) '0.06 (1.50 m thick concrete mat

NS=350° on friction piles)

*Note 1: Na = no. floors above ground level, Nb = no. of floors below ground level **Note 2: H= height of the building, D= distance to epicenter, n= number of channels, . ***Note 3: The orientation of the'ntDminal NS designated direction of the building.

482
TABLE 2. Summary of Identified First i^Iode Characteristics

Periods (T in seconds), Frequencies (fin Hz) and Damping(C).

Building Nominal Pre-LPE LPE Post-LPE

Direction Test & Data Ambient

Analyses

s, I/O J

(1) Administrative NS 1.1 1.07 0.7

Building (CSUH) (0.91) (0.93)

EW 1.2 1.16 0.8

(0.82) (0.86)

(2) Santa Clara Cty. NS 2.4 1.92 +

Office Bldg. (0.42) (0.52)

dculL jose,

EW 2.4 1.92 +

(3) Commercial Uthce NS 0.85 7.2 0.59 2.9

JjUllCUIig [i.ii

San Bruno, Ca

EW 1.02 4.1 0.71 2.3

(0.98) (1.40)

(4) Transamerica Bldg. NS 2.94 0.9 3.6

- 2.1 2.9 1.5 San Francisco, CA (0.34) (0.28) (0.34)

EW 2.94 1.4 3.6* 1.6 2.9 1.9 (0.34) (0.28) (0.34)

(5) Pacific Park Plaza NS 1.70** 2.6 2.63 11.6 2.04 0.4

Emeryville, CA (0.59) (0.38) (0.49)

EW 1.70 2.6 2.63 15.5 2.04 1.0 (0.59) (0.38) (0.49)

Torsion 1.70

0.59

3.8 2.63

(0.38)

*A significant rocking mode is identified at 0.5 seconds.

*'This is torsional-translational mode.

"**Identified by spectral analyses.

Unidentifiable due to noisy signal.

483
Hayward - 13-8tory School Office BIdg.

(CSMtP Slatton No. 68364)

16'-e'

1»'-9

SENSOR LOCATIONS

Roof

Pcnihoutt

9

6

7

6

S tti floor

2nd llooc

I*' Moor

Upper Moll lival

Lower Moll l«¥il

Bosemcnl

2nd floor

1st floor

Basement

1 t2'-»'

Roof

51h floor

N,

Note: 1 ff O.3048 m.

Slructaire R«tar«nc« OrianUlkm: Ns 320®

Figure 1. General instrumentation scheme of Administration Building, California State

University (CSUH), Hayward. Figure is adopted from Shakal and others (1989).

484
Hayward - t3-s(ory School Office BIdg.

(CSUIP Slat Ion SB3S4)

r Baseaent: Center - H

Rtcord S83S4-C0ISa-e9291 02

Max. Accel. - 0.08 r

0.21 g

5 »5th Floor; Center - H

Center

10* iMt Floor: 'Eaet Hall - H

I f / " Centsr.-_M ^ . . 0.09

12 Center - E

1 3* Baseaent; East Hall - W *Tracaa are light, but aost can be digitized

14 Basement: Center - E

0,13 g :

1

0.07 g 1

- . . . J

Max. Accel. - 0.09 g

15 Center - Up 0.05 g

16 Center - N

Structure Reference Orientation: N'320*

^to ir" 20 Sec.

Figure 2. Unprocessed copies of the film records of the accelerations recorded at the

CSUH Administration Building (Figure is adopted from Shakal and others,

1989).

485
£3O

U

CSUH--AMBIENT~ROOF--NS

10

f=0.93 Hz

15 20 25 TIME (SEC)

ROOF (NS)

CENTER

EAST END

30

2 2.5 3 FREQ (Hz)

35

4.5

40

O

>

o

u

xlO^

1.5 xlO'

CSUH--AMBIENT--ROOF--EW

10 15 20 25

TIME (SEC)

30

f=0.86 Hz

ROOF (EW)

CENTER

35 40

t igure 3. Ambient vibration response of the roof (NS and EW directions) of CSUH

Administration Building and corresponding Fourier amplitude spectra

486
San Jose - 13~8tory Qovernment Office BIdg.

(CSMIP station No. 57367) SENSOR LOCATIONS

TOP OF PARAPET

ROOF

12TH

7TH

vr 23

2ND

GROUND

srsyuOWER LEVEL

rtMAT FOUNDATION

E-W SECTION

tflcaiioA tor 1.4,1. 12. I$,20

5l«tl Coluflmt

L*C0lior. for (

L*C«ltOit (or -

2, 10, 14, It, 21, 22

l,*c<l>oi> lor 9, 13, 17 Loolion lor i

167'

-L*c<l>n Ui 1,7,II.IS,I9 TYPICAL FLOOR PLAN

4 18 * 17 19*L,

1 2ND FLOOR PLAN

20ooooooo o

22

21

o EXPOSED

° COLUMNS

LOWER LEVEL PLAN

rt46

ROOF PLAN

rt10

S 1 1

12TH FLOOR PLAN

rt

12

-i 14

15 <

7TH FLOOR PLAN

Note: 1 ft 0.3048 m.

Figure 4. General instrumentation scheme of 13-story Santa Clara County Government

Office Building, San Jose. Figure is adopted from Shakal and others (1989).

487
Son Jo>« - l3-«ter|f Covtrimnt Offtc* Bldg.

(CStOr StaUM S73S7) R«»c« ST»7-C01t3-««>*I 0]

Ouunals 1-22. 0-22 ••cooda

I Lowar Laval: HU CoTnar - Dp

SW Comar - Oj

SE Contai - Up

Vi'i.iJ ^"-Wu i'*>ft|»i' » iifWV" wm^i n m^iii

1 2 7th Floor i Cornar - E

13 fth Moor: bH bornar - t

U SW Cornar - H

SE Comtr - II

inimn^.iV.in,

16 2nd Floor: HU Cornar - E

SU Coroar - N

19 SE Ctfrear - M

20 Lowar Laval i MM Corntr_-.E

21 SW Cornar - E

22 SW Cornar - N

Structura Bafarance Orlantatlon: N»337*

0 1 2 3 4 S IS 20 22 Sac.

Figure 5. Unprocessed copies of the film records of the accelerations recorded at the

13-story Santa Claxa County Government Office Building, San Jose. Figure is adopted from Shakal and others, 1989.

488
xlO* SANTA CLARA COUNTY OFHCE BLDG.-AMBIENT-ROOF

5

O

U

SW CORNER (NS)

2

1

0

-1

-2

-3h

NW CXDRNER (EW)

SW CORNER (EW)

10 15 20 25

TIME (SEC)

30 35 40

xlO*

QD

H

(X

2 2.5 3 FREQ (Hz)

Figure 6. Ambient vibration response of the roof (NS and EW directions) of the Santa — Clara County Government Office Building, San Jose, and corresponding Fourier

amplitude spectra.

489
San Bruno - 6-story Office Bldg.

(CSMIP Station No. 584 90) SENSOR LOCATJONS

CENTER MOMENT FRAME

e

• -

1

1 . ^

Penthouse

Roof

6th

4lh

2nd

.Ground Floor

S/N Elevation

200'

Ground Level Plan

Note: 1 ft 0.3048 m.

Nrel

13

-r

Structure Reference

Orientation: Ns 335*

7

Roof Plan

5th Floor Plan

2nd Floor Plan

Figure 7. General instrumentation scheme of the 6-story Office Building in San B

Figure is adopted from Shakal and others (1989).

runo.

490
SAN BRUNO - 6-STORY OmCE BUXJ. - LPE SAJ'J BRUNO - 6-STORY OFHCE BLDG. - LPE

1200r

1000-

800-

600-

MIS 400- U

_1a 200 -

0-

-200-

-400-

1400r

1200-

1000-

800-

!G 600- S

u

400-

-1

to 200-0-

-200-

-400-

-6O0L

CHIO (NS-155) ROOF FL

^AVWyVWVWw-vv^^wVWwvs —

CHll (NS-li5) 5TR FL

CH12 (NS-155) 2ND. FL

CH13 (NS-155) GRND. FL

) 20 30 40 TIME (SEQ

SAN BRUNO - &CTORY OFHCE BLDG. - LPE

30 TIME (SEC)

SAN BRUNO - 6CT0RY OFFICE BLDG. - LPE

-mi

CH3 (EW.065) ROOF

CH5 (EW-065) 5TH FL

CH8 (EW-065 ) 5TH FL (N-END) CH6 (EW-065) 5TH FL (S-END) CHI (VERT.) GRND. FL)

30 TIME (SEC)

30 r

25 -

20-

y 10-

CL 5

- a

0-

-5 -

-10 - -4

CHlO (NS-155) ROOF

CHll (NS-155) 5TH. FL

) 20 30 40 TIME (SEC)

SAN BRUNO - 6-STORY OFHCE BLDG. - LPE

20 30 40 TIME (SEC)

SAN BRUNO - 6-STORY OFHCE BLDG. - LPE

CH3 (EW-065) ROOF

CH5 (EW-065) 5TH FL

CH8 (EW-065) 2ND FL (N-END) ---Al^rAl^^vAl^'^^ 1

30 TIME (SEC)

Figure 8. Processed strong-motion accelerations and displacements at different levels of

the 6-story Office Building in San Bruno (LPE data).

491
u

J

w

u

500

0

-500

-1000

SAN BRUNO - 6-STORY OFHCE BLDG. - LPE

Output

CH2 (EW) ROOF

Recorded and Calculated

Input 5lf r|' M " ' f ^ ' " GRND. FL. CH9 (EW) Recorded

' —

u

•a3

0 10 20

xi04 OUTPIJT-CH2 (EW)

0.98 Hz ROOF —Recorded

and — Calculated

12 3 4

FREQ (Hz)

30

TIME (SEC)

4000

40 50

INPUT~CH9 (EW)

GRND. FL. Recorded

2 3 FREQ (Hz)

60

5^u

u

3

500

0

-500

1000

I

2

1.5

1

0.5

0

SAN BRUNO - 6-STORY OFFICE BLDG. - LPE

Output

Input

0

CHIO (NS) ROOF

Recorded and Calculated

GRND. FL CH13 (NS) Recorded

10 20

xi04 OUTPUT-CHIO (NS)

L17HZ ROOF —Recorded

and

'J

JWj - — Calculated

30

TIME (SEC)

4000

40 50

INPUT-CH13 (NS)

GRND. FL. Recorded

0 2 3 FREQ (Hz)

2 3 FREQ (Hz)

60

Figure 9. Recorded and Calculated (by system identification) Roof Level Accelerations

and Fourier Amplitude Spectra for San Bruno Office Building with Ground

Level Motions as Input.

492
o

u

Q

xl05 SAN BRUNO OFFICE BLDG.--AMBIENT--ROOF--NS

0.5

10 15 20 25 TIME (SEC)

30

1.5 2 2.5 3 FREO (Hz)

3.5

35

4.5

40

1 1 1 f=1.70 Hz

^^^^^^^^^

'

1 . J

Xl05 SAN BRUNO OFFICE BLDG.--AMBIENT-.ROOF--EW

O

>

o

u

o

xlO^

f=1.93 Hz —I

1 1 1 1 ROOF (EW)

CENTER

f=1.4 Hz 11 NORTH END

- .ly U ( A / \j

\

0 0.5 1 1.5 2 2.5 3 3.5 4 4.5 5 FREQ (Hz)

Figure 10. Ambient vibration response of the roof of the San Bruno Office Building (NS

and EW directions) and corresponding Fourier ampHtude spectra.

493
hvaoowsi3S

CI«3

COC!O

a

bO

C!O

CO

«3

Ah

fl3

:3 o

I-* ^

CI o

CO

'c3 <i

S-1

bO

1)>

-(JCI

eti

494
500

^ 0

a

u

-500

SYSTEM IDENnFICATION--TRANSAMERICA BUILDING--LPE

OUTPUT 49TH FL NS(351) Recorded and Calcvilated

INPUT BASEMAT NS(351) Recorded

0 10 20

6000

I

4000

OUTPUT--49TH FL(351)

30

TIME (SEC)

2000

40 50

INPUT-BASEMAT(35 1)

u 1500

-a1 1000

500

0

0 2 3 Freq (Hz)

500

0

SYSTEM IDENTIFICATION--TRANSAMERICA BUILDING--LPE

g

-500

-1000

OUTPUT 49TH FL EW(261) Recorded and Calculated

INPUT V

" BASEMAT EW(261) Recorded

— —— . . .

0 10 20

1.5 xi04 OUTPUT~49TH FL(261)

u

T33 Ih

0.5

0

30

TIME (SEC)

2000

40 50

INPUT~BASEMAT(261)

Recorded

and

Calculated .

2 3

Freq (Hz)

60

I L Recorded

f

60

Figure 12. Recorded and Calculated (by system identification) 49th Floor Accelerations

and Fourier Amplitude Spectra with Basemat Motions as Input

495
§

5

O

OSu

3

10

5

0

2

1

0

-1

xi05 TRANSAMERICA BIJILDING--AMBIENT-48TH FL. (NS)

Recorded Data

MV V V ' ',; V v' *v •^.1 V,'' < V* • l;'

• V,' • -^1/ Filtered Data

10 30

xlO«

15 20 25

TIME (SEC)

TRANSAMERICA BUILDING--AMBIENT--48TH PL. (EW)

35 40

Recorded Data

Filtered Data

10 15 20 25 30 35

TIME (SEC)

xlO^ TRANSAMERICA BUILDING-AMBIENT RESPONSE~48TH FL.

40

0.:54 Hz

0.78

1

1 Hz

1— ' ! ! 1 !

Isolid : NS

0.61 I

I

[z 2.00 H^ i dashed: ;EW

Ms ;p

i

0 0.5 1.5 2 2.5 3 FREQ (Hz)

3.5 4.5

Figure 13. Ambient vibration response of the 48th floor of the Transamerica Building

(NS and EW directions) and corresponding Fourier ampHtude spectra.

496
a o o a

N. O) O

C4 n to C4

N ^

< «

Q..2

O

<«

CO

O w

CO V •—I -u < 'S

CQn^ CO> ^ 0)

^^ '—

<u

^ ^ ^

<U CO

.2

(-1 6 .S

CO -g 5 CO <V) 0 CO _r2 I 2 S

1 'I

.5 a; ^

G 25

(U > O

O .S

1)

bO

(ujz-68) .9 .262

497
SYSTEM IDENTinCATION-PACIFIC PARK PLAZA-LPi. 1

1 1 1 30TH. FL. (260) Recorded and Calculated

GRND. PL. (260) Recorded

15

xi04 OUTPUT-30TH FL(260)

20 25

TIME (SEC)

30 35 40

6000

INPUT-GRND FL(260)

45

5^U

500

0

SYSTEM IDENTIFICATION-PACIFIC PARK PLAZA--LPE

m -500

u

-1000

OUTPUT

1 n 1

1 r T 30TH. FL. (350) Recorded and Calculated

->w ^ —

INPUT , GRND. FL. (350) Recorded

0

8000

10 15

OUTPUT-30TH FL(350)

20 25

TIME (SEC)

6000 r

30 35 40

INPUT-GRND FL(350)

45

Figure 15. Recorded and Calculated (by system identification) 30th Floor Center Core

Accelerations of Pacific Park Plaza Building and corresponding Fourier Amplitude Spectra with Ground Motions as Input

498
PACIFIC PARK PLAZA--AMBIENT-ROOF--WINGS

South Wing (050) Recorded Recorded I I h I . ji

«h

I

0.5 S

uS OH

-0.5

-1.

North Wing (290) lecorded

Filtered

West Wing (350) Recorded

Filtered

10 15 20 25 30 35 TIME (SEC)

PACinC PARK PLAZA--AMBIENT--ROOF--CENTER CORE

40

«

«o

I

°

oS -0.5

-1

(260) Recorded

(350) Recorded

(350)^^^^^ ^^^.^^^ ^.^ ..^Fil,t.^red^^_^ _^_^^ _^ ^^..^

0 10 15 20 25 TIME (SEC)

30 35 40

•n 4

3

I

2

xlO^ PAQFIC PARK PLAZA-AMBIENT TESTING--ROOF

0.49 Hz solid : south wing (050)

dashed : north wing (290)

i dashdot: west wing (350)

L 1.45 Hz 2.73 Hz 3.78 Hz

0 0.5 1 1.5

xlO^

2 2.5 3 3.5 4 4.5 5 FREQ (Hz)

1

1 1

0.49 Hz

1

solid : center core (350)

dashed : center core (260)

0 0.5 1 1.5 2 2.5 3 3.5 4 4.5 5 FREQ (Hz)

Figure 16. Ambient vibration response of the 30th floor center core of Pacific Park Plaza

Building (NS and EW directions) and corresponding Fourier amplitude spectra.

499

Evaluation of Penetration Resistance for Liquifaction

Assessment at Jackson Lake Dam

by

Jeffrey A. Farrar' and Nario Yasuda"

ABSTRACT

The authors performed a statistical analysis on a large

amoimt of Standard Penetration Resistance Test (SPT) data collected as part of dynamic compaction

construction control at Jackson Lake Dam. Jackson

Lake Dam was modified by the Bureau of

Reclamation to increase seismic stability.

Modifications included treatment of the foundation

with dynamic compaction to increase resistance to

liquefaction. Dynamic compaction was accomplished

in two stages. SPT was used to evaluate effectiveness

with over 300 SPT borings performed. Engineers

performed initial evaluations were performed by direct comparisons among adjacent boreholes. In this report

the SPT data are evaluated in statistical fashion to provide further insight into penetration testing and

construction by dynamic compaction. Analysis is presented on depth of treatment, effect of compaction

phase, improvement by soil type, effect of differing

energy in SPT, influence of gravels in evaluation of

SPT, and increase in liquefaction resistance in potentially liquefiable soils. Further analyses should

be performed to evaluate overburden pressure

correction factors, clean sand equivalent adjustments,

and effectiveness of wick drains. KEYWORDS: penetration resistance testing; standard

penetration test; liquefaction resistance; dynamic

compaction.

1. INTRODUCTION

Jackson Lake Dam located in northwestern Wyoming,

was constructed in several stages from 1906 to 1916. The dam is comprised of a short southern

embankment, concrete spillway and outlet works

structure founded on a bedrock ledge, and a northern

embankment approximately 4,300 ft. long founded on

fluviolacustrine sediments. The dam was investigated

starting in 1975 as part of the Bureau of Reclamation's

Safety of Dams program. The north embankment and

foundation were found to be susceptible to earthquake-induced liquefaction. The design

earthquake was of Richter magnitude 7,5 at an

epicentral distance of 4 miles. Reclamation undertook

modifications of the dam from 1986 to 1988 to increase seismic stability. Modifications included

removal of the northern embankment and foundation treatment consisting of deep dynamic compaction and

reinforcement with mixed-in-place soil cement

columns. A detailed overview of the foundation

treatments was reported by Farrar et al., (1990).

The purpose of this report is to present analyses of

penetration resistimce data collected during the

dynamic compaction program. A preliminary report on evaluation of dynamic compaction was presented by von Thun (1988). These preliminary analyses were

based on a direct comparison of neighboring drill holes. In this report, all data were evaluated in a

statistical manner to provide further insight into the

effectiveness of dynamic compaction and the use of the standard penetration test for liquefaction resistance

evaluations.

2. DYNAMIC COMPACTION PROGRAM

Dynamic compaction was used to improve the

foundation beneath the central portions of the dam as

illustrated on figure 1. The concept of block

treatments was applied, using deeper treatments of a honeycombed pattern of interconnecting soil cement

columns under the toe of the dam. This allowed for

* Civil Engineer, Materials Engineering Branch,

Research and Laboratory Services Division, Bureau of

Reclamation, Denver, Colorado, USA, 80225.

** Research Engineer, Filltype Dam Division, Dam

Department, Public Works Research Institute, Ministry

of Construction, Tsukuba, Japan, 305.

501
significant savings in foundation costs.

The dynamic compaction program consisted of

dropping a 30-ton weight from a height of 100 ft. The

drop grid, illustrated in figure 2, had primary impact

points spaced on 40 ft. centers with 20 drops at each

impact point. The treatment was then completed with

15 secondary and 10 tertiary drops with 40 and 20 ft. closure spacings. A final "ironing" phase was apphed

to the complete construction surface using a 30 ton weight dropped from a height of 40 ft. The final ironing phase was only applied to the downstream

portion of the foundation downstream of centerline

since the working pad was excavated upstream of

centerline for embankment placement.

Several design and construction features assisted in the dynamic compaction program. Drainage trenches were excavated around the perimeter of the dynamic

compaction working surface. These trenches were

helpful in lowering the ground water. Typical

horizontal spacing between trenches was 300 ft. Finegrained soil was excavated to a depth of 7 ft. and silty gravel backfill was replaced to provide a stable working surface. This excavation and backfill was

performed in almost all of the treatment areas. Foundation investigations revealed numerous layers of

lake deposits separating cohesionless soils. In areas where lake deposits would delay dissipation of excess pore pressures, wick drains were installed, von Thun

(1988) analyzed wick drains using the direct comparison approach and found they greatly improved

effectiveness of treatment. Approximately 45 percent

of the 750,000-ft^ area of dynamic compaction was

treated with wick drains.

Dynamic compaction treatment areas were divided

into sectors as shown on figure 3. Construction was

performed in two stages as shown on the figure. Since

the construction was performed in two stages, there were significant differences in construction and

monitoring methods to be considered. The Stage I contractor used modified Manitowac cranes with a

single part line to drop the weight. The Stage II contractor used a lattice boom crane specifically

designed for dynamic compaction. The significant

difference among the contractors was the greater

amount of repair time required by the Stage I contractor which influenced continuity of treatment. As will be shown later, a greater level of treatment was realized in Stage II. This increased improvement

may be also due to more extensive use of wick drains

in Stage II.

3. MONITORING PROGRAM

Principal methods for evaluating improvement in dynamic compaction were SPT and piezometers. A

considerable amount of additional testing was

performed to evaluate improvement. Cone

penetration testing was initially performed in Stage I construction but the program was greatly reduced after

it became evident that the presence of gravel layers

slowed testing. Foundation shear wave velocity was measured by both cross hole and surface wave

methods. There was also a study using several inplace

test methods to evaluate the increases in horizontal

stresses from dynamic compaction. Interpretation of

specialized testing was complicated by complex

layering of the foundation. As a result of complex

stratigraphy, simple penetration tests provided the measure of improvement.

Monitoring of Stage I was performed by drilling 168

SPT drill holes before and after compaction. For

Stage II monitoring 153 SPT drill holes were drilled. SPT holes were located in regular patterns of 50- to

80-ft. spacings. Samples were taken at 2.5 ft. intervals to depths of 35- to 50-ft. Drilling was performed by

contractors. SPT testing was accompHshed generally

within 1 to 2 weeks after completion of the dynamic

compaction phase to be studied. Reclamation drilled an additional 25 SPT holes approximately 9 months

after dynamic compaction to evaluate aging effects. These drill holes are not included in statistical evaluations presented in this paper. A typical layout

of SPT drilling performed in sector N is shown on

figure 4.

Different philosophies were applied to the SPT

monitoring program in the two stages of construction.

There was an equal amount of pre- and postcompaction SPT drill holes. In Stage I, SPT's were

performed after specific phases of compaction. At a

single location where soil conditions were critical there were two and sometimes three post-compaction SPT

borings. In other noncritical locations there were no

post-compaction borings. In Stage II, each pre- compaction boring was matched with a postcompaction

boring drilled after the final phase of compaction. As

a result of these differences, the Stage I data are

useful to evaluate improvement by phase of

compaction.

502
There were very significant procedural differences in SPT among Stage I and II contractors. The Stage I SPT was performed using safety hammers by the rope

and cathead (pulley) method. Stage I drill holes were

advanced using tricone rotary rock bits and drilling

mud. A pump bypass line was used to maintain fluid level in Stage I holes during the removal of rockbit. Casing was rarely required in Stage I. The Stage II contractor performed SPT using a CME (Central

Mine Equipment Company) automatic hammer. The

hammer lift is accompUshed with a hydraulically driven

chain and cam mechanism to lift and drop the

hammer. Stage II holes were advanced with the casing

advancer which is a rotary casing advanced by

circulation of drill mud. With the casing advancer

system, the casing was left inplace during SPT

resulting in less disturbance from fluctuating fluid levels in the drill hole. During SPT testing, a 50-blow

limit was typically observed. With the use of the automatic hammer in Stage II, there was a greater

tendency to continue testing in high blow count

material and blow counts in excess of 50 were

obtained.

The most significant difference in between Stage I and

II SPT testing was the energy delivered to the sampler.

Seed et al. (1984) have shown the significance of SPT

with regard to Hquefaction analysis. In an effort to measure energy. Reclamation performed energy

measurements in accordance with ASTM (American

Society for Testing and Materials) test designation D4633 (1986). A Binary Instruments Model 102

calibrator with piezoelectric load cell was used to check SPT energy. The measurements were difficult due to premature integration times believed to be

caused by the large NW drill rods used. Data for the Stage I contractor were initially collected using an

incorrect rod area factor. Only 10 series of tests of

approximately 200 blows of data were collected for both the Stage I and II contractors. The best estimate

of drill rod energy ratio, ER„ was 73 and 97 percent

for Stage I and II contractors, respectively.

The measured energy delivered by the Stage I contractor appears high since he used a jointed

hammer and rod connection with consistently soft worn rope to drop the hammer. Prior to the discovery

of the rod area correction, energy corrections to SPT

N values in this report were performed using

conservative assumptions of 40 and 80 percent ER^ for Stage I and II contractors, respectively. These values

of energy were also used in this report for calculation

of N,6o values. Details of the energy measurements

are being reported in a thesis by Farrar (1991).

For statistical analyses, data were collected and input on a spreadsheet program Lotus 123 TM. Data were

entered at field offices under the direction of project

geologist Karl Wirkus. For each SPT test, particle size analysis (gradation) and moisture content tests were

performed to identify soil type. The soil sample was

classified according to the Unified Soil Classification System (USCS). Recovery, which is expressed as the

ratio sample length to drive length, was also input to evaluate sample quality. An example of the data base

is shown on table 1. Four spreadsheets were

developed, representing pre- and postcompaction data

for Stage I and II construction. Over 4,000 SPT tests make up the database. Analysis and verification of

this data was undertaken by Mr. Yasuda. A more

comprehensive draft report by is available from

Reclamation. This database is available for further study by request.

Caution should be used in grouping of soil types based on gradation analysis from the SPT test. In many

cases there is low recovery of soil, resulting in a soil classification which is not representative of the material tested. Also, the gradation could be a mixture of layers and may not represent the dominant

layer affecting the N value. Occurrence of no recovery may often represent liquefied sands. In many cases

the sample size is below that required by recognized

testing procedures for a representative gradation.

It is important to note that the frequency distribution

analysis represents the number of occurrences less than or equal to the value shown on the graphs. Most

frequency distribution (histogram) graphs plot N

values in increments of five blows. An apparent

arithmatic mean from distribution bars will plot to the

right (larger value) of the true mean value. Therefore,

the true arithmatic mean is generally five blows less than depicted on the graphs.

4. SUMMARY OF IMPROVEMENTS

4.1 Improvement by Soil Type

A summary of improvement in raw N value by soil type is shown on table 1. Several significant findings can be seen in this table. Predominant soil types

present in the foundation include those with USCS

503
groupings of Poorly Graded Gravels (GP), Poorly

Graded Sands (SP-SM), Silty Sands (SM), and Silts (ML). Frequency distribution graphs of the pre- compaction raw N value for these major soil groupings

are shown on Figures 5 through 12. We have found no explanation for the differences in frequency

distribution between Stages I and II. The data for Stage I appear log-normally distributed. This may be

attributed to thicker layers of gravels encountered in Stage I. A summary of gross improvement by soil type is taken

as the percentage of the increase in mean values

before and after compaction. The amount of

improvement for Stage I appears to be less than Stage

II, but the data from Stage I include primary and

secondary phases of compaction. Since the data for Stage I include interphase testing, it does not represent

final improvement. Improvement is significant for all soil types.

The most significant data trends in table 1 are the

differences in mean values from Stage I and 11 construction. The primary difference is due to the energy delivered in SPT by the Stage I and II contractors. The data provide a unique opportunity to investigate the theorem of N value inverse

proportionality to energy delivered. From this data we

can determine if the theorem holds true, regardless of

soil type.

4.2 Energy Effect on SPT Data

To evaluate the inverse proportionality of N value with

energy, the ratio of pre-compaction Stage I to Stage II data is plotted by soil type on figure 13. This graph

shows the ratio of raw N values ranges from 48 to 57

percent for the most predominant soils present. The

ratio did not seem to correlate with particle size or permeability inferred by USCS symbol. The

differences in mean values vary little by soil type.

These findings indicate that the inverse proportionality

holds regardless of soil type. Extreme values outside

this range were checked and found to be caused by

low sample population.

It is assumed that the ratio of precompaction raw N

values represents the ratios of mean energy delivered

by the two driUing contractors. This assumption would

appear valid if geologic conditions remain constant. The ratio of approximately 50 percent does not agree

with the measured energies mentioned in section 3.

Part of the reason for this disagreement may be due to the differences in drill hole advancement techniques.

The ratio does confirm the assumed energies based on

engineering judgement (ER, - 40 and 80 percent). It

is clearly evident that the different energies are

significant to interpretation of ground improvement.

4.3 Improvement by Phase of Dynamic Compaction

Data from Stage I allow the best opportunity for evaluating the improvement by phase of the dynamic

compaction program. Figures 14 through 16 show the mean raw N values tabulated at 5 ft. depth increments

for different phases of compaction. The data show

that while a significant incr.ease in penetration

resistance is obtained from the primary and secondary

impacts, there is still additional benefit from tertiary

and quaternary impacts. Improvement in silts (ML) is more pronounced from later phases of compaction

which is related to lower permeability of the fine grained soil. Effect of compaction phase is also

illustrated in figure 17 where all Stage I data are

evaluated.

4.4 Effective Depth of Improvement

The determination of average raw N values versus depth aids in ready evaluation of effective depth of

treatment. Figures 14 through 16 show that effective depth of treatment is approximately 40 ft. for Stage I. Effective depth of improvement both for Stages I and

II is summarized by averaging all penetration data on

figures 17 and 18. It is not clear from figure 18 if the

depth of treatment for Stage II is greater. SPT testing was generally stopped at 40 ft during Stage II testing.

Examination of figure 19 for Stage II silty sands (SM)

shows some indication that effective depth is also in the vicinity of 40 ft.

Comparison of figures 17 and 18 indicates that there was a significantly larger occurrence of gravels in Stage

I with higher penetration resistance.

5. EVALUATION OF LIQUEFACTION

RESISTANCE

5.1 Computation of Normalized SPT N value

To evaluate liquefaction resistance, we computed the

Ni6o value as proposed by Seed et al. (1984). To

adjust for energy differences between Stages I and II,

504
drill rod energy ratios of 40 percent for Stage I and 80

percent for Stage II were first applied. Future studies

should be performed using different energy ratio assumptions. This initial assumption of energy seemed

to produce a good match of mean N values with each

stage.

SPT N values at depths less than 10 ft were multiplied

by a factor of 0.75 to account for incomplete energy

transmission in short drill rod sections. For

calculation of overburden stress, the ground water was assumed to be at the base (7 ft) of the dynamic

compaction working surface. The wet unit weight of

the silty gravel (GM) working surface backfill was assumed to be 104 lb(/ft^. This assumption is unreliable for the gravels whose wet unit weights can

reach 140 lb(/ft^ and future studies should use higher

weights. The saturated unit weight of the foundation

soils could vary from 115 to 145 lb(/ft^. In this analysis we assumed a saturated unit weight of

foundation soils of 125 lb(/ft^.

Overburden pressure correction of the SPT N value to the 1 ton/ft^ stress level was performed using the

following equations.

and

where;

n = -0.5 a' - effective vertical stress in ton/ft^

Chamber tests on coarse and fine sands have shown

that the factor, C„, varies with particle diameter.

Chamber tests also show that penetration resistance in coarser sands is larger than that for finer sands at an

equivalent density. In this analysis, we only used a constant C^ factor. Future studies could be performed

to evaluate the data with C„ as a function of mean

particle diameter, Djq. This correction factor may not be too significant since a majority of data is taken near

the 1-ton/ft^ stress level which corresponds with a depth range of approximately 20 to 30 ft.

5.2 Improvement by Soil Type

Figures 20 and 21 were developed to reflect the improvement in mean Nj^q for all soil types. It should

be noted that the Nj^o method proposed by Seed et al. (1985) is only applicable to sands. These figures allow

for examination of improvements to all soil types. All

soils exhibit significant increases in penetration

resistance. Data for Stage II gravels show larger

increases than those for Stage I.

5.3 Improvement in Sands

Figures 22 through 25 illustrate the improvement in clean sands (SP, SP-SM) and silty sands (SM). The

frequency distributions are shown before and after compaction for both Stages I and II. All of the diagrams show a reduction in the lower values. For clean sands, there was a reduction in the number

of test occurrences after compaction. There were more low N values in SM sands in Stage II data.

In the method proposed by Seed et al. (1985), the

limiting N^go value for clean sands for magnitude 7.5 earthquakes is approximately 20 to 25. Njgo values

greater than 25 should not exhibit initial liquefaction.

An approximate correction in Ni^o for silty sands (SM)

is approximately 7.5 blows. Figures 22 through 25

show that the soils have practically no liquefaction

potential after compaction. The data also demonstrate

that improvement goals should be specified in a

statistical manner.

5.4 Effects of Gravels

The presence of gravels presents a difficulty in evaluating liquefaction resistance. Figures 22 to 25

show a large number of data with high N,^ values

prior to compaction. If gravels are present in the test interval, the N values of sands may be increased

greatly. At first impression it would seem that the

data would indicate that most soils are resistant to

liquefaction. However, it is well known that coarse

particles present in significant percentages will result

in higher penetration resistance values.

Figures 26 and 27 were developed to evaluate if removing samples recovered with appreciable gravel contents (>5 percent) would result in a more

reasonable distribution of data. These graphs were

developed for Stage II construction. Examination of

these figures reveals that although the number of high

blow count tests occurrences is significantly reduced,

the distribution of high Nj^ values remains. The

figures do show a significant reduction in the occurrence of low Ni^o values for loose sands. Possibly

gravel can also be present in the bottom of the sample

interval and is not recovered.

505
Figures 28 and 29 were developed to illustrate the dependence of Nj^q on mean particle diameter, D50,

Hydrometer tests were not performed on the SPT soil samples so fine-grained soils were not evaluated. The

data serve to illustrate the extreme variability of N

value at any mean particle diameter. This variability

is caused by many factors including geologic variability

and sample recovery. Although there is considerable

scatter, a trend of decrease in Nj^q with decrease in particle diameter, is evident.

Of all the attempts to evaluate liquefiable sands none

were completely successful in reducing the effects of

gravels on penetration resistance. Recently it has been

recognized that given the correct conditions gravel may

be liquefiable. The conventional SPT is not appropriate for evaluation of these materials. One

approach to separating data by geologic classification was presented by Wirkus (Farrar et al., 1990). SPT

soil samples were grouped into three geologic groups:

Ofg (any soil containing more than 5 percent gravel),

Qfc (fine-grained soils according to USCS group

symbol), and Qfs (sandy soils with less than 5 percent

gravels). Figure 30 illustrates improvement by depth

for these three groupings for Stage II construction.

The improvement for sands shows reasonable

conformance with ranges in penetration resistance for liquefiable sands without the influence of gravels.

5.5 Importance of No Recovery Tests.

For evaluation of liquefaction resistance, penetration

resistance tests may be the most important when no

sample recovery was possible . Lack of recovery may

be caused by very loose clean sands which liquefy

during testing. These same sands have the highest

susceptibility to earthquake induced Hquefaction. No

recovery may also be caused by very coarse gravel

obstructing the barrel. To examine this effect, we took

all Nj^ values of less than 50 blows. Figure 31 shows

the occurrence of no recovery N^^ values for Stage II. There is a significant reduction in no recovery after compaction.

5.6 Clean Sand Equivalent Offset

Recently, there has been an attempt to evaluate post

earthquake residual strength from penetration

resistance data (Seed, 1987). A proposed correction

in N160 to account for increased fines content was

proposed. This correction allows for evaluation of an

equivalent clean sand Njgo value for soils containing

fines contents of up to 75 percent.

We investigated this offset in Nj^q values to see if the

offset may be influenced by increased density and

horizontal stresses induced by dynamic compaction.

Figure 32 shows a comparison of the values

recommended by Seed (1987) and the values obtained

from both stages of compaction before and after treatment. First it can be seen that the factors proposed by Seed are extremely conservative and

could possibly result in a preponderance of low

equivalent N values in dirty sands. Part of the

differences between Seed's values and our data maybe

caused by the higher N values resulting from the

gravels. It is interesting to note that at approximately

35 percent fines, our data show an equivalent offset of

7.5 blows, which was proposed by several groups

including Seed for correcting N values for silty sands.

The clean sand equivalent offset increased significantly

after treatment in Stage II. This trend was not apparent in Stage I data but intermediate treatment phase data are included. The results do point to the

dependence of the offset on density and horizontal

stress level. Further study should be performed to analyze the data using overburden pressure corrections

which are dependent on particle size.

6. CONCLUSIONS

Reclamation performed over 4,000 SPT tests both

before and after a dynamic compaction ground

improvement program at Jackson Lake Dam.

Improvements were performed to increase the

resistance of foundation soils to liquefaction. These

data were analyzed to provide additional insight into the use of penetration testing for assessment of

liquefaction potential. Construction was performed in two stages with each stage using differing construction

and penetration testing methods. Important

conclusions are as follows;

- The data serve to illustrate that at any given site there will be extreme variability in penetration

resistance in any given soil type due to natural deposition conditions. This variability may also be due

to particle size effects. After dynamic compaction,

there is also variability. Evaluations of improvement

of improvement should be performed in a statistical manner. Contract specifications for improvement

should also be expressed in a statistical manner.

506
Penetration resistance data were significantly affected by energy delivery. Drilling contractors used

two different drilling and SPT testing methods.

Adjustments to raw N values were required to accurately evaluate liquefaction resistance.

The effective depth of improvement was

approximately 40 ft.

- Penetration resistance was increased by dynamic

compaction for almost all soil types.

- Liquefaction resistance was evaluated by using the

N160 method proposed by Seed et al. (1984). After

dynamic compaction, there were very few occurrences

of N,go values less than 20 to 25 blows/ft.

- The frequent occurrence of gravels had a large

influence on the penetration resistance data. Data for sands can be more accurately evaluated by omitting

SPT samples where coarse particles are encountered.

- The offset in penetration resistance caused by fines content was evaluated. The offset in mean N value was investigated to see if it was a function of

compactive effort or resulting density range of the

soUs. The offsets were much larger than the

conservative values proposed by Seed (1987). Some of

the data indicate this offset was a function of density range.

6.1 Areas of Future Study

The database of SPT data from Jackson Lake Dam is available for additional study. Additional studies

varying the overburden stress correction factor, energy

corrections, and corrections for particle size effects could be performed. Such analyses would provide

further insight into use of penetration resistance for evaluating ground improvement.

of Jackson Lake Dam Foundation," USCOLD News,

March.

3. Seed, H.B., Tokimatsu, K., Harder, L.F., and R. M.

Chung, (1984), "Influence of SPT Procedures in Soil Liquefaction Resistance Evaluations," Report No.

UCB/EERC/84/15, Earthquake Engineering

Research Center, University of California at Berkeley,

October, 50 pp.

4. American Society for Testing and Materials, (1986),

"Standard Test Method for Stress Wave Energy

Measurement for Dynamic Penetrometer Systems,"

Annual Book of Standards. Volume 4.08, ASTM,

Philadelphia.

5. Farrar, JA., (1991), 'Tield Energy Measurements

of Standard Penetration Testing," Thesis, University of

Colorado at Denver, In preparation.

6. Seed, H.B., (1987), "Design Problems in Soil Liquefaction," Journal of Geotechnical Engineering,

ASCE, Vol. 113, No. 8, August

7. REFERENCES

1. Farrar, JA., Wirkus, K.E., McClain, J., (1990),

'foundation Treatment for the Jackson Lake Dam

Modification," Proceedings of the U.S. Committee on

Large Dams, New Orleans, LA., March 14.

2. von Thun, J.L., (1988), 'Treliminary Results of

Dynamic Compaction Stage I and Stage II Treatment

507
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Table 2. Example of SPT Spreadsheet Data - Pre-compaction Stage I

HOLE feECTOR i WICK wia< DATE 1 GROUND ELEV. ELEV. S1AVEL BAND FINES IWATER RECOVER' N D50 SOIL

* 1

' DEPTH 1 ELEV. START END % 1 % % i 1 % BPF mm CLASS

1 AJ 0 0 861 47 67ffl03 6764 33 67S2 83 63 25 7 10 40 20 10O0 GP-QM

1 AJ 0 0 86147 6769.03 5762.03 6760 23 81 17 2 12 X 11 1760 GkV

1 AJ 0 0 86147 6769 03 6759.63 6757 23 0 0 0 0 0 12 ax

1 AJ 0 0 86147 67^ 03 6754 23 6752.73 0 48 52 33 X 14 0 00 SM-ML

1 AJ 0 0 86147 6763 03 6751 93 6750 43 0 48 53 X 60 6 ox SM-ML

1 AJ 0 0 86147 6769.03 6749 23 6747 73 0 X 65 35 87 16 ax ML

1 AJ 0 0 861 47 57ra.03 6746 83 6745 33 0 56 45 31 80 10 oca SM

1 AJ 0 861 47 5769.03 5744.33 5742.83 X 51 19 17 67 22 027 SM

1 AJ 0 0 861 47 57ra03 6741 .53 6740.03 0 19 91 50 0 10 ax a.

1 AJ 0 0 86147 67ra.03 6739.23 6737.73 0 47 53 29 80 12 ox SC-CL

1 AJ 0 0 86147 67ra.03 6736.B3 6736.33 42 49 9 13 60 34 Z96 SP-SM

1 AJ 0 0 861 47 67^.03 6734 33 6732 83 57 X S 10 47 25 6.00 OlV-QM

1 AJ 0 0 861 47 67^.03 6731 93 6730 43 34 60 7 14 33 26 1 42 SP-SM

1 AJ 0 0 861 47 67^03 6729 33 6727 83 13 77 10 20 80 34 026 SP-SM

1 AJ a 0 86147 5769.03 6726 93 672S 43 16 73 12 17 60 24 069 SP-SM

1 AJ 0 0 86147 6769 03 6724.43 5722 93 0 76 24 28 87 31 012 SM

' AJ 0 0 86147 6769 03 5721 93 6720.43 1 SO 49 35 87 21 OX SM-ML

2 AJ 0 0 86141 6771 01 6766.01 6764 51 85 13 2 11 53 38 11 SO GP

2 AJ 0 0 86141 6771 01 6763 71 6762.21 68 27 5 12 40 48 9X GP-GM

2 AJ 0 0 86141 6771 01 6761 21 6759.71 55 39 5 10 47 24 ax QP-GM

2 AJ 0 0 86141 5771 01 5758 71 6757.21 0 0 0 0 0 48 OX

2 AJ 0 0 86141 5771 01 6756.1

1

6754 61 51 25 1 4 13 7 22 600 GM

2 AJ 0 0 86141 6771 01 6753 81 6752.31 53 33 4 9 27 33 784 OP

2 AJ 0 0 86141 5771 01 6751 21 6749 71 0 0 0 0 0 33 OX

2 AJO 0 86141 5771 01 6748 81 6747 31 72 25 3 9 X 22 11 SO GP

2 AJO 0 86141 6771 01 6746 31 5744 81 0 0 0 0 0 11 ax

2 AJO 0 86141 6771 01 6743 81 6742.31 0 0 0 0 0 11 ox

2 AJO 0 86141 5771 01 5741 31 6739.81 0 0 0 0 0 60 ax

2 AJO 0 86141 6771 01 6738.71 6737.21 83 15 2 4 X 42 2270 QKV

2 AJO 0 86141 6771 01 6736 51 673S.01 0 0 0 0 0 37 ax

2 AJO 0 86141 6771 01 6733.71 6732 21 53 41 6 11 53 34 570 ew-QM

2 AJO 0 86141 5771 01 6731 21 6729 71 33 59 8 11 73 34 ax SP-SM

2 AJO 0 86141 6771 01 6728 61 6727.11 11 78 11 19 73 X 049 SP-SM

2 AJ 0 0 86141 5771 01 6726.11 6724.61 7 79 1 4 20 73 21 038 SM

2 AJ 0 0 86141 5771 01 6722.91 6721 .41 23 56 22 20 47 34 018 SM

3 AJ 0 0 86149 5771 ,80 6762.50 6761 00 0 a 0 0 0 23 ax

3 AJ 0 0 86149 6771 80 6756 20 6754 70 0 S9 31 28 67 13 013 SM

3 AJ 0 0 86149 6771 80 5753 70 5752 50 0 61 39 32 87 16 aio SM

3 AJ 0 0 861 49 5771 80 6751 20 6749 70 5 52 13 X 73 9 ox SM

3 AJ 0 0 86149 6771 80 6748 70 6747 20 61 34 5 8 X 20 7X QP-QM

3 AJ 0 0 86149 5771 80 5746 M 6744 80 52 18 X 17 47 9 7X GO

3 AJ 0 0 86149 5771 80 6743 60 5742 10 3 34 63 X 73 3 ox a.

3 AJ 0 0 86149 6771 80 6741 .20 6739 70 45 41 14 11 BO 36 3X GM

3 AJ 0 0 86149 6771 80 5739 10 6737 60 53 42 5 9 73 15 520 GP

3 AJ 0 0 86149 5771 80 6736.10 6734 60 50 43 7 9 60 40 4 75 QlV-QM

3 AJ 0 0 861 49 5771 .80 6733.90 5732 40 66 28 6 7 47 41 ax QP-GM

3 AJ 0 0 86149 6771 80 6731 3D 6729 80 37 56 7 14 67 34 1 89 SP-8M

3 AJ 0 0 861 49 5771 80 6728 80 5727.x 27 54 9 16 60 34 077 SP-SM

3 AJ 0 Q 861 49 5771 80 5726X 6724 SO 4 63 X 27 67 X 013 SM

3 AJ 0 0 86149 5771 .80 6722.40 6720.70 0 75 25 27 60 32 014 SM

4 AJ 0 86140 6768ra 5757.79 5756.29 0 0 0 0 0 11 OX

4 AJ 0 86140 5758 5755 29 6753.79 0 42 58 X SO 8 ax ML

4 AJ 0 86140 5768 6752.79 6751 29 44 50 5 12 67 21 367 SP-SM

4 AJ 0 86140 6768 69 675049 6748 99 36 54 10 13 X 30 1 X SP-SM

4 AJ Q 86140 5768. 6747 79 6746 29 76 22 2 7 13 27 21 £0 ON

4 AJ 0 86140 6768 69 6746 39 6743 89 59 37 4 9 33 28 7X GP

4 AJ 0 86140 6768 6742 89 5741 39 12 65 23 23 67 19 017 SM

4 AJ 0 86140 5768 69 5740.29 6738 79 0 0 0 0 0 24 OX

4 AJ 0 86140 676669 5737.69 6736 1 9 0 68 32 31 87 14 ax SM

4 AJ 0 86140 6758 6735.29 6733.79 53 42 5 11 60 29 5X Grtf 4 AJ 0 86140 6768 69 6732.79 6731.29 55 38 7 10 27 28 5.90 GW-GM

4 AJ 0 86140 5768 69 67X29 6728 79 42 53 5 13 53 25 2.96 SP-SM

4 AJ 0 86140 6768 89 5727 89 5725 89 11 77 12 20 SO 25 020 SP-SM

4 AJ 0 86140 6768.69 6725.19 6723.69 1 77 22 28 87 29 040 SM

4 AJ 0 861 40 5768 6720.69 671 9 1

9

1 83 16 25 BO X 019 SM

5 AJ 0 0 86147 6771 64 6766.74 6765 24 38 33 29 23 SO 25 023 GM

5 AJ 0 0 851 47 6771 64 6754.44 6752.94 68 22 10 14 27 19 1330 QP-QM

5 AJ 0 0 86147 6771 64 6761 84 6760 34 87 12 1 7 20 34 24 60 GP

509
X.

Figure 1. Cross Section of Jackson Lake Dam as Constructed

O • • O • • O

• O • • O • s- O • • O • • G-^

• o • • o •

10.0

Q • • O •

I G-^

TYPICAL DYNAHIC COUPACTION

GRID PATTERN

IMPACT POINTS PHASE

O PRIMARY 2 SECONDARY 3 A TERTIARY 4

• QUATERNARY 5

Figure 2. Dynamic Compaction Drop Pattern (Spacing Shown in feet)

510
•

1 aw* I I 1 suu gr FUT JACKSON LAKE DAM MODIFICATION

STAGE I DYNAMIC COMPACTION

-Llairs of ayroBiG compocrion

worK artas

JACKSON LAKE DAM MODIFICATION

STAGE II DYNAMIC COMPACTION

UfflltB ol 3t«g« II <lyn*m4c como«etlo«

1

./Oris

1 / iicn» L

J

StCTOii —

,

/

TL«"«J I T T

Figure 3. Plan of Stage I and Stage II - Jackson Lake Dam Modification

511
512
PRESTAGE-I GP

60 -1

50 -

40 -

eo N-VALUE (UNCORRECTED)

Figure 5. Histogram of raw N value of GP, precompaction, Stage I

PRESTAGE-II GP

60 -1

O 5 lO 15 20 25 30 35 40 45 50 55 60 N-VALUE (UNCORRECTED)

Figure 6. Histogram of raw N value of GP, precompaction, Stage 11

513
PRESTAGE-I SP-SM

so -1

50 -

40 -

5

0 5 10 15 20 25 30 35 40 45 50 55 €0

N - VALUE ( UNCORRECTED

)

Figure 7. Histogram of raw N value of SP-SM, precompaction, Stage I

PRESTAGE-II SP-SM

60 -1

50 -

0 5 10 15 20 25 30 35 40 45 50 55 €0

H - VALUE (UNCORRECTED)

Figure 8. Histogram of raw N value of SP-SM, precompaction, Stage II

514
PRESTAGE - I SM

50 -

PRESTAGE -II SM

80

O 5 10 15 20 25 30 35 AO 45 50 55 €0

N- VALUE (UNCORRECTED)

Figure 10. Histogram of raw N value of SM, precompaction, Stage II

515 ^
PRESTAGE - I ML

V77\

10 15 20 25 30 35 40 45

M- VX1.0E (ONCORRECTED)

50 55 SO

Figure 11. Histogram of raw N value of ML, precompaction, Stage I

PRESTAGE - I I ML

10 15 20 25 30 35 40 45 50 55 60

H- VALUE (UNCORRECTED)

Figure 12. Histogram of raw N value of ML, precompaction, Stage II

516
RATIOS OF AVERAGE RAW N VALUE

(PRESTAGE- II/PRESTAC3K- I)

I I I I 1 I I I I I I I CL GM GP GP-GM GW GW-GM ML ML-CL SM SM-hCL SP - SM SW-SM KR SOIL TYPB

Figure 13. Ratio of average raw N value of Stage I and II by soil type, precompaction

AVERAGE RAW N VALUE OF GP AT STAGE-I

517
AVERAGE RAW N VALUE OF SM AT STAGE-I

10 -

0 5 10 15 20 25 30 35 40 45 50 DEPTH (FT)

Figure 15. Average raw N value versus depth, SM soil, Stage I

<

>

Pio

<

>

<

70

60

50

40

30 -

20 --

10

AVERAGE RAW N VALUE OF ML AT STAGE-I

ML

PRESTAGE-I

- o P0STSTAGE-I(PHASE2&3)

• - P0STSTAGE-I(PHASE4&5)

-H \ \ \ h10 15 20 25 30 DEPTH (FT)

35 40 45 50

Figure 16. Average raw N value versus depth, ML soil, Stage I

518
AVE . N FOR EACH DEPTH STAGE -

I

\

\

\

/\

/\

\

T

DEPTH (FT) POST- I (2«t3) POST- I (Its)

Figure 17. Average raw N value of all data versus depth, Stage I

519
N- SM- ST- II

z

Figure 19. Average raw N value versus depth, SM soil, Stage II

AVERAGE (Nl)6 0 STAGE -

I

s -/

ffl. CH CL GM QP GP-GM GW GW-GM ML ML-CL SM SM-MI. SP 3P-3M SW 3W-SM NH

|\ \| POST (PHASE 2t3) SOIL TYPE

POST ( PHASE 4 tS

)

Figure 20. Average Ni^o value for soil types, Stage I

520
(Nl)6 0 (STAGE- II)

n

\

\

\

\

\

\

\

\

\

\

\

\

\

\

\

\

\

\

\ /\

/\

/\

n

\

\

\

\

\

\

\

\

CH Cl. GM GP GP-GM GW GW-GM ML ML-CL SM SM-ML SP SP-SMSW-SM NN

SOIL TYPE

Figure 21. Average Nj^q value for soil types, Stage II

COMPARISON OF HISTOGRAMS OF (Nl)60 VALUE. SP, SP-SM

STAGE-I BEFORE AND AFTER DYNAMIC COMPACTION

80

70

60--

>- 50-

o

I

40

o

^ 30

20

10-f

0

SP, SP-SM m PRE COMPACTION Cn POST COMPACTION

U2L

0 5 10 15 20 25 30 35 40 45 50 55 60

(N1)60 VALUE

Figure 22. Distribution of Nj^ values of SP and SP-SM sands, Stage I

521
o

ZDo

COMPARISON OF HISTOGRAMS OF (N1)60 VALUE, SP, SP-SM

STAGE-II BEFORE AND AFTER DYNAMIC COMPACTION

80

70-

60-

50

40 + ^30 + 20

10-f

0

0

SP. SP-SM m PRE COMPACTION (1) CZ3 POST COMPACTION (17)

NUMBER IN PARENTHESES IS THE NUMBER OF (M )60 WHICH EXEEOS 65

N

IZL

\ /

\

14 X/

MM

Fl

_K1 5 20 25 30 35 40 45 50 55 60

(Nl)60 VALUE

Figure 23. Distribution of N^^ values of SP and SP-SM sands, Stage II

80

70

60

> 50

o

I

40

o

^ 30 Li_ 20

10

0

COMPARISON OF HISTOGRAMS OF (Nl)60 VALUE, SM

STAGE-I BEFORE AND AFTER DYNAMIC COMPACTION

N

J2L

1

\

i

\ 7 \

/ \

/ \

/ \

/ \

/ \

/ \

/

/

\

/

\

\

/

/

\

/

/ ^ / \

/ \

/ \

/ \

/ \

/ \

/

/

\

\

m PRE COMPACTION Cn POST COMPACTION (4)

NUMBER IN PARENTHESES IS THE NUMBER OF (M)60

WHICH EXEEDS 65

/

> Fl

F

/ \ VI/

1

/ \ / /

/

/

/ EL

0 5 10 15 20 25 30 35 40 45 50 55 60

(N1 )60 VALUE

Figure 24. Distribution of Nj^^ values of SM sands, Stage I

522
>-o

80

70

60

50--

40-

o

^ 30

20

10

0

COMPARISON OF HISTOGRAMS OF (N1)60 VALUE, SM

STAGE-II BEFORE AND AFTER DYNAMIC COMPACTION

0\

\

\

ED PRE COMPACTION (1) CZD POST COMPACTION (22)

NUMBER IN PARENTHESES IS TXE NUMBER OF (M)60

WHICH EXEED5 65

\

/

/

N

5 10 15 20 25 30 35 40 45 50 55 60

(N1)50 VALUE

Figure 25. Distribution of values of SM sands, Stage II

19 -

18 -

6 -

5 -

SUMMARY OF N16 0 STAGE I

I

PRE AND POST SP & SP-3M. GRAVEL<5%

71

5 10 15 20 35 30 35 40 45 50 >S5 N160 VALUE - BPF

[/ /\ PRESTAGEir |\ \| POSTSTAGEII

Figure 26. Distribution of N^^ values for SP and SF-SM sands with gravel content less than

5 percent, Stage II

523
SUMMARY OF N16 O VALUES

PRE AND POST SM SAND, GRXVi:L<5»

10 15 20 25 30 35 40 45 50 >50 N16 0 VALUE - BPF

PRBSTAGEII |\ \| POSTSTAGEII

Figure 27. Distribution of Nj^q values for SM sands with gravel content less than 5 percent,

Stage II

(Nl)60 VS. D50, SAND

STAGE-I BEFORE DYNAMIC COMPACTION

70

Logarithmic regression

Y=6.67Log(D50)+24.17

A

A

A A

A

A A A A A A A ^ A

A A

* A^

A

*

A

^ * A-^ \^ ^ A, A A^A a'^A^ a Sa

49 aA 'a^ ——A 7 ^

S-*^ A A '^A " i^"^ *

^ .AA Af^'=', ^A.A -A ^&A^

A A

-

A

«=^A A a & A

S^^A A A ^

^ 1— 1

A ^

1 1 hA

-H 1 1 1 1 1

1 1 1 1 1

1 1 1 0.010 0.100 1.000 10.000 D50 (mm)

Figure 28. Nj^o values versus D50 for sands, Pre-compaction, Stage I.

524
70

60 --

50 --

40

30

20

10

0

(Nl)60 VS. D50, SAND

STAGE-II BEFORE DYNAMIC COMPACTION

Logarithmic regression

Y=9.2Log(D50) + 27 58

0.010 0.100 1.000 D50 (mm)10.000

Figure 29. Nj^q values versus Y)^ for sands, Pre-compaction, Stage II.

UJ

_1

<

>

<

LjJ o

<

UJ>

<

60

50

40

30

20

10

SUMMARY OF STAGE 2 IMPROVEMENT BY SOIL TYPE

PRE AND POST DATA WITH THE SAME SYMBOL

A A Qfc - FINE GRAINED SOILS

Qfs - SANDS

O O Qfg - MORE THAN 5%

GRAVELS

10 ; 20 30 40 DEPTH -FT50

Figure 30. Stage II soil improvement by geologic grouping.

525
SUMMARY OF N16 0 VALUES STAGEI

I

PRE AND POST, NO RKCOVKRY. N160<50

AS. /\1 r/\i r/\i r/\i \/ k

5 10 15 20 25 30 35 40

N16 0 VALUE - BPP

\/ /\ PRESTXGBlr |\ \J POSTSTAGEir

4S 50 >S0

Figure 31. Distribution of N^^ values of No Recovery, Stage II.

OFFSET OF (Nl)60 OF SAND

0 -I 1 \ 1 \ 1 \

1 10 20 30 40 50 60 70 80 FINES (%)

Figure 32. Offset of Nj^^q values versus fines content.

526
Torsional Resistance of Multi-Column Foundation

Yukitake SHIOI* ' , Michio OKAHARA* ^ , Jiro FUKUI*^, Syoj i TAKAGI*"

[Abstract]

In this report, torsional resistance

of a multi-column foundation to

eccentric horizontal load is reviewed.

To rsional resistance characteristics

were clarified by means of loading

tests at fields and in laboratories.

[Key words]

Multi-column foundation. Eccentric

load. Torsional resistance. Inclined

pile. Field test. Model test

1 . Introduction

The multi-column foundation method

is economical and easy to execute

when foundations are to be installed

in the sea or in a lake with

considerable water depth. (Fig. 1)

However, when the free length of a pile is long due to the deep water,

the displacement of a slab caused

by horizontal forces is large.

Therefore, it becomes necessary to

install quite a few piles to keep

the displacement within allowable

tolerances. It was previously

reported that the use of inclined

piles in combination with vertical

piles is very effective in solving

this problem, in the report on "Vibration characteristics of multicoliimn foundation with inclined

piles during earthquakes" at the

12th UJNR joint meeting. (Fig. 2)

In that report, the analysis was made on the premise that horizontal

forces work on the center of a slab. However, on actual foundation,

earthquake forces change moment by

moment and the horizontal forces do

not always work on the center of a slab. There are cases where an eccentric load exists as a result

of a ship collision, or where the

center load is changed into an eccentric load due to the ground

slope or lack of uniformity of pile

lengths and or pile spacing. There

are very few examples in which

analysis or experiments were made

on these eccentric loads.

In this paper we introduce equations

for calculating torsional resistance

to an eccentric load and load sharing

for multi-column foundations, results

at a full size model test, and the

results of a comparison between the

small-sized model test values using

acrylic model and the analysis

values.

2. Torsional resistance and load

sharing of multi-column foundations

Fig. 3 shows the two different

relationships between a torsional

moment and the rotating angle of a slab and between a torsional moment

and the rotating angle of the whole

multi-column group on the ground

surface, when the torsional moment

works at a slab of the multi-column

foundation. When the displacement is

small, these relationships are linear.

It is generally considered that

torsional resistance of a multicolumn foundation consists of

two resisting moments. One is the sum of the torsional resisting

moments Mr i of individual piles.

The other is the sum of the values

obtained through multiplying

horizontal resistance at a pile

head Hr i by the distance from the

center of a slab £ i

.

In this case, theoretical solutions

of torsional resistance and horizontal

resistance are given and their

sharing ratios in the torsional moment

are determined.

* * Chief, Design Department, TransTokyo Bay Highway Corporation. ( Fomer

Chief of Structure and Bridge Department, P. W.R. I . , Ministry of construction)

*^ Head, Foundation Engineering

Division, Structure and Bridge

Department, P. W.R. I., Ministry of

Construction

*^ Senior Research Engineer, Foundation Engineering Division, Structure

and Bridge Department , P . W . R . I .

,

Ministry of Construction

* Chief, Investigation Section,

Kagoshima National Highway Construction Office, Ministry of Construction

527
( 1 ) Theoretical solution of torsional

resistance.

Under the conditions that a pile

with a free length h penetrates into

elastic ground and a pile head

receives a torsional moment Mq (Fig.

3 )

, then if the rotating angle of a pile at the depth X is (p x , the

shear ground reaction force, z x. of,

the pile surface can be expressed

by the following equation.

T ^= D/2- </> X -ks (1)

Here,

X X • shear ground reaction force

between a pile and the ground (tf/m^)

D: diameter of a pile (m)

ks : coefficient of the shear

ground reaction force of the

ground ( tf /m^

)

From the balance of the dx section in

Fig. 3,

7C dx

(2)

Here,

Mx: torsional resisting moment

(tf-m) at the depth x. The following equation is obtained

from the relationship between the

rotating angle <p x and strain 0 x •

d 0

dx

6

(3)

GIp

Here,

G:coeff icient of shear elasticity

of a pile ( tf /m^

)

Ip:pole moment of a pile (m")

The following differential equation

can be derived from (equations 1-3).

0

dx' (4)

£ =( TC D3k3/4GIp)'/2 (5)

By solving the equation 4 the

following is obtained:

0 X = 0 oexp( -ex) ( 6 ) Here

,

0 o : rotating angle at the ground

surface

When (6) is differentiated and x equals 0, equation 7 is obtained.

0 o=- e 0 o (7)

Torsional moment Mo at a pile head

is expressed as follows.

Mo= - 0 oGIp (8)

From equation 7 and 8 the rotating

angle 0 o at ground surface can be

obtained as follows.

0 o- GIp (9)

When 1/ £ equals £ ( Refer

to Fig . 4 ) , jg o indicates a depth of

a virtual fixed point to torsion.

From the above, torsional resistance

Mp t of one pile can be expressed

as in ( 10 ) by using the rotating

angle 0

* of a footing.

i = GIp 0 o = GI

GIp

0

*

h o

1 + eh

0

'

(10)

Here

,

h: free length of a pile (m)

Equation 10 indicates that torsional

resistance of a pile becomes smaller

as the free length becomes longer.

Applying these equations to the case

of an inclined pile shows that

component of cosine of an inclined

angle 6 becomes effective. However,

when torsional resistance in the

axial direction of an inclined pile

is taken into consideration, there

is no significant difference when

10 is applied to the resistance of

all piles.

( 2 )Theoretical solution of horizontal

resistance

A horizontal spring constant at the

pile head of a pile with a free

length h, which penetrates elastic

ground is expressed with the following

equation:

12EI <3 "

' "

(1 + dh)' + 2 (11)

;kD/4Ei; 1 /4

Here

,

Kj: spring constant of a pile for

horizontal displacement (tf/m)

EI:flexural rigidity of a pile (tf-m)

h: free length of a pile (m)

k: coefficient of horizontal

subgrade reaction (tf/m^)

D: diameter of a pile (m)

528
When the rotating angle of the

footing is 4> *

, "the horizontal

force Hi of a pile head can be

obtained from the following equation.

Hi =Ki • £ i

• 0 * (12)

Here,

£ i : distance from the rotating

central axis to pile (i)(m)

In the case where there are vertical

piles only, resisting moment by the

horizontal resisting force is

expressed as follows.

S Hi £ i=i: Hi £ i^ 0

• (13)

In the case where there are inclined

piles, since the axial force of a inclined pile resist a torsional

moment, equation 14 applies. In this

case an inclined angle /3 of a slab

to a horizontal surface due to the

torsional moment is neglected.

S Hi £ i=S Hi £ i 2 0

» (cos^ 0 sin2 Qi

+ cos^ Qi ) + £i^(/)* sin^ 0 sin^ a i (14)

Here,

Kv : spring constant at a pile for

axial direction (tf/m)

d : inclined angle of a pile

a i : angle between the inclination

direction of an inclined pile

and the central direction of

a slab (Refer to Fig. 5)

The equations above indicate that

Ki rapidly decreases with an increase in free length h. Thus a multi-coliimn foundation with long

free length is easily affected by a torsional moment. However, when there

are inclined piles, the longer the

free length becomes, the more

effective the inclined piles become

for both torsional moments and

horizontal loads. Among the various

inclined piles, possible piles at

the comer or at the edge are most

effective as equation 14 shows.

Inclined piles in the central part,

however, have the same properties

as those of vertical piles for

resisting a torsional moment.

( 3 ) Sharing ratio

Sharing of resistance to a torsional

moment which works on a multi-column

foundation is determined as follows.

M H= Z Hpi £ i/M (15)

Hu= SMpi /M (16)

Here,

p. H : sharaing ratio by the horizontal

resisting force of piles

/J, M

' sharing ratio by the torsional

resistance of piles

Hp i : horizontal resistance of one

pile

£ i : distance from a slab center

to a pile head

M: torsional moment working on a multi-column foundation

Mp i : torsional resistance of one pile

When the relations Ip=2l and Mp i = X

Mp i /n are substituted into (15) and

(15) after eliminating 0

* from

the relationship among (equalton 10 ~

13), (17) and (18) can be obtained.

These are applicable only to vertical

piles

.

M M = Z Mp i /M

1 6Z ( £ ./h)2 E r 1 , a^h'

1+ + 1 n G ^eh l(l+(3h)^+2

2n- GI

2n-GI + Z £

I h + j K,

(17)

U H =1- Mm (18)

When inclined piles are used, the

equations change to (19) and (20).

y „=2n-GI/[2n-GI+ E-CiMh+l/ £ ){Ki

(cos^ 6 sin^ a i+ cos^ a i )+ Kv sin^ d

sin^ a i }] (19)

M H= 1- W u = Z & iMh+1/ £ ){Ki (cos^

d sin^ a i+ cos^ a i )+ sin^ d sin^ a i }/[2n-Gl+Z -G i^(h+l/ £){K,(cos2 0

sin^ a i+ cos^ a i )+Kv sin^ 0 sin^ a ,}}

(20)

Here,

£ i : distance from a slab center

to a pile (m)

h:free length of a pile (m)

n: number of piles

E: Young's modulus of pile

materials (tf/m^)

G: shear elastic coefficient of

pile materials (tf/m^)

£ : characteristic value of a pile torsion (tf/m)

( Refer to equation 5

)

/3 : characteristic value of a pile for bending (1/m)

/3 =(kD/4EI)i/^

k: horizontal coefficient of

subgrade reaction

D: diameter of a pile(m)

Ki : spring constant in the

perpendicular direction of a pile axis (tf/m)

12 EI/3 ^ Ki = ; (1 + |3 h)^ + 2

529
Kv : spring constant in the pile

axial direction (tf/m)

Equations 17 through 20 show that

when Ki , Kv and values are determined,

the sharing ratio of torsional moment

by the horizontal resisting force and

that by the torsional resistance of

piles can be determined from the

dimensions of a multi-column

foundation.

3. Field Test Results

While a multi-column foundation is

said to be a structural system

easily affected by torsional moment,

it also can be referred to as a ductile structural system because

torsional resistance largely

depends on the horizontal spring

constant or coefficient of shear

subgrade reaction, although the

torsional resistance is affected by

the free length of a pile as analyzed in the earlier section.

Eccentric horizontal load tests

shown in Fig. 6 were conducted

with full- sized foundation models

in order to examine foundation

movement, stress distribution, load

sharing, etc. when torsional force was applied. The maximum load was

set to be 60tf. The loads on the

foundation models Pi and P2 became

a reaction force to each other. The

measurement was conducted mainly at

P2.Fig. 7 shows load- ( rotational

displacement )-time chart resulting

from torsional load at a slab of

P2 . The amount of displacement is

the average of horizontal

displacemants measured by two dial

gauges on the diagonal line at the

model . The two dial gauges showed

almost the same values. Because

horizontal displacement of the

central pile of the foundation was small (maximum 0.2mm), we concluded

that the horizontal load acted

almost purely as a torsional

moment

.

According to Fig. 7, displacement

shows an almost straight line

relationship and residual displacement

is small. Therefore, as long as the

load is less than 60tf, we conducted

that P2 movement is within the range

of elasticity. A load of 60tf was tested on P2 which corresponds to an allowable horizontal bearing

capacity of 69tf. Displacement on

the ground surface was about 5 mm

at the corner pile which is affected

most by torsional moment. However,

it was about 8 mm in an axial

center loading test. We assume that

the difference is caused by the

torsional resistance of the piles

which worked effect ively to

decrease the horizontal load on the

piles.

Next, torsional angle 0

* was determined by dividing the

displacement by the distance from

the slab center to the measuring

point. Also, the rotating angle 0

' was determined by dividing the ground

surface displacement of a pile by

the distance from the slab center

to the pile center. The relationship

between the torsional moment and

the torsional angle <p * and between

the torsional moment and rotating

angle , 0 ' , are shown in Fig . 3 . The

figure indicates that the ratio of

the rotational angle, 0 ' , to the

torsional angle, 0 *, is about 60%.

Therefore, torsional rotation of a slab forces a horizontal displacement

of a pile on the groundsurface to

take place. As a result, horizontal

resistance of the ground acts as a torsional resistance moment.

The relationship between the two

different angles of torsion, 0 *

,

corresponding to torsional moments

of Pi and P2 respectively is shown

in Fig. 8. When a torsional spring

constant is defined as the ratio,

torsional moment to the angle of

torsion, the torsional constant

gradually decreases as the angle of

torsion becomes larger. However,

since its decrease is small, a fixed

number can represent torsional spring

constants for an extensive range of

of an angle of torsion.

The torsional resistance moment

was calculated from the shear stress

of a pile body. The horizontal

resistance moment was calculated

from multiplyingthe pile head

horizontal force H; by the distance

from a slab center, £ i and pile head

horizontal force was calculated

from the bending moment of a pile.

These results for each pile is shown

in Table 1.

Table- 1 shows that the sharing ratio

of torsional resistance of a pile

is about 60 % and it increases

530
slightly with an increase in load.

We believe it increases because

torsional effects gradually reach

the lower parts of the piles and

increase, shear resistance on the

pile surface due to horizontal

resistance in the ground.

The test results above can be

summarized as follows:

i ) Displacement movement caused by

loads up to 60tf, which is close

to the allowable horizontal bearing

capacity, is within the range of

elasticity as seen in Model P2

.

ii ) The ratio of the torsional

resistance of a pile alone to the

torsional moment of the whole model

is about 60%. The torsional resistance

of a pile is effective in decreasing

the amount of displacement. Although

the sharing ratio increases with

an increase in load, the amount is

small

.

iii) Rotation of a slab has a nearly

completely linear relationship with

displacement of the ground surface.

Since its torsional spring constant

does not respond sensitively to an increase in displacement, it is

possible to set a fixed number as a representative constant for a certain range of displacement.

iv) From these test results, it was shown that a multi-column foundation

is a tenacious structural system

with a desirable response to an eccentric load. It also can be

inferred that the larger the diameter

of a pile, the larger its resistance

will be and that the shorter the free

length of a pile, the smaller its

displacement in rotation will be.

4. Small-sized model test

( 1 )Outline of tests

Most of the loading test of multicolumn foundations have been

conducted on models with vertical

piles only. In this study, loading

tests were performed on one model

with vertical piles and on two other

other models with inclined piles in

addition to vertical piles for

comparison.

As shown in Fig. 9, three types of

models of multi-column foundations

were used. Each had 28 piles

consisting of 4 lines and 7 rows

.

Model 1 . a multi-column foundation

composed of vertical piles only.

Model 2. a multi-column foundation

with inclined piles in the bridge

axial direction ( a short side ) only

Model 3. a multi-column foundation

with inclined piles in the

perpendicular direction to the

bridge axis (a long side) only

Acrylic resin was used as a material because it is easy to build

uniform models with it. From the

tensile test results the Young '

s

E=290N/cm^ , and Poisson's ratio

=0.356 were determined. The linear

relationship between stress and

strain was also confirmed.

It was difficult to make the

elastic ground model appropriate for

the acryic pile model. Therefore,

the diameter of the acrylic resin

part corresponding to a pile in the

ground was adjusted such that the

displacement (deflection) and

inclined angle ( angle of deflection

)

were nearly equal to those values

in the full-scale test. Diameter of

a free length pile is 30 mm, but

diameter of a pile in the ground

should be 50 mm based on our

calculations. On the models, the

lower end of a pile in the ground

was fastened

.

Dimension and form were made as small as a 1/100 scale. After careful

study, we found it difficult to

make an exact acrylic model of a typical foundation. Therefore, we

focused on determining structural

characteristics

.

Loading was executed as follows.

(Refer to Fig. 10)

(1) eccentric loading in the bridge

axial direction HI

( ii )eccentric loading in the

perpendicular direction to the

bridge axis H3

Measurement was concentrated on

one rectangular section (1/4 of the

whole area) out of the four

rectangular sections made by axes

of the bridge axial direction and

that of perpendicular direction to

the bridge axis.

(2) Test Results

In the tests, two series loadings

were carried out on three different

types of models. However, in this

report, only the results of loading

test in the X direction are shown as

531
figures- Other test results, at the

standard loading level, are shovm in

tables

.

In order to compare test results

and calculations, the following

calculation methods were adopted.

• The composite structure model

(SHELL)- a structure analysis model

wiht a slab as a shell and a pile

as a solid beam element

•The solid frame model ( 3 FRAME )- a solid frame structure analysis

( three dimension ) model with a slab

as a plane framework and pile as a bar

member connected to a node of frame

Fig. 11 and Fig. 12 show the

horizontal displacement and the

torsional angle caused by a horizontal

load with an eccentricity of 220mm

in the bridge axial direction (

X

direction) of a slab. Fig. 11 shows

displacement on the central axis

( the X axis ) of a slab of three

different models. Of the two methods

tested, SHELL was used as a standard

method. Fig. 12 shows the torsional

angles of slabs of three models in

radians. Horizontal displacement and

the resulting torsional angle are shown in Table. 2when horizontal

eccentric loading as much as 200kgf

acted in the X direction and the Y

direction. If there were two different

values found at H=200kgf, the

average was taken. Section force of

a pile caused by a horizontal

eccentric load of H=200kgf is shown

in Table. 3. Table 4 shows the

estimated section force caused by

the torsional moment. This force is

estimated by subtracting the section

force obtained in the separately

conducted horizontal axis load test

from the section force obtained in

the horizontal eccentric load test.

( 3 ) Review

Fig. II and 12 show that the

horizontal displacement of a slab in

the X direction and the torsional

deformation respectively have a linear relationship with an increase

in load. Scattering of the test

values and analysis values, other

than those of the torsional angle in

the Y direction, was relatively

small; within 20%.

Table 2 shows that horizontal

displacement of a slab center by

eccentric loading has the same

results as axial horizontal loading

(Table 4). Displacement is half as much because the load is half as

much. Torsional moment in the Y

direction is half of that in the X

direction. However, inclined piles

in model 2 resist torsional moment

well, both in the X direction and

the Y direction judging from torsional

angles. This is due to the effect

of the arm length in the direction

from a torsional center to the

piles. Though inclined piles in

model 3 are effective for countering

the effect of eccentric loading in

the Y direction, they are not as effective for loading in the X

direction. The results above

indicate that the eccentric load in

the horizontal direction can be

analyzed as an axial loading in the

horizontal direction anda torsional

moment. Table 3 shows the torsional

effect from a horizontal eccentric

load on a pile section. Though

there are differences between the

test values and analytical values,

there is the same tendency. Torsional

moment in the X direction is twice

as much as that in the Y direction.

Bending moment decreases by 20-30%

when there are inclined piles in

theX direction ( in plane ) . With

inclined piles in the perpendicular

direction to the load ( out plane )

,

a decrease in the moment by about

10% was obtained in our the analysis

The decrease in the bending moment

in the Y direction is not as remarkable as that in the X direction.

However, a decrease of 10-20% was observed in Model 3 . Bending moment

in the direction perpendicular to a load works in the both X and Y

direction, which becomes a resisting

force against torsional moment on a slab. Some test values of axial

force are abnormal. However, the

difference between test values and

analytical values is relatively

small . This indicates that axial

force in the inclined piles becomes

remarkably large.

5 . Summary

The characteristics of torsional

resistance of a multi-column

foundation against eccentric load in

the horizontal direction which is

said to be a weak point of this

532
foundation was clarified through

theoretical reviews and load tests

in the field and the laboratory.

The results of the reviews and the

tests are as follows.

( 1 ) The resisting moment to a torsional moment caused by an eccentric horizontal load can be

treated as the sum of the total

amoiint of torsional resistances

of unit piles and the total amount

of the moment obtained from

multiplying the horizontal

resistance of a pile by the

distance from the slab center.

( 2 ) When the angle of torsion is very

small, the resisting moment and

the angle of torsion have an almost linear relationship and

the angle of torsion of a slab

and the angle of torsion of pile

groups of a multi-column

foundation on the ground surface

have a directly proportional

relationship

.

( 3 ) The theoretical equations for

the two resisting moments were

obtained, which made it possible

to calculate a sharing ratio of

the two different moments. These

equations can be applied to a foundation which has inclined

piles in combination with

vertical piles.

( 4 ) From these equations it was shown

that though a multi-column

foundation is less resistant to

a torsional moment as the free

length becomes longer, the

foundation can be strengthened

by installing inclined piles.

( 5 ) As a result of the field test

( pure torsional loading test )

,

it was found that multi-column

foundations have structural

characteristics of ductile

resistance to a torsion.

( 6 ) The share of torsional resistance

in a pile unit in the torsional

resistance of the whole

foundation is equal to the moment

resulting from multiplying the

horizontal resistance by the arm length and is nearly directly

proportional. This shows that

the larger the pile diameter is

and the shorter the free length

is, the larger the torsional

resistance will be.

(7) If the torsional spring constant

is de fined as the ratio of the

angle of torsion to the torsional

moment, a fixed number can

represent the constants, because

it is not sensitive to

displacement

.

( 8 ) Also in the acrylic model tests

( eccentric horizontal load test )

,

both horizontal displacement

(central axis displacement) and

torsional displacement

( horizontal rotational angle

)

have a linear relationship with

load. These displacement agree

with the analysis values.

( 9 ) Eccentric horizontal load can be

treated by being analyzed as a horizontal force and a torsional

moment. In this case, horizontal

displacement. agrees with the

values obtained in the horizontal

axial loadingtest

.

(10) Inclined piles effectively resist

to rsional moment. Inclined

piles become more effective as the arm length from the central

axis becomes longer.

(11) With vertical piles and inclined

piles, bending moment caused by

an eccentric horizontal load

occurs in two directions; the

loading direction and the

direction perpendicular to it,

which acts as torsional

resistance.

From these results, the mechanism

of resistance of multi-column

foundation to an eccentric

horizontal load was made clear and

calculations can be made for various

designs.

Finally, we would like to express

our profound gratitude to those who

supported and helped us in the

process of this study.

533
"u uau ur u uuiSnu^

Fig. 1 Multi column foundation

II (lon-i)

>i

(ton)

Fig. 2 Behavior of multi column foundation with inclined piles

a • o o

• It level of slib

at livel of fround surfice

lorilonal inile of <lab lorjionil angle at fround surface ^' dlO'*)

i

: (rad)

I I ) displaceient al (round surface S ()

Fig. 3 Torsional load and deformation Fig. 4 Deformation of pile by torsion

0 o • o o • O Q • • o-er o o • • e-einclined pile

direction of

inclination

Fig. 5 Direction of inclined pile and angle of pile position

534
^ I Location of eccentric loading

Location of eccentric loading

Number: Pile No. O : Uni-aJilal gauge 90 piece

: Trl-axial gauge 13 piece

s

Oi

O o' O; S.

1

Fig. 6 Location of Gauge

i

Fig.

7 Eccentric Horizontal Load Test (Load-Displacement-Time chart)

535
H

^ 3'

o PI Slab

« P2 slab

• 0

0 1

Slab torsional angle

Fig. 8 Slab Torsional Spring Constant-Torsional Angle

Model 1 Model 2 Model 3

Fig. 9 Side view of Models

HI: Horiiontol axis In the X direction (loading in two direc

tions o£ X direction)

H3: Horizontal axis in tha V direction (loading in twj direc

tions of Y direction)

Fig. 10 Loading Location

536
* HODF.L 1 X Horizontal eccentricity

to : 0

.

(0 o

•p =. C 3 0 - N

-AU o 0 ° X 2

/

A•

/

/ - —

SHELL 3FRRHE

meter.

2

0.00 i.»o . Horizontal disolacement

U) Horizontal displacement

» MOUEL X Horizontal eccentricity

X)

ffl 2 0

0 s N

•H

1^ < 0

' X

:

X-OIRECTION Legend

SHELL 3FRRHE

Displacement

meter.

0.00 0. 10 I. IS MO ].I9 Horizontal disolacement

(e) Horizontal displacement

* MODEL 3 X Horizontal eccentricity

(0 P 2 C

0

N

•Hu

0 2 X

X-DIRECTION /

A

/

/

Legend

SHELL

SFRflHE

Displacement

meter

.

2

"O.OO I.IQ J. IS «.!0 Horizontal displacement

Horizontal displacemnt

(0 -

> -

fl0 S

-H ^

4J><

r-lfl o

Q)3H S 5,

* MODEL 1 X Horizontal eccentricity

X-DIRECTION

w

0) _

E-< 0.00

Legend

SHELL 3FRRHE

Displacement

meter.

2

3

> -

0.00 ICO. 00 7(0.00 120.00 Horizontal load

^''^ Horizontal displacement

* MODEL 2 X Horizontal eccentricity

X-DIRECT ION

fl" 2

•H . 4-1 -

ra %\\

Q)3

fl>

°

W o

Q) °. ^ °0.00

Legend

SHELL 3FRnHE

Displacement

meter.

L

no. 00 no. 00 (10. 00 KO.OO Horizontal load

(<i) Horizontal displacement

* MODEL 3 X Horizontal eccentricity

0)

~\ 0 X- DIRECT 1 ON H •

fl "

>

- ^ a

fl0 -

-1-

* •M

I-I 0

fl ~

^ 0 3 2-

-1 0

fl>

- P S W 0

ni 0 00 10 00 lie . 00 :(( .00 no Horizontal load

Legend

SHELL

3FRfiHE

Displacement

meter.

2

("Horizontal displacemnt

Fig. 11 Horizontal Displacement of a Slab by a Horizontal

Eccentric Load (In the X direction)

537
* MODEL 1 X Horizontal eccentricity

10 +jco

NH

U

0X

a

0 R : X ^1.

O

o - —

O —1»- //

o fhr V o /

s7

0.90 9.31

Legend

SHELL

SFRflHE

Displacement

meter.

o 2

1 9.11 a.}< 0 Torsional angle

<•) Deformation

* MODEL 2 Horizontal eccentricity

o

a

o*

(0 o 0H

(0 o C

0

N

•H0

a:

/ -

/

/

70.00 o.oa

Legend

SHELL

3FRfiHE

Displacement

meter.

D 2

1 O.ll O.JI O.IJ Torsional angle

(e) Deformation

MODEL 3 X Horizontal eccentricity

(00

DIR:X':|.T-2

' t /

V

r f1

Legend

SHELL 3FRnHE

Displacement

meter,

o 2

1 0.00 e.oa o.ii o.]« o.i} Torsional angle

(e) Deformation

> -

* MODEL 1 X Horizontal eccentricity

OIR- ' -^l , T»2

roU 2- •H • P

~

>1 •

0)3

(0

:

> . H

«3U , •H ' 4J • >i H

(D

(U3

-I t ID

• > '

Legend

SMELL 3FRnHE

Displacement

meter

o 2

• I 0.00 eo.oo ISO. 00 }io.oo j;o.oo Horizontal load

(b) Deformation

* MODEL 2 X Horizontal eccentricity

OIR: x=i , T=2 Legend

SHELL

- - 3FRflME

Displacement

meter,

o 2

I O.DO ISO. 00 110.00 tlO.OO KO.OO Horizontal load

(d) Deformation

MODEL 3 X Horizontal eccentricity

(0u s

l-l •

4-1 "

>i • H

I:

Q)3

.-I o

(0 ". >

°

V)

il E" 0.00

0IR:xh.t = 2 Legend

SHELL 3FRnHE

Displacement

meter.

a 2

• I 0.00 ito.oo :io.oo sio.oo. Horizontal load

(') Deformation

Fig. 12 Torsional Deformation of a slab by a Horizontal

Eccentric Load (In the X direction)

538
Table 1. Sharing Ratio of Resisting Moments

Load

(t) Torsional

moment

load

(t.ra)

Torsional resisting moment of a pile

( Z M, , ) Horizontal resisting moment of piles (H,,- S , )

* Resisting moment per one pile

(t-m)

Torsional resisting

moment of

of all piles

(t-m)

Sharing

ratio u a (%)

"Horizontal moment of each pile (t-m) Horizontal resisting

moment of all piles

(t-m)

Sharing

ratio

1 2 3 4 5 6 7 8 9 U u ( t )

20 39.5 2.554 22.99 61.5 2.09 1.32 2.87 0.91 0 1.19 1.79 1.32 2.93 14.42 38.5

40 79 5.552 49.97 62.6 4.99 2.73 4.95 2.18 0 2.23 4.47 2.73 5.33 29.61 37.2

60 118.5 8.282 75.54 62.6 7.42 4.09 7.82 3.15 0 3.26 6.78 4.09 7.96 44 . 57 37.4

Remarks) * Average values of piles No. 2, 5, 7, 8.

9

** Valuea of pllea No. 2, 8 are the average value obtained from other veluea

Table 2. Deflection of Slab by Horizontal Eccentric Load H=200kgf

Direction Location kind Model 1 Model 2 Model 3

Horizontal

displaceinent

of axis (nun)

X direction

Eccentricity

220 mm

Test , 3.2 1.4 3.2

Analysis 2.90 1.49 3.00

Y direction

Eccentricity

110 ram

Test 3.2 3.3 1.8

Analysis 2.86 2.95 1.58

Angle of torsion

X 10-' (rad)

X direction

Eccentricity

220 mm

Test 1.60 0.90 1 .25

Analysis 1.66 0.91 1 .38

Y direction

Eccentricity

110 mm

Test 0.65 0.50 0.50

Analysis 0.83 0.45 0.69

Remark : Analytical values are the SHELL values.

539
Table 3. Section Force by Horizontal Eccentric Load H=200kgf

Direc

tion

Locat

ion Pile Kind Model 1 Model 7. Model 3

Bending

moment

(kef -era)

X

Eccentri

ci ty

220tnni

AA

My

Test -185 -128 -171

Analysis -191 -138 -155

Mz*

Test -19 -26 -13

Analysis -18 -8 -17

D

My

Test -159 -141 -173

Analysis -189 -133 -161

Mz*

Test -48 -49 -57

Analysis -55 -31 -50

Y Ef~T"^n f- r- i city

llOrani

AA

Mz

Test -89 -95 -70

Analysis -90 -87 -67

MyTest -34 -26 -32

Analysis -54 -39 -42

B

Mz

Test -98 -113 -91

Analysis -109 -95 -84

MyTest -34 -39 -28

Analysis -53 -39 -41

Axial force

(kgf)

X

Eccentri

City

220mra

B

1 est -36 -79

Analysis -33,3 -104.6 -0.8

F

Test -11 -37 -71

Analysis -27.9 -49.9 -58.

1

Y

Eccentri

city

llOniin <^

Test <24> <31>

17

Analysis 9.6 <15.7>

-21.7

<37.0>

12.4

D

Test 69 Analysis -1.4 -17.3 73.9

Remark) Analytical values are SHELL values. Bending moment with * is the moment in the perpcndic

ular direction to the load.

Table 4. Horizontal Displacement at Slab by H=400kgf

( In mm ) Direction Location kind Model 1 Model 2 Model 3

X direction Eccentricity

Test 6.2 3.0 6.2 Analysis 5.81 2.99 6.02

Y direciton Eccentricity

Test 6.8 6.8 3.5 Analysis 5.72 5.90 3.33

Remark: Analytical values are SHELL values. 540
Nondestructive Evaluation of Civil Structures and Materials

KEN P. CHONG* and JOHN B. SCALZI

PROGRAM DIRECTORS

National Science Foundation

1800 G Street, N.W.

Room, 1108

Washington, D.C. 20550 USA

ABSTRACT

Due to environmental, man-made

and natural hazards,

infrastructure of public works

in the world is deteriorating.

Nondestructive evaluation (NDE)

is essential for condition

assessment for repair,

retrofit, rehabilitation, and

replacement (if needed) to

insure the public safety. In

this paper an overview of

nondestructive evaluation

projects, including a new

initiative and four new awards

on Quantitative NDE for Large

Structural Systems, is

presented.

KEYWORDS: Non-destructive

evaluation; structures,

condition assessment

INTRODUCTION

The world invests about US

$1,430 billion annually in the

Note: Part of this paper is

taken from "Overview of

Nondestructive Evaluation

Projects and Initiative at "NSF

by K. P. Chong, J. B. Scalzi

and O. W. Dillon in Intelligent

Structures . Elsevier, London,

1990.

*to present paper.

construction of structures,

including: buildings

(residential, commercial,

industrial, etc.), civil works,

and utilities, such as

highways, water and sewer

structures , railroads, and

transmission lines.

Construction is the largest

industry in the world amounting

to 10% of the world's gross

domestic product.

Structures are generally the most expensive

investments/assets in any

country. In addition

structures have long service

life compared with any other

kinds of commercial product,

and are rarely replaceable once

they are erected. Yet the

feedback and controls

on the "state of health" of

structures are practically nonexistent (compared with say the

much less costly automobiles)

.

Nondestructive evaluation is an

essential part of this feedback

and monitoring system for

infrastructures

.

During the past several years,

many reports have been

published concerning the sad

state of deterioration of the

nation's public works, such as

bridges, roadways, water and

sewer systems, ports,

harbors, airports, and

buildings of all types.

According to the 1988 National

541
Research Council Report on

"Building for Tomorrow"

estimates of public

infrastructure amounted to $4 09

trillion in

1984, and growing rapidly. The

U.S. Interstate System, built

in the 1950 's needs repair, so

do most of the state and county

secondary roads. Some 42% of

all U.S. highway bridges are

inadequate and half of school

buildings are 50 years or

older. In the U. S. alone,

recent bridge collapses

include:

o Mianus River Bridge

Connecticut, 1986, 3 deaths

o Schoharie Creek - New York,

1987, 10 deaths

o Hatchie River - Tennessee,

1989, 8 deaths

o Miamis Town River - Ohio,

1989, 2 deaths

According to David Aschauer

(Chief Economist of U.S.

Federal Reserve Bank

Chicago) , a 1% increase in

public capital causes 1/3% rise

in productivity. From 1971 to

1985 U.S. productivity growth

fell 1.2%, out of which 1% was

attributed to neglect of the

infrastructure. If the present

infrastructure deterioration

continues, the Department of

Transportation estimates that

the U.S. economy will miss out

on:

o 3.2% growth in GNP

o 5.9% disposable income

o 2.2% employment

o 2.7% productivity

The infrastructure ages and

deteriorates with time. The

deterioration is mostly a

result of aging of the

materials, excessive use,

overloading, climatic

conditions, lack of sufficient

maintenance, and difficulties

encountered in proper

inspection methods. All of

these factors contribute to the

obsolescence of the structural

system as a whole. As a

result, repair, retrofit,

rehabilitation, and replacement

become necessary to insure the

safety of the public.

THE PROBLEM

In order to evaluate the safety

of a structure, it is essential

to determine the remaining

strength of the total system by

a performance analysis based

upon the strength and

inspection data of the

materials and the structural

system.

A visual inspection process

will provide data with respect

to the external physical

condition of the material such

as the degree of corrosion and

general deterioration. By past

experience and judgement an

engineer will evaluate the

remaining life of the material

and the contribution to the

safety of the structure.

Strength determinations may be

made by sampling the material

and testing it in a laboratory.

However, this method destroys

some of the structure which

subsequently must then be

repaired. The repair is

usually noticeable and

disruptive, as a result, many

engineers would prefer a

nondestructive technique.

542
THE SOLUTION agree

.

Although there are many

techniques for nondestructive

evaluation (NDE) currentlyavailable for metals which are

used in defense and

manufactured products, there

are only a few which are

available for the construction

materials such as wood,

concrete, masonry, and steel.

These current methods for

construction materials are not

very reliable and in some

instances they can only be used

on a comparative qualitative

basis. Instruments

specifically designed for

construction materials and with

greater accuracy than those

currently available are

required in order to

quantitatively evaluate the

safety of the structure and to

estimate its remaining life

adequately.

NSF ACTIVITY IN NDE

The NSF Division of Mechanical

and Structural Systems is

encouraging basic research to

investigate the requirements to

develop suitable instruments

for nondestructive evaluation

of strength properties and

other physical conditions for

construction materials. The

Program Directors in the

Division are encouraging m.ulti—

disciplinary activities

involving electrical,

mechanical, chemical, computer,

electronic and structural

engineers, as suitable for the

project to perform group or

team efforts in the research

projects. Unfortunately, this

"encouragement" should not be

interpreted as large sums of

new money for this approach.

Rather this is our personal

assessment with which all of us

NSF is also encouraging close

cooperation and active

participation by the industries

involved in the manufacture of

nondestructive equipment and/or

instrumentation. By joint

efforts viable products can

result which will be of great

help to engineers responsible

for the safety and longevity

evaluation of structural

systems

.

As a source of information on

the type of questions which the

engineering community would

like to have answered by the

use of nondestructive

evaluation techniques, NSF

supported two workshops to

outline specific needs of the

profession. One was held in

February, 1988, at the

University of Southern

California, Los Angeles, and

the Proceedings are titled,

"Nondestructive Evaluation for

Performance of Civil

Structures." The contact

person is Dr. Sami F. Masri.

The other workshop was held in

June, 1989, at the National

Institute of Standards and

Technology, at Gaithersburg,

Maryland, titled International

Workshop on Sensors and

Measurement Techniques for

Assessing Structural

Performance, " organized by Dr.

Richard D. Marshall and his

associates.

CENTERS AND OTHER NDE ACTIVITIES

In addition to NDE research

done at MIT, Illinois, Cornell,

West Virginia, Virginia Tech,

New Jersey Institute of

Technology, Georgia Tech, etc.,

the following centers

concentrate on various aspects

of NDE.

543
I. Johns Hopkins Center for

NDE (R. Green)

o Heavy on experiment

o Advanced sensors

o Artificial intelligence

o Life cycle management - include fracture, fatigue

o Measurement science for

microelectronics

o Process control

The closest project to large

structural system is the

infrared television for wall

thinning of buried pipelines.

II. Iowa State Center for NDE

(D. Thompson)

o NSF Industry/University CRC

o Modeling efforts emphasized

o Advanced ultrasonic NDE

o Advanced electromagnetic NDE

G X^Ray « thermal NDE imaging

o NDE for materials properties

o NDE for advanced composites

o Ultrasonic stress

measurement

III. National Institute of

Standards and Technology

o Transducer standards

o Pulse-echo testing of

concrete

o Automated manufacturing

o Metals, plastics, ceramics,

paints, etc.

IV. NASA - Langley NDE Center

(J. Heyman)

o Space structures

o Aging aircrafts

o Imbedded sensors

o Smart structures

NSF INITIATIVE

Believing that "an ounce of

prevention is better than a

pound of cure," and that NSF

should play a leading role in

addition to be acting as a

catalyst in the "Quantitative

Nondestructive Evaluation for

Large Structural Systems" in

civil engineering, a NSF

initiative was issued in April

1990. The deadline for receipt

of

proposals was set at June 1,

1990. The initiative was

sponsored by the following four

programs, totalling $1M, over

a two year period. Program

directors involved are listed

below.

Dr. John B, Scalzi - Structures

and Large Systems

Dr. Ken P. Chong - Structures

and Building Systems

Dr. Oscar Dillon - Solid

Mechanics

Dr. Thomas Zimmie

Geotechnical , Geomechanics

Academic and research

institutions in the United

States with engineering,

research and educational

programs were invited to submit

proposals. The contents of the

proposal must clearly relate to

the research topics described

in the Programs of the Division

of Mechanical and Structural

Systems of the National Science

Foundation.

Projects were encouraged to be

multidisciplinary in nature

involving investigators from

one or more departments at a

single institution or from

sevei^al institutions wherever

the required expertise may be

located. It was envisioned

that projects may include

researchers from the following

engineering and science

disciplines: structural,

544
mechanical, material science,

electrical, geotechnical

,

computer science, chemical or

chemistry, electronic and

others as appropriated. A team

or group effort was strongly

encouraged to achieve the

desired result and final

product of research.

Contributions of all the

participants must be combined

into a single integrated

proposal which confomed to the

usual requirements and

limitations of a proposal

length. Each researcher must

have a scope of work which

contributes to the entire

project.

Industry collaboration is

highly desirable and may be in

the form of participating

personnel, loan of equipment,

gifts, supplies of various

kinds, and/or financial

contributions. The academic

institution must be the prime research management

organization.

A panel of experts have been

r^^-^tT^-^^^ *- ^ .-^-TT--. 1 ll -.4- ^ -UO..^ proposals on a competitive

basis. The investigators and

the synergistic benefits

derived, as well as how these

interactions are to be

organized and performed, are

examined.

SPECIFIC TOPICS OF THE NSF INITIATIVE

Research efforts are to be

directed toward increasing

fundamental knowledge of

quantitative nondestructive

evaluation of construction

materials and their application

to large structural systems of

wood, concrete, masonry and

steel. The data developed for

equipment may be for the

material properties alone or

for the fabricated forms as

total systems. Some data may

be used to determine how each

system or material

deteriorates with time and

excessive loads. The data and

prototype resulting from the

research should be of such a nature that the resulting

equipment may be useful for

practitioners, and can be

produced as a viable portable

and economical commercial

product.

SUGGESTED RESEARCH TOPICS OF INTEREST

The following topics of

interest were published in the

NSF Initiative.

o Determination of wood

properties - above and below

ground

o Determination of concrete

properties - above and below

ground

o Determination of masonry

properties

o Determination of steel

safety remaining as a result

of corrosion

o Determination of initial

cracking in steel and

concrete, including welds

and rebars due to fatigue,

fabrication, or

design faults

o Rate of deterioration in

steel or concrete

o Early warning of impending

failure

o General inspection of the

545
degree of deterioration in

steel, concrete, masonry,

soil retaining systems and

wood

o Determination of the

effective length of piles of

wood, steel, or concrete, and

the amount of erosion or scour

in foundations

o Methodology for predicting

the remaining safe life of

large structural systems

PkOjii;utS bu'FFORTtiD BY THE INITIATIVE

The recently awarded projects

are briefly described below.

The projects are in four basic

areas of interest to the

Structures, Solid Mechanics,

Geomechanics, and Building

Systems programs of the

National Science Foundation.

1 . Fiber Optic Sensors for

Strength Evaluation and Early

Warning of Impending Failure in

Structural Components ; Dr

.

Edward G. Nawy, Mohamad H.

Maher - Rutgers University, The

State University of New Jersey

The objective of this research

is to develop and verify by

experimentation the optical

fiber strain and stress sensor

systems for on-line, accurate,

and real time monitoring of

critical components of

structural systems under

service load conditions. The

application of these sensor

systems will pennit easier

inspection of structures for

impending failures.

A series of experiments are to

be undertaken on scale models

of reinforced and prestressed

concrete structural beams,

rolled structural steel

sections, and beam to column

connections.

2 . Nondestructive Methods of

Assessincf Damage in Reinforced

Concrete Structures ; Drs

.

Kenneth H. Stokoe, Jose M.

Roesset and James O. Jirsa - University of Texas-Austin

The objective of this research

is to study nondestructive

methods based ' on the

propagation and attenuation of

stress waves for evaluating (1)

the degree of damage to

structural and foundation

elements, (2) the location and

extent of that damage, and (3)

the extent and quality of

repair. One of the key thrusts

of this research will be the

integrated study of both body

and surface stress waves. The

techniques to be used will be

selected in terms of the type

of structural elements and

their accessibility for use of

nondestructive test equipment.

The rapidly-developing

technique based on surface wave

measurements will be used where

there is access to an exposed

surface of the structural

element which has dimensions

similar in magnitude to the

investigated. The Spectral

Analysis of -Surf ace-Waves

(SASW) method will be used.

3 . Micromaqnetic Surface

Studies of Materials for

Nondestructive Evaluation , Dr.

David Jiles - Iowa State

University,

Dr. W. A. Theiner - University

of Saarland, Federal Republic

of Germany

This project is a joint

546
research effort into the use of

magnetic surface studies for

nondestructive evaluation

between the Center for NDE at

Iowa State University and the

Fraunhofer Institute for

Zerstorungfreie Prufverfahren

(NDT) at the University of

Saarland in West Germany. The

objective of the research is to

identify alternating current

magnetic measurements which can

be made on the surfaces of

constructional steel components

to evaluate their mechanical

condition. Such methods, in

which the penetration of the

magnetic field is restricted to

the surface, will be useful for

determining the mechanical

state of the surface as a

result of shot peening,

grinding, surface hardening and

residual surface stress. It is

also known that in fatigue

failure, the fatigue crack

initiates at the surface and

then propagates into the bulk

of the material. Monitoring of

the surface condition is

therefore critical in

identifying the onset of

failure in the material as a

whole.

4 . Fundamentals of Embedded

Optical Fibers In Structural

Concrete; Drs. K. S. Kim, T. F.

Morse - Brown University

The thrust of this project will

be to combine the knowledge and

expertise of faculty members

from solid mechanics, materials

optical sensing to obtain a

more quantitative description

of the response of structural

concrete to external loads. By

embedding optical fibers in a concrete structure during the

fabrication process, it is

possible to sense, through a variety of techniques, the

local internal state of the

structure. This research will

aid in the prediction of the

behavior of such materials,

will point toward directions

for the improvement of

structural concrete, and can be

applied to the retrofitting of

existing structures with novel

sensors

.

NSF SUPPORTED WORKSHOP ON NDE

o "NDE for performance of Civil

Structures," Sam Masri

(USC) , 2/88.

o "International Workshop on

Sensors and Measurement

Techniques for Assessing

Structural Performance"

Richard Marshall (NIST) , 6/89.

o "Colloquium on Facility Monitoring and for

Infrastructure Management"

(Building Research Board/NRC)

.

o "NDE of Construction

Materials and Large

Structural Systems," (ColoradoBoulder), 10/90.

o "NSF-ATLSS Workshop on Sensor

and Signal Processing for

Structural Control" (NSF)

,

2/91.

SUMMARY AND CONCLUSION

An oveirview of non-destructive

evaluation projects at NSF and

U j.^v^ ci.Ok3 J. Wil Wii Ulii:: iicw iNOr

initiative and projects on the

"Quantitative Nondestructive

Evaluation for Large Structural

Systems" are presented. The

authors hope that this paper

will act as a catalyst,

sparking interest and further

547
research in quantitative

nondestructive evaluation of

civil engineering structures.

This paper reflects the

personal views of the authors,

not necessarily those of the

National Science Foundation.

548
NSF Supported Research in Concrete Structures and Materials

>

By

KEN P. CHONG, SHIH-CHI LIU

and HENRY J. LAGORIO

PROGRAM DIRECTORS,

NATIONAL SCIENCE FOUNDATION

WASHINGTON, D.C 20550 USA

ABSTRACT

The U.S. National Science Foundation spends

over U.S. $40 million annually for basic research

in civil engineering, covering areas in structures,

earthquake engineering, building systems, solid mechanics, geomechanics and environmental

engineering. Significant amounts of basic

knowledge and technology have been advanced

through these research efforts. Concrete

engineering, including structures and material

aspects, is an important component of civil engineering research. In this paper current and

recent research projects and findings are

discussed.

Keywords: Concrete structures; high

performance concrete; seismic concrete

structures; fracture of concrete; micromechanics.

1. INTRODUCTION

Structural failures emphasize the need for material research. The short-term solution is to have innovative NDE (nondestructive

evaluation) sensors to evaluate and diagnose the

integrity and predict the remaining useful life of

various structures. The long term goal should

involve creating and designing high-strength,

high tech (HSHT) materials with desirable

structural properties and reliable performance

specifications.

Advanced ceramics and high performance

concrete are HSHT materials that have a lot of

potential. Among the desirable properties,

ceramics is highly resistant to corrosion and

heat. However, one of the major drawbacks of

such materials is its brittleness, failing suddenly

without any warning. Researchers at Michigan

and others have tried mixing fiber in ceramics,

improving its ductibility, with some degree of

success. Others have concentrated on the

micro-mechanics, metallurgy, surface chemistry

and characterization. High strength concrete is a form of ceramics having compressive strength

as high as 40 ksi (about 10 times normal

concrete) which would enable future high rise buildings to be built with concrete. One of the

major problems with high strength concrete is that the fracture toughness (K;,.), which is a measure of the energy related to fracture, does

not increase proportionally, and needs to be

improved for structural performance.

The National Science Foundation, through its engineering programs in Structures,

Geomechanics and Building Systems, and the

Earthquake Hazard Mitigation programs, in addition to the Engineering Research Centers

(e.g. at Lehigh University) and the Science and

Technology Centers (NSF Center for Advanced

Cement-Based Materials), supported research in concrete materials and structures.

Some of the NSF supported projects are

listed below:

* Mathematical Modeling of Concrete Creep

and Shrinkage

* Research Needs for Concrete Masonry

* Rehabilitation, Renovation and

Reconstruction of Buildings

* High Strength Concrete

* Fiber Reinforced Concrete

* Fracture Toughness

* Micro-mechanics of Concrete

* Shear in Reinforced Concrete

549
* Physics and Chemistry of Cement-based

Materials

* Behavior of Concrete in Cold Climate

* NDT of Concrete

* Fiber Optics in Concrete

* Continuous Lightly Reinforced Concrete Joist Systems

* Highway Bridges

* Construction Methods

* Automation/Robotics

* Non-Distinctive Evaluation

* Seismic Precast Structures

* Structural Controls

* International Cooperative Initiatives

* Full-Scale Seismic Pseudo Dynamic Testing

of Reinforced Concrete Structures HIGH PERFORMANCE CONCRETE

More than half of the NSF supported

research projects are related to high

performance concrete (HPC), which is defined

as "Concrete having desired properties and

uniformity which cannot be obtained

routinely using only conventional constituents

and normal mixing, placing, and curing

practices" (Carino and Clifton, 1990). These

properties include: high strength, high

toughness, durability, enhanced workability,

freeze-thaw durability, etc. National programs on HPC exist in Japan, Canada, Norway, and

France. Generally chemical admixtures or

fibers can be added in concrete to achieve

certain mechanical properties. Concrete

admixtures (Rosenbaum, 1991) include: highrange water reducers (superplasticizers),

retarders, accelerators and air-entraining agents.

In the United States, W.R. Grace and Master

Builders are the two largest suppliers of the concrete admixtures, accounting for about twothirds of the U.S. admixture market. A new

development (2) is on extended-set retarders

and activators, which indefinitely keeps fresh concrete from setting (until it is mixed with an

activator).

Besides the basic concrete research sponsored

by the NSF, other federal programs, such as the

Strategic Highway Research Program (SHRP),

supports HPC and concrete research (3,4).

NSF PROJECTS IN PROGRESS

The following is a list of examples of ongoing NSF research projects in concrete. Highlights of some of these projects will be

presented at the Symposium.

* Bazant (Northwestern) "Nonlinear and

Probabilistic Theory for Concrete Creep."

The linear modeling of concrete creep with

aging has matured, and it is now appropriate

to develop a nonlinear theory. This research

will establish such a theory, based on

micromechanics analysis of the solidification process of portland cement and nonlinearities

due to strain-hardening and development of

backstress. The model will first be developed

for basic creep at constant humidity and

temperature, and is then generalized to include the effects of variable humidity and

temperature. This model is based on the

hypothesis that the rate of bond ruptures

(which is the source of creep) depends on the

microdiffusion flux of water molecules in cement gel. The model will be extended to include probabilistic aspects of material

parameter uncertainty, including correlation of

random parameters, and fluctuation of

environmental conditions. Random effects of

environment are handled by a new sampling

scheme based on spectral approximations

solutions from diffusion theory. A numerical

algorithm for step-by-step integration in time,

in finite element programs will also be

developed. Creep experiments are to be

conducted with a novel large triaxial-torsional testing machine with temperature control. Hollow test cylinders will be subjected to

triaxial loading at various temperatures with

confining pressure which prevent cracking or simultaneous strain-softening due to drying or temperature change. The results will improve

creep predictions for radioactive waste

storage, large span bridges, nuclear reactor

accidents, concrete shells and tanks, buildings

and other concrete structures.

* Brock (Kentucky) "Transient Studies of

Dislocation-based Micromechanical Effects in Fracture."

* Buyukozturk (MIT) "A Study of Fracture Behavior in High Strength Concrete."

550
Darwin/McCabe (Kansas) "Improving

Development Characteristic of Reinforcing

Bars" (Co-sponsored by Civil Engineering

Research Foundation).

The purpose of the research is to investigate

the shear strength of lightly reinforced concrete flexural members and develop

recommendations for modifications in current design provisions. The research includes

tests of 21 continuous two-span T-beam and

joist systems to identify the effects of flexural and shear reinforcement on shear capacity.

Flexural reinforcement ratios of 0.75 and 1.0 percent and levels of shear reinforcement

ranging from 0 to 75 psi are included. The

T-beams extend earlier work at the University

of Kansas, and are aimed at directly

addressing the level of safety provided by

current shear design provisions for lightly

reinforced members. The joists represent

typical configurations used in practice.

Efforts during the first year have

concentrated on evaluating the shear capacity

of continuous lightly reinforced T-beams.

Recently a literature survey has been

completed, forms have been constructed, and

six specimen cross-sections without shear

reinforcement have been tested. Of the six sections, only one provided a capacity in excess that predicted by the ACI Building

Code. Strengths between 80 and 90 percent

of the predicted strength were obtained in the other five cases. If borne out by the next

series of tests, these observations will have a major impact on design practice, for they

firmly establish for the first time that a wide

range of reinforced concrete beams have been

designed with a reduced margin of safety for shear. Analytical evaluation will include a comparison of the cracking shears with

existing empirical and analytical expressions.

The contribution of concrete, shear

reinforcement, dowel action, and aggregate

interlock to shear strength will be estimated. Fracture mechanics concepts are to be

applied to the failure of beams without shear

reinforcement and evaluate the applicability

of truss models to lightly reinforced members

with shear reinforcement.

Gopalaratnam (Missouri), Shah

(Northwestern), Batson (Clarkson), Criswell

(Colorado St.), Ramakrishnam (SDSM&T)

and Wecharatana (NJIT) "Fracture Toughness

of Fiber Reinforced Concrete (FRC)"

- study specimen size, loading

configuration/rate.

- evaluate currently used toughness and

specifications.

- provide basis for the use of

strength/toughness data.

- recommend guidelines in specifications.

* Hawkins Kobayashi (Washington) "Mixed

Mode Fracture of Concrete."

The objective of the proposed research is to continue prior University of Washington

studies on mixed-mode concrete fracture aspects of diagonal tension failure in reinforced concrete members. Specifically,

the shear transfer mechanism, which is thought to resist stable crack growth, will be

studied using crack-line wedge-loaded, double

cantilever beam specimens with short starter cracks and various loading histories which are

crack-length dependent. Moire

interferometry will be used to measure

simultaneously two orthogonal displacement

fields from which the crack opening

displacement and the slip of stably growing

diagonal tension cracks will be determined.

Moire data together with the loading data

will be used interactively with a finite element code to deduce the crack closure and

shear forces in the fracture process zone

under mixed mode loading. Mixed mode

stable crack growth criterion will be deduced

from this hybrid experimental-numerical

analysis. ITie results will be correlated with

fractography and profilometer analyses of the

fractured surfaces in order to assess the

relative micromechanic contributions of the

various constituents (hardened cement paste, coarse aggregate and fine aggregate, etc.) to mixed mode concrete fracture.

* Hinze/Holt (Washington) "Expansive

Cements"

- mix with water in bore holes, etc.

- use in liu of explosives

- measure expansive stresses

- geomechanics

551
Hover (Cornell) - To construct durable

concrete.

- 60 ksi automatic mercury porosimeters were built, have been used in a microsilica-based concrete parking garage.

Hsu (Houston) "Shear in Reinforced

Concrete."

Shear action is an important factor in the

failure of reinforced concrete structures. A

softened truss model theory has been

developed that is capable of predicting the

behavior of various types of structures when

shear is predominant. The first objective is to generalize the softened truss model theory

so that it can be applied to situations where

shear (torsion) is combined with axial loads. The second objective is to perform tests on

40 full-size reinforced concrete panels in order to determine the material laws

required in the general theory, including the

softening law, the constructive law of cracks,

and the law governing the contribution of

concrete. The research are being carried out

in a $440,000 panel-testing facility recently

completed at the University of Houston.

This facility consists of a giant steel frame

embedded with 40 jacks of 100-ton capacity

each, which can test reinforced concrete

panels 55 inches square and up to 16 inches

thick. Establishment of a general truss model theory in conjunction with the finite element method will permit one to predict

the behavior of a whole range of structures subjected to membrane stresses (walls and

shells). A few full-size panels made of high

performance concrete are being tested.

Kangari (Georgia Tech) "Advanced

Technologies and Materials in Concrete Slipforming."

Lee and Chung (SUNY-Buffalo) "Effect of

Freezing Cycles on the Mechanical Behavior

of Concrete."

- 9 to 12 ksi concrete is planned for year 2.

- low cycle

Li (Michigan) "Fracture Testing Technology

for Advanced Ceramic and Cementitious

Matrix Based Composites."

This project investigates the bridging processes of quasi-brittle materials by means

of a unique fracture testing technology

developed by the Principal Investigator, with

specific focus on high strength concrete,

advanced ceramics and their composites (e.g. metglas and polymer fiber reinforced high

strength concrete; and silicon carbide

reinforced silicon nitride to be manufactured

and provided free of cost by Norton

Ceramics). The objectives of the proposed

work are (1) to refine and extend a testing

technique based on sound theoretical

principles for testing of non-linear fracture parameters, including the post-cracking force- displacement curve; and (2) to develop a data

base of accurate fracture parameters for high

strength concrete and ceramics which exhibits

the development of process zones large

enough to invalidate classical linear elastic fracture mechanics treatment.

The test technique requires stable measurement of load versus load point

displacement and crack tip opening

displacement beyond the peak load. Successful load tracking at this stage requires

system stiffness from both the loading frame,

loading train and specimen.

* Monteiro (Berkeley) - Micromechanics in Concrete.

* Naaman (Michigan) "High Performance Fiber

Reinforced Cement Composites:

International Workshop." 6/23-26/91,

Stuttgart, Germany (with Reinhardt and

German DFG).

* Nanni (Penn State), "Aramid Fiber

Reinforced Concrete."

* Nawy/Maher (Rutgers) "Fiber Optic Sensors

for Strength Evaluation and Early Warning

of Impending Structural Failure."

The main objective of this proposed research

is to develop and test optical fiber strain and

stress sensor systems for on-line, accurate,

and real-time monitoring of critical components of structural systems under

service load conditions.

552
In this proposed research the combined

resources of Rutgers' Civil Engineering

Department and those of its center for Fiber

Optic Material Research Program (FOMRP)

will be devoted for development and

implementation of both discrete and

continuous optical fiber strain/stress sensors

for detection and warning of imminent

structural failures for both concrete, and

steel structures. In the case of concrete

structural components, the sensor system will be evaluated under both embedded (internal)

and exposed (external) condition; where the

latter is attached to existing structures and

the former is embedded in the structure during its construction.

Ross (New Mexico) "Dynamic Fracture in Quasi-Brittle Materials: An Experimental

Study."

Sansalone/Pratt (Cornell) "Neural Network in Non-Destructive Testing."

- impact-echo develop by Carino and

Sansalone.

- automating and simplying impact-echo

signal analysis/reduction.

- applicable to HFC.

Englekirk (Englekirk & Hart) "Concept

Developments for Precast Concrete

Structural Systems."

This is a part of the coordinated research

program in seismic precast concrete

structures. This project will develop

structural concepts for using precast concrete

in seismic zones of the United States. Due

primarily to current building codes and lack

of relevant analytical and experimental data,

precast concrete is not currently used in U.S.

seismic zones.

Initially, several regional research workshops

will be held around the country. These will be limited in size to facilitate interaction

among the participants who will be invited to represent various contributors to the

design/construction industry (including

architects, developers, researchers, etc.)

Next, the economic viability as well as

structural feasibility of the design concept

will be investigated. Design documents will be prepared using the concepts resulting from

the workshops. Feedback about economical

ways to improve the details proposed from

general contractors will be obtained.

Potential use of automated processes will also be considered.

The analytical model development and design

criterion development projects will require a feedback from design engineers and other

industry representatives. The required

information to academic researchers and

other means will be provided throughout the

duration of this project.

* Stanton (Washington) "Connection

Classification and Modelling for Precast Seismic Structural Systems."

* Swartz (Kansas State) "Influence of

Concrete Constituents on Mode 1 Fracture

Characteristics"

Since the processes of initiation and growth

of cracks which lead to fracture in concrete must surely be influenced by the various

components in the mixture (as well as state of stress, curing conditions, moisture content,

etc.), it would be desirable to verify that the

proposed testing standards are suitable for a wide range of key mix parameters.

Concomitantly, the influence of different mix

parameters on fracture properties including

the shape of the crack front and process zone

also may be determined.

The successful completion of this work will provide further evidence that fracture mechanics is a viable and useful tool for the

characterization of the behavior of

cementitious materials subjected to stress and

for the development of design formulas for shear strength, bond strength, deflection

response and crack width of reinforced concrete structures in which size and

brittleness effects can be considered explicitly.

* Keiser (Illinois) "An Experimental Setup to Investigate Dynamic Torsional-Translational

Response of Reinforced Concrete."

* Azizinamini (Nebraska) "Design and

553
Detailing of Transverse Reinforcements for High Strength Reinforced Concrete Columns

Subjected to Seismic Loading."

The development and investigation of high

strength reinforced concrete applications

comprises a major advance in building

planning and design because of its increased

capacities. The present state of high

strength reinforces concrete design needs in- depth study to accurately predict the

anticipated performance of its many

component parts under dynamic lateral loading such as earthquake loads. High

strength reinforced concrete is also a

sophisticated and complex building material

due to the large number of design and

construction constraints which influence its quality, long term behavior, and final stress values. In order to more accurately model

its structural response, it is necessary to develop a fundamental data-base of

performance standards.

The proposed project is concerned with the

development of design and detailing criteria for transverse reinforcement in high

rectangular columns with compressive

strengths exceeding 15,000 psi. Its objectives are to determine methods to provide

sufficient ductility for earthquake-resistant

structures. The research plan includes both

experimental and analytical studies. At least twelve large-scale column specimens will be

tested and be subjected to axial and repeated

lateral loads at increasing displacements.

From these tests, data will be collected for the analytical documentation of the effects of

concrete compressive strength, axial load, and

types of transverse reinforcement on the

ductility of high strength concrete columns.

Wood (Illinois) "Experimental Investigation

of the Strength Stiffness and Deformation

Capacity of Slender Reinforced Concrete

Walls."

During the past 20 years, the satisfactory

behavior of many reinforced concrete

buildings during damaging earthquakes has

been attributed to the presence of structural

walls. Structural walls provide stiffness to the building systems, and control the drift and damage in the buildings. A critical

concern for the design of structural walls in seismic regions has been the specification of

reinforcement details to insure ductile response. The performance of buildings in Vina del Mar during the March 1985 Chilean

earthquake appears to contradict professional

opinion, because the buildings performed well

and the reinforcement details were essentially

nonexistent.

This research project develops methods for determining the strength, effective stiffness,

and deformation capacity of reinforced concrete walls. Walls are used as the

primary lateral-load resisting system in many

reinforced concrete structures, yet many

fundamental concepts have not been verified

experimentally.

The experimental phase of the project

involves lateral load tests of reinforced concrete walls. Major variables include the

geometry of the cross section, amount of web

reinforcement, type of loading, and amount

of confinement reinforcement. Walls are subjected to static load reversals and dynamic

loading.

The analytical phase of the project uses the

observed response and the measured response

of the experiments to develop a hysteresis

model suitable for reproducing nonlinear

response in shear and flexure. Methods are developed to determine the strength, likely

mode of failure, and effective stiffness of

structural walls during the design process.

* Wight (Illinois) "Earthquake Type Loading on R/C Beam-Column Connections: Special

Cases of Wide Beams and Eccentric Beams."

* Moehle (Berkeley) "Seismic Resistance and

Retrofit of Post-Tensioned Flat-Plate Floors."

The flat-plate floor system is generally

recognized as an unsuitable structural system

for resisting high-intensity lateral loads. Nevertheless, in regions of high seismicity in the U.S. this structural system is still one of

the most widely used systems in multistory

buildings braced by stiffer ductile frames of

structural walls. Recent research has shown

that even in these braced structures the

conventionally-reinforced flat plate is

554
susceptible to failures if gravity shear stresses on the floor and lateral drifts are not

controlled to acceptable values. Performance

of the unbonded post-tensioned flat-plate floor in a similar situation is not well known.

This research studies the behavior of this structural system.

The research involves design, construction,

experimental testing, retrofitting, and analysis

of unbonded post-tensioned flat-plate

construction subjected to gravity and reversed

cyclic lateral loads simulating earthquake

effects. Both isolated connections and a three by three bay model floor are studied. Effects of gravity load intensity, lateral load

intensity, and retrofitting techniques are the

primary variables in the study. The

experiments determine the suitability of this structural system to support gravity floor loads when subjected to cyclic lateral drifts typical of those occurring in buildings during

intense ground shaking. One of the primary

objectives is to establish guidelines for design

of this structural system.

* Eberhard/Sozen (Illinois) "Experiments and

Analysis to Study the Seismic Response of

Reinforced Concrete Frame-Wall Structures with Yielding Columns."

* Hsu (Houston) "Reinforced Concrete Framed

Shearwalls in Earthquake."

REFERENCES

1. N.J. C::arino and J.R. Clifton, "Outline of a National Plan on High- Concrete: Report on

the NIST/ACI Workshop, May 16-18, 1990,"

NISTIR 4465, National Institute of Standards

and Technology, USA, Dec, 1990.

2. D. B. Rosenbaum, "Concrete Admixtures,"

Engineering News Record, pp. 35-40, Jan. 21,

1991.

3. Strategic Highway Research Program,

"Concrete and Structures: Progress and

Products Update," National Research

Council, USA, Nov., 1989.

4. "Nationally Coordinated Program of Highway

Research, Development, and TechnologyAnnual Progress Report

FY 1990," FHWA-RD-90-109, Federal

Highway Administration, USA

5. T. Hsu and S. Mau, "International Workshop

on Concrete Shear in Earthquake."

University of Houston, January 1991.

NSF CENTER FOR ADVANCED

CEMENT-BASED MATERIALS

Northwestern (Shah) Michigan

niinois NIST

Purdue

Areas of Investigation

* Processing procedures (control of initial/final structure).

* Microstructure (porosity, damage, interfaces)

* Bulk Properties (fracture mech., durability,

FRC)

* Modeling (generic, interfacing, simulation,

database)

* Design of New Materials (improved , ^ properties through material science)

555

Development of Water Level and Wave Heiglit Design Data

by

Martin C. Miller^ William E. Roper^,

Leon E. Borgman^, and Joannes J. Westerink'

ABSTRACT

The historical record of tropical storms at specific sites in the Gulf of Mexico is too short

to provide adequate statistics for estimates of

extreme waves and water levels for

engineering design and project economic

evaluation. This paper describes a method that

has been applied by the Coastal Engineering

Research Center (CERC), U.S. Army Engineer

Waterways Experiment Station to expand the

available suite of storms and sample the

historical storm data set to obtain reliable,

statistical parameters of storm effects.

Parameters from historical storms were used,

along with additional, hypothetical storms, to obtain relationships between the parameters

associated with storm characteristics and storm

effects of wave height, water level, and

shoreline recession. Water level estimates were

obtained using a finite element storm surge,

tide and current model which was recently

developed and tested for the Gulf. This model

offers considerable advantages over the finite

difference models that were previously used

for this purpose. The extreme conditions at the site were obtained using a "bootstrap"

statistical technique to reduce the number of

storms modeled. The methods described were

used to evaluate beach fill design alternatives

at Panama City Beach, Florida in the northern

Gulf of Mexico.

KEYWORDS: Panama City Beach, Florida;

modeling; water levels; storm surge; wave

heights; "bootstrap."

1. INTRODUCTION

Panama City Beach, Florida is a recreational

beach located in the northern Gulf of Mexico

along the panhandle of west Florida (Figure 1 ). The beach is heavily developed over much of

its eighteen mile length, with single family

homes and condominiums built along the waters' edge. Inlets at each end of the beach

provide access to small boat harbors and quiet

coastal lagoons. The beach is susceptible to flooding from hurricane related storm surge,

which could result in great loss of life and

property. Beach and flood protection is considered necessary but the recreational and

aesthetic value of the area must also be

preserved. An artificial berm is being

considered along beach front to protect the

region from flooding. In order to determine

the economic justification for the project and

to effectively design the berm, it is necessary

to evaluate the degree of protection it will

provide during storms of various magnitude

that may impact the site. From the designers

standpoint, it is desirable to know the return

period of the storm that would be protected

against for a selected design. From the

economic standpoint, project justification

requires that the benefit of the project exceed

the cost. This requires an economic model that

allows assessment of the damage for an

expected flooding level.

^ Chief, Oceanography Branch, Coastal

Engineering Research Center, U. S. Army

Engineer Waterways Experiment Station,

Vicksburg, MS 39180-6199

^ Assistant Director, Directorate of Research

and Development, Civil Works Program,

Headquarters, U. S. Army Corps of Engineers,

Washington, DC 20314-1000

^ L. E. Borgman, Inc., 2526 Park Avenue,

Laramie, WY 82070

Assistant Professor, Civil Engineering,

University of Notre Dame, Notre Dame, IN

46556-0767

557
In the northern Gulf of Mexico, the design

storm is the hurricane. These non-frontal

storms enter the Gulf of Mexico from the

Atlantic (and sometimes the Pacific) during the

months of June through November. The

months of most frequent occurrence are August through October. During the period

from 1871 to 1973, a total of 281 tropical

cyclones entered the Gulf coast of the United

States, an average of 2.75 storms per year. These were, by no means, evenly distributed in

time or space, however. A detailed analysis of

these storms and their characteristics can be

found in Ho, et al. (1987) and later publications of the National Oceanic and

Atmospheric Administration (NOAA),

National Weather Service.

Though records exist of many storms that have

occurred throughout the Gulf, the historical occurrence at any specific site is usually

insufficient to provide a statistically reliable

basis by which to calculate extreme values of

waves, water levels, and the resulting shoreline

recession for long return periods. Designs of

coastal structures and beach fills are often

based on the fifty or one-hundred year storm

condition. There are virtually no areas where

measurements of the desired parameters have

been made for that length of time.

Designers must, therefore, rely on numerical

models to obtain the necessary statistical data. The spectral wave model, SHALWV,

has been developed at CERC and has been

used to determine wave conditions in a number

of coastal applications including hurricanes

(Jensen, et al., 1987; Hughes and Jensen, 1986;

Abel, et al., 1989). It will reliably provide a time series of the wave height and period on a

grid overlying the area of interest provided the

wind distribution for the region is known.

The winds are obtained from direct

measurement, in the case of more recent

storms or they can be calculated using the, well

established. Standard Project Hurricane (SPH)

model and the relevant parameters of storm

path and landfall location, speed of movement,

central pressure deficit, maximum wind speed,

and radius to maximum winds.

Historical data on the tropical storms and

hurricanes that impact the northern Gulf of

Mexico have been collected and published

since the late 1800's. An archive of storm

parameters is maintained by the National

Hurricane Center in Miami, Florida. These

data can be used to obtain the landfall

location, path, central pressure, maximum

winds, and (sometimes) radius to maximum

winds for storms that impact the site of

interest. The time histories of wave conditions

and water levels, required for engineering

design or project evaluation, are not included

in the data and, except on rare occasions

during the most recent storms, measurements

of these parameters do not exist. Reliable

measurements of recession rate are virtually

nonexistent.

2. WATER LEVEL MODEL

Until quite recently, finite difference models

have been used to calculate the water level due

to tides and storm surges in coastal areas. These models require that the area of interest

be discretized with regular spatial resolution,

typically one degree by one degree, or with

nested grids of smaller resolution in areas of

particular interest. A fully non-linear finite

element approach has now been formulated

that allows greater flexibility in the

representation of complex boundary geometry.

This model, developed by Westerink, et al. (1991) and by Luettich, et al. (1991) has been

applied to calculate the tide in the entire Gulf

of Mexico and has been calibrated to match

measured tidal response at twenty tidal stations

around the Gulf. A finite element grid was

created to cover the entire area (Figure 2) with

the particular area of interest for water level

calculations along the northern Gulf of Mexico

from Mobile Bay to the Florida Bight (Figure

3).

The Mobile Bay area is the site of intensive

investigations in the Army Corps of Engineers,

Dredging Research Program. One of the

elements of this $35m dollar, five year

program is to determine the potential for

resuspension of dredged material disposed of

in the nearshore zone, within the potential

reach of waves and currents. This has required

the determination of the water level during all

558
tidal stages and during periods of hurricanes

that affect the Gulf, in order to describe the

detailed structure of the bottom boundary

layer under the combined influence of waves and currents. The second area of interest is the beach berm construction at Panama City

Beach, Florida. Though no additional projects

are presently planned, the northern Gulf of

Mexico is heavily used for business and

recreation and it is expected that detailed

knowledge of the water level along the

coastline will eventually be required.

A grid of triangular elements was constructed over the entire Gulf of Mexico. The density

of triangulation was determined in relation

with the depth over the different areas investigated and the need for accurate results

at selected sites. The grid coverage for

Panama City Beach and the adjacent offshore

area are shown in Figure 3. Typical grid

spacing in the open ocean is about 27 nautical

miles (50 km), while nearshore the spacing was

reduced to 6600 yards (2000 meters) for a

coastal segment. The smallest grid element was

300 yards (330 meters). The ocean tide was

induced at the boundary of the Gulf, i.e. at the

Straits of Yucatan and the Straits of Florida, as well as by tidal potential forcing within the

interior domain. The input variables include

earth rotation and a constant bottom friction term. No additional tuning of the model was performed for the calculation of the tides. Comparison of model and measurements for

several tide stations along the Gulf are shown

in Figure 4. The tides are shown to be

accurately calculated at these stations from

Mobile in the west to the Florida Bight in the

eastern side of the Gulf.

Once the grid was set up and confirmed for

the tides, the model was applied to determine

the water level due to hurricane storm surge at Panama City Beach. The finite element

approach allows the calculation of storm surge over the entire Gulf using the, well

established, tidal boundary conditions at the

measurement stations. This has the

considerable advantage that the boundary

conditions do not have to be tuned for

individual storms and, therefore, provides a much greater predictive capacity for the

model. The great flexibility of the triangular

finite element grids allowed the establishment

of the proper resolution of coastal features. If these had not been properly resolved, the

storm surge would have been seriously underpredicted.

Water level data have been collected at tidal

stations located near the beach, and

observations of high water were available for

post-storm surveys made by Corps of Engineer

personnel and others. The water levels from

hurricane Kate (November 19-22, 1985) were

simulated and are compared to measurements

at locations east and west of the landfall

location (Figure 5). Storm track, central

pressure deficit and radius to maximum winds were used in the (SPH) model to calculate the

wind field over the Gulf of Mexico. These

show that the model accurately predicts the

surge elevation, particularly on the eastern, or "dangerous" semi-circle, of the storm. The

results were only slightly less successful on the

western side. This was judged to be a result of

the SPH model's inability to accurately

represent the changes in wind speed in the

transition zone from land to water.

3. STORM SELECTION

There are several ways in which reliable

statistics for determining storm effects may be

obtained. The reconstruction of historical

storm activity provide a direct, concrete

summary of what is likely to have occurred in

the past storms. When studies are concentrated on a particular local site, however, it is unlikely that a sufficient number of storms

have occurred over the period of record to provide reliable statistical measures. The

population of storms for the last 75 years may

show apparently random variation in number

of storms and storm intensity that are difficult

to ascribe to local climatology. Rather, it seems more likely that future storm activity

will possess sufficient sampling variability to

fill in the present gaps in the historical data.

There are at least two statistical procedures

that can be employed to utilize the general

statistics of the historical storm data and

analyses, but which allow the future to have

realistic variation from the past. (1) The

559
storm climatology can be used to establish

joint probability laws for the geometric and

intensity parameters of the storms, and these can be used with substantial subjective

engineering judgement to select typical storms

with estimated probabilities of occurrence as input to the engineering studies. (2) The

vectors of geometric and intensity parameters can be used in an extension of statistical procedures, known as "bootstrap" techniques to simulate a large population of storm activity

(Efron, 1982; Borgman, 1990).

The pure bootstrap method uses random

selections from the historical population to compute some statistic of interest which

depends on the data set. For example, suppose

that Xj, Xj, X3,....,Xjj are independent

observations in the sample taken from some

population. It is a simple matter to calculate

the 95th percentile of the data. The single

sample of data can be used to compute one

value of this percentile. If many such samples were available, it would be possible to calculate the 95th percentile for each and to determine the standard deviation. Since only one sample is available, however, it is not

possible to calculate the standard deviation.

A surprisingly accurate estimate of the

standard deviation can be obtained by

repeatedly drawing n observations from the

sample population with replacement. This can

be done rapidly in a computer until a large

number (say, 10,000) of synthetic samples of n

observations are assembled. It has been shown

that the standard deviation for the statistic,

which was prepared from the synthetic

bootstrap samples, is not much different from

what would have been obtained if one had

taken a large number of independent samples

from the original population, calculated the

statistic from each of these real samples, and

computed the standard deviation (Efron,

1990).

The basic bootstrap procedure does not allow

for the occurrence of values outside of the

historical data set. This is probably

satisfactory for relatively large historical data

sets when dealing with statistics primarily

related to the mean or middle behavior of the

data. Most designs are based on the extreme

values that exist at the upper tail of the

distribution. There may also be "holes" in the

data which could be filled by likely future occurrences. It may be necessary to apply some sort of smoothing or interpolation

procedure to extend the storm data set to include additional storms. Of course this step

must be made with considerable engineering

judgement and evaluation of the results as they

are generated.

The result of the extended bootstrap procedure

is a large population of possible future storm

activity. This method provides a population of

several thousand storms to which coastal

engineering models can be applied to determine the probabilities, bias and standard

variation of coastal flood stage, wave

conditions, beach recession, or whatever other

statistic is of importance to the engineering

study. Because of time and money constraints,

it is usually only possible to carry out complete

model calculations for 30 to 40 storms. Even

if computer time is free, the data file manipulation required in the cascade of

numerical models applied to achieve the final

result makes the simulation of all storms

impractical.

4. APPLICATION TO PANAMA CITY

BEACH, FLORIDA

Tropical storms and hurricane data from the

National Hurricane Center indicated that

forty-two storms made landfall within

seventy-five miles of Panama City Beach since

1886. Of these, fourteen were of hurricane

strength at the time they were within that

radius. The historical record does not supply

enough hurricanes to obtain statistically

significant extrapolations to extreme

conditions. Though the record shows the

storms that actually occurred and provides

information on the frequency of occurrence, it does not provide a record of the storms that

might have occurred with small random

variations in conditions. The relationship

between the storm parameters and storm

results can be investigated by "inventing"

storms to fill in the parameter space.

560
The number of hurricanes that were used to determine extreme conditions along Panama

City Beach was expanded, first, by path

translation in which the track of the storm was

shifted parallel to itself in such a way as to make landfall at the same stage of storm

development. The path of each historical

storm was translated 50 miles to the right and

left of the recorded track. The procedure

increased the number of storms but did not

triple the result since some storms were

translated out of the area of influence on the

site and others provided near duplicate

conditions. Some judgement must be applied

at this stage to eliminate duplicate storms and

to develop synthetic storms to fill in the

parameter space.

The storm selection procedure resulted in

approximately fifty storms for which the

parameters of maximum wind speed, radius to maximum winds, central pressure deficit,

forward speed, and landfall location were known. These parameters were used in the

SPH model to obtain the wind field

surrounding the storm. The wind field was

then used to calculate the time series of wave

conditions using the spectral wave model

SHALWV. -The water level was calculated

using the finite element model described

above, and the recession of the shoreline

determined from the SBEACH model (Larson

and Kraus, 1989; Larson, et al., 1990).

Multiple regression equations were developed

to determine the relationships between wave

height, water level, and shoreline recession as dependent variables and the storm parameters

as independent variables. The nearest- neighbor, extended bootstrap procedure was

then used to select the parameters the storms

had and the regression relationship was used to determine the result of the parameters. A very

large number of storm sets was bootstrapped in

this way to provide guidance for the conditions

for return periods of interest.

5. CONCLUSIONS

The Corps of Engineers has the responsibility

to provide design criteria for coastal protection

projects in hurricane prone areas. An

innovative statistical approach has been

developed to provide the data required for

extrapolating historical hurricane storm effects

to long return periods. The water levels generated by these hurricanes has been

modeled using a finite element model,

ADCIRC. This model allows the computation

of water level due to hurricane surge over the

entire Gulf of Mexico without tuning the

boundary conditions to provide a match of the

surge elevation at specific locations. These

new techniques have been evaluated by the

Corps, and have recently been applied to the

design of a protective beach fill at Panama

City Beach, Florida.

6. ACKNOWLEDGEMENTS

The research presented in this paper, unless

otherwise noted, was sponsored by the Office

of the Chief of Engineers, U. S. Army, and is published by its permission. Panama City

Beach studies were conducted for the Mobile

District, U. S. Army Corps of Engineers.

7. REFERENCES

1. Abel, C E., B. A. Tracey, C. L. Vincent,

and R. E. Jensen, "Hurricane Hindcast

Methodology and Wave Statistics for

Atlantic and Gulf Hurricanes from

1956-1975," Wave Information Studies of

US Coastlines, WIS Report No. 19,

Coastal Engineering Research Center, US

Army Engineer Waterways Experiment

Station, Vicksburg, MS., March 1989.

2. Borgman, L. E., "An Algorithm for

Optimal Efficiency in Selecting Storms

for Finite Element Modeling of Surge,

"Contractor Report, Coastal Engineering

Research Center, U.S. Army Engineer

Waterways Experiment Station,

Vicksburg, MS., 18 November 1990.

3. Efron, B., "The Jackknife, The

Bootstrap, and Other Resampling Plans,"

Regional Conference Series in Applied

Mathematics. No.38 . Society for

Industrial and Applied Mathematics,

Philadelphia, PA., 1982.

561
4. Efron, B., "More Efficient Bootstrap

Computations," Journal of the American

Statistical Association, 85(409), pp.79-

89, March 1990.

5. Ho, F. P., J. C. Su, K. L. Hanevich, R. J. Smith, and F. P. Richards, "Hurricane

Climatology for the Atlantic and Gulf

Coasts of the United States," NOAA

Technical Report NWS 38, National

Oceanic and Atmospheric

Administration, National Weather

Service, Washington, DC, April 1987.

6. Hughes, S. A. and R. E. Jensen, "A

User's Guide to SHALWV: Numerical

Model for Simulation of Shallow Water

Wave Growth, Propagation, and Decay,"

Instruction Report CERC-86-2, Coastal

Engineering Research Center, US Army

Engineer Waterways Experiment Station,

Vicksburg, MS., June 1986.

7. Jensen, R. E., C. L. Vincent, and

C. E. Abel, "A User's Guide to SHALWV: Numerical Model for

Simulation of Shallow-Water Wave

Growth, Propagation, and Decay; Report

2, SHALWV-Hurricane Wave Modeling

and Verification," Instruction Report

CERC-86-2, Coastal Engineering

Research Center, US Army Engineer

Waterways Experiment Station,

Vicksburg, MS, June 1983.

8. Larson, M. and N. C. Kraus, "SBEACH:

Numerical Model for Simulating StormInduced Beach Change, Report 1, Empirical Foundation and Model

Development," Technical Report

CERC-89-9, Coastal Engineering

Research Center, US Army Engineer

Waterways Experiment Station,

Vicksburg, MS., July 1989.

9. Larson, M., N. C. Kraus, and

M. R. Byrnes, "SBEACH: Numerical

Model for Simulating Storm-Induced

Beach Change, Report 2, Numerical

Formulation and Model Tests," Technical

Report CERC-89-9, Coastal Engineering

Research Center, US Army Engineer

Waterways Experiment Station,

Vicksburg, MS., May 1990.

10. Luettich, R. A., J. J. Westerink, and

N. W. Scheffner, "ADCIRC: An

AdvancedThree Dimensional Circulation

Model for Shelves, Coasts and Estuaries,

Report 1: Theory and Methodology of

ADCIRC-2DDI and ADCIRC-3DL,"

Coastal Engineering Research Center, US

Army Engineer Waterways Experiment

Station, Vicksburg, MS., In Prep.

11. Westerink, J. J., R. A. Luettich, and

N. W. Scheffner, ' "ADCIRC: An

Advanced Three DimensionalCirculation

Model for Shelves, Coasts and Estuaries,

Report 2: ADCIRC-2DDI Users

Manual," Coastal Engineering Research

Center, US Army Engineer Waterways

Experiment Station, Vicksburg, MS., In Prep.

562
Pensacola

GULF OF MEXICO

Figure 1. Northern Gulf of Mexico from Louisiana to the Florida Bight.

563
Figure 2. Finite element grid covering the Gulf of Mexico. Grid elements are concentrated in the northeastern Gulf from Mississippi Sound to the Florida Bight.

564
565
Figure 4. Comparison of the measured tidal signal (dashed line) to the numerical

solution (solid line) using the program ADCIRC for tidal stations to the east and

west of the study area.

566
HURRICANE KATE HURRICANE KATE

PANAMA CITY - OESTIN

16

Figure 5. Comparison of the measured (solid line) and computed (dashed line)

water level (tide plus surge) during hurricane Kate (18-24 November, 1985).

Panama City and Destin are to the west of landfall of the storm center. Turkey

Point and Carrabelle are to the east of landfall.

567

U.S. Coordinated Program for Masonry Building Research:

Sixth Year Status

by

James L. NolandABSTRACT

The U.S. program, which consists of

twenty-eight specific research tasks, is

coordinated with a parallel program in

Japan. Both are conducted under the

auspices of the UJNR Panel on Wind and

Seismic Effects. The primary purpose of

the U.S. program is to support the development of a limit state design procedure for masonry as well as provide ex- perimental data, analytical methods,

etc. Several technical accomplishments

have been realized in both modeling and

experimental areas. The technology

transfer plan includes dissemination of

research reports, topical reports and a summary report on the specific research

areas as well as by presentations and

papers. Design and criteria recommendations will be an extremely important

mode of technology transfer and are ex- pected to support limit state design

standard development. Current status of

the program is presented. Completion is

expected by spring 19 93.

KEYWORDS: masonry, progress,

structures, technology transfer, UJNR

1. INTRODUCTION

The U.S. Coordinated Program for Masonry

Building research is an integrated set

of 28 specific research tasks being con- ducted by the U.S. Technical Coordinating Committee for Masonry Research

(TCCMAR/US) . The U.S. program is coor- dinated with a parallel program in Japan

to exchange information and concepts for

the mutual benefit of both countries.

Both programs are conducted under the

auspices of the United States-Japan

Natural Resources Panel on Wind and

Seismic Effects.

2. PROGRAM OBJECTIVES

Primary program objectives are:

1) Development of design and criteria recommendations for limit

state design of reinforced masonry buildings and components.

2) Development of a consistent

experimental database on the behavior of masonry materials, com- ponents and systems.

3) Development of analytical non- linear models for research and

design office use for detailed

. analysis, system analysis, and

dynamic loads determination.

4) Improved material and subassemblage experimental procedures for obtaining masonry

properties

.

5) Improved masonry fabrication

procedures and standards.

6) Developing an increased awareness among engineers, architects, code

bodies and the public of the

capabilities of reinforced masonry in all seismic zones.

7) Interfacing with standards development groups to support

development of a consensus limit

state standard for masonry.

Atkinson-Noland & Associates, Inc.

2 619 Spruce Street

Boulder, CO 80302

569
3. APPROACH 4. CUBBENT STATUS

With NSF support, the Technical Coordinating Committee for Masonry Research/U.S. (TCCMAR/U. S . ) was formed in

February 1984. TCCMAR/U. S. was and is

comprised of researchers from academic

and industrial organizations who have

strong backgrounds in research into the

properties and characteristics of reinforced masonry materials, structural

components and systems, analytical techniques, structural dynamics, building

codes, and earthquake engineering.

TCCMAR researchers are listed in Table

1. TCCMAR participants defined the re- search program conducted, the research

and will analyze and interpret the re- sults .

The U.S. research plan is a phased stepby-step program of separate, but

coordinated research tasks. Emphasis is

being placed upon intra-task information

exchange and consultations so the experimental and analytical efforts are mutually supportive. The research tasks

comprising the coordinated program are listed in Table 2.

The program includes both experimental

and analytical research tasks. The ex- perimental and analytical efforts are conducted in parallel with each becoming

more complex and sophisticated as knowledge and experience accumulate. The

U.S. plan is thus a "building block" approach. The program began in October

1985 and will be completed in the spring

of 1993.

As of March 1991, the program is approximately 75 percent complete. Remaining

work consists of refining analytical

models and programs, further, experimental studies on reinforcing bar bond and

splices in grouted masonry, development

of masonry material tests compatible

with the needs of nonlinear analysis and

limit state design, experimental and an- alytical studies of the behavior of the

five story research building, development of limit state design and criteria

recommendations and final technology

documentation. Research task reports

which have been published are listed in

Table 3. ....

Partial funding for planning and con- ducting experiments on a five-story

full-scale reinforced masonry research

building has been arranged. Researchers

at the University of California-San

Diego will commence detailed planning in

mid-April. Specimen construction will

begin in August 19 91 and testing is ex- pected to begin in the early fall of

1991.

Funding for development of design recom- mendations and technology documentation

is expected by May 1991. Work in that area will begin shortly thereafter.

5. ACKNOWLEDGEMENTS

Primary funding for the program was and

is being provided by the National Sci- ence Foundation (U.S.). The U.S. Army

Corps of Engineers has provided support

for extended studies on reinforcing bar

bond and splicing and material tests.

The U.S. masonry industry has provided

masonry materials and funds for specimen

construction

.

570
TABLE 1

TCCMAR RESEARCHERS

Daniel Abrams

University of Illinois

Urbana, IL "

Richard Atkinson

Atkinson-Noland & Associates

Boulder, CO

Robert Englekirk

Englekirk, Hart & Sabol

Los Angeles, CA

Ahmad Hamid

Drexel University

Philadelphia, PA

Gary Hart

Englekirk, Hart & Sabol/UCLA

Los Angeles, CA

John Ka riot is

Kariotis & Associates

South Pasadena, CA

Ronald Mayes

Computech Engin . Services

Berkley, CA

Max Porter

Iowa State University

Ames, lA

Freider Seible

Univ. of CA-San Diego

La Jolla, CA

Leonard Tulin

University of Colorado

Boulder, CO

Samy Adham

Agbabian Associates

El Segundo, CA

Russell Brown

Clemson University

Clemson, SC

Robert Ewing

Ewing & Associates

Rancho Palos Verdes, CA

Mike Hammons-'- US Army Corps of Engineers

Vicksburg, MS

Gilbert Hegemier

Univ. of CA-San Diego

La Jolla, CA

Richard Klingner

Univ. of Texas-Austin

Austin, TX

James L. Noland

Atkinson-Noland & Assoc.

Boulder, CO

M.J.N. Priestley

Univ. of CA-San Diego

La Jolla, CA

P.B. Shing

University of Colorado

Boulder, CO

New Member

571
TABLE 2

RESEARCH TASKS

Subject Researcher

Grouted Hollow-Unit Clay & Concrete

Masonry in Compression

R. Atkinson

Grouted Hollow-Unit Concrete Masonry

Under Combined Compression & Flexure

A. Hamid

Grouted Hollow-Unit Clay Masonry Under

Combined Compression & Flexure

R. Brown

Masonry Test Methods for Limit State

Design

R. Atkinson

M. Hammons

Nonlinear Force-Displacement Models of

Reinforced Masonry Components & Systems

R. Englekirk

Nonlinear Finite Element Models for

Analysis of Reinforced Masonry

R. Ewing

Nonlinear Lumped Parameter Models for

Dynamic Response Analysis of Reinforced

Masonry Systems

J. Kariotis

Nonlinear Models of Floor Diaphragms

Nonlinear Models of Reinforced Masonry

Walls Acting Out-of-Plane

M. Porter

R. Mayes

Experimental Study of One-Story

Reinforced Masonry Shear Walls

B. Shing

Experimental Study of Two-Story

Reinforced Masonry Shear Walls

R. Klingner

Experimental Study of Three-Story

Reinforced Masonry Shear Walls and

Development of Automated Feed-Back Test

Method for Stiff Structures

G. Hegemier

F. Seible

Experimental Study, Static & Dynamic,

of Out-of-Plane Behavior of Reinforced

Concrete Masonry Walls

A. Hamid

S . Adham

Experimental Study, Static & Dynamic,

of Out-of-Plane Behavior of Reinforced

Clay Masonry Walls

Research Tasks to be completed

R. Mayes

572
Subject Researcher

Experimental Study, Static & Dynamic, N. Priestley

of Flanged Reinforced Masonry Walls

*Analytical Study Based on Prior Tests, G. Hegemier

of Wall-Floor Intersections

Experimental Study (Preliminary) of L. Tulin

Reinforcing Bar Bond & Splices in

Grouted Masonry

*Experimental Study (continuation) of

Reinforcing Bar Bond and Splices in

Grouted Masonry

Experimental Study, Static & Dynamic D. Abrams

Behavior of Scale Model Three-Story

Reinforced Masonry Buildings

R. Atkinson

M. Mammons

Development of Reliability-Based

Limit-State Design Methodology for

Masonry Buildings

G. Hart

Development of Reliability Factors for

Masonry Materials, Components, & Systems

G. Hart

Design of Five-Story Reinforced Masonry

Research Building

J. Kariotis

Design and Construction of Large

Specimen Research Building

G. Hegemier

F. Seible

Experimental Study of Nonlinear

Behavior of a Five-Story Building

Under Cyclic Lateral Loads

G. Hegemier

F. Seible

Design & Criteria Recommendation for

Limit State Design and Documentation and

Technology Documentation

J. Noland

Overall Program Coordination J. Noland

*Research Tasks to be completed

573
TASK REPORTS

Task No . Author/s - Title

1-1-1: Atkinson and Kingsley, Comparison of the Behavior of Clay & Concrete

Masonry in Compression , September 1985.

1.2(a)-l: Hamid A.A. , Assis, G.F., Harris, H.G., Material Models for Grouted Block

Masonry , August 1988

1.2(a)-2: Assis, G.F., Hamid A.A., Harris, H.G., Material Models for Grouted Block

Masonry , August 198 9

1.2(b)-l: Young, J.M., Brown, R.H., Compressive Stress Distribution of Grouted

Hollow Clay Masonry Under Strain Gradient , May 1988

2.1-1: Hart, G. and Basharkhah, M. , Slender Wall Structural Engineering Analysis

Computer Program (Shwall, Version 1.01), September 1987.

2.1-2: Hart, G. and Basharkhah, M. , Shear Wall Structural Engineering Analysis

Computer Program (Shwall, Version 1.01). September 1987.

2.1-3: Nakaki, D. & Hart, G., Uplifting Response of Structures Subjected to

Earthquake Motions , August 1987.

2.1-4: Hart, G., Sajjad, N. , and Basharkhah, Inelastic Column Analysis Computer

Program (INCAP, Version 1.01), March 1988.

2.1-5: Hong, W.K., Hart, G.C., Englekirk, R. E ., Force-Deflection Evaluation and

Models for University of Colorado Flexural Walls , December 1989.

2.1-6: Hart, G.C. Jaw, J.W., Low, Y.K., SCM Model for University of Colorado

Flexural Walls , December 1989.

2.1- 7: Hart, G., Sajjad, N. , Basharkhah, M. , Inelastic Masonry Flexural Shear

Wall Analysis Computer Program, February, 1990.

2.2- 1: Ewing, R. , El-Mustapha, A., Kariotis, J., FEM/I - A Finite Element

Computer Program for the Nonlinear Static Analysis of Reinforced Masonry

Building Components , December 1987 (Revised June 1990).

2.3- 1: Ewing, R. ; Kariotis, J.; El-Mustapha, A., LPM/I, A Computer Program for

the Nonlinear, Dynamic Analysis of Lumped Parameter Models , August, 1987.

2.3-2: Kariotis, J., El-Mustapha, A., Ewing, R., Influence of Foundation Model

on the Uplifting of Structures , July 1988.

2.3-3: Kariotis, J., Rahman, M., El-Mustapha, A., Investigation of Current

Seismic Design Provisions for Reinforced Masonry Shear Walls , January

1990.

574
3.1 (a) -1: Scrivener, J., Summary of Findings of Cyclic Tests on Masonry Piers , June

1986. , .

3.1(b)-l: Seible, F. and LaRovere, H., Summary of Pseudo Dynamic Testing , February

1987.

3.1{c)-l: Merryman, K. , Leiva, G. , Antrobus, N., Klingner, R., In-Plane Seismic

Resistance of Two-Story Concrete Masonry Coupled Shear Walls , September

1989.

3.2(a): Hamid, A., Abboud, B., Farah, M. , Hatem, K. , Harris, H., Response of

Reinforced Block Masonry Walls to Out-of-Plane Static Loads , September

1989.

3.2(b)-l: Agbabian, M. , Adham, S., Masri, S., Avanessian, V. Traina, Out-of-Plane

Dynamic Testing of Concrete Masonry Walls , Volumes 1 and 2, July 1989.

4.1-1: Limin, H., Priestley, N., Seismic Behavior of Flanged Masonry Shear

Walls , May 1988.

4.2-1: Hegemier, G. Murakami, H. , On the Behavior of Floor-to-Wall Intersections

in Concrete Masonry Construction: Part I: Experimental

4.2-2: Hegemier, G. , Murakami, H. , On the Behavior of Floor-to-Wall

Intersections in Concrete Masonry Construction: Part II: Theoretical

5.1- 1: Porter, M., Sabri, A., Plank Diaphragm Characteristics , July 1990.

5.2- 1: Porter, M. , Yeomans, F., Johns, A., Assembly of Existing Diaphragm Data , July 19 90.

6.2-1: Scrivener, J., Bond of Reinforcement in Grouted Hollow-Unit Masonry: A

State-of-the-Art , June 1986.

6.2-2: Soric, Z. and Tulin, L. , Bond Splices in Reinforced Masonry , August 1987.

7.1-1 Paulson, T., Abrams, D., Measured Inelastic Response of Reinforced

Masonry Building Structures to Earthquake Motions , October, 1990.

8.1-1: Hart, G., A Limit State Design Method for Reinforced Masonry , June 1988.

9.1- 1: Kariotis, J.C., Johnson, A.W., Design of Reinforced Masonry Research

Building , September, 1987

9.2-1: Seible, F., Report on Large Structures Testing Facilities in Japan , September 1985.

9.2.2: Seible, F., Design and Construction of the Charles Lee Powell Structural

Systems Laboratory , November 1986.

9.2.3: Seible, F., The Japanese Five-Story Full Scale Reinforced Masonry

Building Test , January 1988.

575
11.1-1: TCCMAR, Summary Report: U.S. Coordinated Program for Masonry Building

Research , September 1985 to August 1986.

11.1-2: Status Report: U.S. Coordinated Program for Masonry Building Research , November 1988.

1 Task Reports are available through

the Earthquake Engineering Research

Library, Berkeley, CA

576
Appendix

Task Committees A-J Reports

Summary of Task Committee Workshop Reports

Report of Task Committee on

(A) STRONG-MOTION INSTRUMENT ARRAYS AND DATA

Date

:

•May 15, 1991

Place

:

Public Works Research Institute (Tsukuba Science City)

Attendees

:

Japan Side S. Noda (Chairman)

K. Ohtani

K. Tamura

(PHRI)

(NIED)

(PWRI)

U.S. Side A.G. Brady (Chairman) (USGS)

MISSION STATEMENT

1 . Objective

:

The objective of the Task Committee is to coordinate the provision of

strong-motion earthquake data to researchers and to the practicing

engineering community. The Task Committee focuses on instrumentation;

recording, processing, and analyzing strong-motion data; research to ensure high-quality data dissemination; and analysis of ground motion

and structural motion in dynamic response to earthquakes.

2. Scope of Work

The T/C coordinates strong-motion research, processing strong-motion

data, and disseminating information on the recorded behavior of

structures. In addition to regular exchange of data and publications,

the T/C's technical approach includes:

Planning and conducting T/C workshops and meetings, held

generally in conjunction with UJNR Joint Meetings.

Conducting related technical sessions at professional society

meetings

.

Creating procedures for the dissemination of significant

strong-motion digital data with regard for the rights and

expectations of (a) the owner, (b) the users of data, and (c)

the earthquake engineering community.

Creating an up-to-date information exchange on computer

processing procedures and related research so all users of

processed data are aware of valid ranges and practical

limitations.

Developing a research program for appropriate instrumentation

and dynamic analysis of structures with base isolation or other

response-reducing systems, under earthquake excitation.

Developing a research program using both permanent and

temporary arrays and data derived to study site response.

579
3 . Accomplishments

* Catalogs of strong-motion earthquake records observed both in

the United States and Japan are being exchanged. U.S. data are published in the "Seismic Engineering Program Report", and the

Japanese data are published in "Strong-Motion Earthquake

Records in Japan". Digital data is exchanged on tape,

floppies, and CDROMs. Many other publications were also

exchanged

.

* Data tapes on the Loma Prieta earthquake of October 17, 1989,

processed by USGS and CDMG, were delivered to Japanese side.

* Disks containing data on the Chiba-Toho-Oki earthquake of

December 17, 1987, processed by the National Research Center

for Disaster Prevention, were delivered to U.S. side.

* The Japanese Technical Study Team for the Loma Prieta

Earthquake visited northern California, USA, between Nov. 16 &

29, 1989 to study the effects. The team and the U.S. hosts

included many panel members.

* Dr. Sasaki visited the U.S. during March 20-29, 1990, visiting

U.C. Berkeley, USGS, Stanford, CALTRANS, CDMG; Loma Prieta

Earthquake effects in San Francisco, Oakland, and epicentral

area; and Washington (AID, NIST, NSF ) . Panel members hosted

throughout; T/C A contacts included A. G. Brady and R. D.

Borcherdt, of USGS, and A. G. Franklin of USAEWES.

(Other Panel members included J. Gates, R. Hanson, H.S. Lew and

N. Raufaste).

* Mr. Keiichi Tamura of PWRI was a Visiting Scholor at Stanford

University for one year ending October 1990 working on strong

motion instrument arrays and data.

* Mr. Susumu lai of PHRI visited California for one month in

March 1991 for joint research with California Division of Mines

and Geology on recording and processing of strong earthquake

motion

.

4 . Future Plans

* The T/C Chairmen will seek funding sources and finalize the

planning for a 3rd Workshop on Processing Strong Motion Records

in spring of 1992.

* Coordinate the analysis of records from arrays: downhole 3-D

arrays (geophysics and earthquake engineering) and structural

arrays (structural engineering).

* The T/C recognizes the efforts of the International Working

Group on the Effects of Surface Geology and encourages wider

participation by Panel members.

580
Information Exchange

* Will be performed within the domestic community through

technical meetings, conferences, seminars, and workshops.

* Will be performed through participation in post-earthquake

field investigations with concentration in areas where

strong-motion stations have recorded data.

* Will be performed with the counterpart side through T/C

workshops; visitor exchanges, seminars, lectures; and

participation in annual Joint Panel meetings.

* The most crucial information exchange is rapid description of

recovered strong motion (and subsequently, digital

strong-motion data on tape or disk) from significant

earthquakes

.

Impact

* Most managers who process strong-motion recordings use techniques based, or closely associated with, the work of this

T/C and its members. Foreign networks fitting this category

include

:

AUSTRALIA - Australian Seismological Center, Bureau of Mineral

Resources, PC Box 378, Canberra ACT 2601.

CANADA - Pacific Geoscience Center, PC Box 6000, Sidney, BC.

V8L 4B2.

ITALY - The Italian National Commission for Research and

Development on Alternative Energy Sources (ENEA) , Rome.

MEXICO - Japan-Mexico Earthquake and Disaster Prevention

Center, Mexico City.

PERU - Japan-Peru Earthquake and Disaster Prevention Center,

Lima

.

TAIWAN - SMART Array

TURKEY - University of Bogazici , 80815 Bebek, Istanbul.

. YUGOSLAVIA - Institute of Earthquake Engineering and

Engineering Seismology, Skopje.

* Use of seismological and strong-motion data, in convenient

form, is facilitated in countries other than where recorded.

* Dissemination of digital data from significant earthquakes,

taking into account the conditions of the 3rd paragraph in the

Scope of Work, increases the research data base; application of

this data in design practice aids in reduction of earthquake

hazards

.

581
7. Barriers

Lack of adequate funding for T/C joint endeavors.

Establishing joint research efforts and maintaining close

interaction between participants are difficult unless visiting

research appointments are established. (Funding and staff

availability are often barriers).

582
Report of Task Committee on

(B) LARGE-SCALE TESTING PROGRAM

Date : May 15, 1991 . .

Place : Public Works Research Institute

Tsukuba , Japan

Attendees: Japan Side - K. Ohtani (Chairman) (NIED)

H. Yamanouchi (BRI)

S. Nakata (BRI)

Y. Koga (PWRI)

U.S. Side - H. S. Lew (Chairman) (NIST)

MISSION STATEMENT

1. Objective

The objective of the Task Committee is to develop performance data of

full-scale dynamic properties of structures to better resist wind and seismic

loads through both laboratory testing of prototype structures and field

testing of structures in situ. Improved full-scale test data will validate

the results of small-scale model tests and substantiate the results of

computer analyses of structural behaviour.

2 . Scope of Work

The Task Committee develops its research agenda in coordination with

other appropriate T/Cs, such as T/C's "C" and "D", for full-scale evaluation

of buildings, and other structures except bridges.

* Plans and conducts workshops and joint meetings to identify research

topics and develops joint research programs.

* Coordinates research projects carried out by various laboratories in the

U.S. and Japan. Facilitates publication of research results and

implementation of findings in codes and standards.

* Facilitates exchange of research personnel, technical information and

available testing facilities.

3. Accomplishments

* The National Science Foundation (NSF) has funded the third phase (5-story

structural test) of the U.S. -side research of the U.S. -Japan Coordinated

Program for Masonry Building Research. The testing of five-story

full-scale masonry structure will be carried out at the University of

California, San Diego during 1991-1993.

* The U.S. Technical Coordinating Committee on Masonry Research was held in

August 1990.

583
The 6th Joint Technical Coordinating Committee on Masonry Research

(JTCCMAR) was held in August 1990 at Seattle, Washington. Progress on individual research projects were presented by principal investigators

and future coordination of the Joint Program was discussed.

The first phase of the U.S. -Japan Joint Technical Coordinating Committee

on Precast Seismic Structural Systems Program (PRESSS) was held in

November 1990 At San Diego, California.

The Japan PRESSS Research program is continuing at several institutions

including BRI , University of Tokyo and Yokohama National University. The

Japan TCC Meetings were held during the past fiscal year.

Future Plans

The second Joint Technical Coordinating Committee (JTCC) on PRESSS will

be held in Japan in October 1991. The U.S. TCC on PRESSS will meet prior

to the second JTCC on PRESSS in June 1991.

In support of International Decade for Natural Disaster Reduction (IDNDR)

program, techniques will be explored to disseminate findings of Joint

Research Projects carried out under the auspices of T/C "B" to countries

with high seismic risks

.

A workshop on Composite, Mixed an Hybrid Structural Systems is being

planned to be held in summer of 1992 in the U.S.

Information exchange of following topics between appropriate U.S. and

Japan Organizations will be continued;

a. Application of high strength construction materials, such as concrete

and steel, to structures in high seismic zones,

b. Testings of large-scale structures other than buildings and bridges.

Information Exchange

Technical reports and research documentation were exchanged with

participating organizations in both countries.

Impact

The results of the past joint research projects on reinforced concrete

and steel structures have been published in both U.S. and Japanese

technical journals. A number of key findings have been incorporated into

U.S. and Japanese building codes and design guidelines.

As a results of joint research projects, the state of practice and

understanding in earthquake engineering has been improved in both

countries. The results from the joint program above further stimulated

research in both countries.

Facilitate joint research projects utilizing available testing facilities

in both countries.

584
Barriers

There are no major technical barriers for the attainment of the Task

Committee mission. However, different building practices and needs

require careful planning for effective joint research projects.

Securing funding to support exchange of research personnel is difficult

in both countries. Funding agencies of both countries should be informed

about the technical benefits resulting from the joint panel activities

and encouraged to participate researcher exchange and joint research

programs

.

585
First Meeting of the Joint Technical Coordinating Committee

on Precast Seismic Structural Systems (PRESSS)

San Diego, USA

Nov. 29, 30, Dec.1, 1990

DRAFT RESOLUTIONS

1. The first meeting of JTCC-PRESSS was held at the University of California

San Diego at the Whispering Palms Country Club, San Diego on

November 29 and 30, and December 1, 1990. The meeting was attended

by 22 researchers and 5 observers from Japan, and 15 researchers and 7

observers from the USA.

2. At the meeting there was a successful exchange of information and ideas

between the researchers from both countries on the subjects of precast

structural system behavior, behavior of assemblies, testing procedures,

analytical modeling methods and seismic design procedures.

3. The need for, and importance of, continuing experimental and analytical

research on PRESSS was reaffirmed.

4. The common goal of Japan and USA research efforts is to develop design

guidelines for PRESSS.

5. Continued cooperation and interchange of design concepts will be carried

out based on the 'test building' concept.

6. The Hachinohe ^Q6Q N-S record amplified to 0.33 g peak ground

acceleration will be used as the base record for comparing dynamic

response.

7. The US side will develop at least one shear wall building concept and at

least one ductile joint concept and report results of design studies at the

next US/Japan meeting.

8. The US side will carry out dynamic inelastic time-history analyses of the

test-design buildings and report results at the next US/Japan meeting.

9. The US side will carry out inelastic collapse mechanism analyses of its design buildings and report at the next US/Japan meeting.

10. The Japan side will carry out dynamic inelastic time-history analyses of its Case II and Case Hi buildings and report at the next US/Japan meeting.

586
Page 2

Resolutions

11. The Japan side will carry out collapse analyses and dynamic inelastic

time-history analyses of the US frame test building to ensure correct

calibration of analysis methods between the two countries.

12. Both sides will investigate correlation between full structural simulations

and simplified structural condensation models for dynamic inelastic

analyses.

13. Both sides will continue to develop and classify new connection concepts.

The US and Japan connection groups will interact and agree on the format

for presenting connections for the data base report. Agreement on the

format will be reached by May 1 , 1 991

.

14. The completed data base report will be finalized by 1992.

15. The Japan side has made considerable progress on its new design

guidelines. The US side will prepare and present a detailed framework for

its recommendatiohs for the 1991 meeting.

16. The two sides will exchange information on test load patterns to simulate

seismic response of connections and members and investigate the

practicality of adopting a common load pattern for experimental research.

17. The technical coaimunication which has carried out between Japan and

US PRESSS teanls has been valuable and should be continued. There

should be a continljation of the exchange of researchers so as to facilitate

dialogue on concepts and design.

18. The next meeting will be held in Japan on October 30, 31st and November

1,1991.

587
REPORT OF TASK COMMITTEE ON

(C) REPAIR AND RETROFIT OF EXISTING STRUCTURES

Date May 15, 1991

Place Public Works Research Institute

Tsukuba, JAPAN

Attendees

:

Japan Side - M. Hirosawa (Chairman) (BRI)

(BRI)

(BRI)

(BRI)

(BRI)

(MRI)

S. Nakata

S . Okamoto

H. Yamanouchi

T. Murata

S. Ohtsuka

U.S. Side S.C. Liu (Acting chairman) (NSF)

M. Shinozuka (NCEER)

T. Lew (NCEL)

A. Smith (S.U.)

M. Celebi (USGS)

D. Eagling (DOE)

MISSION STATEMENT

1 . Objective

The objectives of the Task Committee are to encourage research and

provide technology transfer activities in materials, components, and

whole buildings and structural systems related to the repair and

retrofit of existing buildings and other structures. The scope will

encompass studies related to materials of construction, structural

components as well as prototype structural systems.

2. Scope-of-Work

The T/C work includes information exchanges, planning, and hosting

workshops and seminars. Workshops are generally held in conjunction

with the annual Panel's meeting. They provide T/C members with an opportunity to review related research and to direct plans for future

research, share and implement results. The T/C conducts related

technical sessions at professional society's meetings. The T/C studies new materials and methods to accomplish repair and retrofit operations

on buildings e.g., fiberglass reinforced plastics, high performance

concrete, post-installed anchor bolts, carbon-fibre wire, structural

adhesives, and exploits use of automation and robotics for repair and

retrofit. Work focuses on coordinating research projects in U.S. and

Japan to minimize duplication and to maximize benefits.

3. Accomplishments

* Status reports were delivered on recent related NSF initiatives:

588
o QUANTITATIVE NONDESTRUCTIVE EVALUATION FOR LARGE STRUCTURAL SYSTEMS

This initiative was issued in June, 1990. The responses from the

U.S. research communities were overwhelming. Plans are underway for

the second year announcement; interdisciplinary research teams are encouraged. The following four projects were selected for support

in the first round:

1. Fiber optic sensors for strength evaluation and early warning of

impending failure in structural components.

2. Nondestructive methods in assessing damage to reinforced concrete

structures

.

3. Micromagnetic surface studies of materials for nondestructive

evaluation.

4. Fundamentals of embedded optical fibers in structural concrete.

Details of these four projects and a description of the initiative are included in the proceedings of the 23rd UJNR Joint Meeting in Tsukuba,

Japan.

o REPAIR AND REHBILITATION RESEARCH FOR THE SEISMIC RESISTANCE OF

STRUCTURES

In early 1990, NSF announced a new 5-year research program that focuses

on a coordinated research effort in the seismic repair and

rehabilitation of existing structures. The general goals of the program

are to : (a) initiate a concentrated effort to integrate the existing

body of knowledge on the repair and rehabilitation of buildings, (b)

carry out additional research as required to develop practical

engineering measures for immediate implementation, and (c) to promote

closer ties between research and practice by facilitating transfer of

knowledge and techniques through the development of design guidelines

applicable to U.S. practice.

During the first year of the research program, FY 1990, NSF funded a total of nine projects.

- Strengthening and repair of nonductile reinfo.rced concrete frames

using external steel jackets and plates.

- Seismic retrofit with energy dissipators.

- Evaluation of the strength, stiffness, and ductility of older steel

frame structures.

- Experimental evaluation of slab-column frame and the use of the

ductile steel.

- Behavior of clay brick and concrete masonry infill walls.

- Repair and rehabilitation research program coordination.

589
In addition, cooperation and coordination efforts were initiated with

other on-going seismic repair and rehabilitation programs, conducted by

the National Center for Earthquake Engineering Research (NCEER) , the

Earthquake Engineering Research Institute (EERI) through its Existing

Building Committee, the Federal Emergency Management Agency (FEMA) , the

Building and Fire Research Laboratory (NIST), The California Seismic

Safety Commission (CSSC), and other organizations. The first

coordination meeting of FY 1991 was held in Salt Lake City on February

13, 1991. •

* 3-1 day ASCE seminars on Rehabilitation, Planning and Structural

Repair of Concrete and Masonry were held in, Oakland, CA (3/90); Boston

(4/90); and Washington D.C. (5/90).

* A workshop on Nondestructive Evaluation of Civil Structures and

Materials was held in October, 1990, in Boulder, Colorado. 406 pages

hardcover proceedings are available from the University of 'Colorado,

Department of Civil Engineering. Participants came from many

disciplines of engineering in academics and industries.

There were five working groups:

- concrete

- masonry

- steel

- timber

- advanced equipment and techniques.

* Proceeding of the Workshop on Evaluation, Repair, and Retrofit of

Structures , (NISTIR 4515), held at NIST in May 1990, was published by

NIST and distributed at the 23rd UJNR Joint Meeting in Tsukuba, Japan.

Copies of the proceedings were distributed also to authors, workshop

participants, and other researchers and practicers in the U.S. 22

papers were presented at the NSF sponsored workshop, of which 17 papers

were on repair and retrofit of structures. 24 delegates attended, 12 were from the U.S. and 12 from Japan.

* In the U.S., FEMA published ATC 21, A Handbook for Seismic Evaluation

of Existing Buildings , and ATC 22 Techniques for Seismically

Rehabilitating Existing Buildings . The U.S. Department of Energy and

U.S. Postal Service are involved in extensive repair and retrofit

projects

.

* The Architectural Institute of Japan (AIJ) have revised its Ultimate

Strength and Deformation Capacity of Buildings in Seismic Design in

December 1990 after 13 years since the first edition. This book contain

many recent research results mainly on evaluation method of bearing

capacity and ductility of structures and structural members such as foundation structure, reinforced concrete structure, steel framed

reinforced concrete structure, steel structure, wooden structure, vessel

structure and structure with passive control.

* The Architectural Disaster Prevention Association has revised its

Guidelines for Evaluating and Strengthening Existing Reinforced Concrete

Buildings in October, 1990 after 13 years since the first edition and

the Guidelines for Evaluation of Earthquake Damage and Retrofitting

590
Methods of Buildings in March, 1991. '

. In the Guideline for Evaluating, a simplified rigorous evaluation

method for school buildings vulnerable to earthquake was added as one of

appendixes

.

. In the Guideline for strengthening, structural calculation method

to strengthen existing reinforced concrete buildings by steel panels or steel braces was newly regulated based on recent research results.

* In Japan several prefectures such as Kanagawa prefecture and

Shizuoka prefecture where a severe earthquake is anticipated to occur,

continue to discuss countermeasures after the earthquake, and in case of

Kanagawa prefecture, it will be decided to start to cultivate engineers

especially for urgent evaluation of damages applying the guidelines.

* The guidelines were presented to U.S. side at the 23rd UJNR Joint

Meeting ; the English version of the guidelines was translated by Dr. M.

Ohkubo and will be printed in March, 1992 at UCSD.

4. Future Plans

* A second round of related NSF initiatives are planned for FY 92:

o Quantitative Nondestructive Evaluation for Large Structural Systems

o Repair and Rehabilitation Research for the Seismic Resistance of

Structures

* Repair and retrofit of existing structures are being studied in

association with a possible federal initiative on adaptation due to

the global changes (MARS) program.

* New initiatives are being developed by NSF to focus on central and

eastern U.S. existing structures/buildings.

* In support of IDNDR activities, T/C "C" plans to participate in task

committee activities related to repair and retrofit of damages due

to various disasters.

* A possible workshop is being planned on the Design Approaches and

Construction Methods for Repair and Retrofit of Buildings and

Structures after 1992. The Workshop will focus on the use of newly

developed methods for repair, retrofit and evaluation of

buildings/structures

.

5. Information Exchange

* Publications of the Joint T/C "C" of "D" Workshops on Evaluation

Repair and Retrofit of Structures were distributed, to all delegates

and will be made available to design professionals through technical

presentations at national meetings of professional societies.

* A U.S. Panel (Japanese Technology Evaluation Center Panel) visited

29 sites in Japan in July, 90, to study and exchange information on construction technologies.

591
* Published workshop reports and individual research papers; they were exchanged with the U.S. and JAPAN-side members.

* Numerous papers presented at the 23rd UJNR Joint Panel Meeting are of interest to T/C "C".

6. Impact

* Contributed to development of U.S. and Japan design criteria for

repairing and strengthening reinforced concrete, steel, and other

buildings.

* T/C recommendations were incorporated into building codes and

professional practices, as illustrated by rehabilitation programs

used by the City of Los Angeles to upgrade existing masonry

buildings to resist earthquake forces.

* Influenced changes to U.S. and Japan design practices to upgrade

existing structures.

7. Barriers

* Greater progress needs be made to implement available research

results. Additional funding sources need to be identified to

support technical activities to support committee missions and to

support participate attendance at Joint Panel Meeting workshops.

Creative technical information services, including translation of

Japanese files, abstracts or entire reports are needed to improve

the effectiveness of information exchange.

592
Report of Task Committee on

(D) EVALUATION OF STRUCTURAL PERFORMANCE

Date: May 14, 1991

Place: Public Works Research Institute

Tsukuba , JAPAN

Attendees: Japan Side -

U.S. Side

s. Okamoto, Chairman (BRI)

M. Hirosawa (BRI)

S. Nakata (BRI)

H. Yamanouchi (BRI)

T. Murota (BRI)

S. Ohtsuka (MRI)

S. C. Liu, Acting chairman (NSF)

T. Lew (NCEL)

H. S . Lew (NIST)

M. Shinozuka (NCEER)

D. Eagling (DOE)

M. Celebi (USGS)

A. Smith (Stanford)

1 . Objective

The objectives of this task committee are to develop disaster mitigation

policies and programs to improve the capacity of existing structures to

resist wind and seismic forces. To provide adequate performance

evaluation, each country will coordinate development of condition

assessment, screening, and structural analysis methodologies. Structures

will be analyzed and instrumented, and then evaluated after disasters.

2. Scope of Work

Develop an inventory of "benchmark" structures that have been

analyzed and instrumented to measure wind and seismic

responses

.

Develop a uniform system for screening and analyzing wind and

seismic resistance capacity of structures in each country.

Develop sensor technology, instruementation , and "expert"

systems to provide condition assessment of existing

structures.

Evaluate masonry and precast concrete buildings, and seismic

base isolation systems; include results in each country's

catalog of "benchmark" structures for post-disaster

performance evaluation

Conduct workshops at future UJNR Joint Panel Meetings,

cooperatively with T/C "B" and "C".

593
3 . Accomplishments ' '

* The Proceedings of a Workshop on Evaluation, Repair, and

Retrofit of Structures (Joint Task Committee "C" and "D"),

held at NIST in May 1990, was published and distributed to the

authors, workshop participants, and other researchers and

practicers in the U.S.

* The U.S. Side (NIST) conducted the translation and published

the Japanese MOC documents on base isolation:

Report of the Expert Committee on Advanced Technology for

Building Structures:

Part 1, Damping Structures and Trends of Future

Technological Developments , and

Part 2, Survey Report on Framing of the Guidelines for

Technological Development of Base-Isolation

Systems

* Status reports on recent U.S. legislation were presented at

the 23rd Joint Panel Meeting:

1990 National Affordable Housing Act , Sec. 947 on Seismic

Safety Standards. Requires HUD to conduct risk assessments of

all HUD-assisted housing, and to develop or adopt aseismic

standards in accordance with Executive Order 12699. A first

report to Congress is due by May 21, 1992. the

Secretary shall consult with the Director of the Federal

Emergency Management Agency (FEMA) and may utilize the

resources under the National Earthquake Hazards Reduction

Program . . .

.

National Earthquake Hazards Reduction Program Reauthorization

Act; Public Law 101-614, Nov. 16, 1990 . Sec. 8 on Seismic

Standards. "The President shall adopt, not later than

December 1, 1994, standards for assessing and enhancing the

seismic safety of existing buildings constructed for or leased

by the Federal Government...." "The President shall ensure

the issuance, before February 1, 1993, by all Federal agencies

of final regulations required by .... Executive Order numbered

12699, issued January 5, 1990."

* Information was exchanged on the activities of each country in

the International Decade on Natural Disaster Reduction

( IDNDR) . Several U.S. Agencies are participating on the

Federal Subcommittee on Natural Disaster Reduction (SNDR),

under the Committee on Earth and Environmental Sciences

(CEES)

.

A report entitled: Hazard-Proofing the Nation, Strategy to

Reduce Natural Disasters was drafted by SNDR, and will be

submitted to all Federal Agencies by June 1991. Included in

the research and implementation strategies are risk assessment

and mitigation programs. When completed, copies will be

mailed to the Japanese members.

594
* status reports of several FEMA - ATC projects were discussed,

including those on seismic risk assessment of existing

structures. The U.S. Dept. of Energy and U.S. Postal Service

are actively conducting risk assessment studies, and planning

or conducting extensive repair and retrofit projects.

4. Future Plans

* Develop a reporting system and data collection • form for

building inventory and performance standard, and compile a computer data file for exchange between countries. Report at

each Joint Panel Meeting. A special subcommittee shall be

established to formulate a working plan for "benchmark"

structures

.

* Continue to encourage participation by private industry,

consulting engineering firms, universities, funding agencies

and State and local government agencies involved in

instrumentation, evaluation, and condition assessment of

existing buildings.

* Monitor development of sensor technology and instrumentation

of structures for condition assessment, and to measure

long-term aging effects on building components.

* Participate in coordination of Joint Panel activities in the

International Decade for Natural Disaster Reduction (IDNDR),

pertaining to structural performance evaluation.

* A workshop on Structural Performance Evaluation Methodologies

will be discussed to be held in the U.S. after 1992. The

workshop will place emphasis on a synthesis of current

methodologies used in both sides and comparative studies of

these methodologies.

5 . Information Exchange

* Technical documents exchanged, and pertinent Japanese papers

delivered at 23rd Joint Panel Meeting.

* Japense side will send to the U.S. T/C Chairman a copy of the

Final Revision of "Design Guidelines for Medium Rise

Reinforced Masonry Buildings" developed by TCCMAR-Japan . U.S.

Side will explore feasibility of translating this manuscript

into English. Summary report of the design guidelines was printed in the Pre-Proceedings of *22nd Joint Meeting (3-3

Okamoto )

.

6 . Impact

* Enables the development of revised standards and building code

provisions to prevent destruction from high winds and

earthquakes.

595
* Provides engineers with the capability to conduct condition

assessment of existing structures for the purpose of

developing cost effective repair, retrofit, strengthening or demolition programs.

* Will increase the confidence level of the public, regulators,

policy makers and engineers, to be asured that the majority of

major structures will perform adequately to resist high wind

and earthquake forces.

7 . Barriers

* Difficulty in identifying priorities and inadequate resources

to document load carrying capacity of existing buildings

subjected to highly probable extreme winds and earthquakes.

Difficulty in developing probability risk maps for tornadoes,

typhoons, and hurricanes.

* Limited resources for instrumenting buildings, for conducting

post-disaster evaluations, and for participating in foreign

travel for coordinaton of activities.

596
Report of Task Committee on

(E) GROUND MOTION AND SEISMIC DESIGN FORCE

Date: May 14, 1991

Place: Public Works Research Institute

Tsukuba

Japan

Attendees: Japan Side- K. Kawashima (Chairman) (PWRI)

T. Dohi (GSI)

Y. Kitagawa (BRI)

K. Ohtani (NIED)

T. Uwabe (PHRI)

H. Sugita (PWRI)

K. Tamura (PWRI)

G. A. MacRae (PWRI)

U.S. Side- S. T. Algermissen (Chairman) (USGS)

A. G. Brady (USGS)

J. R. Hill (DOE)

MISSION STATEMENT

1. Objectives

The objectives of this Task Committee are to develop techniques and

exchange information and experiences on: 1) engineering characterization of

ground motion, 2) mapping techniques for displaying ground motion data, 3)

microzonation, 4) siting and design applications, and 5) earthquake hazard

and risk assessment.

2. Scope of Work

Plan and conduct cooperative research, T/C workshops and meetings. These

meetings and workshops are generally held in conjunction with the annual

Panel Meeting on Wind and Seismic Effects. Activities include:

- Conduct cooperative research programs

- Exchange research personnel

- Exchange information on a timely basis

597
3. Accomplishments

All final preparations have been made for the workshop "Hazard/Rislc

Assessment and Design Earthquake Loading" to be held on May 28-29, 1991 at

the Public Works Research Institute, Tsukuba, Japan. Fourteen papers have

been accepted for presentation.

4. Future Plan

i) The workshop "Hazard/Risk Assessment and Design Earthquake Loading" will

be held on May 28 and 29, 1991. A study tour will precede the workshop.

ii) Proceedings for the workshop will be published and distributed to the

panel members and Interested parties.

ill) The Task Committee co-chairmen will develop a system for the exchange of

appropriate technical material.

5. Information Exchange

Technical reports and research documentation were exchanged with

participating organizations in both countries.

6. Impact

Task committee efforts Influence the national seismic codes concerning Input

seismic forces.

7. Barriers

Lack of funding for travel to conduct workshops and related meetings has

been the principal barrier In the past.

598
Report of Task Committee on

(F) DISASTER PREVENTION METHODS FOR LIFELINE SYSTEMS

Date: May 15, 1991

Place: Public Works Research Institute, Tsukuba, Japan

Attendees : Japan Side - Y. Sasaki (Chairman) (PWRI)

(BRI)

(PHRI)

(PWRI)

(PWRI)

T. Murota

T. Uwabe

K. Toklda

H. Suglta (Observer)

U.S. Side J. Cooper (Acting Chairman)

M. Celebl

J. Hill

(FHWA)

(USGS)

(DOE)

(NCEL)

(LBL)

T. Lew

D. Eagllng

MISSION STATEMENT

1. Objective

The objectives of Task Committee "F" are to improve 1) the behavior of

lifeline systems during earthquakes and 2) engineering and other seismic

countermeasures such as damage estimation techniques and inspection procedures.

These technologies provide users with improvements to existing standards of

practice for safety and serviceability of lifeline systems. Lifeline systems such

as gas, oil, water, and sewage pipelines; and power, communication, and

transportation systems are crucial to the survival and health of a city or community. Earthquakes affecting such systems cause severe social and economic

disruptions to communities and cause human suffering to the residents.

2. Scope of Work

» Plan and conduct T/C workshops and meetings; they generally are held in

conjunction with the annual UJNR Joint Meeting. The purpose is to

exchange the state-of-the-art knowledge and practice, and to identify

cooperatively key opportunities for exchange and studies which effectively

Implement research results.

* Conduct related technical sessions at professional society meetings.

599
» Create a research program on a selected lifeline system (similar in

concept to the Large-Scale Testing Program). For example, through joint

demonstration, produce criteria for emergency design and loss evaluation

and on-line monitoring systems for lifeline operation.

* Conduct special activities such as translating, printing, and distributing

MOC PWRI's publication Manual for Repair Methods for Civil Engineering

Structures Damaged by Earthquake.

« Develop performance standards and a manual on repair, restoration, and

retrofit of lifelines.

3. Accomplishments

» Researchers from both sides have contributed technical papers on lifeline

systems at the 23rd joint panel meeting.

• Researchers from both sides have participated in the post-event field

investigations on the collapse and damage of lifeline structures in the

San Francisco Bay area following the Loma Prieta earthquake in October

1989. A number of in-depth studies on earthquake behavior, safety, repair

and strengthening of the lifeline structures have been undertaken.

* Encourage active perticipation in promoting the goals of IDNDR,

4. Future Plan

Conduct the 4th Joint Workshop on Earthquake Disaster Prevention for

Lifeline Systems on August 19-21, 1991 at Los Angels.

* Encourage the identification of existing specifications used for the

seismic design and retrofit of lifeline structures in both countries and

encourage the development of criteria for those lifeline structures

where insufficient design guidance exists.

« Continue observation of instrumented buried pipes in Parkfield,

California, and Chiba, Japan and exchange data.

5. Information Exchange

• Domestic community: through technical meetings, conference sessions,

seminars, and workshops.

Between U.S. and Japan: through T/C workshops and annual Joint Panel

Meetings.

600
Visits by researchers and participation in technical meetings sponsored

by organizations such as NCEER, EERI and ASCE.

» Short-term (3 to 12 months) researcher exchanges to participate in

onrgoing lifeline research.

Impact

Contributed to increased field engineers' technical knowledge on repair

methods, through seminars with lectures by T/C members using the MOC

Manual in nine major cities in Japan, including Tokyo and Osaka; more than

two thousand participants attended seminars.

* Contributed to the assessment of design, construction, and operating

standards for lifeline earthquake engineering (TCLEE/ASCE) via T/C

members participation in the committee work.

Barriers

» Insufficient funds to conduct annual Task Committee Workshops and to

publish proceedings and reports.

Lack of participation by U.S. industrial organizations and utilities in

research or in implementation of findings due to a reluctance to adopt

new technologies and practices.

601
TASK COMMITTEE REPORT

(G) Passive, Active, and Hybrid Control Systems

Date May 15 1991

Place Public. Works Research Institute, Tsukuba, Japan

Attendees

Japan Side

K.

S«

H«

M.

G.

Kitagawa (Chairman)

Kawashima

Yokoyama

Nakata

Yamanouchi

Ozaki

MacRae

(BRI

)

( PWRI

)

( PWRI

)

(BRI

)

(BRI ) (Chiba

( PWRI

)

University)

U.S. Side S.

H.

Mo

H.

C. Liu (Chairman)

S . Lew

Shinozuka

Allison Smith

(NSF)

(NIST)

(NCEER, Buffalo)

(Stanford University)

MISSION STATEMENT

( 1 ) Objective

The objectives of T/C "G" are to: (1) develop research plans in

passive, active, and hybrid control of structures and equipment, (2)

implement such control techniques as computerized motion reduction or

modification devices/systems in actual design and construction of

structural engineering facilities under seismic or wind environments,

(3) promote U.S. -Japan cooperation in structural control research,

(4) bring together governmental, academic, and industrial participants in joint pursuit of such efforts, and (5) contribute to IDNDR

by organizing joint research and other technical activities in structural control on the basis of international cooperation.

( 2 ) Scope of Work

The UJNR T/C "G" works closely with other organizations to provide

the leadership in control research efforts by facilitating the exchange of technical data and information through the established UJNR

mechanisms. Work includes:

Providing technical assistance, consultation and coordination to UJNR affiliated research agencies in the initiation,

development, and execution of their programs in structural

control research.

Promoting joint government-university- industry collaborative

efforts to facilitate technology transfer and practical

implementation

.

Sponsoring and conducting multidisciplinary workshops and

602
meetings to identify key areas of research and opportunities

for cooperation, and to exchange new knowledge and experience in practice.

Developing promotional and demonstrative activities to

stimulate public awareness and interest in this emerging

field of research.

Establishing performance standards, design specifications,

guidelines and code recommendations for application in new

construction as well as retrofit/rehabilitation of existing

structures with control systems.

Initiating research in new intelligent materials, high-tech

sensors, optimal control systems design, and laboratory and

field experiments of prototype and full-scale structures.

( 3 ) Accomplishments

The U.S. Panel on Structural Control Research interacted

actively with its counterpart Japanese Panel in exchanging

technical information and developing joint research plans.

An "International Workshop on Intelligent Structures" was

held in July 1990 at the Academia Sinica, Taipei, Taiwan.

The Workshop was organized by S.C. Liu and K.P. Chong - both

UJNR panel members - and J.C. Li of the National Central

University in Taiwan, and was well attended by researchers

from the U.S., Japan, Korea, Italy, and other countries.

The major theme areas of the Workshop are: (a) Sensing and

Monitoring, (b) Structural Control, and (c) Intelligent

Systems. The workshop proceedings (460pp. hard bound) was

published by Elsevier Applied Sciences (London) including

technical contributions in the development, application, and

modification of active/hybrid control theories and technologies for civil engineering structures.

A "U.S. National Workshop on Structural Control Research"

was held at the University of Southern California, Los

Angeles, professors George W. Housner and Sani Masri. UJNR

panel members S. C. Liu and K. P. Chong actively participated in the workshop's technical activities. The workshop

identified four specificareas related to earthquake safety

in it's technical program.

They are: (a) Analytical research, (b) Experimental research, (c) Building applications, and (d) Non-building

applications. Proceedings of the Workshop is in press and

will be available by the Summer of 1991.

Workshop on Sensors on Signal Processing for Structural

Control was held at NSF, February 25 and 26 1991. Proceedings which include research needs are being distributed by

Lehigh University.

603
( 4 ) Future Plans

A "U.S. -Japan Workshop on Passive and Active Control of

Bridges Against Earthquakes" sponsored by T/C "G" is scheduled for September 4 and 5, 1991 at NCEER, Buffalo, New

York. A study tour to inspect base-isolated bridges across

the U.S. will be conducted by participants prior to and

following the workshop (workshop co-chairmen: Ian

Buckle/NCEER and K. Kawashima/PWRI )

.

A major multi-year research initiative on "Structural Control Research for Performance, Safety and Hazard

Mitigation" is being developed by NSF. The program encourages team research projects and lays out a multidisciplinary

research agenda to achieve the following objectives:

. encourage innovations in passive, active, and hybrid

controlKsystems including the adaptation of existing

technology.

. advance sensor and actuator technology and signalprocessing techniques and their applications.

. develop intelligent- robots, other devices, advanced

computer systems and control algorithms for structural

safety

.

. develop innovative systems for energy absorption,

addeddamping, and variable stiffness.

. study, the robustness and system reliability of control

systems.

develop guidelines, standards and practical approachesand techniques for design, fabrication and field

installation.

. investigate emerging technology subjects in innovativestructural systems, and smart materials, sensors,

anddevices for detection, sensing, monitoring, and

diagnosis

.

Joint research projects on hybrid control of bridges are

beingdeveloped by PWRI, NIST, FHWA, NCEER, and University of

California, Irvine.

Joint research projects on active and hybrid control of

buildings are being developed by BRI, NCEER and University

of California, Irvine.

Develop specific work plans for coming year (1991-1992) and

forthe next five years.

In cooperation with T/C "J", continue to support development

and implementation of base-isolation technology for bridge

604
structure.

( 5 ) Information Exchanges

Technical materials on structural response and control systems were

exchanged

.

( 6 ) Impact

The T/C "G" has stimulated research communities from both sides to

initiate multidisciplinary studies and cooperative investigation in

structural control.

( 7 ) Barriers

No technical barriers are identified at the present time. However

lack of sufficient financial backing from both sides may present

difficulties in achieving the objectives of T/C "G"

,

605
Report of Task Committee on

(H) SOIL BEHAVIOR AND STABILITY DURING EARTHQUAKES

Date: MAY 14, 1991

Place: Public Works Research Institute

Tsukuba, Japan

Attendees; Japan side - K. Tokida (Chairman) (PWRI)

(PWRI)

(PWRI)

(PHRI)

(PWRI)

(BRI)

(PWRI)

Y. Koga

S. Sasaki

S. Noda

T. Fujisawa

Y. Yamazaki

Y. Yamaguchi(Observer)

U.S. side - A. G. Franklin (Chairman) (WES)

MISSION STATEMENT

1. Objective

Government agencies with responsibility for public works have the need

to assure their seismic safety and to provide economical protection against

earthquake hazards. The objective of this Task Committee is to assist in

meeting those needs by enhancing the available technology for predicting

the dynamic behavior of soils, analyzing dynamic soil-structure interaction,

and modifying the earthquake behavior of foundations and earth structures

in order to assure their safe performance during earthquakes.

2. Scope of Work

» Present technical papers at annual joint panel meetings on technological

developments and the state-of-the-art and practice related to soil

behavior and stability during earthquakes.

* Exchange information and technical data relating to field performance,

research and methods of practice.

* Plan and conduct T/C workshops in coordination with proposed or ongoing

cooperative research programs.

* The Task Committee plans and conducts programs of cooperative research,

exchange of researchers between U.S. and Japanese research institutions,

publication of research results and recommended practice. The benefits

realized from available research funds can be amplified by bringing

606
greater resources to bear on problems of mutual interest, and by

coordinating research tasks to make the best use of the respective

strengths of the two sides' research capabilities.

» Exchange visiting researchers using mechanisms provided by formal and

informal cooperative research programs. Such exchanges promote

technical information exchanges, familiarization with methods of practice

on the respective sides, and mutual understanding and cooperative

relationships between researchers on the two sides.

3. Accomplishments

« U.S. researchers presented 6 papers and Japanese researchers presented

11 papers, on soil behavior and stability during earthquakes, at the 22nd

and 23rd joint panel meetings. 5/90, 5/91

* Mr. Nario Yasuda of the PWRl completed a one-year guest researcher

assignment at the U.S. Bureau of Reclamation doing research in Seismic

Behavior of Soils. 1/90

* Dr. Yasushi Sasaki of the PWRI visited sites of damage around the San

Francisco Bay Area and the epicentral area of the Loma Prieta

Earthquake. He was assisted by several U.S. Panel Members including the

T/C (H) U.S. Chairman. 3/90

* A set of 35mm slides showing the effects of the Loma Prieta Earthquake was

provided to the Japanese side by the U.S. side. 3/90

* Dr. Ashok K. Chugh, of the U.S. Bureau of Reclamation, visited the PWRI in

March and April 1990, to study the Dynamic Behavior of Rock-Fill Dams.

4/90

* Mr. Keiichi Tamura of the PWRI completed a one-year visit as a guest

researcher at Stanford University working on Seismic Response

Characteristics of Ground. - 10/90

* Mr. Motoki Kazama of the PHRI completed a one-year visit as a guest

researcher at the Scrlpps Institute of Oceanography working on Effects of

Offshore Environments on Seismic Response of Soil Deposits and Pile

Foundations. 10/90

* The International Workshop on Remedial Treatment of Potentially Liquefiable Soils was conducted at PWRI in Tsukuba from 22 to 25 January, 1991.

The U.S. side was represented by Professor W. D. Liam Finn of the

University of British Columbia. Nineteen papers were presented. 1/91

» Mr. Ken-ichi Tokida, the T/C (H) Japan Chairman, visited St. Louis and met

Mr. R. H. Ledbetter of the U.S. Army Engineer Waterways Experiment

Station and discussed the proceedings of the International Workshop held

in January and plans for the next workshop. 3/91

607
» Proceedings of the International Workshop on Remedial Treatment of

Potentially Liquefiable Soils was published and distributed to the

members of Task Committee (H) and participants of the Workshop.

5/91

» The English language report on the expert team of Japan Disaster Relief

Team (JDR) on the earthquake in the Philippines of July 16, 1990, was

provided to the U.S. side. 5/91

4. Future Plans:

« Identification of research tasks on preventive measures against soil

liquefaction that are of common Interest, in accordance with the results

of the planning meeting held at Tsukuba in May 1989. 9/91

• T/C (H) proposes to hold the next U.S. -Japan Workshop on Remedial

Treatment of Potentially Liquefiable Soils in conjunction with the 25th

annual UJNR meeting. 5/93

• T/C (H) has relinquished its responsibility for consideration of marine

structures subject to earthquakes and the measurement of sea floor

earthquake motions to the new Task Committee (K) on Wind and Earthquake

Engineering for Offshore and Coastal Structures, recognized at the 23rd

Joint Panel Meeting. 5/91

» T/C (H), at the 22nd Joint Panel Meeting, discussed the need for

technological innovation and research to assist in saving lives following

natural disasters caused by earthquakes. The principal areas of need

were identified as improved technology for locating survivors and the

development of new equipment and systems for safe excavation of debris to

remove survivors from collapsed structures. T/C (H) recognizes this

subject is outside their area of responsibility but feels it is important

and should be discussed by the full Joint Panel.

The Task Committee suggests the following options for implementation:

(1) Establish a new Task Committee which will address this topic area

and provide the focal point to share experience and research in

this area.

(2) Incorporate this topic into another existing T/C.

T/C (H) will continue to explore methods whereby this topic will receive

appropriate visibility.

« T/C (H) will support the proposed visit of Mr. Yoshikazu Yamaguchi to the

U.S. Army Engineer Waterways Experiment Station as a visiting researcher

during 1991 and 1992.

» T/C (H) will support the proposed visit of Mr. Junichi Koseki to the

Massachusetts Institute of Technology as a visiting engineer during 1991

and 1992.

608
5. Information Exchange:

* T/C workshops, exchange of visitors, participation in annual joint panel

meetings, and presenting technical papers.

* Exchange of researchers between U.S. and Japanese government laboratories.

* Visits by researchers and participation in technical activities such as post-earthquake field investigations.

* Professional community in the U.S. and Japan, publication of papers in

journals, publication of research reports by the respective agencies,

participation in professional society meetings and conferences, and

workshops.

6. Impact

* Through the exchange of researchers and the diffusion of earthquakerelated technical data, experience, and information on methods of

practice, the state of technology in geotechnical earthquake engineering

has been raised on both sides.

* Through the cooperative research program on in-situ testing of soils

which was carried out after the 1983 Nihonkai-Chubu Earthquake, improved

and more accurate methods of using in-situ tests for evaluating the

liquefaction potential of soils were achieved; these have been in general

use in the United States since 1985. The Loma Prieta Earthquake

stimulated renewed awareness of the need for further research in soil

liquefaction and amplification effects of soft soils.

* Through the workshops on remedial treatment of potentially liquefiable

soils held in 1988 and 1991, beneficial information on recent technology

were exchanged and contributed to progress future research of both U.S.-

and Japan- sides.

7. Barriers

» Insufficient U.S. and Japan Government funds allocated to adequately

address earthquake engineering research.

* Administrative limitations on foreign travel.

609
Resolution of the International Workshop

on

the Remedial Treatment of Potentially Liquefiable Soils

Public Works Research Institute

Ministry of Construction

1 Asahi, Tsukuba-shi, Ibaraki-ken 305

Japan

January 22-25, 1991

The following resolutions are hereby adopted:

1. The U.S. Panel members of the UJNR Meeting were unable to participate in

this particular meeting because of effect of the Gulf War. Consequently, the

title of this particular meeting was chcuiged to International Workshop from

the Second Workshop of the Task Committee "H" of the UJNR Meeting which was

originally scheduled.

2. About 35 participants from Camada, Philippines and Japan joined this

Workshop, discussed and exchanged information eagerly and effectively.

19 papers were presented and discussed according to the workshop program.

Technical visits were made to the sites in and around the Tokyo Bay where

the remedial treatment of soil liquefaction and soft ground were conducted . All the participants to this Workshop recognized that this particular

meeting gave a beneficial chance to exchange the current findings cind

knowledge, which are inevitable to establish a more reliable technology and

standards concerning the remedial treatment of the potentially liquefiable

soils.

3. Some importcint findings of the workshop are as follows;

3-1 It is important to recognize what is the liquefied state, in order to find

the best solution. There was considerable discussion about the different

liquefied states; 1) Near surface liquefaction with upward seepage gradients

and very low liquefied strength, and 2) The deformation of soil structures

with liquefied zones at their residual strength. This kind of recognition

should be reflected in the algorithms of the analyses as well as in the back

calculation of the past cases.

610
2 A sophisticated technical method was presented to estimate the

deformation of soil structures such as embankment dams caused by the soil

liquefaction. Efforts like this kind of development are expected to refine

the current technology in analysis.

3 There is a great need to document the performance of treated ground

during earthquakes and to compile the information into a data base.

4 Further investigations for estimating the effects of liquefaction on

bridges and underground structures are required for establishing simplified

procedures and reasonable design methods.

5 Although various types of remedial treatment methods have been

developed and applied in practice, the standards and specifications for

selecting reasonable countermeasures against liquefaction and the induced

lateral ground displacement have not been established. For establishing the

effective countermeasures, the liquefaction phenomena, the degree of

liquefaction and its effect need to be investigated further.

6 The continuation of information exchange by the Joint Meetings, UJNR and

its future Workshops is considered essential, although the U.S. members

were unable to participate on this particular occasion.

611
Report of Task Committee on

(I) STORM SURGE AND TSUNAMI

Date: May 15, 1991

Place

;

Public Works Research Institute

Tsukuba, Japan

Attendees: Japan Side - T. Uda (Chairman) (PWRI)

K. Tanimoto (PWRI)

S. Ohtsuka (MRI)

M. Okada (MRI)

N. Ohyagi (NIED)

U.S. Side - J.F. Lander (Acting Chairman)

(Univ. of Colorado)

W.E. Roper (U.S. Corps)

MISSION STATEMENT

1. Objective

The objective of this T/C is to mitigate damages from storm surges and

tsunamis through shared technologies, research, information sharing, and

cooperative work. Storm surge and tsunami are hazards capable of

inflecting damage of disastrous proportions. Storm surges are associated with hurricanes and typhoons where high winds and the

mounding of water under the low barometric pressure of the storm's eye can cause extensive flooding and severe wave action. Tsunamis are predominately caused by underwater earthquakes and, to a lesser extent

volcanic activity. Depending on the distance from the source tsunamis

arrive within minutes to up to a day after generation.

2. Scope of Work

* Exchange results of research on storm surge and tsunami

occurrence, generation, propagation, and coastal effects. This

includes observations on historical, current, and theoretical

tsunamis. Of particular interest is the effort by U.S. and

JAPAN to acquire deep ocean tsunami measurements.

Exchange results and status of anti-storm surge and tsunami

activities including analysis of the problem, planning,

warning, and engineering approaches.

Exchange information on planned and ongoing projects relating

to storm surge and tsunamis.

Exchange information on development of technologies such as computer programs to predict travel times, land-fall locations,

run-up heights, and wave characteristics and analysis; improved

instrumentation, and use of satellite communications for

detection and warning.

Facilitate dissemination through exchange of literature,

technical reports at joint meetings, special workshops, joint

projects, and direct interaction among participants.

612
Accomplishments

* Exchanged digital bathymetric data with the Japanese side; U.S.

side received a computer program to calculate tsunami travel

time charts.

* Expanded the number of active U.S. participants with

representation from COE, University of Colorado, NOAA, and NSF.

* The U.S. side has compiled detailed information on all tsunamis

affecting the U.S. and its possessions. The publication,

"United States Tsunamis 1690-1989" has been printed and given

to the Japanese side.

* A Second Tsunami Workshop was held on November 5-7, 1990 in

Hawaii. Nine members participated from the Japanese side and

about 20 from the U.S. side. Site visits were made to Hilo as part of the workshop. Proceedings were published and they were

sent all related members.

* NOAA has adopted the travel time program developed by Mr. Okada

(MRI/JMA) using NOAA gridded bathymetric data to run on their

computers. The first ever computed travel time charts have

been prepared for the Caribbean Sea using this program

developed by Mr. Okada. A copy of the published paper was sent

to the Japanese side.

Future Plans

* Develop exchanges of data and information on activities related

to tsunami measurements including deep ocean measurements.

* Explore and promote joint undertakings such as tsunami gauge

testing, scientist exchange, satellite communications for

tsunami warning, and assistance in contacts within the U.S. and

JAPAN.

* The U.S. and Japanese sides agree that there is a need to hold

workshops at about 3 year intervals to exchange the latest

research and technology relating to various aspects of storm

surge and tsunami hazards. A third workshop including both

storm surge and tsunamis is being planned for August, 1993 in

Japan near the time of the International Union of Geodesy and

Geophysics (lUGG) Tsunami Commission Symposium.

* Develop activities and information exchanges to support the

International Decade of Natural Disaster Reduction.

Information Exchange . '

* Continue exchange of publications, bibliographic data,

technical reports, and personal contacts.

613
6 . Impact

* Increased cooperation among scientists and engineers working in

these fields.

* Hasten implementation of new modeling and design techniques

developed by the U.S. and Japan.

* Facilitated wider dissemination of information and technology

to scientists and engineers in both countries.

7 . Barriers

* Insufficient funding from government and private sector

organizations to participate in T/C workshops and the Panel's

annual joint meetings.

* The tsunami problem is less severe in the US than in JAPAN

which leads to a smaller community of researchers. There are few U.S. researchers focusing on tsunamis at any specific

location; thus highlighting the importance of T/C periodic

meetings and workshops.

* There are different approaches and degree of significance in

dealing with problems in the U.S. and JAPAN. For example, in

the U.S. much of the hazard is from remote sources which leads

the government to emphasize developments of warning and

evacuation schemes. In JAPAN the engineering approach is needed

because there are more locally generated tsunamis which often

do not provide sufficient times for adequate warning and

evacuation. This barrier is partially being addressed by

inviting U.S. engineers and non-government specialists to

participate during the annual meetings.

614
TASK COMMITTEE I

Report of the 2nd UJNR Tsunami Workshop

November 5 - 7, 1990

Honolulu and Hilo, Hawaii

Activities

1. The workshop was held over two days at the East-West Center,

University of Hawaii, followed by a one day field trip to Hilo,

Hawai i

.

2. Three topics were addressed

A. Predictions and Observations of Tsunami Behavior in Coastal

Waters and on Land.

B. Applications of Numerical Modeling to Improved Travel Time

Predictions, Source Region Definition, Inundation, and

Run-up Prediction.

C. Protective Measures in Japan and the U.S. With Regard to the

Use of Numerical Models and Design Tsunami Concept.

3. Discussions followed each paper and a longer round table

discussion followed each topic.

4. Twenty-five papers were presented and about 30 people were in

attendance including nine from Japan.

5. On the field trips, briefings were made by Dr. Gordon Burton,

Director of the Pacific Tsunami Warning Center at Ewa Beach, Oahu,

by the staff of Hawaiian Volcano Observatory and by Mr. Harry Kim,

Director of Hawaiian Civil Defense on Hawaii. The participants

toured the Kilauea Crater, the Tsunami Memorial and Hilo Bay area,

and the site of the destruction of the village of Kalapana by

recent volcanic activity.

6. The proceeding have been published and distributed to the

participants, task committee member and other involved with the

workshop. Additional copies will be available to tsunami

researchers world wide on a cost basis.

615
Comments

1. There was much good interaction among the participants. Among the

topics stressed for future work were:

a) the importance of having a true wave profile, perhaps from deep

ocean sensors both to determine the tsunami source structure and

to project for run-up models.

b) The need for observational data from the shore line and on shore areas where the waves have become non-linear.

c) The importance of continuous development of models simulating

harbor effects for specific harbors as a training and planning

tool. This includes accurate inundation models, life line

protection, tsunami forces on structures and effect of defense

measure such as dikes and breakwaters.

d) Quick evaluation of the tsunami potential perhaps using seismic

moment is needed for warning purposed as well as a more deliberate

evaluation which may involve historical data for planning

purposes.

2. Workshops and symposia are useful ways to involve Japanese and

U.S. government and academic researchers in the solution of the

Task Committee I's mission area in storm surge and tsunami and

should be convened at reasonable intervals,

3. Acknowledgment is made to the U.S. National Science Foundation

which supported most of the U.S. participants under grant

BCS-8908688 and NOAA's National Geophysical Data Center which

provided important clerical and other support including the

publication of the proceedings.

Takaaki Uda

Public Works Research Institute

Tsukuba, Japan

May 1991

616
Report of Task Committee on

(J) WIND AND EARTHQUAKE ENGINEERING FOR TRANSPORTATION SYSTEMS

Date: May 14, 1991

Place: Public Works Research Institute, Ministry of Construction

1-banchl Asaiii, Tsukuba-shi, Ibaraki-ken 305, JAPAN

Attendees: Japan Side M. Fujiwara (Chairman) (PWRl)

M. Okahara (PWRI)

T. Fujitani (MRI)

K. Nishikawa (PWRI)

Y. Nlnomiya (PWRI)

K. Hasegawa (PWRI)

U.S. Side - J.D. Cooper (Chairman) (FHWA)

W.E. Roper (Corps of Eng.)

MISSION STATEMENT

1. Objectives •

The objectives of this Task Committee are to plan, promote, and foster

research on the behavior of highway bridges when subjected to wind and seismic

forces and to disseminate research results and provide specifications and

guidelines based on the task committee's finding. Surface transportation systems

for movement of goods and people play a vital part of commerce and intercourse

between people. Highway bridges are especially influenced b3'^ the forces of wind

and earthquakes because of their open exposure to those forces.

2. Scope-of-Work

The scope of the work is applicable mostly to highway bridges without any

limitation on their size and function: such as existing bridge and new bridge

designs; whole system of bridges; and/or to single components of a bridge.

The mission is performed through:

Conducting workshops, exchanging researchers, developing methods for design

evaluations and test procedures, developing inspection techniques,

rehabilitation and maintenance specifications and policies, and performing

cooperative research programs and other relevant cooperative administrative

activities.

3. Accomplishments (past 2 years)

• Japanese side hosted an ad hoc T/C(J) Retrofit Subcommittee meeting in

December 17 to 22, 1990 to discuss and review practical seismic retrofit

applications in both countries. Nineteen technical presentations were made by

5 U.S. researchers and 30 Japanese researchers. Technical topics included the

617
history of developing retrofit techniques for bridges; damage assessment from

the Loma Prieta earthquake; discussion on condition assessment and

prioritization of bridges for seismic retrofit; seismic strengthening methods;

and research in progress. Site visits were made to view retrofit technology

applied to bridges on Tokyo metropolitan expressways.

United States and Japanese post earthquake investigative teams collaborated to

investigate the damage to transportation systems with emphasis on highway

bridges caused by the Loma Prieta Earthquake of October 17, 1989. Technical

information was exchanged at the 22nd and 23rd Joint Panel Meetings.

Conducted coordinated research studies on seismic performance of bridge piers

and columns. Details of this work involved determining the performance of

reinforced concrete piers and columns subjected to dynamic cyclic loading,

performing model tests on the failure of reinforced concrete piers, testing

full and large-scale concrete columns and for the behavior of concrete filled

steel tubes.

• Held the seventh bridge workshop in Tsukuba, Japan in 1991, which was followed

by site visits of Higashi Kobe bridge, Kansai international airport, and Akashi

Kaikyo bridge. Held the sixth bridge workshop in Lake Tahoe Nevada U.S. in

1990, dealing with the determination and evaluation of the performance and

strengthening of bridge structures, structural monitoring, repairs, safety and

nondestructive evaluation. Special attention was given to the seismic and wind

loadings of modern cable-stayed bridges, vibration suppression, base isolation,

dynamic control techniques, construction techniques of long span bridges,

maintenance of existing bridges and limit state design.

• Both sides presented technical papers at the 22nd and 23rd UJNR Joint Meeting

and participated In technical study tours to buildings, bridges, factories

construction sites, and instrumented structures.

4. Future Plans

Conduct the 8th bridge workshop Just prior to the 24th Joint Panel meeting in

the Chicago Illinois area of the United States.

• United States side plans to host a follow-up ad hoc T/C(J) Retrofit

Subcommittee meeting to study practical seismic retrofit measures.

• Continue to investigate and exchange technical information on improved seismic

retrofit and strengthening procedures for highway bridges based on

experimental, analytical, and field studies. This exchange should Include

information on maintenance of existing bridges.

• Continue experimental research study on the seismic performance of bridge

piers and columns, and support emphasis on base isolation of bridges In cooperation with T/C (G).

• Develop a coordinated research study on seismic, aeroelastic, and aerodynamic

response of cable-supported bridges with emphasis on cable Inspection,

vibration control and corrosion protection.

618
Conduct a coordinated research study to compare the seismic design criteria

for bridges in Japan and the U.S. and discuss the method and analysis for

bridge column design. Information on application of limit state design method

should be exchanged.

Continue a coordinated research study on seismic response control, system

identification techniques, and nondestructive evaluation of bridge structures.

Encourage a coodinated research study on use and performance of structural

materials including new materials.

5. Information Exchange

• Exchanged technical reports, research program documentation, construction

logs, design plans, and assorted photographs and videos, including technical

data appropriate for investigating the Philippine earthqualce of 1990.

• Mr. Keiichi Tamura who studied at Stanford University, Mr. Kinji Hasegawa who

studied at University of California and Mr. Shigeki Unjoh who studied at

University of Southern California, returned to PWRI.

Mr. Mohsen Sultan, California Department of Transportation is on a one year

exchange with PWRI In Tsuliuba.

6. Impact ,

• The free exchange of literature, instrumentation technology, and earthquake

response data has led to the advancement in design technology, and

development of preliminary retrofit guidelines, and to the calibration and

verification of specifications.

• Greater uniformity was achieved in wind engineering test procedures and

modeling resulting in more efficient solutipns to the aerodynamic bridge

problem.

• The workshop and study tours in the two countries have facilitated technology

transfer, as an example, in the area of cable protection and vibration

suppression of cable-stayed bridges.

A set of priorities for needed research from a more international viewpoint has

reduced the conduct of duplicative research efforts.

7. Barriers

There are no major technical barriers to achieve the Task Committee mission.

However, financial barriers, especially funding needed to participate in

workshops and panel meetings and for conducting actual bridge tests and large

scale model tests, continue to limit the effectiveness of the work in this Task

Committee's work.

619
APPENDIX1

RESOLUTION OF THE FIRST U.S. -JAPAN WORKSHOP ON SEISMIC RETROFIT

BRIDGES

TSUKUBA, JAPAN

December 17 and 18, 1990

The following resolutions are proposed by the organizers of the

First U.S. -Japan Workshop on Seismic Retrofit of Bridges to the

Task Committee J of UJNR:

1. Both Japanese and U.S. participants gained insight into the

need for bridge retrofit methods and technical solutions as a

result of the workshop. It is recommended a second workshop

on seismic retrofit of bridges be held in one or two years.

2. Considerable progress has been made in implementing retrofit

techniques to restrain bridge spans from failures. As a

consequence, future research and implementation should be

directed into more critical areas, such as substructure

stability.

3. Both countries are actively involved in developing retrofit

techniques for improving performance of bridge piers. The

research being carried out has been complementary rather than

duplicative which emphases the value of the current

information exchange. Japanese research is directed primarily

to problems associated with premature termination of

longitudinal reinforcement. U.S. research has been directed

toward improving flexural ductility and shear strength of

piers. Continuing cooperation and information exchange is

recommended in this area.

4. Both sides recognize the importance and difficulty of

effective retrofit of bridges on foundations of inadequate

strength or stability. Increased efforts to develop and

implement effective solutions is urged.

620
APPENDIX-2

RESOLUTION OF THE SEVENTH U.S. -JAPAN BPvIDGE ENGINEERING WORKSHOP

ANALYSIS, DESIGN, PERFORMANCE AND STRENGTHENING

TSUKUBA, JAPAN

May 8 and 9, 1991

The following resolutions are proposed by the organizers of the 7th U.S.

-Japan Bridge Engineering Workshop to the Task Committee J of UJNR:

The seventh U.S. -Japan Bridge Engineering Workshop provided valuable exchange of

technical information which was beneficial to bridge engineering communities of

both countries. 81 participants representing government, academia, industry and

private practice presented 32 papers in 10 technical sessions. Main themes

included : 1) Seismic Design; 2) Wind Resistant Design; 3) Bridge Construction; 4)

Design Methods; and 5) Bridge Maintenance. In view of the importance of

cooperative programs on the subject of wind and seismic effects on bridges and on

performance, strengthening, structural monitoring, safety, non-destructive

evaluation and design innovation of these transportation structures, the

continuation of technical cooperation and Interaction is considered essential.

The Workshop recognizes the importance of continued exchange of personnel,

technical information, research data and bridge response data from significant

earthquakes and strong winds, and the use of available unique research facilities

in both countries and encourages cooperative efforts in mutually agreeable areas of bridge research.

The primary purpose of the Workshop is to exchange technical information on a timely basis that is usable by bridge engineering profession. The specific

objectives of this Workshop are :

3.1 To ensure a first-hand exchange of information resulting from the unique

problems each country is faced with; for example, the development and

implementation of limit state or load and resistant factor design methods.

3.2 To stimulate collaborative research work in the important area of bridge

engineering for better design, construction, performance, rehabilitations and

strengthening of bridge structures in both countries.

3.3 To facilitate the exchange of technical information and ideas and to promote

the broadening of viewpoints for those involved.

621
4. The Workshop identifies the following technical topics for continued cooperative

research programs :

4.1 Dynamic control and seismic isolation of bridge structures including : active,

passive and hybrid control, damping augmentation, experimental studies and

application to long-span cable-supported bridges.

4.2 System identification techniques for wind and seismic engineering including : data interpretation studies, development of mathematical models, identification

of non-linear systems, reduced-order models, computational algorithms and

optimum sensor locations.

4.3 Non-destructive evaluation techniques including damage detection by sensor

monitoring, expert systems, inspection techniques, structural monitoring,

condition assessment for safety, performance evaluation and repair procedures.

4.4 Experimental and analytical studies on the effects of dynamic soil

characteristics, including bridge-soil interactions, on the behavior of bridge

structures.

4.5 Retrofitting and strengthening methodologies for bridge structures.

4.6 Development of design criteria based on limit-state concepts.

5. The Workshop identifies the following technical topic for future cooperative

technical programs : i

5.1 Use and performance of structural materials including new materials.

6. It is anticipated that the Interaction among the delegates from the United States

and Japan during the round-table discussions and the technical study tour will

lead to an enhanced exchange of data and innovative ideas, and will also lead to

the establishment of cooperative research efforts, both theoretical and

experimental, on bridge engineering. Because of the importance, benefits and

potential of these Workshops, it is recommended to conduct the 8th bridge

engineering workshop in Chicago Illinois, Just prior to the 24th UJNR joint meeting

in the United States, in May 1992. A study-tour will be planned to visit research

laboratories and bridge-construction sites.

622

•i' U.S^ GOVERNMENT PRINTING OFFICE 1991 - 281-557 401+91
: NIST-114A U.S. DEPARTMENT OF COMMERCE

1 (REV. 3-89) NATIONAL INSTITUTE OF STANDARDS AND TECHNOLOGY

RIRi DIDL.IVJVarlMr lOr^PAPHIP niV^ HATA UMIAOntlEli QHPFT

1. PUBUCATION OR REPORT NUMBER

NIST/SP-820

2. PERFORMINQ ORGANIZATION REPORT NUMBER

3. PUBUCATION DATE September 1991

4. TITLE AND SUBTITLE Wind and Seismic Effects--

Proceedings of the 23rd Joint Meeting of the U.S. -Japan Cooperative Program in Natural

Resources Panel on Wind and Seismic Effects

5. AUTHOR(S)

Noel J. Raufaste, Editor

j6. PERFORMINQ ORGANIZATION (IF JOINT OR OTHER THAN NIST, SEE INSTRUCTIONS) 7. CONTRACT/GRANT NUMBER

U.S. DEPARTMENT OF COMMERCE

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11. ABSTRACT (A 200-WORD OR LESS FACTUAL SUMMARY OF MOST SIGNIFICANT INFORMATION. IF DOCUMENT INCLUDES A SIGNIFICANT BIBUOGRAPHY OR

UTERATURE SURVEY, MENTION IT HERE.) This publication is the proceedings of the 23rd Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects. The meeting was held at the Public Works Research Institute,

! Tsukuba, Japan, during May 14-17, 1991. The proceedings include the program, list of

members, panel resolutions, task committee reports, and 42 technical papers.

The papers were presented under seven themes: (I) - Wind Engineering, (II) - Earthquake

Engineering, (III) - Storm Surge and Tsunamis, (IV) - Joint Cooperative Research Program,

(V) - Performance of Nonstructural Systems, (VI) - International Decade for Natural

Disaster Reduction, and (VII) - Summaries of Task Committee Workshop Reports (oral presentations only).

12. KEY WORDS (6 TO 12 ENTRIES; ALPHABETICAL ORDER; CAPITALIZE ONLY PROPER NAMES; AND SEPARATE KEY WORDS BY SEMICOLONS)

accelerograph; bridges; buildings; concrete; design criteria; disasters; disaster reduction;

earthquakes; geotechnical engineering; ground failures; lifelines; liquefaction; masonry;

repair and retrofit; risk assessment; seismic; soils; standards; storm surge; structural

engineering; tsunamis; and wind loads.

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