_ _ I. Btdg,

MAY i4 1981

1 9032S \

OClOO NBS SPECIAL PUBLICATION

,U5l

477

U.S. DEPARTMENT OF COMMERCE / National Bureau of Standards

fIND AND SEISMIC EFFECTS

Proceedings of the

Eighth Joint UJNR

Panel Conference
NATIONAL BUREAU OF STANDARDS

The National Bureau of Standards' was established by an act of Congress March 3, 1901. The Bureau's overall goal is to strengthen and advance the Nation's science and technology and facilitate their effective application for public benefit. To this end, the Bureau conducts research and provides: (1) a basis for the Nation's physical measurement system, (2) scientific and

technological services for industry and government, (3) a technical basis for equity in trade, and (4) technical services to pro- mote public safety. The Bureau consists of the Institute for Basic Standards, the Institute for Materials Research, the Institute for Applied Technology, the Institute for Computer Sciences and Technology, the Office for Information Programs, and the

Office of Experimental Technology Incentives Program.

THE INSTITUTE FOR BASIC STANDARDS provides the central basis within the United States of a complete and consist- ent system of physical measurement; coordinates that system with measurement systems of other nations; and furnishes essential services leading to accurate and uniform physical measurements throughout the Nation's scientific community, industry,

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1 Headquarters and Laboratories at Gaithersburg, Maryland, unless otherwise noted; mailing address Washington, D.C. 20234.

' Located at Boulder, Colorado 80302.
HAY 2 6 1977 WIND AND SEISMIC EFFECTS qo

Proceedings of the Eighth Joint

Panel Conference of the U.S.-Japan ^

Cooperative Program in ^1. • Natural Resources

May 18-21, 1976

National Bureau of Standards

Gaithersburg, Md.

H.S. Lew, Editor

Center for Building Technology

Institute for Applied Technology

National Bureau of Standards

Washington, D.C. 20234

U.S. DEPARTMENT OF COMMERCE, Juanita M. Kreps, Secretary

Dr. Sidney Harman, Under Secretary

Va C national bureau of standards, Ernest Ambler, Acting Director

Issued May 1977
Library of Congress Catalog Card Number: 77-600015

National Bureau of Standards Special Publication 477

Nat. Bur. Stand. (U.S.), Spec. Publ. 477, 626 pages (May 1977) CODEN: XNBSAV

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1977

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, D.C. 20402

(Order by SD Catalog No. CI3. 10:477). Stock No. 003-003-01772-8
PREFACE

The Eighth Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects was held

in Gaithersburg , Maryland on May 18-23, 1976. This panel is one of the eleven panels in the

U.S. -Japan Cooperative Program in Natural Resources (UJNR) . The UJNR was established in

1954 by the U.S. -Japan Cabinet-Level Committee on Trade and Economic Affairs. The purpose

of the UJNR is to exchange scientific and technological information which will be mutually

beneficial to the economics and welfare of both countries. Accordingly, the purpose of the

annual joint meeting of this panel is to exchange technical information on the latest research and development activities within governmental agencies of both countries in the area

of wind and seismic effects.

The proceedings include the program, the formal resolutions, and the technical papers

presented at the Joint Meeting. The papers were presented in the respective language of

each country. The texts of the papers, all of which were prepared in English, have been

edited for obvious errors and clarity. The contents of the papers are the sole responsibility of the individual authors. The illustrations were reproduced from the working documents used at the Joint Meeting. The formal resolutions were drafted at the closing

session of the Joint Meeting and adopted unanimously by the panels of both countries.

It should be noted that throughout these proceedings certain commercial equipment,

instruments or materials are identified in order to specify adequately experimental

procedure. In no case does such identification imply recommendation or endorsement by the

National Bureau of Standards, nor does it imply that the material or equipment identified is

necessarily the best available for the purpose.

Pages of the technical papers are numbered with a prefix corresponding to the theme

number. The texts are consecutively numbered in each theme.

H. S. Lew, Secretary

U.S. Panel on Wind and

Seismic Effects

iii
SI Conversion Units

In view of present accepted practice in this technological area, U.S. customary units

of measurements have been used throughout this report. It should be noted that the U.S.

is a signatory to the General Conference on Weights and Measures which gave official status

to the metric SI system of units in 1960. Conversion factors for units in this report are:

Customary Unit International

(SI) , UNIT

Conversion

Approximate

Length

Force

Pressure

Stress

Energy

Torque

or Bending

Moment

Weight

inch (in)

foot (ft)

pound (Ibf)

kilogram (kgf)

pound per square

inch (psi)

kip per square

inch (ksi)

inch-pound (in-lbf)

foot-pound (ft-lbf)

pound- inch

pound- foot (Ibf-ft)

pound (lb)

meter (m) meter (m) newton (N) newton (N)

2 newton/meter

2 newton/meter

joule (J) joule (J) newton-meter (N-m)

newton-meter (N-m)

kilogram (kg)

1 in=0. 0254m*

1 ft=0. 3048m*

1 lbf=4.448N

1 kgf=9.807N

2 1 psi=6895N/m

1 ksi=5895xlO^N/m^

1 in-lbf=0.1130 J

1 ft-lbf=1.3558 J

1 lbf-in=0.1130 N-m

1 lbf-ft=1.3558 N-m

1 lb=0.4536 kg

Unit Weight

Velocity

Acceleration

pound per cubic foot

(pcf)

foot per second

(ft/sec)

foot per second

^ per second (ft/sec )

kilogram per cubic

meter (kg/m^)

meter per second

(m/s)

meter per secon^

per second (m/s )

1 pcf=16. 018 kg/m

1 fps=0.3048 m/s

2 2 1 ft/sec =0.3048 m/s

^Meter may be subdivided. A centimeter (cm) is 1/100 m and a millimeter (mm) is 1/1000 m. *Exact.

XV
ABSTRACT

The Eighth Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects was held

in Gaithersburg, Maryland on May 18-21, 1976. The proceedings of the Joint Meeting include

the program, the formal resolutions, and the technical papers. The subject matter covered

in the papers includes wind effects on structiires and design criteria; extreme winds for

structural design; earthquake ground motions and instrumentation; seismicity and earthquake

risk; seismic effects on structures and design criteria; lessons learned from recent natural

disasters; design of nuclear reactor facilities.

Key Words: Accelerograph; Bridges; Buildings; Codes; Disaster; Dynamic Analysis; Earthquakes; Ground Failures; Nuclear Facilities; Seismicity; Soils; Standards;

Structural Response; and Winds.

CONTENTS

Preface

SI Conversion ^jy Abstract Y Program of the Joint Meeting

Opening Session

List of Members

Photographs of Participants -^-jj^ Formal Resolutions -jo^

Themes and Technical Papers I-l

Theme I: WIND EFFECTS ON STRUCTURAL AND DESIGN CRITERIA I-l

Aspects of Hurricane Winds as Recorded at an Instrumented Suspension Bridge . . I-l

R. H. Gade, R. H. Scanlan

Study on the Wind Effect on Eaves , 1-12

Tatsuro Murota

Measurements of Wind Loads and Tie Down Forces on Mobile Homes 1-21

Richard Marshall, Robert A. Crist

On the Wind Resistant Design Specifications for the Proposed Honshu-Shikoku

Bridges (1975) 1-34

Tadayoshi Okubo, Nabuyuki Narita

Equivalent Static Wind Loads for Tall Building Design 1-49

Emil Simiu

A Wood House Will Resist Wind Forces 1-66

Billy Bohannan

Wind Loads on Low Rise Buildings 1-70

Noel Jo Raufaste

Theme II: EXTREME WINDS FOR STRUCTURAL DESIGN II-l

Characteristics of the High Wind at Hachijojima Island on the Occasion of

Typhoon No. 7513 II-l

Seiji Soma

High Winds in the United States-1975 11-15

Arnold R. Hull, Thomas D. Potter, Nathaniel B. Guttman

Mean Speed Profiles of Hurricane Winds 11-30

Emil Simiu, Virendra C. Patel, John F. Nash. ....

Theme III: GROUND MOTIONS AND INSTRUMENTATION III-l

Planning and Design of Strong Motion Instrument Networks III-l

R. B. Matthiesen

Observation of Earthquake Response of Ground With Horizontal and Vertical

Seismometer Arrays III-16

Satoshi Hayashi, Hajime Tsuchida, Eiichi Kurata

Building Strong Motion Earthquake Instrumentation III-26

Christopher Rojahn

Characteristics of Underground Seismic Motions at Four Sites Around

Tokyo Bay II 1-41

Toshio Iwasaki, Susumu Wakabayashi, Fumio Tatsuoka

vii
THEME IV: SEISMICITY AND EARTHQUAKE RISK IV-1

Relationship Between Earthquake Damage of Existing Wooden Houses and Seismic. IV-1

Intensities

Eiichi Kuribayashi, Takayuki Tazaku, Takayuki Hadate

A Method for Calculating Nonlinear Seismic Response in Two Dimensions . . IV-18

William B. Joyner

A New Scale Representing the "Quake Sensitivity" at a Certain Region. . . IV-47

Tsutomu Terashima, Tetsuo Santo

THEME V: LESSONS LEARNED FROM RECENT NATURAL DISASTERS V-1

Damage to the Civil Engineering Structures on Hachijojima Island by

Typhoon 7513 V-1

T. Okubo, N. Narita , K. Yokoyama

Cyclone Tracy V-21

Richard D. Marshall

On the Damage to Buildings in Hachijojima Caused by Typhoon No. 7513. • . V-54

Tatsuro Murota

THEME VI: SEISMIC EFFECTS ON STRUCTURES AND DESIGN CRITERIA VI-1

Seismic Response of Reinforced Concrete Highway Bridges ...... VI-1

Joseph Penzien, W.G. Godden, M.C. Chen, D. Williams, K. Kawashima

An Evaluation Method for the Earthquake Resistant Capacity of Reinforced

Concrete and Steel Reinforced Concrete Columns . . VI-10

Masakazu Ozaki, Yuji Ishiyama

The Earthquake Engineering Program of the National Science Foundation . , VI-28

John B. Scalzi

Large-Scale Testing Programs Related to Wind and Seismic Effects Currently

Underway in Japan . . . . . . VI-37

Seiichi Inaba

Earthquake Damages to Earth Structures VI-44

Kenkichi Sawada

Dynamic Test of a Circuit Breaker for Transformer Substation VI-50

Seiichi Inaba, Shegio Kinoshita

Comprehensive Seismic Design Provisions for Buildings - A Status Report . . VI-61

Charles Culver

Retrofitting of Vulnerability in Earthquake Disaster Mitigation Problems. . VI-69

Kaoru Ichihara, Eiichi Kuribayashi, Tadayuki Tazaki

Dynamic Response Characteristics of a Model Arch Dam. VI-85

Charles D. Norman, Roger D. Crowson, Jimmy P. Balsara

The Measurement of the Dynamic k-Value in Site and Its Application

To Design VI-118

Tetsuo Kunihiro, Kaname Yahagi, Michio Okahara

Laboratory Investigation of Undisturbed Sampling and Standard Penetration

Tests on Fine Sands VI-141

W. F. Marcuson III, S.S. Cooper, M.A. Bieganousky

Dynamic Soil Properties with Emphasis on Comparison of Laboratory

Tests and Field Measurements VI-158

Toshio Iwasaki, Fumio Tatsuoka

Design Earthquakes VI-179

Ellis L. Krinitzsky, Frank K. Chang

viii
Relation Between Seismic Coefficient and Ground Acceleration for Gravity

Quaywall VI-192

Satoshi Hayashi, Setsuo Noda , Tatsuo Uwabe

Investigation of Earthquake Resistance of Structural (Shear) Wall

Buildings Carried out at Portland Cement Association ....... VI-199

Mark Fintel

A Philosophy for Structural Integrity of Large Panel Buildings .... VI— 220

Mark Fintel, Donald M. Schultz

THEME VII: DESIGN OF NUCLEAR REACTOR FACILITIES VII-1

Wind and Seismic Design of U.S. Nuclear Power Plants . VII-1

Lawrence C. Shao, Richard J. Stuart, Charles H. Hofmayer

Outline of Basic Philosophy and Practices of Aseismatic Design for Nuclear

Facilities in Japan ............... VII-29

Makoto Watabe, Yorihiko Ohsaki

Tornado-Borne Missiles .... VII-36

James F. Costello

THEME VIII: SPECIAL SESSION ON GUATEMALA EARTHQUAKE . . VIII-1

Structural Damage to Bridges Resulting from the Guatemala Earthquake . . . VIII-1

James D. Cooper

ix
EIGHTH JOINT MEETING

OF THE

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS

MAY 18-21, 1976 AT THE NATIONAL BUREAU OF STANDARDS

TUESDAY-May 18

OPENING SESSION

10:00 Call to order by Dr. H.S. Lew, Secretary, U.S. Panel

Remarks by Mr. Robert S. Walleigh, Acting Deputy Director

National Bureau of Standards

Remarks by Mr. Kiichiro Nagara, Counselor, Embassy of Japan

Remarks by Dr. Edward O. Pfrang, Chairperson, U.S. Panel

National Bureau of Standards

Remarks by Dr. Kaoru Ichihara, Chairperson, Japan Panel

10:30 Introduction of U.S. Panel Members by U.S. Chairperson and

Japan Panel Members by Japan Chairperson

10:45 Election of Conference Chairperson

11:00 Adoption of Agenda

11:15 Adjourn

11:30 Group Photograph

12:00 Lunch-Dining Room "C"

SESSION II - WIND EFFECTS ON STRUCTURES AND DESIGN CRITERIA

Chairman: Dr. Edward O. Pfrang

1:30 Aspects of Hurricane Winds as Recorded at an Instrumented Suspension

Bridge — R.H. Gade and R.H. Scanlan

2:00 Study of Wind Effects on Eaves —

Tatsuro Murota (Presented by Masakazu Ozaki)

2:30 Measurements of Wind Loads and Tie-Down Forces on Mobile Homes —

Richard Marshall and Robert A. Crist

2:40 On the Wind Resistant Design Specifications for the Proposed Shikoku

Bridges (1975) — Tadayoshi Okubo and Nobuyuki Narita

2:50 Equivalent Static Wind Loads for Tall Buildings —

Emil Simiu (Presented by Robert A. Crist)

3:00 Break

X
3:20 A Wood House Will Resist Wind Forces —

Billy Bohannan

3:30 Wind Loads on Low- Rise Buildings —

Noel J. Raufaste

3:40 Discussion

4:10 Adjourn

4:15 Informal Discussion of Resolution and Task Committees

WEDNESDAY-May 19

SESSION III: EXTREME WINDS FOR STRUCTURAL DESIGN

Chairman: Dr. Kaoru Ichihara

9:00 Characteristics of the High Wind at Hachijojima Island on the Occasion

of Typhoon No. 7513 — Seiji Soma (Presented by Masakazu Ozaki)

9:30 High Winds in the United States 1975 —

Arnold R, Hull, Thomas D. Potter, Nathaniel B. Guttman

10:00 Mean Profiles of Hurricane Winds —

Emil simiu (Presented by Robert A. Crist)

10:10 Discussion

10 : 30 Break

SESSION IV: EARTHQUAKE GROUND MOTIONS AND INSTRUMENTATION

Chairman: Dr. Edward O. Pfrang

10:45 Planning and Design of Strong-Motion Instrument Networks —

R.B. Matthiesen

11:15 Observation of Earthquake Response of Ground with Horizontal and

Vertical Seismometer Arrays — Satoshi Hayashi, Hajime Tsuchida

and Eiichi Kurata

11:45 Building Strong-Motion Earthquake Instrumentation —

Christopher Rojahn

11:55 Characteristics of Underground Seismic Motions at Four Sites

Around Tokyo Bay — Toshio Iwasaki, Susumo Wakabayashi and

Fumio Tatsuoka

12:05 Discussion

12:30 Lunch-Dining Room "C"

SESSION V: SEISMICITY AND EARTHQUAKE RISK

Chairman: Dr. Kaoru Ichihara

2:00 Relationship Between Earthquake Disaster on Existing Houses and

Seismicity — Eiichi Kuribayashi, Tadayuki Tazaki and Takayuki Hadate

xi
2:30 A Method for Calculating Nonlinear Seismic Response in Two Dimensions —

William B. Joyner (Presented by R.B. Matthiesen)

2:40 A New Scale Representing the "Quake Sensitivity" at a Certain Region —

Tsutomu Terashima and Tetsuo Santo

2:50 Discussion

3:30 Break

SESSION VI: LESSONS LEARNED FROM RECENT NATURAL DISASTERS

Chairman: Dr. Kaoru Ichihara

3:45 Damage to the Civil Engineering Structures on Hachijojima Island

by Typhoon 7513 — Tadayoshi Okubo, Nobuyuki Norita and Koichi Yokouama

(Presented by Masakazu Ozaki)

4:15 Cyclone Tracy —

Richard Marshall

4:45 The Damage to Buildings on Hachijojima Island Caused by

Typhoon 7513 — Tatsuro Murota

4:55 Discussion

5 : 15 Adj ourn

THURSDAY-May 20

SESSION VII: SEISMIC EFFECTS ON STRUCTURES AND DESIGN CRITERIA

Chairman: Dr. Edward O. Pfrang

9:00 Seismic Response of Reinforced Concrete Highway Bridges —

J.A. Penzien, W.G. Gooden, M.C. Chen, D. Williams, and Kikawashima

9:30 An Evaluation Method for the Earthquake Resistant Capacity of

Reinforced Concrete and Steel Reinforced Concrete Columns —

Masakazu Ozaki and Yuji Ishiyama

10:00 The Earthquake Engineering Program of the National Science Foundation

John B. Scalzi

10:10 Large-scale Testing Programs Related to Wind and Seismic Effects

Currently Underway in Japan

Seiichi Inaba (Presented by T. Kinoshita)

10:20 Earthquake Damages of Earth Structures —

Kenkichi Sawaua

10 : 30 Break

10:40 Dynamic Tests of Circuit Breakers for Transformer Substation —

Seiichi Inaba and Shigeo Kinoshita

11:00 New U.S. Recommendations on Seismic Design Criteria —

Charles G. Culver
11:10 Discussion

12:00 Lunch-Dining Room "C"

1:30 Retrofitting of Vulnerability in Earthquake Disaster Mitigation Problems - Kaoru Ichihara, Eiichi Kuribayoshi and Tadayuki Tazaki

2:00 Dynamic Response Characteristics of a Model Arch Dam —

C. Dean Normans Roger D. Crowson and Jimmy P. Balsara

2:30 The Measurement of the Dynamic k-Value In Site and Its Application to Design — Tetsuo Kianihiro, Kaname Yahagi and Michio Okahara

2:40 Laboratory Investigation of Undisturbed Sampling and Standard Penetration

Tests in Fine Sands — William F. Marcuson III, S.S. Cooper and W.A. Bieganousky

2:50 Dynamic Soil Properties with Emphasis on Comparison of Laboratory

Tests and Field Measurements — Toshio Iwasaki and Fumio Tatsuoka

3:00 Break

3:20 Design Earthquakes —

Ellis L. Rrinitzsky and Frank K. Chang

3:30 Relation Between Seismic Coefficient and Ground Acceleration for

Gravity Quaywall — Satoshi Hayashi, Setsuo Noda and Tetsuo Uwabe

3:40 Investigation of Earthquake Resistance of Structural (Shear) Wall

Buildings Carried Out at Portland Cement Association — Mark Fintel

3:50 A Philosophy for Structural Integrity of Large Panel Buildings

Mark Fintel and Donald M. Schultz

4:00 Discussion

4:45 Adjourn

FRIDAY-May 21

SESSION VIII: DESIGN OF NUCLEAR REACTOR FACILITIES

Chairman: Dr. Kaoru Ichihara

9:00 Wind and Seismic Design of U.S. Nuclear Power Plants —

Lawrence C. Shao > Richard J. Stuart and Charles H. Hofmayer

9:30 Outline of Basic Philosophy and Practice of Aseismometer Design for

Nuclear Facilities in Japan — Makoto Watabe and Yorihiko Osaki

(Presented by Masakazu Ozaki)

10:00 Tornado Borne Missiles —

James F. Costello

10:10 Discussion

10:40 Break

xiii
SESSION IX: SPECIAL SESSION ON GUATEMALA EARTHQUAKE

Chairman: Dr. Kaoru Ichihara

Structural Damage to Bridges Resulting from the Guatemala Earthquake —

James D. Cooper

Informal Discussion of Guatemala Earthquake —

Panel: Masakazu Ozaki, James D. Cooper, Charles G. Culver, Mark Fintel and

R.B. Matthiesen, and Eiichi Kuribayashi

Lunch-Dining Room "C"

SESSION X: TASK COMMITTEE REPORTS

Chairman: Dr. Edward O. Pfrang

Strong Motion Instrumentation Arrays and Data —

R.B. Matthiesen and Hajime Tsuchida (Represented by Satoshi Hayashi)

Large Scale Testing Program —

Charles G, Culver and Seiichi Inaba (Represented by Takeo Kinoshita)

Repair and Retrofit of Existing Structures —

John J. Healy and Kiyoshi Nakano (Represented by Masakazu Ozaki)

Structural Performance Evaluation —

John Scalzi and Masakazu Ozaki

Land Use Programs for Controlling Natural Hazard Effects —

S.T. Algermissen and Eiichi Kurbayashi

Disaster Prevention Methods for Lifeline Systems —

Charles Scheffey and Tadayoshi Okubo (Presented by Eiichi Kuribayashi)

High Speed Wind Recorded Data —

Thomas D. Potter and Seiji Soma (Presented by Masakazu Ozaki)

Discussion of Task Committee Reports

SESSION XI: CLOSING SESSION

Chairman: Dr. Edward O. Pfrang

Adoption of Formal Resolution

Closing Remarks —

Dr. Kaoru Ichihara

Dr. Edward O. Pfrang

Adj ourn
U.S. Panel on Wind and Seismic Effects

Dr. Edward O. Pfrang

Chairman

Chief, Structures, Materials and

Safety Division

Center for Building Technology, lAT

National Bureau of Standards

Room B368, Bldg. 226

Washington, D.C. 20234

Dr. S.T. Algermissen

Office of Earthquake Studies

Denver Federal Center

Branch of Earthquake Tactonics

uses

Stop 978, Box 25046

Denver, Colorado 802 25 303-234-4014

Mr. Billy Bohannan

Assistant Director

Wood Engineering Research

Forest Products Laboratory

US DA

P.O. Box 5130

Madison, Wisconsin 53705

608-257-2211

Dr. Roger D. Borcherdt

Office of Earthquake Studies

Branch of Earthquake Hazards

uses

345 Middlefield Road

Menlo Park, California 94025

415-323-8111 (FTS 467-2755)

Dr. Charles G. Culver

Disaster Research Coordinator

National Bureau of Standards

Room B212, Bldg. 226

Washington, D.C. 202 34 301-921-3126

Mr. Jerry Dodd

Bureau of Reclamation

Mail Code 230

P.O. Box 25007

Denver Federal Center

Denver, Colorado 80225

Membership List

May 1976

Dr. Michael P. Gaus

Head, Engineering Mechanics Section

Engineering Division

National Science Foundation

1800 G. Street, N.W. Washington, D.C. 20550

202-632-5787

Mr. John J. Healy

Department of the Army

DAEN-RDM

Washington, D.C. 20314

202-693-7287

Dr. William B. Joyner

Office of Earthquake Studies

Branch of Earthquake Hazards

USGS

345 Middlefield Road

Menlo Park, California 94025

413-323-8111 (FTS 467-2754)

Mr. John W. Kaufman

Aerospace Environment Division

ES42-Space Science Laboratory

National Aeronautics and Space

Administration

Marshall Space Flight Center, Alabama

35812, 205-453-3104

Dr. Noel E. LaSeur

Director

National Hurricane and Experimental

Meteorology Laboratory

National Oceanic and Atmospheric

Administration

P.O. Box 248265

Coral Gables, Florida 33134

305-350-4150

Dr. Richard D. McConnell

Office of Construction

Veterans Administration

811 Vermont Ave., N.W. Washington, D.C. 20234

202-389-3103

Dr. Richard D. Marshall

Research Structural Engineer

National Bureau of Standards

Room B168, Bldg. 226

Washington, D.C. 20234

301-921-3475

XV
Dr. R.B. Matthiesen

Chief, Branch of Seismic Engineering

Office of Earthquake Studies

USGS

345 Middlefield Road

Menlo Park, California 94025

415-323-8111 (FTS 467-2881)

Mr. Howard L. Metcalf

Staff Engineer

Construction Standards and Design

Directorate

Office of the Assistant Secretary

of Defense

Room 3E763

The Pentagon

Washington, D.C. 20301

202-695-2713

Mr. Keith O. O'Donnell

Assistant Chief, Structural Branch

Engineering Division, Civil Works

Directorate

Office of The Chief of Engineers

ATTN: DAEN-CWE-D

Washington, D.C. 20314

202-693-7311

Dr. Thomas D. Potter

Director

National Climatic Center

Environmental Data Service

National Oceanic and Atmospheric

Administration

Federal Building

Asheville, North Carolina 28801

704-258-2850 x236

Dr. John B. Scalzi

Program Manager, Earthquake Engineers

Division of Advanced Environmental

Research and Technology

National Science Foundation

1800 G Street, N.W. Washington, D.C. 20550

202-632-0648

Mr. Charles Scheffey

Director

Office of Research

HRS-1

Federal Highway Administration

Department of Transportation

Washington, D.C. 20590

202-426-2943

Mr. Lawrence C. Shao

Chief, Structural Engineering Branch

Nuclear Regulatory Commission

Washington, D.C. 20555

202-492-8035

xvi

Dr. Warren A. Shaw

Head, Civil Engineering Department

Civil Engineering Laboratory

Naval Construction Battalion Center

Port Hueneme, California 93043

805-982-5407

Mr. William J. Werner

Energy Building Technology and Standards

Department of Housing and Urban Development

Room 8158

451 7th Street, S.W. Washington, D.C. 20410

202-755-0642

ALTERNATIVES

Dr. A. Gerald Brady

Physical Scientist

Office of Earthquake Studies

USGS

345 Menlo Park, California 94025

415-323-8111 (FTS-467-2881)

Dr. James D. Cooper

Structures & Applied Mechanics Division

Federal Highway Administration

Office of Research HRS-11

Washington, D.C. 20590

202-557-4315

Mr. G. Robert Fuller

Architectural and Engineering Division

Department of Housing and Urban Development

Room 6176, 451 7th Street, S.W. Washington, D.C. 20411

202-755-5932

Dr. H.S. Lew

Secretary

Structures, Materials and Safety Division

Center for Building Technology, lAT

Washington, D.C. 202 34 301-921-3851

Dr. Charles T. Thiel

Deputy Director

Division of Advanced Environmental

Research and Technology

National Science Foundation

1800 G. Street, N.W. Washington, D.C. 20550

202-632-5734

Mr. Drew A. Tiedemann

Bureau of Reclamation

Engineering and Research Center

Denver Federal Center

P.O. Box 25007

Denver, Colorado 80225

303-234-3029
Japanese Panel on Wind and Seismic Effects

Membership List

May 1976

Dr. Kaoru Ichihara, Chairman

Director

Piablic Works Research Institute

Ministry of Construction

28-32 Honkomagcme 2-chome

Bunkyo-ku, Tokyo, 113

Dr. Sat.oshi Hayashi

Head, Structure Division

Port and Harbour Research Institute

Ministry of Transport

1-1 Nagase 3-chome

Yokosuka-shi , Kanagawa-ken , 2 39

Mr. Seiichi Inaba

Chief, Earthquake Engineering Laboratory

National Research Center for

Disaster Prevention

Science and Technology Agency

4489 Kurihara, Sakura-mura

Niihari-gun, Ibaragi-ken, 300-32

Mr. Toshio Iwasaki

Chief, Earthquake Resistant Structure Sec. Structure and Bridge Division, Chiba Branch

Public Works Research Institute

Ministry of Construction

12-52 Anagawa 4-chome

Chiba-shi, Chiba-ken, 280

Dr. Tetsuo Kunihiro

Head, Structure and Bridge Division

Chiba Branch

Public Works Research Institute

Ministry of Construction

12-52 Anagawa 4-chome

Chiba-shi, Chiba-ken, 280

Mr. Eiichi Kuribayashi

Chief, Earthquake Engineering Section

Structure and Bridge Division, Chiba Branch

Piiblic Words Research Institute

Ministry of Construction

12-52 Anagawa 4-chome

Chiba-shi, Chiba-ken, 280

Mr. Tatsuro Murato

Chief, Structure Section

Structure Division

Building Research Institute

Ministry of Construction

28-8 Hyakunin-cho 3-chome

Shinjuku-ku, Tokyo 160

Dr. Kiyoshi Nakano

Assistant Director

Building Research Institute

Ministry of Construction

28-8 Hyakunin-cho 3-chome

Shinjuku-ku, Tokyo, 160

Mr. Nobuyuki Narita

Chief, Structure Section

Structure and Bridge Division, Chiba Branch

Public Works Research Institute

Ministry of Construction

12-52 Anagawa 4-chome

Chiba-shi, Chiba-ken, 280

Dr. Masakazu Ozaki

Head, Structures Division

Building Research Institute

Ministry of Construction

28-8 Hyakunin-cho 3-chome

Shinjuku-ku, Tokyo, 160

Dr. Tetsuo Santo

Head, International Institute of Seismology

and Earthquake Engineering (I.I.S.E.E.)

Building Research Institute

Ministry of Construction

28-8 Hyakunin-cho 3-chome

Shinjuku-ku, Tokyo, 160

Mr.Kenkichi Sawada

Chief, Soil Dynamics Section

Construction Method and Equipment Division,

Chiba Branch

Public Works Research Institute

Ministry of Construction

12-52 Anagawa 4-chome

Chiba-shi, Chiba-ken, 280

Dr. Seiji Soma

Chief, First Research Section

Physical Meterology Laboratory

Nakano Branch

Meteorological Research Institute

Meteorological Agency

35-2 Nakota 1-chome

Nakano-ku, Tokyo, 165

Mr. Akira Suwa

Head, Seismological Laboratory

Meteorological Research Institute

Meteorological Agency

35-8 Koenji, Kita 4-chome

Suginami-ku, Tokyo, 166

XV ii
Dr. Tsutomu Terashima

Chief, Seismology Section

International Institute of Seismology and

Earthquake Engineering (I.I.S.E.E.)

Building Research Institute

Ministry of Construction

28-8 Hyakunin-cho e-chome

Shinjuku, Tokuo, 160

Mr. Hajime Tsuchida

Chief, Earthquake Resistant Structures

Laboratory

Structure Division

Port and Harbour Research Institute

Ministry of Transport

1-1 Nagase 3-chome,

Yokosuka-shi , Kanagawa-ken , 239

Dr. Makoto Watabe

Chief, International Institute of

Seismology and Earthquake Engineering

(I.I.S.E.E.

)

Building Research Institute

Ministry of Construction

28-2 Hyakunin-cho 3-chome

Shinj loku-ku , Tokyo, 160

Mr . Kaname Yahagi

Chief, Foundation Engineering Section

Structure and Bridge Division

Chiba Branch

Public Works Research Institute

Ministry of Construction

12-52 Anagawa 4-chome

Chiba-shi, Chiba-ken, 280

Dr. Tadayoshi Okubo

Secretary-General

Head, Planning and Research Administration

Division

Public Works Research Institute

Ministry of Construction

28-32 Honkomagome 2-chome

Bunkyo-ku, Takyo, 113

xviii
xix

RESOLUTIONS OF EIGHTH JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS

U. J.N.R.

May 18-21, 1976

The following resolutions for future activities of this Joint Panel are hereby proposed:

1. The Eighth Joint Meeting was an extremely valuable exchange of technical information

which was beneficial to both countries. Considering the importance of cooperative

programs on the subjects of wind and seismic effects, the continuation of the Joint

Panel Meeting is considered essential.

2. The resolutions of the Seventh Joint Panel Meeting are considered important to the continued cooperative program and are hereby incorporated as a part of these resolutions:

a. Disseminate technical reports and state-of-the-art reports continually.

b. Publish proceedings of each joint meeting for benefit of other organizations

in each country.

c. Solicit state-of-the-art papers on wind and seismic effects from other governmental and private technical organizations.

d. Promote dissemination of technological information to developing countries.

e. Promote cooperative research programs through the Task Committees including

exchange of personnel and available equipment. Publish reports of these

Task Committees.

f. Incorporate SI Metric System for units of measurement.

3. Establish a new Task Committee to consider problems associated with soils such as liquefaction, landslides, subsidence, etc. An exchange of complete soils reports from areas

of liquefactions should be encouraged. These should include accurate boring logs with

description of soils and field and laboratory test data.

4. The date and location of the Ninth Joint Meeting on Wind and Seismic Effects will be

May 1977 in Tokyo, Japan. Specific dates and itinerary of the meeting will be determined

by the Japanese Panel with concurrence by the U.S. Panel.

xxi
4
Aspects of Hurricane Winds as Recorded at an

Instrumented Suspension Bridge

by

R. H. GADE

Federal Highway Administration, McLean, Virginia

and

R. H. SCANLAN

Princeton University, Princeton, New Jersey

ABSTRACT

Velocity spectra of the winds of tropical storm Doria in its passage over

the Newport, Rhode Island, Suspension Bridge is presented for four mean wind

speeds. Data was acquired with five 3-component Gill propeller anomometers,

spaced at approximately 350 feet (108 m) intervals along the semispan of the

bridge. This paper presents final information on the wind mean angle of attack,

turbulence levels, vertical and horizontal spectra of wind of various velocity

levels, derived surface drag coefficients, and a comparison of results with

standard spectra and other applicable data from the Japanese, United States,

and Canadian wind literature.

KEY WORDS: Wind; wind velocity spectra

INTRODUCTION

In 1971 wind-speed measuring instruments were temporarily installed, under the

supervision of the Federal Highway Administration, and with permission of the Rhode Island

Turnpike and Toll Bridge Authority, on the Newport Bridge, Rhode Island. Figure 1 depicts

the geographical location of this bridge. The instrumentation consisted of 3-component Gill

windmill-type anemometer groups at each of five locations on the bridge. These locations

are shown in Figure 2, with their respective elevations and span positions designated. The

bridge, completed in 1969, satisfactorily withstood the winds of tropical storm Doria which

I-l
passed in the vicinity in August 1971. During this passage, peak horizontal wind gusts as

high as 75 mph (33.5 m/s) were recorded by the instrumentation on the bridge. Some characteristics of the project and wind data obtained therefrom are discussed below.

INSTRUMENTATION

Three-component anemometers were placed in five locations designated by numbers 5

through 9 in Figure 2. The anemometers of the Gill windmill-type extended on booms out

about 35 feet (10.7 m) from the lower wind truss at each location on the seaward side of

the bridge. The distance constant of the Gill anemometer is 1.07 meters. The frequency

response characteristics of the anemometer, relative to frequencies in the variable velocity

signal of the wind, are not completely known. Research continues on this question.

However, the data from the anemometers is treated in the present paper as faithfully representing the component wind speeds.

The system used accomplished repeated sampling of the analog velocity signals at 14.3

Hz, resulting in a Nyquist data cut-off frequency of about 7 Hz, which was largely adequate

for the wind variations observed. In the final processing, data compression permitted a

reasonable array size. An upper frequency range of 1.8 Hz was obtained with the processed

data set.

WIND DATA

Each transducer group furnished wind data representing the following time-varying

quantities

U: horizontal velocity component perpendicular to the bridge roadway

centerline

V: horizontal velocity component parallel to the plane of the roadway

centerline

W: vertical velocity component normal to V

Only the U and W components are discussed in this paper; i.e., velocities along the bridge

span are not considered in present analyses.

J. M. McMichael and P. S. Klebanoff, "The Dynamic Response of Helicoid Anemometers,"

National Bureau of Standards Report NBSIR 75-772 November 1975, (Available as PB 246-861

from the National Technical Information Services, Va. 22l6l.)

1-2
The basic wind data selected for analysis consisted of 40 time-histories of approximately

10 minutes each. The more convenient record length of 10 minutes was found to give consistent

results to 20-minute records, which were also examined initially. The records were selected

in nominal mean horizontal velocity ranges of 15, 30, 50, and 60 mph (6.7, 13.4, 22.4, and

26.8 m/s) from each of the U- and W-transducers at positions 5 through 9, respectively.

Broadly speaking, the rotating storm passed in such a manner that the lowest recorded velocities approached the bridge from the southeast, while the highest velocities were almost

directly perpendicular to the span. The actual mean wind velocity components together with

associated variances about the mean are presented in Table I.

PRELIMINARY RESULTS AND METHOD OF DATA PROCESSING

The basic W/U represents the tangent of the instantaneous angle of attack of the wind

crossing the bridge deck. This was calculated for each of the transducer locations, and

the results were plotted. Figure 3. It is observed that a band of values is produced which

represents a steady diminution of mean-wind angle to the horizontal at transducer height on

the bridge (average about 205.6 feet = 62.7 meters). This angle drops from about 20° in

mild winds down to 3° or 4° in high winds. These results tend to confirm the validity of

the aerodynamic testing of bridge models at angles to the wind not usually exceeding 6°.

The evidence of Figure 3 also suggests that there is probably also a decrease in turbulence level a/U (where U is the mean horizontal velocity) with increasing wind speed.

Figure 4 is a plot of this specific trend. Somewhat analogous data are presented in [2],

though the possibility of much higher turbulence occurring at extremely high velocities

over a sea fetch ("50 m/s) is suggested there also.

The basic time-history data referred to in Table I was digitally processed according

to the flow chart presented in Table II. In order to smooth and reduce the digital data

available at 0.07 second intervals to more manageable proportions, new samples were taken

at every fourth point only; i.e., a sampling interval of 0.28 seconds. While basic timehistory selections were visually made of winds with sustained characteristics, it was found

[2] M. Shiotani and H. Arai, "Lateral Structures of Gusts of High Winds," Paper 20 VHI Proc.

Int'l. Res. Seminar on Wind Effects on Buildings and Structures, NRG, Ottawa, Canada,

September 1967, pp 535-555, University of Toronto Press1-3
necessary to remove slight "ramp" trends from such selections prior to evaluating means.

Standard Fourier transform techniques with taper window were employed, followed by conversion to power spectral density S(n) in the "logarithmic" form ^^^'^'^

^ where n is frequency,

a which has become conventional in the presentation of wind spectra. The abscissa used for

eventual display was the mean-wind wave number n/U , U being taken at bridge mean elevation.

HORIZONTAL WIND SPECTRA

One ten-minute sample wind time-history from transducer 7U at nominal 60 mph (26.8 m/s)

together with its derived spectrum presented in the standard form nS-^(n) /o^ versus n/U, is

presented in Figure 5, where U is the mean wind velocity at transducer height. Altogether,

20 spectra of this type were derived from the similar time-histories recorded at the five

transducer locations 5 through 9 for the four selected nominal wind speed ranges.

Figures 6 through 9 present the combined spectral results for all transducers at each

of the four nominal wind speeds, respectively. Along with each mean spectral plot is presented a comparative plot based on the following f ormula:

KU^CIO) (2+n2)5/6

where K is the surface drag coefficient, and U(10) is the mean velocity, both referred to

10-meter height, and N = nL

, L being a peak wavelength "constant." This formula is

U(10)

converted for comparative purposes to the desired form as follows;

nSy(n) Uk2u2(z)

CT, 22n— In

^'^'^

1

X(z)

In 62.7

2 -.5/6

2 +

X z, In 10

where K, Karman's constant, is taken as 0.4, U(z) is averaged mean wind velocity at bridge

height (av. 62.7 m) , and 1/X(z) = n/U(z); Zq, the surface roughness length, is related to K

by the formula: ,

iln i^f

[3]

R. I. Harris, "The Nature of the Wind," Paper 3 Proc. CIRIA Seminar: The Modern Design

of Wind-Sensitive Structures , London, England, 1971, pp 29-55.
The peaks of the sample and reference spectra were arbitrarily equated to obtain the

values of K and L. The resulting relations are:

1.812 KU^(IO) = [n S^(n)]^^^a^^

and

^ ^ /3 ^°/zo

_1 7 62.

T

A(z) z

VERTICAL WIND SPECTRA

A sample spectrum obtained from a ten-minute time-history of wind vertical velocity at

transducer 9W at the nominal mean wind speed of 50 mph (22.4 m/s) is presented in Figure 10

in the form

'^

.^W^'^) versus n/U for U at transducer altitude. Altogether, 20 such spectra

were developed. Figures 11 through 14 present mean combined spectra of vertical wind at

each of the nominal horizontal wind speeds. On each of Figures 11 through 14 is presented,

for comparative purposes, an empirical spectrum derived from the following formula

•^SyC^) 6f

KU^(IO) (1 + 4f)^

where f = /y , z being the height of the transducer, and U being the mean horizontal

velocity at that height.

This formula is converted for comparative purposes to the desired form as follows:

nS^(n) k2 U^(z )(i/X(z))6z ^ W

The surface drag coefficient is similarly obtained by

0.375 KU^(IO) = [ns^(n)]^^^ ^^2

and Zq as above.

H. A. Panofsky and R. A. McCormick, "The Spectrum of Vertical Velocity Near the Surface,"

Quart. Jnl. Royal Meteorological Soc. , 1960, Vol. 86.

1-5
DISCUSSION

It was noted earlier that the wind fetches were different (see Figure 1) as the wind

direction swung from the 15 mph (6.7 m/s) values over to 60 mph (26.8 m/s) . Thermal effects

probably were more strongly present at the lowest velocities, though some evidence

(Figure 3) suggests that there was an unstratified (rising) condition during all measurements

taken. It is of interest to note that horizontal spectral peaks at all wind speeds in the

presently analyzed data occurred at practically the same frequency (0.06 Hz); similarly,

the vertical peaks grouped around 0.13 Hz. In Figures 6 through 14, values of K and (where

applicable) L for the reference spectra are given. The K values are reasonable with respect

to published values for the expected fetches; values of L are shorter than values chosen

for spectral matching elsewhere. '^^ Owen''^'' and Simiu'-^-' comment that spectral peaks

may be expected to shift widely in location. However, the right-hand, downward slopes of

the spectral curves, Figures 6 through 14, exhibit fair agreement with the expected boundary

-5/3 layer energy cascade results (N law) . The spectra exhibited show reasonable similarity

to spectra published elsewhere of which [8] and [9] refer specifically to typhoons.

A. G. Davenport, "The Dependence of Wind Loads Upon Meteorological Parameters," Paper

2, Proc. Int'l. Res. Seminar on Wind Effects on Buildings and Structures , Ottawa,

Canada, September 1967, University of Toronto Press.

[6] P. R. Owen, "Buildings in the Wind," Quarterly Journal of the Royal Meteorological

Society . Vol. 97, 1971, pp 396-413.

E. Simiu, "Wind Spectra and Dynamic Alongwind Response," Jnl. Struct. Div. Proc , ASCE,

Vol. 100, No. ST9, September 1974, pp 1897-1910.

[8] M. Hino, "Spectrum of Gust Wind," Proc. 3rd Int'l Conf. on Wind Effects on Buildings

and Structures , Tokyo, 1971, pp 69-77.

J Y. Mitsuta, Preliminary Results of Typhoon Wind Observation at Tarama Island, Okinawa,

Proc. U.S. -Japan Res. Seminar on Wind Effects on Structures , Kyoto, Japan, September

1974.

1-6
TABLE I

MEAN AND VARIANCE OF WIND VELOCITIES OF THE 10-MINUTE DORIA DATA SEGMENTS USED FOR SPECTRAL ANALYSIS

NOMINAL

VELOCITY

HORIZONTAL VERTICAL

15 MPH

(6.7 m/s)

30 MPH

(13.4 m/s)

50 MPH

(22.4 m/s)

60 MPH

(26.8 m/s)

MEAN

U

VARIANCE MEAN

W

VARIANCE

9U

MPH

15.29 11 37 9W

MPH

4.36 3. 68

8U 18. 41 8 47 8W 3. 72 3. 44

7U 16. 74 8.84 7W 4. 42 4. 86

6U 15. 94 9 71 6W 4. 01 3. 52

5U 19. 97 9 77 5W 3. 52 3. 58

Average 17. 270 9 632 Average 4. 006 3. 816

9U 31.00 19. 94 9W 4. 25 3. 93

8U 31. 85 28 29 8W 3. 46 5. 83

7U 32. 35 18 27 7W 4. 05 5. 38

6U 27. 84 20 10 6W 4. 26 4, 83

5U 34. 62 28 77 5W 3. 48 4. 94

Average 31. 532 23 074 Average -3 ~J • 900 4. 982

9U 52. 69 61 87 9W 2. 91 9

.

23

8U 51. 70 58 02 8W 2. 38 9. 10

7U 48. 63 44 68 7W 3. 56 9

.

93

6U 48. 63 40 50 6W 4. 23 7

.

30

5U 53. 29 39 60 5W 2. 96 7. 51

Average 50. 988 48 926 Average 3. 208 8. 614

9U 51. 22 39 52 9W 3. 12 6. 93

8U 52. 37 27 34 8W 1. 91 9. 49

7U 49. 19 29 02 7W 3. 46 9. 11

6U 49. 89 24 26 6W 4. 51 5. 13

5U 55. 86 26 96 5W 2. 69 7. 78

Average 51. 706 29 420 Average 3. 138 7. 688

1-7
TABLE II

PROCESSING OF WIND DATA

Time-Series Data (Sample interval 0.07 sec)

Data Resampling (Sample interval 0.28 sec)

Demean & Detrend

o Apply Window Taper (10% cosine )

Time-to-Frequency

Transform

Convert Amplitude Spectra

to Reduced

Power Density Spectra

Convert Frequency

to Wave Number

RPD. =

X

ampl-^ X freq.

VARIANCE

(WN. =

freq.^/mean velocity)

Spectral Smoothing (19 pt.

averaging)

1-8
Figure 1 Site Location

Newport, R. I. Bridge

NEWPORT BRIDGE HURRICANE "DORIA" MEAN ANGLE OF ATTACK OF

WIND vs MEAN WIND VELOCITY

Figure 3

2000 30-00 40.00 50.00 MEAN CROSS-DECK VELOCITY, mph

(I mph = 0.44715 m/s)

0.20

HORIZONTAL VELOCITY ACROSS DECK

(I mph = 0.44715 m/s) TURBULENCE LEVEL vs, MEAN WIND VELOCITY

Figure 4

1-9
I-IO
1 m

VERTICAL WIND SPECTRUM VERTICAL WIND SPECTRUM

50 mph 60 mph Figure 13 Figure 14

I-ll
STUDY ON THE WIND EFFECT ON EAVES

T. MUROTA

Chief, Structure Section

Building Research Institute

Ministry of Construction

Japan

ABSTRACT

Recently in Japan damage to flat roofs of light weight roofing is increasing.

The damage is often very severe to the extent that complete removal of the roofs occurs. Th

dynamic behavior of the eaves is related to the damage. A current research project is

also described.

Key Words: Dynamics; Failure; Roofs; Wind.

1. Introduction

Eaves, and ridges of dwellings often suffer damage from severe winds. This local

damage to roofs is caused by the high wind suction acting on these various parts. These

local failures, however, have not caused severe damage to traditional wooden houses in

Japan. This is because traditional Japanese wooden houses are constructed with heavy

clay- tile roofings. When such clay-tiles are lifted away, the roof structures composed

of bed boards and rafters still remain undamaged and therefore the interiors of houses

are well protected from the wind and rain.

In Japan, the use of clay-tile on roofs is decreasing, and the use of light weight

thin metal plate roof material is increasing. Roof structures are therefore also becoming

light due to the change of the weight of the roof material. Accordingly, the damage to roofs is changing, and in some instances, these light weight roofs have shown perfect

wind resistant performance. Those roofs which suffered severe damage such as complete

removal of roofs, have been flat roofs or slightly pitched roofs of gentle slope. Typical

damage of this type has recently been reported, in which complete removal of the roofs

of about 30 prefabricated houses occurred. These houses were located in Aichi prefecture,

and were subjected to an average wind of about 30 m/s as caused by Typhoon No. 7220 in 1972

The characteristics of the roofs which were damaged are as follows:

i) flat roofs

ii) long projection of eaves

iii) low flexural rigidity of the roof structures

Examination of some of the damage which occurred also indicates that one of the causes of

the roof removals was vibration of the eaves.

In 1975, Typhoon No. 7513 caused similar damage to flat roofs which were constructed

of light roofing material. After this storm, the Building Research Institute initiated a 1-12
research project which would study techniques to prevent damage to light weight roofs. The

following will describe the ideas of this project relative to wind excited oscillations of

flat roof eaves.

2. Characteristics of Wind Pressure on Eaves

It is well known from wind tunnel tests that large negative pressures develop on the

wind ward edge of eaves. The values of this pressure differ according to each investigator, however, it is not iinusual that a pressure coefficient of -5 be recorded near windward edge.

Shape factors for eaves have been estimated by some investigators. These factors

have been determined by assuming that the pressure on eaves is static and uniformly distributed along the span. The maximum shape factor value is reported is about -3. The

pressure coefficient is 0.7 to 0.8 for the under surface of eaves and the upper surface

coefficient is -2.2 to -2.3. However, such large values have not been obtained in wind

tunnel tests, except in the case when the eaves are very short.

The Building Code of Japan has specified a value of -0.5 prior to 1973, and after

1973 the value of -1.5 has been adopted.

In order to discuss the appropriateness of these values as a design shape factor for

eaves, it is necessary to know the characteristics of the pressure fluctuation on eaves

and the corresponding dynamic behavior of eaves. In the past, many investigations on the

pressure fluctuation on square cylinders induced by Karman vortices, have been made.

However, only the investigations which will now be described, give details on the pressure

fluctuation near the edge of buildings

.

In order to generate Karman vortices actively, the flow around a body has to be twodimensional and also the cross-section of the body must not be slender in the direction

of the wind. However, in case of ordinary buildings, these conditions are not satisfied.

It is considered, therefore, that the turbulence induced by the separation of flow will be

predominant in the pressure fluctuation generated near edges of buildings. From this point

of view Ishizaki et al"*"' have measured pressure on the surface of slender square cylinders

in a smooth wind tunnel flow, as shown in Fig. 1. They obtained the following results, as given in Fig. 2 and 3.

i) At points near the windward edge, high negative pressures were observed but their

fluctuation was not intense,

ii) Pressure fluctuation increased on the leeward surfaces,

iii) Spectra of fluctuations showed that pressures were not composed of a single frequency component, as in case of Karman vortex excited fluctuations, but were

composed of widely ranged frequency components,

iv) The reduced frequency, of the spectra of fluctuations, has a peak value 0.16,

which agrees with the Strouhal number for square cylinders.

2) Mori has measured pressures on a flat roof model set in turbulent boundary layers

as shown in Fig. 4. He obtained the following results, where h is the height of the roof

model and Zo is the roughness length:

1-13
i) When h/z =6, the pressures contained non-convective fluctuation components of

o frequencies around fh/V=0.07 and also convective fluctuation components of higher

frequencies (Fig. 5) .

ii) When larger values of h/z^ occur, non-convective fluctuations observed near the

windward edge are diffused rapidly downstream and convective fluctuations have a spectral peak at a reduced frequency of fh/V-0.2 to 0.25 which became predominant

(Fig. 6 and 7)

The value of the central reduced frequency of the non-convective pressure fluctuation

components , observed near the windward edge , is 0.14 where the characteristic length is

taken as twice the roof height h, this value agrees with the Strouhal number for square

cylinders, as in the case of the experiments of Ishizaki et al as stated before.

From the results of these two experiments, it can be concluded that the pressure

fluctuation near the windward edge of flat roofs (whose lengths of the horizontal direction perpendicular to wind are large) is induced by the turbulence of flow separation at

the windward edges . A spectra of these pressures have energy -peaks at a reduced frequency

2fh/V, which is about the same value as the Strouhal number for square cylinders.

3. Power Spectrum of Pressure Fluctuations on Eaves

Width and/or depth of plan of ordinary houses are not large compared to their height.

Therefore, air-flows round their roofs will have different properties from those of the

above two experiments. It is considered, however, that pressure fluctuations on flat

roof eaves are induced by turbulence of separated flow and have a random nature, which is

the same for ordinary houses having flat roofs

Then we can assume the spectra of pressure fluctuation on eaves, S (f )

, as a form

P

f-S (f) = o^-g(f) (1) P P where

f = fh/V (2)

It is considered that the universal function g(f) can be obtained by wind tunnel

experiments. This function g(f) depends on the aerodynamical shape of the houses, however,

we expect the function to be constant which is applicable to flat roofs of most common

sizes and shapes. In addition it will be better if pressure fluctuations under surface

of eaves (this is considered to have the same nature as that on windward wall) are measured in the experiments and the effects are taken into account when determining the

form of g ( f )

4. Response of Eaves in Turbulent Winds

If the form of the power spectrum of pressure fluctuations on the eaves was shown

in eq. (1) , it is possible to obtain the statistical characteristics of the response by

1-14
applying the probabilistic theory of dynamics. This is a well-known method and the outline

is as follows

:

The equation of eaves is

Z(y,t) + 26w Z(y,t) + W^(y,t) = - P(y,t) (3) o o m

where Z = deflection of windward edge of eave at a position y along the edge and at time t. 1/2

= (k/m) , where K is spring constant of eave and m is mass per unit length in y direction

.

6 = damping constant.

P(y/t) = total pressure along eave span per unit length in y direction.

The solution of eq. (3), in terms of normal mode ym(y), is

Z(y,t) = Z am (t) -y (y) (4) m

where

, , sin mlTy

y„(y) =

and I is length of eave in the direction perpendicular to wind.

Here we need a cross spectrum of pressure fluctuation (y,y' ,f) , which gives the

correlation of the pressure fluctuations between two points y and y' . If this function is

obtained by experiments, the spectra of the mode coefficients m is given by the equation

S (f) = |h (f) l^'Jmm (f) (6) am ' m II2 H (f) is the mechanical admittance function of the form m

'

|h (f) P = 1/M^ [(l-W^/W^)"* + (26W/W )^] (7)

' m ' m m m

and Jmm (f) is the joint admittance function,

Jmm (f) =

j |sp(y^,y2,t) y^ (y^) y^ (yz ) dy idy2 (8) o o Then the variation of a is obtained as follows m

J? a = \ S ( W )dw (9) m J um

5. Concluding Remarks

Recently in Japan damage to flat roofs of light weight roofing is increasing. The

damage is often very severe such as bodily removal of roofs. The dynamic behavior of the eaves is considered to relate to this damage. In this paper is described a research project presently being conducted by the Building Research Institute concerning the problem.

1-15
Reference

1) H. Ishizaki and J, Katsura, Intensity of Fluctuating Wind Pressures on the Side

Surfaces of Models with Long Rectangular Section, Transactions of A.I.J. , No. 214, Dec. 1973, pp. 1-6.

2) T. Mori, Pressure Fluctuations on Steps in Turbulent Boundary Layers, M.E. Sc. Thesis, Faculty of Engineering Science, Univ. of Western Ontario, 1971.

3) Y.K. Lin, Probabilistic theory of Structural Dynamics, McGraw-Hill.

1-16
Fig. 1 Arrangement of

a model-'^^

Fig. 2 Distribution of static

pressure coefficient-'-)

• 2x15.10

ix30 ,20

. 8x60. iO

|X1U

I ,. ..»»» t » ....

.

4

[

Rsint A

1

10

i 1-

15

icH f

L

-i.

L „^

* <

.

H h

2 6.3 20 63 2C0 630 2000 HZ 2 6J 20 63 2;. 630 2Ca Fig. 3 Power spectra of fluctuating pressure coeff icients^

Vc

I = 6h

Th

0 X

Fig. 4 Mori s wind tunnel test 2)

1-17
1-18

1-20
MEASUREMENTS OF WIND LOADS AND TIE-DOWN FORCES ON MOBILE HOMES

RICHARD D. MARSHALL

National Bureau of Standards

Gaithersburg , Maryland

and

ROBERT A. CRIST

National Bureau of Standards

Gaithersburg, Maryland

ABSTRACT

This paper describes instrumentation, experimental techniques and progress to date

on a program of research into the effects of wind on mobile homes. Direct measurements

of lift and drag forces on a nominal 12 x 60 ft. (3.66 x 18.3 m) mobile home provide more

reliable information on load fluctuations than is possible with the usual approach of

measuring surface pressure alone. Results of spectral analysis conducted on time histories of overturning forces suggest that a simple quasi-static approach, when used with

appropriate gust factors, can be employed to calculate loads for the design of mobile home

anchoring systems.

Key Words: Aerodynamics; Buildings; Full-scale testing; Mobile homes; Wind loads.

1. Introduction

Although the general area of wind research has made tremendous advances over the past

decade, most studies have been carried out on tall buildings and only recently have the

problems associated with wind forces on low-rise buildings been given proper attention.

The purpose of the research effort described in this paper is to establish a source

of data upon which to base criteria for the design of mobile homes and their anchoring

systems to resist wind forces. Mobile homes account for a significant percentage of single

family dwellings in the United States and, next to fire, wind is 'ihe primary cause of

deaths, injuries and property damage.

The problem of wind forces on mobile homes is complicated by the fact that, near

ground level, the local terrain and upwind buildings have a pronounced effect on the variation of mean wind speed and turbulence intensity with height. In addition, their limited

size makes them more sensitive to wind gusts than is the case with conventional low-rise

buildings where spatial averaging can substantially reduce the effectiveness of load

fluctuations

.

This project is being funded by the Department of Housing and Urban Development (HUD)

under the Mobile Home Construction and Safety Act.

1-21
2. Description of Experiment

a) The Test Site

The field tests are being carried out on the Gaithersburg Campus of the National

Bureau of Standards in an area which has a clear exposure to the prevailing winds. Strong

winds in the Gaithersburg area are most frequent during the winter months from November

through March and are usually associated with the passage of low pressure systems to the

north. The strongest winds [50 mph (22 m/s)] generally blow from the west-northwest, although several events can be expected each winter from the southwest. The surrounding terrain

is gently rolling and mean velocity profiles can best be described by a power law [V = V^^{ZI1^q) ] with values of a ranging from 0.16 to 0.25. This exposure corresponds approximately to Exposure C of the American National Standards Institute Standard A58. 1-1972. A

plan of the test site and prevailing wind directions are shown in figure 1.

b) Design and Installation of Mobile Home Support System

The test site is located on a paved parking lot which facilitated mounting the

mobile home on a supporting frame and turntable so that various relative wind directions

could be selected for each wind storm. This allowed an expedient method of gathering the

necessary data for the forces on the mobile home during one wind season at the NBS site.

Consultations were held with the project sponsor (HUD) and representatives of the mobile

home manufacturing industry to determine the most representative mobile home for the research. A nominal 12 ft (3.66 m) by 60 ft (18.3 m) mobile home with a roof to ground height

of 10.8 ft (3.30 m) was used. The test setup also allowed the open space between the ground

and the floor level to be enclosed or open to study the effects of these two configurations

on lift and drag.

The support system was designed so that the lowest first mode natural frequencies

(vertical or horizontal) of the supports and the linking components were at least the / 2 times the largest first mode natural frequencies of the mobile home (vertical and horizontal)

This prevented resonance of the mobile home and supports and approximately uncoupled the

dynamic response of the mobile home and the supports, i.e., the foundation stiffness >> mobile home- stiffness . This provided a realistic simulation of the in-service foundation

conditions of a mobile home. The mobile home natural frequencies were determined by

temporarily mounting the home on rigid supports similar to those to be used in the experiment, exciting the home and measuring the free vibration with an accelerometer . The highest

first mode natural frequencies in the place of the floor were 8 Hz vertically and 4 Hz horizontally

The support system can best be described as three separate systems, the foundation

system which consists of four beam and column assemblies, a force measurement system, and a main support frame. The four beam and column assemblies simulate an actual installation which includes diagonal and vertical tie-down cables (figure 2) . The

force measurement system consists of four force links to measure horizontal forces

(drag) and eight force links to measure lift and overturning. All force links are connected

to a main support frame with four leveling points and a turntable (figure 3) . The

turntable is staked to the ground through a set of wooden ties. The turntable allowed

1-22
216 degrees of rotation (figure 1) of the mobile home. A set of vertical and horizontal

steel wheels on the turntable holds the home in place for rotation during windstorms as well as when it is in the stationary position. The lowest calculated first mode natural

frequencies of the foundation system and main support frame are 80 Hz horizontally and 32 Hz vertically. The force link lowest calculated first mode natural frequencies are 10 Hz horizontally and 12 Hz vertically.

3. Field Measurements

a) Wind Speed

Two types of anemometers are being used to obtain wind speeds at the test site, the

standard 3-cup Mod. F420C used by the National Weather Service and Mod. 27100 Gill 4-blade

propeller anemometers. The Mod. F420C is mounted at the 10-meter level of an 18-meter

guyed mast and provides wind speed records which can ultimately be related to extrememile speeds referenced in ANSI Standard A58. 1-1972. Gill propeller anemometers are mounted at the 18, 10, 5, 3 and 1.5 meter levels to provide a good definition of the mean velocity profile and to allow flow characteristics at the height of the mobile home to be

expressed in terms of mean wind speeds at the 10-meter level. These anemometers are conventionally used in pairs to measure the u and v components of turbulence and to establish the mean speed vector. However, the present installation uses only one anemometer

at each level, the anemometers being mounted on a vertical support pipe which runs the

height of the mast. The support pipe is manually adjusted prior to recording of data so that the anemometers are faced directly into the wind. A direction vane is mounted at the

top of the mast.

In addition to the anemometers on the mast, two Gill anemometers are mounted on tripods which are positioned upwind of the mobile home for obtaining along-wind and

cross-wind correlations. The tripods are adjustable to cover a range of heights from

near ground level to the roof height. Nominal sensitivity of the anemometers is 10 mv

per mph (22.4 mv per m/s)

.

b) Surface Pressure

Surface pressures are measured by means of differential pressure transducers of

the variable reluctance type. Nominal output is + 10 volts at the full-scale range of

2 0.1 psi (689 N/m ). This allows direct recording without signal conditioning, although

the transducer demodulators are provided with switch-selectible filters which are usually

set at a cutoff frequency of 10 Hz. A conventional tap- tube-transducer arrangement is

used, the transducers being mounted inside the mobile home and connected to the 3/8-in

(95 mm) diameter pressure taps with short lengths of flexible tubing. The frequency re- sponse of this system is flat to approximately 20 Hz. The pressure transducers are referenced to outside ambient pressure by means of an omnidirectional probe mounted on the guyed mast. Transducer offsets are measured just

prior to recording of data by means of solenoid valves which shunt the active pressure

line to the reference side of the transducer. The transducer and solenoid valve are

1-23
built into an assembly that can be quickly attached to a wall or ceiling tap. A total of

27 transducers are available and can be placed in various arrays, depending upon the

pressure patterns being investigated. The pressure tap layout is shown in figure 4.

c) Deformation and Acceleration

Racking deformations are measured at the approximate mid- length of the mobile

home with displacement transducers mounted on telescoping rods rimning along the diagonals

of the mobile home cross-section. Recordings of horizontal acceleration have been obtained at floor level and along the bottom chord of the roof trusses for the purpose of

estimating structural damping.

d) Lift and Drag Forces

The force links previously described are instrumented with strain gages in a fullbridge configuration to produce maximum sensitivity and to provide temperature compensation. Nominal sensitivity of the force links after amplification is 1 mv per Ibf (225

yv per N) . Bias voltages at the amplifier inputs allow the bridges to be nulled for "at rest" conditions.

e) Data Acquisition

The primary data acquisition system is a computer- controlled digital and analog

system with a total capacity of 112 data channels and an effective maximum sampling rate

of 13 samples per second per data channel . A programmable gain control allows the selection of five ranges for each data channel; +^0.5, j^l, + 2, and +_ 10 volts. Normal

operation utilizes approximately 40 data channels and the 20 kHz scan rate reduced time

skew to a negligible amount for the frequencies of interest in this study. Data channels

are usually sampled at the rate of 12 samples per second, but this rate is doubled for

the measurement of surface pressures at points on the mobile home where large fluctuations

and high frequency components are observed. Record lengths range from 500 to 1000 seconds,

depending upon the sampling rate. All recordings, both digital and analog, are identified

by time code

Certain data channels are also recorded in analog form on a 14-channel recorder

under control of the system computer. This allows analog analysis of selected records

and digital analysis at higher sampling rates by selecting the proper ratio of record to reproduce speeds

4. Data Reduction and Analysis

a) Methods

Because of the large amounts of data involved in this study, it is important that

the data handling be as efficient and reliable as possible. An extensive package of com- puter software has been developed at the NBS over the past few years for the analysis of

wind load data. Although the techniques used depend upon the particular variable being

analyzed, the analysis generally consists of the following steps:

1. Sorting of samples by data channel.

2. Calculation of means and removal of trends.

3. Calculation of peak departures from the mean and their cumulative distributions.

1-24
4. Area integration of surface pressures.

5. Spectral analysis.

To aid in the selection of records which are to undergo detailed analysis, continuous

stripchart recordings of wind speed and direction are taken during strong wind conditions

and the digital recording intervals are noted on the stripcharts . A subjective assessment

is then made as to the stationarity of the records and their general suitability for

additional analysis.

The analog recordings provide additional information on data quality but are primarily

used for spectral analysis. Auto- and crosscorrelations are obtained by use of a hybrid

analyzer which accepts an analog input and computes a 400-point correlation function. By

making repeated entries of the analog records, extremely stable estimates of the correlation function can be obtained. A Fourier transformation of the 400 points is then carried

out. This technique has the advantage of an early look at the form of the correlation

function without the expense of computer time and, if the form appears to be satisfactory,

extremely reliable estimates of the spectral densities with a minimum of computational

time required.

b) Typical Response Analysis

As previously described, the support system was designed to measure drag (horizontal)

and overturning (lift) forces. A typical overturning force response time history is shown

in figure 5 with the corresponding wind speed record from the 10-meter standard 3-cup

anemometer. Refer to figure 1 for the location of the anemometer in plan. Note that the

major force response correlates well in time history with the slow increases and decreases

in the wind speed. To further analyze such records statistical analyses as previously

discussed were performed on the force link time history of approximately 17 minutes, a portion of which is shown in figure 5. The following parameters were determined, mean wind speed = 21.6 mph (9.7 m/s)

y , = mean force = 119 lb (530 N) xl

peak force = 199 lb (885 N) = standard deviation, force = 29 lb (130 N)

This record represents relatively low magnitudes compared to others obtained in this

experiment and to anticipated design wind speeds of approximately 100 mph (45 m/s)

.

However, it can be shown that this experimental data can be scaled with confidence to

the more intense wind speeds to provide data for extreme design wind speeds.

Further analysis of the time history data (figure 5) provide autocorrelation and

spectral density functions. The autocorrelation function is given in general form as

R(T) = ^ x(t)x(t + T)dt (1) T O

and the spectral density function is the Fourier transform of the autocorrelation

function or

1-25
S (f) = 4 f° R (T)cos 2-nf TdT

X ox (2)

Normalized autocorrelation and spectral density functions are given in figure 6 where

T = time lag = 0.08 seconds

R(I) = R(T) a'X

and

2 a = variance of x X

Also

NDS = fS (f) X

2 aX

These typical records and analyses form the basis to transform the random data gathered

from the experiment to forces on the mobile home in the form of design loads. It is con- venient to express the wind load in terms of mean and fluctuating load components. The

contribution of load fluctuations over specified frequency ranges can be determined from

the spectral density function where

^x " S^(f)df (3)

Using the spectral density functions of figure 6 for two frequency ranges 0.01 < Af < 10

2 2 and 0.01 < Af^ 1.0, values of and were obtained. The contribution to the fluctuating

force by the resonant response of the mobile home is eliminated in the Af^ range . Note

that the spectral density function shows a peak, the highest first mode natural frequency

of the mobile home at 3.7 Hz which is approximately equal to that measured when the mobile

home was mounted on temporary supports as previously discussed. By integration the ratio

''^xl *xl

= 1.044 or 7^ = 1.022

<2

was determined. This shows that approximately 2 percent of the fluctuating force is

contributed by the resonant response of the mobile home (f > 1 Hz) and the remainder of

the fluctuating force is contributed by the wind load which is considered, in this case,

as a quasi-static force. In terms of the mean force, u , = 119 lb (530N) , the contribuxl

1-26
tion by the resonant response of the mobile home and fluctuating wind load is 0.5 and 24

percent respectively. These observations are important in transforming the response data

to design criteria through the appropriate use of parameters such as gust factors.

5 . Summary and Acknowledgements

A research project has been described which is currently gathering data for the

development of design criteria for mobile homes to resist wind forces. The test setup,

instrumentation, data acquisition and analysis, and typical response analyses have been

presented. The entire program when completed will provide data for transformation to wind load design criteria in the areas of wind profiles below 10 meters, pressure distributions on mobile homes, forces on foundation systems transferred through mobile homes,

and structural response.

The assistance in the conduction of the program by Dr. Bruce Ellingwood, Mr. Charles

Yancey, Mr. Randolph Williams, and Dr. James Shaver of the National Bureau of Standards is

gratefully acknowledged.

NOMENCLATURE

f Frequency

R(T) Autocorrelation

S(f) Spectral density

V Mean wind speed

Z Height above ground

a Power law exponent

Mean force

0 Standard deviation of force

fluctuations

T Lag time

Fraction of total variance

1-27
H = 5m

SOLAR TEST HOUSE

H = 2lm

REACTOR BUILDING

H= 10 m H = 45m

ROADWAY

I'igure 1. Test site plan

1-28
it; o

<

1-29
<

I-zo

fsl o

I

<

UJ

q:

1-30
E

i£) 00

X

CC

Oo

a:a

UJmOo

(£I

I-

<rmoo

a:Q

CDoo

a:a

LUm— Ul

O.g9:*ri

—i r 1

L_.ij

LJ

IT<

bJ

a: 0 t - 1 ^

CO

0)

CO

CO

(Up.

toOOO3O

0)

00 H

1-31
1-32
R(I)

0.4

0.2

\ L_

0.00 44.44 88.89

\ \ L

133.33 177.78 222.22 266.67 311.11 SECONDS

AUTOCORRELATION FUNCTION

355.58 400.00

0.50 —

Q40

NDS

0.30

0.20

010 —

0.00

0.010 0.100

f, Hz

SPECTRAL DENSITY FUNCTION

10.000

Figure 6. Typical autocorrelation and spectral

density functions for force measurement

1-33
ON THE WIND RESISTANT DESIGN SPECIFICATIONS

FOR

THE PROPOSED HONSHU-SHIKOKU BRIDGES (1975)

T. OKUBO

Head, Planning and Research Administration Division

Public Works Research Institute

Ministry of Construction, Japan

N. NARITA

Chief, Structure Section

Structure and Bridge Division, Chiba Branch

Ministry of Construction, Japan

ABSTRACT

The newly revised specifications for the wind resistant design of the proposed

Honshu-Shikoku Bridges are introduced in this paper. The previous specifications (the

first edition was issued in 1964, and the second edition in 1972) have been revised by

taking into account recent research activities, resulting in final edition as established

in the fall of 1975.

The main subjects discussed in this paper are:

1. Scope of application

2. Basic wind and design wind speed

3. Drag coefficients

4. Gust response

5. Wind resistibility of structures under construction.

Key Words: Bridges; Specifications; Wind.

1. Introduction

The design of bridges to connect the Honshu (the mainland of Japan) with the Shikoku

(the fourth largest island) was initiated after formation of the Honshu-Shikoku Bridge

Authority in July 1970. The wind resistant design specifications, the first edition of

which was formulated in 1964, have been revised and applied in the preliminary design of

structures

In a previous paper, the authors introduced the main features of the wind resistant

design specifications for the proposed Honshu-Shikoku Bridges (1972) and related problems.

The significant points that were revised are as follows;

(1) To clarify the design procedure and the constitution of articles in the specifications .

(2) To reexamine the basic wind speed in due consideration of the recent wind data

in the Seto Inland Sea, and to set the basic wind speed for the respective zone

or route.
(3) To determine the modification factors for converting a basic wind speed into the

design wind speed, where the analytical method of the gust response was tentatively

adopted

.

(4) To revise the values of the drag coefficients, structural damping, and vertical

wind inclination by using the data which was recently obtained.

As the design work, progressed, the authority requested the formulation of design

specifications which could be uniformly applied to all of the bridges in the project.

In order to meet this request the former specifications had to be extensively revised.

These revisions are given in the latter part of this paper, the significant topics will

now be described.

2. Expansion of Scope

The (1972) specifications were applicable to a truss stiffened suspension bridge

with a span length of 500 meters or more. However, the scope of the (1975) specifications

have been expanded to bridges which may be affected by strong winds in the Inland Sea. Then the boundary for the respective route, in which the specifications are applicable, is

definitely stipulated. The basic wind speed, therefore , for an on-shore bridge, shall be

the same as used for the off-shore bridge in an identical route.

3. Basic Wind Speed

The basic wind speed should be determined by considering the observed wind speed,

durability of the structure, and the return period of the wind. The period of the wind

observation in the Inland district has not been extensive enough to estimate the return

values during a longer period. These wind periods have, therefore, been derived as follows

;

(1) Estimation from the correlation between wind record at the construction site

and that of the standard weather stations.

(2) Estimation from direct application of the Gringorten method to the observed

data at the construction site.

(3) Estimation from a multi-regression method.

The estimated return wind value has not been established, and only the basic wind

speed in the northern district of the Onomichi-Imabari route has been evaluated and is

equal to 37 m/s.

4. Design Wind Speed - Gust Response

The effects of the statistical characteristics of the natural wind on the dynamic

behavior of a structure should be taken into account on determining the design wind speed.

The fundamental meterorological data, indispensable in analyzing gust responses, were

collected in 1973. These data include power spectrum of the wind speed, frictional constant of the ground roughness, index of the space correlation of the wind, and other

variables

.

1-35
Parametric calculations of the gust response has been carried out for a series of

preliminary designs with different types of bridges, spans, and rigidities. Consequently

the following two types of wind speeds are defined; 1) basic wind speed, which is the ten minute averaged wind speed at a height of ten meters above sea or ground level, and 2) the

design wind speed. The design wind speed can be deduced from the basic wind speed by

modifying the effect to the altitude of the structure and the horizontal or vertical

length of the structure. The safety factor of self-excited oscillation (flutter) may

then be defined as the ratio of the flutter speed against the design wind speed. The load

magnification effect due to the gust is then considered in the design wind load. In this

context the consistency between the revised specifications and the design specifications

for highway bridges can be preserved.

5. Drag Coefficients

a) Longitudinal Wind Load

The stiffening truss of a suspension bridge moves longitudinally due to the wind

component in that direction, where its magnitude will have a great influence on the design

of the expansion joints and wind shoes. It has been specified in the specifications, as indicated by wind tionnel experiments , that sixty percent of the transverse wind load shall

be applied longitudinally to the stiffening truss of a suspension bridge.

b) Drag Coefficient of Solid Girder

The drag coefficient of a solid girder is given by the following empirical formula;

Cj^=2.1-0.1x B/D (1 £ B/D < 8)

= 1.3 (B/D > 8)

where C^^ B, and D indicates the drag coefficients, width and depth of girder respectively.

6. Wind Resistibility Under Construction

How much wind load should be applied to a structure during the construction state?

This is an important consideration since many accidents have occurred during the con- struction stage. The static and dynamic characteristics of a structure changes gradually

during the progress of the construction. Therefore, the aerodynamic stability of the

structure at each stage should be confirmed by a wind tunnel test using the full-model.

However, if the unit construction period is assumed as two years, the return value in

five years given the equivalent safety to a structure under construction as the completed

structure. The intensity of the wind loading for a structure under construction is

stipulated as half of the design wind load.

7. Conclusive Remarks

The following are the problems which should be examined in order to enhance the wind

resistant design method of bridges;

1-36
(1) To establish a unified method for estimating the return value of the wind speed.

(2) To establish quantitative evaluation of the wind load with regard to gust effects.

(3) To improve wind tunnel techniques with regard to gusty wind.

(4) To establish a more rational factor of safety.

I. General

A. Scope of Application

The following specifications as shown in Table 1.1.1, were applied to the wind

resistant design of the bridges in the Honshu- Shikoku Bridge Project.

Table 1.1.1 The district for scope of application

Route District

Kobe-Naruto

Maiko Viaduct ^ Akashi Straits Br. ^ Matsuho

Viaduct

Tozaki Viaduct ^ Onaruto Br. Muya Br.

Koj iraa-Sakaide Shimotsuiseto Br. Bannosu Viaduct

Onomichi-Imabar

i

Onomichi Br. Kurushima 3rd Br. (Limited

to the bridges across the straits)

B. Definitions and Symbols

1. Definitions

The following terminology is used in these specifications:

(a) Basic wind speed

Basic wind speed is the wind speed used as the basis of the wind resistant design, and is the average wind speed over a ten minute interval at a height of ten

meters above the ground or water surface at the site where the bridge is to be constructed.

(b) Design wind speed

Design wind speed is the wind speed used in computing the design wind load

on the structures and in establishing the wind stability of the structures against wind.

The basic wind speed shall be converted to the design wind speed by multiplying appropriate

1-37
modification factors to consider the height of structure, and the horizontal linear dimension or vertical linear dimension of the structure.

of the self-excited oscillation which occurs in the structure in relation to its aeroelastic

characteristics

(d) Wind direction

Wind direction is the direction of wind in the horizontal plane.

(e) Wind inclination angle

Wind inclination angle is the angle of wind inclination measured from

horizontal axis in the vertical plane, and is taken as positive when the wind blows upward.

(f) Design wind load

Design wind load is the computed air force obtained from the design wind

speed, and is the mean drag component of the air forces acting in the direction of bridge

axis or in the horizontal direction perpendicular to bridge axis

which is enclosed by the surrounding lines of the structural part under consideration. In computing the circumscribed area, handrail and guardfence are to be considered. However,

the effect of the transverse slope of the floor deck and floor system shall be ignored.

(h) Projected area Projected area is the average area per unit length of a structure projected

to the plane which is at a right angle to the acting direction of the wind load.

(i) Solidity ratio

Solidity ratio is the ratio of the projected area to the circumscribed area,

(j) Structural damping

Structural damping is the vibration damping of a sturcture in still air,

and is expressed by a logarithmic decrement,

(k) Angle of attack

Angle of attack is the angle between the direction of wind acting on the

structure and the horizontal reference axis perpendicular to the bridge axis. This angle

is taken as positive when the wind blows upward.

(1) Standard sea level

Standard sea level is the level for computing the elevation of the structure, where the datum of standard sea 1 evel is the mean sea level of Tokoyo Bay (T.P.)

2 . Symbols

In these specifications symbols are defined as follows:

(c) Critical wind speed for self-excited oscillation

Critical wind speed for self-excited oscillation is the onset wind speed

(g) Circumscribed area Circumscribed area is the average area per unit length of a structure.

(a) V.10

Basic wind speed (m/s)

Wind speed modification factor for the elevation of structure

Wind speed modification factor for the horizontal linear dimension

of structure.

1-38
(d) V : Wind speed modification factor for the vertical linear dimension

of structure.

(e) : Wind load modification factor for the horizontal line-like structure

(f) : Wind load modification factor for the vertical line-like structure

(g) : Design wind speed (m/s)

(h) Z : Elevation above ground or sea level (m)

(i) a : Power index in the power law of wind profile

2 (j) : Mean drag force on unit length of structure (kg/m )

(k) Drag coefficient

(1) 6 : structural damping

2

(m) A : Circumscribed area (m /m) g 2 (n) A : Projected area (m /m) n

(o) ^ : Solidity ratio

2 -4,

(p) p : Air density (kg's 'm )

II. Static Design

A. Basic Wind Speed

The following basic wind speed shall be applied to the bridge structure of the

Honshu-Shikoku Bridge according to the district in which the bridge is to be constructed.

Table 2.1.1 Basic wind speed

Route District Basic wind

speed (m/s)

Kobe-Naruto

Akashi Straits district 43

Naruto Straits district 50

Koj ima-Sakaide Whole Route 43

Onomichi-Imabari

Kurushima Straits district 40

Others 37

B. Design Wind Speed

The basic wind speed V^^ shall be converted to the design wind speed by

multiplying by modification factors as given in the following equations:

for horizontal line-like structure: = ^2*^10

for vertical line-like structure: V = V^' ^^2*^10

1-39
Modification factors V^, V^f and shall be taken as shown in table 2.2.1, 2.2.2 and

2.2.3 respectively. When the value is not shown in the table, it shall be derived from

proportional allotment and be computed to three decimal places and counting fractions of

.5 and over as a whole number and disregarding the rest. In the area where the wind

convergence due to special topographical conditions such as cape or peninsula is anticipated, the modification factor may be modified by taking into account the results of

field observations of the wind profiles or results from wind tunnel experiments on topographical models. The elevation and the length to be used in computing the modification

factors V , V , and V are shown in Table 2.2.4.

Table 2.2.1 Modification factor Vi

Elevation (m)

< 10 20 30 40 50 60 70 80 90 100 120 140 160 180 200

^1 1.00 1.10 1.17 1.22 1.26 1.29 1.32 1.35 1.37 1.39 1.43 1.46 1.49 1.51 1.53

Table 2.2.2 Modification factor V2

Horizontal length (m) 200 300 400 500 650 800 1,000 1,200 1,500 1,800

V2 1.21 1.19 1.18 1.17 1.16 1.15 1.14 1.13 1.12 1.12

Table 2.2.3 Modification factor V3

Vertical length (ra) 80 100 120 160 200 260

\)3 1.26 1.25 1.24 1.24 1.24 1.23

1-40
Table 2.2.4 Definition of elevation and length of structural element

'"'Bridge

\ype \ Factorx^

Suspension bridge. Cablestayed girder bridge Others

Suspended structure of

suspension bridge and

girder of cable-stayed

bridge:

mean elevation of

center span.

The higher elevation between

the mean elevation of support + [ (the maximum elevation of

structure) - (the mean elevation of support)] X 0.8 and

the maximum elevation of

designed road level.

Cable and hanger:

^1 mean height between the

elevation of the tower

top and the lower

surface at the midpoint of stiffening

structure or bridge

girder at the center

span

.

Tower:

65 percent height of

tower

V2 Center span length or maximum span length

^^3 Height of tower

C. Design Wind Ijoad In the static design, only the mean drag component of the air force shall be

considered. The wind load shall be determined as follows:

1. Bridges with a maximum span length more than 200 meters for horizontal linelike structure

:

D 2^ D 4 D n

for vertical line-like structure :

^

1 2 D 2 D 5 D n

2 -4

in which p, air density, shall be taken as 0-12 kg-s 'm . The drag coefficient C^^ and the

projected area shall be calculated by data given in Table 2.3.1 and 2.3.2. The modification factors V. and shall be chosen from Table 2.3.3.

4 D

1-41
•H

60co

I3 C

00 o

e -H

td -u u o

o 01

0) to u

o o

!-i O

-a

0)

-a

•w

COCo

a

0)

S-i 4) -a

(LI

0) TD

•H

cnc

oo

o

0)

•a

<u •H

tocoo

c

•HQ ID O to c

!-J O

4-1 (U -H O >

cn u

4-1 C 0)

01

cnu

cu>

cnC

cfl H

O

T3

C

(U 1-1

03

4-1 4-1 •H 4-1

cn iH

I ^ o

r-l O -U 4-1 •H

4-< 5

O

cn O CD

0) cn cn cn O D

•H O O

iH D 4-1

P. 4-1 ca <; CO 4-(

c 0) cn -H B

<f c 0)

• 0) rH O 4-1 0)

4-1 c; 0 o

•H

SJ 4-) 01 O

3 (U o cn

4-1 CO

4-4 CO O O

1-1 4-1 O

4-iCO M ^ CO CO rH3

0) 00 C C

0 cd

4-1 u u

01 0)

P-i *-i

-ecao

4-1 4-J O o

CO U 0)

4-J -H

c: 4-1 01

CJ <f o c

4-1 qj

01 OJ

CO £

cn cs

A I Qm

A I CO

O Q

T-H pa

CCJ 1-1 p 4-1

4-i c

CJ 0) 3 e u 01

4-1 .H

CO 01

0)

t-l00 Ai C a

•H 01 CO T3

0)

3 O

O 01 Q -d

J33op

-a cu

01 u

3

C 4-1 01 CJ O. 3

cn3 4-1 CO cn

U

0)00C

nj

cd

01CO

T3

•H

l-l

cn

cn3H

01 a!-i •H

00

-O

•H

i-lO

CO

0300 a 0) •Hm 4-)

01oo 3 -a

1-42
Table 2.3.1 (Supplement)

Shape of cross section

Drag coeff ic lent

Shape of cross section

Drag coefficient

\ circular

c y -L xnu cL 1.2

]^ rectangular

"^H—

^

[( ): r>d/12]

a 1.5(1.1)

flat plate 2.2

1 square

^ cylinder

[( ): r>d/12]

2.1(1.5)

-> 1

1 flat plate 1.8 _^ L rectangular

1 j cylinder

0.7

2.7

/\ square

cylinder

1.5 ->

1

rectangular

2 cylinder

[( ): r>d/29]

2.3(2.1)

Table 2.3.2 Projected area

Bridge

type

Structural

element

Transverse Longitudinal

Suspension

bridge

Suspended

structure

Windward side of stiffening frame, floor deck,

curb and handrail

Tower

Windward and leeward side

of tower shafts

Two tower shafts

and web members

Cable

Windward and leeward side

of cable

Not considered

Hanger

Windward and leeward side

of hanger

Bridges

except

suspension

bridge

Truss

bridge

Windward side of truss,

floor deck, curb and

handrail

Not considered

Solid

girder

Windward side of girder,

curb and handrail

Others Effective cross section

Table 2.3.3 Modification factor , V5

^~^\Direction

Factor

Transverse Longitudinal

1.3 1.0

^5 1.0

1.0

free-standing tower: 1.2

1-43
Moreover, the designed sections based on the herein specified value of drag

coefficients, shall be examined by wind tunnel model test as occasion demands and these

drag coefficients obtained from such wind tunnel tests can then be adopted. Wind tiinnel

model tests shall always be carried out for the suspended structure of a suspension bridge,

and when the drag coefficient to the transverse wind of zero degree angle of attack

measured by wind tunnel tests differs more than 5% of herein specified value, then the

structural section shall be redesigned by taking into account the measured values.

When calculating wind load together with live load, V^^ shall be limited to 30 m/s.

2 . Bridges with Maximum Span Length Less Than 200 m

In the case of live loading: P = P

' D D where P^^' is the wind load specified in "the Specifications for Highway Bridges".

3 . Wind Load Acting Upon Live Loads

When wind effects on live loads are to be considered, the following wind load

perpendicular to the bridge axis shall be applied.

D. Way of Loading

The wind load is in principle to be considered as a horizontal uniform load acting

in the transverse longitudinal direction of bridge axis. But, if necessary, the full and

half wind loads shall be applied to the most unfavorable part and the remaining part of the

structure, respectively.

The wind load shall be determined as follows:

1. Suspension bridge, cable-stayed girder bridge and the bridge with maximum span

length greater than 200 m

(a) Transverse wind load

Suspended structure of a suspension bridge, deck girder of a cable-stayed

bridge and truss bridge:

In the case of a storm:

road deck 150 kg/m

railway deck: 450 kg/m

Full wind load shall be applied on the windward side of structure. In case of a suspension bridge, wind load acting upon hangers shall be

equally divided to the stiffening girder and cables.

Cables of a suspension bridge and cable-stayed girder bridge:

In the case of a suspension bridge the wind load shall be applied to the

axial line of the cable. In the case of a cable-stayed girder bridge,

the wind load on the cables shall be equally divided to the girder and

tower

I

Towe r

f

The wind load shall be applied to the axis line of the windward and

leeward pillar of the tower, respectively.

r

1-44
(b) Longitudinal Wind load

Suspended structure of a suspension bridge

:

The longitudinal wind load is to be considered as a uniform load. In this case, the wind load on cables and hangers shall be ignored.

Tower

:

The wind load on a tower shall be calculated using the projected area, including tower shafts and web members, and shall be considered as uniform

load.

Cables, hangers, and girders of cable-stayed girder bridges;

Ignored.

2. Solid bridge with main span lengths greater than 200 m

(a) Transverse wind load

Full wind load shall be applied to the windward side of the girder.

(b) Longitudinal wind load

Ignored.

3. Bridges with main span lengths less than 200 m

The wind load shall be applied in accordance with the Specifications for

Highway Bridges, except the design wind load is corrected as specified in article 2.3.

However, the wind load for a railway bridge shall be applied in accordance with the Standard

Specifications for Steel Railway Bridges.

E. Combination of Design Loads and Allowable Stress

Combinations of the wind load with other design loads will permit an increase in

the allowable stress for each combination as given in Table 2.5.1.

W : Wind load

D : Dead load

W(L) : Wind load applied together with live load

L(W) : Live load applied together with wind load W(L) T : The effect of temperature change

B : Braking load, Starting load

LF : Rolling stock lateral load

LR : Longitudinal load of long rail

Each load shall be determined by "The design criteria of superstructures of the

Proposed Honshu-Shikoku Bridges", but the maximum temperature combined with wind load can be considered as 35°C.

F. Proofing of Lateral Buckling

In the case of a suspension bridge, the lateral buckling of the suspended , structure

due to wind force shall be proofed. The critical load for the lateral buckling must

exceed 1.7 times of the design wind loads. The aerostatic coefficients used in the

estimation of the critical load are to be obtained from wind tunnel model testing.

1-45
Table 2.5.1 Combination of loads and coefficients of

increase of allowable stress

Coefficients of increase

of allowable stress

Combination of loads

Main structure of

bridge with span

length more than

200 m and stiffening girder and

tower of suspension

bridge

Other structures

D+W+T 1.50 1.35

W 1.50 1.35

D+W(L)+L(W)+T 1.25

B-)-W(L) 1.25

LF+W(L) 1.25

D+W(L)+L(W)+LR 1.40

D+W(L)+L(W)+LF 1.40

D+W(L)+L(W)+B 1.40

D+W(L)+L(W)+B+LR 1.50

III. Proofing of Dynamic Behavior

A. Outline

For the design of flexible structures, like suspension bridges or cable-stayed

girder bridges, the investigation of the aeroelastic characteristics of the structure

and the confirmation of the safety against wind action shall be performed. For this

purpose, the dynamic proofing of the structural system, obtained through the static

analysis, shall be conducted. The wind tunnel testing for the dynamic proof is to be

based upon "the Standards of wind tunnel tests for the proposed Honshu- Shikoku Bridges

(1973) ".

B. Structural Damping

Relative to the vibration mode, the damping capacity of a structure can be

assumed as following:

(i) for the whole system of a suspension bridge and a cable-stayed truss

stiffened bridge: ^ -

(ii) for the whole system of a cable-stayed solid girder bridge

6 = 0.02

(iii) for the free standing tower: <5 = 0.01

1-46
C. Vertical Inclination of Wind

The relation between the wind speed and the vertical inclination of the wind

shall be assumed to lie in the domain shown as in Fig. 3.3.1. When the wind characteristics at the site where the structure is to be constructed shows a singularity, this

condition shall be considered.

-50 .

/ ^

speed - 25 N

Wind

— j~ — -7 -3 0 3 7

Vertical inclination of wind (degree)

Fig. 3.3.1 Wind speed versus vertical inclination of wind

D. Critical Wind Speed for Self-excited Oscillation

The critical wind speed for a self-excited oscillation, derived from a dynamic

analysis or wind tunnel test, should lie beyond the domain enclosed by the line of the

wind speed of 1.2 times of design wind speed, and the lines shown in Fig. 3.3.1.

The critical wind speed shall be defined as the lowest wind speed at which the

self-excited oscillation, with amplitude of one to five degrees, takes place.

E. Restricted Oscillation

When the occurrence of the restricted oscillations of the structure or elements

are anticipated at the wind speed less than the design wind speed, a sufficient countermeasure relative to fatigue of the elements of disfunctioning of the structure, shall be

adopted.

IV. Others

A. Examination to be taken during Construction

Enough care shall be exercised to prevent serious wind effects on the structure,

not only at the time of its completion, but at each stage during construction.

Special care of free-standing towers, cables or stiffening girders during con- struction should be noted.

1-47
The static and dynamic safety of a structure against wind effects during each step

of construction shall be confirmed through appropriate techniques such as the wind tunnel

tests, and suitable countermeasures shall be taken if necessary.

Wind load during construction as a standard is to be half of design wind load.

Reference

Wind resistant design specifications (1975) for the proposed Honshu-Shikoku Bridges

and its commentary: JSCE, March 1975.

1-48
EQUIVALENT STATIC WIND LOADS FOR TALL BUILDING DESIGN

EMIL SIMIU

Structural Research Engineer

Center for Building Technology

National Bureau of Standards

Washington, D.C. , U.S.A.

ABSTRACT

Certain shortcomings of current procedures for computing along wind structural response

have been shown to result in unrealistic estimates of tall building behavior under the

action of strong winds. Differences between predictions of fluctuating response based on various such procedures have been shown to be in certain cases as high as 200%. In recent

years, advances in the state of the art have been made which provide a basis for significantly improved alongwind response predictions. The purpose of the present work is to present a procedure for calculating alongwind response, including deflections and accelerations, which incorporates these advances. The meteorological and aerodynamic models on which the procedure is based are briefly described. The practical use of the procedure is

illustrated in a numerical example. Estimates are provided of errors inherent in the

models employed. The range of applicability of the procedure is defined, and it is indicated that for structures with unusual modal shapes or for which the influence of higher

vibration modes is significant, a recently developed computer program should be employed

in lieu of the procedure presented herein.

Key Words: Design; Equivalent loads; Tall Buildings; Wind loads.

1. Introduction

Several procedures for computing alongwind response have been proposed in the last

decade []^, 2^, 3^] . The purpose of these procedures is to calculate equivalent wind loads

whose effect upon the structure is the same as that of the gusty wind.

As part of an effort aimed at evaluating and improving building code provisions on design for wind loads, an analysis of current procedures for computing alongwind structural

response - including the procedures described in the American National Standard A58.1 [4] and in the National Building Code of Canada [5_, 5_] - has been presented in ref. 7. From

this analysis, the following points have emerged. First, the alongwind cross-correlation

of the fluctuating pressures is represented in current procedures by models which are

f\andamentally inadequate [8^] . It is primarily for this reason that the differences between the values of the fluctuating part of the response calculated in accordance with

the A58.1 Standard, on the one hand, and the National Building Code of Canada, on the

other hand, may in certain cases be as high as a few hundred percent [7^] . Second, current

1-49
procedures do not take into account the dependence of the longitudinal wind velocity

spectra upon height above ground. This was shown to result in significant overestimates

of the components of the fluctuating velocity which produce resonant excitation in windsensitive structures []_, 9] . Third, current procedures do not enable the analyst to estimate the contribution of the higher vibration modes to the structural response. Fourth,

existing procedures do not provide an estimate of the alongwind accelerations induced in

structures by the gusty wind.

In the years following the development of the procedures described in refs. 2^, 3^, 4_ and 6, developments of significance in the context of determining structural response to wind have taken place notably in the area of boundary layer meterology. Progress has also

been made in understanding and evaluating the effect of the alongwind pressure cross- correlations upon building response. Finally, effective numerical methods have been

developed, from which economical computer programs, suitable for the analysis of structures

subjected to wind loads, could evolve. In ref. 10, a procedure was presented for calculating alongwind response which incorporates and utilizes these advances and is thus free of

the shortcomings mentioned above of the procedures currently in use. If the modal shapes

of the structures analyzed are unusual or if the contribution of the higher modes to the

response is significant, the procedure requires the use of a computer program. A machineindependent, Fortran version of this program is listed in Ref. 10. However, in a large

number of cases of practical interest, it may be assxamed that the fundamental modal shape

is linear snd that the response is dominated by the fundamental mode. In such cases, the

procedure may be simplified, i.e., graphs can be developed which may be used for rapidmanual calculations of the alongwind deflections and accelerations.

The purpose of this paper is to present this simplified version of the procedure and

to illustrate its practical use. The meteorological and aerodynamic models on which the

procedure is based are briefly described. Results of numerical calculations are used to discuss some of the approximations and errors inherent in the models employed, as well as the range of applicability of the procedure presented herein.

2. Meteorological and Aerodynamic Models

a) Mean Wind Profile . In most structural engineering applications, mean wind profiles

in horizontal terrain of uniform roughness over a sufficiently large fetch are currently

modeled by the so-called power law, first proposed by Hellman in 1916 (11) . However, as noted by Pasquill (12) , the logarithmic law,

U(z) = 2.5u.\n- — (i)

where U(z) = mean speed at height z above ground, a^ = zero plane displacement, z^ = roughness length and u^ = friction velocity, is a superior representation of mean wind profiles

in the lower part of the atmospheric boundary layer. Balckadar and Tennekes (13) , Tennekes

(14), Carl, Tarbell and Panotsky (15), Helliwell (16) and others have shown that for high

wind speeds, such as are assumed in design, the logarithmic law may be applied throughout

1-50
the height range of interest to the structural engineer.

The use of the logarithmic law in structural design is believed to be justified for

the following three reasons. First, as was just mentioned, eq. (1) represents an improvement over the power law. Second, since the logarithmic law is universally employed in

meteorological work, results of meteorological research may readily be used in engineering

applications. Errors inherent in attempts to reformulate such results in terms of the

power law are thus avoided, as is unnecessary and costly duplication of research. Third,

in the logarithmic representation, the mean speed - just as the spectrum of the longitudinal velocity fluctuations (see eq. (5)) - is explicit function of u^. Thus, if the

logarithmic law is used, an internally consistent representation of the flow structure is

achieved. This is not the case if the mean profiles are modeled by the power law, in which

the exponent is not clearly and uniquely related to u^. For these reasons, the logarithmic

law will be employed in this work. To account for flow disturbances by the building near

the ground, following common practice (6) , it will be assumed that, for 0<z<z,+10 d

meters, U(z) = U{z, + 10).

a Suggested values of the roughness length z are given in table 1 (see refs. 16, 17, 18), o For example, at Sale, Australia, for terrain described as open grassland with few trees,

at Cardington, England, for open, farmland broken by a few trees and hedge rows, and at Heathrow Airport in London, z = 0.08m (16,17,19). At Cranfield, England, where the ground

o upwind of the anemometer is open for a distance of half a mile across the corner of an airfield, and were neighboring land is broken by small hedged fields, z^ = 0.095m (17,20).

3. Equivalent Static Wind Loads for Tall Building Design

The zero plane displacement may in all cases be assumed to be zero, except that in

centers of large cities the smaller of the values z^ = 20m and z, = 0.73h, where h =

a a average height of buildings in the surrounding area, may be used (16)

.

In designing tall buildings it is reasonable to use mean wind speeds averaged over a period of one hour (4,6). If the mean winds are averaged over periods t different

from one hour, the mean winds averaged over one hour may be obtained using Table 2 (21)

.

For values of t not included in Table 2, linear interpolation is permissible. If the

wind speeds are given in terms of fastest miles, the averaging period t in seconds is given

by

t = 3600/v^ (2)

where v^ is the fastest mile in miles per hour.

a) Relation between Wind Speeds in Different Roughness Regimes . Consider two adjacent

terrains, each of sufficiently large fetch, the roughness lengths of which are denoted

z^^, z^, and over which the mean wind speeds may be described by eqs. (3) and (1) , respectively: z-z,,

U,(z) = 2.5m., In (3)

1-51
The relation between the friction velocities u^^, may be obtained, as a function

of z ,

, z , from Fig. 1. The latter was developed in Ref . 17 on the basis of theoretical

ol o and experimental work reported by Csanady (22) and others (23)

b) Longitudinal Velocity Fluctuations . The cross-spectrum of the longitudinal

fluctuating velocities at two points P^, may be expressed in the form:

S„(P,,P2.") = S'J^ {P,.n)S['HP:,,n) R„(y , , y^, z

, . z^, ") N„(P, , Pj. ") (4)

in which y^, z^ are the coordinates of point P^(i=l,2) in a plane perpendicular to the

direction of the mean wind and 5^(P,n) is the spectral density of the longitudinal velocity

fluctuations. Expressions of the spectrum S^(P,n) which take into account its dependence

upon height above ground and are consistent with theoretical and experimental results of

recent micrometeorological research have been presented in Ref. 9. In this work, the

expression

nS„(r!) 200/

ul (1 + 50/)

(S)

developed in Ref. 9, was used. In eq. (5), f = n(z-z)/U(z). The mean square value of

2 2 the fluctuations described by eq. (5) is u (z) = 6u . The functions R and N in eq. (4) are defined as the acrosswind and the alongwind

u u cross-correlation coefficient, respectively. By definition, if the points P^, P2 are in

the same plane perpendicular to the mean wind direction, N iV^iP^'^^ ~ Otherwise,

N^(P^,P2,n) < 1; its magnitude decreases as the alongwind separation between P^jP^, and

as the frequency n, increase. The acrosswind cross-correlation coefficient is a complex

quantity, the modulus of which is known as the square root of the across-wind coherence

function (24). However, its imaginary part has been determined by measurements to be

negligible for wind engineering purposes (k,2,3,25). The following expression for the

acrosswind cross-correlation coefficient, proposed by Davenport (3) , has been used

herein:

«J>',.z,.y2.Z2'n) = exp 1 ilLj (g, l/(z,) + t/(z,)

in which C = 10, C =16 (3).

z y

c) Relation Between Wind Pressures and Wind Speeds. The fluctuating pressures, p',w

p^i on the windward and leeward sides of a building are assumed to be linearly related

to the velocity fluctuations u, i.e.,

p;(P) . pC^U(z)u(:) (7) p'(P) - pCj^U(:)u(:) (8)

1-52
in which

(9)

(10)

p^, are the average values of the mean pressure at elevation z on the windward and the

leeward faces of the building, respectively. The model represented by eqs. (7) -(10) is

discussed in some detail in Ref. 10. In this work, it was assumed C = 0.8, C„ = 0.5.

W X, It is implicit in eq. (8) that the pressure fluctuations on the leeward side are proportional to the velocity fluctuations in the upstream flow. As noted in Ref. 8, in

reality the flow in the wake of the building is quite dissimilar from that in the oncoming

flow. Measurements suggest that the fluctuating pressures on the leeward side are small

compared to those on the windward slide and, therefore, that eq. (8) may be slightly con- servative (8,26,27).

d) Cross-Spectra of Fluctuating Pressures . If eq. (7) or (8) is used,

Sp(P^.P,.n) . Cp(P^)C^,crjp^U(:^)U(:3)

^^^^ S^(Pj.P,.n)

in which the cross-spectral density of the velocity fluctuations can be written as

S^(Pl.P..n) » Sy^P^.n)Sy^^,.a)R^(Pl,P,.^)

N (P,,P,,n)

u 1 •!

and in which C (P.) = C or according to whether P. (P, ,2) is on the windward or leeward

p 1 w 2 1 1 side. Useful information regarding the magnitude of N^(P^,P2,n) is provided by results

of measurements of cross-spectra of pressures on the windward and leeward sides of structures. Such measurements, carried out on full-scale buildings, have been reported by

Lam Put (29) and by Van Koten (27) who found that except for very low frequencies,

N (P. ,P„„,n) < 0.2 (the subscripts w, I indicate that points P., P„ are on the windward

u Iw zJc L z and leeward sides, respectively; the points P, ,P, „ and P^ ,P^n have coordinates y,

,

Iw Ix, 2w 2ii- 11

and y^, z^, respectively). From the results of wind-tunnel measurements reported by Kao

(Fig. 5.1, 5.19, 5.21 of Ref. 29) it also follows that, except for extremely low frequencies corresponding to eddies of negligible energy, N(P^^,P2£fn) is nearly zero. It was

shown in Ref. 10 that it is reasonably conservative to assume N = 1 for 0 < n < 0.9n^ and

N = 1 for 0.9n^ < n < °°. This assumption will be used in the present work.

4. Alongwind Deflections and Accelerations

Let B, H denote the building width and height (in meters) ; m, its mass per unit height

(kg/m) ; n , its fundamental frequency of vibration (cycles/s) ; n, the damping ratio; g , X. a

g", the ratios of maximiup probable value to r.m.s value of the fluctuating deflections and

accelerations: a' (z) and a"^ (z) ,^r .m. s . of the alongwind deflections and accelerations

(in meters); G.F., the gust response factor; a(z) and a(z), the maxim\im probable alongwind

deflection and acceleration, respectively (in meters) . Using the assumptions stated in
3 this paper, and the air density being p - 1.25 kg/m , the expressions developed in Ref. 9 can, as shown in Ref. 10, be written in the form;

g (z) = (2tn 3.600 v^]'^^

0.577

(13)

(14)

(15)

[iVn 3,600 V ]^^'

^^^^^ ' ^ITF^'''"^

'^^'^

G.F..l*g, (17) * i(0

a (2) = (G.F.) 3(:) (18)

jrjTjl/2 so 09)

,1/2 O.iTT

5 (I) = [Cin 3.600a ] - ^ ^ [2in 3.6C0n,l

1/2 a^^^(z) . E^a'C) (21)

The equivalent static wind load is the product of the gust factor, G.F., by the mean wind load.

J h] = j ('-/-A + ^000 + + iz]) (22) \H H J \H Hi

z. ^ = (^1 - -j ^ (23)

where all dimensions are in meters. J can be obtained form Fig. 2 and k^, k^, k^ are given in Table 3. ^ can be obtained from Fig. 3. For z, ^ 0, eq. (23) is approxid mate; the approximation has been verified by numerical calculations to be of the order of

1: (10). In eqs. (14), (15), and (17), the quantity J2/ , associated with the resonant part

f the response, may be written as

Ci + 2C^CiN^ + Cf-nf, . /B z„ D .

1-54
if C = 0.8. C, = O.S. N = 0..;,

' w J.

A. ' .562 (2S)

in which f = . The values of Y can be found in Figs. 4 thourgh 7 (linear inter1 u^ 11

polation within and between Figs. 4 through 7 is permissible) . The Y^^ curves are identified by capital Roman letters to where there correspond ratios z /H, z,/H, listed

o d

in Table 4. Suggested values for the mechanical damping ratios n of steel frames and

reinforced concrete frames are 0.01 and 0.02, respectively (4,6). Lower values of n may

have to be used, for example in the case of welded steel stacks, of certain prestressed

structures (6) or of structures of the framed tube type (33) . In addition to the mechani

cal damping, the aerodynamic damping may, in principle be also taken into account. Aerodynamic damping is associated with changes in the relative velocity of the air with res- pect to the building as the latter oscillates about its mean deformed position (see Ref.

25) . It appears that it may be unconservative to rely upon the effect of the aerodynamic

damping; for this reason, the latter is not taken into account in Refs. 4, 6, and will

also be neglected in this work.

5. Numerical Example

Consider a building for which the height H = 400m, the width B = 66m, the depth (i.e.

the dimension in the alongwind direction) D = 45m, the damping ratio n = 0.01, the funda3 mental frequency n^ = 0.09 cycles/s, the weight per unit volume w = 1,500 N/m . The

basic wind speed (i.e., the fastest mile wind at 10m above ground in open terrain for

which z , = 0.07m) is v^ = 75mph (1 mph = 0.4474 m.s) . The building is located in the

o: r center of a large city for which it may be assumed = 0.65m, z^ = 20m.

The mass of building per unit height is m = w xBxD/g = 445,000 kg/m, where g = 9.81

, 2 m/s . The friction velocity, u^ for the builtup terrain, is obtained as follows:

Averaging time for fastest mile: t = 3600/75 = 48s (eq. (2)); ration uVu": 1-26 (Table

2); mean hourly speed at 10 m elevation in open terrain: U^(IO) = v^/1.26 = 26.8 m/s;

friction velocity in open terrain: u... = U.,(10)/(2.5 in (lO/z )) = 2.16 m/s (eq. (3)); *1 1 ol

= l-l"^ (Fig- 1) '• = X 2.16 = 2.54 m/s.

The quantities, j, ^, .g?, are obtained as follows:

.6 -X i 0

= in. 5. 35.2^, k. - 3.76 (table,3); J = 17.1 (fig.

2); a 5.9 (fi3. (table 4 ar.d - S.45 [eq. {23)

]

3)

;

0;

s.

5) ; J -"17

.8 X lO"

,1 [eq. (::}];3~

1-55
The quantities sought are

i(z) - 0.76 i/H (meters] [eq. (13]] -1/2

- 0.273 ls<l. (14)]

» 3.46 [eq. (IS) and (16)

G.F. . 1.96 [eq. (17)]

"llii

' (neters) [eq. (IS)]

rrir^''^ • 0.04S z/H (m/s^) [eq. (19)]

g- - 3.57 (Jq. (20)]

2

- 0.16 n/s = 0.016 s [eq. (21)]

a) Comparison Between Gust Response Factors Calculated Using Various Current Procedures . For the building just considered, the values of the gust response factors calculated

in accordance with the procedures described in the ANSI A58.1 Standard (4), the Canadian

Structural Design Manual (6) and Ref. 3 are 1.33, 2.83 and 3.38, respectively, versus

1.98 as calculated herein. As noted in Ref. 7, the gust factors calculated using Refs.

3 and 5 are overestimated, first, because the pressures on the windward and leeward sides

are conservatively assumed to act in phase (i.e., to be perfectly correlated) and second,

because the variation with height of the spectrum of the longitudinal velocity fluctuations

is ignored. Gust factors are underestimated by the procedure of Ref. 4 because the alongwind pressure correlation coefficient, rather than being applied as a reduction factor to

just the cross-spectrum of pressures acting on opposite sides of the building, is applied

to the entire alongwing response, in violation of basic random vibration theory (see Ref.

3) .

6. Approximations and Errors in the Estimation of the Alongwind Response

In this section, results of numerical calculations are presented on the basis of

which estimates can be made of the errors associated with uncertainties regarding certain

features of the models employed. The calculations were carried out for two typical

buildings selected as case studies (Table 5) . The wind speed at 10m elevation in open

terrain (z = 0.07m) was assumed to be v^ = 75 mph (1 mph = 0.447 m/s) , where v. = fastest

o f f mile of wind.

a) Contribution of Higher Vibration Modes to the Response. Let n^, n^ , n^ denote

the natural frequencies in the first three modes. For both buildings in both open and

center of large city exposures, the percent contribution of the higher vibration modes

to the r.m.s. of the flutuating alongwind deflections was of the order of 5-15%, if it was assumed that n^/n^ = 1.2, n^/n^ = 1.5, and about 1% if n^/n^ = 2.5, n^/n^ = 5. The

contribution to the r.m.s. of the alongwind accelerations was of the order of 10-20% if

n./n = 1.2, n /n = 1.5 and about 3-10% if n /n = 2.6, n /n = 5. Typical modal shapes XI, 21 31

have been used in the calculations, as indicated in Ref. 10. 1-56
b) Influence upon Calculated Response of the Deviation from a Straight Line of

Fundamental Modal Shape. A convenient means for estimating the influence upon response

of the shape of the fundamental mode of vibration is provided by the expression,

-rrl/2

"

1 . V - ^ ? (25)

derived by Vickery (3) on the bases of the ass\jmptions that the power law holds, that the

spectra of the turbulent velocity fluctuations are independent of height and that the

fundamental modal shape is described as follows

:

where y is a constant. In eq. (26) , a = exponent of the power law and Q = function of

geometrical, mechanical and environmental parameters, independent of Y- It may be

assumed, roughly, that a can vary between 0.10 for open exposure and 0.40 for centers of

—TT 1/2 large cities. It follows from eq. (26) that, for a = 0.10, the ratios a /a calculated assuming y = 0.5 and y = 1.5 differ by less than 1% from that calculated assuming y

= 1.0 (i.e., a linear fundamental modal shape). For a = 0.4, the corresponding differences

are about 3%. It is thus seen that deviations from a straight line of the fundamental —TT" 1/2 modal shape have an insignificant effect upon the calculated ratio a /a.

c) Influence Upon Calculated Response of Errors in the Estimation of the Roughness

Length. To estimate the magnitude of the error associated with uncertainties regarding

the actual value of the roughness length, the response of buildings 1 and 2 was calculated for coastal (z = 0 . 005-0 . 01m) , open (0 . 03-0 . 08m) , suburban (z = 0.20-0. 20m), center

o o of town (z = 0.40m) and center of large city (z = 0.60-0. 80m) exposures. The zero

o o plane displacement was in all cases assumed to be zero. The results of the calculations

are shown in Table 6. It is seen from Table 6 that the sensitivity of the results to even large errors in

the estimation of the roughness lengths is tolerably small. It is also noted that the

alongwind deflections and, consequently, the design wind loads are higher by about 15%

in the case of the coastal, than in the case of the open exposure.

d) Spectra in the Lower Frequency Range and Alongwind Response . It was shown in Ref

.

9 that no universal relation exists that described the shape of the spectral curve in the

lower frequency range and that the peak similarity coordinate appears to vary, strongly

between sites and between atmosphere and laboratory. To estimate the effect of this

variation, the response of buildings 1 and 2 was calculated in the cases z^ = 0.07m, =

0 and a = 0.80 m, = 20m assuming values of the peak similarity coordinate f^ = 0.033

(eq. (5)) and f^ = 0.01, f^ = 0.10, f^ = 0.19 (eqs. (11) through (13) of Ref. 9). The

ratios (a )f,/(a )„ of the maximum probable response calculated using the values msx 1 max 0.033

of the peak similarity coordinate f^ and 0.33, respectively, are given in Table 7.

1-57
The results of Table 7 indicate that eq. (5) (to which there corresponds = 0.33)

is slightly conservative. It is noted that, in view of Eq. (19) , the influence of the

spectral curve shape in the lower frequency range upon the value of the acceleration is

negligible

e) Acrosswind Correlation of the Pressures and Alongwind Response . In current procedures for calculating alongwind response (1,2,3,4,6) it is assumed that the values of

the acrosswind cross-correlation function are nearly the same for the pressures and for

the fluctuating velocities in the undisturbed flow. According to Refs. 25, 29, this

assimiption has been verified in the wind tvinnel. However, full-scale measurements (30)

suggest that if the acrosswind cross-correlation of the pressures is represented by an

ejjponential function as in eq. (12), the values of the decay coefficients C^, are smaller than in the case of the fluctuating velocities, i.e., the acrosswind correlation

of the pressures is higher than that of the velocity fluctuations which generate them.

It also appears that, rather than being constants, the decay coefficients may be functions of the frequency n. For low frequencies, values as low as = 4 have been

reported (30)

It is therefore of interest to estimate the errors in the calculated alongwind

response that correspond to possible errors in the values C^, C^. The alongwind response

of buildings in open (z^ = 0.07m, z^ = 0) and large city (z^ = 0.80m, z^ = 0) exposures

was therefore calculated separately for C = 10, C =16 (case 1) , for C = 4, C =6.4

z y z y (case 5) and for four intermediate cases in which C , C were assumed either constant

z y throughout the frequency range (case 4) or to have lower values at low frequencies and

higher values near and beyond the fundamental frequency n^, (cases 2,3,5). Graphs similar

to Figs. 3-7, corresponding to C = 6.3, C =10 and C = 4, C =6.4 have been presented

z v z y

in Ref. 10. The ratios r. = (a )/(a ),, r. = (a )./(a ) , in which cases 1 and i are

1 max max 1 i max i max 1 denoted by the indices 1 and i, respectively (i = 2,3,4,5,6), are given in Table 8 for

large city exposure. The values of these ratios in open exposure are of the same order

of magnitude or slightly lower.

It is seen that even a considerable change in the values of C , C in the lower frez y quency range affects little the calculated response (cases 1,2,3). However, large

decreases in these values near the fundamental frequency of the structure do increase the

total response, by as much as almost 10-20% in the case of the deflection and 30-80%

in the case of the acceleration.

As mentioned previously, the data available so far suggest that the acrosswind

cross-correlation of the low frequency components of the pressures may be higher than

was reported in Refs. 25, 29 and assumed in Refs. 1,2,3, in the ANSI A58.1 Standard

and in the National Building Code of Canada. However, no persuasive evidence appears to exist to the effect that the correlation is also higher for frequency components in the

range of natural frequencies of structures. Even if this were the case, it must be

recalled that these components are actually attenuated to some extent (31,32); the possibly increased correlation would then, at least partly, be compensated by this attenuation.

1-58
The sensitivity of the calculated accelerations to changes in the values of exponential

decay coefficients suggests that measurements of accelerations, which are relatively

simple and inexpensive, might offer useful information on the actual magnitude of these

coefficients

.

7. Conclusions

In this paper a practical procedure for calculating alongwind deflections and accelerations has been presented, which accounts for the variation of wind spectra with height

and makes allowance for the imperfect correlation in the alongwind direction of pressures

acting on the windward and leeward building sides. For a specific case, a numerical ex- ample is given, and the gust response factor obtained is compared to those calculated

using procedures described in current codes and standards.

The procedure presented herein may be applied to typical tall structures for which

the ratios of higher to fundamental frequencies are not unusually low (e.g., for which

n^/n^ > 2) and for which the fundamental modal shape may be expressed as y^{z) = (z/H)Y,

where 0.5 < y < 1.5. Results of numerical calculations have been presented which show

that the effect upon the response of the possible variation of the frequency, n, for

2 which the reduced spectrum nS(n)/)j , reaches a maximum, is small. Information is also

provided on the errors in the estimation of the response associated with uncertainties

regarding the actual terrain roughness and the values of the exponential decay parameters.

It is noted that for structures with unusual modal shapes or for which the influence of

the higher vibration modes is significant, the computer program presented in Ref . 10

should be employed in lieu of the procedure presented herein.

Acknowledgements

The writer wished to express his appreciation to E.O. Pfrang, of the Center for

Building Technology, National Bureau of Standards, for his encouragement and support,

and to R.D. Marshall, of the Center for Building Technology, for useful comments and

suggestions offered during the preparation of this work. The capable assistance of R.L. Hocker, of the Center for Building Technology, is also gratefully acknowledged.

References

1. Davenport, A.G. , Gust Loading Factors, Journal of the Structural Division , ASCE,

Vol. 93, No. ST3, Proc. Paper 5255, June 1967, pp. 11-34.

2. Vellozzi, J. and Cohen, E. , Gust Response Factors, Journal of the Structural

Division , ASCE, Vol. 97, No. ST6, Proc. Paper 5980, June 1968, pp. 1295-1313.

3. Vickery, B. J. , On the Reliability of Gust Loading Factors, Civil Engineering

Transactions , April 1971, pp. 1-4.

4. American National Standards Institute, American National Standard Building Code

Requirements for Minimum Design Loads In Buildings and Other Structures , New York, New

York, 1972.

5. National Research Council, National Building Code of Canada , Ottawa, 1970.

1-59
6. National Research Council, Canadian Structural Design Manual , Supplement No. 4 to the National Building Code of Canada, 1970.

7. Simiu, E. and Marshall, R.D., Wind Loading and Modern Building Codes, Meeting

Preprint 2268, ASCE National Structural Engineering Meeting, Cincinnati, Ohio, April 22-26,

1974.

8. Simiu, E. , Gust Factors and Alongwind Pressure Correlations, Journal of the

Structural Division , ASCE, Vol. 100, No. ST9, Proc. Paper 9686, April 1973, pp. 773-783.

9. Simiu E., Wind Spectra and Dynamic Alongwind. Response, Journal of the Structural

Division , ASCE Vol. 100, No ST9 , Proc. Paper 10815, Sept. 1974, pp. 1897-1910.

10. Simiu, E. and Lozier, D.W. , The Buffeting of Tall Structures by Strong Winds , Building Science Series, Center for Building Technology, National Bureau of Standards,

Washington, D.C., 1975 (to be published).

11. Hellman, G. , Ueber die Bewegung der Luft in den untersten Schichten der Atmosphaere, Meteorol. Zeit . , Vol 34, pp. 273, 1916.

12. Pasquill, F. , Wind Structure in the Atmospheric Boundary Layer, Philosophical

Transactions of the Royal Society of London , A. 269, 1971, pp. 439-456.

13. Balckadar, A.K. and Tennekes, H. , Asymptotic Similarity in Neutral Baratropic

Boundary Layers, Journal of the Atmospheric Sciences , Vol. 25, No., 1968, pp. 1015-1020.

14. Tennekes, H. , The Logarithmic Wind Profile, Journal of the Atmospheric Sciences , Vol. 30, 1973, pp. 234-238.

15. Carl, D.M. , Tarbell, T.C., and Panofsky, H.A. , Profiles of Wind and Temperature

from Towers over Homogeneous Terrain, Journal of the Atmospheric Sciences , Vol. 30, July

1973, pp. 788-794.

16. Hellwell, N.C. , Wind Over London, Proceedings , Third International Conference of

Wind Effects on Building and Structures, Tokyo, Japan, 1971.

17. Simiu, E., Logarithmic Profiles and Design Wind Speeds, Journal of the Engineering

Mechanics Division , ASCE, Vol. 99, No. EMS, Proc. Paper 10100, October, 1973, pp. 1073-

1983.

18. Fichtl, G. , and McVehill, G. , Longitudinal and Lateral Spectra of Turbulence in

the Atmospheric Boundary Layer , NASA TN D-5584, Washington, D.C., 1970.

19. Deacon, E.L. , Gust Variation with Height up to 150m, Quarterly Journal of the

Royal Meteorological Society , Vol. 91, 1955, pp. 562-573.

20. Harris, R.I., Measurements of Wind Structure at Heights up to 598 ft above

Ground Level, Proceedings , Symposium on Wind Effects on Buildings and Structures, Loughborough University of Technology, Leicestershire, England, 1968.

21. Durst, C.S., Wind Speeds over Short Periods of Time, Meteorological Magazine , Vol.

89, 1960.

22. Csanady, G.T. , On the Resistance Law of a Turbulent Ekamn Layer, Journal of the

Atmospheric Sciences , Vol. 24, Sept. 1967, pp. 467-471.

23. Tennekes, H. and Lumley, J.L., A First Course in Turbulence , The MIT Press,

Cambridge, Mass., 1972.

24. Panofsky, H.A. and Singer, I. A., Vertical Structure of Turbulence, Quarterly

Journal of the Royal Meteorological Society , Vol. 91, July, 1966, pp. 339-344.

25. Vickery, B.J. and Kao K.H., Drag of Alongwind Response of Slender Structures,

Journal of the Structural Division , ASCE, Vol. 98, No. STl , Proc Paper 8638, January

1972, pp. 21-36.

26. Dalgliesh, W.A. , Statistical Treatment of Peak Gusts on Cladding, Journal of the

Structural Division , ASCE, Vol. 97, No. ST9, Proc. Paper 8356, Sept. 1971, pp. 2173-2187.

27. Van Koten, H. , The Fluctuating Wind Pressures on the Cladding and Inside a Building, Proceedings

,

Symposium on Full Scale Measurements of Wind Effects on Tall

Buildings, The University of Western Ontario, June 23-29, 1974.

1-60
I

28. Lam Put, R. , Dynamic Response of a Tall Building to Random Wind Loads, Proceedings , International Conference on Wind Effects on Buildings and Structures, Tokyo, Japan, 1971,

pp. III-4.1-II.

29. Kao, K.H., Measurements of Pressure/Velocity Correlation on a Rectangular Prism

in Turbulent Flow , Engineering Science Research Report BLWT-2-20, University of Western

Ontario, London, Canada, 1971

30. Newberry, C.W. , Eaton, K.J., and Mayne, J.R., Wind Loading on Tall Buildings - Further Results form the Royex House, Industrial Aerodynamics Abstracts 4, July-August,

1973, pp. 1-16.

31. Bearman, P.W., Some Measurements of the Distortion of Turbulence Approaching

a Two-Dimensional Body, Journal of Fluid Mechanics , Vol. 53, part 3, pp. 431-467, 1972.

32. Marshall, R.D., Surface Pressure Fluctuations Near an Axisymmetric Stagnation

Point , NBS Technical Note 563, National Bureau of Standards, August 1971.

33. Taoka, G.T., Hogan, M. , Kahn, F.R. and Scanlan, R.H. , Ambient Response Analysis

of Some Tall Structures, Journal of the Structural Division , ASCE, Vol. 101, No. STl, Proc. Paper 11051, January, 1975, pp. 49-65.

1-61
Table 1 - Suggested values of z ioT various typei of exposure T/pe of exposure I (meters)

I o Coastal

I O.OOS-0.01

J Open I 0.03 -0. 10 . Outskirts of to^-ns, Suburbs | 0.20 -0.30 | Centers of towns | 0.40 ^ Centers of lar^e ci:ies i 0.60 -O.SO

Table 2. Approxirate ratios of probable maximum speed averaged over period t to.- that avcrajed over one 'i\our (open terrain)

t (seconds) 2 5 j 10 30 60

} 1.S3 1.-:- i 1.42

1 i.:s 1.24[

( jccon as ) , 1 j

1

^ ::oo

u'/u*^ 1 I.IS 1.13 ! '.or ! 1.^3 1.-7

Table 3. - Value of k^.Jc.lCj

z [meter si 0.01 0.07 jo. 20 0.40 O.coj 1.00 1.20

"l 320 223 jiri |133 U4 j 1 50

"2 1 -

- 1- 1- 30 1 26 1 22

- j - 1 - |3.«|2.d5|2.25

In cqs. (14) and (16)

Table 6 - Deflections and accelerations

for various roughness Icrigths

a

txposure

C0,i<:t;li ! O.COS O.Ol] 0.03| j.^'S 1 Juilding 1 a (H)"' " r.ax

'

A (H)

1 —IX 0.020

1.69 1 1.39 j

1 o.o2o! o.oig'c.;:? 1

1

1 ""'"'^jCaii^Kci:

.33 i .33 '

. : 1 o.:i-' ^ci^ :.:'5 j

a 0.20 1 0.3J 1 0. - . Building 1 3 (H)-

a (h)"

1.3? 0.013

1.30 o.oi;

1.2"

0 . : 1

Building 2 a (!-:) '

iriax

^

.27 0.014.26 0.013

. 25 0.012

a

Exnosurc

Center o: lj r ; o : • . 0.60 1 -J. iu 1 i . ^ . Building 1 ^.nax^")" 1.22 0.016

1. 16 O.OIS1.14 O.OIS

'

Building 2 a (,Hj' rax , .23 0.011 0.011

.21 0.010

^»OTE : ncters; -.mils of g (g ^ acceleration

01 fravity)

Tiblc 4 - Values z to various cur

^/H, i^/H correspondi

CURVE : ,. r. 1 : / ri ill A 1 3 X 0. J' 3 4 X 10-^ 0. 04

B 3 i X 0. 3 4 X 0.

C- 3 3 X 0. S 4 X 10 : 0. 15

D 1 1 X 0. r 3 C X lo-f 0.

E 1 3 X 0. L' 3 0 X 0. 10

7 X 10 . 0. M 1 3 X 10

' 0.

G 1 C 0. M

1 3 X 10'; 0. 20 H T e I 0. N 1 2 X 10": 0.

I 2 X 0. M' 3 X

10'-, 0. 45 r T -» X 10-f O.Cb 0 7 X 10 ; 0 J 3 4 X 10- .J. 0' T X 10'" n 30 Note: If ; /H is o less and : /H < 10"

i; ccy be assumed :

t.'ie order of 0. 1.

ir. Jetemming 1

• 0. U

Table S - Description of buildlnjs

selected as case studies

3ui Idin

H

1 : ^ i ^1 1 n

3 1 .ictonZ-neter I

1 ou i3|j. iJlU.Ol

2 ISC 60 45[o.:o!j.qi l.SOO

Table 7 - Ratios [a 1- /[a 1, ' nux-'t

J

' m,i.>;' 0.033

EuilJir. ; I : Exposure 0;'cn L.ir.c Li:;. I. pen L.i

f « 0.01

f, = 0.10

f| = 0.19

1 .00

.93 .97

1.00

.57

.96

1.00

.97 .96

l.CO

.94 .93

Tabii 3 - 7-1 'c~o 5 e ;

r icu J va.lu23 C ,C to response calculared using

C - 10, -C 15

Case C. C.,

] Larg: City

Interval ! Suildin? l| 2uildin<; 2

1

i r

i 1 ^ 1 r. 1 r". V 1 11

1 jlO lo 1 G<n<» 1 1

1. ; 1. 1.

2 6.3

10

1016

0<r,< . 9n. : , .

, .9n,<n<i!^-°'

1 1. 1 1.01

1 1.

3

410 6.4

16

0<;-.<.:n, , _

.9n,<n.-:-'^-

1. j 1.02

' 1 1 • ,

6.3 10 0<n«.^ , 1 . Oo 1.3 1 1 .03 14

S 46.3

6.4

10 0<n<.."n , .9n,<n<i ^

1 1.3 1 l.OS

! 1.4

0 •'. C.4 0<a<^ 1 1 . 1 1 1 1 . 3 I I . i3 ~—:—

1-62
1-63
5 10 IS 20 2S 30 35 40

FIG. 4 —Function Y„, B/H = 0

1-64
O' n' 0 N m' M L.L' h

FIG. 5.—Function Y„, B/H = 0.167

FIG. 6.-Funrtion Y„. B/H = 0.4 f^'G ? -Function Y,„ B/H = 1.0

1-65
A Wood House Will Resist Wind Forces

by

Billy Bohannan

Forest Products Laboratory, Madison, Wisconsin

Experience tells us that wood-frame houses built by conventional construction

practices will be long-lived structures. Why then should we be doing research on lightframe wood construction.

The answer seems simple. There are reasons to believe that the conventional wood

frame systems are overbuilt in many respects. Initial evaluations have indicated that

significant material savings are possible without sacrificing any of the structural

integrity of wood-frame systems.

The actual structural performance of the conventional wood-frame house is not well

understood. Little or no structural engineering goes into its design. While a lot of

research effort has gone into defining pieces and parts of houses, much of this research

has been aimed at very narrow objectives. Little attention was given to the house as a

complete structural system. The interaction of components has not been evaluated.

It is assumed that the resistance to wind-caused racking or shear would be provided

solely by the end-walls of a conventional wood-frame house. There are no known formulae

for calculating the racking resistance of these walls. In fact, there is even some

question as to how much of the total wind force actually reaches the end-walls.

KEYWOEU)S: Structural performance; Mode of failure; Racking resistance; Design criteria;

Concentrated loads; Uniform loads.

On-going Research

One primary objective of on-going research in this area at the Forest Products

Laboratory is to develop engineering criteria for the structural responses of a conventional house to simulated wind forces. Because of the many facets of the unknown character of

a house, a study was broken down as follows:

1-66
1„ Measure the racking resistance of walls under concentrated loads during

progressive stages of construction, as the house advances from a simple wall panel to a

complete 3 -dimensional structure.

2. Correlate the relationship between uniform loads (wind load) and concentrated

forces as they affect racking distortion.

3. Determine the ultimate state limits or weak links in the house that may lead to

structural failure.

To consider these points, one full-scale house was tested under simulated wind

loads. A series of six tests were conducted under horizontal forces to assess the racking

properties of the house. The first five utilized concentrated loads, during progressive

stages of construction, to measure stiffness of walls; the final test employed uniform

loads to determine both stiffness and strength of the complete house.

The conventionally constructed wood frame house was 24 feet wide and 16 feet long.

The size was limited by the dimensions of the structural test facility. The 24-foot

width is fairly typical, but the 16-foot length is no more than one-half the house in

length. This is, however, a reasonable distance from an end wall to a partition. The two

end walls (parallel to the load) contained windows and doors.

A detailed discussion of test procedures and results is given in "Testing of a Full

Scale House Under Simulated Snow and Wind Loads," USDA Forest Service Research Paper

FPL-234, Forest Products Laboratory, Madison, Wisconsin.

Results

The average racking stiffness of the walls during the five progressive stages of

construction are given in Table I.

1-67
Table I. Average Racking Stiffness of Two Walls During Progressive

Stages of Construction.

Stage of

Construction

Description

Load at 0.1-in.

deflection

Relative

stiffness

Lb Pet

No. 1 3/8--inch plywood sheathing 1,480 100.0

2 Cut--in window and door openings 910 61.5

3 Add drywall and siding 1,770 119.6

4 Add sidewalls 2,220 150.0

5 Add trussed roof system 2,760 186.5

The sequence and modes of failure in the house subjected to simulated uniform loads

are given in Table II.

Table II. Sequence and Modes of Failure in House Subjected to

Horizontal Uniform Loads.

Sequence of failures

Pressure Equivalent wind

p velocity—

Lb/ft^ Mi/h

1. Sole plate broke loose on

loaded wall 63 157

2. Center stud on loaded wall broke

in bending (estimated) 70 165

3. House slid off sill plate

(foundation) 123 220

1/ From V = 19.8 ^Tp

1-68
The first structural failure occurred along the sole plate of the loaded wall under

a uniform pressure of 63 pounds per square foot. At this point it was necessary to repair

the sole plate and reinforce the loaded wall to resume testing. Finally, at a pressure

of 123 pounds per square foot, the connections between the floor system and the foundation

failed. At this load, which equates to a wind force of approximately 220 mph, there were

no visible structural failures to the end walls which were resisting racking forces.

With proper anchorages to the foundation, the house would certainly have sustained a

considerably larger load.,

Firm conclusions cannot be derived from a single test, but observations are possible.

The racking resistance of end walls were more than adequate. The wall-to-floor connection

was the first weak link, but this could be easily rectified with minor design modifications. The final weak link in the resistance to wind forces was in the connection between

the house and foundation. About one- fourth of the simulated wind load was transmitted to

the foundation and three-eighths of the load was resisted by each end wall.

To complement the empirical evaluations of the structural performance, theoretical

engineering design criteria are being developed to predict the racking strength and stiffness of walls. The theory is based on an energy approach whereby the racking force and

displacement are equated to the energy absorbed by the fasteners. It considers the

geometry of the sheathing panel, the number of fasteners, and the lateral resistance of

fasteners in covering materials. The theoretical analyses should be published this

calendar year.

1-69
Wind Loads

On

Low-Rise Buildings

Noel J. Raufaste, Jr. National Bureau of Standards, Gaithersburg, Maryland

ABSTRACT

The National Bureau of Standards is continuing its project to develop improved design

criteria for low-rise buildings in developing countries to better withstand the effects of

extreme winds. This paper is an overview of the results of the project, some of which have

occurred. since the Seventh meeting of the U.S. -Japan Panel on Wind and Seismic Effects, that

is to say, the third year of this 3 1/2 year project. To date, most data analyses have been

completed and presented in several NBS published progress reports. The final report is

expected to be published in the Fall of 1976.

KEYWORDS: Codes and Standards; Disaster Mitigation; Housing; Low-Rise Buildings; SocioEconomics; Structural Connections; Wind Loads.

INTRODUCTION

This paper supplements High Wind Study in the Philippines , which was presented at the 7th

Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects. It presents the results

of continuing activities associated with this 3 1/2 year project to develop improved design

criteria for low-rise buildings to better resist extreme winds.

The project originated in 1972 from a recognition by the Agency for International Development

(AID) and the National Bureau of Standards (NBS) that additional research on wind was needed

to reduce human suffering, property losses, disruptions to productive activities, and

expenditures for disaster relief. This is the primary goal. The objectives are to:

1) learn more about the effects of high winds on low-rise buildings; 2) develop improved

siting, design, and construction information, which would improve the resistance of buildings

to extreme winds and which would be culturally acceptable to the user; 3) provide training

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to local professionals and technicians in performing wind measurement and analysis of

full-scale- and wind-tunnel testing; and 4) provide a large-scale transfer of technology to

make use of these improvements in design and construction, as well as new climatological,

sociological and economic findings.

The activities and accomplishments associated with the project have produced the following

benefits: 1) improved design criteria; 2) focused an awareness of the need for improved

ways to design against effects of extreme winds and on the methods required to improve

building designs; 3) essential documentation of new and under-used existing information;

4) excellent working relationships with public and private decision makers in several

developing countries and continued excellent working relationships with building professionals in developed countries; 5) training professionals and technicians in developing countries

for carrying out wind measurement and' analysis procedures; and 6) methods to transfer

information to users in developing countries.

The results of these efforts are significant. These activities provided the first step

toward the preparation of improved wind codes and standards for developing countries.

The Philippine National Building Code is currently undergoing revision that will lead to

improved building practices. The project's test data including improved gust factors

and mean pressure coefficients are being incorporated into Section 2.05, "Wind Pressures"

of the National Structural Code for Buildings. Also, the project's test results will be

made available to the subcommittee on wind loads of the American National Standards Institute,

Inc. (ANSI) for possible incorporation into the American National Standard A58.1—"Minimum

Design Loads for Buildings and Other Structures."

The project is scheduled for completion in September with final report printing during the

Fall of 1976.

BACKGROUND

Wind loading on tall buildings has been extensively studied during the past 10 years.

However, very little parallel work was devoted to buildings less than 10 meters high. Post1-71
disaster investigations of wind damage suggest that coefficients contained in most current

codes and standards do not adequately describe the wind characteristics near the ground and

the pressure distributions on low-rise buildings. Highly localized wind pressures on buildings tend to be underestimated while overall pressures tend to be overestimated. This

observation was confirmed from full-scale tests during this program.

Equally important to the understanding of wind pressures on buildings, is the selection of

design wind speeds in geographic areas having a high frequency of tropical storms. By

taking account of local frequencies and terrain effects, it is anticipated that design

speeds can be selected that will accurately reflect the risk of wind damage to buildings.

It will be possible for building designers to specify more realistic design loads (from

the NBS-improved pressure coefficients)"''.

Field Test Sites

Three field test sites in the Philippines were selected for recording wind load data on

buildings. The sites are: Quezon City, where three houses were instrumented with wind

pressure equipment; Laoag City, about 500 kilometers north of Quezon City, is the site

of two test houses; and Daet, about 230 kilometers south east of Quezon City, is the location

of one test house. The paper High Wind Study in the Philippines contains additional information on field test sites, wind tunnel studies, and other background information.

Wind Tunnel Studies

In addition to wind tunnel studies performed at the University of the Philippines (UP) in

Dr. Richard D. Marshall, the project's principal investigator is responsible for the

development of the test data leading to improved design pressure coefficients. This

paper is based in large part on the National Bureau of Standards Interaeencv Reoort yS-VQn,

which was authored in part by Dr. Marshall and Mr. Raufaste.

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Quezon City, NBS project staff initiated a series of wind tunnel tests at the Virginia

Polytechnic Institute and State University (VPI and SU) . These tests used a flow simulation

technique similar to that which was perfected at the University of Philippines. Since VPI

and SU's wind tunnel test section was larger than that at the UP (2.14 x 2.14 meters square

and 5.5 meters long) the scale ratio was increased from 1:80 to 1:70. The tests, begun in

April, 1975, have been completed. They cover 32 combinations of roof slope, height-to-width

ratio, length-to-width ratio and size of roof overhang. The test results are still being

analyzed. Preliminary analysis suggests that additional studies should be conducted on roof

slopes near 10 degrees. The wind tunnel test and the full-scale test results obtained during

previous studies in the state of Montana, United States, and studies conducted by the Building Research Establishment in England and including research by NBS in the Philippines provided the basic data used in developing tentative design pressure coefficients. These and

other results are discussed in more detail in the National Bureau of Standards Interagency

Report (NBSIR) 75-790, FY 75 Progress Report on Design Criteria and Methodology for Construc -

tion for Low-Rise Buildings to Better Resist Typhoons and Hurricanes .

Additional wind tunnel studies are being performed by the VPI and SU. Measurements are

presently being conducted at 18 locations on a scale model of a quonset shape building.

These will present the characteristics of the turbulent wind flow, mean wind speed turbulence

intensity at the scaled 10 meter level, and integral scale of the turbulence and typical

spectra of along-wind fluctuations. Results are expected during this summer. The building

being tested is a model of a low-cost house designed by CARE, Inc. , Bangladesh. It is built

in 1-meter modules bolted at the top and along ridges at its sides. It is about 4 meters

wide and 3 meters high.

Regional Conferences

Project results were discussed at two regional conferences; one in Manila, Philippines (May

1975) and the other in Kingston, Jamaica (November 1975) . These conferences provided a

channel to transfer results to the building community including government agencies, private

developers, design professionals, regulatory officials, and university staff from Asian and

Caribbean wind-prone countries. The conference nrovided an opportunity for the participating

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countries to cominent on the design criteria and to offer suggestions for their implementation. Members from developing countries were provided an opportunity to better understand

wind and building design problems common to their countries and discuss methods for solutions.

Technology Transfer

The NBS/AID high wind project results should have an impact on several key sectors of the

local developing country's population—the professional community, the government policymaking sectors, and the general public—building owners and tenants. The technology developed has and is continuing to be transferred to the participating country's building community.

To stimulate the general public to improve their buildings to better resist extreme winds,

consumer-oriented material has been developed containing concise and graphic descriptions

of technological issues, research findings , and conclusions. The use of brochures and

pamphlets containing consumer-oriented material is an excellent method to transfer such

information. This material consists of recommendations on siting, design, and construction,

checklists of recommended construction practices, summarized technical reports prepared

by NBS staff and consultants, and a selective list of resources.

A 16 mm movie, another consumer-oriented package, completed in 1975 was modified in 1976

to include information on project results. The movie summarizes the high wind research

project. Starting with the destructive effects of winds on buildings, it progresses to

the NBS field-testing activities and the wind tunnel-testing program and the project outputs.

The movie also uses computer graphics to illustrate the effect of wind on buildings. The

movie and the consumer material are two of the various ways to continue information exchanges

among the users of this project and between the U.S. and other country participants. Through

March 1976 over 10,000 persons—both technical and lay audiences and school children—viewed the movie.

PROJECT RESULTS

The following presents a brief description of the results of the project. The reader should

reference the NBS report NBSIR 75-790 for additional information about the following project

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results. NBSIR 75-790 can be obtained from the National Technical Information Service,

Springfield, VA 22161 as PB 250-848.

A comprehensive review of the probablistic techniques for the analysis of extreme wind speeds

was developed by Dr. Emil Simiu of NBS and published as Estimation of Extreme Wind Speeds in

the Philippines . Corrections of annual extreme speeds for the type of instrumentation used,

averaging time, height above ground, and type of exposure are presented so that homogeneous

sets of wind records can be established. Probabilistic models are described and applied

to wind records for several Philippine stations using a computer program developed at the

NBS. The results are compared with design speeds presented in the National Structural Code

which is part of the National Building Code of the Philippines,

Analysis of wind speed data show that design speeds currently used can be reduced in some

areas of the Philippines and should be increased in other areas, including coastal plains.

Similar analysis may be performed for other countries as well. The computer program is

available for use by other wind-prone countries.

Simplified procedures for the calculation of wind pressures acting on building surfaces

were developed. These improved design criteria provide building professionals with more

accurate design loads. This information is contained in the report, A Guide to the Determination of Wind Forces , by Dr, Richard D. Marshall of NBS. The report describes the basics

of wind flow around buildings and the pressures created by these flows on building surfaces.

Effects of such features as roof slope, roof overhang, and building openings are discussed.

It refers to material covered in Dr. Simiu' s report and assumes that the basic wind speed

is known or can be calculated by the designer. A procedure for determining the design

speed is presented. It takes into account the general terrain roughness, local topographical

features, height of building, expected life of building, and risk of failure.

Once the design speed has been obtained, the mean dynamic pressure is calculated and

pressures acting on walls, roofs, and internal portions of buildings are determined by use

of appropriate pressure coefficients tabulated in the report. Finally, corrections are

applied to the calculated pressure to account for terrain roughness and height of building.

The steps required to calculate pressures and total drag and uplift forces are summarized

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and an example is presented. It is intended that these data form the basis for wind load

design standards in developing countries.

Another output of the project was a set of recommendations for good construction practices

and details, especially for countries experiencing extreme winds. This material is found

in A Guide for Improved Masonry and Timber Connectors In Buildings by Dr. S. George Fattal

of NBS and Messrs. Gerald Sherwood and Thomas Wilkenson of the Forest Products Laboratory.

A survey and evaluation of some current masonry practices as used in the tropics are presented. Local practices, investigation of building products, and standards compliance are

discussed and evaluated with the aid of photographs of buildings during various stages of

construction. Current masonry practices in the United States are presented. Also, various

types of connectors used in U.S. construction are discussed. This serves as a reference

for improving building practices. The chapter addressing timber fasteners is similar in

format to that for masonry connectors. It presents a general introduction to timber fasteners and a review of Philippine timber-fastener construction practices. Recommendations

follow. The report also addresses suggested improvements in masonry and timber construction.

The suggestions are based on a review of many masonry and timber documents, related reports,

building codes, specifications, working drawings and photographs of Philippine houses and

on-site post-disaster surveys of typhoon-damaged buildings.

The study also produced a method to allow a country's planners, economists, public officials,

and other decision-makers to assess housing needs up to 20 years into the future. The report.

The Economics of Building Needs - A Methodological Guide , by Dr. Joseph Kowalski of NBS,

presents this information. The data serves as a tool to assist the decision-maker in assessing the impact of the development of buildings to better resist winds on local housing needs.

The report attempts to identify those parameters that define a low-cost house. This is

determined by a simple mathematical procedure based on readily available data, i.e., family

income, housing expenditure, and the established regional proverty level.

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The study examined cultural and socio-economic factors that affect building practices. It

also shows how strong, inexpensive locally available building material can be integrated

with good building design. The report, Housing in Extreme Winds , by Stephen Kliment, architectural consultant, presents this information. The report is consumer-oriented in nature

and is intended for residents, owners, planners, and builders of low-cost/low-rise housing

in developing countries. The report addresses the Philippines, Jamaica and Bangladesh. A

general review considers socio-economic conditions prevailing in developing countries.

Suggested solutions include the placement of building to exploit existing land contours and

vegetation, and the design of buildings so as to conform to research recommendations

concerning preferred configurations of roofs, walls, overhangs and openings. In addition,

construction techniques are briefly covered. Materials that are cheap, strong and locally

available are recommended, and several innovative methods of construction are mentioned.

Plastics have been produced using native and imported raw materials and could serve as an

excellent low-cost house once their configurations become more familiar to the population.

At first, they may be used for clinics, schools and warehouses. Rural and urban components

I of population trends for each of the three countries are reviewed. Finally, housing characteristics peculiar to each country are summarized through text and in thumbnail sketches,

in both architectural terms and in terms of occupancy characteristics. The socio-economic

functions of family ties and housing needs are reviewed. The community structure, the

economics of housing, the housing characteristics, and the housing programs are summarized

by country.

SUMMARY

This project produced test data and information for improving current low-rise building

practices. They are:

o , J -, a methodology to estimate extreme wind speeds for

wind-prone countries

;

o improved design criteria to determine wind forces

I on low-rise buildings;

! o recommended good building construction practices;

methodology to forecast housing needs; and

! 1-77
cultural, architectural, and socio-economic factors

that affect building practices

Thus, the study shows it is possible for wind-prone developing countries to design their

low-rise buildings to better resist wind and build them at an acceptable cost.

THE FUTURE

It is anticipated that the utilization phase of this project will begin in September 1976.

The purposes are: 1) continue transferring technology to participating countries;

2) provide spare parts on an as-needed basis to maintain field and laboratory equipment;

3) explore channels for expanding this technology to other developing countries subjected

to extreme winds and to other professionals involved in aspects of low-rise building

technology; 4) continue field and laboratory data analysis based on new information;

5) investigate further the mechanisms of information transfer at the level of the general

public; and 6) assist as necessary in the incorporation of test results into codes and

standards—both domestically and internationally.

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CHARACTERISTICS OF THE HIGH WIND AT HACHIJOJIMA ISLAND

ON THE OCCASION OF TYPHOON NO. 7513

SEIJI SOMA

Chief, Meteorological Research Institute

Meteorolocical Agency

Japan

ABSTRACT

A typhoon that attacked Hachijojima Island on October 5, 1975 brought to this island

very strong winds which were beyond expectations . The value of the peak gust observed at

the weather station, located at the central part of the Island, was 67.8 m/s . Moreover,

in the premises of the lighthouse situated on the southeastern part of the Island, an incredible wind of 82.4 m/s was recorded. This wind was extraordinarily strong and was

the third strongest recorded wind in our country. Due to this high wind, a great deal of

damage was caused in various parts of the Island. Houses and buildings suffered the most

from the high wind. The damage rate to houses which varied in different areas, had a high of 7.3%, which were completely destroyed, and 21% in which half of the houses were

completely destroyed. Wether this strong wind was innate in the typhoon, or whether it was brought about due to the peculiar topographic features of the Island was the subject

of this study. However, it was not possible to clarify these points quantitatively.

Although, in certain areas, it was realized that topographic influences were apparent.

While making this survey on the typhoon's effect, it was noted that severe damage

occurred to the windowpanes in tall buildings. An administrator of the building told us that these panes were broken by fragments of houses that were scattered in the high wind.

This, therefore, indicates that a problem exists when tall buildings are surrounded by

houses which have low wind-resistance in their vicinity. Therefore, the damage indicates

that in town planning, not only a wind resistant design of individual buildings is re- quired but also wind resistant planning of the entire environment is essential.

Key Words: Damage; Effects; Houses; Typhoon; Wind speeds.

1. Introduction

On October 5, 1975, typhoon No. 7513 passed over the northern part of Hachijojima

Island and brought to this Island a wind which had a maximum speed of 67.8 m/s. This wind

was the strongest ever recorded, since the weather station was established. This high

wind caused a great deal of damage to houses and other strui.. !"ures on the Island. This

typhoon was a typical wind typhoon according to newsmer , whi'jr, have often attacked the

Mainland of Japan in the late autumn. The amount of precipitation that was produced by

the typhoon was merely 26 mm, but the wind was extremely strong. The wind speed of 67.8

m/s was only the value recorded at the weather station, and a far stronger wind of 82.4 m/s

II-l
was recorded at another part of the Island. Even in Japan, which is called one of the

typhoon countries , a wind record of over 80 m/s has been experienced only twice in the

past. The first was 84.5 m/s recorded at the Murotomisaki Weather Station in Shikoku

Island on September 16, 1961 and the other was 85.3 m/s observed at the Miyakojima Weather

Station on the Nansei Islands on September 5, 1966. The wind record of 82.4 m/s due to Typhoon No. 7513 was observed not by a station under the Japan Meteorological Agency, but

at the lighthouse under the control of the Maritime Safety Agency. However, it was recorded by the same type anemometer as is used by the stations under the Japan Meteorological Agency. Therefore it is quite trustworthy.

The question concerning this strong wind, was it a natural phenomenon, or did the

formation cause such an extreme wind. This is an interesting problem, which at present,

still remains unsolved. I wish to describe below the characteristics of the high wind

at the time of the Typhoon No. 7513. This will be achieved by referring to data of strong

winds which blew on the Island in the past, and also referring to those two typhoons that were accoiipanied by extremely high wind, as mentioned before.

2. Topographic Features of Hachijojima Island and Wind Records in the Past

2 Hachijojima is a small island covering 68.3 km and is situated in the Pacific Ocean

250 km south of Tokyo, and administratively belongs to the Tokyo Metropolis. Its population is 10,320. Because of its warm climate and scenic beauty, the Island is visited by

a great number of tourists from the mainland. In order to accommodate these people, many

high-storied hotel buildings were constructed during the past several years.

On this Island, there are two comparatively high and steep mountains in proportion to

the small area, i.e., Mt. Hachijo Fuji (854 m) and Mt. Miyake (700 m) situated in the

northwestern and southeastern parts of the Island, respectively. The area lying between

these two mountains consists of flat land. This land is positioned in the southwestern

and northeastern directions, which lead to the sea coasts. The main city area of Hachijojima is on this flat part of the land. The weather station, that recorded the wind

speed of 67.8 m/s is located in the southwestern part of this flat area, which is about

1,000 m from the west coast. Due to such peculiar topographic features of the Island,

the wind direction on the level area is SW-NW, and NE-ESE. Because Hachijojima lies in the middle of the Ocean, it is prone to typhoons, and

so far strong winds have been recorded on a number of occasions. Table 1 lists those

recorded winds which have exceeded 50 m/s.

3. Approach of the Typhoon and the Change in Weather Elements

Photo 1 taken by radar on Mt. Fuji, 250 km north of the Island, shows the way

Hachijojima was enveloped in the eye of the typhoon. Also, according to the barographic

record obtained at the Hachijojima Weather Station, the funnel-shaped pressure change

which is characteristic of a typhoon was noted, suggesting that the center of the

typhoon was very close (Fig. 1)

II-2
From this chart, it shows that the typhoon was closest to the Island at 16.30 on the

5th, and the lowest recorded pressure was 947 mb. Wind speeds observed prior to, at and

after this time are shown on Table 2

.

The column on the left hand side of Table 2 shows the wind speed recorded at the

Hachijojima Weather Station and the column on the right shows those values which were

recorded by the lighthouse in the Sueyoshi District in the Southern part of the Island.

The peak gust value in the brackets were obtained by the average gust factors , between

13.00 and 16.00 hrs. on that day, multiplied by the average wind-speed. As can be seen

from this table, peak gust (or the maximum instantaneous wind speed) reached its maximum

at 15.30 hrs. at both observation points. The difference between these two values, shown

as the peak gust at 15.30 at the lighthouse, is due to different people who read the selfregistering paper.

If one looks at the recording paper carefully, a gust of over 50 m/s wind speed was recorded five times between 16.20 and 17.00 hrs. However, the wind record obtained near

the lighthouse, utilized a pen which went beyond the scale. Therefore, judgement was required in estimating the speed. It appears that a peak gust of 75 m/s occurred several

times during 20 minutes initiated at 15.20.

4. Outline of Damage

The largest damage caused by the typhoon was to houses. The detailed number of damaged

houses will be mentioned in the next section. Other noticeable damage included collapse of

electric poles, the Loran nautical mark towers, and overturning of motor vehicles. Trees

that were completely up rooted throughout the Island. Damage unique to the area was

caused by the salty wind, which killed the leaves of the trees all over Mt. Hachijo Fuji,

up to a height of 854 m and changed the color of the mountain to red.

According to the distribution map of the damage prepared by the weather station (Fig.

2) , the most afflicted area was the flat land lying in between the two mountains. Needless

to say, this is due to the fact that the principal cities and towns are located in this

part of the Island. Also the southern part and the southeastern part of the Island suffered

from damage. The arrows shown on the map indicate the direction in which houses and

structures collapsed (Photo 2)

.

a) Damage Rate of Buildings and Air Current Characteristics

In order to show how the buildings suffered from the gust, it is more appropriate

to show the damage rate rather than by the number damaged. The damage caused by the high

wind, in various areas, was investigated resulting in the data given in Table 3. Shown is

the damage ratio, which is a value calculated by dividing the number of damaged houses by

the number of households.

As mentioned in Section 3, the wind direction at the peak gust time was SSW. When

the wind blows in this direction, the Mitsune District should have been right behind Mt. Mihara and accordingly less windy. On the contrary, however, according to Table 3, the

damage rate is higher than one would expect (51.2%)

.

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As it happened, approximately one year before this typhoon, we had made an observation

of the wind distribution on Hachijojima Island, by spreading the observation network

throughout the flat area of the Island and on top of Mt. Mihara. The number of observation

points at that time was nine, including the weather station.

First of all, we tried to obtain from this data what wind direction and speeds were

observed at the eight other check points, where the wind at the weather station showed an SSW direction. The analytical material used at this time was restricted to the wind speed

over 10 m/s. The result of the survey is shown by the diagram on Fig. 3, in which the

wind direction and speed are indicated by vectors . This diagram shows that when the wind

direction is SSW at the weather station (B point) , it becomes 100% SW at the Mitsune District (G point) . In other words, the air current that comes in from the SSW direction proceeds to the right along Mt. Mihara, keeping the strong wind speed to its back, and goes

around the mountain. The diagram also shows that the wind direction on top of Mt. Mihara

(A point) stands high from the S, as against the SSW at the weather station. The topographical experiments carried out by use of a wind tunnel also proved a similar air current

characteristic when the SSW wind was given. Photo 3 shows an example of the experiment by

the wind tunnel.

b) Wind Speed Distribution at the Time of Typhoon

At only three locations during 16.30 hrs. on the 5th, the wind speed was actually

recorded; at the weather station, the upper air observation point belonging to the same station (point F) , and at the lighthouse. The extent of the damage by the high wind, however, covered a very wide area and it is necessary to find out, from the viewpoint of the

investigation on wind damage, what wind speed had actually blown at those respective points.

Therefore, relying on past observation data, a wind speed ratio at each point as against

the value observed at the Hachijojima Weather Station (in case of SSW wind) was sought, and

by using this percentage ratio, values of wind speed at various points at 16.30 were estimated. These are shown on Table 4. According to this table, it is estimated that on top

of Mt. Mihara (point A) the wind was 1.89 times as strong as the weather station. Therefore, the wind speed at this point at 16.30 hrs. on the 5th, is estimated to be 60.0 m/s.

The next task was to work out the peak gust at various points by reading gust factors

from the past observation records on the wind recording papers and by the estimated wind

speed, as shown on Table 4. These results are given in Table 5. The wind speeds and the

peak gusts estimated at 16.30 hrs. are shown in Fig. 4. According to this table or the

figure, the peak gust on top of Mt. Mihara at 16.30 hrs. is estimated as 77.4 m/s. The

Loran nautical mark tower could not have stood against such a terrific gust. Also by

this table, at point G (Mitsune District) the peak gust is estimated to have been 54.4 m/s.

Considering that this district is located leeward of the mountain when an SSW wind blows,

this figure seems to be extremely large. This is due to the fact that because the mountains

are situated in the northwest and southeast, the level area served as a corridor between

these mountains and the SSW wind blew right through to the east coast.

II-4
5. Extremely High Wind in Sueyoshi District

"As mentioned before, an extremely strong peak gust of 82.4 - 85.2 m/s was recorded in

the premises of the lighthouse in Sueyoshi District on the southeastern part of Hachijojima.

The high rate of house destruction up to 79.1% as shown in Table 3 was due to this extreme

peak gust. This gust also destroyed the lantern panes of the lighthouse, which is an unprecedented occurrence. According to the report given by the Hachijojima Lighthouse, six sections of the

lantern panes out of twelve, which sheltered the luminant components and lenses of the

lighthouse, were broken, i.e. three entirely and the other three cracked. Most of the

broken panes had been placed at the windward side of the lighthouse (Fig. 5)

.

It may have been possible that fragments of the broken glass had caused the damage to other glass on the leeward side, but the details of the situation at the moment of this

destruction have not been clarified. It may be added that the lantern panes employed for

this lighthouse was 7 mm thick, 1,200 mm long and 800 mm wide, reinforced against wind and

curved outward as a convex lense. There have been few cases when lantern panes of lighthouses were broken by strong winds. However, in most of these cases, pebbles or other

fragments blown and scattered by the wind were considered to be the immediate cause of the

breakage. It is difficult to reason this situation in the Hachijoj ima ' s case when the

location of the lighthouse is taken into consideration. The lighthouse is located at the

windward edge of the breakwater stretched out on a cape into the sea. Consequently, it

may be correct to conclude that the extremely high wind, which was the third highest wind

speed in observatory history, was the exclusive factor of the destruction.

6. Records of Strong Gusts of More Than 80 m/s

Considerable lead time is needed to decide whether the record of 82.4 - 85.2 m/s

in the Sueyoshi District can be attributed only to the Typhoon, or to some topographical

feature which contributed to its strength. In this section, analysis will be given for

reference on the situations where peak gusts of more than 80 m/s occurred,

a) The Miyakojima Typhoon II The strongest gust ever recorded on flat ground and at a regular weather station

was a maximum of 85.3 m/s. This was recorded at the Miyakojima Weather Station located

on the Nansei Islands at 6:30 on September 5, 1969 during the Miyakojima Typhoon II. The

direction of the wind was NE and the mean wind speed was 60.8 m/s. The distance from

Miyakojima to the Typhoon was 20 km at the closest time. This was very similar to the

relative distance observed in the case of Hachijojima and Typhoon No. 7513, while the

Miyakojima Typhoon showed the lowest atmospheric pressure of 930 mb, which was far below

that seen in Typhoon Na 7513. Miyakojima is a flat island, 108 m above sea- level at its

highest point, and 2.5 times larger than the area of Hachijojima. The Miyakojima Weather

Station is located on this flat island and is not influenced by the configuration of the

surrounding land. Consequently, the value recorded on Miyakojima can be considered to have been brought about by the typhoon itself and not by any other influences.

II-5
Mr. Mitsuta says that the Typhoon passed slightly outward with respect to Miyakojima

and consequently the island was in the proximity of the typhoon's maximum wind speed zone which produced the extremely strong winds on Miyakojima. This assumption will give some insight in the analysis of the Sueyoshi District of Hachijojima.

b) The Muroto Typhoon II At 11:30, September 16, 1961, a maximxim wind speed of 84.5 m/s (wind direction:

WSW, average speed: 66.7 m/s) was recorded at Murotomisaki Weather Station. At this time

the typhoon passed rather close to the station, i.e. 7 km to the west. The lowest atmospheric pressure observed at this time was 931 mb. The extreme wind speed can be naturally

attributed to the energy of this large scale powerful typhoon itself, however, some consid

eration should be given to the geographical characteristics of the location. Murotomisaki

is shaped like a half-cylinder with its top, where the station is, 184 m above the sea.

It may be natural to consider that the air current was accelerated when the wind blew in a crosswise direction over the cape. Thus, the observed value of 84.5 m/s may well include

the influence of the topographic factor described above. In the case of Sueyoshi District

the site where the anemometer was set was similar to that of Murotomisaki, while the

height above sea-level was 95 m, about half the height of Murotomisaki Weather Station.

c) The Factors of the Extremely High Wind Observed in Sueyoshi District

It may be misleading to determine the factors of the extremely high wind observed

in the Sueyoshi District based on these two cases, but it may be worthwhile to do so at any rate, as a start of this study. Some of the important factors may be:

1. Sueyoshi District was located in the proximity of the maximiim wind speed zone of

Typhoon No. 7513.

2. The wind blew in a crosswise direction over the cape of the island.

3. The district is located southeast of Mt. Mihara and a corner effect of airflow

may have occurred.

7. The Characteristics of Wind Damage in Hachijojima

There has not been any strong typhoons which have hit big cities for a period of

sixteen years, since the Isewan Typhoon. During this period many new forms of houses

and structures have been rapidly constructed and the life-style has changed significantly.

Wind resistant design should have been taken into account in cases of these new houses and

structures. However, we do not have experience on new types of construction, and only

when they are exposed in severe condition of nature, do they show their weakness,

a) Destruction of Glass in Large Buildings

A typical example of glass damage in large buildings was seen in the case of

this typhoon. This occurred in the Hachijojima Kokusai Kanko Hotel, a six-story building.

The curtainwall of the large lobby facing the sea on the first floor of this hotel is 90%

constructed of fixed windowpanes. The size of each windowpane is 3,000 mm long, 2,000 mm

wide and 10 mm thick. The typhoon seemed to have generated a wind speed of 68 m/s, and

destroyed first 13 windowpanes in the windward side of the lobby. The destruction of

these windowpanes caused a strong influx of the storm's wind into the lobby. Furniture

II-6
and in the shops of the lobby were blown away and they destroyed the windowpanes on the leeward side of the lobby. The hotel personnel testified that the destruction of glass on the windward side of the lobby was not due to wind pressure, but due to objects that were blown away from the destroyed houses in the windward direction. If this

is true, the damage in Hachijojima seems to imply that not only is wind resistant design

for glass required for buildings, but also a more general wind resistant design, including

the overall environment in the city is required.

b) Collapse of Electric Poles

This typhoon caused collapse of many electric poles; this is very common damage

caused by typhoons. The number of totally torn down poles was counted at 47. 26 poles

out of 47 were concrete poles and 21 were wooden, which shows that the destruction rate

of concrete poles was slightly higher than that of wood. The number of poles which were

half torn down or left in a leaning position reached 420. The collapse of electric poles

not only crushes things under the poles, but also interrupts traffic. Moreover, the most

serious effect of the destruction of electric poles is power blackouts. Power blackouts

causes serious effects, not only to households, but also to the industrial facilities and

medical institutions when it is prolonged over a long period of time in the big cities.

c) Overturning of Motor Vehicles

The exact number of motor vehicles that were overturned is not known when this

typhoon struck Hachijojin>s, but the several cases were reported. It is anticipated that

when motor vehicles are overturned, and if it happens in the big cities and in their

suburbs where traffic is heavy, it will cause serious traffic problems. This is a problem

which should be examined in the future.

8. Conclusion

The characteristics of the high wind brought about by Typhoon No. 7513 to Hachijojima,

taking into account the wind and the configuration of the land, have been described.

And the possible damage by typhoons hitting the big cities were examined through an analysis of wind damage caused by this typhoon. Among other things, the destruction of

large windowpanes in large buildings, the collapse of electric poles, and the overturning

of motor vehicles were discussed.

Finally, the fact that this typhoon did not cause any death tolls in spite of serious

damage to houses and structures should be mentioned. The Miyakojima Typhoon II which re- corded a maximum wind speed of 85.3 m/s caused no death tolls either. This is very fortunate, but this cannot be expected in all typhoons. In 1934, the Muroto Typhoon I with

a peak gust of 60.0 m/s in Osaka destroyed a primary school building and caused the death

of 400 pupils of that school. The fact that the typhoon hit Hachijojima on Sunday and

that the peak gust was recorded in the day time, was part of the reason for the few nxamber

of casualties. It is also of interest to know what kind of refuge measures were taken

by people to minimize the damage of the high wind which caused the total destruction of

275 buildings.

II-7
References

Ishizaki, H. et al (1968) : The Structural Damage Caused by the Miyakojiina Typhoon II Disaster Prevention Research Institute Annuals, No. llA.

Mitsuta, Y. et al (1968) : Severe Wind Storm Caused by the Miyakojima Typhoon II,

Disaster Prevention Research Institute Annuals No. IIA. Meteorological Agency (1967) : Report on the Muroto Typhoon II (No. 6118, Nancy) , Technical Report of the Japan Meteorological Agency No. 54, Mar. 1967.

II-8
TYPHOON No. 13 OCTOBER 5. 1975

Photo. 1. Typhoon No. 7513 taken

by the Radar at top of

Mt. Fuji; Hachijojima

is located on the southern part of the typhoon

eye vail.

Photo. 2 Examples of houses damage on Hachijojima.

II-9
Photo. 3 An example of distribution of wind by wind tunnel

experiment. The fixed wind is SSW, and the measuring equipment is small flags made of styloform.

11-10
Fig. 2 Distribution of wind damage on Hachijojima.

Fig. 3 Distribution of wind shown by vector formu- : ic:. L la various sites, when the Weather Station I;oti:u:;id :ns_>.E r.-eJ. ( :cr indicated SSW (Based on special observation ^ -.eal-; -.u* {r._r, : i'- - loct- data betVBen 1973 and 191^). cicca ac lo.ZC, o:i . 5

11-11
TuDle 1.

:-i^^K COST C3SER'-.';:0 AT H-iC-TIJCJi: A i'IZA:>{£R STATIC-''!

PERICD

1238, 9. IS \ S5>7 i 50.0

I 1S38, 10,2i ; W ' 57.0 i

I : \ \ I

I

I lS-40, 9, IS ^^^!E ' 54.3 ^

I

] : ! 1337-1574

I 1543, 10,10 I ^'B'^ 51.2

I

1354, 11,23 ! NN^7 ! 51.0

!

i

'

'

; i:'75, 10, 3

I SSV: £7.3 1 i973

11-12
"A'C-HIJ OJ IMA WEATHER STATION HACHIJOJIMA LIGHTHOUSE

TIME W. D. U fm/s ^ P. G. (m/s ) G.E. W. D. U (m/s

)

F. G.

' m. s ' G. F.

1 3. 0 0 S 1 0. 0 2 1.3 2. 1 3 S 1 7. 4 2 5. 5 1.4 7 14.00 ssw 1 2. 2 2 5. 5 2. 0 9 S 1 9. 4 2 7. 5 1.4 2 14.30 s 1 3.5 2 6. 9 1. 99 s 2 2. 8 3 6. 0 1.58

1 5. 0 0 s 1 5. 9 3 2. 5 2. 0 4 s 2 5. 6 3 2. 5 1.27

1 5. 3 0 s 1 5. 3 3 5. 4 2. 3 1 s 3 4, 3 4 8. 5 1. 4 1 16.00 s 2 2. 1 5 0. 5 2. 2 9 8 4 2. 7 5 8. 8 1.38

1 6. 1 0 s 2 3.0 5 4. 6 2.3 7 SSW 4 7. 4 ( 6 7. 3 1 —

1 6. 2 0 s 2 9.9 6 0. 0 2. 0 1 ssw 5 3.2 (' 7 5. 5 > —

1 6. 3 0 ssw 3 1.7 6 7. 8 2. 1 4 SSW 5 8.0-6 0.0 (82.4-85.2) -

16.40 wsw 2 9. 3 5 9. 5 2. 0 3 SW 7 7. 0 -

1 6. 5 0 w 2 8.0 5 5. 5 1.98 SW 5 8. 5 —

1 7. 0 0 w 3 5. 5 6 6. 2 1.86 wsw - 4 2, 0 -

1 7. 1 0 w 2 5. 9 4 6. 0 1.78 WSW 4 3. 6

1 7. 2 0 w 20. 9 3 6. 6 1.75 wsw 2 1.1)

1 7. 3 0 w 1 9.8 3 7. 0 1.8 7 WSW

1 8. 0 0 w 1 4. 4 2 7. 4 1.90 wsw 1 5. 5 MEAN

2. 0 7 MEAN

1.4 2

( ) ESTIMATED VALUES O^'' WIND SPEED FROM MEAN WIND SPEED

AND GUST FACTOR

;CCT. 5, 1575}

1

i -"'0. ol "id

To-cal

i

1

; 1 or

,)

(-t

)

15S7 i lis 7 312 51.2 1247 37 7 16

.

. ;-2.4

1 ~

270 20 : 7 3.3

'27 ' ' ~ ! 3

r, c. 213

23 1 11 .1 4 201

v.

: 37o5 _ 3

1 w 1

"»

1 13, 7 ::73

11-13
Table 4. WIND RATIO TO THE VALUE OBSERVED

AT HACHIJOJIMA WEATHER STATION (B)

SITE Mt.M B

w.s. c

A.P. D E F G H I

WIND RATIO TO (B) 1.89 1.10 0.97 0.78 0.88

27.9

1.07

33.9

1.00

ESTIMATED 31.7 VUND SPEED (iVs) 60.0 |31.7| 34.9 30.7 24.7 ]29.5

] OBSERVED VAUUE AT 16:30, OCT. 5, 1975

Table 5. GUST FACTOR AIP PEAK GUST(ESTIMAm))

SITE A 6

W.S. c

A.P. D E F G

1.95

H

1.67

I

1.69

53.6

GUST FACTOR 1.29 2.14 1.97 1.99 1.98 1.62

ESTIMATED PEAK GUST(ll\/s) 77.4 |67.8| 68.8 61.1 48.9 47.7 54.4 56.6

I

I OBSEaWED VAUJE AT 16:30, OCT. 5, 1975

11-14
I High Winds in the United States, 1975

by

Arnold R. Hull*

Environmental Data Service

National Oceanic & Atmospheric Admin., Washington, D.C. 20235

and

Thomas D. Potter

National Climatic Center, EDS

National Oceanic & Atmospheric Admin., Asheville, N.C. 28801

and

Nathaniel B. Guttman

National Climatic Center, EDS

National Oceanic & Atmospheric Admin. , Asheville, N.C. 28801

During 1975, high winds in the United States were associated with tornado

and thunderstorm activity, major extra-tropical storms and Hurricane Eloise.

Hurricane Eloise reached the coast of Florida with an observed minimum

pressure of 95.5 kPa and sustained surface winds estimated at 202 km/h (56

m/s) . Associated gale winds were reported from Cedar Key, Florida, to southeastern Louisiana and northward over most of Alabama, western Georgia, and

extreme southeastern Tennessee. A unique set of hourly meteorological data

and some oceanographic data was collected by two buoys during the approach and

passage of the eye of Hurricane Eloise. These data, along with simultaneous

ship, aircraft, and satellite data are being assembled and packaged for

hurricane research studies.

During 1975, 918 tornadoes and numerous storms with high winds caused

damage estimated in excess of 2 billion dollars.

A devastating January blizzard dumped up to 58 cm (.58 m) of snow in some areas of the North Central States and was accompanied by winds of up to 130

km/h (36 m/s). In contrast, a severe sandstorm struck southern California on June 17-18, bringing near zero visibility and windspeeds of 144 km/h (40 m/s).

Damage to power lines, poles, and other facilities exceeded $100,000.

Key Words: Extra-tropical storms; Hurricanes; Thunderstorms; Tornadoes;

Wind; Wind damage

Deceased.

11-15
High Winds in the United States, 1975

1. INTRODUCTION

During 1975, high winds in the United States were associated with Hurricane Eloise,

tornado and thunderstorm activity, and major extra-tropical storms. High winds occurred

somewhere in the United States during each month of the year, and almost every state received

extensive damage at least once. Property losses exceeded two billion dollars from the more

than 1,500 separate storms that were reported.

2. 1975 TORNADO SEASON

Last year 918 tornadoes were reported^ throughout the United States on 204 days. They

occurred in all states except Alaska, Idaho, New Hampshire, Rhode Island, Utah, Vermont, and

Washington. The season began on January 7 and ended on New Year's Eve. During the year, 60

people were killed, 1,504 were injured and property damage exceeded 500 million dollars.

Over 1,120 mobile homes were destroyed by tornadoes.

A total of 53 tornadoes was reported in the United States as a whole during the month

of January, eclipsing the old record of 39 in 1967. New state records were set as follows:

Month

January

February

March

April

State

Alabama

Florida

Oklahoma

New York

Mississippi

North Carolina

Virginia

Florida

Louisiana

New Record

Number

13

10

61

13

84

13

17

Previous

Record

5 (1974)

8 (1970)

4 (1961)

0

7 (1950)

7 (1956)

2 (1969)

12 (1973)

7 (1967)

National Oceanic and Atmospheric Administration, National Climatic Center,

Storm Data , Vol. 17 (1975).

11-16
New Record Previous

Month State Number Record

May Louisiana 12 10 (1950)

June Montana a

9 6 (1971)

Virginia 4 2 (1969)

West Virginia 2 1 (1973)

July Maryland 6 2 (1971)

August Delaware 2 1 (1967)

Florida 13 8 (1974)

Virginia 4 2 (1962)

September Florida 9 7 (1967)

November Iowa 9 3 (1960)

December Iowa 1 0 North Carolina 3 2 (1967)

On 10 January twenty-six tornadoes reported in Alabama, Louisiana, and Mississippi

killed 10 people, injured 286 and caused over 30 million dollars worth of damage. The

severe outbreak resulted from a squall line preceding a cold frontal passage.

In March, a tornado struck Warren, Arkansas. Seven people lost their lives. A major

industrial area sustained extensive damage. A total of 151 buildings were destroyed, 100

others received major damage and 200 others sustained minor damage. Later in the month a

tornado moved through Atlanta, Georgia, at 79 km/h (22 m/s) . It killed three people, injured

152, and did 56 million dollars worth of damage. The storm struck a major industrial area,

two large apartment complexes, many businesses, hundreds of fine homes and the Governor's

Mansion. There were 550 families left homeless.

The worst tornado disaster in Omaha, Nebraska, since Easter Sunday 1913 occurred on

6 May 1975. Three tornadoes killed three, injured 133, and caused damage in excess of 400

million dollars. The storms moved through residential and business areas causing extensive

damage to apartments, homes, businesses, schools, autos, trucks, and trees. In places the

damaged area was a quarter-mile wide.

11-17
Row of houses twisted off foundations

(courtesy R. Paskach, Omaha World-Herald).

11-18
Figure 2. Houses twisted off foundations

(courtesy R. Paskach, Omaha World-Herald)

.

Figures 1-4 depict some of the damage in Omaha. The first two photographs show frame

houses twisted off cement foundations. Winds in this area were estimated at approximately

2 324 km/h (90 m/s) . The pattern of debris in Figure 3 is also indicative of strong vortex

motion. The fourth photograph was taken in an area where residences were completely flattened. Figure 5 vividly portrays the human suffering caused by the tornadoes.

T. Fujita, personal communication (1976).

11-19
Debris pattern indicative of strong vortex

(courtesy R. Paskach, Omaha World-Herald).

11-20
I

Figure 5. Human suffering from tornadoes

(courtesy S. Melingagio, Omaha World-Herald).

11-21
On 23 July, Canton, Illinois, was hit by a tornado that killed two and injured 69.

This was the first storm to kill anyone in the State of Illinois during the month of July

since records have been kept. About 20 million dollars worth of property was damaged in

Canton. In a five block area in the central portion of the city, 127 businesses were

damaged; some buildings were totally destroyed while others were so badly damaged that they

had to be demolished. About 100 homes and 50 mobile homes were destroyed; 300 other homes

and 100 other mobile homes were damaged.

3. EXTRA-TROPICAL STORMS

A devastating mid-January blizzard wreaked havoc over Illinois, Indiana, Iowa, Michigan,

Minnesota, Missouri, Nebraska, North Dakota, South Dakota, and Wisconsin. The low pressure

system moved across Oklahoma into eastern Kansas on the 9th and reached Missouri by daybreak

on the 10th. It then curved sharply to the north-northeast, deepening rapidly and intensifying as it moved over eastern Iowa into southeastern Minnesota. The storm track is shown in

Figure 6. During the night of the 10-llth, record low pressures were reported at Rochester

(97.0 kPa) , Minneapolis /St. Paul (96.9 kPa) , and Duluth (96.7 kPa) , Minnesota.

Wind gusts of up to 130 km/h (36 m/s) were observed in Minnesota, 122 km/h (34 m/s) in

Iowa, 112 km/h (31 m/s) in Indiana and Wisconsin, and 97 km/h (27 m/s) in Illinois, Nebraska,

and North Dakota. Strong winds and cold temperatures resulted in wind chills below -62°C

(211°K) in Minnesota, North Dakota, and South Dakota. Snowfall measured up to 58 cm (.58 m)

in Minnesota, 48 cm (.48 m) in Nebraska, 38 cm (.38 m) in Iowa, and 25 cm (.25 m) in North

Dakota.

11-22
1/11/15Z • 96.6

35

Fargo»

-SOUTH DAKOTA

a^id City

J^^ternational

1/11/09Z

Duluth^A 96.6 MINNESOTA Tw

1/11/06Z|» 97.0 ^1

1/11/03X I

Lander

i/9/03Z

99.2 j

Cheyenn

Huron

t

NEBRASKA

North Platte

^ -^J^ IOWA

-^m6t_l ,8.9 I

fJ^"-* ^"notion

1 Minneapolis^ • 97.2

f ^TJ GreenBa^

|\ WISCONSIN

1/11/OOZ • 97.6 /-^

1/10/21Z •

97.6 V

Dubuque*. I ; Chicago;

^ J ILLINOIS

isz'm 97.7 f Springfielc

Lincoln J

1/10/1 ^

KANSAS

. , ' # i Dodge i^uuge City uv f

1/9/12^/^^^ p

1/10/12ZI#'

I-JLJ^-"^ J_ 1/10/09Z

1/9/15Z •!^1"9"8.9 t

'/10/06Z-#^',

i/io/isz«^sas City

' 98.3 St. Louisf

1 Jii

17 MISSOURI

98.6

98.6 99.2

ARKANSAS JPfTPj?

Figure 6. Storm track and central pressures (kPa) of the

mid-January blizzard.

11-23
The blizzard caused 31 deaths and hundreds of injuries. Damage from the wind, snow and

cold temperatures consisted of towers blown down, roofs blown off, broken windows, fallen

trees, severe livestock losses and collapsed buildings. Power outages were common and

transportation was stalled.

Another January extra-tropical storm swept across the State of Washington. Gusts near

112 km/h (31 m/s) toppled trees, flipped aircraft and caused widespread power outages.

Considerable wind damage to irrigation equipment in the eastern part of the State was

reported.

One of the worst March blizzards since 1951 struck Wisconsin, Minnesota, and Michigan

on the 23-24th. Official observations recorded gusts in excess of 130 km/h (36 m/s), while

unofficial reports exceeded 162 km/h (45 m/s). Wind-driven waves of 6.1 m pounded the shore

of Lake Superior washing away large sections of the beach and causing extensive flooding.

Wind, snow, ice, and water resulted in collapsed buildings, towers and signs, power outages,

and cancellations of most activities. Six days later the area was struck by another blizzard

with winds up to 112 km/h (31 m/s).

A severe storm struck Maryland, West Virginia, Pennsylvania, New York, and New England

on 2-5 April. Western New York recorded the worst April snowstorm since 1894 as 61 cm (.61

m) of snow fell. Winds associated with the storm reached 241 km/h (67 m/s) atop Mt. Washington, New Hampshire, 162 km/h (45 m/s) in Maryland, 148 km/h (41 m/s) in Pennsylvania, 140

km/h (39 m/s) in Maine, and 126 km/h (35 m/s) in Massachusetts. Buildings were destroyed,

power lines were downed, trees were broken, mobile homes were blown over, windows and signs

were broken, and fishing equipment along the Atlantic coast was extensively damaged.

11-24
An intense low pressure system spread across the Midwest on 9-10 November. Winds in

excess of 97 km/h (2 7 m/s) were reported at several stations in Illinois, Indiana, and Iowa.

One hundred twenty-six km/h (35 m/s) winds were recorded in Michigan. On 29-30 November

another low pressure system moved across the Midwest with similar wind speeds reported.

During both storms, damage was widespread to power lines, trees, windows, buildings, and

mobile homes.

4. OTHER SEVERE WINDSTORMS

Thunderstorms in the United States during 1975 often produced locally damaging winds in

excess of 79 km/h (22 m/s). One of the more intense thunderstorm outbreaks occurred on 22

May in Minnesota and South Dakota, Several areas in these states reported winds above 162

km/h (45 m/s) that destroyed farm buildings, mobile homes, aircraft, trees, and power

lines. An outbreak on 16 June in Nebraska and Oklahoma also caused winds of about 162 km/h

(45 m/s). Damage to crops, farm equipment, and mobile homes was extensive. On 13 December,

thunderstorms in Kansas resulted in winds estimated at 184 km/h (51 m/s) that damaged farm

buildings, mobile homes, and trees.

Dust-laden winds gusting to 115 km/h (32 m/s) in the northern two-thirds of Arizona in

May caused widespread property damage. Many mobile homes were damaged, sections of roofs

were blown away, storm windows were blown out, signs were blown down and large trees fell on

houses, cars, and power lines. Traffic accidents were numerous as dust reduced the visibility to zero.

California experienced a sandstorm in June. At the height of the storm a wind speed of

144 km/h (40 m/s) was recorded. Damage to utility company facilities was about $100,000.

In addition thousands of automobiles and trucks had paint blasted off, pitted windshields

and chrome, and damaged engines. In one area more than 100 automobiles were half-buried in

sand

.

11-25
5. HURRICANE ELOISE

The track of Hurricane Eloise across the Caribbean Sea and Gulf of Mexico is shown on

Figure 7. On 23 September, she moved onshore in the Florida Panhandle with a minimum

pressure of 95,5 kPa. Above normal tides and high winds caused considerable damage, even

though the storm rapidly lost strength as it moved northward over the land.

11-26
Northwest Florida was battered by sustained winds as high as 202 km/h (56 m/s) and

gusts as high as 248 km/h (69 m/s). Tides along an 80 km strip of shoreline were 3.7 to

4.9 m above normal. Damage in this part of Florida was estimated at 150 million dollars.

Wind-driven storm tides eroded 9.1 to 21.3 m of sand and undermined beach front structures.

Eighty-five to 90 percent of the buildings between Fort Walton Beach and Panama City on

the water side of the coastal highway were severely damaged or destroyed. Airplanes and

boats were also destroyed. The hurricane affected one million acres of timber, and crop

damage was severe.

Away from the center of the hurricane, wind gusts of up to 194 km/h (54 m/s) were

reported in Alabama and 130 km/h (36 m/s) in Georgia. Gale winds associated with the

storm were reported as far west as Louisiana and as far north as Tennessee. Crop damage

in these states exceeded one million dollars - 50,000 pecan trees were uprooted, the corn

crop was virtually a total loss, and the cotton and soybean crops were severely damaged.

In addition, hundreds of buildings were damaged and some were destroyed, numerous mobile

homes were either damaged or destroyed, power outages were widespread, and thousands of

trees were felled.

A unique set of hourly meteorological data and some oceanographic data was collected by

two buoys during the approach and passage of the eye of Hurricane Eloise. These data,

along with simultaneous ship, aircraft, and satellite data are being assembled for hurricane research studies.

From 19-25 September buoy EB-04 at 26.0 N, 90.0 W collected air pressure, air and surface water temperature, dew point, solar radiation, wind speed and direction, surface

salinity and surface current direction data. The second buoy, EB-10, at 27.5 N, 88.0 W

collected from 20-26 September the above data plus wave height, wave period and precipitation; and subsurface temperature, pressure, salinity and current information. In addition,

wave spectra are available from 20-24 September. Most of the data collected are 15 minute

11-27
averages. Time series of sea level pressure and wind speed at EB-10 are shown on Figure 8.

The passage of the eye on 23 September is distinctly represented by the sharp drop and rapid

recovery of both the pressure and wind speed.

Figure 8. Time series of the sea level pressure and wind speed

for buoy EB-10 during the passage of Hurricane Eloise.

The assembled data package from the National Climatic Center will contain digitized

buoy data, land station data (surface and upper air), ship observations and satellite

data. It will also contain microfiche of records of climatological observations. National

Hurricane Center bulletins and advisories, barograms, hourly precipitation data and

meteorological observations forms. Surface and upper air synoptic charts and forms,

satellite photographs, reconnaissance data and radar logs will be microfilmed, and 16 mm

films of satellite and radar data will be available. In short, a complete data package on

Hurricane Eloise will be available at the National Climatic Center for researchers.

11-28
6. SUMMARY

Several severe windstorms caused loss of life and extensive damage in the United

States during 1975. During each month some location in the United States was battered by

strong winds from either relatively short-lived meso-scale storms (like tornadoes) or longlived large scale storms (like hurricanes or severe extra-tropical storms) . Some new

records on frequency of storms and on damage were established during 1975. But the year was

not highly unusual — new records of some sort are established somewhere in the United

States almost every year. Engineers designing systems to withstand wind damage must be

cognizant of the likelihood of such extreme winds and of the establishment of new records.

Meteorologists must continue work to better define the extreme wind environment.

7 . ACKNOWLEDGEMENTS

The help of Mr. Henry Vigansky of the National Climatic Center is greatly appreciated.

The cooperation of the management of the Omaha World-Herald, Omaha, Nebraska, for granting

permission to use their photographs is also appreciated.

11-29
MEAN SPEED PROFILES OF HURRICANE WINDS

EMIL siMIU

Structural Research Engineer

Center for Building Technology

National Bureau of Standards

Washington, D.C.

VIRENDRA C. PATEL

Associate Professor

Division of Energy Engineering

University of Iowa

J Iowa City, Iowa

JOHN F. NASH

Vice President

Sybucon , Inc

Atlanta, Georgia

ABSTRACT

A numerical solution of the hurricane boundary layer problem is presented in which

the hurricane is modeled as a steady, axisymmetric , neutrally stratified flow. The turbulence effects in the flow are accounted for by the phenomenological relations proposed by

Bradshaw et al, and Nash, which provide a considerably more realistic picture of the actual

flow than the pseudolaminar model used in previous solutions of the boundary layer problem.

The results of the calculations obtained on the basis of the model just described suggest

that: (1) in the height range of interest to the structural designer, say up to a height

of 400 m above ground, it is permissible to use the logarithmic law to represent the mean

velocity profile of hurricane winds and (2) if the relation between wind speeds in different

roughness regimes which is valid in extratropical storms is applied to hurricane winds, the

speeds over built-up terrain, calculated as functions of speed over open terrain, may be

underestimated by about 10% and 10-20% in suburban and in urban exposure, respectively.

The corresponding mean loads are then underestimated by about 15% and 30%, respectively.

Key Words: Boundary layer; Hurricanes; Loads (forces); Natural analysis; Tall buildings;

Wind Profiles.

1. Introduction

In view of the considerable influence upon tall building design of the assumptions

regarding the variation of mean winds with height, the question has been raised as to whether or not mean wind profiles in mature hurricanes differ significantly from profiles

typical of extratropical storms [6, 16, 21]. In the Southern Standard Building Code [19]

it is assumed that hurricane wind speeds increase with height only up to an elevation of

100 ft (30.5 m) above ground and are constant above that elevation, i.e., that hurricane

winds are characterized by flatter profiles than winds associated with inland storms. 11-30
On the other hand, in both the 1955 and the 1972 versions of the American National Standard

A58.1 [1, 2] it is assumed that there are no differences between profiles in the two types

of storms. Several attempts to provide a description of the boundary layer in a mature hurricane

have been reported in the literature. However, as noted in Ref. 16, the results obtained

so far have been largely inconclusive. Hurricane wind speed data recorded in Florida

suggest that mean profiles are considerably flatter in hurricanes than in extratropical

storms. However, these data were discounted by Gentry [6, 21] as being of dubious reliability. It has also been argued [2] that no inferences valid for mature hurricanes can be

drawn from data measured at Brookhaven, N.Y. in 1950, which are representative of a hurricane in the decaying stage. Finally, existing analytical solutions [4, 14, 18] are based

upon the unrealistic assumption of pseudolaminarity and are therefore capable of providing,

at best, a qualitative picture of the flow.

The occurrence of a hurricane at any one location is a rare event. It is therefore

difficult to obtain relevant measurements on the variation of hurricane speeds with height

and on the relation between hurricane wind speeds in different roughness regimes. This

being the case, it appears reasonable to attempt an analytical solution of the hurricane

boundary layer problem, based on a realistic representation of the turbulent flow. Such

a solution is described herein. Numerical results corresponding to typical situations of

interest to the designer and code writer are presented. Current building code provisions

on hurricane winds are then assessed in the light of these results.

2. Analytical Model Method of Solution

Following Rosenthal [14] , Carrier et al [4] and Smith [18] , it is assumed that the

hurricane boundary layer flow is steady, axisymmetric and neutrally stratified. Except

in the immediate proximity of the eye, where the boundary layer assumptions break down [18]

,

the Reynolds equations may be written as

3U

3z

V

R

2 (1)

U

3V_

3R

+

R p 3z

V = 0 (2)

The equation of continuity is

(3)

11-31
Here, R is the radial distance measured from the storm center, z is the distance measured

normal to the earth surface, U, V and W are the radial, circumferential and normal (vertical) mean velocities, respectively, p is air density, p is static pressure, f is the

Coriolis parameter (in this work it was assumed f=5xl0^s''") and and are the

turbulent Reynolds stresses (the molecular stresses being negligibly small) in the radial

and circumferential directions, respectively.

According to Rosenthal [14] , measurements indicate that the expression

= exp (-R /R) W p dR pj^2 vm ^ ' vm

is quite representative of the pressure field in hurricanes. In Eq. 4, R^ = v.alue of the

radius, R, at which the gradient wind, ^g^,' is a maximum and Ap = difference between the

high pressure in the far field (R - 1000 km, say) and the low pressure at the storm center.

The relation between pressure gradient and gradient velocity is given by the classical

frictionless balance equation in cyclonic flow [7]

,

V

2 _l_dp^^r^ (5) p dR R ^ ^^gr

The boiHidary conditions, at the top and bottom edges of the boundary layer region were

expressed as follows

z = 0 U = 0 V = 0 W = 0

(6)

grin actual computations, condition (7) is satisfied with sufficient accuracy at some finite

distance of the order of the boundary layer thickness 6. Previous experience [4, 18] has shown that the solution of the equations of motion

is not unduly sensitive to small variations in the boundary conditions at the far field

radius. In this work, it was assumed that at R = 1000 km, the velocity and shear stress

profiles are the same as in horizontally homogeneous flow over a rough plane surface, with

a boundary layer thickness 6 = 1.4 km. To solve the equations of motion, phenomenological relations must be specified that

describe the Reynolds stresses T^, T^. In this work, the relations proposed by Bradshaw

et al. [3] and Nash [8] have been employed. These relations have been shown to result in

successful predictions in a variety of boundary layer problems [11, 17] and have recently

been also employed in micrometeorological work [13] ; as shown by Nash [8] , they are particularly suitable for use in the three-dimensional boundary layer problems. For the case

of the hurricane boundary layer, and assuming, as in Ref. 8, that the directions of Reynolds

stress vector and of the mean rate-of-strain vector coincide, the closure relations may be

11-32
written in the form

•J^+^-^+^a It — +t ^} + 2a , — a ,T + - / - = 0 dK OZ I V 02 U o2 I oz ^ L p(8)

and

T T

u

31J_ 8V_ 3z 3z

In Eq. 8, which may be interpreted as an empirical model of the equation expressing the

2 2 1/2 conservation of the turbulence kinetic energy, T = (t + T ) , a, is an absolute convecu V 1 tive constant (= 0.15), and a^ and L are universal functions of z/6 which model, respectively, the diffusion and dissipation of turbulence energy across the boundary layer. The

empirical functions L and a^ used are the same as those given in Ref. 9. The system consisting of the Reynolds equations, the equation of continuity and the

closure relations, with the given boundary conditions, was solved numerically using a time-relaxation procedure developed by Patel and Nash and described in some detail in Ref.

12. The numerical solution of the boundary layer equations was terminated at a height

above ground of 0.05 times the local boundary layer thickness and then matched with the

logarithmic law

"- o

1/2 assumed to be valid below that height. In Eq. 10, u. = shear velocity = (T /p) , T = resultant shear stress at the ground, p = air density,

* Q = resultant mean wind

WW

velocity,

K = von Karman constant, = roughness length.

3. Numerical Results

Nine cases were investigated, covering four different roughness regimes. Values of

the parameter were used in the calculations that correspond, roughly , to flow over

open water (z = 0.0023 m) , flow over open terrain (z = 0.012 m and z = 0.03 m) , flow

o o o over suburbs or towns (z = 0.35 m) and flow over large cities (z = 0.90 m) . For simplio o

city, the terrain roughness was assumed to be uniform. The profiles were calculated at a distance, R, from the storm center equal to the radius of maximum gradient winds, R , vm

except that in one case the profile was also calculated at R = 2 R , in addition to R = vm

R . Typically, it was assumed R = 30 km and Ap = 100 mb. These assumptions are convm -''^ vm

sistent with observed data (see for example, Ref. 20) in hurricanes of great intensity.

To assess the influence upon the results of the magnitude of A and R , additional calcu- ^ p vm

lations were performed with Ap = 60 mb, Ap = 140 mb and R^ = 50 km. The values of the

11-33
parameters for the various cases investigated are given in Table 1. The results of the

calculations, based on Eqs. 1-9, are listed in Table 2 [columns (1) , (2) , (3) , (5) , (7)

,

(9)]. Table 2 also includes mean speeds calculated in accordance with the logarithmic law

[columns (4) , (6) , (8) , (10) ] .

V(z) =^.„

4. Interpretation of Results

a) Logarithmic Profiles and Calculated Hurricane Wind Speeds.

It can be seen from Table 2 that, in the height range of interest to the structural

designer, hurricane wind speeds, as determined on the basis of Eqs. 1-9, are smaller by

less than 3% than speeds determined in accordance with Eq. 11.

(It is noted that the results reported in Ref. 12 show that the differences between

the resultant mean speed Q(z) and the tangential component V(z) are negligibly small - of the order of 1% or less) . Since the logarithmic law is an excellent representation of

wind profiles in the lower few hundred meters of extratropical storms [10, 15], it

follows, according to the results of Table 2, that mean profiles in hurricanes and in

extratropical cyclones may be assumed to have approximately the same shape. It thus

appears that, insofar as the power law is an approximation of the logarithmic law, the

assumption regarding mean wind profiles used in the A58.1 Standard [1,2] is justified,

and that, on the other hand, the assumption implicit in the Southern Building Code [19]

,

according to which, at elevations higher than 30 m above ground, hurricane winds do not

increase with height, may be unsafe from a structural design viewpoint.

b) Wind Speed Reduction due to Increased Roughness of Terrain.

In terrain that is built-up over a sufficiently large getch, the higher friction

at the ground will retard the flow more effectively than is the case over open terrain.

Thus, at any elevation within the boundary layer, mean wind speeds will be lower in builtup, than in open terrain.

Wind maps included in the A58.1 Standard 1972 [2] specify wind speeds at 30 ft (9.15

m) above ground in open terrain. To calculate the corresponding speeds over built-up

terrain (suburban or urban) it is necessary that a model be assumed, describing the relation between wind speeds in different roughness regimes. Two such models are currently in use [2, 15] . The model presented in Ref. 15, referred to as the similarity model, is

based on recent results of theoretical and experimental boundary layer research [5] , and

rests on the assumption that the flow in the free atmosphere is geostrophic. In accor- dance with this model, the friction velocities, u^^ and u^, over terrains of roughness

length z and z , respectively, are related as follows [15]

11-34
'J,

2 20.25 + 2n —r z o 1/2 (12)

2

"•••1 20.25 +

o 1

where f = Cordis parameter. The model included in Ref. 2, referred to as the power law

model, assumes the validity of the equation

v(2) = v(io) {^)- (13)

where a = 1/7 in open terrain, a = 1/4.5 in suburban terrain and a = 1/3 in urban terrain.

This model assumes, in addition, that the gradient height (i.e., the thickness of the

boundary layer) is z = 275 m, z = 366 m and z = 458 m in open, suburban and urban terrain, respectively. The magnitude of the wind speed at 10 m above ground in open terrain

being specified, the corresponding speed over suburban or open terrain is easily obtained

using Eq. 13, in which for all values of a, the gradient wind, \J{z^), is the same [2].

Implicit in the power law model, which is in effect an empirical approximation of

the similarity model, is the assumption that the flow in the free atmosphere is geostrophic

This assumption does not hold in the case of hurricane winds. Nevertheless, the power

law model is applied in the A58.1 Standard [2] not only to extratropical storms, for which

it may indeed be assumed that the flow is approximately geostrophic, but to hurricane

winds as well.

The question thus arises as to whether it is permissible to apply the same relation

between speeds in different roughness regimes - based on the assumption of geostrophic

flow - to both extratropical storms and hurricanes. This question will now be examined

in the light of the results of Table 2. A direct comparison between these results, based on the model consisting of Eqs.

1-9, and those obtained if the power law model is employed would present difficulties

insofar as the two models use different roughness parameters, vix., the roughness length,

z^, and the exponent, a, respectively. It will therefore be more convenient to compare

the results of Table 2 with those obtained if the similarity model is used. Let z = 0.03 m and z , = 0.0023 m, z , = 0.012 m, z , = 0.35 m and z , = 0.90 m. o ol ol ol ol

The ratios ^^*]_/^*) gjj^j-' obtained using Eq. 12 are given in Table 3, in which the ratios,

(u.Vu.), , based on Eqs. 1-9 and obtained from Tables 1 and 2 (cases 3, 6, 7, 8, 9) are *1 * hurr

listed for purposes of comparison (the subscripts extr, and hurr, indicate that the quantities involved are calculated for extratropical storms and for hurricanes, respectively)

.

The quantities [V(z)], and [V(z) ] , in Table 3 are mean speeds over terrain of

hurr extr

roughness z^^, calculated in accordance with Eqs. 1-9 and with Eq. 12 (in which = 0.03 m

respectively. Assuming the model consisting of Eqs. 1-9 to be correct, it is seen that the

similarity model (Eq. 12) results in underestimates of the mean speeds in suburban and

urban terrain of the order of 5-15% and, therefore, underestimates of the mean load (which

11-35
is proportional to the square of the mean speeds) of the order of 10-30%.

The results of Tables 2 and 3 are based on the assumption that the terrain roughness

is uniform throughout the area swept by the hurricane. In reality, the roughness corres- ponding to suburban or urban terrain only prevails over a small part of this area and

thus the flow over rougher terrain is retarded somewhat less effectively than indicated

by the results of Table 2. It may therefore be expected that the actual ratios (u^l/u^),

* * hurr

and [V(z)] ^ for the suburban and urban terrain cases are larger than indicated in Table

extr

3. It thus appears that if Eq. 12 - which does not take into account the presence of

inertia forces in the hurricanes - is used in structural engineering applications for

calculating hurricane wind speeds over suburban terrain, the error involved is for z^^ = 0.20 - 0.35 m, of the order of 5-10%. If Eq. 12 is applied for calculating hurricane

wind speeds over urban terrain (z^^ = 0.40 - 0.90 m) , according to the results reported

herein, the calculated speeds may be underestimated by about 10-20%.

5. Conclusions

A numerical solution of the hurricane boundary layer problem was presented in which

the hurricane was modeled as a steady, axisymmetric , neutrally stratified flow with a pressure field represented by Rosenthal's expression [14] (Eq. 4). The turbulence effects

in the flow are accounted for by the phenomenological relations proposed by Bradshaw et al

[3] and Nash [12] , which provide a considerably more realistic picture of the actual flow

than the pseudolaminar model used in previous solutions of the boundary layer problem

[4, 14, 18].

The results of the calculations obtained on the basis of the model just described

suggest that:

1. In the height range of interest to the structural designer, say up to a height

of 400 m above ground, it is permissible to use the logarithmic law to represent

the mean Velocity profile of hurricane winds. Accordingly,

(a) the assumption used in the A58.1 Standard [2] , according to which the variation of mean winds with height follows the same law in both extratropical

storms and hurricanes is acceptable.

(b) the assumption implicit in the Southern Standard Building Code [27]

,

according to which, at elevations higher than 100 ft (30.5 m) above ground,

hurricane wind speeds do not increase with height, may be unsafe from a structural design viewpoint.

2. If the relation between wind speeds in different roughness regimes which is valid

in extratropical storms is applied to hurricane winds , the speeds over built-up

terrain, calculated as functions of speed over open terrain, may be underestimated

by about 5-10% and 10-20% in suburban and in urban exposure, respectively. The

corresponding mean loads are then underestimated, roughly, by 15% and 30%, re- spectively .

11-36
Acknowledgements

The work presented herein is part of a study of hurricane effects on port facilities

conducted by the Center for Building Technology, National Bureau of Standards and supported

by the Maritime Administration, U.S. Department of Commerce. The calculations presented

herein were planned and carried out by Sybucon, Inc.

Appendix I - References

1. "American Standard Building Code Requirements for Minimxim Design Loads in

Buildings and Other Structures," American Standards Association, New York, N.Y., 1955.

2. "American National Standard Building Code Requirements for Minimum Design Loads

in Buildings and Other Structures A58-1," American National Standards Institute, New York

N.Y., 1972.

3. Bradshaw, P., Ferris, D.H. , Atwell, N.P., "Calculation of Boundary-Layer Development Using the Turbulent Energy Equation," Journal of Fluid Mechanics , Vol. 28, 1967,

pp. 593-616.

4. Carrier, G.F., Hammond, A.L. , George, O.D., "A Model of the Mature Hurricanes,"

Journal of Fluid Mechanics , Vol 47, pp. 145-170, 1971.

5. Csanady, G.T., "On the Resistance Law of a Turbulent Ekamn Layer," Journal of the

Atmospheric Sciences , Vol. 24, Sept., 1967, pp. 467-471.

6. Gentry, R.C. , discussion of "Nature of Wind" by R.H. Shelock, Journal of the

Structural Division , ASCE, Vol. 85, No. ST3 , Proc. Paper, 1988, Part 1, Mar., 1959, pp. 170-172.

7. Haltiner, G.J. and Martin, F.L. , Dynamical and Physical Meteorology , McGraw-Hill Book Co., Inc., New York, N.Y. , 1957.

8. Nash, J.F. , "The Calculation of Three-Dimensional Turbulent Boundary Layers in

Incompressible Flow," Journal of Fluid Mechanics , Vol. 37, 1969, p. 625-642.

9. Nash, J.F. and Patel, C.V. , Three Dimensional Turbulent Boundary Layers , S.B.C.

Technical Books, Atlants, 1972.

10. Pasquill, F. , "wind Structure in the Atmospheric Boundary Layer", Philosophical

Transactions of the Royal Society of London , A. 269, 1971, pp. 439-456.

11. Patel, V.C. and Nash, J.F., "Some Solutions of the Unsteady Turbulent BoundaryLayer Equations," Proceedings , lUTAM Symposixom, Quebec, 1972.

12. Patel, V.C. and Nash, J.F. , "Numerical Study of the Hurricane Boundary Layer

Mean-Wind Profiles," Report prepared for the National Bureau of Standards by Sybucon, Inc.,

Atlanta, Georgia, June, 1974.

13. Peterson, E.W., "Modification of Mean Flow and Turbulent Energy by a Change in

Surface Roughness," Quarterly Journal of the Royal Meteorological Society , Vol. 95, 1969,

pp. 560-575.

14. Rosenthal, S.L. , "A Theoretical Analysis of the Field of Motion in the Hurricane

Boundary Layer", Report No. 56 , National Hurricane Research Project, U.S. Department of

Commerce, Washington, D.C.

15. Simiu, Emil, "Logarithmic Profiles and Design Wind Speeds," Journal of the

Engineering Mechanics Division , ASCE, Vol. 99, No. EM5 , Proc. Paper 10100, October, 1973,

pp. 1073-1083.

16. Simiu, E. , "Variation of Mean Winds with Hurricanes," Journal of the Engineering

Mechanics Division , ASCE, Technical Note, Vol. 100, No. EM4, Proc. Paper 10692, Aug., 1974,

pp. 833-837.

17. Singleton, R.E. and Nash, J.F. , "A Method for Calculating Unsteady Turbulent

Boundary Layers in Two- and Three-Dimensional Flows," AIAA Journal , Vol. 12, No. 5, May

1974.

11-37
18. Smith, R.K. , "The Surface Boundary Layer of a Hurricane," Tellus , Vol. 20, 1968,

pp. 473-484.

19. Southern Standard Building Code , Birmingham, Alabama, 1965, p. 5„12 . 20. "Survey of Meteorological Factors Pertinent to Reduction of Loss of Life and

Property in Hurricane Situations," Report No. 5 , National Hurricane Research Project,

U.S. Department of Commerce, Washington, D.C. , 1957.

21. "Wind Forces on Structures," Final Report by Task Committee on Wind Forces,

Transactions , ASCE, Vol 126, Paper No. 3269, Part II, 1961, pp. 1124-1198.

Appendix II - Notations

vm

UV

Vw

z

zoAp

6K

PTU

T V

Convective constant in the turbulent shear stress equations

Diffusion constant

Dissipation length

Coriolis parameter

Static pressure

Resultant mean wind velocity

Radial distance from the storm center

Radial location of maximum pressure gradient

Radial location of maximum gradient wind

Shear velocity in terrain of roughness length z z , respectively

ol

Radial mean wind velocity

Circumferential mean wind velocity

Gradient wind

Vertical mean wind velocity

Vertical distance above the ground

Terrain roughness length

Pressure depression

Boundary- layer thickness

Karman constant in the Law of the Wall

Air density

Reynolds shear stress in the radial direction

Reynolds shear stress in the circumferential direction

Resultant shear stress at the ground

11-38
Table 1 Values of Parameters for the

Various Cases Investigated

Number R Ap z vm o

ki lometers mill i bars meters ki lometers

1 30 60 0.0023 30

2 30 lAO 0.0023 30

3 30 100 0.0023 30

4 30 100 0.0023 60

5 50 100 0..0023 50

6 30 100 0.012 30

7 30 100 0.03 30

8 30 100 0.35 30

9 30 100 0.90 30

^radius at which profile was calculated

11-39
Table 2 - Calculated Hurricane Wind

Speeds in meters/second [12]

Case 6 meters meters/ second

z (meters)

100 200 300 400

(1)

1 ,590

1 ,600

1 ,610

3.^30

2,i»80

1 ,950

2,200

3,200

3,790

(2)

1 .A5

2.2k

1 .89

1 .59

1 .80

2.16

2.36

3.06

3.kh

(3)| W

38.5|(38.8)

59.3|(59.8)

^9.9|(50.3)

42.3!(A2.3)

48.0(48.0)

48.8(48.8)

I hl.Uhl.l)

I 43.2j(43.2)

40. 6X40. 6)

(5) (6) (7)

40.6(41 .3)

62.6|(63.6)

52.7(53.7)

44.8'(45. 1)

50.6j(51 .2)

51 .7,(52.6)

51 .o|(51 .8)

48.0(48.5)

46.o'(46.5)

(8)

41 .8(42.8)

64.3'(65.9)

I 53.2j(55.5)

46.1j(46.7)

52.2j(53.0)

53.5|(54.8)

53. 1:(54.2)

I 50.5:(51 .6)

48.9;(50.0)

(9); (10)

42.3,(43.9)

65.l':(67.5)

I

I 54.9|(56.9)

47.0(47.9)

I 53.2(54.3)

i 54.5(56.3)

54.2X55-9)

52.3,(53.8)

50.9(52.5)

Note. Numbers in parentheses represent wind speeds calculated using the

logarithmic law (Eq. 11).

Table 3

- Calculated Ratios u,,. ^/u.,. and

f^(^)Jhurr/f^(^)^extr

"

[V(z)]^^^^ tV(^)^hurr

z , (meters) (u. /u.J, (u, ,/u,) ^

0.0023 0.862 .800 .93 .86

0.012 0.951 .916 .96 .91

0.03 1.000 1.000 1.00 1.00

0.35 1.189 1.229 1.09 1.19

0.90 1.257 1.463 1.16 1.36

11-40
Planning and Design of Strong-Motion Instrument Networks

R. B. MATTHIESEN

U.S. Geological Survey, Menlo Park, California

ABSTRACT

The types of research studies that utilize strong-motion data may be

classified as: source mechanism studies, ground motion studies, soil

failure studies, studies of the response of typical structures (including

soil-structure interaction effects), and studies of the response of equipment .

In planning networks and arrays to make these studies, criteria must be

established based on the tectonic setting, the seismicity or recurrence of

strong ground motions, the reliability of operations in different regions,

and a cost/benefit analysis of the data that may be obtained. A review of

the strong-motion records that have been obtained during the past 40 years

indicates significant variations in the recurrence of strong ground

motions in the seismically active regions of the western United States.

When combined with instrument costs, maintenance costs, and the reliability

of operations, these recurrence relations can be interpreted in terms

of the cost per record for different levels of motion. The benefits to be

derived from each type of study in each region need to be established.

Current plans call for additional arrays to be installed in California,

the Mississippi embayment, the Yellowstone Park region, and Alaska to study

the spectral characteristics of strong ground motions in these regions.

Special studies of local site effects and structural response are being

planned in the more seismically active regions of California. Similar

criteria and planning should be applied in the establishment of arrays of

strong-motion instruments on a worldwide basis.

Key Words: Cost effectiveness; ground motion; network design; strong-motion record.

I

III-l
INTRODUCTION

The primary input for the design of strong-motion instrument arrays comes from the

research needs in strong-motion seismology and earthquake engineering. The impetus hehind

much of that research is the application of the research results in engineering design and

reduction of earthquake hazards. This input defines general objectives to be accomplished

by an appropriately designed network but does not constrain the development of the network

geographically. A secondary input comes from the mission-oriented and regulatory agencies

that desire to monitor the response of critical facilities to assess their response during

potentially damaging earthquakes. The location of this instrumentation is constrained to

the specific structure or system being monitored but may add significant data to that

obtained from a network of research instruments. The mission-oriented agencies also

influence the design of the network through their need for additional research results on a

timely basis.

The types of studies that utilize strong-motion data may be classified as follows:

- Studies of the source mechanism.

- Studies of the spectral characteristics of strong ground motion

and of the variations of these characteristics with the nature of

the source, the travel path and regional geology, or the local

site conditions.

- Studies of soil failures such as soil liquefaction or landslides.

- Studies of the response of representative types of structures and

interconnected systems at potentially damaging levels of response

and of the influence of the foundation conditions on this response.

- Studies of the response of equipment which may be free-standing

or mounted on structures.

The first of these studies is a fundamental study in seismology. The remainder conveniently divide into ground motion studies and structural response studies and may involve

the use of instrumentation for either the research or monitoring function.

III-2
CRITERIA FOR NETWORK DESIGN

The development of criteria for the planning and design of networks and arrays of

instrumentation to measure ground motions involves the following steps: (1) the ground

motions must be estimated; (2) the costs of operations must be evaluated; and (3) the

benefit to be derived from the data must be assessed. A similar process is also required

as the first step in the process of planning instrumentation arrays for structures

(Rojahn, 1976).

The estimation of ground motions involves a determination of the tectonic setting,

the seismicity of the region, and the recurrence of strong ground motions. This is

similar to the process of obtaining ground motions for use in analyses of seismic risk or

for use in establishing design levels for critical facilities (Algermissen and Perkins,

1972; Hays, et al, 197 5). In this basic approach to estimating ground motions, the

source characteristics are modeled in terms of the recurrence of earthquakes of different

magnitudes; the transmission of the motion is modeled as an attenuation of peak acceleration; and the motion at the site is obtained as a recurrence relation for particular

site conditions (see fig. 1). More refined techniques are the subject of current research

in which the source characteristics are modeled in terms of the expected stress drop and

source dimension; the transmission of the motion is modeled in terms of the wave propagation, attenuation, and dispersion; and the site effects are modeled in terms of their

influence on the spectral characteristics of the motion.

Several authors have gathered a considerable amount of data indicating that for

appropriately large source regions, the recurrence of earthquakes of different magnitudes

can be represented as straight lines on semi-log plots (see Algermissen, 1969, for

example). On the other hand, the existing data on the attenuation of strong ground

motions indicate that there is a considerable amount of scatter in the relation of the

peak acceleration, velocity, or displacement to the distance from the source. Figure 2

presents the attenuation of maximum acceleration with distance from the source for all of

the data recorded during the San Fernando earthquake (see Maley and Cloud, 1971). An

order of magnitude difference may be seen in the peak accelerations recorded at any one

distance. Figure 3 presents the attenuation of peak accelerations with distance from the

epicenter for all data recorded at Ferndale, Calif., during the past 40 years. A considerable amount of scatter is evident in this plot also. This large amount of scatter in

III-3
the data casts some doubt on the validity of the simplified model of transmission of

motion.

The availability of data from several stations that have been installed for about

40 years provides a more direct approach to the evaluation of the recurrence of strong

ground motions. For example, the results obtained from Ferndale are shown in figures 4

and 5. In figure 4, the cumulative numbers of events for which peak accelerations have

exceeded selected levels are plotted versus the year in which the event occurred. The

levels of peak acceleration used in figure 4 were selected to illustrate the approach

used. Similar results can be obtained for each of the levels of peak acceleration

recorded at this site. No attempt has been made to distinguish between foreshocks, main

events, and aftershocks in compiling these data. Straight lines have been fitted to the

data, and the slopes of these lines define the "events per year" for each of the selected

values of peak acceleration.

All levels of peak acceleration from the complete set of records obtained at

Ferndale were used to construct figure 5, in which the cumulative distribution of events

per year is plotted versus the peak acceleration values. The end points are shoxm as

circles in this figure, signifying an insufficiency in these data. (The value at a peak

2 acceleration of 10 cm/sec appears to be too low as a result of the fall off in the number

of low level events that are recorded by an Instrument that is triggered by the event

being recorded, whereas the values for the highest peak accelerations are obtained from

fewer than five events, and this is considered to be an insufficient amount of data). The

amount of scatter in the data in figure 5 is relatively small compared with the amount of

scatter indicated by the attenuation plots in figures 2 or 3.

Data of the type shown in figure 5 have been obtained for all of the strong-motion

instrument sites that have been in place for about 40 years (table 1). Only in three

cases (Ferndale, Hollister, and El Centro) are the data sufficient to provide statis2 tlcally meaningful results for peak accelerations up to 100 cm/sec . Of equal importance,

however, is the fact that at several sites in these "seismlcally active" areas, no estimate

of recurrence could be made after 40 years of recording. For example, in the San Francisco

Bay region, no reliable estimates of recurrence could be made for Golden Gate Park or San

Jose, although a maximum value of greater than lOu cm/sec has been recorded at each site.

In the Los Angeles basin, no reliable estimates can be made for Westwood or Pasadena.

III-4
Similarly, although 12 records have been obtained at Helena, Mont., only three of these

have been recorded since 1940 and none since 1960. The results in California are in sharp

contrast to other estimates of recurrence that yield equal rates along most segments of

the San Andreas fault. These results are also an indication of the serious difficulties

in any attempt to provide a rational plan for obtaining the desired strong-motion records:

potentially damaging earthquakes in any one area occur infrequently, and our basic understanding of the processes and recurrence of potentially damaging ground motions is therefore inadequate.

The cost of maintenance has been found to be about three times the cost of the

instruments themselves (depreciated over a 20-year life). As a result, the procedures

used in instrument maintenance need to be critically evaluated. In particular, the

service interval may be lengthened if an evaluation indicates that this will not result

in serious depreciation in either the quality or number of records recovered. The results

of a study of the length of the service interval is shown in figure 6. In the early days

of the program, a service interval of 2 months had been established. As the numbers of

instruments being maintained dramatically increased in the late 1960's, this interval was

perforce increased to 3 months. More recently, a general policy of servicing at a nominal

4- month interval has been adopted (a selected group of instruments are being serviced at

6-month intervals). Since the current cost figures are based on a nominal 3-month service

interval, an increase to a 6-month interval will significantly decrease the maintenance

costs, although they are not expected to decrease by half since it is planned that more

time should be spent at each instrument when the service interval is lengthened. As a

part of the evaluation of maintenance procedures, all of the older instruments are being

replaced with modern instruments, and the modern types of instruments in service are being

modified to bring them up to present specifications. This upgrading of the instruments

should raise the lower of the two sets of lines shown in figure 6.

From the recurrence data summarized in table 1 and the average costs of instruments

and maintenance ($400 per year), the costs per record for records with peak accelerations

greater than specified amounts can be obtained (see table 2). At most sites, the cost

doubles as the level of peak acceleration doubles. Estimates of these costs for other

sites must be determined if planning criteria are to be firmly established. Since peak

2 accelerations on the order of 200 cm/sec are the minimum levels of potentially damaging

motions, significant costs must be anticipated if we are to record potentially damaging

III-5
levels of ground motion at many of these sites.

The benefits that will be derived from the data that will be obtained must be estimated in order to assess the proper significance of these costs. Obviously, the first set

of data that will permit some of the unanswered questions regarding the nature of the

strong ground motions from earthquakes in the eastern part of the United States will be

2 of considerable benefit, whereas additional records at 50 cm/sec obtained at many of the

sites in California are of little benefit. It is clear, in general, that those studies

that can be accomplished in the more active areas may cost one tenth as much as they would

in other areas of California. Thus, studies of local site effects should be planned for

Ferndale, Hollister, and El Centro, if the local soil conditions permit. Studies of lowrise buildings should be conducted in these same regions, whereas studies of high-rise

buildings can be conducted only in the San Francico or Los Angeles areas.

GENERAL OBSERVATIONS

Most of the strong-motion records obtained to date have been obtained in California,

and the techniques for estimating ground motion spectra are largely based on these records.

Preliminary evaluations for other regions of the United States suggest that the Mississippi

embayment and Yellowstone Park regions may provide as much data, and as inexpensively, as

some of the less active areas of California. On the other hand, high maintenance costs

in Alaska offset the advantage of the generally high level of activity in that region.

General information on the influence of local site conditions on the spectral amplitudes of ground motion may be obtained from the regional networks by placing instruments

in different geologic settings or at sites with different soil conditions. More detailed

studies will require an expensive instrijmentation program including dow-hole instruments.

These should be conducted in regions where the seismic activity is sufficiently high to

insure an adequate return on the investment in instrumentation and its maintenance.

Similarly, instrumentation designed to study soil failures through liquefaction or

landsliding can be incorporated into the regional arrays if areas subject to soil failure

are identified. Remotely recording instruments should be placed on the area of potential

landslide or liquefaction as well as on nearby stable ground. Extensive instrumentation

for these studies should be installed only in highly active areas.

III-6
The information necessary to assess the costs of obtaining strong-motion data from

all parts of the world should be assembled so that the greatest benefit may be derived

for all concerned with earthquake hazards and the loss of life that has occurred in past

earthquakes

.

REFERENCES

Algermissen, S. T. , "Seismic Risk Studies in the United States", Proceedings Fourth

World Conf. on Earthquake Engr ., Santiago, Chile, 1969.

Algermissen, S. T. and Perkins, D. M. , "A Technique for Seismic Zoning", Proceedings

Conference on Microzonation , Vol II, Seattle, 1972.

Hays, W. W. , et al, Guidelines for Developing Design Earthquake Response Spectra ,

TR M-114, Construction Engineering Research Laboratory, June 1975.

Maley, R. P., and Cloud, W. K. , "Strong-Motion Accelerograph Records", Strong-Motion

Instrumental Data on the San Fernando Earthquake of Feb. 9, 1971 , Seismological

Field Survey and Earthquake Engr . Res. Lab., California Institute of Technology,

Pasadena, 1971.

Rojahn, Christopher, "California Building Strong-Motion Instrumentation Program",

Proc. Conference on Dynamic Reponse of Structures: Instrumentation, Testing

Methods, and System Identification , UCLA, Los Angeles, March 1976.

III-7
Table 1 - Recurrence times for stations Installed for 40 years

Total Numb er Maximum Station Location Years /Event

Number

Years

of

Records

Accel

/ 2 cm/sec a > 25 a

a > 50 „ a m cm/sec

> 100

40 45 274 FERNDALE 1.5 3 6 40 11 230 EUREKA 8 (15)

40 3 124 GOLDEN GATE PARK -

39 6 52 ALEXANDER BUILDING (20)

39 11 48 SOUTHERN PACIFIC BLDG (10)

40 6 45 OAKLAND CITY HALL (12)

40 6 56 BERKELEY (15) -

41 4 138 SAN JOSE _

30 il lyi nULLis J.hK o

i. oo

40 9 172 SANTA BARBARA 10 -

40 3 100 WESTWOOD _

40 7 220 HOLLYWOOD (30) -

40 9 110 OCCIDENTAL BLDG 14 - 41 11 210 VERNON 10 20

41 10 250 LONG BEACH 10 (25)

40 4 100 PASADENA

40 10 46 COLTON 12 (25) -

41 24 314 EL CENTRO 3 6 12

41 9 30 SAN DIEGO (20) -

40 13 38 BISHOP 8

36 7 42 HAWTHORNE 10

38 12 115 HELENA

Numbers in parentheses are based on an insufficient amount of data.

III-8
Table 2 - Summary of costs per record

Station Location Maximum

Accel„

cm/ sec

Cost/Record in Dollars

a > 25 a > 50 a > 100

/ 2 a m cm/sec

FERNDALE 274 600 1200 2400

EUREKA 230 3200 (6000)

GOLDEN GATE PARK 124 - ALEXANDER BUILDING 52 (8000)

SOUTHERN PACIFIC BLDG 48 4000

OAKLAND CITY HALL 45 / /. Q n n ^ BERKELEY 56 \ bUOU )

SAN JOSE 138 -

HOLLibihK 800 1600 3200

SANTA BARBARA 172 4000

WESTWOOD 100

HOLLYWOOD 220 (12000) - OCCIDENTAL BLDG 110 5600 - VERNON 210 4000 8000

LONG BEACH 250 4000 (10000)

PASADENA 100

COLTON 46 5000 (10000)

EL CENTRO 314 1200 2400 4800

SAN DIEGO 30 (8000)

BISHOP 38 3200

HAWTHORNE 42 4000

HELENA 115

Numbers in parentheses are based on an insufficient amount of data.

III-9
JDoA/N^oi

III-IO
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III-12
III-13
5 .2

FERNDALE

PEAK ACCELERATION - cm/sec^

Figure 5. EVENTS PER YEAR VERSUS PEAK ACCELERATION - FERNDALE RECORDS

III-14
SiN3lNnaiSNI dO lN30a3d

III-15
OBSERVATION OF EARTHQUAKE RESPONSE OF GROUND WITH

HORIZONTAL AND VERTICAL SEISMOMETER ARRAYS

SATOSHI HAYASHI

Head, Structures Division

Port and Harbour Research Institute

Ministry of Transport, Japan

HAJIME TSUCHIDA

Chief, Earthquake Resistant Structures Laboratory

Structures Division

Port and Harbour Research Institute

Ministry of Transport, Japan

EIICHI KURATA

Earthquake Resistant Structures Laboratory

Structures Division

Port and Harbour Research Institute

Ministry of Transport, Japan

ABSTRACT

A horizontal seismometer array, having six observation points along a straight line

of 2 500 meters in length, have been established at the Tokyo International Airport. Each

observation point is equipped with two horizontal seismometers. Downhole seismometer

arrays have also been established at two points, one at the end of the observation line

and the other at a point 500 meters inside from the other end of the line. The observation started in April 1974 and since then 28 earthquakes has been recorded as of June

1975. Correlations among the ground motions at the points on the ground surface and the

cwo points in the ground where the downhole seismometers have been installed, have been

studied. The relative displacements between the points have also been studied.

It was assumed that there had been a straight pipeline made of steel along the observation line and the pipe motions had been equal to the observed ground motions. Then, the

stresses in the pipe were estimated, and it was found that the stresses due to axial deformations were remarkably larger than those due to the bending deformations.

Key Words: Arrays, Earthquakes, Ground motion, Records, Results, Seismometers.

1. Introduction

The Earthquake response analysis of ground over an area where a structure extends has

been recognized to be important and investigated by researchers and practicing engineers.

This is very important especially for extending structures ' such as pipelines and subaqueous

tunnels. Stresses in such structures due to earthquake ground motions depend not only on acceleration but also on the relative displacement of the ground along the route. For that reason, the earthquake response analysis of the ground along extended structures is important

I I I- 16
The data to be collected during earthquake studies should include the simultaneous

recording of earthquake ground motions at several points distributed along a line or in an

area. The strong motion earthquake instruments currently being operated in the world,

however, are installed to record ground motions at separate points. Only a few observations of earthquake ground motions, with seismometers arranged along lines, have been

utilized to obtain basic field data (1, 2).

The authors have established a horizontal seismometer array which consists of six

sets of seismometers located at an equal spacing of 500 meters along an observation line

of 2500 meters in length. Each set has two horizontal seismometers. Two downhole

seismometer arrays also have been established. The observation system was briefly re- ported at the 6th Joint Meeting, U.S. -Japan Panel on Wind and Seismic Effects, U.S. -Japan

Cooperative Program in Natural Resources (3).

The observations were started in April 1974 and as of June 1975, 28 earthquakes have

been recorded. In this paper, some observation results and analyses of three of the 28

earthquakes will be presented. Observations, however, are being continued as is the

analysis.

2. Seismometer Arrays

The horizontal seismometer array has been established in the Tokyo International

Airport, parallel to the C- runway, as shown in Fig. 1. The soil profile beneath the

array is shown in Fig. 2; the profile was constructed by engineering geologists based on

the two borings made for installation of the downhole seismometers and 59 borings which

had been made for planning and construction of the airport facilities.

The seismometers, to observe ground surface motions, are installed at six observation points, designated as points A through F shown in Fig. 1. The distance between

adjoining points is 500 meters and total length of the observation line is 2500 meters.

Each point is equipped with transducers to observe the longitudinal and transverse components of ground accelerations. At the points A and E the vertical seismometer arrays

are installed; the lower seismometers are at depths 67.2 meters and 49.6 meters below the

ground surface respectively. The downhole seismometers have three transducers to observe

longitudinal, transverse, and vertical components of the ground accelerations.

All the seismometers are of moving coil type where the output are recorded with four

electro magnetic oscillographs. Frequency ranges of the seismometers (within 10%) are

0.1 to 35Hz for the seismometers on the ground surface and 0.5 to 50Hz for the downhole

seismometers. The oscillograms were digitized for analyses with a computer on-line

digitizer at equal time intervals of 0.01 seconds.

3. Observation Results

Parameters of the three earthquakes, which are designated as TIA-3, TIA-6, and TIA-9

,

included maximum accelerations at the observation points and other parameters are listed

in Table 1. The locations of the epicenters are shown in Fig. 3, in which the direction

of the horizontal seismometer array is indicated as D.H.A. The ground acceleration

III-17
records (longitudinal component) of the earthquake TIA-3 at the points A, C, and E are shown in Fig. 4, which is a computer plot of the digitized records.

4. Displacement

Displacements of the earthquake ground motions were calculated from the observed

acceleration time histories. For the calculation, operation of a seismograph to the

observed ground acceleration was simulated with an analog computer. The seismograph has

a pendulum with a natural period of 6.0 seconds and amped at 55% of critical. The seismograph that was used is also being used in the network by the Japan Meteorological Agency.

Filtering was applied to the input and the output of the simulated seismographs in order

to eliminate components with periods longer than 10 seconds. The procedure used to calculate the displacement was compared with the procedure developed at the California

Institute of Technology in the United States in their processing of strong motion accelerograms. The displacements calculated by both procedures agrees very well, when the difference between the frequency ranges of the integrations in both procedures are taken

into account.

The calculated displacements for the earthquake TIA-3 are shown in Figs. 5 and 6. It can be seen from these figures that the displacement time histories, at all the observation points, are very similar to each other. The wave forms of the earthquake TIA-6

was similar to earthquake TIA-3 as shown in Figs. 5 and 6. However, earthquake TIA-9

did not show any similarity as the case of the other earthquakes . The maximum ground

displacements at the points are listed in Table 1. From the calculated ground displacements, the relative displacements of the two points adjoining each other were examined. The maximum relative displacements are listed

in Table 1.

5. Wave Propagation

The cross correlation functions were calculated between the time histories of the

calculated ground displacements at the points. The time-shifts for maximum correlation

were then obtained, as given in Table 2. The time histories at point F were used for all

calculations of the cross correlation function.

If it is assumed that the waves propagated from the epicenters straight along the

ground surface, the velocities of the wave propagations are estimated as 5.3, 5.9, and

4.4 km/sec for the earthquakes TIA-3, TIA-6, and TIA-9, respectively. It should be

noted, however, that regardless of the locations of the epicenters the cross correlation

analysis implies that the waves will arrive at point F sooner than at point A. Fig. 7 is a hodograph at point E from 10 to 20 seconds after the origin of the record

of the earthquake TIA-3.

III-18
6. Stresses in Underground Pipeline

Based on the earthquake records, induced stresses in steel pipelines due to the

ground motions were calculated. In perfojrming these calculations the following assumptions

were made: 1) steel pipes of different diameters were buried in the ground along the

observation line. 2) displacements of the pipes at the observation points were equal to

the ground displacements calculated from the recorded ground accelerations.

Maximum axial strains of the pipes were calculated from the maximum relative longitudinal displacements between two adjoining observation points. Where the axial strain means

a strain due to the longitudinal component of the ground motion. The axial strain was assumed to be uniform between the two points.

In calculating the maximum bending stresses, using the deflection curves of the

pipelines, were estimated as follows; three adjoining observation points were selected

and a parabolic curve was fitted to the displacements at the three points. This curve

was considered as the deflection curve of the pipelines, and the maximum bending strain

in a segment of the pipelines under consideration was then calculated. This procedure

was applied to four combinations of the observation points and for different 0.01 second

time intervals, over significant part of the ground motion records. Then, with the

maximum bending strain of the earthquake under consideration, the largest value was selected among the maximum bending strains calculated for different segments and times.

The calculated maximum axial strains and the maximum bending strains were converted

into stresses and listed in Table 3. Engineering requirements for oil pipelines have been established under Oil Pipelines

Enterprise Law (5, 6) , and a design formula to estimate stresses in steel pipelines due

to earthquake ground motions is given. Comparison ratio of the maximum acceleration

recorded by the downhole seismometers to the acceleration of gravity was taken as a design

seismic coefficient. Stresses were then calculated by the formula in the assumed pipelines along the observation line with the results shown in Table 3. In these calculations

the displacement of pipeline was considered to be equal to that of the ground. It will

be noted, however, that other engineers may reach slightly different results with the same

formula, since the details of application of the formula depends on judgement of each

engineer.

7. References

1) Akio Sakurai, Tadashi Takahashi, Hajime Tsutsumi, Hiroshi Yajima, Toshiro Noguchi

,

and Toru Iwakata: Dynamic Stresses of Underground Pipelines during Earthquakes (A Study

Based on the Observed Records in the Matsushiro Earthquakes) , Report No. 67058, Technical

Laboratory, Central Research Institute of Electric Power Industry, 1967.

2) Choshiro Tamura, Katsuyuki Kato, and Hiroshi Maeda : On Characteristics of Earthquake Ground Motion along a Line on Soft Ground, Proceedings of the 29th Annual Meeting

for Research Presentations, The Japan Society of Civil Engineers, Part I, October 1974.

3) Satoshi Hayashi and Hajime Tsuchida: Observation and Analysis of Ground Response

in Earthquakes, Proceedings of the 6th Joint Meeting, U.S. -Japan Panel on Wind and

Seismic Effects, U.S. -Japan Cooperative Program in Natural Resources, May 1974, Washington,

D.C.

III-19
4) M.D. Trifunac and V.W. Lee: A Note on the Accuracy of Computing Ground Displacements from Strong-Motion Accelerograms, Bulletin of the Seismological Society of America,

Vol. 64, No. 4, August 1974, pp. 1209-1219.

5) Ministrial Ordinance on Engineering Requirements for Facilities for Oil Pipelines

Enterprise, Ordinance No. 2 of Ministry of International Trade and Industry, Ministry of

Transport, Ministry of Construction, and Ministry of Automomy, 1972.

6) Ministrial Notification on Details of Engineering Requirements for Facilities

for Oil Pipelines Enterprise, Notification No. 1 of Ministry of International Trade and

Industry, Ministry of Transport, Ministry of Construction, and Ministry of Autonomy, 1973.

III-20

B

-20 » 20

Point, A

20 30 40 Time - seconds

Fig. 4 Acceleration records at poiacs A, C, and S Earthquake TIA-3, Lcnff,i cudir ai cov.oor.er.C

III-22
III-23
Fig. 7 Hodograph at poip.t E, Earuhquake TlA-2, frcm 10 to 20 seconds

after origin of recora

Zar thquake

Obser. or desi^

T I A

TI ATA

Cos. DesObs. Des.

.\xial stress 3040 9120 2340 6080 2G60 21280

3endir.g stress

Diameter: 20 cm 1 30 0 21 1 71

: 60 91 61 213

: 100 era 6 152 2 101 310

Total itress'^"' 304C 16G8C 2840 10790 2060 375oC

Unit:lcN/m^ (Interr.ar.ional unit system, Ikg/cm'^ =9 S . C665k;; /tt?" ) " Obser .: S tress derived from obierved ground motion-..

Design : S tress derived fro.n design formula.

* In calculanior. of total stress only axial stress is taken into ^ccci

Table 3 Comparison of observed and design stresses in

underground pipeline cue to earthquake ground motions

III-24
Record casignaeicn

Oaca

Time

Loc-, L. ion

Laci L i.. da

Long:. Lade

De;,. -.^

Zpicinw-'al discanca

Coj.po".. _nt

Ma;-ciniu:n accalarauion in jc

Point

3

C

D

£

F

Ma:cin;Lij. displ£cer.iant in c.a Point A

3

CD

S

"f

Maximam relative aiaplacaL-.ent in c::. A C 50 0 59 .) 16 0 0

B C c

'J "7 0 64 0 19 0 C & D 0 71 J 61 0

1 3 0 25 0 51

D & 0 74 0 53 Q 21 0 34 E & F G 54 G 64 53 2^ 0 22

'i^icle 1 Eartnquakas, maxirauiT. accelerations, r.",a.-.i;r.ur:; displacer.ents J and rusxir^un: relative diso lacemen ca

Earthcj " a

TIA-3

TIA-o

Conoonent

Longitudinal

Tranaverse

Lcr ;:itadinal

Transverse

Larii' itudinal

Tra.-jversa

May 1974

C3:33

Near ,3 Ccaat

of I:: i Pen.

33.5':: 133.

6.9

140 krr:

Lcn:' . T r an .

15. G 9.1

14.0 2.3

16 . 0

2

1

. 7 11 . 2 22 19 IS .5 ^3 .0

1. 91

1. 48 0

.

76 33 85

2. li 0. 31

1. 96 26

1. 52 0. 94

F- F- p

0. 13 0. 15

0. 20 0. 17

0. 15 0. 11

0. 13 0

.

12

0. 44 0. 3S

0. 43 0. 25

TiA-o

J a ly 3 , ^ . 4 14: 15 Jif Ibar._.';i Prefecture

36 . 4''v 14 1 . 2

" E 43 kn

6.3

161 krr.

Lens., Tr2n_.

6 4 5. 11 4

c

9 6 11 / 9 ., IL 7 11. 11 7 11.

0 52 0 6:^ c C

0 5; c 0 0 79

J 49 0 33

c 51 0 CO

Poin cs :-C F-3

0 . J. 5 0 . J. Id O.Gl

..08 G.J2

G. 11 0

0.17 C.12

-.23 G..0

. A9

Aur; . :. c 4 , 1 03:

3'v iL;aragi

Pref .jttare

36. 139. '2

50 '.en,

54 k.n

Lan., . Tran.

20.3 24. c 47.2 35.

S

36.0 31.0

-:9.3 22.5

51.8 32.^

40. 9 34.7

0 40 c G G 37

0 6 3 0 52

0 46 0 39

0 4 0 0 58

0 5a 0 55

3.31

0. K7 0.3^

0.27

0.26

F-L

O.Oii

00.C2

Tabae 2 Ti.v.e-shift tor T.ajii.ricm correlation ir. second

111-25
Building Strong-Motion Earthquake Instrumentation

CHRISTOPHER ROJAHN

U.S. Geological Survey, Menlo Park, California

ABSTRACT

Based on the recommendations of a special ad -hoc committee, twenty-one

geographic areas will be instrumented under the building instrumentation

phase of the California Strong-Motion Instrumentation Program, a statewide

program established by law in 1971 and funded through an assessment of

estimated construction costs collected statewide from building permits.

The areas were selected on the basis of population density, locations of

buildings already instrumented, and the probability for potentially damaging earthquakes. Buildings to be instrumented will be of typical

construction, simple in framing and design, and of various heights with the

instrumentation of low-rise buildings emphasized. Remote recording instrumentation, consisting of single or multiaxial accelerometers connected via

data cable to a central recorder, will be installed in each building. The

accelerometers will be placed on the lowest level, at the roof level, and,

in many cases, at one or more intermediate levels. The instrumentation

will be situated st as to separately record both translational and

torsional response.

On the basis of current projected revenues, and instrument procurement, installation and maintenance expenses, it is estimated that as many

as 400 buildings may be instrumented under the State program.

Key Words: Buildings, Instruments, Strong Motion

III- 26
INTRODUCTION

In 1971 the State of California established a Strong-Motion Instrumentation Program

to assure the development of a scientifically sound distribution of strong-motion instruments throughout the State. The California Division of Mines and Geology (CDMG) operates

the program with the advice of an Advisory Board. Funding is provided by a .007 percent

{7(^ per $1000) assessment of estimated construction costs collected statewide from

building permits.

During the first two years the program was in effect (1972-73), the instrumentation

of ground or "freefield" sites was emphasized and all available funding was used to purchase and install strong- motion accelerographs for that purpose. In 1974, with the

"freefield" instrumentation phase of the program well established, a large segment of the

available funding was channeled for use in instrumenting buildings, the second priority

of the program. At the request of the CDMG and its Advisory Board, a set of detailed

guidelines for selecting and instrumenting buildings was developed by a special ad-hoc

committee. Those guidelines, their background and subsequent implementation, are the

subject of this paper. Similar guidelines, not discussed herein, have also been developed

for a third phase of the program in which dams throughout the State are being instrumented.

The basic objectives for the building instrumentation phase of the State program

were adopted by the Advisory Board's Site Selection Committee in mid-1973 as follows: to

place a high priority on instrumenting buildings in Zone III of the Preliminary Map of

the Maximum Expectable Earthquake Intensity in California, figure 1 (Alfors, Burnett, and

Gay, 1973), and a lower priority on instrumenting buildings in Zone II; to place the

highest priority on instrumenting buildings located within five miles of the major faults

along which there is significant activity; to seek the assistance of the Structural

Engineers Association of California in selecting buildings to be instrimiented under the

program; to select representative types and heights of buildings; to instrument many

buildings moderately rather than a few buildings extensively; and to use remote recording

accelerograph systems with accelerometers located so as best to record both translational

and torsional response of each building. With these objectives and at the request of the

I Site Selection Committee, the specially appointed ad-hoc committee (the Subcommittee on

j Instrumentation for Structures) developed a series of guidelines defining where, which

1 tjrpes, and how buildings should be instrumented under the program. Those guidelines,

discussed in detail below, are now being implemented.

III-27
GEOGRAPHIC DISTRIBUTION OF INSTRUMENTED BUILDINGS

In adopting its objectives for the building instrumentation program, the Site

Selection Committee stipulated that areas within Zone III (zone of maximum expected

intensity) of the Preliminary Map of Maximum Expectable Earthquake Intensity in California,

and especially areas within five miles of the major active fault zones, should be given

highest priority for the instrumentation of buildings. In light of these stipulations

and considering population densities, locations of buildings in which strong-motion

instrumentation had already been installed, and a best-educated-guess on the probability

for potentially damaging earthquakes in various areas throughout the State, the Subcommittee on Instrumentation for Structures established a list of 21 areas recommended for

instrumentation and suggested how many buildings should be instrumented in each area

(expressed as a percentage of the total number of buildings to be instrumented). With

the exception of the highly populated San Diego area, all selected areas were either in

or immediately adjacent to Zone III. The area along the Hayward fault between Milpitas

and Geyserville was selected for the largest number of buildings because of the high concentration of buildings adjacent to the fault, high probability for potentially damaging

earthquake activity, and lack of existing instrumented buildings. The area along the San

Andreas fault between Los Gatos and Fort Bragg was chosen for similar reasons. In Los

Angeles, where a large number of mid- and high- rise buildings have already been instrumented under the terms of the city ordinance, the concentration of buildings and probability for potentially damaging earthquake activity were also considered to be high and

the need to instrument low-rise buildings was considered to be great. A slightly lower

number of buildings were recommended along the Calaveras fault system as it extends from

San Jose to Napa, along the San Jacinto fault from Cajon Pass to Hemet and from Hemet to

El Centre, and along the San Andreas fault from Cajon Pass to Calipatria. These areas

were considered to have a high probability for potentially damaging earthquake activity,

moderately dense populations, and few instrumented buildings. In the remaining selected

areas, a relatively small number of buildings were recommended because of lower population

densities and postulated lower potential for seismic activity.

Because of the manner in which seismic waves propagate and attenuate in the California

region (high frequencies tend to attenuate more quickly with distance than lower

frequencies, which tend to be more pronounced away from the source of energy release) and

III- 28
the fact that the natural periods of vibration of buildings are approximately proportional

to the number of stories (the higher the building, the longer the fundamental period),

recommendations were made on the distribution of the height of instrumented buildings

relative to the distance from potentially active faults. Those recommendations, shown in

table 1, suggest that instrumented low-rise buildings (one to six stories) should be

within 10 miles (16 km) of the fault of interest, that most of the Instrumented mid-rise

buildings (seven to fifteen stories) should be within a 25-mile (40-km) range, and that

most of the instrumented high-rise buildings (greater than fifteen stories) should be in

the 5- to 25-mile (8- to 40-km) range.

SELECTION OF TYPES AND HEIGHTS OF BUILDINGS

In its interpretation of the legislation creating the State program, the Advisory

Board indicated that one of the primary objectives should be to provide data on which to

base improvements in engineering design practice. In light of this objective and the

general stipulation that representative buildings throughout the State should be instrumented, it was recommended that: all major building types, construction techniques, and

materials should be equitably represented; each instrumented building should be relatively simple in framing and design so that the response can be readily interpreted; and

the instrumentation of low-rise buildings should be emphasized.

In regard to building type, an attempt was made to ascertain an equitable distribution of the types of buildings shown in table 2. No recommendations were made with

regard to date of design or construction although it was believed that the instrumentation

of buildings designed since the Long Beach earthquake of 1933 should be emphasized.

With respect to the percentage distribution of instrumented building heights, it was

suggested that 28% of the instrumented buildings should be in the one to two-story range,

32% in the three to six-story range, 26% in the seven to fifteen-story range, and 14%

greater than fifteen stories (table 1). This distribution reflects the attitude that

low-rise buildings (one to six stories) should make up 60% of the total number of

buildings instrumented under the State program. The emphasis on low—rise buildings stems

from the fact that they are vastly more numerous throughout the State than their high-rise

counterparts, and that they have historically been more hazardous when subjected to strong

ground shaking (all known deaths in the 1971 San Fernando earthquake, for example,

III-29
occurred in low-rise buildings). Furthermore, relatively few low-rise buildings had been

instrumented prior to the development of the State program.

SELECTION AND PLACEMENT OF INSTRUMENTS IN BUILDINGS

In general, it was recommended that all buildings instrumented under the State

program should be instrumented using remote recording instrumentation, consisting of

single or multiaxial accelerometers connected via data cable to a central recorder (s),

and that the accelerometers should be installed at locations prescribed by the guidelines

for instrumenting buildings developed by the author and R. B. Matthiesen (Rojahn and

Matthiesen, 1975).

Remote recording systems were recommended rather than triaxial optical-mechanical

self-contained accelerographs (triaxial accelerographs are presently required by the city

of Los Angeles and other municipalities that adopted similar ordinances) because remote

recording systems give greater flexibility for accelerometer placement, space requirements

are minimized, and the recorder can be centrally located for easy maintenance and record

retrieval. Furthermore, triaxial systems like those presently required by the city of

Los Angeles do not provide enough data to isolate translational and torsional response, a

capability that forced-vibration tests (Goebler, 1969; Hart, DiJulio and Lew, 1974;

Jennings, Matthiesen and Hoerner, 1972) as well as analyses of records from the 1971 San

Fernando earthquake (Blume and Associates, 1973; Gates, 1973) indicate is vital even in

highly symmetrical buildings.

As a minimum, accelerometers should be placed on the lowest level and at the main

roof level. On the lowest level, it is recommended (as a minimum) that three orthogonal

accelerometers (two horizontal and one vertical) be attached firmly to the foundation or

floor near the center of plan with the horizontal accelerometers oriented parallel to the

transverse and longitudinal axes of the building. If the foundation conditions are such

that differential horizontal motion may occur (Yamahara, 1970), one or more additional

horizontal accelerometers are recommended. In a building that is large and relatively

square in plan, two additional accelerometers should be positioned along and parallel to

two adjacent outside walls, whereas in a building that is very long in comparison with

its width, one additional accelerometer positioned and parallel to one of the outside end

III-30
walls may be sufficient (figure 2a) . If the building has a rigid mat foundation and

rocking motion is expected, two additional vertical accelerometers are recommended. These

should be positioned so that rocking motion can be recorded along any azimuth, i.e., one

vertical accelerometer should be positioned in each of three corners of the building

(figure 2b). In a building that is quite large in plan with significantly varying

foundation conditions, it is recommended that additional triaxial packages be installed

on the different foundation materials.

Instrumentation at the main roof level, as well as at all instrumented intermediate

floors, should consist of an array of remote horizontal accelerometers arranged so as

separately to record both translational and torsional motion. If the roof or instrumented

floor is very stiff and is expected to be rigid in the horizontal plane, only three

horizontal accelerometers are required. A biaxial pair should be located at the predicted

or known center of rigidity so as to record pure translational motion along the transverse and longitudinal axes of the building. The third accelerometer should be positioned

along and parallel to the most distant outside end wall so as to record torsional motion

(figure 3a). If the roof is not expected to be rigid in the horizontal plane, one or more

additional horizontal accelerometers is recommended. The location of each of these will

be dependent upon the expected response of the roof (floor). For example, in the case of

a rectangular-plan exterior shear-wall building with the roof (floor) diaphragm flexible

in the transverse direction and not in the longitudinal direction, one additional accelerometer is recommended. It should be positioned so as to facilitate the interpretation of

relative motion in the transverse direction between the end walls and the center of the

roof (floor) diaphragm (figure 3b). Because the most significant motions in building

response to strong ground shaking are normally in the horizontal direction, vertical

accelerometers are not felt to be as crucial above ground level as horizontal accelerometers. If vertical response is of interest, however, vertical accelerometers sufficient

in number to determine all significant relative motions should be installed. In masonrywall buildings, for example, where ultimate strength is a function of bearing stress

(Mayes and Clough, 1975) , the vertical accelerometers should be aligned vertically on the

wall(s) of interest at various heights throughout the building, including the lowest level.

Likewise, if the vertical response of a floor slab or beam is of interest (in any type of

building) , multiple vertical accelerometers should be installed at the slab edges, or III-31
beam ends, and the mid-span.

The number of intermediate levels at which instrumentation should be installed is a

functon of the structural framing system, number of stories, architectural configuration

and known dynamic characteristics of the building. Unless mode shapes have been predetermined by forced-vibration or in-depth ambient vibration tests and intermediate level

instrumentation is not considered to be necessary, instrumentation should be placed at as

many intermediate floors as is economically feasible because the accuracy with which a

building's response to earthquake motion can be determined is largely proportional to the

number of levels instrumented. As a minimum, it is recommended that at least two intermediate levels be instrumented in buildings having more than six stories above ground and

at least one be instrumented in buildings having three to six stories. The level (s) should

not coincide, if at all possible, with a nodal point of any of the modes of predominant

response (usually some or all of the first four modes) . Close examination of the mode

shapes in figure 4 indicates the most optimal "anti-node" areas for buildings uniform in

plan with height are located at about 25%, 40% and 70% of the above ground building height.

If no other information is available, it is recommended that one or two of these levels

be instrumented. If, however, mode shapes based on a computerized model of the building

are available, such mode shapes should be used to determine the optimal locations. Stiffness discontinuities must also be considered in the process and should be instrumented

whenever their effect on mode shapes is unknown. Instrumentation at such locations could

either serve as or complement the intermediate level instrumentation recommended as

minimal. In some buildings, such as one having a slender tower on a wide base, the discontinuity is obvious, whereas in others, stiffness discontinuities may be revealed only

through a thorough investigation of the structural framing system.

Other recommendations designed to enhance and facilitate the analysis of records

obtained from instrumented buildings are as follows: "free-field" accelerographs should

be located near each instrumented building in order to obtain data on site-structure

resonance effects and soil-structure interaction; the CDMG should collect and archive

plans, specifications, calculations, and pertinent construction and inspection records

for each instrumented building as well as site soil and geology descriptions.

III-32
IMPLEMENTATION OF THE PROGRAM

-The above guidelines were submitted to the Advisory Board in mid-1974 and approved

shortly thereafter with the stipulation that they be regularly updated (perhaps annually)

as more geologic, seismological, engineering analysis and other pertinent information

becomes available. After approval, the CDMG solicited and received from the Structural

Engineers Association of California (SEAOC) a list of 54 buildings to be instrumented

under the first phase of the program. The size and location distribution of this first

set of buildings (figure 5) are not in exact adherence to the criteria established by the

above guidelines, though the basic intent of the guidelines certainly has been met.

Significant deviations from the recommended distributions will be rectified in subsequent

phases of the program.

On the basis of current projected revenues, and instrument procurement, installation

and maintenance expenses, it is estimated that as many as 400 buildings may be instrumented

under the State program (California Division of Mines & Geology, unpublished report to

the California Legislature). The installation phase of the program is expected to be

completed in the year 2035, the time at which program revenues are expected to be sufficient

only for covering instrument maintenance (a major cost), personnel (minimal staff), and

data analysis expenses.

ACKNOWLEDGMENTS

The author gratefully acknowledges Dr. R. B. Matthiesen of the U.S. Geological Survey

and the Subcommittee on Instrumentation for Structures —L. Leroy Crandall, Jack F. Meehan,

John 0. Robb, Roland L. Sharpe, William M. Wells, and Edwin G. Zacher—for providing the

support and advice that made these guidelines possible. Thanks also go to the SEAOC StrongMotion Instrumentation Committees, chaired by Kenneth K. Honda and John 0. Robb, who have

assumed responsibility for selecting buildings to be instrumented under the program.

III-33
REFERENCES

Alfors, J. T., Burnett, J. L. and Gay, T. E., Jr., 1973, Urban geology master plan for

California: California Division of Mines and Geology Bulletin 198, p. 20.

Blume, John A. and Associates, 1973, Sheraton-Universal Hotel, Bank of California, Holiday

Inn (Orion Avenue), Holiday Inn (Morengo Street), Bunker Hill Tower: San Fernando,

California Earthquake of February 9, 1971, ERL/NOAA/U.S. Dept. of Commerce, v. 1,

p. 307-444.

Bowkamp, J. G. and Blohm, J. K. , 1966, Dynamic response of a two-story frame structure:

Seismol. Soc. America Bull., v. 56, p. 1289-1303.

Gates, W. E,, 1973, KB Valley Center, Muir Medical Center, Kajima International Building,

Certified Life Building: San Fernando, California Earthquake of February 9, 1971,

ERL/NOAA/U.S. Dept. of Commerce, v. 1, p. 449-574.

Goebler, H. W. , 1969, Three-dimensional modeling and dynamic analysis of the San Diego

Gas and Electric Company Building: California Univ., Los Angeles, M. S. thesis, 72 p.

Hart, G. C, DiJulio, R. M. , Jr. and Lew, M. , 1974, Torsional response and design of highrise buildings: Cincinnati, Ohio, ASCE Structural Engineering Convention Meeting

Preprint, 30 p.

Jennings, P. C, 1969, Spectrxjm techniques for tall buildings: World Conference on Earthquake Engineering, (4th), Santiago, Chile, Proc, p. 61-74.

Jennings, P. C, Matthiesen, R. B. and Hoerner, J. B. , 1972, Forced-vibration of a tall

steel-frame building: Earthquake Engineering and Structural Dynamics, v. 1, p. 107-132.

Mayes, R. L. and Clough, R. W. , 1975, Cyclic shear tests on fixed-ended masonry piers:

New Orleans, Louisana, ASCE National Structural Engineering Convention Meeting

Preprint, 30 p.

Moran, D. F. and Bockemohle, L. W. , 1973, History and philosophy of California earthquake

codes and elements of lateral force design: San Fernando, California Earthquake of

February 9, 1971, ERL/NOAA/U . S . Dept. of Commerce, v. 1, p. 24.

Rojahn, C. and Matthiesen, R. B. , 1975, Guidelines for strong-motion instrumentation of

buildings: presented at the ASCE National Structural Engineering Convention, New

Orleans, Louisiana.

Yamahara, Hiroshi, 1970, Ground motions during earthquakes and the input loss of earthquake

power to an excitation of buildings: Soils and Foundations (The Japanese Society of

Soil Mechanics and Foundation Engineering), v. X, n. 2, 14 p. Ill- 34
Table 1 - Proposed Distribution of Instrumented Buildings

as a Function of Distance from Fault of Interest,

in percent

Number of

^"""-^^.^^ stories

Distance

from fault ^""-^--^^^^^ in miles (km)

1-2 3-6 7-15 >15

0-5 (0-8) 16 16 4 Ol5-10 (9-16) 12 16 12 4

10-25 (16-40) 0+ 0+ 8 8

>25 (>40) 0+ 0+ 2 2

Table 2 - Recommended Types of Buildings to be Instrumented

I. One and two-story buildings.

A. Open frame type - gymnasiums, auditoriums.

B. Continuous frame - school classrooms, offices.

C. Box structures - commercial masonry or concrete wall

structures with flexible diaphragms.

II. Three to six-story buildings.

A. Frame.

B. Shearwall.

C. Combination.

D. Precast structural elements.

III. Buildings over six stories.

A. Frame.

B. Shearwall.

C. Combination.

Ill --35
Figure 1.- Map showing historically active fault zones and preliminary

maximum expectable earthquake Intensity zones (Alfors, Burnett,

and Gay, 1973).

III-36
Suggested additional

accelerometer for recording differential

horizontal foundation^

motion

LEGEND: Horizontal accelerometer

Triaxlal accelerometer

package (minimum

suggested Instrumentation)

Vertical accelerometer

Figure 2a.- Suggested strong-motion Instrunentatlon scheme for recording

differential horizontal foundation motion In a building

wfaoae length la Tarj large In coB|>arl8on with Its width.

Plan of Lowest Level LEGEND: Horizontal accelerometer

Vertical accelerometer

Figure 2b.- Suggested strong-notion Instruaentatlon scheme for recording

rocking motion at lowest level.

111^37
Plan of Upper Floor or Roof

LEGEND: Horizontal accelerometer

Figure 3a.- Suggested strong-motion Instrunentatlon scheme for roof (or

floor) expected to be rigid in the horizontal plane.

Shear wall

Plan of Upper Floor or Roof

LEGEND: Horizontal accelerometer

Figure 3b.- Suggested strong-motion instrustentation scheme for roof (or

floor) where relative diaphragm ntotion Is expected in the

transverse direction.

III-38
O 00 «o ^ CVJ o

•-^

O

O

O 11

III-39
Figure 5.- Locations of buildings recommended for Instrumentation under

the California Strong-Motion Instrumentation Program (1974-

1975).

III-40
CHARACTERISTICS OF UNDERGROUND SEISMIC MOTIONS

AT FOUR SITES AROUND TOKYO BAY

TOSHIO IWASAKI

Chief, Ground Vibration Section

Piiblic Works Research Institute

Ministry of Construction

4-12-52 Anagawa, Chiba, Japan

SUSUMU WAKABAYASHI

Research Engineer

Public Works Research Institute

Ministry of Construction

4-12-52 Anagawa, Chiba, Japan

FUMIO TATSUOKA

Research Engineer

Public Works Research Institute

Ministry of Construction

4-12-52 Anagawa, Chiba, Japan

ABSTRACT

This paper discusses the dynamic behavior of siibsurface soil and rock layers on the

basis of acceleration records triggered during actual small to moderate earthquakes.

Borehole accelerometers are installed at four sites around the Bay of Tokyo. These were

installed in 1970-74, in connection with the Tokyo Bay Loop Highway Project proposed by

the Ministry of Construction.

Important acceleration records were obtained during sixteen moderate earthquakes

(Magnitude - 4.8 - 7.2) which occurred near the area in September, 1970 through February

1975. From distributions of maximum accelerations at the four stations, it seems that

the surface magnification factors (ratios of the surface acceleration to the base acceleration) are large (2.5 to 3.5) at the soft clayey soil site, small (about 1.5) at the rocky

site, and medium (1.5 to 3) at sandy soil sites.

Response spectrum curves from typical acceleration records are shown. Comparison of

the spectral curves from records obtained at three (or four) levels of one station during

an earthquake suggests that frequency characteristics at the several depths are comparatively similar. Also it seems that frequency characteristics of earthquake ground motions

are influenced by seismic conditions (such as magnitudes, epicentral distances, etc) as well as soil conditions at the sites.

Key Words: Acceleration records; dynamic behavior; earthquakes; response spectra, soils.

1. Introduction

At the present there are more than 200 strong motion accelerographs (mostly SMAC type)

installed at the ground surface of various sites throughout Japan, which provide us with

ground motion records during strong earthquakes. The number of stations, however, where

borehole seismometers are installed in the underground is comparatively less. To analyze

the dynamic behavior of the soil layers during earthquakes , the measurements of the earthIII- 41
quake motions at the deeper levels below the ground surface seem very important. This

paper intends to present the results of seismic observation with use of borehole accelerometers installed at four stations aroiind Tokyo Bay.

2. Outline of Observation Stations

The Public Works Research Institute and the Kanto Regional Construction Bureau have

been measuring underground seismic motions at four stations around the Bay of Tokyo (1)

As shown in Fig. 1 the stations are located in Futtsu Cape, Chiba Prefecture; Ukishima

Park; Kawasaki City; Kannonzaki, Yokosuka City; and Ohgishima, Yokohama City. These were

installed in 1970 through 1974, in connection with the Tokyo Bay Loop Highway Project proposed by the Ministry of Construction.

Depths of the borehole accelerometers and soil profiles at the four stations are shown in Figs. 2 and 3 respectively. At the site of Futtsu Cape, where a comparatively

uniform sand layer exists, 3-component borehole accelerometers are set up at three levels,

the surface, -70 m, and -110 m. At Ukishima Park, consisting of reclaimed soils, silts,

clays, and sands, four pickups are installed at the surface, -27 m, -67 m, and -127 m. At Kannonzaki, indicating a rocky layer (siltstone) with a shallow topsoil, three accelerometers are set up at the surface, -80 m. and -120 m. Furthermore, at Ohgishima, a newly

reclaimed island consisting of deep reclaimed sands and stiff clays, four acceleration

transducers are installed at the surface, -15 m, -38m, and -150 m. Each accelerometer

has three components - one vertical and two horizontal. Time dependent acceleration

waves are triggered on oscillograph papers.

3. Acceleration Records

Table 1 indicates a list of the earthquake records obtained during sixteen small and

moderate earthquakes (Magnitude = 4.8 to 7.2) which broke out near the area between 1970

and 1975. In the table the maximum accelerations in the three directions are listed, to- gether with the dates and magnitudes of the earthquakes . In the colvimn of remarks of the

table H denotes the hypocentral depths and D means the ranges of the distance from the epicenter to the stations. An example of the records triggered is shown in Fig. 4, which is

a set of acceleration records obtained at Ukishima Park during the earthquake (M = 4.8) of

September 30, 1970. Most earthquake records including those in Table 1 are digitized,

and their spectral characteristics are obtained (2)

4. Distribution of Maximum Accelerations Relative to Depth

The variation of maximum accelerations with respect to depths and the three components,

is illustrated in Fig. 5 for six typical earthquakes. The normalized distribution of maxi- mum accelerations are obtained by averaging all data in Table 1. In the upper part of the

figure the maximum acceleration at the surface is taken as a unit, and in the lower part

the deepest is taken as a unit. From the figures the following conclusions can be derived,

relative to the distribution of underground accelerations during small and moderate earthquakes . 111^2
(a) The vertical distribution of underground accelerations with respect to depths varies with stations, directions, earthquake properties, etc.

(b) Depths where the accelerations become one half of the surface accelerations at Futtsu Cape, Ukishima Park, Kannonzaki, and Ohgishima are about 70*^, 50^, more than 120^,

and 150TM, respectively.

(c) Accelerations at the deeper layers are close among the values at the four stations.

Accelerations at the surface, however, differ greatly among the four stations. Furthermore,

accelerations at the deeper layer of each station do not change with respect to depths.

(d) Ratios of the surface acceleration to the deepest acceleration are large (2.5 to

3.5) at the soft clayey soil site (Ukishima Park), small (about 1.5) at the rocky site

(kannonzaki), and medium (1.5 to 3) at the sandy soil sites (Futtsu Cape and Ohgishima).

(e) Vertical accelerations are generally smaller (1/3 to 1/2) than the horizontal

acceleration.

5. Characteristics of the Response Spectra for Underground Motions

In order to show the frequency characteristics of underground motions , response spectral

curves for horizontal components of typical records are illustrated in Fig. 7 (3) . The

abscissa denotes the natural period of a single degree of freedom system. The ordinate

indicates the maximum absolute response acceleration of the system with a damping ratio of

2%, when subjected to time dependent accelerations recorded at respective levels of each

station. These spectral curves indicate the following:

(a) It is apparent that frequency characteristics of acceleration records obtained

at one station during an earthquake are very similar among the three or four layers of the

station.

(b) By comparing (A) with (B) of Fig. 7, it ia seen that frequency characteristics at one station considerably differ between records obtained during the two different earthquakes. Fig. 7 (A) shows the spectra at Ukishima Park for a small and near earthquake

(M = 4.8, A = 10 km) , and Fig. 7 (B) shows the spectra at the same station for a moderate

and distant earthquake (M = 7.2, A = 320 km). The characteristics of these two spectra

are very different.

(c) By comparing the spectra at two stations (whose soil conditions are considerably

different, for example. Fig. 7 (B) and (D) during an earthquake, it is seen that the general

characteristics are very close, and that the responses in the range of longer periods are predominant at the soft soil site (Ukishima Park)

.

6. Conclusions

From the analyses of underground seismic acceleration records, obtained during small

and moderate earthquakes with magnitudes of 4.8 to 7.2, the following conclusions can be

drawn.

(a) The distribution of the maximum accelerations, with respect to depths, changes

considerably with the change of soil conditions near the ground surface. Ratios of the

III- 43
surface acceleration to that at the deeper layer (110 to 150^) are about 1.5 at a rocky

ground, 1.5 to 3 at sandy grounds, and 2.5 to 3.5 at a very clayey ground.

(b) Although the acceleration values are smaller at deeper layers, frequency characteristics of underground seismic motions are close to those of the surface motions.

(c) It appears that the characteristics of earthquake ground motions are influenced

by seismic conditions such as magnitudes of earthquakes, epicentral distances, etc., as well as soil conditions at the site.

Acknowledgements

The observation of the underground seismic motions described in this paper was realized under the strong promotion of Dr. Tadayoshi Okubo and Mr. Eiichi Kuribayashi at

the Public Works Research Institute. The authors appreciate their achievements to attain

this project. Acquisition of acceleration records was made by the staff members at the

Kanto Regional Construction Bureau, and digitization and computer analysis were conducted

with an assistance of Mr. Shunichi Horiuchi, former engineer at the Public Works Research

Institute. The authors wish to thank them for their strenuous efforts.

References

1) Toshio Iwasaki, Susumu Wakabayashi, Fumio Tatsuoka, "Seismic Behavior of Subsurface

Ground Layers", Proceedings, 4th Japan Earthquake Engineering Symposium, November, 1975,

Tokyo , Japan

2) Toshio Iwasaki, Susumu Wakabayashi, Shunichi Horiuchi, "Strong-Motion Earthquake

Records at Undergrounds (Vol 1, Vol. 2)", Technical Memorandum of the Public Works Research

Institute No. 1103, March, 1976.

3) Eiichi Kuribayashi, Toshio Iwasaki, Yutaka lida, Katunari Tuji, "Effects of Seismic

and Subsoil Conditions on Earthquake Response Spectra" , 5th World Conference on Earthquake

Engineering, Rome, Italy, June, 1973.

III-ii4.
Table 1 Maximum Accelerations during 16 Moderate Earthquakes

Since Sept. , 1970 (in gals)

EQ

Da t

>• Ma gn i t u d e Ep i center

- ompo- Fu t t su C, P e U k i s h i m a Pa rk Kan ^ 0 n z a k i Ohgi si i m a Rem arks

n e n t 0 m -70m - UOm Om - 2 7ni - 6 7ni - 127m 0 m -80m -120m 0 m - 1 5m -38m -150m

1

9.30.1 9 70 M=4 8 Eas t ern Kanagawa-ken

NSEW

UD

28 2 29 4

8,4

1 6 0

9.8

4.2

7 8

1 12

6.4

1 1 8 178

7.8

1 0 7

1 1,2

3.8

5 5

6.0

1.8

4,3

3,4

2,0

H=4 0km 0~30km

Mouth of Tama River

723.1 97 I M= 5.3 East ern

Yamanashi -ker

NSEW

UD

5,6 86

2 2

2.5

2,6

1,2

2.7

4 3

1.3

1 6.8

1 2 3

5.7

1 3.3

1 1833

5.6

5.0

3.0

6 4

3,5

2,6

1 2-4

6-7

5.4

5 3

5.5

4.1

4.9

8.5

3.4

H = 4 0 km ^=i60~80km

3

2 2 9.1 9 7 2 M= 7 2 E OFF

Hach i jo

NSEW

UD

46 1

5 1 6

1 7 2

3 2.3

2 3.6

1 0 9

1 8.6

21 6

1 3 3

1 0.3

8.6

6.0

1 3,1

1 4,9

1 1 3

1 2,2

1 4,6

1 04

H= 7 0 km

^!=;300km

<!

1 0,1 8.1 97 2 M- 5 1 Nor t he rn

(' h i ba-ken

NSEW

UD

6 4

8 0 28

3,5

3,3

1 3

2,1

12

6.5

1 2.0

4.2

5.8

9,6

2 8

3-7

2 1

16

3,1

3,6

1 1

6 1

1 7 2 40

4 6

3,0

2 1

3 1 35

1 8

H= 8 0 km ^^50-70 km

5

1 1 6,1 972 M- 5,1 SW

I bar ak i -ken

NSEW

UD

2,8

32

18

16

1 4

0,9

0,8

1409

3.3

8.1

2.2

3,2

6,3

1,4

2 0

2-5

0.9

1-8

16

0.9

H= 4 0 km

.£^^60-90 km

6

1 2 4.1 9 7 2 M= 7 2 E Off Hach I J o

NSEW

UD

20.0

15.6

4.1

1 4.6

1 1 3

3.6

7.4

7 1

3.0

6.4

4 7

3.0

6.5

7 3

10,1

1 0 6

1 0,4

1 0 7 H=5 0km

^*=i300 - 320kir

7

1 2 8,1 9 72 M = 4 8 Northern

Tokyo Bay

NSEW

UD

1 2 3

7 0

4,0

7 2 36

18

3.2

8.5

2.1

1 0.1

1 3.4

3.7

9-3

1 0-4 24

5.3

4,2

16

4.4

3.3

1.5

8.3

1 6.8

6-1

5,0

5,9

2,5

4,2

5,3

2,5

H-9 0km

^^20~40 km

8

3.2 7, 1 9 7 3 M= 4 9 Tokyo Bay

NSEW

UD

4 75

42 1

7.1

3 0,7

1 2 3

6 5

7 4

1 3 7

5 8

18.6

3 15

8.5

2 5.9

2 6.7

9,3

16,8

5.4

H= 6 0 km

^=. 20- 40 km

9

1 2,2 2.1 9 7 3 M = 5,0 Southern

C h i ba-ken

NSEW

UD

1 0 5

7 6

3.4

8,5

4,3

1,8

2 6

7 8

2.3

1 3.0

1 8.4

6.3

6-4

7 3

3-0

60

7 4

3.9

H= 7 0 km ^^50^ 55 km

1 0

5 9.1 9 7 4 M - 6 9 S Coast of

Izu Pen,

NSEW

UD

38 0

4 0.6

13.9

22,3

2 6.4

6,8

2 8.7

1 3 7

7 1

3 0,0

3 77

1 2,5

14-7 177

6,9

1 15

22 4

5.5

H- 1 U km

10^1 2U km

Ot I Izu

Pen. Eq.

1 1

7. 8.1 9 7 4 M- 63

Ol f Ibarak

-ken

NSEW

UD

4.1

3.3

2.3

2,4 19

1 7

0.9

2-0

1,3

4.2

3.3

1.5

2-0

0.7

1 0

2,9 13

10

1.0 17

1 0

3,5

3-6 20

1,9 19

1 9

1 5 20

1.2

H— 4 0 km

km

^5^160-190

1 2

8. 4.1974 M= 5,8 SW

I bar ak i -ken

NSEW

UD

1 3.6

1 2.9

5.8

1 0.6

7.7

3.1

6.5

5.5

2.0

5.1

4.4 19

1 2.5

1 5.0

4.5

6.5

7.1

4.5

4.9

6.5

3.2

1 2.1

8.2

1 0.7

71

6.0

4.7

8,5

1 0.8

6.2

4.8

3.4

H- 5 0 km

^:^60~90km

1 3

9 2 7,1 9 7 4 M= 6,5 SE 01 f B 0 s 0 pen

NSEW

UD

1 18

1 0.2

6.9

7,0

7 1

4.0

5,0

5 9

3,0

7.0

5.5

3.1

59

5 0

2.0

3.7

3,3

1.3

3.1

1 9

1.3

4.7

5.9

4.5

5.0

3.9

2,5

5 0

5.5

2.3

2.9

3.2

1 6

H= 4 0 km

2 40 km

1 4

1 1 1 6,1 97 4 M= 6 1 Ne a r Cho sh i C h i ba-ken

NSEW

UD

4.8

4.8

1.7

4.0

2.6

1-2

3.0

2,2

1,6

2.8

1 7 12

5.4

3.6

3-1

3,4

4,2

2 8

2.9

3.3

2.9

18

1914

H= 4 0 km

140 km

1 5

1 1 30,1 97 4 M- 7,5- 7.7 5 Near Tor i shima

NSEW

UD

1 1 6

8.5

4.7

8.5

5 8

3.1

4,0

4 5

2,8

6.2

5.3

2.3

7-0

5-0 17

2 5

2 9 13

2.8

2 2

11

6-2

6-2

3,7

3,2

3,4

2,2

5.6

6.8

2 4

4.2

3.6

1 5

H=-3 50km

^!=! 550 km

1 6

2. 8,19 75 M= 5,4 Northern

Chiba-ken

NSEW

UD

14.2

1 8.9

5.6

9-3

1 6-6

3 1

4,8

7 5

2,6

4.2

5-5 20

H-50km

^==5 5 0 km

III-45
O Ichikawa

Ichiharo

Chlba Pref,

Kannonzakill97l)

Fig. 1 Stations of Borehole Accelerometer Installation

om

i

-70 m

-IIO"^

Fjtt'ij Cape

-127m

Ukishima Park

om

;-80m

Kannonzaki

38"^

Ohgishimo

Fig. 2 Depths of Acceleration Pickups

III-46
Depth (m) IT'

10

20

30

40

50 -

60 -

70

80 -

90 -

100-

110

Soil Profile

Coarse Sond

with Shells

Medium Sond

with Shells

Silty Fine Sand

Silty Fine

Sand

Fine Sond

Medium Sand

Legend *- indicotes depth

of borehole accelerometer

instolled.

Futtsu Cape

Depth (m)

Soil Profile N-Volue (No /cm) 0

Recloimed Soil

- 2 /30

, 5 /30

- 8 /30

10

8

1 4 Sandy Silt 2 /3C 3/30

20

23

Cloyey Silt

3 /30 3/50

30

Silty Cloy 4/52

4 /30

40

34-

4?

49-

Silt

8 /30 3/30

Silty Fine Sand - 5/30 50

60

52

- 61 * 68 -

Silly Sond

Sond

5 /30

- 90/22

100/26 Muddy Sond - 75/30

70 Sond

- 54/30

- 100/22 80 - 79

90 i_ 89-

91

98

Grovel

_Hymjcfi Fine Sond

Qiitu jllLy Fino riMc ^nnH ouiiu

- 87/30

- 100/17 100

104 Sondy Silt

- 106/26

110

108

113

122

133

Jilty Fine Sond Sand

- 23/30

120

Silty Fine Sond

- 58/30

130 Sand

WO

146 -

Silty Sond

150

149 Fine Sond

" 152 Silty Fine Sond

Depth (m)

32

37.5

57

60.5

71.5-

76

Soil Profile ^op Soil

Alternotive Loyers of Sandstone

and Siltstone

Sandstone

Siltstone

Sandstone

Siltstone

Sandstone

Siltstone

Depth (m) Soil Profile

"z^ond

Gravel Sond

Sand

Grovel Sand

Alternative Layers of Sond and Silt

Clay or Cloyslone

Stiff Cloy

Gfovel Silt Stiff Silty Cloy Grovel Sond

Stiff Cloy

Siltstone

N-Volue (NoVcm)

2/30

3/30

2 /30

5 / 30

7/30 8/30

40/30

25/30 45/30

47/30

50/ 0 50/20

50/25

Very Stiff Siltstone

S Wove

Velocity

Im/secl

160

220

270

^130

2 180

4

310

2 270

I 460

14 370

)2

500

Ukishima Pork Kannonzaki Ohgishimo

Fig. 3 Soil Profiles at Four Observation Stations

III-47
III-48
EQ NO I Sept 30. 1970 ( M -- 4.8 ) EQ N0.3 Feb. 29. 1972 (M = 7.2)

Futtsu (i=30km) Ukishima ( A - lOkm ) Konnonzaki (fl; 300km) Futlsu (/i=,300km)

Fig. 5 Vertical Distribution of Maximum Accelerations

during Six Earthquakes

III-49
90D)jnS dm )D U0IJDJ9I90DV iWf] ( \/

(sadasQ aqj (o uouDJ3|aDD\/ jiuf] (8)

1 i I I 1 I I L I I 1 I I I I 1 I o ooooooo EO o oooopoo £o

III-50
(A) Ukishima Park, EQ. NO, 1

Fig. 7 Response Spectral Curves

III-51
(B) Ukishima Park, EQ. NO. 6

Fig. 7 Response Spectral Curves

III-52
Natural Period, T

(C) Kannonzaki, EQ. NO. 9

Fig. 7 Response Spectral Curves

III-53

gals

Natural Period, T

(E) Futtsu Cape, EQ. NO. 3

Fig. 7 Response Spectral Curves

III-55
(F) Ohgishima, EQ. NO, 12

Fig. 7 Response Spectral Curves

III-56
RELATIONSHIP BETWEEN EARTHQUAKE

DAMAGE OF EXISTING WOODEN HOUSES

AND SEISMIC INTENSITIES

EIICHI KURIBAYASHI

Chief, Earthquake Engineering Section

Public Works Research Institute

Takayiiki Tazaku

Research Member

TAKAYUKI HADATE

Research Member

ABSTRACT

This report discusses a quantitative relationship between a ratio of earthquake damage

of existing wooden houses and seismic intensities.

The ratio of earthquake damage of the houses is useful not only for understanding

house and building design criteria but for presuming the damage ratio of other structures

such as bridges, roads, public utilities, etc. Using statistics of disaster documents on short distant earthquakes; Fukui Earthquake

(M = 7.3, 1948), Isuhanto-Oki Earthquake (M = 6.8, 1974) and Ebino Earthquake (M = 6.1, 1968)

the relationship between the ratio of razed houses and epicentral distances or magnitudes

of earthquakes was analyzed. The equivalent ratio of razed houses in earthquakes (Dl) and

the original ratio of razed houses (D2) are defined as;

^

_ Nunber of Razed Houses 0.5 :< Number of i-ialf Razed H ouses _ ^ 1 Total Number of Existing Houses

Number of Razed Houses

X 100 (/i) Total Number of Existing Houses

Conclusively D^ in the area of diluvium or tertiary (given as Dq^j) and in the area

of alluvium (given as D^^^) can be tentatively expressed as follows:

Dii = Kj X

10*^1

where and K^^ are respective constant values, and K^^ are functions of epicentral

distances and earthquake magnitudes.

Moreover D^ and D^ correlate with each other and ^2x1 area of alluvium)

is greater than D^^ (in the area of diluvium or tertiary)

.

Key Words: Classification, Damage, Seismic, Intensity, Wooden Houses.

IV1
1. Introduction (1, 2) This report discusses relationships among damage of houses, loss of human life,

epicentral distances and earthquake magnitudes, using documents obtained in earthquake

disasters relative to the Fukui Earthquake (M = 7.3, 1948), Isuhanto-Oki Earthquake (M =

6.8, 1974) and Ebino Earthquake (M = 6.1, 1968).

2. Classifications of Damage Due to Earthquake (2) Damage due to earthquakes can roughly be classified from the disaster documents, as shown in Fig. 1.

3. Analyses on Distributions of Earthquake Damage Ratios

In order to analyze earthquake damage of existing wooden houses , earthquake damage

ratios are defined as follows,

^ _ number of razed houses + 0.5 x number of half razed houses

1 total number of existing houses

^ _ number of razed houses

2 total number of existing houses

^ _ number of houses burnt

3 total number of existing houses

^ _ number of houses lost by floods

4 total number of existing houses

^

_ number of houses inundated

5 total number of existing houses

where

equivalent ratios of razed houses due to earthquakes,

original ratios of razed houses due to earthquakes,

ratios of houses burnt due to earthquakes,

ratios of houses lost by floods due to earthquakes,

ratios of houses inundated due to earthquakes.

Similarly, earthquake damage ratio for human lives are defined as follows.

number of death by razed houses + number of injured by razed houses

D

6 population respected

number of death by razed houses

°7 = population respected

^ _ number of heads burnt to death + number of injured by fires

8 population respected

(6)

(7)

(8)

IVnumber of heads burnt to death

9 population respected

number of heads drowned

I =

10 population respected

where

^6 ' ratios of death and injured due to earthquake. ^

'• ratios of death by razed houses.

°8 '• ratios of heads burnt and injured by fires.

ratios of heads burnt to death.

°io

'• ratios of heads drowned

.

b) Earthquakes Employed

In this analyses, three earthquakes, so called short distant earthquakes are employed as follows,

(1) 1948, June, 28 Pukui Earthquake (5, 6); magnitude M = 7.3, focal depth H =

20 km-

(2) 1968, February, 21 Ebino Earthquake (7); magnitude M = 6.1, focal depth H =

O km.

(3) 1974, May, 9 Isuhanto-oki Earthquake (8); magnitude M = 6.8, focal depth H =

20 km.

c) Distribution of the Ratios of Razed Houses

The relationships among the ratios of the razed houses , the epicentral distances

and magnitudes of the subgrounds are classified in two types;

type 1; deluvium and Tertiary

type 2; alluvium (9) According to preliminary analyses, relationships between the equivalent ratios and

epicentral distances can approximately be expressed by exponential functions as shown in

Fig. 2 (5, 10). Since the above, the damage ratio is assumed as follows,

Di ^-a . 1 o-"'-^ ( logDi . j4-|oga ) , (11)

where a and 3 are coefficients statistically obtained from topographical distributions

of actual damage ratio and A is the epicentral distance.

a and 3 with respect to the earthquakes and the types of subground properties , are estimated as shown in Fig. 3. Substituting the values of a and 3 into Eq. (11) , with

respect to the type of siabground properties can be expressed as follows

,

IV3
/ . M ^ 1 V 1 rt0625M4-0.02 54M-0.2 72 4 , (12)

Dil ( ^ • M-) ^ 164"

D1JI(J.M)= _L_ X 1 0 05 76M+0.02 6 .M- 02 4 8i ,(13)

where D^^; is the equivalent ratio in respect of type 1 of subground properties and

D^^^; is the equivalent ratio in respect of type 2. Equations (12) and (13) are applicable to following cases,

6.0 g M < 7.0 and 5.0 KM 4 A ^ 40.0 KM

where Fig. 4 shows various relationships.

d) Correlations Between the Equivalent Ratios of Razed Houses and Original Ratios

of the Razed Houses

In Fukui Earthquake, correlation between the original ratios and the types of

subgro\inds, are shown in Fig. 5. Further details on the sxabground type 1 are shown in Fig.

6, where the numbers with circles represent the following;

2 and 3 4 and 5

6 and 7

mean values of the ratio,

limits for confidence levels with five percent,

limits for confidence levels with two percent and

limits for confidence levels with one percent.

As shown in Fig. 6 four out of fourteen data are scattered beyond the domain of the one percent interval because of the rough classification of the types of subgrounds and

nonuniformity of capable loads of the houses.

Regression coefficients of the ratios (gradients) increase with increases of earthquake magnitude, as shown in Fig. 7.

4. Losses of Hxoman Lives (5, 12)

H. Kawasumi relationships between damage of houses and losses of human lives are used

j and are given as follows, (11)

i

i number of razed houses ; N 0. 2 X 1 0^'" ^ (14)

number of houses burnt ; H i 0" 5 |\j2 , (15)

number of death ; ,^ ^ q~2 \i i 2 5

i qii-S |{0.6 2 5

^ (16)

where M means earthquake magnitudes.

IV-4

I

i
Eq. (16) is available only to estimate total numbers of death and not the ratio of

death. With respect to the type 2 subground properties in the Fukui Earthquake, relationships between the equivalent damage ratio and the ratio of death can be expressed as follows

,

the ratio of death and injured ; 2 >. 1 0" I'lir' » ^^^^

the ratio of death ; 3 ^ 1 0 ' Dm- , (18)

where D^^^ means the equivalent ratio of razed houses.

The above indicated that death and injury increases progressively with increase of

the damage ratio.

5. Conclusions

a) According to the statistical analyses using documents concerning earthquake disasters, the damage ratios of wooden houses can be expressed as functions with respect to

the types of subground properties, epicentral distances and magnitudes of earthquakes.

The trend between the ratio and the components is as follows;

subground properties ; Relatively small damage ratios for subground type 1 (stiff and sound) and

relatively large damage ratios for subground type 2 (soft)

.

epicentral distances ; Larger damage ratios for shorter distances , and magnitudes of earthquakes

(Richter scale)

;

Larger damage ratios for larger magnitudes

b) Although the value of the functions of the ratios near epicenters statistically

exceeds 100 percent in some cases it seems that the capacity of the structural material

is smaller than the actual vibratory loads induced in the earthquakes concerned.

c) Under a constant value of equivalent damage ratio, the trend of the original

damage ratio is as follows

,

subground properties ; Relatively small damage ratios for stibground type 1 and relatively large

damage ratios for subground type 2 in Fukui Earthquake , and magnitudes of earthquake

Larger damage ratios for larger magnitudes when comparing Fukui and Ebino Earth

quakes

.

References

!• E. Kuribayashi et aJ ; Quantitative Relations Between Magnitudes of Earthquake

and Damage (in Japanese), proc, Kanto District, Japan Society of Civil Engineers, Tokyo,

January 1976.

2- K. Kuribayashi et a]; Distributions of Damage Due to Recent Short Distant Earthquakes (in Japanese) , Technical Memorandum of Piiblic Works Research Institute No. 1106, P.W.R.I. Ministry of Construction, March 1976.

IV5
3- N. Koiwai and K. Miyawaki, Determination of Earthquake Hazards in Densely Populated

Areas and Preventive Measures to be Considered in Urban Planning, Proc. of 5th Joint Meeting U.S. and Japan Panel on wind and Seismic Effects, UJNR, Tokyo, May 14-16, 1973.

4- Earthquakes Experienced and Their Disasters in Nagoya City (in Japan), Disaster

Prevention Council, Nagoya City, June 1974.

5- Fukui Earthquake Disaster History (in Japanese), Fukui prefecture, June 28, 1949.

6- Report on Hokuriku Earthquake Disaster (in Japanese) , Report of Public Works

Research Institute, P.W.R.I., Ministry of Construction, March 1949.

7- Memorandum of Ebino Earthquake (in Japanese), Miyazaki prefecture, September 30,

1969.

8- Preliminary Report of 1974 Isuhanto-Oki Earthquake (in Japanese) , Building Research

Institute, Ministry of Construction, May 23, 1974.

9- Classified maps of subground (in Japanese) , Development Bureau, Economy and

Planning Agency, Prime Minister's Office, 1974.

10. Report of Statistic Survey on Housing (in Japanese) , Japan Statistics Agency,

1948, 1968.

11; H. Kawazxami, Intensity and Magnitude of Shallow Earthquakes, Bureau Central

Seismology, International Series A, Tran., Sci. , 19, (1954).

12. Report of Census 1970 (in Japanese) , Japan Statistics Agency.

IVEarthquake

Damage

I Direct Damage

I— Loss of Life — Material Damage

Indirect Damage

Fig. 1. Classifications of earthquake damage

100

30

Equivalent ratio of razed houses (D]) in %

10

—^-^ j-tr^—\ — -A

—\

SB

L ^

\ —\

\

v o

A Subground type O Subground type e— V

V —A —\ V 0— \ — A—-ek \

e

— —

V

—V^-Ae-

\

\

t^

ee

0 0<A

<

\

\

• 10 20 30

Epicentral distances (J) in km

Fig. 2. Relationships between Di and J in Fukui Earthquake

Subground a a

Type 1 O A ai= 1 O''"^"/! 6 4 ^1= 0.025M-0.272

Type 2 e A au= 1 0""6M/^

2 /?n= 0.026M-0.24 8

Fig. 3. Relationships between M and coefficients a or /S"

IV7
note) Dij ; Equivalent ratio in respect of type 1 in %

Dijj; Equivalent ratio in respect of type 2 in % A ; Epi central distance in km

i1 ; Earthquake magnitudes

Fig. 4. Equivalent damage ratios by Eq.(12) and (13)

(7o) 100

I Ql \ I I I I \ I I f 20 60 100 (7o)

j Equivalent ratio of razed houses I

Fig. 5. Correlations between D] and ' in respect of subground types '

I

i IV- 8

'

i

I*
(%) 60

Hosorogi

0 20 40 60 (%) Equivalent ratio of razed houses (Di)

Fig. 6. Correlations between D-] and D2 in respect of subground tyne 1 in Fukui Earthquake

(7o) 100 I 1

Original ratio of razed houses (D2)

60

10

Fukui Earthquake

iobe Earthquake 4)

20 60 100 (7o) Equivalent ratio of razed houses (DO

Fig. 7. Correlations between D] and D2 in respect of subground type 1 in Fukui Earthquake

IV9
IV- 10
O ratios of houses burnt less than 1 % A ratios of houses burnt greater than or equal to 1 %

( ) ratios of houses burnt

Ratios of death and injured

40

30

20

10

7

6

4

3

2

0

- Nishifuzishima

o

Maruok*

L 170%!

-

A (24 72%!

^15,/^,

\

10''« Of i.^'+x^

Fukui — 0 4(1352 %) ^^-'''''^ 0

-

o

o

o

\ Matsuoka ^^^-"^^ 0 0 \ aI5£%)^^- ^^^^^ ° 0 °o

0 ^ ^ n 0 ^ ^

0 ° 0 ° ° °

Ratios of death

-

A 074

Ha rue (5%)

-

Maruoka

i (70°U

Kanazu Morita

A (24 72%) A (2,43%) Matsuoka

A (5 6%)

v3xlO"\0?

\ ^ 0 \ ^^^^ ° \ 0 \

°

0 ^^^^Fukui ° 0 § ^^""^ (13 52%) ° „ ° ^— °

_o__^ O 0 1

oo

0 ^ 00

0 ° 00 0

1 1 \ 1 1 -

0 10 20 30 40 50 60 70 80 90 (00 %

Equivalent ratios of razed houses

Fig. b. Relationships between D] and losses of lives

IV- 11
Fig. c Distribution of earthquake damage of Wooden houses

in Fukui Earthquake

IV- 12
IV- 13
g. d Distribution of earthquake damage of Wooden houses

in Ebino Earthquake

Kumamoto Prefecture

v_y Sakamoto

Ubatake

Machubar^^

Mivazaki Prefecture

Higashi-kamakiCa o Nishi-kamakita ;

_

' jf^XT^ . Nishinoshlta--^^

, /" 5p ^ae-aa thubara

if' -"^•Ruda ^•Naklliiia 0 Saigo

Daimyozi

Mlhamisyomyozi;

Nagayaina*V^V^

Kagoshima

Nakauchitate Shimoshlmauchi

\ /'-#.« /#,,ON^kaura "Chl

^. Halzuka(^^

Mlnamlokamatsu Shlmoura ^

- (^Yanagithuru

Hannyaziushiro

Harag Yamasb'ita^'^^ /^Harag--^- , i^/^Thuru / ^^Hara t jKusunobe

* VnFumqaj^ guchl-tnae ^^"^^ni^rsfub^^*'^d\r)^akaguchl \

Yothueda mac^Ji^^shijno-Nakathugawa y tclicli^ V ^tKami-kawazoe tfisanAjP^Kaklgl

ZlnzenLQlchlhara^IkeOa Korlnoha^M ,1^ ^^^^jj^.^aa

„athu*maV»\ USl""°

Nakanagayama Yadokoro

Yadate

iguchl-ushlro'-^A) . r-^;z-^.." V aguchi-nishi^O^ ^

V

/

Sendai River

IV- 14
IV- 15
IV- 16
IV- 17
A METHOD FOR CALCULATING NONLINEAR SEISMIC RESPONSE

IN TWO DIMENSIONS

WILLIAM B. JOYNER

Office of Earthquake Studies

Branch of Earthquake Hazards

U.S.G.S

Menlo Park, California

ABSTRACT

A method is presented for calculating the seismic response of two-dimensional configurations of soil testing on bedrock. The method, which is based on a rheological model

suggested by Iwan, takes account of the nonlinear, hysteretic behavior of soil and offers

considerable flexibility for incorporating laboratory data on soil behavior. An approximate treatment of the boundary conditions is employed which permits energy to be radiated

into the underlying medium. Examples are shown to illustrate the method.

Key Words: Bedrock, Modeling; Rheological, Two-dimensions, Seismic Response.

1. Introduction

Numerical calculations of the effect of local site conditions on strong earthquake

ground shaking play a major role in the design of important structures for earthquake

resistance. Most such calculations at present employ one or both of two major simplifications. The first simplification is the use of one-dimensional models. In such models

conditions at the site are represented by a system of horizontal soil layers resting on

a semi- infinite elastic medium. The source of earthquake energy is represented by a vertically-incident shear wave in the underlying medium. Serious reservations have been expressed about such models (Hudson, 1972; Newmark and others, 1972), and their range of applicability is a subject of considerable dispute.

The second major simplification has to do with the nonlinearity of the stress-strain

relationships of soils. The one-dimensional problem was solved for the case of linear

visco-elastic layers by Kanai (1952) some years ago. When we are dealing with input motion

sufficiently intense to cause severe damage to structures, however, we cannot assume linear

behavior over the entire range of strain. To do so would imply stresses many times greater

than the strength of typical materials as measured in the laboratory. To circumvent this

difficulty, the method in common use is what is called the "equivalent linear method"

(Idriss and Seed, 1968; Schnabel, Seed, and Lysmer, 1972) . It is based on the assumption

that the response of the system can be approximated by the response of a linear model whose

properties are chosen in accord with the average strain that occurs in the model during

excitation. The average strain level is found by iterative calculation. The equivalent

linear method has been applied to two-dimensional as well as one-dimensional models

(Idriss and Seed, 1974).

IV- 18
Recently, a number of workers have developed truly nonlinear methods of solving the

one-dimensional ground response problem (Streeter, Wylie, and Richart, 1974; Constantopolous,

1973; Faccioli, Santoyo V., and Leon T. , 1973; Papastamatiou, written communication; Chen

and Joyner, 1974; Joyner and Chen, 1975) . Streeter, Wylie and Richart (1974) have extended

their method to two dimensions by replacing the continuum with a latticework of one-dimensional elements.

The method of Chem and Joyner (1974) is based on a rheological model proposed by

Iwan (1967). In his paper Iwan extended the one-dimensional model to three dimensions.

Application of the extended Iwan model leads to a rigorous and efficient method for computing the seismic response of two-dimensional soil configurations. The present paper describes that method.

Two-dimensional computations have potential application for determining free-field

site response in cases where one-dimensional models are inadequate. In addition, such

computations have a special application of particular importance in evaluating the response

of earth dams.

2. Basic Computational Scheme

The computations are done step by step in time over a grid of discrete points. The

displacements of grid points are considered small, and the movement of material relative

to a fixed frame of reference is neglected. The soil mass is divided into prismatic

elements of infinite length, either square or right- triangular in cross section. The soil

is iinderlain by a semi-infinite elastic medium representing bedrock. The velocities of

both P and S waves in the underlying medium are assumed large compared to the velocities

in the soil. The sources of seismic disturbance lie in the semi-infinite medium. Figure

1 is a schematic illustration of a possible configuration.

For square elements the mass is divided equally among the adjoining nodes. In some

of the examples presented later, the mass of triangular elements is divided equally

between the two nodes on the horizontal leg of the triangle. In others the mass is

divided equally among all three adjoining nodes. The latter procedure is preferable, but

the difference should not be significant.

The behavior of the system is described in terms of the three components of particle

velocity V. at the nodes and the nine components of total stress S. . and strain E. . for

1 ID i: the elements. The state of initial stress prior to seismic disturbance may be included

in the formulation of the problem. Mean stress o and deviatoric stress o. , are defined

M ±j

by the equations

M

= \k/3

IV- 19
where the repeated index denotes summation and S. is the Kronecker delta. Mean strain

11 e and deviatoric strain e. . are defined by the corresponding equations M

e.. = E.. - J...

For both deviatoric stress and deviatoric strain only five components are independent.

A simplified outline of the sequence of computations is as follows:

1. At each node the components of particle velocity are known at time t, and,

for each element, the mean stress a,, and deviatoric stress O. . are known at time (t - At/2). M 1]

2. The values of V. at time t are used to compute the change in mean strain Ae and

1 M deviatoric strain Ae . . for a time interval At centered about t.

3. The values of Ae,, and Ae . . are used in conjunction with the constitutive relations

M 13 described in the next section to obtain the change in mean stress Aa^^ and the change in

deviatoric stress Ao^ ^ , which are used to update the stress components to time (t + At/2)

4. From the stress components at time (t + At/2), the corresponding force on each

node is determined and this is added to the gravitational force to give the total force on

the node

5. Values of the total force on a node at time (t + At/2) are used to update the

particle velocity to time (t + At)

The increment in strain over a time At is given by

3V . 8V AB. . - 1/2 ^ ^) At (2)

n ox. ox.

Once the increment in total strain is known, the increment in mean strain and deviatoric

strain can readily be found using equations (1)

Particle velocities are assumed to be independent of the x^ coordinate, so that

8V. 8X2

The quantity 9v^/ Bx^ , for values of j other than 2 , is approximated in terms of the

particle velocities at the nodes. Consider the two elements diagrammed in Figure 2 with

the nodes numbered as shown. For the triangle assume that the velocity within an element

is a linear function of position. This leads to the expression

|^%(V.(3) - V.(l))/Ax

^^(V. (3) - V (2))/Ax (3) i 3 where Ax is the nodal spacing and the numbers in parenthesis refer to the nodes. Corres- ponding expressions can be derived for triangles of other orientations. The square

IV-20
elements are considered as two triangles within each of which velocity is a linear function

of position. It does not matter which diagonal is used to divide the gquare into triangles,

the result is the same. Averaging the values from the two triangles gives for the square

elements

V. (2))/2Ax

V.(3))/2Ax (4)

Equations (1) through (4) are used to obtain the strain increments Ae and Ae . . for

M

the elements from the velocities V. at nodes. The strain increments are used with consti1 tutive relations described in the next section to obtain stress increments Ao and Ao. ,, M i: which are accumulated to give the stresses aMi: and a. . in the elements.

In order to advance from one time step to the next, it is necessary to compute the

forces on the nodes. The force on each boundary segment of an element resulting from

stress in that element is computed using Cauchy ' s formula (Fung, 1965, p. 63). Half the

force on the boundary segment is applied to the node at one end of the segment and half

to the node at the other end. The total force on an internal node is simply the sum of the

contributions from all the boundary segments connected to that node plus the gravitational

force on the node. The nodes along the outer boundary of the basin are subject to an additional force resulting from stresses in the external medium. The part of the additional

force that is constant in time is computed from the initial state of the system by applying

the requirement of equilibrium. The dynamic boundary resulting from seismic disturbance

is, handled by methods described in a subsequent section of the report.

From the total force acting on an internal node, the velocity of that node in the

next time step is given by

V, (t + At) = V. (t) + F.At/m

i

' ' 1 (5)

where m is the mass lumped at the node. Computation of for nodes along the boundary

with the underlying medium is described in the section on boundary conditions.

3. Constitutive Relations

The stress-strain model used in the computations was given by Iwan (1967) . The one- dimensional version of that model for simple shear is shown in Figure 3. It is composed

of simple linear springs and Coulomb friction units arranged as shown. The friction

xmits remain locked until the stress on them exceeds the yield stress Y^. Then, they

yield, and the stress across them during yielding is equal to the yield stress. Generally,

the yield stress of the first unit is set to zero. The spring constants are chosen

so as to reproduce stress-strain behavior measured in the laboratory.

1 o< (V. (4) + V. (3) - V^(l)

^^i s (V.(l) + V.(4) - V.(2) -

IV- 21
The type of hysteresis loops that such a model produces is shown in Figure 4, which

illustrates the behavior of a single soil element subject to cyclic shear loading of increasing amplitude. The loops of Figure 4 are plotted in terms of reduced stress and

reduced strain - scaled in such a way that the maximum stress on the sample is one unit

and the low-strain modulus has a value of one. The spring constants of the model were

chosen to give the behavior as a function of strain indicated by the experimental work of

Hardin and Drnevich (1972a, 1972b) . The model of Figure 4 contains 50 springs and 50

friction units.

In order to extend the model to three dimensions, Iwan (1967) introduced an extension

of the standard incremental theory of plasticity (Fung, 1965). Instead of a single yield

surface in stress space, Iwan postulated a family of yield surfaces. The development of

the constitutive relations given below follows Iwan with minor modifications.

The relationship between mean stress and mean strain is presumed elastic so that

where K is the bulV^ modulus. The deviatoric stress a. . is considered as a vector in ninedimensional space, and a family of yield surfaces is postulated, represented by the yield

functions

2 F (a,, - a ,,) = k

n ij nij n

where k is a constant characteristic of the nth surface and a . . represents the "origin"

n — ni:

of the surface. The total deviatoric strain e. . is presumed to consist of the sum of an

ID elastic strain e„ . . plus plastic strain components e . . each associated with the nth

yield surface. Kinematic hardening of the Prager type is assumed so that

da , , = C de- ^ . (7) nij n Pnij

where C is a constant associated with the nth surface,

n —

The plastic strain increments must be normal to the corresponding yield surfaces.

This gives

de_. . . = L h n (8) Pnij n n

dO . .

where

L = 0 if F < k

^ or ^^n do. < 0: n n — .J il

L = 1 if F = k

^ and ^^n do > 0. n n n ij '

3-J

IV- 2 2
The requirement that loading from a plastic state must lead to another plastic state can

be used to determine h^ (Fung, 1965, p. 147-151).

, OF /3a ) da

h = _1 n rs rs (S) " C (3F-/3a, J(^F /3o, ^

)

n n kl n kl

The elastic deviatoric strain is given by

e = 0../2G

Eij ij °

where G is the low-strain shear modulus,

o Summing elastic and plastic strain increment components and substituting from equations (8) and (9) gives the total deviatoric strain increment

de, . = Q, . da + da . ./2G

ij ^ijrs rs ij o (1'))

where Q . . =1

ijrs n L (3F /3a. .)(3F /3a ) n n i j n rs

Equation (10) is to be solved for da in terms of de . . , but first it should be

rs 1]

noted that only five of the components of do^^ are independent. Five independent components are selected, and using the relationships

and

o = a rs sr

°11 + °22 ^ °33

equation (10) is rewritten in terms of the independent components.

- hni " '^n2 '^12 ' ^313 ^°13 - ^123 ^"23 * ^333 ^° 33

^

where (i,j) takes on the values (1,1), (1,2), (1,3), (2,3), and (3,3). The coefficients

are given by

IV- 2 3
ij 11 n

L OF /3a, . C OF /3a, , n n kl

L (3F /3a,

n n ii ^^^^ " C OF /3a

ijl3 n

ij23 n

lj33 n

n' kl

L (3F /3a, . n n Ij C (3F /3a

n n kl

L Of /3o. . n n ij

C OF /3o n ,n kl

^ L OF /3a. . L n n 11

Ti -J C (3F /3a

OF /9c^^ - 3F /3a„)

n 11 n 22 (3F^/3a^^)

(3F /3a, ^ + 3F /3a-,

)

n 12 n 21

(3F /3a ) n kl

(3F /3a,, + 3F /3a,,)

n 13 n 31

(3F^/3a^^)

(3F /3a„_ + 3F /3a, „) n 23 n 32 OF /3a, ,) n kl

OF /3a,, - 3F /3o,-)

n 33 n 22 (3F /3a, J n n kl n kl

By solving equation (6) and (11) , the stress increments are obtained from the strain

increments. Gaussian elimination is used in solving equation (11). Since the equation

must be solved for each element at each time step some care was taken to insure that the

programming of the solution was efficient.

The coefficients in equation (11) depend on 9f /9a. ., C , and L . The yield condition

n n n of von Mises is used so that

F = 1/2 (a. . - a . ,) (a. . - a . .) n ij nij ij ni3

(12)

and

9Fn

= a

nij

With that choice the parameter in the equation that described the yield surface repre- sents the initial yield stress in simple shear. The values of k and C are chosen in

n n order to fit laboratory data on the stress-strain behavior of the material. In order to determine the value of L we need to know the value of F and the sign of the quantity

n n Of /9a. .) da. .. The first is no problem, but the second requires a knowledge of da. ., which is the unknown in equation (11) . This difficulty is avoided by a procedure that is

the equivalent of choosing L in accordance with the sign of ( 9f /9a. .) da. . during the

n n 13 13 previous time step. The adequacy of this approximation has been checked by comparing

with the results of one-dimensional calculations for which the approximation was not made.

IV- 2 4
If during a given time step yielding takes place on the nth surface, the quantities

a^_j^j are changed. In order to ensure that the stress point always remains exactly on a yielding surface, equation (7) is not used to calculate the change in O ... Coincidence

ni]

of the stress point and the yielding surface is maintained by using the following

equation

a- .

. = g' . . - ^^n^" ij nij^ (13) nij ij ^,2

kl nkl kl nkl

where the unprimed quantities represent values at time {t-At/2) and the primed quantities

values at time {t+At/2)

.

The actual sequency of computations is given below, where the primed quantities are computed in the current time step and the unprimed quantities are inherited from the previous time step.

1. New values of the stress components 0' . . are obtained from

2. The quantity T^ is evaluated as

3. The quantities 8f /do. . are evaluated. The equation used can be derived by

noting that form (12)

3F

Combining with equation (13) gives

3F k

_D_ = -iL (a- - a ). (1^)

4. The coefficients of equation (11) are evaluated using (14) . Equation (11) is

then solved to give the new stress increments Ao . . . Factors of (k /T ) from (14) appearing

1 j n n in the numerator and denominator of the expressions for the coefficients of (11) cancel

out, simplifying the calculations.

5. Equation (13) is used to evaluate a' ... nij

IV- 2 5
Determination of the constants is accomplished by considering the behavior of the

system in simple shear. Iwan (1967) showed that the behavior of the three-dimensional

model in simple shear is equivalent to the behavior of the one-dimensional model, which is

illustrated in Figures 3 and 4 of this paper. It is assumed that the loading curve from an initial state of zero deviatoric stress and strain is known. In principle it could be

determined from laboratory measurement. With the proper choice of coordinate axes the

non-vanishing components of stress and strain in simple shear are

^13 = °31

^13 = ^31

The parameter represents the yield stress in simple shear for the nth yielding surface.

A set of values of k^ is chosen to cover the range of stress the system is expected to encounter, and the values are distributed so that the initial loading curve can be faithfully represented. From the initial loading curve a set of values e^ (n=l, N) is obtained

for the shear strain corresponding to the stress values k^^. A stress value ^-^^-^ is

chosen larger than k^^ and larger than any stress that is expected to occur. A corresponding

strain value is obtained from the loading curve. For the case of simple shear

equation (10) becomes

^^3

^^3 =2^^ ^1313 '^13 ^ ^1331 '^"31- o

Evaluating

Q-j^^-j^^ ^1331 ^^^^ ^^'^ equations (9) and (12) , gives

^^13 = ^ ^ C^)

da^3 (15)

o n=l n

where j is the index of the surface with the largest yield stress of all the surfaces

that are in the yielding state. Equation (15) leads to a simple recursion relationship for determining C_.

The evaluation of C is the same except that a , is used in place of the undefined k

N ^ N+1 N+1

In order to simplify the computations, stress and strain are normalized in a way

similar to that used by Hardin and Drnevich (1972b) . Stress is normalized by multiplying

by 1/0 , where a is the maximum stress the material can withstand in simple shear, max max

and strain is normalized by multiplying by 2G /O . The normalized stress-strain curve

o max

has a limit of 1.0 for high strain and an initial slope at the origin of 1.0. The computaIV-26
tions are simplified by using the same normalized initial loading curve, and therefore

the same set of normalized C and k , for all of the soil elements in the model. The

n n differences in soil behavior from one element to another result from differences in the

values of a and G assigned to the different elements, max o For the examples presented later in this report, the following set of values was adopted for k and aN+1

k = 0.1 n 1 < n < 9 n — —

'^N+l = 0-^5^^

A hyperbolic initial loading curve (Hardin and Drnevich, 1972b) was used and normalized

strain was expressed in terms of normalized stress by the equation

k

e = —il- n 1-k

n

The method described here in no way depends on the hyperbolic relationship; a purely

empirical stress-strain curve derived from laboratory measurements could be used just as well.

The method of normalization and the use of a hyperbolic stress-strain curve is similar

to the approach of Hardin and Drnevich (1972b) , but some differences should be noted. The

stress-strain curve considered by Hardin and Drnevich is the curve formed by the loci of

the endpoints of hysteresis loops formed in cyclic loading experiments at differing peak

strain levels. The stress for the Hardin and Drnevich curve is the shear stress on a horizontal plane applied to a sample with an initial geostatic stress state characterized

by non-vanishing deviatoric stress. The stress-strain curve referred to in this report,

on the other hand, is an initial loading curve in simple shear from an initial state of

zero deviatoric stress. Given the constitutive relationships adopted in this report, an initial loading curve is identical to the loci of endpoints of hysteresis loops. The

difference in initial state of stress, however, may be significant.

Total stress rather than effective stress is used in this report to describe soil

behavior under seismic loading conditions. That is essentially equivalent to using the

undrained strength principle (Whitman, 1961, p. 587-589) , the presumption being that un- drained conditions prevail during seismic deformation. The parameter O is therefore max

related to the undrained strength , which is commonly used property in soils engineering.

Seismic deformation implies relatively rapid loading conditions, however, and values of

undrained strength measured by quasi-static methods should not be used for O without ^ max

considering the effect of loading rate. The assumption of an elastic relationship for volumetric stress and strain coupled

with the use of the von Mises yielding condition is about the simplest set of choices that

IV- 2 7
could be made. It does not incorporate in any direct way the effects of dilatancy, the

tendency of soils to change in volume when undergoing shear. The effect of dilatency upon

shear strength (through its effect on pore pressure) can be incorporated indirectly, however, by choosing values to satisfy laboratory data on the stress-strain relationship.

With the flexibility that comes from the choice of C and k values, this rather simple

n n model may give an adequate representation of the behavior of a wide range of material.

Extension to more elaborate models poses no obvious difficulties except for possible

increase in computation costs.

4. Dynamic Boundary Conditions

Problems involving the response of soils to seismic motion in the underlying bedrock

are commonly solved by assigning the particle velocity (or acceleration or displacement)

along the boundary between rock and soil. This is an approximation, because in the real

case the reaction of the soil will modify the motion of the boundary. The accuracy of the

approximation depends on the contrast in properties across the boundary. Assigning particle velocity at the soil-rock boundary is the equivalent of assuming infinite rigidity

and incompressibility for the rock. Such as assumption has the disadvantage of allowing

no energy to be radiated out of the soil system. If the rate of energy dissipation within

the soil is small, multiple reflections may give rise to resonances that are significantly

stronger than would occur in a system that allowed energy to be radiated from the boundary

This may be true even for systems with substantial contrasts in rigidity and incompressibility at the boundary. The method presented in this paper, which assumes zero energy

loss in compressional deformation, is particularly sensitive to this problem. For low

amplitude motion the energy loss in shear would be small and the problem would occur for

shear waves also. To minimize this problem, the writer has adopted a method for approximating the

boundary conditions that is based on the method used in one dimension by Papastamatiou

(written communication) . The approach is similar to that of Lysmer and Kuhlemeyer (1969)

At every point on the boundary a vector function V is defined which is the particle

F velocity that would be produced by the given seismic sources if the boundary were a free

surface. The actual motion V of the point is then approximated by substituting for the sources an equivalent combination of plane P and S waves normally incident on the boundary

The velocity V of a boundary node is obtained by solving an equation involving all of

the forces acting on the node. In general, a node along the boundary is associated with

two boundary segments (Figure 5) . The forces on half of each segment are considered to be applied to the given node. Consider one segment at a time. Define a new coordinate

system ' , rotated so that x^' is normal to the boundary (Figure 5). The specified

input function when transformed to the prime coordinate system is given by

Fi ij Fj

IV-28
where 6. . is the matrix of the transformation. Since the x„ axis is the axis of rotation

ID 2 = 1.0

The motion of the boundary node is considered the sum of incident and reflected P, SV, and SH waves with ray paths normal to the boundary node is considered the sum of incident and reflected P, SV, and SH waves with ray paths normal to the boundary segment in

V

3 = V PI + V

PR

V

1 = V

SVI + V SVR (16)

v 2 = V SHI + V SHR

V . ; is defined such that

Fi

V ' = 2V ' F3 PI en ) V ' = 2v • ^

FI svi

V ' = 2V ' F2 ^SUl

To compute for forces on the node due to the input motion, note that V ' and V

' ^ ^ PI PR

are considered plane waves in the underlying medium. The corresponding particle displacements can be expressed as follows:

"pi' = "pi' (^3' -

"pr' = "pr' (-3' ^-p^)

where v^ is the compressional wave velocity in the underlying medium.

^"Pi' _ -"'pi

3x^' v

3- p

^"pk' "'pr

3x ' V

3 P

The strain components in the underlying medium due to the hypothesized incident and re- flected P waves are

IV- 2 9 L
^ PI33 —V

P

E PR33

" V

P

The corresponding stress components are

p

V

P

The equations above can be rewritten

^'PI33 = ^Pl'

^'PR33 = ^Pr'

Substituting from equations (15) and (17) and adding gives

.^P33' = (^3' - ^F3')-

A corresponding argument gives

^'SV31 = ^'SV13 = ^^s - ^Fl')

S'SH32 = S'SH23 = ^^2

' " ^F2')

where v^ is the shear wave velocity in the underlying medium. The other stress components

are zero. The force ' applied to the given node from boundary segment a_ is then

^33' = -^"p ^^3' - ^F3'^^^a/^

^al' = -P-s - ^Fl')^^/^

^a2' =-P% (^2' -^F2')^^a/^

where As is the length of boundary segment a.

ci —

Transforming to the original coordinate system and adding the forces from boundary

segments a and b gives

F , + F, = -A.^ (V^ - V„.) (18)

ai bi ij j Fj

IV- 30
where the non-zero components of A^^ are given by

^1 = ^ ^11 ^ ^13 ^ Si

^ ^^n hi ^ hi \ hi

h3 = \ h3 ^ h\ % h3

^ ^h\ ^3 ^ ^13 ^3

^1 = ^^31 \ hi ^ ^33 \ hi

+ ^h\ \ hi ^ ^33 hi

h3 = Kl \ h3 + ^33 ^p ^3

^ ^hl ^3 + ^33 ^p ^3

pAs /2

a a

pAs

D b

a P^'J^

b

a

b P^V>^

pAs /;!. a a

, pAs./i

b o

s a

All the other forces acting on the given node point are accounted for in the basic

itational

requires that

computational scheme. Denoting the sum of those other forces by F^, Newton's second law

F, + F . + F, . = ma^ i ai bi i (19)

where m is the mass and a the acceleration of the node.

a, = (V^ - V^.)/At (20)

where V^^ is the value of from the previous time step. Substituting from equations

(18) and (20) into equation (19) gives

^ij ^h - ^Fj) ^h - \i)WAt.

IV- 31
Slight rearrangement gives

(21)

The solution of equation (21) for in terms of the other quantities is

= ^Fi + ^ij + (\j - Vp.)m/At)

where N. . is the inverse of the matrix (A. . + 6. . m/At)

i: ID i: Components of the matrix N^_. are computed for each boundary node at the beginning of

the computer program and are stored for use at each time step.

Possible methods of synthesizing realistic input motion are not taken up in this

paper. There has been much progress recently in computing expected ground motion using

dislocation theory (Boore and Zoback, 1974; Trifunac, 1974), and it is hoped that such

methods could be used for generating the required input motion. For the examples presented

later in this report, simple plane waves are used as input. Calculations using simple

input motion may give valuable insight into the character of earthquake ground response

even though the motion itself is not totally realistic.

5. Resolution, Stability, and Convergence

In choosing the mesh size Ax, the rule suggested by Boore (1972b) was adopted which

requires 10 mesh points per wavelength for the highest frequency wave for which faithful

representation is desired. Once Ax is chosen, the choice of At is limited by the re- quirement of stability. No attempt was made to do a stability analysis of the nonlinear

system. The stability requirement derived by Alterman and Loewenthal (1972) for the

corresponding linear elastic system is

V At ^ Ax [1 + ( —) ] (22)

p . \vp/

In the nonlinear system, v^ is a function of strain level, but for the strain-softening

behavior characteristic of soils v never exceeds the low-strain limit. Values of At

s chosen in accordance with (22), using the low-strain limiting value for v^ , have not led

to any stability problems in nonlinear calculations done by the method described in this

report.

Convergence was investigated by repeating an experimental run with Ax and At reduced

to one-half. The agreement was satisfactory.

IV- 3 2
6. Examples

Examples are shown for the purpose of illustrating and testing the method. They are not intended as a basis for any general conclusions regarding the response of soil to earthquake motion.

The first example is a linear elastic problem for which Boore, Larner , and Aki (1971)

have published a solution calculated by a finite difference method. The problem is a symmetrical basin filled with soft material, excited by a shear wave vertically incident

from below with particle motion parallel to the basin axis. A diagram showing half of the

symmetrical basin is given in Figure 6. The shear velocity is 0.7 km/sec in the basin

and 3.5 km/sec in the underlying medium. The rigidity ratio between the basin material

and the underlying medium is 0.0286. The incident wave has the form of a Ricker wavelet.

As explained in a previous section, the input motion for the method described in this

report is specified in terms of the motion that would be produced on the soil-rock interface if that interface were a free surface. In this example the motion is approximated

at each point on the interface by doubling the amplitude of the incident wave. The approximation neglects the diffractions generated in the underlying medium by the non-planar portions of the interface. The calculations were made using a nodal spacing of 100 meters and

a time step of 0.01 second.

The solid lines in Figure 7 show particle displacement calculated by the method of

this report. The dashed, lines give the results of Boore, Larner, and Aki (1971) for

comparison. The agreement is satisfactory, in view of the approximate specification of

the input motion and the approximate boundary conditions used in computing the solid lines.

The second example demonstrates the method for a case in which the input motion is

strong enough to give a highly nonlinear response. As in the first example, input particle

motion is parallel to the long axis of the two-dimensional configuration. The grid is

shown in Figure 8. It represents a simple, hypothetical basin 90 meters deep and 540 meters

wide at the surface. The nodal spacing is 9 meters and the time step 0.005 second. The

basin is filled with cohesive alluvial material with a plasticity index of 20 percent,

3 a density of 2.05 gm/cm , and a P velocity of 1.69 km/sec. The material is overconsolidated

near the surface, having a maximum past vertical effective stress of 2.94 bars. Complete

3 water saturation prevails throughout. In the underlying medium the density is 2.6 gm/cm , the P velocity 3.7 km/sec and the S velocity 2.0 km/sec.

The key material parameters are the maximum shear stress O and the low strain

max

shear modulus G^. Both parameters are strongly dependent upon the effective stress prior

to seismic excitation, which in turn depends upon depth. An attempt was made to assign

values to these parameters as functions of depth in a reasonably realistic manner. The

approach taken relies heavily on the work of Hardin and Drnvich (1972b) . The approach is

described in some detail to show how the parameters might be estimated for a real problem.

In the normally consolidated part of the section a was assimied proportional to

^ ^ max

effective stress,

0 = C P max s ve

IV- 3 3
where P is the vertical effective stress prior to seismic excitation. P can readily

ve ve be calculated given the density and the depth. The coefficient was evaluated by

assuming that no pore-pressure change occurs during shear and using the Mohr's circle

construction in Figure 9. This gives

1 + K max —X sin iP

2 ve

where 6 is the angle of drained shear resistance and K is the coefficient of earth preso sure at rest. Estimation on the basis of the plasticity index of the material gives a value of 31° for d) (Terzaghi and Peck, 1967, p. 112) and 0.55 for K (Brooker and Ireland,

o 1965). Carrying out the computations gives a value of 0.4 for C^.

The assumption of no pore pressure change will tend to lead to an over-estimate

of the strength and thereby of the capacity of the material to transmit shear. Normally

consolidated clays tend to contract with shear, increasing the pore pressure and reducing

the strength.

For the overconsolidated part of the section it was assumed that 0 obeyed the max

following relationship

o = C P (OCR)"^ max s ve

where OCR is the overconsolidation ratio in terms of vertical effective stresses. Analysis

of undrained strength data from Ladd and Edgers (1972) gave a value of 0.75 for T. Hardin and Black (1969; Hardin and Drnvich, 1972b) have shown that for a wide

variety of undisturbed cohesive soils, and sands as well, the low-strain shear modulus

is given by an equation which can be rewritten in the form

G = n P

^''^ (OCR

o G me m

where P is the mean effective stress, OCR is the overconsolidation ratio in terms of

me m

mean effective stresses, C is a constant depending on void ration, and H depends on

G plasticity index. For a PI of 20 percent H has a value of 0.18.

It is more convenient in the present case to deal in terms of vertical stresses.

Using values from Brooker and Ireland (1965) to convert from mean stress to vertical

stress, it was found that the values of equation (23) could be reproduced within a few

percent over the range of 1.0-32 in OCR by the following equation

G = P (OCR)^

o G ve

where Q takes on a value of 0.28 for a PI of 20 percent. A value of 0.9 x 10^ (dynes/cm^

was chosen for C based on shear velocity on shear velocity measurement in a 180 meter

G drill hole in alluvium in the San Francisco Bay Area (Warrick, 1974) . With these choices

IV- 3 4
the low-strain shear velocity in the basin ranges from about 200 to 300 meters/sec. The

vertical transit time through the basin for shear waves at low strain is 0.3 second. This

implies a fundamental resonant frequency of 0.8 Hz for the equivalent plane layer at low

strain. The resolution for low strain excitation is approximately 2.5 H^. At high strain

levels this will be reduced somewhat.

Initial values for the total vertical and horizontal stress are specified for each

element. Vertical total stress is readily calculated given the depth and the density.

Vertical total stress is converted into vertical effective stress by si±)tracting the fluid

pressure. Horizontal effective stress is then calculated from vertical effective stress

using data; from Brooker and Ireland (1965) for the assumed PI of 20 percent. Finally,

horizontal total stress is obtained by adding the fluid pressure.

The input consists of a plane SH wave incident from the left at an angle of 30° from

the vertical. The time history of the particle velocity, after doubling to allow for the

effect of the bedrock surface, is that of the integrated N21E component of the Taft

accelerogram multiplied by a factor of 4, giving a peak velocity of 67 cm/sec. (The Taft

accelerogram was recorded during the 1952 Kern County, California, earthquake.) No

attempt was made to correct the input motion to account for departure of the bedrock

surface from a plane. This approximation should be adequate because the dimensions of

the sloping portions of the basin are small compared to a wavelength in the underlying

medium (Boore, 1972a)

.

Results are shown in Figure 10, which gives the horizontal component of particle

velocity in the direction of the long axis of the basin at the points on the surface

designated in Figure 8. The peak particle velocity at the midpoint of the basin is

larger than that of the input motion by about 25 percent. Computations with a one- dimensional method (Joyner and Chen, 1975) give a particle velocity time history almost

identical to that for the two-dimensional calculation at the midpoint of the basin. The

peak values differ by less than one percent.

Even though the input motion has a component only in the direction of the basin axis

,

the components of particle velocity in the plane perpendicular to the axis do not vanish,

as they would in the elastic case. Because of the initial stress and the nonlinear material

behavior, antiplane strain motions (motions parallel to the long axis of the structure)

are not uncoupled from plane strain motions (motions in the plane perpendicular to the

long axis) . The plane strain components (x^ and x^) are very small, however, the peak

being no more than three percent of the peak antiplane strain component for nodes 1 through 5 on Figure 8. This suggests the feasibility of developing simplified models in

which only antiplane strain motion is permitted. Such models would reduce computation

time by a large factor, in part because the number of computations per time step would be

significantly reduced, but also because larger time steps could be used. This is true

I because, if compressional deformation is excluded from the model, the stability requirement on At (Boore, 1972b) is much less restrictive. Plane strain models are subject to

I the same stability requirement as the general model, but they, too, offer savings in

IV- 3 5
computation time by reducing the number of computations per time step.

For a model restricted to antiplane strain it is necessary to ignore initial stress

and carry out the computations in terms of stress departures from an unspecified initial

state rather than in terms of absolute stress. Computations made with both antiplane

strain and plane strain input motions indicate that this is feasible. The results of runs made with initial deviatoric stress set to zero show little difference from those with

initial stress assigned by the methods described previously. Ignoring initial stress and

dealing in terms of stress departures from an initial state is convenient in the plane

strain case, though not necessary as with antiplane strain. For the example of Figures 8 and 10, the computations required 740 seconds on the

CDC 7600 at the Lawrence Berkeley Laboratory. This is the equivalent of 360 microseconds

per element per time step and indicates that the method is fast enough to be useful.

(Running time is essentially proportional to the product of the number of elements and

the number of time steps.) The larger time step permissible with an antiplane strain

model would reduce the computation time on this example by a factor of six. A further

reduction by about a factor of two would result from the reduction in the number of computations per time step in an antiplane strain model. Even with the reductions, however,

the cost of the computations will place a practical limit of the size of the problem that

can be solved for a given frequency resolution.

The final example involves plane strain input motion and illustrates the application

of the method to calculating the response of earth dams. As in the previous example, the

input motion is strong enough to produce a highly nonlinear response. The configuration,

as shown in Figure 11, represents a simplified model of an earth dam. The nodal spacing

is 9 meters and the time step 0.005 second. The elements marked with dots represent the

clay core of the dam and the other elements represent granular material. For simplicity

3 all elements are assigned a density of 2.05 gm/cm , a P velocity of 1.69 km/sec and a low

strain shear velocity of 300 meters/sec. This gives a low-strain frequency resolution of

3.3 Hz. The maximum shear 0 is 0.9 x 10^ dynes/cm^ for the clay core and 1.2 x 10^

2 "^^^3 dynes/cm for the granular material. In the underlying medium the density is 2.6 gm/cm , the velocity 3.7 km/sec and the S velocity 2.0 km/sec. The initial stress is ignored

and the computations are carried out in terms of stress departures from the unspecified

initial state. Only plane strain motions are permitted.

The input motion consists of a vertically incident plane shear wave with particle

motion confined to the plane of the section shown in Figure 11. The time history of

particle velocity for the input motion is the same as in the previous example, with the

amplitude doubled, as in the previous example, to allow for the effect of the bedrock

surface

The results are shown in Figure 12 , which gives the horizontal particle velocity for

the points on the surface indicated in Figure 11. The peak velocity at the crest is

approximately 30 percent greater than the peak for the input. The computations required

200 seconds on the CDC 7600, which represents 210 microseconds per element per time step.

IV-36
7. Conclusions and Applications

The Iwin rheological model leads to an ef f icientmethod for two-dimensional soil response calculations that takes account of the nonlinear, hysteretic behavior of soil and

allows flexibility for incorporating laboratory data on soil behavior. The method can be

used to compute free-field response at sites where the soil configuration cannot be

adequately represented by simple plane layers. For a given frequency resolution, however,

the size of such models is limited by computer time requirements . The method has particular promise for use in calculating the response of earth dams. It could also be

adapted to the calculation of soil-structure interaction in two or three dimensions

.

8. Acknowledgments

The writer is indebted to Harold W. Olsen for his advice on questions relating to soils dynamics and to David M. Boore and D.J. Andrews for helpful suggestions concerning

two-dimensional finite mesh calculations. D.J. Andrews and Albert T.F. Chen read the

mainuscript and suggested a number of improvements.

References

1. Alterman, Z. , and D. Loewenthal (1972). Computer generated seismograms , in

Methods in Computational Physics, Vol. 12, Seismology, Body Waves and Sources , Bruce A. Bolt (ed.). Academic Press, New York, 35-164.

2. Boore, D.M. (1972a). A note on the effect of simple topography on seismic SH waves. Bull. Seism. Soc. Am., 62, 275-284.

3. Boore, D.M. (1972b). Finite difference methods for seismic wave propagation

in heterogeneous materials, in Methods in Computational Physics, Vol. 11, Seismology , Surface Waves and Earth Oscillations , Bruce A. Bolt (ed.) , Academic Press, New York, 1-37.

4. Boore, D.M. , K.L. Larner, and K. Aki (1971). Comparison of two independent

methods for the solution of wave-scattering problems: Response of a sedimentary basin

to vertically incident SH waves, J. Geophys . Res., 76 , 558-569.

5. Boore, D.M. , and M.D. Zoback (1974) . Two-dimensional kinematic fault modeling of

the Pacoima Dam strong-motion recordings of the February 9, 1971, San Fernando earthquake.

Bull. Seism. Soc. Am., 64 , 555-570.

6. Brooker, E.W., and H.O. Ireland (1965). Earth pressures at rest related to

stress history, Canadian Geotechnical Journ. , 2, 1-15.

7. Chen, A.T.F., and W.B. Joyner (1974). Multi-linear analysis for ground motion

studies of layered systems. Report No. USGS-GS-74-020 , NTIS No. PB232-704/AS , Clearinghouse, Springfield, VA 22151.

8. Constantopoulos , I.V. (1973). Amplification studies for a nonlinear hysteretic

soil model. Research Report R73-46, Department of Civil Engineering, Massachusetts

Institute of Technology, 204 p.

9. Faccioli, Ezio, Enrique Santoyo V., and J.T. Leon T. (1973). Microzonation

criteria and seismic response for the city of Managua, Proceedings of the Earthquake

Engineering Research Institute Conference on the Managua, Nicaragua, Earthquake of

December 23, 1972 , Vol. I, 271-301.

10. Fung, Y.C. (1965). Foundations of Solid Mechanics , Prentice-Hall, Englewood

Cliffs, New Jersey.

11. Hardin, B.O., and W.L. Black (1969). Vibration modulus of normally consolidated

clay (discussion) , Proc. Am. Soc. Civil Engin. , Journ. Soil Mech. and Found. Div. , 95,

1531-1537.

IV- 3 7
12. Hardin, B.O. , and V.P. Drnevich (1972a). Shear modulus and damping in soils:

Measurement and parameter effects, Proc. Am. Soc. Civil Eng., Journ. Soil Mech. and Found . Div . , 98, 603-624.

13. Hardin, B.O. , and V.P. Drnevich (1972b). Shear modulus and damping in soils:

Design equations and curves, Proc. Am. Soc. Civil Eng., Journ. Soil Mech. and Found. Div ., 98, 667-692.

14. Hudson, D.E. (1972) . Strong motion seismology. Proceedings of the International

Conference on Microzonation , Seattle, Washington, Vol. I, p. 29-60.

15. Idriss, I.M. , and H.B. Seed (1968). Seismic response of horizontal soil layers,

Proc. Am. Soc. Civil Eng., Journ. Soil Mech. and Found. Div ., 94, 1003-1031.

16. Idriss, I.M., and H.B. Seed (1974). Seismic response by variable damping finite

elements, Proc. Am. Soc. Civil Eng., Journ. Geotechnical Eng. Div., 100 , 1-13.

17. Iwan, W.D. (1967) . On a class of models for the yielding behavior of continuous

and composite systems. Jour, of Applied Mechanics , 34, 612-617.

18. Joyner, W.B. , and A.T.F. Chen (1975) . Calculation of nonlinear ground response

in earthquakes. Bull. Seism. Soc. Am. (in press)

19. Kanai, K. (1952). Relation between the nature of surface layer and the amplitudes of earthquake motions. Bull. Earthq. Res. Inst., Tokyo Univ., 30 , 31-37.

20. Ladd, C.C., and Lewis Edgers (1972). Consolidated-undrained direct-simple

shear tests on saturated clays. Research Report R72-82 , Department of Civil Engineering,

Massachusetts Institute of Technology, Cambridge, Mass.

21. Lysmer, John, and R.L. Kuhlemeyer (1969). Finite dynamic model for infinite

media, Proc. Am. Soc. Civil Eng., Journ. of the Eng. Mechanics Div., 95 , 859-877.

22. Newmark, N.M. , A.R. Robinson, A.H.-S. Ang, L.A. Lopez, and W.J. Hall (1972).

Methods for determining site characteristics. Proceedings of the International Conference

on Microzonation , Seattle, Washington, Vol. I, p. 113-129.

23. Schnabel, P., H. Bolton Seed, and J. Lysmer (1972). Modification of seismograph

records for effects of local soil conditions. Bull. Seism. Soc. Am., 62, 1649-1664.

24. Streeter, V.L., E.B. Wylie, and F.E. Richart (1974). Soil motion computations by

characteristics method, Proc. Am. Soc. Civil Eng., Journ. Geotechnical Eng. Div., 100,

247-263.

25. Terzaghi, Karl, and R.B. Peck (1967) . Soil Mechanics in Engineering Practice

(2d ed.), Wiley, New York.

26. T.rifunac, M.D. (1974) . A three-dimensional dislocation model for the San

Fernando, California, earthquake of February 9, 1971, Bull. Seism. Soc. Am., 64 , 149-172.

27. Warrick, R.E. (1974) . Seismic investigation of a San Francisco Bay mud site

Bull. Seism. Soc. Am. , 64 , 375-385.

28. Whitman, R.V. (1961) . Some considerations and data regarding the shear strength

of clays, in Research Conference on Shear Strength of Cohesive Soils , American Society

of Civil Engineers, New York, 581-614.

Figure Captions

Figure 1 - Schematic diagram of computational grid.

Figure 2 - Diagram showing the different types of elements and the numbering of the

nodes

Figure 3 - Rheological model in one dimension (Iwan, 1967).

Figure 4

- Hysteresis loops produced by the model of Figure 3 under cyclic loading

with increasing amplitude.

Figure 5 - A portion of the boundary.

IV-38
Figure 7 - Horizontal particle displacement for the soft basin problem. Solid lines

represent results of the method described in this report and dashed lines

are from Boore, Larner, and Aki (1971)

Figure 8 - Grid for hypothetical basin described in text. Figure 9 - Mohr's circle construction for determining a . The inner circle represents the intital geostatic stress state and "Me outer circle represents

the limiting state. Figure 10 - Horizontal component of particle velocity in the direction of the long axis

of the basin at the points on the surface designated in Figure 8. Figure 11 - Grid for simplified model of earth dam described in text. The elements

marked with dots represent the clay core. Figure 12 - Horizontal particle velocity at the points on the surface designated in

Figure 11.

i

IV- 39
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IV-41
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IV-44
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IV-46
A NEW SCALE REPRESENTING THE "QUAKE- SENSITIVITY"

AT A CERTAIN REGION

TSUTOMU TERASHIMA

Chief, Seismology Section

International Institute of Seismology

and Earthquake Engineering

Building Research Institute

Ministry of Construction

Japan

TETSUO SANTO

Head

International Institute of Seismology

and Earthquake Engineering

Building Research Institute

Ministry of Construction

Japan

ABSTRACT

The nature of ground motions due to earthquakes, depends on the property of the

superficial materials of the ground, which vary in different areas. In this paper,

the different characteristics of the ground motions is normalized by a new scale designated

as "Quaker Sensitivity. " This scale is defined as a ratio of N(I)/S, where N(I) is the

annual mean frequency of seismic intensity of more than III (in J.M.A. scale) and S is

seismicity index which has been defined previously by the authors.

A seismic zoning map was then made for Japan Islands relative to the "Quake- Sensitivity". The area having a large value of N(I)/S means that the area is sensitive to earthquake motions or the neighboring area has moderate earthquakes.

Key Words: Earthquake; Index; Quake- Sensitivity; Seismic Index; Seismicity.

1. Introduction

In the previous seventh joint meeting in Tokyo of this Panel (May 20-23, 1975)

,

we introduced "Seismicity Index S" as a quantitative seismicity scale being defined as follows

:

S» (N(M)/T) ^ < ^^^^^ T= 100 years,

in which N(M) is the frequency of shallow earthquake with the magnitude M greater than 6. If a geological structure in a certain region is uniform, the earthquake with the

magnitude greater than 6 shakes the ground with seismic intensity of more than approximately III inside a circular area with diameter of 100 km. The distribution of the

seismicity index S above defined, therefore, might also represent a regionalization with

regard to the annual frequency distribution of a seismic intensity of more than III due to the occurrence of earthquake of M > 6.0 during 100 years.

IV-47
In practice, however, the intensity distribution due to a certain earthquake is

almost always distorted from a circular configuration because of the different superficial conditions from region to region (Fig. 1) . Moreover, if we also take into account

the intermediate or deep earthquakes, the distortion of the isoseismal distribution from

the circular shape due to the passing and/or the absorption of short period seismic waves

caused by the penetration of a rigid slab (Fig. 2) (Utsu et al., 1968). In other words,

the distribution of the seismicity index S is not directly related to the distribution

of the intensity larger than III. The present paper is a first step in determining the distribution of the regions

which were subjected to an intensity of more than a certain level, say the intensity III on the Japanese scale. Such intensities are caused by special superficial conditions and/

or the geographical situations against the distributions of deeper shocks and abnormal

situations in the upper mantle being underthrusted by a slab. For this purpose, a new

scale representing the "Quake-Sensitivity" of the ground, is proposed.

2. Definition of the "Quake-Sensitivity"

As was mentioned in the previous section, our final purpose is to select special re- gions where the ground is subjected to abnormally high intensities. The following quantity

will be used to describe this intensity;

Q.I, = NCl)/3,

where the quake intensity is given by the ratio of the frequency of seismic intensity

more than a certain level N(I) to the seismicity index S. Taking this ratio, we can exaggerate the regions where the grounds were frequently quaked either due to the occurrence of near (A < 100 km) earthquakes of M < 6.0 or due to particularly worse situations

related to the superficial and/or upper mantle conditions.

3. Analytical Example

Distribution of "Quake- Sensitivity" thus defined is surveyed for Japan. As for N(I),

frequency of seismic intensity greater than III (J.M.A. scale) in period 1921-1973 at every stations (Terashima, 1976) are used. As for Seismicity Index S, the result by T. Terashima (1973) is also used directly.

Fig. 3 shows the contour lines of the annual mean frequency of seismic intensity

greater than III. Fig. 4 shows the distribution of N(I)/S as derived by superimposing Fig.

3 on Fig. 5 and shows the distribution of seismicity index S in and near Japan. Fig. 4 is further rearranged as given in Fig. 6. From the distribution of high numbers in Fig.

4 and Fig. 6, we can suggest the regions where the ground can be easily quaked. From

these maps, the following is recognized:

1< There are sensitive regions for seismic motions even if the seismicity is low. For instance, seismicity indexes S of northern and middle regions in Hokkaido, Niigata

prefecture and western part of Kyushu are smaller than Pacific side of Tohoku and Kwanto

IV-48
regions. But, N(I)/S of those regions are larger than those of Pacific side of Tohoku and

Kwanto regions. This means that the regions quite often suffer from the seismic motions

comparec^ with Pacific side of Tohoku and Kwanto regions due to smaller events or due to the

geological situations.

2. There are non-sensitive regions even if seismicity is active. For instance,

seismicity indexes S on Pacific side of Tohoku and Kwanto regions show the largest values

in and near Japan. But N(I)/S are not large compared with the other regions. This means

that these regions have been quaked mainly due to the events with magnitudes larger than 6.

4. Conclusions

A new scale of "Quake- Sensitivity" which is a ratio of the annual mean frequency of

seismic intensity to seismicity index, has been proposed. Regionalization of this Q.I. were made for the Japan Islands (Fig. 4 and Fig. 5) for an example. These maps show the

regions in which the ground is frequently quaked due to the following situations:

1. Seismic activity of the earthquakes with the magnitude less than 6 is high in

the neighborhood (A < 100 km) of the position.

2. The region with bad superficial geological conditions or the region whose locations are such that the high intensities due to the deeper shocks easily appear by the

penetrating slab.

Among these conditions, regionalizations due to the second cause is more important by

eliminating the effect due to the first cause. We shall aim to make this kind of

regionalization as a next step for various intensity levels.

References

1. Utsu, T. and H. Okada, 1968, Anomalies in Seismic Wave Velocity and Attenuation

Associated with a Deep Earthquake Zone (II). Jour. Faculty of Science, Hokkaido Univ.,

Ser. VII, (Geophysics), Vol. 3, No. 2, 65-84.

2. Terashima, T. , 1973, Quantification of Seismicity, Bull. Inter. Inst. Seism.

Earthq. Engg. , 11, 33-41.

3. Terashima, T. , 1976, Frequency Distribution of Seismic Intensity in Every

Station Belongs to Japan Meteorological Agency, Kenchiku Kenkyu Shiryo, (in press)

.

IV-49
The East Off Fukushima Pref. Earhquake, IMS

134» 136' 138° '"0°

Fig.l. Examples of complicated distribution of seismic intensity.

IV-50
1

Fig. 2. Constitutions of crust and upper mantle of the earth and regions

of abnormal seismic intensity.

Fig. 3. Mean annual frequency of seismic intensity more than III (J.M.A.

scale) at every station and the isoplethic curves,

IV- 51
IV- 5 2
IV-53

DAMAGE TO THE CIVIL ENGINEERING STRUCTURES

IN HACHIJOJIMA ISLAND

BY TYPHOON 7513

T. OKUBO

Head, Planning and Research Administration Division

Public Works Research Institute

Ministry of Construction, Japan

N. NARITA

Chief, Structure Section

P.W.R.I

K. YOKOYAMA

Research Engineer

Structure Section, P.W.R.I.

ABSTRACT

Typhoon 7513 passed across Hachijojima Island on 16:40, October 5, 1975 and caused

damage to property, houses and public service. The Public Works Research Institute performed a field investigation in Hachijojima Island. In this paper the record of strong

wind and the state of damage are introduced. Typhoon 7513 is classified as a small but

strong typhoon, and the maximum wind speed, averaged over ten minute intervals was 35.5

m/s, and the maximum instantaneous wind speed was 67.8 m/s

.

The number of the wounded was 85, but fortunately no one- was killed. A great deal

of damage was done to houses, trees, poles and fences. The damage, with regard to public

service was the destruction of a school and hospital building, and the interruption of

electric power, telephone, and roads. The damage to large scale structures consisted of

the destruction of a guyed tower 30 meters high, a panzer-mast, a bridge and four arc

lamp standards.

Key Words: Damage; Structures; Typhoon; Wind; Wind Effects.

1. Introduction

Typhoon 7513 passed the southeast of Okinotorishima Island on October 2, 1975, then

continued to the north, and grew to a small but strong typhoon. The typhoon then advanced

to the sea south of Honshu (Mainland of Japan) and passed through the Hachijojima Island

on 16:40, October 5. This typhoon 7513, caused damage to property, houses, and public

service such as electric power, telephone and roads.

The Public Works Research Institute performed a field investigation throughout

Hachijojima Island, from October 9 through October 11. In this paper, the record of the strong winds and the damage situation on Hachijojima

Island are introduced.

V-1
2. Strong Wind Characteristics

a) Abstract of the Typhoon 7513

According to the announcement of the Meteorological Agency, typhoon 7513 passed near Hachijojima Island on 16:40, October 5 as shown in Fig. -1. At that time the central

pressure of the typhoon was 945 mb, the maximum wind speed was 45 m/s. The scale of storm

area, in which wind speed was more than 2 5 m/s, was about 2 30 km long to the southeast

and 140 km long to the northwest as shown in Fig. -2. Typhoon 7513 can be classified as

"a small but strong storm" according to the standard classification of typhoons shown in

Tables-1 and -2. The observed wind at the Hachijojima Weather Station is shown in Table -3. This

data indicates that the characteristics of the typhoon, are as follows:

1. The maximum wind speed, at its center, was high in comparison to its scale. (No

damage occurred to the Aogashima Island, 60 km from the Island.)

2. Typhoon 7513 had a strong intensity and a clear eye, until it came near Hachijojima Island.

3. The stormy zone area was small and the travelling speed was as high as 60 km/h

,

a strong wind of 15 m/s was observed for a short period of time. (3 hours from 15:00 to

18:00, October 5 in Hachijojima Island).

4. The total precipitation on Hachijojima Island was only 25.5 mm, and most part of

the damage was caused by the strong wind.

The strong wind caused by typhoon 7513 and previous strong wind observations are given in Table -4 and Fig. -3. From these data it can be stated that;

(a) There are more possibilities to have hard winds on the Hachijojima Island than

on the Honshu Island.

(b) The maximum wind speed caused by typhoon 7513 is the seventh worst during

the past 47 years, such a storm has not occurred during the past 15 years.

(c) The frequency of such a strong wind due to a typhoon, is about 7 years.

(d) The time when the maximum instantaneous wind speed was observed did not coincide with the time when the maximum wind speeds averaged over a ten-minute intervals

was observed. The gust factor was 1.9 relative to values of 1.2-1.5 as usually observed.

b) The Characteristics of Topography and Wind on Hachijojima Island

Hachijojima Island is located 300 km south of Tokyo and has a circumference of

about 60 km. As shown in Fig. -4, Mt. Hachijo-Fuji (854.3 m high) is located in the

northwest and Mt. Mihara (704 m high) is located in the southeast, and has facilities such

as an airport and town offices, between two mountains. The wind direction was south during

the typhoon, and was west when it was leaving; it appears that the wind was influenced by

the two mountains.

Wind observation was performed, not only at the Hachijojima Weather Station, but

the Hachijojima Lighthouse in Sueyoshi and at the Hachijojima Airport in Ohkago. The

wind record traces at these stations, are shown in Fig. -5. The Hachijojima Weather Station is on a hill 79.7 m above the sea level and 1 km distance from the seashore in

Ohkago, and has instrijmients which are on a building as shown in Photo -1. The observed

V-2
wind speed is thus influenced by the topography.

Hachijojima Lighthouse is located on a bluff located at the southeast tip of the island as shown in Photo -2. The intensity of the wind turbulence is very low, in comparison

with the wind record trace of the weather station shown in Fig. -5 (1) , (2) . When the

typhoon passed through, the wind direction changed from south to west and the wind speed

became weak due to the effect of the topography.

Hachijojima Airport is located at the foot of Mt. Hachijo-Fuji. Anemometers have

been installed on the top of the building and at the east tip and west tip of the runway

at this airport. The wind speed observed at the airport was nearly the same at that

recorded at the weather station.

The wind data at each station, as given in Fig. -6, 7 and Table -5, reveal the following comparison:

1. The wind speed observed in the lighthouse is higher than that in the weather

station, where the magnification factor is from 1.25 to 2.0.

2. The wind speed observed in the airport is nearly the same as that in the weather

Station

.

The wind characteristics at Hachijojima Island, due to typhoon 7513, are summarized

as follows

:

1. The wind direction was south while the typhoon was approaching, with strong

wind of low intensity and turbulence blowing in the Yaene and Sueyoshi district near the

seashore

,

2. The wind direction changed from south to west accompanied with passage of the

typhoon. The wind speed became weak in the Sueyoshi district, because the wind from

the west is blocked by the mountains. The wind was negligible in the Ohkago and Mitsume

district located between Mt. Hachijo-Fuji and Mt. Mihara.

3. At the district located in the rear of Mt. Hachijo-Fuji and Mt. Mihara, the wind

became so weak that these areas had minimal damage.

3. Examination of Damage

a) Abstract of Damage

The statistics of damage is shown in Tables -5 and-7. The number of the wounded

was 85, but fortunately no one was killed. On the other hand a great deal of damage

occurred to the houses, where the number of fully and halfly collapsed houses were 304

(8.0%) and 539 (14.2%), respectively. The damage occurred when wooden houses leaned and

fell, the roofs was peeled off or blown away and the window glass was broken into pieces

as shown in Photos -5 and-6. Galvanized steel roofing is widely used in Hachijojima, and

the damage of the roofs, for the most part, occurred to these galvanized steel roofs. As

an example of the damage to window glass, the Hachijojima Kokusai Kanko Hotel had extensive damage as shown in Photo -7. The window glass with 8 mm thick at the south side of

building and was broken into pieces when hit by blowing pebbles etc. It was also noted

that in some instances the aluminiun window frames yielded, in spite of this the glass did

not break as shown in Photo -8. The damage relative to public service was the destruction

V-3
of a school and hospital building, and the interruption of electric power, telephone and

roads. The number of damaged major structures, except buildings, has few, and consisted

of destruction of a guyed tower, a panzermast, a bridge and four arc lamp standards, this

is because there are only a few large structures in Hachijojima.

b) Damage to Roads

Roads were lightly damaged and no large scale landslide occurred as the total

precipitation was only 25.5 mm. The roads were cut off at 150 places by such obstacles

as fallen trees and pieces of broken houses, but they cleared in a few days. Many trees,

at the slope in Yokomagaura, fell due to the typhoon. The inclination of the slope at Yokomagaura is steep and therefore, preventative work had been conducted in order to avoid future landslides.

c) Damage to Structures

(1) Bridges and Breakwater

A girder bridge, with 15 m span at the Yaene Harbor, was damaged by waves

as shown in Photo -10. This bridge consisted of 17 PC beams which fixed on the abutment

with bolts and a concrete slab. Due to typhoon 7513 the slab was lost and some anchor

bolts sheared so that some of girders fell into the sea. The breakwater at the Aegae Harbor was partially destroyed at the concrete joint. The

state of damage is as shown in Photo. -11

(2) Guyed tower of the Loran Long Range Navigation Station

Hachijojima Laran Station, of the Maritime Safety Agency is located on the

top of Mt. Mihara, and has a guyed tower 80 m high and a panzer-mast 20 m high. During

the storm both the tower and mast collapsed.

The guyed tower was constructed in 1966. It consisted of 16 blocks of triangular

truss with a side length of 60 cm, and block length of 5 m. The tower is stiffened by 4 layers and 3 direction guys as illustrated in Fig. -8. The state of damage is shown in

Photo -12. The tower failed at the base in the north direction (leeward). The 4th block

from the base of the truss yielded and the others remained without deformation. The guy was cut near the tower as illustrated in Fig. -8.

The state of destruction of this tower was presumed as follows:

1. Excessive wind loads acted on the tower, and the guy on the windward and at the

lowest layer fractured.

2. This resulted in an induced severe bending moment to act on the truss and thus

caused failure of the 4th block of the truss.

3. Consequently, the whole tower fell in leeward direction.

This tower was designed using the standard for structural designs of steel towers

(Architectural Institute of Japan) . Namely, the design wind load is calculated by the

following equation and the design wind speed is 60 m/s.

P = qCA where P: wind pressure (kg)

2 q: design velocity pressure (kh/m )

C: coefficient of drag

q = 12o'^/h~ A: projected area (m^)

V-4
h: height above the ground (m)

C = KFMCq K: coefficient of wind force for latticed structure

F: aspect ratio

M: modification factor of wind pressure

C^: modification factor for circular member

Because the highest tension acts on the highest guy according to these calculations,

it is difficult to assume that the lowest guy had first featured. But these calculations

assume that the vertical wind profile is peculiar.

The wind at this site influences its height and the effect of the topography, namely

the tower existed on the top of Mt. Mihara and its height is 704 m. As an example, the

convergence of the wind at the top of a mountain is shown in Fig. -9, as a result of

wind tunnel experiments.

(3) Arc Lamp Standard

Arc lamp standards at the Mihara Junior High School and Hachijo High School

broke near the base. The panzer-mast consisted of 8 pipes, at length 2 m, thickness of

2 mm, diameter at the base of 55 cm and a material of high- tensile-steel . The dimension

of the lamp is as shown in Table -8. The layout of school is as shown in Table -8, and

the state of the damage is shown in Photos -14 through 16. The lamp standard fell down toward the north which coincides with the wind direction,

therefore, the standards fell down due to an excessive wind load. A gap, shown in Photo

-14, was left at the base between the ground and the standard did not fall down. The

direction of this gap is perpendicular to the wind direction, thus the standards vibrated

due to Karman vortex where the amplitude at the top of the standard reached more than

10 cm. The fact that only 2 standards among 6 standards were damaged, shows that wind

affects were very local, thus, the existence of buildings or others blocks or the wind,

had an effect on the intensity of the wind.

(4) Poles and Fences

Many concrete telegraph poles broke and fell in the Ohkago and Mitsune district

(Photos -17, 18) . Most of damaged poles broke near the base, but some of them broke in

the middle. The standard for concrete poles is provided by JIS (Japanese Industrial

Standard) and the configuration and design loads are as shown in Table 9. The length

and design loads on poles used in Hachijojima are 12 m and 350 kg, respectively, and the

design wind speed is 50 m/s (for reference, 40 m/s in Tokyo).

Many concrete block fences and net fences fell. Concrete block fences, shown in

Photo -18, broke at the boundary between the foundation and blocks. Net fences, shown

in Photo -19 were damaged, although its frequency ratio was small.

I 4. Conclusion

Results from this examination are summarized as follows:

1. According to the observed data, the wind speed over a ten- minute interval was

35.5 m/s and the maximum instantaneous wind speed was 67.8 m/s, therefore the gust factor

V-5
was over 1.9 and is larger than the usual value.

2. After performing some calculations on the return period of annual maximum wind

speeds, the return period of a strong wind due to the typhoon 7513 is 7 years, which is

very small. Hence, it is necessary to consider wind effects in future design of structures

in Hachijojima.

3. A great deal of damage was done to pxiblic service. This interruption to public

service produces a powerful effect on the civilian populations, thus the design wind speed

has to be determined with due regard to the importance of the public.

4. The design wind speed for the design of structures is considered to be the

wind speed over a ten^ minute interval. However the destruction of structures is related

to the maximum instantaneous wind speed. Therefore it is necessary to examine such

effects by engineers.

Acknowledgments

Many persons have assisted the writers in the collection and evaluation of the

information presented herein. We thank each of these people for their help.

V-6
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Fig.-9 Wind profile over the summit ol moart;iin

(After the result of wind tunnel exporin cnt)

The general view of the guyed tower

Fig. -10 (1) Tae plane view of Mihara Junior Hiegh Scnwol

Fig. -10 (2) Tl-ie plane view of Hachijo Hiyh Scr.ool

V-13
Table- 1 Classification of scale of typhoon

Grade The radius of isobar "lOOOmb"

The radius of area in which

wind speed is more than 25m/s

(for reference)

The radius of maximum cir- cular isobar

(for reference)

Extremely large More than 600 km 400 km or over More than 800 km

Large 300 - 600 about 300 km 400 - 800 Medium 200 - 300 about 200 km 300 - 400

Small 100 - 200 about 100 km 200 - 300

Extremely small Less than 100 km Less than 100 km

Table-2 Classification of strength of typhoon

Grade

Central atmospheric pressure Maximum wintf speed

(for reference) Weak

mean

Strong Extremely strong

Violent

More than 990 mb

950 - 989

930 - 949

900 - 929 Less than 900 mb

Less than 25m/s

25 - 34

35 - 44 45 - 54 More than 55m/s

Hachijo-jima

T«ble-3 Observed data at Weather Stations Mlyake-jlma

Data X. *

wind Jl- m/i ma dwr

4 21 1010.2 26.5 87 3W 5.8 o

2A 1009.7 25.3 3W S.2 o

5 3 1008 7 26.2 87 SW 5.5 o

6 1008.6 25.6 3 3W 5.1 @

g 10O7.B 26 I 87 3 67 0.0 @

11 1005.2 263 87 8 SW 7.6 o

12 1002 7 265 86 S 3W &7 o

13 1000.3 264 84 3 10.0 0.0 o

14 99<l0 16 3 87 S 8W 122 0.0 •

16 990.4 2s. 6 88 8 15.9 10.5 •

16 972.7 24.7 94 3 221- 5.6 •

17 966.3 25.4 93 W 366 4.6 •

18 991.0 26.3 96 W 14.4 1.5 •

19 998.4 23.4 86 NW 12 2 1.0 ©

20 1002 3 23.8 82 WNW 12.0 o

21 1005.0 243 78 NW 7.0 o

24 1007.8 226 NW 3.8 ZO •

6 3 1009.4 21.6 89 0.0 0.5 6 1011.4 21.4 N R 1.9 ®

8 1013.5 2S0 65 BN E 5.4 0.0 ®

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997.7 26.1 90 8 E 68 5.5 993.9 2S.0 92 BSI 94 6.0 989.8 261 92 g 15.8 160 9891 26.1 88 NN E 17.5 1.5 993,6 21.9 91 NNE 16.2 126 •

9969 23,0 89 N 11.2 3 5 20 1000.6 23.6 86 N 8 8 0.0 21 10017 23.3 79 N 6.9 0.0 24

6 1008.5 22.7 66 E 4 8 •

1014.1 24.5 50 N E 4 4

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Photo-1. Hachijojlma Weather

Station

Photo-2, Hachijojlma Lighthouse

Photo-3. Hachijojima Airport Photo-4. Hachijojima Airport

(control office) (west side of runway)

V-17
!

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Photo-7. Damage to Hachijojima Photo-8. Damage to window frame

Kokusai Kanko Hotel (Hachijo High School)

Photo-9. Fallen trees in slope (Yokomagaura)

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Photo-10. Damage to bridge Photo-11. Damage to breakwater

I

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i V-18
Photo-13. Damage to arc lamp

Photo-12. Damage to guyed tower standard (Mihara

(on top of Mt. Mihara) Junior High School)

School)
Photo-18. Damage of concrete Photo-19. Damage of net fence

block fence

V-20
CYCLONE TRACY

RICHARD D. MARSHALL

Center for Building Technology

Institute for Applied Technology

National Bureau of Standards

Washington, D.C.

ABSTRACT

During the early morning hours of December 25, 1974, the city of Darwin was devastated

by the most damaging cyclone ever to strike the Australian Continent. Winds of up to 75

m/ s caused extensive damage to housing in particular, requiring the evacuation of approximately half of the 45,000 residents to other major cities in Australia. This report is a result of the author spending several days on temporary assignment with the Department of

Housing and Construction-Australian Government to inspect the damage, and to participate

in discussions regarding the establishment of new design criteria and construction practices

for cyclone areas. The fact that most of the damage was caused by wind forces rather than

a combination of wind and storm surge greatly simplified the assessment of damage and

structural performance. The experience at Darwin points out the danger in depending too heavily upon past experience and intuition in the design of housing. It also makes clear

the need for additional research into the behavior of certain building materials under

repeated loads and missile impact, and the racking strength of walls subjected to uplift

loads

.

Key words: Buildings; Cyclones; Disasters; Structural Engineering; Tides; Wind.

V-21
1.0 INTRODUCTION

Cyclone Tracy, which struck Darwin, N.T. in the early morning of December 25, 1974, was the most damaging cyclone ever to strike the Australian Continent. In the four to five hours required for the storm to pass over Darwin, approximately 50 lives were lost and damage in excess of $300,000,000 (US) was done to buildings and other structures. This report is an attempt to describe the more common

types of buildings in Darwin and their performance under extreme wind conditions . Some residential

areas suffered almost complete destruction while other areas of the city experienced only moderate

damage. The wide variety of construction and the fact that most damage was due entirely to wind effects rather than a combination of wind and storm surge make possible some important and meaningful

deductions and point up several areas that should be given closer attention in designing structures

for wind-prone areas.

2.0 STORM SUMMARY

2.1 DEVELOPMENT AND LANDFALL

Cyclone Tracy originated as a tropical depression in the Arafura Sea near 8S-134E early on December 20, 1974. The storm system moved off in a southwesterly direction and was classified as a tropical cyclone

during the evening hours of the 21st. On the morning of the 22nd, Tracy was located at 9S-131E, or approximately 200 km north of Darwin. It continued to move toward the southwest for the next two days,

entering the Timor Sea and passing some 50 km west of Bathurst Island on the morning of December 24. Shortly thereafter Tracy made an abrupt change in direction and headed southeast, directly toward Darwin.

At midnight Tracy was located just off Point Charles or 30 km WNW of Darwin, moving to the ESE at approximately 8 km/hr (see Fig. 1) . Landfall (passage of the leading eye wall) is believed to have occurred somewhere between East Point and Nightcliff shortly after 0300 hours on the 25th, the eye passing directly over the Royal Australian Air Force (RAAF) Base and Darwin International Airport

(see Fig. 2). Extreme winds persisted over the greater Darwin area until approximately 0600 hours. Tracy continued on a southeasterly course, dissipating over Arnhem Land during the remainder of the

25th.

2.2 WIND SPEEDS

The only detailed wind speed record obtained during the passage of Tracy was that of the Bureau of Meteorology at the airport. The Dines anemometer was mounted on a 10-meter mast and the recorded

gusts correspond to an averaging period of from 2 to 3 seconds. Gusts of 12 m/s were recorded over

the period 1300 to 1930 hours on the 24th with a steady increase up to 24 m/s at midnight. Gusts

of 40 m/s were being recorded at 0200 hours on the 25th and the peak recorded gust of approximately

57 m/s occurred just after 0300 hours. At this point, the anemometer was struck by flying debris and

no further wind speed measurements were obtained.

Estimates of maximum wind speeds (3-second gust at 10 meters) in Cyclone Tracy range from 60 to 75 m/s. Based upon peak departures from the mean just prior to failure of the airport anemometer and the observed passage of the eye wall shortly thereafter, a maximum speed of 70 m/s seems reasonable.

Several observers are of the opinion that wind speeds on the rear edge of the eye exceedeS those on

the leading edge. There is no way to confirm this, but many structures apparently suffered the heaviest

damage just after passage of the eye. The average speed of passage is believed to have been 6 to 8

V-22
km/hr. Winds in excess of 50 m/s extended outward 10 to 12 km from the center.

2.3 BAROMETRIC PRESSURE

The barograph record obtained by the Bureau of Meteorology dropped below 990 mb at 0030 hours, reached

a minimum of 940-950 mb at 0400 hours and recrossed 990 mb at approximately 0730 hours. This record was obtained at the city office which is located just to the south of the estimated eye path. The diameter of the eye is believed to have been about 10 km, based on radar pictures obtained at 0400 hours while Tracy was passing over Darwin. The diameter of the calm area would have been slightly less than

this since the radar image represents the location of the rain bands.

2.4 STORM SURGE

The maximum surge level (departure from predicted tide) recorded at Darwin Harbor was 1.6 meters at 0215 hours on December 25 [1]^. The maximum predicted tide of 5.3 meters occurred at 0116 hours [2]. Since the astronomical tidal range at Darwin is approximately 8 meters, the storm surge caused very

little structural damage. The eye of Cyclone Tracy passed to the north of Darwin Harbor, thus creating

higher surge levels at the northern beaches due to the clockwise circulation. It is estimated that a maximum water level of 9 meters (8 meters above town datum) occurred at Casuarina Beach [1]. This level includes wave effects and is approximately 4 meters above predicted high tide. Had Cyclone Tracy

coincided with the highest astronomical tide, some of the lower sections of Darwin would have been

flooded to a depth of approximately 3 meters.

2.5 COMPARISON WITH OTHER INTENSE TROPICAL STORMS

Whatever the maximum wind speeds. Cyclone Tracy must be ranked as an extremely intense tropical storm

and compares with two notable U.S. hurricanes; Hurricane Camille which struck the Mississippi Gulf

Coast in 1969 and Hurricane Celia which caused heavy damage to Corpus Christi, Texas in 1970 [3, 4, 5].

Significant characteristics of these storms are listed below:

Tracy Camille Celia

(1969) (1970)

Central pressure (mb) 940-950 905-915 940-950

Maximum wind speed (m/s) 60-75 75-85 70-75

Diameter of eye (km) 8-10 8-10 20-30

Radius of winds > 50 m/s (km) 10-12 30-50 25-30

Forward speed (km/hr) 6-8 20-25 20-25

Maximum storm surge (meters) 1.6 7.4* 2.8*

(*) Includes astronomical tide.

Other notable cyclones affecting Darwin occurred on January 7, 1897 and on March 10, 1937 [6]. The

1897 event is believed to have been the most intense cyclone to strike Darwin prior to Tracy with

extensive damage to the city and trees either flattened or stripped of branches. Twenty-eight

deaths were recorded and a minimum barometric pressure of 950 mb was observed at Point Charles

Lighthouse. As with Tracy, the direction of travel was SE. The 1937 cyclone approached from the NW

and caused very rough seas with waves breaking over the cliffs at Fannie Bay and Emery Point. Wind

figures in brackets indicate literature references

V-23
gusts of 44 m/s and a minimum pressure of 975 mb were recorded. Most trees were stripped of branches.

3.0 TYPES OF BUILDINGS AND DISTRIBUTION OF DAMAGE

3.1 GENERAL

All areas of Greater Darwin suffered substantial wind damage, the severity depending upon the type

of construction and degree of wind exposure. The purpose of this section is not to describe the damage in detail, which will be covered in subsequent sections, but rather to describe the predominant

types of buildings by subdivision and the general distribution of damage.

3.2 TYPES OF BUILDINGS BY SUBDIVISION

The following tabulations are necessarily subjective and are intended to convey a general description

of types of buildings located in the various subdivisions of Darwin. See Figure 2 for subdivision

locations

Darwin City : Commercial buildings up to three stories, government administration buildings, city office buildings and several historic buildings constructed prior to 1900. Three recently constructed

reinforced concrete buildings of 10 and 11 stories, one 8-story load-bearing brick masonry building, several apartment buildings of two and three stories. Some industrial buildings and petroleum storage

tanks in the railway area along Francis Bay. Several transit sheds and a power plant next to Darwin

Harbor. Very few single-family dwellings.

Larrakeyah : Military area on Emery and Elliot Points, Darwin Hospital, one-and two-story commercial

buildings, apartment buildings and single-family dwellings.

Stuart Park : Railway workshops, some light industrial buildings, several apartment buildings and single-family dwellings.

Fannie Bay and Parap : Single-story commercial buildings, primarily single-family dwellings and some apartment buildings of two and three stories along Fannie Bay. Darwin High School.

Winnellie : Large steel-framed warehouses and multiple-bay industrial buildings.

Narrows : Single-family dwellings and apartment buildings.

Ludmilla : Single-story commercial buildings and single-family dwellings.

Coconut Grove and Nightcliff : Single-family dwellings, apartment buildings up to 4 stories and some single-story commercial buildings.

Northern Suburbs : Includes all subdivisions north of airport and east of Nightcliff. Primarily single

family dwellings, schools, shopping centers and a few apartment buildings.

Darwin International Airport : Terminal building, aircraft hangars, RAAF operations buildings and military housing adjacent to Narrows and Ludmilla subdivisions.
3.3 DISTRIBUTION OF DAMAGES

Shortly after the passage of Cyclone Tracy, a survey of damage to houses in Darwin was carried out by

the Northern Territory Housing Connnission [7]. Approximately 85 percent of the houses in Darwin

(estimated at 8,000 units) were surveyed, the subdivisions of Tiwi and Anula (see Figure 2) being

excluded because many of these houses were still under construction and the destruction in these subdivisions was very close to 100 percent. RAAF housing was also excluded from this survey. The

results are as follows:

Subdivision Destroyed Intact

(Percent of houses in each subdivision)

Darwin City 38 9 Larrakeyah 29 4 Stuart Park 24 10 Fannie Bay and Parap 33 11 Narrows 5 29 Ludmilla 32 6 Nightcliff 54 4 Mxllner 33 3 Rapid Creek 37 6 Jingili 51 7 Moil 75 2 Alawa 57 3 Wagaman 72 2 Nakara 97 0 Wanguri 86 0

It is fairly obvious from the above tabulation that the major damage occurred in the northern suburbs

and that Narrows and, to some extent, Stuart Park suffered less damage to housing than did most other

areas of Darwin. The Narrows area runs along the upper reaches of Ludmilla Creek and is at a lower elevation than adjacent subdivisions. Stuart Park is located on fairly irregular terrain and while

many buildings were directly exposed to the wind, several sections were protected by hills and ridges.

The northern suburbs, on the other hand, had very little protection, particularly after the surrounding

trees had been stripped of their branches. The entire area rises gently from Casuarina Beach north of Nightcliff and thus experienced scant reduction in wind speeds due to terrain roughness. It is quite

likely that Winnellie, Stuart Park, Larrakeyah and Darwin City did benefit from terrain effects,

particularly in the early stages of the storm passage.

Because of the different styles of construction and changes in design criteria and construction materials

that accompanied the growth of Darwin, it is not possible to deduce local maximum wind speeds on the basis of housing damage alone. A comprehensive study of simple structures such as road signs and flagpoles which showed no evidence of missile impact provided valuable information on extreme speeds and

wind direction [8] . Based on this study the following description of wind conditions during the passage of Tracy was deduced :

.

Location Maximum Speeds Range of Direction

Darwin City & Larrakeyah 50-60 NE to SSE

Fannie Bay & Parap 55-65 NE to S Winnellie & Narrows 55-65 NE to SSW

Airport 60-70 NE to SW Nightcliff 60-70 NNW to SW Northern Suburbs 65-75 NW to WSW

V-25
4.0 PERFORMANCE OF BUILDINGS

4.1 HOUSES

4.1.1 Background

The vast majority of houses in Darwin were constructed for the Department of Housing and Construction

or for the Northern Territory Housing Commission. There are some privately constructed houses and several prefabricated housing systems which were introduced within the past two or three years. The design of housing was largely based upon intuition and experience with very little attention given to the rational estimation of wind loads and load capacity. An exception to this was the newer prefabricated housing systems whose designs were subjected to a careful review prior to acceptance by local authorities. Design wind speeds for Darwin varied somewhat through the years and included adjustments for type of exposure, expected life of structure and height above ground. Design speeds

of 45 to 50 m/s were used from 1951 to 1972 at which time the new code provisions for wind loads

(AS CA34.2) were adopted [9]. This code, subsequently re-issued as AS 1170 Part 2-1973 specifies a basic design speed of 56.4 m/s for Darwin and allows reductions for terrain roughness. Following the adoption of AS CA34.2, a category 3 (mean velocity profile described by a power law with an exponent of approximately 0.25) terrain was assumed with a corresponding basic design speed of 39.4 m/s. Hindsight suggests that category 2 terrain (exponent of 0.15) or a design speed of 56.4 m/s would have been

more appropriate.

Of approximately 7,000 houses surveyed, more than 50 percent were damaged beyond repair and only some

400 units could be considered to be intact with minor damage to roofing, wall cladding and windows. Of those considered to be repairable, the vast majority suffered substantial damage to roofing. The following sections describe some of the more common housing systems and their performance during Cyclone

Tracy.

4.1.2 High-Set Houses

By far the most popular style of house in Darwin is the so-called "high-set" house constructed for the Department of Housing and Construction. Although there are several variations the basic unit consists of an elevated platform and light timber frame with a gable roof (see Figure 3) . The plan

dimensions are approximately 7 x 16 meters with the platform some 2.5 meters above ground. Prior to 1968 galvanized steel pipe columns with cross-bracing were used to support the platforms (see Figure 4)

These were superceded by concrete columns on simple spread footings and the cross-bracing was replaced

by concrete block, shear walls installed between 4 adjacent columns, the resulting enclosure serving as

a laundry room or storage space.

The frames of the older high-set houses were constructed of 2x4 in. (nominal) wood studs with let-in diagonal bracing and the 6 mm (1/4-inch) asbestos cement wall cladding was not counted on to provide

principal resistance to racking. In more recent construction the let-in bracing was eliminated,

the stud sizes were reduced to 2 x 3-in. (nominal) and both the corrugated galvanized iron (CGI) roof cladding and asbestos cement wall cladding were assumed to provide sufficient resistance

to racking in the roof and wall planes. Following Cyclone Althea which struck Townsville in 1971,

certain changes were introduced to improve the wind resistance of new housing in tropical areas. These changes included the addition of steel rods or "cyclone bolts" extending from the roof trusses

V-26
down through the walls to the underside of the floor joists. Nails were replaced by self -drilling

wood screws with neoprene washers for the attachment of CGI cladding and reinforced concrete bond beams were added to the concrete block shear walls. Another important feature of the high-set houses, and most

other houses in Darwin for that matter, is the widespread use of prefabricated wall panels containing

jalousie windows and ventilation louvers. The frames and mullions in these units are extruded

aluminum sections.

The high-set houses performed reasonably well in the older subdivisions such as Fannie Bay and Larrakeyah where there was some protection provided by other buildings and trees and where the wood

frames contained diagonal bracing. Many of these older houses were roofed with corrugated asbestos

cement sheets. In the northern suburbs, however, the new high-set houses suffered extensive damage.

The failure of upwind buildings created a source of missiles which undoubtedly played a major role in the widespread damage observed in the northern suburbs. This tended to be a self-sustaining effect

as other structures failed downwind. The sequence of failure is believed to have been loss of roof cladding and purlins, followed by collapse of roof trusses and large unsupported walls, particularly

in the living rooms. The racking strength of the walls was reduced by missiles fracturing the asbestos

cement cladding and these sheets completely failed once racking deformations occurred. In many cases

the entire wood frame was removed from the elevated platform (see Figure 5) . The piers and platforms

generally remained Intact, but there were cases where the entire structure collapsed or where the piers

were racked following failure of the concrete masonry shear walls due to overturning (see Figure 6)

.

No failures of the older steel pipe columns were observed (see Figure 4) . A typical connection between

the concrete pier and floor beam is shown in Figure 7. Note that the floor joists are toenailed to the nailing strips which are attached to the steel beams by "Ramset" studs. Some high-set houses

used metal cladding on the endwalls as seen in Figure 6 and this generally performed very well, being

much more resistant to missile impact than the asbestos cement cladding. The cyclone bolts which

were installed in all houses built after mid-1972 (see Figure 8) do not appear to have made any significant difference in performance since the main problem was a lack of racking strength once the asbestos cement cladding had failed. Many instances of endwall failure were observed (see Figure 9), suggesting inadequate corner connections and roof-wall connections at the gable ends. The performance

of roof cladding is dealt with in a separate section of this report.

4.1.3 Conventional Single-Story Houses

Houses constructed for the Northern Territory Housing Commission were, for the most part, single-story

slab-on-grade structures. There were several variations of this system but the cavity wall with bond

beams was by far the most common. A typical cavity wall is shown in Figure 10. In several cases

bond beams were provided only along the sidewalls and were inadequately tied through the walls to the footings (see Figure 11) . This system generally performed very poorly with many cases observed in which the entire roof structure was removed intact and the ties between inner and outer leafs or wythes (each continuous vertical section of wall one masonry unit in thickness) of the cavity walls

appeared to be inadequate. Both brick and concrete block were used for the exterior wythe. A few of

the newer houses were of timber frame construction with brick veneer on the endwalls. Overall performance

of this type of construction was generally poor. Some units in Wanguri and Anula, under construction

at the time of Cyclone Tracy, utilized precast-tiltup concrete panels and performed exceptionally

well.

V-27
4.1.4 Prefabricated Houses

Of the several prefabricated housing systems used in Darwin, only two were investigated in some detail

and are described herein. The first is a high-set house with the same general floor plan and overall dimensions as the timber high-sets constructed for the Department of Housing and Construction (see Figure 12). All components are of precast concrete with welded and grouted beam-column connections. Columns are 200 x 200 mm in cross-section and the beams are 300 mm deep and 75 mm wide. The floor and

roof panels are approximately 50 mm thick except for the edges which are 100 mm thick. A screed course

is placed over the floor panels and the roof panels are bolted into sockets provided in the top edge

of the beams. Wall panels lock into slots provided in the beams and columns. Conventional aluminum

window frames, jalousie windows and metal louvers are also used in these houses. A conventional

truss-purlin-CGI roof is placed over the ceiling panels, the timber plate being bolted down with the precast ceiling panels. This system would have performed very well had it not been for the fact that anchor bolts were installed only along the perimeter of the roof, even though the interior ceiling

beams contained bolt sockets (see Figure 13) . The result was that many of the panels were lifted from

the interior beams, causing the perimeter anchors to fail.

Some of the ceiling panels shifted enough to drop off of their supporting beams, causing damage to wall and floor panels. In some cases the ceiling panels were flipped over and fractured adjacent

panels (see Figure 14) . This occurred after loss of the CGI roof cladding and roof trusses. One house had obviously received a heavy missile impact, causing failure of a second story corner column. The beam-column connection detail revealed by this failure is shown in Figure 15

The second prefabricated system inspected is a single-story dwelling with a gable roof as shown in Figure 16. The walls are sandwich panels of 6 mm asbestos cement with a polyurethane core. The panels are framed with steel "H" sections (two channels back to back) and the roof system consists of welded, light-gage steel members. Roofing varied from structure to structure, the type shown here being "Decramastic , " a steel sheeting stamped into the shape of tiles and coated with an asphaltic

compound. Sandwich panels are also used for the ceiling, the channel framing also serving as the bottom chords of the trusses. This detail allowed the ceiling to act very effectively as a diaphragm.

Although there were some window failures and substantial loss of roofing, this system performed

exceptionally well. The walls and roof trusses exhibited no permanent deformations and the resistance

of the sandwich panels to missiles was far superior to the ordinary asbestos cement cladding used on

the high-set houses.

4.1.5 Privately Constructed Houses

Several privately constructed houses performed well with damage being limited to loss of CGI roof cladding or tiles and some broken windows. The house shown in Figure 17 is located in Nakara and has

a welded steel frame with brick veneer. Note that the roof tiles have been replaced since the storm. Commercially produced timber connectors were used in many of these houses to attach purlins and trusses

V-28
A. 2 SCHOOL BUILDINGS

4.2.1 General

Several school buildings were constructed in Darwin over the past two years, particularly primary

schools in the northern suburbs. Time did not permit a detailed study of all school buildings, but it

is generally agreed that most of these buildings performed very well. Darwin High School served as

the main center for recovery operations and provided temporary shelter for victims of Tracy while

evacuation plans were being formulated. Design wind speeds for the newer schools in Darwin are not known to the author, but were probably between 50 and 55 m/s. It is important to note that most of these structures would have served well as community shelters during the passage of Tracy. The buildings described in this section were quite closely inspected and are believed to be representative

of recent school construction in Darwin,

4.2.2 Millner Primary School

This complex consists of an auditorium and two-story classroom buildings connected by covered walkways.

Construction is steel frame with concrete and brick masonry walls and roofing is steel pan-type sheet

with hidden locking clips. The entire auditorium roof and most of the cladding over the covered

walkways failed as did some portions of the roofing on the classroom buildings (see Figure 18) , It

is believed that the loss of roofing followed the failure of window panels on the windward side of

the buildings. Two large infill masonry walls were blown in, falling onto the stairway leading to second-floor classrooms.

Of particular interest at this school were the temporary classrooms shown in Figure 19. Construction

is cold-formed steel frame and steel cladding. They suffered a number of missile impacts and some

loss of windows, but remained intact. The anchorage system depended solely on the dead weight of the concrete piers and footings and performed very well. One case of partial anchorage failure due to overturning of the piers was observed. Again, these units would have provided adequate refuge.

4.2.3 Darwin Community College

Darwin Community College is located just west of Nakara and was one of the more interesting complexes

inspected because of some rather unusual structural failures. The main classroom building is a twostory reinforced concrete structure with masonry spandrel walls. The roof system consists of pantype steel sheet carried by bar joists which are bolted to the column tops. A view of one-half of

the north or windward wall is shown in Figure 20 and it is obvious that cracking has occurred in all

of the columns just above the level of the second floor. The other half of the windward face of the

1 building is shown in Figure 21. The roof structure has been folded over the exposed interior wall,

and the second-story spandrel wall and several columns have failed. It was originally believed that

the column failures were due to a poor splice detail, i.e., insufficient bar lap and column ties.

However, a closer inspection revealed a quite different mode of failure. Following loss of the

windows and/or spandrel walls the bottom chords of the bar joists failed in compression due to the

high uplift pressures on the roof cladding and absence of any lateral bracing of the bottom chords.

This, in turn, displaced the column tops and caused them to fail in combined bending and tension.

This sequence of events was quite clear, both in the failed section and in the portion which was

still standing.

V-29
A number of failures of masonry walls were observed in the buildings housing the metal and woodworking

shops of Darwin Community College. These buildings have structural steel frames and concrete block walls as seen in Figure 22. The walls were unreinforced and are believed to have failed by direct

wind loading since there was no source of debris upwind of these structures. There was no apparent

damage to the steel frames.

Damage to student housing, located downwind of the main classroom building, is covered in a subsequent

section.

4.3 APARTMENT BUILDINGS

No damage statistics on apartment buildings were available at the time the author visited Darwin. However, it was quite obvious that apartment buildings performed considerably better than did houses.

It was observed that a substantial number of the people remaining in Darwin after Tracy were being

lodged in apartments and that the highest density of apartments is in the subdivisions of Stuart

Park, Larrakeyah and Darwin City. These subdivisions are believed to have had slightly lower wind

speeds than did other parts of the city, but the relatively good performance of this class of structure

can be largely attributed to better structural detailing. These buildings are typically two or three stories with reinforced concrete floor slabs and loadbearing walls of brick or concrete masonry.

Roof decks are either steel purlins and CGI cladding or reinforced concrete. Figure 23 shows typical

units in Darwin City.

Structural failures generally involved loss of roof decks and, in a few cases, collapse of masonry

walls on the top floors following loss or buckling of purlins. Most roof failures are believed to have been initiated by breakage of windows or failure of doors on windward faces. No cases of primary structural damage were observed in apartment buildings with reinforced concrete roof decks. The building shown in Figure 24 is located in Millner, very close to the center of estimated storm track. Damage was limited to some broken windows, loss of an infill wall in the stairwell and partial failure

of an equipment enclosure on the roof.

4.4 PUBLIC BUILDINGS

4.4.1 Government Office Blocks

These buildings are concentrated in the SE portion of Darwin City and are typically two and three story structures with steel or reinforced concrete frames. Design wind speeds are believed to have been in the range 49 to 58 m/s and maximum speeds in this area during the passage of Tracy are estimated

at 50 to 60 m/s. Damage was generally limited to broken windows and partial loss of roofing. The amount of debris available for missile damage was markedly less than in other parts of Darwin. A very interesting observation was the effectiveness of sunscreens in protecting windows from missiles. Note the impact points in Figure 25. It is also possible that the sunscreens served to reduce the wind loading on cladding panels and windows.

V-30
4.4.2 Darwin Hospital

Darwin Hospital is located in Larrakeyah adjacent to the military area on Emery Point and was designed

for a wind speed of 58 m/s. It is a 3-story reinforced concrete structure with sunscreens (see Figure 26) and has a relatively clear wind exposure. This building performed very well and continued

to function during the storm, switching over to emergency power when the main distribution system

failed. Maximum wind speeds in this area were probably in the range 55-60 m/s. As with the city area, relatively small amounts of debris were available to cause missile damage.

4.4.3 Historic Buildings

Several buildings constructed prior to 1900 have been preserved as historic buildings and are of interest because of their previous exposure to intense cyclones. These buildings are all quite

similar in physical size and type of construction, having timber roof trusses and load-bearing walls

of native stone approximately 400 mm thick. Most of these buildings suffered extensive damage in the cyclones of 1897 and 1937. Typical damage is shown in Figure 27.

4.5 TALL BUILDINGS

Several new hotels and office buildings have been constructed in Darwin since 1970. These include

the 10-story MLC Building and the 11-story Travelodge (Figures 28 & 29) in the Darwin City area. Both are of reinforced concrete flat plate construction with concrete shear walls. Most wind damage

was superficial, being limited mainly to loss of pan-type roof cladding and broken windows. There

proved to be no substance to the rumor that a large missile caused the partial failure of the brick

and concrete block endwalls of the Travelodge (see Figure 29) . An identical failure occurred at the same level on the opposite end, undoubtedly due to extremely low surface pressures downstream of the separation points. The change in wind direction in this area would have been sufficient to create

identical pressure distributions at each end of the building. The design speeds are not known, but inspection teams reported no evidence of cracking or plastic action in either structural system.

Occupants did report noticeable oscillations during the storm passage. Water damage was extensive in these and other tall buildings following loss of roof cladding and glass breakage.

4.6 INDUSTRIAL BUILDINGS

Most of the industrial buildings are located in Winnellie and Stuart Park. Maximum speeds in Winnellie

are believed to have been 55-60 m/s and slightly lower in Stuart Park. Structural details and materials

used in most of these buildings are very similar to U.S. practice; i.e., cold-formed rafter sections,

purlins and girts with 26 ga. CGI cladding. Wind bracing was usually limited to diagonal rods in

one or two wall panels, the cladding being counted on for most of the in-plane stiffness. Most

builidngs had a concrete masonry infill wall and small office area at one end, the other end consisting

of girts and CGI wall cladding. In many cases the purlins were tied directly into the concrete

masonry wall, the end rafter being eliminated.

v-31
The overall performance of industrial buildings was good, but damage to these light-frame structures

in some areas of Winnellie was heavy. It was quite apparent that these buildings had little variation

in ultimate load capacity; if one building of a particular type failed, most others of that same type also failed. It is quite likely that failure of rollup doors, windows, and flashing led to gross cladding failure because of sudden increases in internal pressure. Many cases of purlins being

buckled upward were observed and this often led to failure of the masonry endwalls to which they were

anchored. In other cases the roof cladding was removed by direct pullout of the fasteners, but was more often due to fatigue failure in the material adjacent to the fasteners. This often resulted in total collapse, particularly where the cladding provided the only source of racking strength in the roof planes. A typical failure is shown in Figure 30. Several buildings suffered complete loss of roof and wall cladding without the frame collapsing. In these cases, however, a conventional system

of diagonal bracing had been installed in the roof and wall planes.

The buildings shown in Figure 31 are located in Stuart Park and, excepting some loss of wall cladding,

they performed very well. These buildings have 22 ga. CGI cladding with 16 mm washers under the fastener heads. This detail, in conjunction with the ridge ventilators undoubtedly had much to do with the overall good performance. Ridge ventilators effectively lower the internal pressure and reduce the uplift load on the roof. This observation has been made in previous surveys of wind

damage.

5.0 PERFORMANCE OF MATERIALS AND STRUCTURAL SYSTEMS

5.1 GENERAL

The behavior of certain materials and subsystems under repeated loading and attack by windborne

debris was an important factor in the overall performance of buildings in Darwin and deserves special

attention in this report. It raises serious questions as to the validity and application of design

criteria based solely upon static load tests of builidng elements and structural subsystems. Although

many of the failures observed can be explained by extreme loading, poor workmanship or poor design,

there was overwhelming evidence of consistently poor performance of certain materials and structural

subsystems during the passage of Cyclone Tracy. These are discussed in the following sections.

5.2 CORRUGATED GALVANIZED IRON CLADDING (CGI)

As with other tropical regions of Australia and, in fact, throughout the world, CGI cladding is widely used in Darwin. There are many variations in the sheet profiles and types of fasteners, by

far the most common being a sinusoidal cross section with self-drilling, self-tapping fasfeners equipped

with a neoprene sealing washer. Immediately following Cyclone Tracy, laboratory tests and field inspections of single-family dwellings and industrial buildings were carried out by the Housing

Research Branch, Department of Housing and Construction to establish the reasons for the widespread

failure of CGI in Darwin [10]. These investigations are continuing, but preliminary findings indicate

that low-cycle fatigue was the predominant failure mode. Repeated load tests carried out at approximately

30 percent of the ultimate static load capacity produced failures after 2000 to 3000 cycles. A typical failure is shown in Figure 32. Data obtained in recent full-scale investigations of wind

V-32
loading [11] suggest that this combination of load and number of cycles would have been exceeded in slightly less than one hour for the wind speeds occurring in the northern suburbs. This suggests

that the fastening system in current use (typically 15 nmi diameter screw heads with screws installed

at every second or third crest) will have to be radically altered. Possible improvements include:

a) Installation of fasteners in valleys rather than on ridges

b) Larger washer or screwhead diameter

c) Battens placed over the CGI and bolted into the purlins

5.3 ASBESTOS CEMENT CLADDING

The asbestos cement cladding used on most of the high-set houses and some of the single-story dwellings

proved to be very susceptible to missile impact. This was a major factor in the poor performance of those structures which depended entirely on cladding for racking strength. Static load tests did not reveal this deficiency and it is likely that performance would have been substantially better had a tougher material such as plywood been used. It is interesting to note the drastic improvement in impact resistance where the same material is used in sandwich panel construction (see Figure 33)

.

Conventional nailing of asbestos cement sheets subjected to repeated racking loads deserves additional

study.

5.4 MASONRY CAVITY WALLS

Many failures of masonry cavity walls were observed. Lack of bond beams and lateral loading due to purlin buckling were often the cause of these failures. However, a number of cases were observed

where the crossties were either inadequate or improperly installed and the inner and outer wythes did

not act together in resisting lateral loading. Since Australian practice with respect to masonry

cavity wall construction is quite similar to that here in the U.S., this high incidence of failure is cause for concern. Most load tests on masonry walls have been carried out with vertical compression

loads applied to simulate deadload. This load condition overestimates resistance to racking and

lateral loads when uplift forces are sufficient to offset the deadload. In many cases observed in Darwin, it is likely that the uplift forces transmitted through the purlins were sufficiently high to place the entire wall in tension.

6.0 UTILITIES

6.1 POWER GENERATION AND DISTRIBUTION

6.1.1 Stokes Hill Power Station

The Stokes Hill Power Station supplies the entire greater Darwin area with electrical power and is located at the extreme southern tip of the city. It does not function as part of a grid system and

was thus essential to recovery operations. Damage was primarily due to water entering switch gear

following loss of roof cladding. A trawler was washed ashore directly over the cooling water outlet

structure, but this did not require any immediate repair work. The one reinforced concrete and three

steel stacks at the station suffered no apparent damage and the station was placed in partial operation

within two days after the storm.

V-33
6.1.2 Electrical Distribution

With the exception of the main transmission line running across Darwin to the northeast (conventional

steel space-truss towers), most distribution lines were heavily damaged and required complete replacement. Most power poles were either steel lattice structures or prestressed concrete and failures were generally due to accumulation of CGI cladding on the conductors and poles. Many lamp standards also

failed due to debris loading (see Figure 34) . Primary distribution lines were put back in service within one to two weeks after the storm.

6.2 COMMUNICATIONS

Darwin is connected through a microwave link to the national communications network and most lines in the city are buried. The microwave tower and related equipment in Darwin City experienced only light

damage and telephone communications were thus partially restored within hours after the passage of Tracy.

6.3 WATER AND SEWERAGE SYSTEMS

Damage to water and sewerage systems was minimal, primarily due to the fact that there was very

little erosion due to storm surge and wave action. Most of the disruption was caused by lack of electrical power at pumping stations. Of the half-dozen or so large elevated water tanks in Darwin, none suffered any noticeable structural damage. The extremely heavy damage in the northern suburbs required that most of the water mains in that area be shut down until individual service taps could

be closed.

7.0 CONCLUSIONS

Cyclone Tracy was an extremely intense storm which passed directly over the city of Darwin. The unusually slow forward speed (6-8 km/hr) and long duration of extreme winds is believed to have been a key factor in the extensive damage to buildings, housing in particular, and revealed weaknesses in structural systems that would quite likely have resisted an equally intense storm of shorter duration. While certain construction practices and building materials used in Darwin are not commonly found in

the U.S., much of what happened should be given proper attention in future revisions of wind load design criteria and indicates a need for additional research in the following areas:

a) Behavior of CGI cladding and fastening systems under repeated loads

b) Resistance of brittle materials such as asbestos cement and fiber board to missile impact

c) Racking strength of wall panels subjected to uplift loads

d) Behavior of housing systems subjected to unusually high internal pressures resulting from failure of doors and windows

e) Performance of brick and concrete masonry cavity walls for various conditions of edge

support and load combinations, including uplift.

V-34
8 . 0 ACKNOWLEDGEMENTS

The author wishes to express his special thanks to Dr. George R. Walker, Senior Lecturer, Department

of Engineering, James Cook University of North Queensland, for making most of the technical data contained in this report available and for providing invaluable assistance and guidance during the brief period available for the inspection of damage. This report is the result of the author spending

several days on temporary assignment with the Department of Housing and Construction, and the assistance

of members of the engineering staff, both in Darwin and Melbourne, is gratefully acknowledged.

9.0 REFERENCES

1. Wilkinson, F.L., "Interim Report - Cyclone Tracy Storm Surge at Darwin, 25 December,

1974," Report No. 57, Department of Housing and Construction, Maritime Works Section,

Melbourne, Vic, January 1975.

2. "The Port of Darwin Tide Tables - 1974," Compiled at Naval Headquarters, Darwin, NT.

3. Thorn, H.C.S. and Marshall, R.D., "Wind and Surge Damage Due to Hurricane Camille."

Journal of the Waterways, Harbors and Coastal Engineering Division, ASCE, Vol. 97, No. WW2, May 1971.

4. DeAngelis, R.M. and Nelson, E.R. , "Hurricane Camille, August 5-22." Climatological

Data, National Summary, ESSA, Vol. 20, No. 8, Washington, D.C. 1969.

5. Condon, C.R. "North Atlantic Tropical Cyclones - 1970" Climatological Data, National

Summary, NCAA, Vol. 21, No. 13, Washington, D.C, 1970.

6. "Is Darwin Prepared for a Cyclone?" Report by the Cyclone Subcommittee of the Institution of Engineers, Australia - Darwin Group, April 1974.

7. Walker, G.R., "Report on Cyclone Tracy - Effect on Buildings." Vol. 1, Department of Housing and Construction, Melbourne, Vic, March 1975.

8. Halpern Click Pty. Ltd., "Indications of Wind Velocities During Cyclone Tracy From the Behavior of Simple Structures." Report on Cyclone Tracy-Effect on Buildings, Vol. 2, Appendix 1, Department of Housing and Construction, Melbourne, Vic, March 1975.

9. SAA Loading Code, Part II - Wind Forces (AS CA34, Part II - 1971). Standards Association

of Australia, Sydney, NSW, 1971.

10. Beck, V.R., "Appraisal of Roof Cladding Under Dynamic Wind Loading - Cyclone Tracy,

Darwin, 1974." Technical Report No. 1 (Preliminary), Housing Research Branch,

Department of Housing and Construction, Melbourne, Vic, February 1975.

11. Marshall, R.D., "A Study of Wind Pressures on a Single-Family Dwelling in Model and Full-Scale." NBS Technical Note 852, National Bureau of Standards, Washington, D.C,

40 pages (October 1974) .

V-35
V-36
Beagle Gulf

^ Casuarina r^-^ /~— -/

Beach /^Brinkin^-

/ Nakara'^^^"^""/

'

-, ' Alawa I

I \ / , Wagaman/

'\ V J'laiii 1 Moil / '^^''^i

|Creek^-V' \

. 7 v

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1

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Fig. 2. Approximate cyclone track across Darwin

I

V-37
Fig. 3, Typical high-set house

ig. 4. Pipe columns used in older construction

V-38
Fig, 5. Severe racking and loss of timber deck

V-39
Fig. 7. Typical floor systems and pier anchorage

Fig. 8. Cyclone bolt and rafter connection 1

Ij

I

V-40
Fig. 9. Failure of endwall

V-41
Fig. 11. Concrete bond beam and timber plate

( single- story construction)

V-42

Fig. 15. Beam to column connection - precast concrete high-set

Fig. 16. Prefabricated house with asbestos cement-polyurethane

sandwich panels and steel wall and roof frames

V-44

!
V-45
Fig. 19. Temporary classrooms - Millner Primary School

V-46
Fig. 21. Failure of second story (view immediately to right of Fig. 20)

Fig, 22. Woodworking shop - Darwin Community College

V-47
Fig. 23. Apartment buildings in city area

Fig. 24. Apartment building in Millner

V-48
V-49
Fig, 27. Historic building in Darwin City - hip roof completely

removed

Fig. 28. MLC Building - Darwin City

V-50
Fig. 29 Travelodge - Darwin City (similar damage at opposite end)

Fig. 30. Industrial building - Winnellie

V-51
Fig. 32. Typical fatigue failure of CGI cladding

V-52
V-53
ON THE DAMAGE TO BUILDINGS IN HACHIJOJIMA

CAUSED BY TYPHOON NO. 7513

TATSURO MUROTA

Chief, Structure Section

Structure Division

Building Research Institute

Ministry of Construction, Japan

ABSTRACT

On October 5, 1975 Typhoon No. 7513 hit Hachijojima Island, Tokyo and the maximum

peak gust of 67.8 m/s was observed. The damage to this island was severe and widespread,

damage to buildings, services, crops, trees, electric power lines and telephone lines

occurred. The wind records and the rate of damage to buildings was one of the largest in

Japan

The Building Research Institute investigated the damage to the buildings in Hachijojima Island from 9 to 16, October. This paper describes results of the investigation.

Description of the typhoon and meteorological environment in the Island during the

passage of the typhoon are referred in detail by Soma (1) and therefore are not referred

to herein.

Key Words: Damage; Typhoon; V7ind; Wind Effects.

1. Definition of Tein^as In order to express the extent of damage the following terms and notations will be

used in this paper:

"completely destroyed" = damaged to the extent that the floor area damaged exceedes

70% of the total floor area, or the expense of repairs is more than 50% of the current

price of the building.

"badly damaged" = damaged to the extent that the floor area damaged are 20 to 70%

of the total floor area or the expense of repairs is more than 50% of the current price of

the building.

"rate of completely destroyed houses = R^ = rate of the number of completely destroyed houses to the total number of households in a certain district.

"rate of badly damaged houses" = = rate of the number of badly damaged houses

to the total number of households in a certain district.

"rate of damaged houses; = R^ = sum of the two rates defined above.

2. Description of Damage to Buildings

a) Statistics and Regional Distributions of Damage to Houses

Quoting from the damage statistics published by Hachijo-Cho Town Hall under date

of October 9, damage to houses in every administrative district in Hachijojima Island is

V-54
as shown in Table 1. Concerning the rate of completely destroyed houses, the average was

7.9% and by administrative district the highest rate was 11.4% in Sueyoshi. Concerning

the rate of damaged houses it was 18.6% by administrative district 21.8% in Okaho which

was the largest. It is worthy to note that in Nakanogo district both of these rates were

considerably lower than others.

In each administrative district in Hachijojima there are small sections where scores

of people are living. The rate of damage houses for each section was investigated and

the result is illustrated in Fig. 1, where the rate of damaged houses was classified into

5 steps 10% each.

The maximum rate was 58.1% (25/43, Section No. 6, Obata, Mitsune District) and the

minimum rate was 1.5% (1/64, Sec. No. 14, Sakuradaira, Mitsune Dist.). In addition to

the large difference between these extremes, it was very surprising that these two sections

are located side by side. Similar behavior of damage rates were observed in several other

places. One of the reasons for this inconsistency was the effect of the topography.

b) Relation Between the Rate of Damage and the Maximum Peak Gust

According to Takahashi (1) the amount of damage to structures is proportional to

the wind speed. Ishizaki et al. (2) applied this theory to damage of houses caused by

three severe typhoons: the 2nd and 3rd Miyakoj-ma Typhoon (1966 and 1963) and Isewan

Typhoon (1959) , and plotted the relation between the rates of completely destroyed houses

R^ and maximum peak gusts as shown in Fig. 2. A mark A in this figure was plotted by the

present author for comparison. It will be noted from this figure that the value of R^ 7.9%

in a severe wind of maximum peak gust 67.8 m/s was not exceptional when comparing those

damages caused by the past three typhoons.

3. Features of the Damage to Buildings

a) Wooden Buildings

Wooden buildings are most popular in Hachijojima Island. Most of the houses have a gable roof or a hip roof with a slope of 2/10 to 2.5/10. Their roofings^ almost without

exception, consist of a thin metal plate roof nailed to bed rafters where "bed rafters"

are structural members of square cross-sections and are set on common rafters in a direction

perpendicular to the common rafters.

Damage to the wooden buildings occurred to the roofs and openings on the external walls.

Roof damage in which only the metal roof was removed was rare, in most cases, removal of

the bed rafter, common rafters, purlins, roof posts or ridges occurred together with the

removal of metal roof. Such damage was caused by the imperfection of connections between

roof structural members.

b) Reinforced Concrete Buildings

Damage to reinforced concrete buildings occurred only at openings on the external

walls. Such damage occurred at a hotel in Hachijojima. This hotel was six stories and

suffered severe damage to its windows. From the 2nd to 6th floor, 30% of the windows on windward surface were broken. At the 1st floor of the hotel there was a lobby surrounded

by glass walls and these glass walls had been broken on both of windward and leeward

V-55
sides. This damage was caused by wind generated missiles and also the small flexural

rigidity of the window sashes.

In the design of buildings having many windows on exterior walls, it is necessary to

take the breakdown of window glasses into account, because the breakdown of windows lead

to an increase of interior pressure and for some part of the building the overload may be

caused by this pressure increase.

c) Steel Frame Buildings

Steel frame buildings suffered damage to the finish of the external surfaces.

Most of the damage was precipitated by the structural engineer for not checking the safety

of those finishings into detail. It is necessary that structural engineers recognize that

the wind pressure acts on finishings on external surfaces of buildings.

The other point which is considered to be important, is that several buildings of

flat roofs using steel roof decks suffered serious damage: bodily removal of the roofs.

These roofs had considerably long eaves and roof surfaces were more flexible than other

types of roofs. In wind resistant design of these roofs it is necessary to take the

dynamic effect of pressure fluctuations into account.

4. Concluding Remarks

On October 5, 1975 Typhoon No 7513 hit Hachijojima Island. Maximum peak gust of

67.8 m/s was observed at Hachijojima Weather Station. 7.9% of houses in the island were completely destroyed by the severe wind caused by this typhoon. This damage was one of

the severest in Japan.

Lessons obtained from this disaster are as follows:

1. Light weight roofs suffer damage very easily. When flexural rigidity of the

roof surface is small, like some flat roofs with steel roof decks, it is necessary in

structural design to take the dynamic effect of wind pressure fluctuations into account. Investigations will be necessary concerning this problem.

2. Damage to openings and finishings on external walls was observed on many buildings. Such damage may be a trigger to more serious damage and therefore it is necessary

to recognize the importance of the design of details

References

1. K. Takahashi, et al., Reports of the Damage Caused by Isewan Typhoon, jour. of

Architecture and Building Science, Vol. 75, 1960, pp. 111-115.

2. H. Ishizaki, et al.,On the Damage to Houses Caused by the Third Miyakojima

Typhoon, Annals, Disaster Prevention Institute, Kyoto Univ., No. 13A, 1970, pp. 449-461.

V-56
Fig. 1 Distribution of the rate of damaged houses, Rt

V-57
Fig. 2 Relation between Ilia rate of completely destroyed i;ouG2s, and

the maxir.-.u';: peak gust,

o Isewan Typhoon

K The 2nd Miyakoj ima Typhoon

^ The 3rd Miyakojin-.a Typhoon

A Typhoon No. 7513

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MAXIMUM PEAK GUST

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V-58
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V-59
i
SEISMIC RESPONSE OF REINFORCED CONCRETE HIGHWAY BRIDGES

J. PENZIEN

Professor of Structural Engineering

University of California, Berkeley

W.G. GODDEN

Professor of Civil Engineering

University of California, Berkeley

M.C. CHEN

Assistant Research Engineer

University of California, Berkeley

D. WILLIAMS

Assistant Research Engineer

University of California, Berkeley

K. KAWASHIMA

Assistant Specialist

University of California, Berkeley

ABSTRACT

Presented is a brief progress report of an investigation entitled "An Investigation

of the Effectiveness of Existing Bridge Design Methodology in Providing Adequate Structural

Resistance to Seismic Disturbances" which was initiated in 1971 within the Earthquake

Engineering Research Center, University of California, Berkeley, under the sponsorship of

the U.S. Department of Transportation, Federal Highway Administration.

Key Words: Bridges; Design; Experiments; Nonlinear Response; Reinforced Concrete.

1. Introduction

In the past, numerous highway bridges have suffered extensive damages due to strong

motion earthquakes (1,2). The older bridges, consisting of single or multiple simple

truss or girder spans supported on massive piers and abutments, were particularly vulnerable to the action of strong ground motions. Seismic damages were most commonly caused

by foundation failures resulting from excessive ground deformation and/or loss of stability

and bearing capacity of the foundation soils. As a direct result, the substructures often

tilted, settled, slid, or even overturned; thus severe cracking or complete failure was

often experienced. These large support displacements also caused relative shifting of and

damage to the superstructures, induced failures within the bearing supports, and even

caused spans to fall off their supports. It is significant to note that very little damage

occurred to these older structures as a direct result of structural vibration effects.

Certain types of modern highway bridges may, on the other hand, be quite susceptible

to damage from strong ground vibration effects. This fact became very evident during the

VI-1
San Fernando earthquake of February 9, 1971, when numerous reinforced concrete highway

bridges whidi were designed by the traditional elastic approach, i.e., by the equivalent

static seismic coefficient method similar to that previously adopted for buildings,

suffered severe damages (3,4,5). Because of this experience, it was immediately apparent

that the seismic requirements used were inadequate and that the design methodology should

be critically examined. Action was quickly taken following the earthquake to correct cer- tain design deficiencies (6) ; however, the basic static approach to design still remains

in effect.

Recognizing the urgent need for both theoretical and experimental research related

directly to seismic effects on bridge structures, an investigation entitled "An Investigation of the Effectiveness of Existing Bridge Design Methodology in Providing Adequate

Structural Resistance to Seismic Disturbances" was initiated in 1971 within the Earthquake Engineering Research Center, University of California, Berkeley, under the sponsorship of the U.S. Department of Transportation, Federal Highway Administration. This investigation consists of the following six phases

1. A thorough review of the world's literature on seismic effects on highway bridge

structures including damages to bridges during the San Fernando earthquake of February 9,

1971.

2. An analytical investigation of the dynamic response of long, multiple-span,

highway overcrossings of the type which suffered heavy damages during the 1971 San Fernando

earthquake

3. An analytical investigation of the dynamic response of short, single and multiple

span highway overcrossings of the type which suffered heavy damages during the 1971 San

Fernando Earthquake.

4. Detailed model experiments on a shaking table to provide dynamic response data

similar to prototype behavior which can be used to verify the validity of theoretical

response predictions.

5. Correlation of dynamic response data obtained from shaking table experiments

with theoretical response and modification of analytical procedures as found necessary.

6. Preparation of recommendations for changes in seismic design specifications and

methodology as necessary to provide adequate protection of reinforced concrete highway

bridges against future earthquakes.

Final reports covering Phases 1, 2, and 3 have been published (1,7,8) and similar reports

covering Phases 4 and 5 are now in preparation. Phase 6 of the overall investigation will

be completed during summer and fall of 1975.

The brief progress report presented in this paper concentrates on Phases 4 and 5 with some mention of the results obtained in Phases 2 and 3.

2. Analytical Studies

a) Long Multiple- Span Bridges (7,9,10)

The primary objective of the investigation in Phase 2 was to develop suitable proVI-2
cedures for linear and nonlinear three-dimensional earthquake response analysis of long,

multiple-span, modern highway bridges of the type which suffered heavy damages during the

1971 San Fernando earthquake and to identify and investigate the important parameters

which significantly affect the response of this type of bridge system. To achieve these

objectives linear and nonlinear mathematical models suitable for modelling the dynamic

characteristics of this type of bridge were defined and analytical procedures and computer

programs were developed for determining the seismic response of the complete bridge system

to arbitrary, but prescribed, earthquake excitations.

Linear and nonlinear seismic analyses were performed on three typical major overcrossings using earthquake ground motions of several intensities. These bridges were (1) the South Connector Overcrossing of the Golden State freeway and Antelope Valley freeway

interchange (5/14 South Connector Overcrossing) , (2) the proposed curved Figueroa Street

Undercrossing Connector in the Los Angeles area, and (3) a straightened version of the

Figueroa Street Undercrossing Connector. Various parameter studies were carried out for

these bridges and the results were correlated with the apparent behavior of this type of

structure during the San Fernando earthquake. In particular, the causes of collapse of

the 5/14 South Connector Overcrossing were identified and examined.

Based on the results of this investigation, the following conclusions were deduced:

1. The basic mathematical models, analytical procedures, and computer programs developed provide a rational and effective means for determining the seismic response of

curved and straight, multiple-span, long, highway overcrossings

.

2. The collapse of the 5/14 South Connector Overcrossing during the San Fernando

earthquake was due to the inability of the longitudinal restrainer ties to constrain the

joint separation at expansion joints; thus allowing two central spans to fall off one end

support, resulting in the collapse of the spans along with their common supporting column.

3. The proposed curved Figureoa Street Undercrossing Connector can safely withstand

severe ground shaking of an intensity similar to those experienced during the San Fernando

earthquake; however, some flexural damage is likely to occur at the bases of the columns

supporting the central spans.

4. The strength and distribution of bridge columns in each deck segment between

expansion joints are the primary factors which most significantly affect the seismic response characteristics of the bridge structure. Thus expansion joints in the deck should

be located in a manner such that each substructural segment possesses a reasonably adequate

amount of lateral stiffness.

5. Under severe shaking conditions similar to those experienced during the San

Fernando earthquake, the maximum horizontal acceleration response at the deck level substantially exceeded those values specified by the "Bridge Planning and Design manual,"

California State Division of Highways (5) which was in effect prior to the San Fernando

I earthquake of February 9, 1971.

j 5. Because of their effectiveness in reducing the amplitude of the seismic displace-

, ment response of curved bridges , longitudinal restrainer ties should be provided at expansion joints. They should have sufficiently high strength and ductility to constrain the

VI-3

I
joint separation within the limit of the supporting ledge dimension. Vertical restrainers

should also be provided at expansion joints in order to prevent the lifting of spans from

their supports.

7. For curved bridges, a bigger expansion joint gap appears to be desirable to prevent biased seismic response and to avoid a shake-down situation under severe shaking

conditions

8. The required expansion joint ledge dimension and joint gap can be estimated from

the maximum horizontal displacement obtained from a seismic analysis.

9. Under moderate to severe shaking, the overall seismic response of a bridge

structure is considerably affected by vertical ground accelerations. The effect is particularly noticeable for curved bridges in which coupling exists between vertical and horizontal modes of vibration. Thus the vertical component of ground acceleration should

always be included in a seismic analysis.

10. For a normally proportional bridge structure under moderate to severe shaking,

flexural yielding is more likely to occur at the base of the coliimns. Yielding in columns

should be avoided under moderate shaking but can be permitted to limited levels under

severe shaking conditions. It is therefore important that column reinforcement be sized

and detailed in such a manner that the columns can develop full yield strengths and maintain their yield capacities throughout their specified ranges of ductility. Flexural

yielding should be prevented from developing at the top of columns in order to avoid damage

penetrating the deck.

11. The effect of expansion joint Coulomb friction on high seismic response of the

bridges considered is small.

12. Linear seismic responses analysis provides a reasonable estimate of the maximum

displacement response of a bridge system; however, siibstantial error may result in predicting the internal forces in the structure. Thus it is essential that nonlinear seismic

analyses be carried out for major bridge structures under severe shaking conditions.

b) Short Single or Multiple-Span Bridges (8) The primary objective of the investigation in Phase 3 was to develop suitable

procedures for linear and nonlinear two- and three-dimensional earthquake response analysis of short single or multiple-span highway bridges of the type where soil-structure

interaction effects are important.

For the two-dimensional studies, the bridge system was modelled mathematically

using six different sub-elements (1) finite modelling was used for the backfill soils,

(2) bridge deck, piers, and abutments were modelled using prismatic beam elements, (3) a frictional element was used to model the discontinuous behavior at the interface of

backfill soils and abutments, (4) discontinuous type expansion joint elements were

included, (5) linear spring elements were used to provide flexibility at the vertical soil

boundaries, and (6) the soil foundation flexibilities under columns were determined using

elastic half-space theory. In the non-linear mathematical model, the effects of separation and impact at the interface between abutments and backfills, the yielding at concrete

VI-4
columns and backfill solid and slippage at the expansion joints were taken into consideration.

Parameter studies were first carried out considering a rigid wall backfill soil system

A short, stiff, three-span bridge was then investigated with full soil-structure interaction effects included. It was found from these studies that soil-"tructure interaction

effects must be considered when analyzing the dynamic response of short, stiff, single

or multiple span bridges and it was shown that the mathematical modelling and computer

programs developed provide an effective means of conducting such analyses.

For the three-dimensional studies, the bridge system was modelled by four different

sub-elements (1) the linear elastic three-dimensional beam element was used to model the

bridge girder, pier columns, and pile cap, when necessary, (2) the linear elastic threedimensional solid finite element was used to model backfill soils and abutment walls, (3) a nonlinear three-dimensional friction element was used to simulate separation and impact

between backfills and abutment walls, and (4) linear elastic spring elements having six

degrees of freedom, 3 translational and 3 rotational, were used to model column foundations.

The same bridge used in the above two-dimensional analysis was treated again in a three-dimensional skewed form to study the influence of torsion about a vertical axis.

Parameter studies included effects of skewness, impact and separation between abutment

walls and backfill soils, and type of model (linear vs. nonlinear) used for the backfill

soils. Results of the analyses showed that significant torsional response can occur due

to skewness of the deck coupled with out of phase abutment forces resulting from various

causes

.

3. Experimental Studies (11,12,13)

a) Model Design

A 1/30 true-scale microconcrete model of a simplified symmetrical version of a typical high curved bridge structure was constructed and subjected to appropriately scaled

simulated seismic motions of varying intensity using the 20 ft x 20 ft shaking table at

the U.C. Earthquake Simulator Laborabory. The model consisted of three deck segments,

separated by expansion joints and supported at the abutments and on 3 ft high columns.

The radius of curvature was 9 ft. , the deviation angle was 125° and the total span

approximately 22 ft. In order to gain as much experimental data as possible from the basic model, it was

designed as an assemblage of replaceable components. This procedure has the obvious

advantages of making the structure easily repairable when damaged, of facilitating certain

property changes between test series, and of enabling linear and nonlinear characteristics

to be determined by prior component testing.

In relation to the prototype, the quantities scaled included all bending and tor- sional stiffnesses, self-weight effects and inelastic effects both in the expansion

joint restrainers and in the coliamns. The model was designed so that damage would be

confined to these two locations while the remainder of the deck system remained elastic.

VI-5
Inelastic cyclic bending tests of the column elements have confirmed that the nonlinearity

of well designed prototype reinforced concrete columns was adequately simulated.

The model was weight-distorted; thus deficiency weight in the form of lead blocks

equivalent to 30 times the self-weight of the model (based on the prototype box girder

section) was added to the model. The resulting time ratio was 1/5.5. The lower mode

shapes and frequencies of the scaled prototype structure were consequently reproduced in

the model.

b) Model Response

Three types of simulated earthquake excitation were applied to the model, (1) horizontal excitation alone in the asymmetric longitudinal direction, (2) horizontal excitation alone in the symmetric transverse direction, and (3) horizontal excitation in the

symmetric transverse direction with simultaneous vertical excitation. The response of

the model, primarily the global relative displacements, was measured by means of LVDTs and

recorded in digitized form by the data acquisition system. The response of the model was

also recorded by a high speed movie camera.

In all cases of high intensity excitation, severe damage occurred to the bridge deck

at the zone of the expansion joints. The inclusion of ductile steel ties across the joints

to prevent excessive separation of the girders and subsequent collapse of the structure

did not prevent damage to the expansion joint region. The damage generally considered con- sisted of failure of the shear key and severe shear cracking across the hinge seat. Furthermore, the joint restrainers were subjected to large ductility demands. In the case

of longitudinal excitation, spalling of the adjacent joint faces also occurred. It was caused by the large shear and impacting forces at the joints during the dynamic response

of the structure and served to emphasize the undesirability of such discontinuities in

this type of structure. For these particular test structures the column damage was insignificant

The objective of these tests was to indicate trends of behavior, and also to provide

test data on which to check the validity of computer analyses developed in Phase 2 of the

project. The experimental studies clearly demonstrated the complexity of the 3-dimensional

nonlinear response of curved bridges to earthquakes, some of the problems associated with

the mathematical modelling of the phenomena, and the need for a more careful spatial

arrangement and detailing of expansion joints in bridges when conditions necessitate their

use

4. Correlation of Experimental and Analytical Response

a) Formulation of Analytical Model

The experimental bridge model was idealized by a discrete parameter system consisting of 55 nodal points, 24 three-dimensional curved beam elements representing the superstructure, 24 straight beam elements representing the columns, 2 expansion joint elements,

and 2 boundary elements representing the abutments. The stiffness properties of the

superstructure, columns, and abutments were determined either from the results of static

VI-6
load- deformation tests or by calculation using section and material properties. Since the

expansion joints have many constraints as provided by shear stiffness of support pads,

friction at support interfaces, transverse shear key, etc., their overall constraint to the

girder is difficult to define accurately. Therefore, it was decided to determine this

constraint so that the dynamic properties of the analytical model would match those of

the experiment model. As a first requirement the natural frequencies of the analytical

model should match the frequencies of the experimental model derived from low-amplitude

free vibration and forced harmonic tests. The constraint of the expansion joints was modelled by a set of elastic springs which resist relative translational and rotational

motions between adjacent girders. The stiffnesses of these springs were adjusted so that

the calculated natural frequencies of the analytical model matched the measured natural

frequencies of the experimental model. These frequencies were approximately 5.0, 6.6,

and 9.5 for the longitudinal, transverse, and vertical normal modes, respectively.

This procedure established the basic linear analytical model which matched the low-amplitude vibration characteristics of the experimental model.

b) Experimental Tests Chosen for Correlation Purposes

Six basic tests were carried out in the experimental investigation, i.e.. Phase 4 of the overall research program. Three of these designated as Hi, H3, and HV2 were

selected for correlation purposes. The primary features of these tests are shown in Table

1. Tests HI was carried out using only horizontal excitation in the transverse direction

of low intensity (peak acceleration - O.llg). During this test the expansion joint tie

bars remained elastic, no collisions occurred in the expansion joints, and no failures

occurred to the shear keys. Test H3 used high intensity (peak acceleration = 0.50 g)

excitation in the horizontal transverse direction. In this test, yielding of tie bars

occurred, multiple collisions took place in the expansion joints, and failure of a shear

key resulted. Tests HV2 was carried out using both horizontal transverse and vertical

excitations having peak accelerations equal to 0.47 g and 0.27 g, respectively. In this

test yielding of tie bars occurred and multiple collisions took place in the expansion

joints. No failure of shear key occurred in this test.

c) Correlation of Experimental Results with Linear Analytical Model

The linear analytical model described above was used in the correlation of

experimental results obtained in Test Hi. A damping ratio of 6% of critical was used

for the analytical model which agreed with the measured damping ratio of the experimental

model. Upon comparing time histories of displacement of the analytical model with

experimental results , it was found that the overall stiffness of the analytical model

was greater than that of the experimental model. Further adjustments of stiffnesses in

the expansion joints were then made to minimize the observed phase lag discrepancies in

response. The analytical time histories of displacement now agreed quite well with the

experimental results except that the calculated responses showed symmetric motions relative to the outward and inward directions while the measured responses were somewhat unsymmetrical due to the fact that the tie bars resist outward motion but not inward

motion

.

VI-7
The linear analytical model was also used in the correlation Studies of tests H3 and

HV2. However, due to the presence of multiple collisions and yielding of tie bars, very

poor correlation of analytical and experimental responses were obtained. Obviously, linear

analytical modelling is unacceptable for cases of high intensity excitation which produces

significant nonlinearities to develop in the structural system.

d) Correlations of Experimental Data With Nonlinear Analytical Model

Based on the experience derived from the linear correlation studies, it was apparent that linear modelling of expansion joint behavior was inadequate for cases of

high intensity excitation. The linear expansion joint element was then replaced by a nonlinear element which hopefully would properly model the yielding of tension tie bars and

the collision phenomenon. Coulomb friction was also introduced into the analytical model

to present the sliding forces at the support interfaces. Correlation of the nonlinear

analytical responses now showed good agreement with the experimental nonlinear responses

measured in tests H3 and HV2, except for the later part measured in test H3. In this

case, failure occurred to a shear key midway in the test which allowed a drift of the

displacement response which was not predictable by the analytical model.

5. Concluding Statement

Based on the excellent correlations finally achieved in Phase 5 of the overall

investigation between analytical and experimental results, it is concluded that accurate

data was obtained in Phase 4 and that the analytical procedures develop in Phase 2 , which

were modified and improved in Phase 5, predict realistically the nonlinear response of

bridge systems under high intensity seismic excitation.

Acknowledgement

The investigation summarized in this paper was sponsored by the U.S. Department

of Transportation, Federal Highway Administration, under contract No. DOT-FH-7798.

Disclaimer

The contents of this paper reflect the views of the authors but do not necessarily

reflect the official views of the U.S. Department of Transportation.

References

1. Iwasaki, T. , Penzien, J., Clough, R. , "Literature Survey - Seismic Effects on Highway Bridges," EERC Report No. 71-11, Earthquake Engineering Research Center, University of California, Berkeley, November, 1972.

2. Iwasaki, T. , Penzien, J., Clough, R. , "An Investigation of the Effectiveness of

Existing Bridge Design Methodology in Providing Adequate Structural Resistance to Seismic

Disturbance, Phase I: Literature Survey," FHWA-RD-73-13 , Federal Highway Administration,

November, 1972.

3. Penzien, J., "Seismic Effects on Bridges, San Fernando Earthquake, 9, February

1971," III Congreso Nacional de Ingenieria Sismica, Acapulco, Gro. Mexico, November,

1971.

Vl-8
4. Jennings, P.C., "Engineering Features of the San Fernando Earthquake, February 9,

1971," EEKL 71-20, Earthquake Engineering Research Laboratory, California Institute of

Technology, June, 1971.

5. "Bridge Planning and Design Manual," California State Division of Highways," Vol.

1 - Design Specifications, March, 1968.

6. Elliot, A.L. , "The San Fernando Earthquake: A Lesson in Highway and Bridge Design,"

Civil Engineering, ASCE, Special Issue, September, 1972.

7. Tseng, W.S. and Penzien, J., "Analytical Investigations of the Seismic Response

of Long, Multiple-Span Highway Bridges," Report No. EERC 73-12, Earthquake Engineering

Research Center, University of California, Berkeley, June, 1973.

8. Chen, M.C., and Penzien, J., "Analytical Investigations of Seismic Response of

Short, Single or Multiple-Span Highway Bridges," Report No. EERC 75-4, January, 1975.

9. Tseng, W.S., and Penzien, Jr., "Seismic Analysis of Long Multiple-Span Highway

Bridges," International Journal of Earthquake Engineering and Structural Dynamics, Vol. 4,

1975.

10. Tseng, W.S. and Penzien, J., "Seismic Response of Long Multiple-Span Highway

Bridges," International Journal of Earthquake Engineering and Structural Dynamics, Vol.

4, 1975.

11. Williams, D. , and Godden, W.G., "Seismic Behavior of High Curved Over-crossings,"

Proceedings U.S. National Conference on Earthquake Engineering, Ann Arbor, Michigan, EERI

,

June, 1975.

12. Williams, D. , and Godden, W.G. , "Multidirectional Seismic Response of a Curved

Highway Bridge Model," Bulletin of the New Zealand National Society for Earthquake Engineering, Vol. 8, No. 3, September, 1975.

13. Williams, D. and Godden, W.G. , "Seismic Response of Bridge Model," Transportation

Research Board Record, National Research Council. (Accepted for piiblication.

)

TABLE - 1 FEATURES OF SEISMIC EXCITATION TESTS

EXCITATION

TEST NO.

PEAK

TABLE ACCELERATION [Gj OBSERVED FEATURE^ IN EXCITATION

HORIZONTAL VERriCAL

YIELDING

OF TIE BARS

COLLISIONS

OF GIRDERS

FAILURE

OF SHEAR KEY

HI 0.11 NO NO

1

NO

0.50 YES YES YES

liV2 0.47 o.;^7 YES YES NO

4

VI-9
AN EVALUATION METHOD FOR THE EARTHQUAKE RESISTANT CAPACITY OF REINFORCED CONCRETE AND STEEL REINFORCED CONCRETE COLUMNS

MASAKAZU OZAKI

Head, Structure Division

Building Research Institute, Japan

YUJI ISHIYAMA

Chief Research Engineer, Structure Division

Building Research Institute, Japan

ABSTRACT

An evaluation method for the earthquake resistant capacity of reinforced concrete and

steel reinforced concrete columns by utilizing the force deflection relationship of column

specimens subjected to axial force and repeated and reversed lateral loading of considerable intensity is proposed.

An approximate response analysis for a non-linear structural system was developed

based on random vibration theory and was applied to models represented by a single-degreeof- freedom system subjected to a constant white noise acceleration. The mean expected

maximum response values of the models with two different natural periods 0.1 and 0.5 sec. were calculated. Each model has a degrading stiffness system and various hysteretic

envelope slopes after four different yield point levels. The viscous damping ratio is

considered to be 5% of the critical damping for the entire processes of the models, and

the hysteretic damping ratio after yielding is assumed to increase according to the

increase of ductility factor. The ductility factors of the models calculated by the non- linear response analysis are shown in tables and figures.

The maximum strength of the linear model having the equivalent earthquake resistant

capacity of a column specimen can be assumed if the yield point, hysteretic envelope

slope tangent, coefficient of hysteretic damping ratio and ductiltiy factor of the speci- men are measured by testing. A facility was designed for testing large models of reinforced concrete and steel

reinforced concrete columns under action simulating gravity load and ground motion in order

to standardize the testing techniques and to forestall possible errors that may be induced

by the use of different types of testing facilities.

An example of evaluation for the earthquake resistant capacity of a reinforced con- crete column is presented by utilizing the force deflection relationship of the specimen

obtained by the testing facility.

It is confirmed that yield point level, hysteric envelope slope, hysteretic damping

ratio and ductility factor are the most important components of earthquake resistant

capacity.

Keywords: Columns; Earthquake; Concrete; Hysteretic Envelope; Reinforced; Strength.

VI-10
1. Introduction

A number of modern reinforced concrete buildings have had severe damage during recent

strong earthquakes. Some reinforced concrete buildings have suffered damage in the columns,

due to the shearing instituted by the earthquakes, resulting in collapse of entire buildings.

Large scale model tests of reinforced concrete and steel reinforced concrete column

specimens, have been conducted in order to investigate the earthquake resistant capacity

of such columns. Force deflection relationships were obtained from these test results and

were reported in various technical papers

.

There is, however, no reasonable evaluation method for testing of column specimens

with ductility after yielding.

This paper deals with a method for determining the earthquake resistant capacity of

reinforced concrete and steel reinforced concrete columns with ductility after yielding.

The method utilizes the force deflection relationship of column specimens subjected to

axial force and repeated and reversed lateral loading of considerable intensity.

It is considered that the same method is applicable to the evaluation for the earthquake resistant capacity of girders, joints of girders and columns shearing walls and

any building structures constructed of reinforced concrete and steel reinforced concrete.

2. Assumptions

a) Stationary acceleration spectrum of duration 30 sec. without any predominant

frequency (white noise) was utilized to represent strong motion earthquake excitation.

One of the most significant contributions to the field of earthquake engineering was

the introduction of the idea of the earthquake response spectrum which was first introduced by M.A. Biot, E.G. Robinson, G. Housner and others. Using the eight components

of the four strongest ground motions recorded to date, Housner normalized each accelerogram to a common intensity level and by averaging the velocity spectra resulting therefrom

obtained what is now commonly known as Housner 's standard velocity spectra. Even though

j the number of samples may be small in this case, Housner 's standard velocity spectra give

I the mean value of maximum (or peak) response due to earthquakes of an intensity level

I which is generally considered a good measure of the expected damage to structures having

j fundamental period in the range 0.1 < Tn < 2.5 sec.

1 ~

' ' . [

•

. : an analog computer, Bycroft studied the possibility of using a white noise

acceleration process to represent earthquake ground motion of a given intensity level.

Bycroft 's results would seem to indicate that white noise is a reasonable simulation of

earthquake excitation.

A more accurate method of predicting the maximum or peak response when concerned with

long period and low damping ratio and when assuming a white noise acceleration process

! of finite duration is the method presented by Rosenblueth and Bustamante. The results were

I compared with Housner 's standard velocity spectra and quite good agreement between them

I over the entire ranges of damping and period was reported in Reference (5). Therefore,

' VI- 11

i
a stationary acceleration represented by a white noise was utilized for this evaluation.

b) Non-linear structural models with a degrading stiffness system represented by a single-degree-of-freedom system were used as shown in Fig. 1. A non-linear dynamic

response analysis for the models subjected to a stationary white noise acceleration was carried out by utilizing the force deflection curves obtained by large model tests of

column specimens. An idealized force deflection curve with a degrading stiffness system

is shown in Fig. 2. In general, reinforced concrete and steel reinforced concrete column specimens show

a tendency that the maximum strengths at each step of deflection after yielding reduced due

to repeated and reversed loading. An envelope line which plots the reduced maximum strengti

corresponding to each step of deflection can be approximated by a straight line , as shown

in Fig. 2. The slope of the straight line is defined as the hysteretic envelope slope,

and the ratio of the hysteretic envelope slope tangent to the initial stiffness slope

tangent is named the slope tangent ratio of ductility.

The damping factor induced by viscous damping (viscous damping ratio ^v) is considered

to be 0,05 of the critical damping for the entire processes of the models. The damping

factor induced by hysteretic damping (hysteretic damping ratio ^H) can be calculated by

measuring the areas of hysteretic loops corresponding to each step of deflection of

specimens. For example, the hysteretic damping ratio at the deflection 26y is simply

obtained, if the hysteretic loops are symmetrical with respect to the original point, as follows;

1 Area 2. 3. 5 '^H{d = 2Sy) JT Area 2. di 5 + Area 2.(2^3 *

'

where 6y is the deflection at the yielding point of specimens.

In the case of reinforced concrete and steel reinforced concrete specimen tests,

the hysteretic damping ratio can be assiamed to increase proportionally to the increase

of / C6-6y) /6y after yielding. Therefore, the hysteretic damping ratio at the deflection

6 is approximated by

- jS-Sy :H-cCh^—— (3^3y) (2)

where ci^H is defined as the coefficient of hysteretic damping ratio.

By using the method of least squares, the coefficient of hysteretic damping ratio can

be obtained from the hysteretic damping ratios ^Hn at the deflections 6n(n=l,2, . . . .n) as

n <;Ch-^ I (3)

1

VI-12
In calculation, the hysteretic damping with a loop was transformed to the equivalent

viscous damping for a linear system. Therefore, the equivalent viscous damping ratio ? was obtained by adding the viscous damping ratio to the hysteretic damping ratio ^H3. Numerical Calculation

The mean expected maximum response values of the structural models having a degrading

stiffness system shown in Fig. 2 were calculated by a non-linear response analysis when

they were subjected to a stationary white noise acceleration.

2 The models with natural period Tm 6.1 sec. (m = 1.0 ton sec /cm, K = 4 942.0 ton/

cm) were divided into four types according to the four different yield point levels, where

the yield point levels of Model 1, 2, 3, and 4 are 1/2, 1/3, 1/4 and 1/5 of the mean

expected maximum response values in a linear system respectively. Where the mean expected

maximum response value in a linear system is defined as the mean of the maximum response

values, the structural model has a linear system without the yield point. In other words,

the mean maximum response value in the degrading stiffness system can be replaced by the

strength of the linear model with the equivalent earthquake resistant capacity.

In each model, the slope tangent ratio of ductility varies as 0.8, 0.6, 0.4, 0.2, 0.1,

0.0, ...-0.1 after yielding. The ductility factors of the models were calculated by a

I non- linear response analysis when the coefficient C,H of hysteretic damping ratio varies

1 from 0.03 to 0.12.

For example, if the maximum response value of a column having a linear system is

three times the yield point level of the column having a non-linear system, the ductility

indicated in Fig. 3b will be required to prevent the complete collapse from the structure,

when the structure is subjected to a strong earthquake. In other words, if the yield

point, hysteretic envelope slope tangent, coefficient of hysteretic damping ratio and

ductility factor of a column are measured by testing, the maximum strength of the linear

model having the equivalent earthquake resistant capacity of the column can be presumed.

Table 3 and 4 show the increment ratio of the equivalent maximum strength of linear

models to the yield point levels of column specimens due to the increase of the earthquake resistant capacity by ductility for the cases of Type 1 and Type 2. Therefore, a new

allowable strength level of columns can be defined by taking the collapse of structures

j into consideration.

The models with natural period Tr = 0.5 sec. were also calculated in the same manner.

There is however, little difference between the models with the different natural periods

for the calculated response values.

In this calculation, an approximate non-linear response analysis was developed

based on random vibration theory and was applied to models with a single-degree-of-freedom

system subjected to a constant white noise acceleration. The non-linear response analysis

based on random vibration theory is introduced in Reference (4) and examined by the method

of simulation in Appendix I.

VI-13
4. Testing Facility

A facility was designed for testing large models of reinforced concrete and steel

reinforced concrete columns iinder action simulating gravity load and ground motion in order

to standardize the testing techniques and to forestall possible errors that may be induced

by the use of different types of testing facilities. The elevation of this facility is

shown in Fig. 5. Main features of this facility are as follows;

a) Models of columns with sectional area of 25 cm x 25 cm for reinforced concrete

and 30 cm x 30 cm for steel reinforced concrete and length varying fron 50 cm up to 150 cm

can be tested.

b) Models of columns can be subjected to axial force and repeated and reversed lateral

loading of considerable intensity.

c) Models can be subjected to very large lateral deformations by the repeated and

reversed lateral loading.

d) In case the top of a specimen of the models is considerably displaced in lateral

direction, the specimen is subjected to bending moment and shearing force induced by the

eccentric gravity in addition to the bending moment and sheering force by the repeated

lateral load acting at the top of the specimen. In other words, P - 6 effects are considered

e) The facility is safe even in the case of sudden and brittle failure of the speci- mens .

f) Stresses and deformations induced in the specimens under test can be easily

measured and progressive pattern of damages and mechanism of failure up to collapse can be

visually observed.

g) The setting or removal of the specimens can be accomplished easily.

The earthquake resistant capacity of a reinforced concrete column specimen is evaluated

by utilizing the force deflection relationship obtained by the testing facility shown in

Fig. 5. The dimensions and bar arrangement of the specimen are indicated in Table 5 and

Fig. 6. The force deflection curve of the specimen is shown in Fig. 7.

5. Examples

Section B X D = 25 cm X 25 cm Length L = 50 cm Axial force ratio N/BD = 26.25 kg/cm^

Strength of concrete

shear span ratio M/QD = 1.5 Fc = 245 kg/cm'

Main reinforcing bars 6 - D 13 (SD30)

Hoops 9 0 - (3 45.5 (SR24)

Tensile bar ratio Pt = 0.61%

Web reinforcement ratio Pw = 1.12%

VI-14
This specimen showed a shear- type collapse at the lateral deflection of 4.0 cm. The

yield strength of 18.5 ton and the corresponding yield displacement of 0.2 cm were observed.

The hysteretic envelope slope defined in Section 2 was determined by a straight line as shown in Fig. 7. Measuring the hesteretic damping ratios at the displacements 6 =0.5, 1.0, 1.5 and

2.0 cm, the coefficient of the hysteretic damping ratio can be calculated by the method

of least squares given in Eq. (3)

.

The four important components of earthquake resistant capacity were obtained as follows

:

1. Yield point level 19.25 ton

2. Slope tangent ratio of ductility -0.054

3. Coefficient of hysteretic damping ratio 0.073

4. Ductility factor at collapse 20.0

The specimen with the above values of ductility satisfies the ductility requirement

shown in Fig. 3b. The maximum strength of a linear model having the equivalent earthquake

resistant capacity of this specimen is presumed to be about 57.75 ton by multiplying the

yield strength times the increment ratio of maximum strength by ductility shown in Table

3. Let us consider a particular problem of an irregular frame model shown in Fig. 8a. This model has two columns with different length. In order to evaluate the earthquake

resistant capacity of this model, each column cannot be discussed separately, because the

whole structure will be destroyed if one of the columns collapses. Therefore, the superposed force deflection curve must be taken into consideration, as shown in Fig. 8b. It should be noted that the smaller deflection capacity of the shorter column gives

the deflection capacity of the whole structure in the case of this irregular frame model.

6. Conclusions

It is confirmed that yield point level, hysteretic envelope slope, hysteretic damping

ratio and ductility factor are the most important components of earthquake resistant

capacity, when reinforced concrete and steel reinforced concrete columns are siibjected to

a white noise ground motion.

In order to maintain desirable hysteretic envelope slope after yielding, it will be

recommended to avoid the shear- type failure of columns. It is also important to use sufficient ductile materials such as steel and reinforcing bars in columns to increase hysteretic damping ratio and ductility factor of columns. When a column without sufficient

ductility is used, the allowable strength level should be estimated lower than the conventional one in the case of strong earthquakes.

It will be reasonable to use this evaluation method for the earthquake resistant

capacity of reinforced concrete and steel reinforced concrete columns, if it is difficult

to determine a specific strong earthquake input for designing a building structure.

VI-15
Acknowledgments

The authors are deeply indebted to Dr. Minouru Wakabayashi, Professor of Kyoto University for his valuable suggestion for designing the testing facility.

References

1. Tajimi, H.» Introduction to Structural Dynamics, Corona-sha, Tokyo, Japan, 1964.

2. Ozaki, M. and Others , Earthquake Response Prediction and Failure of Reinforced

Concrete Buildings due to Repeated Shearing Forces, Part I. Theoretical Analysis of Earthquake Response Prediction, Research Paper of the Building Research Institute, No. 40, 1969,

Japan.

3. Ozaki, M. and Yamazaki, Y., Dynamic Response Prediction of Buildings subjected

to Earthquakes, Part II. Numerical Analysis, Research Paper of the Building Research

Institute, No. 42, 1970, Japan.

4. Ozaki, M. and Ishiyama, Y., Earthquake Response Prediction of Reinforced Concrete

Buildings, The 5th World Conference on Earthquake Engineering, 1973.

5. Penzien, J., Application of Random Vibration Theory in Earthquake Engineering,

Bulletin of the International Institute of Seismology and Earthquake Engineering, Vol. 2,

1965.

6. Bycroft, C.N., White Noise Representation of Earthquakes, Proceedings of the

American Society of Civil Engineers, Vol. 86, No EM4, April, 1960.

7. Housner, G.W., Behavior of Structures during Earthquakes, Proceedings of the

American Society of Civil Engineers, Vol. 85, No. EM4, October, 1959.

VI-16
Pig. 2 Hysteretic Curve

M»>nEipect«j Mai

Force in Lrnrar System

T 2 3 I 5 6

Fig. 3a Ductility Requirement

^Model l)

Mean Eiipecled Mai.

Fig. 3b Ductil ity Requirement

(Model 2)

Mean Eipected Mai.

T 5 5 I 5 6 5 8 9 » 11 0 u

Fig. 3c Ductility Requirements

(Model 3)

Mean Eipecttd Mai

Force In Unrar S)it«n

1 i 3 4 % S 7 « » 10 I I 12 13 U is

Fig. 3d DuctSIBty Requirements

VI-17
MunExptcled Mii

Force in LoKir Systun

I 23^56789

Fig. 4b Ductility Requirements

(Model 2 of Type 2) Tn = 0.5sec

Mem Exp«cte<j Mai

Force in Lrtear Sy^lem

l23i5G789K)t112UUIS

Fig. 4d Ductility Requirements

(Model 4 of Type 2) Tn = 0.5sec

VI-18
N

t-013

Pi -OH* M«H5.5

P».M2

HIBD. 26 25kO'cm

Fig. 6 Specimen

Fig. 7 Force-Deflection Relationship of Specimen

(A)

(B)

Ki C,

7777777

KjCz

777777

(a) Model

Collapse

(A) Collapse

(b) Hysteretic Curves

Fig. 8 Irregular Frame

VI-19
Slope Tangent Ratio

Coef f i c i ent of Hystcrctic Damping /VS P 9 t110! /I c C n

0.0 0 0. 0 3 0.0 4 0.0 5 0.0 6 0.0 7 0. 0 8 00 9 0. 1 0 0. 1 1 0 1 2

1.0 2.0 0 0.8 2.1 4 1.7 6 1.6 8 1.6 2 1.5 8 1.5 4 1.5 0 1.4 7 1.4 4 1.4 2 1.4 0

0.6 2.3 5 1.8 6 1.7 7 1.7 0 1.6 4 1.5 9 1.5 5 1.5 1 1.1 8 1.4 5 1.4

'

0.4 2.6 9 2.0 1 1.9 0 1.8 1 1.7 3 1.6 7 1.6 2 1.5 8 1.5 4 1.5 0 1.4 0.2 3.3 4 2.2 6 2.,1 0 1.9 7 1.8 7 1.7 9 1.7 2 1.6 7 1.6 2 1.5 7 1.5 4 0.1 4.0 4 2.4 7 2.2 6 2.1 0 1.9 8 1.8 8 1.8 0 1.7 3 1.6 7 1.6 2 1.5 8

0.0 0 6.0 2 2.8 4 2.5 3 2.3 0 2.1 4 2.0 0 1.9 0 1.8 2 1.7 4 1.6 8 1.6 3 - 0.0 2 7. 1 2 2.9 5 2.6 0 2.3 6 2.1 8 2.0 1 1.9 3 1.8 1 1.7 6 1.7 0 1.6 5

- 0.0 4 9. 9 2 3.0 8 2.6 9 2.4 2 2.2 2 2.0 8 1.9 5 1.8 6 1.7 8 1. 7 1 1.6 6

- 0.0 6 3.2 6 28 0 2.5 0 2.2 7 2.1 2 1.9 8 1.8 9 1.8 0 1.7 3 1.6 7 - 0.0 8 3.4 9 2.9 3 2.5 8 2.3 4 2.1 6 2.0 2 1.9 1 18 3 1.7 5 1.6 9

- 0.1 0 3.8 2 3.0 9 2.6 8 2.4 1 2.2 1 2.0 0 1.9 4 18 5 1. 7 7 1.7 1

Table la Due t i 1 i ty Requ i r erne n

for Mode 1 1 of Type 1 (Tn = 0.1 sec. , Py = l/2 Pmax • )

Slope Tangent Ratio

Coefficient of Hysteretic Dam p i ng Ratio Of h 0.0 0 0.0 3 0.0 4 0.0 5 0.0 6 0 0 7 0 U 8 0.0 9 0. 1 0 0 1 1 0 1 2

1.0 3.0 0

0.8 3.3 1 2.5 3 2.3 9 2.2 8 2.1 9 2.1 2 2.0 5 1.9 9 1.9 4 1 9 0 1 8 6

0.6 3.7 8 2.7 7 2.6 0 2.4 7 2.3 6 2.2 7 2. 1 9 2.1 3 2.0 7 2 0 1 1. 9 6. 0.4 4.5 5 3.1 5 2.9 3 2.7 6 2.6 2 2.5 0 2.4 0 2.3 2 2.2 4 2 1 7 2 1 1 0.2 6.2 7 3.8 7 3.5 4 3.2 8 3.0 7 2.9^0 2.7 6 26 4 2 5 3 2 4 4 2 3 6

0.1 8.5 0 4.6 1 4.1 4 3.7 8 3.5 0 3.2 7 3.0 8 2.9 2 2 7 8 2 6 6 2 5 6 0.0 2 1.1 0 6.5 5 5.5 5 4.8 8 4.4 6 3.9 9 3.6 8 3.4 3 3.2 3 3 0 5 2 8 3 - 0.0 2 75 4 6.2 0 5.3 2 4.7 1 4.2 4 3.8 9 3.5 9 3.3 6 3 1 7 3 0 - 0.0 4 9.7 4 7 3 1 6.0 2 5. 1 7 4.6 0 4.1 5 3.8 1 3 5 3 3 3 1 3 1 - 0.0 6 1 0.9 7 7 4 4 5.9 9 5. 1 1 4.5 4 4.0 9 3.7 6 3 4 8 3 2 7 - 0.0 8 8.2 8 6.1 5 5. 1 4 4.5 4 4.0 7 3 7 4 3 4 5 - 0.1 0 6.9 2 5.3 6 4.6 0 4 0 8 3 4

Table lb Ductility Requ i reme n t for Mode 1 2 of Type 1 (Tn = 0.1 sec .

,

Py = 1/3 Pmax. )

Slope

Tangent Ratio

Coefficient of Hy s t e r e t i c Damp 1 ng Ratio cch

0.0 0 0.0 3 0.0 4 0.0 5 0.0 6 0.0 7 0.0 8 0.0 9 0. I 0 0.1 1 0 1 2

1.0 4.0 0

0.8 4.4 8 3.2 7 3.0 7 2.9 2 2 7 9 2 6 8 2 5 9 2 5 1 2. 1 4 2 3 8 2 3 2

0.6 5.2 1 3.6 5 3.4 1 3.2 2 3-0 7 2.9 4 28 3 27 3 2 6 5 2.5 7 2 5 0

0.4 6.4 5 4.2 7 3.9 5 3.7 0 3.5 0 3.3 3 3.1 9 3.0 7 2.9 6 2.8 6 27 8

0 2 9.3 1 5.5 0 5.0 0 4 6 2 4.3 2 4.0 7 3.8 6 3.6 8 3.5 2 3.3 8 3.2 6

0.

1 3.3 0 6.9 0 6. 1 7 5.6 1 5. 1 8 4.8 2 4.5 3 4.2 8 4.0 7 3.8 8 3.7 2 0.0 0 5 0.7 0 1 1.8 2 9.8 3 8.4 5 7 5 0 6.7 5 6. 1 5 5.6 8 5.2 9 4.9 6 4.6 6

- 0.0 2 1 6.7 4 1 2.6 5 1 0.3 1 8.8 2 7 7 2 6.9 4 6.3 1 5.7 9 5.3 9 5.0 5

- 0.0 4 2 0.3 6 12 3 8 9.8 2 8.4 1 7 3 6 6 6 5 6 0 5 5.5 8

- 0.0 6 1 0.8 9 8 5 8 7 3 6 6.5 6

- 0.0 8 1 1.8 2

- 0. 1 0

Table Ic Ductility Requirement for Mode I 3 of Type 1 (Tn -= 0. 1 sec.. Py = 1/4 Pmax )

VI-20
Slope Coef f i c i ent of Hysterc t i c Damp i ng Ratio cf h Tangent Ratio 0. 0 0 U. u o 0.0 4 0.0 5 0.0 6 0.0 7 0.0 8 0.0 9 0. 1 0 nil A 1 9

1.0 5.0 0

0.8 5.6 5 3.9 9 3.7 4 3.5 4 33 7 3.2 3 3.1 2 3.0 2 2.9 3 2.8 5 2.7 7

0.6 6.6 4 4.5 1 4.2 0 3.9 5 3.7 5 3.5 9 3.4 4 3.3 2 3.2 1 3.1 2 3.0 3

0.4 8.3 5 5.3 5 4.9 4 4.6 2 4.3 6 4.1 4 3.9 6 3.8 0 3.6 6 3.5 4 3.4 3

0.2 1 2.3 8 7. 1 1 6.4 5 5.9 6 5.5 6 5.2 3 4.9 6 4.7 2 4.5 2 4.4 3 4. 1 8

0.1 1 8.1 9 9.2 1 8.2 3 7.4 9 6.9 0 6.4 3 6 0 4 5.7 1 5.4 2 5.1 7 4.9 4 0.0 9 7.8 0 1 8.5 6 1 5.2 3 1 2.9 8 1 1.4 1 1 0.2 2 9.2 7 8.4 9 7 8 7 7. 3 5 6.9 0 - 0.0 2 2 7.64 1 9.4 7 1 5.5 4 1 3.2 0 1 1.5 0 1 0.2 7 9. 3 2 8. 5 4 7. 8 8 - 0.0 4 1 8.3 3 1 3.5 4 1 1.5 2 1 0.0 3 — i\ f\ a. U. U D

- 0.0 8

- 0.1 0

Table Id Ductility Requirement for Model 4 of Type 1 (Tn = 0.1 sec •

,

Py = 1/5 Pmax- )

Slope

Tangent Ratio

Coef f i c ient of Hy s te re t i c Dampi ng Ratio cch

n A A (\ Cl 1 0.0 4 0.0 5 0.0 6 0.0 7 0.0 8 0.0 9 U. I u A 1 1 U. 1 ^

1.0 2.0 0

0.8 2.1 3 1.7 7 1.7 0 1.6 4 1.6 0 1.5 6 1.5 2 1.4 9 1.4 7 1.4 5 1.4 3

0.6 2.3 1 1.8 6 1.7 8 1.7 1 1.6 6 1.6 2 1.5 8 1.5 4 15 1 1.4 8 1.4 6

0.4 2.6 0 20 0 1.8 9 1.8 1 1.7 4 1.6 9 1.6 4 1.6 0 1.5 6 1.5.3 1.5 1 0.2 3.1 1 2.2 1 2.0 7 1.9 5 1.8 7 1.7 9 1.7 3 1.6 8 1.6 4 1.6 0 1.5 7 0.1 3.6 3 2.3 7 2.2 0 2.0 6 1.9 5 18 7 1.8 0 1.7 4 1.6 9 1.6 4 1.6 0 0 0 4.7 2 2.6 4 2.3 9 2.2 2 2.0 8 1.9 7 1.8 8 l.S 1 1.7 3 1.7 0 1.6 5

- 0.0 2 5.1 5 2.7 1 2.4 4 2.2 6 21 1 1.9 9 1.9 0 1.8 3 1.7 6 1.7 1 1.6 6

- 0.0 4 5.7 6 2.8 0 2.5 0 2.3 0 2.1 5 20 2 19 2 1.8 5 1.7 8 1.7 2 1.6 8

- 0 0 g 6. 7 3 2.8 9 2.5 7 2.3 5 2.1 8 2.0 5 1.9 5 1.8 7 1.8 0 1.7 4 1.6 9

- 0.0 8 1 0.2 3 3.0 1 2.6 5 2.3 9 2.2 3 2.0 9 1.9 7 1.8 9 1.8 2 1.7 5 1.7 0

- 0. 1 0 3.1 4 2.7 5 2.4 6 2.2 7 2.1 2 2.0 0 1.9 1 1.8 4 1.7 7 l.i? 1

Table 2a Ductility Requirement for Model I of Type 2 (Tn = 0.5 sec.. Py = 1/2 Pmax. )

Slope

Tangent Ratio

Coef f i c i en t of Hys tere t i c Damping Rati 0 cch

o n A 0.0 4 0.0 5 0.0 6 0.0 7 0 0 8 0.0 9 0 10 Oil 0 12

1.0 3.0 0

0.8 3.3 0 2.5 4 2.4 1 2.3 1 22 2 2.1 4 2.0 8 2.0 3 1.9 8 1.9 4 1.8 9 0.6 3.7 2 2.7 7 2.6 1 2.4 8 23 8 2 2 9 2 2 1 2.1 5 20 9 2.0 4 2.0 0

0.4 4.4 3 3.1 2 2.9 1 2.7 5 2.6 2 . 2.5 1 2.4 1 2.3 3 2.2 6 2.1 9 2.1 4 0.2 5 9 2 3.7 6 3.4 5 3.2 1 3.0 2 2.8 7 2.7 3 2.6 2 2.5 2 2.4 4 23 6

0.1 7 6 9 4.3 8 3.9 6 3.6 4 3.3 8 3.1 7 3.0 1 2.8 6 2 7 4 2.6 3 2.5 4 0.0 1 4.5 3 5.7 7 4.9 9 4.4 5 4.0 5 3.7 4 3.4 8 3.2 6 3.0 9 2.9 5 2.8 2

- 0.0 2 2 3.5 1 6.3 4 5.4 0 4.7 3 4.2 7 3.9 2 3.6 3 3.3 8 3.1 9 3.0 4 2.9 0

- 0.0 4 7.1 9 5.9 6 5.1 3 4.5 4 4.1 3 3.8 1 3.5 4 3.3 1 3.1 4 2.9 9

- 0.0 6 9 1 >; 6.8 5 5.7 2 4.9 5 4.4 2 4.0 3 3.7 2 3.4 6 3. 2 6 3.0 9

- 0.0 8 9.4 4 6.7 2 5.5 9 4.8 3 4.3 3 3.9 6 3.6 6 3.4 0 3.2 2

- 6.1 0 6.8 5 5.5 8 4.7 6 4.2 8 3.9 2 3.6 1 3.3 7

Table 2b Ductility Requirement for Model 2 of Type 2 (Tn = 0.5sec ., Py = 1/3 Pmax. )

VI-21
Slope Tangent Ratio

Coef f 1 c i ent of Hys tere t i c Da m p i ng Ra t i 0 cc h 0.0 0 0.0 3 0.0 4 0 0 5 0.0 6 0.0 7 0.0 8 0 0 9 0. 1 0 0. 1 1 0. 1 2

1.0 4.0 0 0.8 4.4 6

0 O Q 3.0 9 2.9 5 2.8 2 2.7 2 2.6 3 2 5 5 2. 4 8 2.4 3 2. 3 7 0.6 5. 1 4 3.6 5 3.4 2 3.2 3 3.0 9 29 6 2.8 5 2.7 6 2.6 8 2.6 1 2 .5 ••

0.4 6.2 9 4.2 3 3.9 2 3.6 8 3.4 9 3.3 3 3.1 9 3 0 8 2 9 7 2.8 8 2.8 I 0.2 8.8 4 5.3 5 4.8 8 4.5 3 4.2 4 4.0 0 3.8 1 3.6 4 3.4 9 3.3 6 3.2 5

0.1 1 2.1 3 6.5 5 5.8 9 5.3 8 4.9 8 4.6 6 4.3 8 4. 1 6 3.9 7 3.8 0 3.6 5 0.0 3 2.0 8 1 0.1 0 8.5 8 7. 5 0 6.7 1 6.1 3 5.6 6 5.2 6 4.9 3 4.6 4 4.4 1 - 0.0 2 1 2.3 4 1 0.0 6 8.5 8 7. 5 1 6.7 3 6.1 5 5.6 8 5.2 9 4.9 6 4.6 7 - 0.0 4 1 3.5 1 1 0.6 5 8.9 4 7. 7 5 6.8 8 6.2 8 .5.7 0 5. 3 8 5 0 3 - 00 6 1 3.3 4 1 0.1 0 8.4 2 7 2 9 6. 5 6 6.0 1 5 5 6 - 0.0 8 1 1.0 3 8.4 4 7. 1 2 6 4 3 - 0.1 0

Table 2c Due t i 1 i ty Requ i rement for Mode 3 of Type 2 (Tn = 0.5 sec . Py = 1/4 Pmax- )

Slope Tangent - Coef f i c i en t 0 f Hys t e re t i c Damp i ngr Ratio cch

Ratio 0.0 0 0.0 3 0.0 4 0.0 5 0.0 6 0.0 7 0.0 8 0.0 9 0. I 0 0. 1 1 0. 1

1.0 5.0 0

0.8 5.6 3 4.0 1 3.7 6 3.5 7 3.4 1 3.2 7 3.1 6 3.0 6 2.9 7 2.9 0 2 8 3 0.6 6.5 6 4.5 1 4.2 0 3.9 6 3.7 7 3.6 1 3.4 7 3.3 5 3.2 5 3. 1 6 3.0 8

0.4 8.1 6 5.3 1 4.9 1 4.5 0 4.3 4 4. 1 4 3.9 6 3.8 1 3 6 8 3.5 6 3.4 6

0.2 1 18 0 6.9 2 6.3 0 5.8 3 5.4 6 5.J 5 4.8 9 4 6 7 4.4 7 4.3 0 4. 1 6

0.1 1 6.7 8 8.7 7 7. 8 6 7. 1 8 6 6 4 6.2 0 5.8 4 5 5 3 5. 2 7 5 0 4 4 8 4 0.0 5 7.9 1 1 5.5 1 1 2.9 9 1 1.2 5 1 0.0 1 9.0 8 8.3 3 7 7 1 7 1 9 6 7 5 6.3 - 0.0 2 2 3.0 5 1 7. 4 3 1 4.2 7 1 2. 1 7 1 0.6 7 9.6 2 8 7 9 8. I 1 7 5 1 7.0 - 0.0 4 2 2.0 0 1 5.4 7 1 2.6 7 1 0-8 6 9 7 5 8.8 9 8.1 - 0.0 6 1 3.5 9 1 0.8 6 - 0.0 8

- 0.1 0

Table 2d Ductility Requirement for Mod el 4 of Type 2 (Tn = 0.5 sec .

, Py = 1/5 Pmax. )

VI-22
2

3

4

56

78

9

2

3

4

56

78

9

2

3

4

56

78

9

2

3

4

56

78

9

2

3

4

56

78

9

2

3

456

78

9

Coefficient of Hystcretic Damping Ratio cch

0.0 3 0.0 4 0.0 5 0.0 6 0.0 7 0.0 8 0.0 9 0. 1 0 O.I 1 0. 1 2

2 2 2 2 2 2 3 3 3 3

3 3 4 4 4 4 4 5 5 5 5 5 5 5 5 5 5

1 1 1 1 2 2 2 2 2 2

2 2 2 2 2 2 2 2 2 3

2 2 2 2 3 3 3 3 3 3 2 2 3 3 3 3 3 3 4 4 2 3 3 3 3 3 4 4 4 4

3 3 3 3 4 4 4 4 4 5

3 3 3 4 4 4 4 5 5 3 3 4 4 4 4 5

VI-23
Slope Duct i 1 i ty Coef f i c ient of Hysteretic Damping Ratio cch

Tangent Rst io Factor 00 3 0.0 4 n n R n n An? n n n A n Q 0. 1 0 0.1 1 0. 1

1X1 1 1 1 2 2 £. *> 0£. 3 2 2 2 2 2 2 2 2 2 3 4 2 2 2 2 2 3 3 3 3 3 _ /\ ft o eD 2 2 2 3 3 3 3

•a

ft 0 2 3 3 3 3 3

A A 7 3 3 3 3 4 4

4 i* •1 0 0 3 3 4 4 4

ji*l C Q9 3 4 4 4 4

A*\ C

111i 1 1 1 2 20C 3 1 2 2 2 2 2 2 2 2 2

4 2 2 2 2 2 2 3 3 3 3 » i\ t\ A c £• 0£• 2 2 3 3 3 0

0 V 9 £. 2 3 3 3 3

f C 3 3 3 3 3

A 1 Q0 3 3 3 3 4

AH A 0

C 0 3 3 3 4 4 •* A A

1

1 1 1 1 2 2 £. £. 3 1 2 2 2 2 2 2 2 2 2

4 2 2 2 2 2 2 2 3 3 3 M A A A c0

0£. £. 2 2 2 3 3 D

0£, 0£, 2 3 3 3 3 >> 5 •J

f 0£. 0 2 3 3 3 3 *• 00

0£, 3 3 3 3 3 *t Q 0 3 3 3 3 3 *\

11 1 1 1 1 1 2 £. 3 1 2 2 2 2 2 2 2 2 2 4 2 2 2 2 2 2 2 2 3 3

_ A A O z 2 2 2 2 2

Q0 0c. c 2 2 2 3 3

Q

f 0£, 2 2 3 3 3 0

0 0£. 2 2 3 3 3 3 3 9 2 2 2 3 3 3 3 3 3 3

2 1 1 1 1 1 1 2 2 2 3 1 1 2 2 2 2

0 2 2

4 2 2 2 2 2 2 2 2 2 3 - O.J 0 5 2 2 2 2 2 2 2 3 3 3 6 2 2 2 2 2 2 2 3 3 3

7 2 2 2 2 2 3 3 3 3 3 8 2 2 2 2 2 3 3 3 3 3 9 2 2 2 2 2 3 3 3 3 3

Table 3 Increment Ratio of Maximum Strength by Ductility

Type 1 ( Tn = 0 1 sec . ) Blank indicater 5 of acre than 5-

VI-24
3

4

56

78

9

2

3

4

56

78

9

2

3

456

78

9

23

4

56

78

9

2

3

436

78

9

2

.34

56

7

Coefficient of Hysteretic Damping: cch

0.0 3 0.0 4 0.0 5 0.0 6 0.0 7 0.0 8 0.0 9 0. 1 0 0.1 1

2 2 2 2 2 2 2 3 3 3 3 4 4 4 4 4 5 5

4 5 S 5 5 5 S

0. 1 2

5

4 4 5 5 5

5 5

4455555555 2222222223 3333444444 5 5 2333333444 2222222222

43444455555 5 5 5 5 2222333333 1 122222222 333334 3 4 4445 4444 5555

2222223333 1 1 1 2 2 2 2 2 2 2

42333333444 3334444445 344445555 4 4 5 5

4 5 5

5

1 1 1 1 2 2 2 2 2 2

2 2 2 2 2 2 2 2 3 3 2 2 2 2 3 3 3 3 3 3 2 3 3 3 3 3 3 4 4 4 3 3 3 3 3 4 4 4 4 4 3 3 3 4 4 4 4 4 5 5 3 3 4 4 4 4 5 5

3 4 4 4 4 5

VI-25
Slope'

L/ucl 1 1 1 ty Coeff ic en t of Hys te re t i c Damp i cch

Tanc^nt Ratio Fac tor 0 0 3 0. 0 4 0.0 5 0.0 6 0. 0 7 0.0 8 0.0 9 0. 1 0 0. 1 1 0. 1 2

2 1 1 1 1 o o

£, 2 2 2 2 3 2 2 2 2 2 2 2 2 2 3

4 2 2 2 2 3 3 3 3 3 3

- 0.0 2 5 2 2 3 3 3 3 3 3 4 4 6 2 3 3 3 3 3 4 4 4 4 7 3 3 3 3 4 4 4 4 4 4 8 3 3 3 4 4 4 4 4 5 5 9 3 3 4 4 4 4 5 5

2 1 1 1 1

11 o 2 2 2 2 3 2 2 2 2 2 2 2 2 2 3

4 2 2 2 2 2 3 3 3 3 3 - 0.0 4 5 2 2 2 3 3 3 3 3 3 3 6 2 3 3 3 3 3 3 4 4 4 7 2 3 3 3 3 4 4 4 4 4 8 3 3 3 3 4 4 4 4 4 4 9 3 3 3 4 4 4 4 4 5 5

2 1 1 I 1

1i o 2 2 2 2 3 2 2 2 2 2 2 2 2 2 2

4 2 2 2 2 2 2 3 3 3 3 - 0.0 6 5 2 2 2 3 3 3 3 3 3 3 6 2 2 3 3 3 3 3 3 3 4 7 2 3 3 3 3 3 3 4 4 4 8 2 3 3 3 3 3 4 4 4 4 9 2 3 3 3 3 4 4 4 4 4

2 1 1 1 1

1i 2 2 2 2

3 1 2 2 2 2 2 2 2 2 2 4 2 2 2 2 2 2 3 3 3 3 - 0.0 8 5 2 2 2 2 3 3 3 3 3 3 6 2 2 2 3 3 3 3 3 3 3 7 2 2 3 3 3 3 3 3 3 4 8 2 2 3 3 3 3 3 3 4 4 9 2 2 3 3 3 3 3 4 4 4

2 1 1 1 1 1 oI. 2 2 2 2 3 1 2 2 2 2 2 2 2 2 2 4 2 2 2 2 2 2 2 3 3 3 -0.1 0 5 2 2 2 2 2 3 3 3 3 3 6 2 2 2 2 3 3 3 3 3 3

7 2 2 2 3 3 3 3 3 3 3 8 2 2 2 3 3 3 3 3 3 3 9 2 2 2 3 3 3 3 3 3 3

Table 4 Increment Ratio of Max imum Strength by Ducti i ty

Type 2 (Tn = 0 5 sec . ) Blank indicates 5 or more than 5.

VI-26
«

COc • » O 0) *J fH « to>

0)

0) J= c V «J CO c o

0) p. ^ U ' (U M o

01 0)

eti 1-1

U

N o C CO

t-i E-i

a j:

•2

"O c 3 S

0) -H 10 o c O -O 00 o C -H o u •l-( -u U r) •o y

*->

C .-1 c c e H 3 3 U4 U e >^ U (U 0) i-i 4-) M W CO U (AC

(0 cn 1-1 c;

dj

O -3 E 3 O .-1 >> •a CO >>

0) > iJ T-" 0) •H 1—1 U 0) c

IM (ft 3 1 c

jj ^ o a. tn o 1 M O V- E O.

1-1 to .-H«

•a c •H « O aco • C t^ c •H

»-i o iJM 0 -H CO 3

l4-l i-l to > M CJ CO Q) 3 3 e C

f-l -H o c

to to G.

O > !/) -H o -u

<u o Ij to to ;J C T3 0) -H O O J= tj ex jz J-i X

V) i-> OJ

14-1

u

B o o u a tn CO a s: o; 3

0) 4-) 3 tzr c ^ x;

•H (0 J-J 1 > 1-1 to c -n OJ 01 o <u z c

•w o E •H

CO oe

OJ CM r4 CM <Nl o o c o o o

CM CM CN

m

CM CO <Mm fSI <jCM (M OJ m rj CM

COn cn CO

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VI-27
THE EARTHQUAKE ENGINEERING PROGRAM OF THE NATIONAL SCIENCE FOUNDATION

JOHN. B. SCALZI

National Science Foundation

Washington, D.C.

ABSTRACT

The general research program, supported by the National Science Foundation, is given

in general terms. The required interaction between the social, economic and technology

is described .

Key Words: Earthquake^ Program; Research; Seismic.

1. Introduction

Earthquakes pose a regularly determined threat to man and his total surroundings . The

relatively recent earthquakes which devasted Anchorage, Alaska in 1964 and crippled San

Fernando, California in 1971 bear stark, vivid evidence to the vulnerability of our most

modern cities and communities. Earthquake destruction need not be widespread chaos and

disruption of life if we can learn to design our buildings, bridges, life lines, and

total systems to resist the forces caused by the abrupt crustal movements of the earth's

surface. Although complete life safety and zero property damage is a goal worthy of

achievement, as realists, we can only hope to approach that objective. Technology and

research capabilities are available today throughout the country to sucessfully support a major program of disaster mitigation to a respectable level to the advantage of the general

public and the nation.

The popular conception of the earthquake hazard in the United States often limits it to the Pacific Coast, especially California, and to such well known disasters as the San

Francisco earthquake in 1906, Long Beach in 1933, the Alaskan Coast in 1964, and San Fer- nando in 1971. Major earthquakes are by no means unknown to the rest of the country where

seismic events have occurred in the St. Lawrence River region on several occasions, in the

vicinity of Boston in 1755; the central Mississippi Valley (New Madrid, Missouri) in 1811

and 1812; Charleston, South Carolina in 1886 and Hebgen Lake, Montana in 1959.

The dense population of the United States , particularly in those areas of substantial

earthquake 'vulnerability, is a relatively recent development. While there have been

many reported earthquakes prior to 1860; the quality of the reported characteristics and

resultant damage associated with these events is variable and unreliable. During this

period, 36 damaging earthquakes have occurred resulting in cumulative property losses

of two billion dollars and over sixteen-hundred lives lost.

In estimating earthquakes losses, it should be noted that property damage is only one aspect of the loss which accrues. Other important losses include; loss of life, injuries,

economic loss due to injuries, loss of income due to business disruption, cost of emergency operations, and many others. There is little available data on the extent of these

VI-2B
costs although they most certainly exceed the direct costs involved.

Losses from earthquakes are not limited to the direct effects of ground faulting and

shaking. Tsunamis are sea waves generated by submarine crustal disturbances, often

associated with earthquakes and volcanic eruptions. They are scarsely perceptible at sea, but as they enter the coastal margin their crests may build to great heights depending on

the energy stored in the waves and on the coastal configuration. Tsunamis are limited in

destructive effects to areas immediately adjacent to the coastline where they destroy by

the impact of water and the subsequent inundation. Affected areas are not necessarily

proximate to the earthquake site. For example, in 1960, Hilo, Hawaii suffered from the

Chilean earthquake and Northern California was hit by the effect of the Alaska event. Thirteen twentieth century tsunamis affected the United States with a loss of 400 lives

and property damages of more than one-hundred sixty million dollars.

Although the most earthquake prone areas of the United States are along the CirciamPacific belt, many other sections of the country are also affected. Parts of the coastal

portion of California are among the most densely populated and rapidly urbanizing sections

of the country and therefore become more vulnerable. Much of the rest of the United

States has had notable seismic events also. Maps have been developed by seismologists and

engineers to delineate in a approximate way zones of earthquakes "vulnerability". From a study of the maps and population distribution, it was found that thirty-five percent of

the population of the United States resides in areas subject to major or moderate risks.

Table I presents a population analysis by risk zone. Thirty-nine states have regions

that are in either zone 2 or 3 or both.

We need only to look at the record of losses of two relatively recent earthquakes

to realize the impending devastation which is now lying dormant, and would be quickly

awakened, if another strong motion earthquake were to occur somewhere in the United States.

The San Fernando Earthquake of 1971 and the Alaska Earthquake of 1964 have the following

statistics

:

San Fernando 65 Lives Lost ^;511 Million, Property Damage

Alaska 131 Lives Lost $500 Million, Property Damage

Although the magnitudes for lives lost and property damages appear to be low, we must

recall that the Alaska event was in a sparsely populated area and the San Fernando region

is a residential siiburto of Los Angeles. If the San Fernando shock had struck closer to

the business district of Los Angeles, the gruesome statistics may have been more frightening.

Although there was an intervening period of seven years between the San Fernando and

Alaska Earthquakes, evidence of damage and study of design and construction practices appear

to indicate that very little new information was brought to bear on the methods of design

and construction. In the general codes for buildings and bridges only a few small changes

were noted, except in those areas where earthquakes are generally considered to be a more

frequent occurrence. Other areas of the country generally remained complacent.

Estimates of an annual investment in construction in the United States are in excess

of fifteen-billion dollars in regions of high seismic activity and the year two-thousand

VI-29
an investment of three-hundred-seventy-five billion dollars of new construction will be

exposed to the possibility severe earthquakes. It is prudent to hope that in the interim

buildings will be designed for maximum safety, economy and functional operation.

Estimates of future damage to be expected in the year two-thousand, if a high intensity earthquake were to reoccur, are indicated below:

San Francisco Property loss $20 Billion, Lives Lost 10,000

New Madrid Property loss $ 6 Billion, Lives Lost 3,000

If the New Madrid event were to shift northward by one hundred miles, the damage

estimate would be tripled because a more populated region including Chicago would be

effected.

Other factors would also affect the estimates, such as increased growth rates, greater

population density, construction standards used in the next twenty-five years, the extra

physical wealth, the ages of buildings, the materials and quality of construction. All of

these factors may not be controlled by good engineering practices, but those factors that

can be controlled should be undertaken soon to mitigate all possible losses.

To assist the professionals in their efforts, research is underway by many organizations, private and public, which will lead to a better understanding of earthquakes

forces and their effect on man made structure. Foremost among the organizations which

sponsor research in earthquake engineering are the various agencies of the Federal Government of which the National Science Foundation (NSF) is a prominent participant. NSJ

supports basic and applied research projects principally at universities. The program in-- eludes two major. areas of interest, Geophysics and Earthquake Engineering.

The NSF Geophysics program supports basic research in the areas of seismology, gravity,

geodesy, geomagnetism, paleomagnetism, earth currents, heat flow and materials properties.

The dynamic process of the earth's crust leading to earthquakes are studied by investigations in plate tectonics, crustal mechanisms, faulting mechanisms, local material re- sponse and propagation characteristics and location of the events. From this research,

realistic assessments of the risk of an earthquake at a particular site and in a given

time span can be pursued.

The NSF Earthquake Engineering program supports a broad, basic and applied research

activity in all aspects of engineering, geotechnical geology, social and behavioral sciences

which are aimed at developing methods and procedures for the mitigation of earthquake

disasters. as such, the NSF program is the lead Federal agency in earthquake engineering

research

The general objectives of the NSF Earthquake Engineering program are to support

research projects leading to:

1. Development of economically feasible design and construction methods for

building earthquake resistant structures of all types.

2. Development of procedures for integrating information on seismic risk with on- going land use planning processes.

VI-30
3. Development of an improved understanding of social and economic consequences of

individual and community decisions on earthquake related issues.

4. Presentation of program results in forms useable by the affected interest

communities to control the vulnerability to earthquakes.

The general objectives are supported by several specific technical program elements

which are described below. These elements have been developed to treat the specific

technical problems of analysis, design, and synthesis of engineering and scientific knowledge to control the consequences of earthquakes which are measured in terms of life loss,

property and functional loss.

2. Ground Motion and Data Services

Destructive ground motions resulting from earthquake actions are of several different

types, including heavy ground shaking, slow or rapid fault slip, subsidence and land slides.

Rindamental to an understanding of any of these damaging phenomena, however, is an accurate

knowledge of the actual earthquake ground motion. Fundamental to the validation of structural design and analysis procedures is the actual structural response generated by the

ground motion.

The measurement of destructive ground motions in the epicentral region of large

earthquakes and the associated response of structures is achieved by placing a network of

instruments at a variety of sites where earthquakes are likely to occur. These instruments

include both passive and active recording devices. The strong motion accelerograph record the history of acceleration in three component directions. It begins recording after

an acceleration threshold has been exceeded. These instruments are the most generally used

types and find application in most activities of data gathering. Other active recording

systems include pore pressure gauges to measure the liquid pressure in saturated soils,

with pressure gauges to measure the inertial effects of soil usually on a foundation wall,

strain gauges and displacement meters. Among the passive instruments are the seismoscope,

a conic pendulum that records motion on a smoked glass, scratch strain gauges and extensometers

.

The measurements obtained are directed at achieving three principal objectives:

1. To support the research program by measuring pertinent quantities to validate,

calibrate and/or formulate theories of earthquake response

.

2. To support the designer's need for earthquake motion information at varying

geological and seismological sites.

3. To obtain a comprehensive data base to perform microoregionalization of earthquake risk areas based on events in the area.

3. Soils Response and Analysis

One of the greatest potential sources of property loss is damage to structures that rest

on soils or on foundations which, although adequate to support the structures under ordinary

circumstances, might fail during an earthquake. A great potential for loss of life resides

VI-31
in the possibility of a dam failure. The following areas represent forms of soil failure

under consideration by the program: settlement of cohesionless soils, bearing capacity

failures, embankment failure; soil liquefaction, and water front bulkhead failures.

The dynamic behavior of structure during an earthquake depends on the shaking transmitted to it by the surrounding soil. These motions are imparted to the structure through

the interaction between its foundation and the supporting soil and/or bedrock and include

amplification effects.

The objectives of this element are to:

1. Develop methods to evaluate and control soil failure potentials.

2. Develop analysis and design methods to evaluate and control soil amplification.

3. Develop design and analysis methods to characterize the loadings transmitted to structures through soil foundation interaction.

4. Structural Response and Analysis

The realization of a structure rests on two complementary activities, analysis and

design. Analysis forms the basis for design. The ability to analyze a hypothetical

structure and determine the stresses and displacements that would be produced by a specified loading is an essential part of the design process. The more accurately this can be done the resulting design will be more efficient and economical with a more reliable

design factor of safety. The analysis of a structure can be exceedingly difficult because ordinary structures are exceedingly complex dynamic system; and because the ground

motions which the structure will be subjected to during its lifetime are probabilistic,

and the construction process leads to a structure which is not precisely known. The

analysis of structural response requires knowledge of the full system, including foundations, adjacent soils and in some cases the properties of adjacent structures.

The objectives for this program element are to:

1. Determine appropriate models for element (e.g., beam, column, plate, etc.)

response to strong motion excitation from analytical and experimental studies.

2. Develop digital computer methods of analysis that predict earthquake response of

structural systems comparable in complexity to real structures.

3. Validate and calibrate these models by comparison with earthquake measured motions

and damage of structural systems and elements.

The basic problem of earthquake design is to synthesize the structural configurations; the size, shape, and materials of the structural elements; and the methods of

fabrication, so that the structure will safely and economically withstand the action of

earthquake ground motions. The object of design is to control the effects of an earthquake

on a structure and keep damage within acceptable limits. A further series of objectives

for this element are:

4. Determine the dynamic properties of structures, elements and materials under

conditions of large strains, beyond the yield point and up to failure.

VI-32
5. Develop reliable, practical and simplified methods of earthquake design for

widely used and special important structures: for example low rise residential, low rise

commercial, school houses, high rise buildings, dams, bridges and industrial structures.

6. Thoroughly study structures that failed during an earthquake, as well as those

that did not fail, to refine design principles.

7. Develop design and analysis methods for non-structural elements and mechanical

components of structures.

5. Systems Response

Previous elements of the research program have dealt almost exclusively with localized

structures, that is, a single building, dam, etc. The operation of the community during

the emergency and recovery periods is dependent on the functioning of utility and public

service facilities that function as a system with elements located at many sites in the

affected area. The failure of an element can cause the total system to malfunction or be

inoperative. Thus the design of system elements must consider the seismic characteristic

requirements of the extended system. Both physically connected and non-connected systems

are to be investigated. An example of a connected system is fire water distribution

(storage, pumping stations, water mains, etc., while a non-connected system is represented

by emergency health care facilities (hospitals, clinics, laboratories, etc.).

The objectives of the element are to:

1. Develop principles of planning and designing facilities for minimum disruption

of operations due to earthquake ground displacements.

2. Develop design procedures for special structures and equipment of each type of

utility and public service facility, among these are:

a) Fire fighting and emergency transportation

b) Emergency power and communications

c) Hospitals and emergency agencies

d) General Communications

e) Water services and sewage disposal

f ) Electrical power and natural gas or fuel supplied

g) General transportation facilities.

6. Coastal and Inland Waterways

An earthquake at sea may generate a tsunami, ( tidal wave as it is commonly known),

that presents a real danger to coastal and island regions of the United States. The risk

posed by tsunamis may best be controlled by land use regulation in vulnerable areas. Verification of the occurrence of a tsunami permits the evacuation of potential affected areas.

If a dam fails, an inundation wave may be generated when the reservoir empties, and

can cause serious downstream damage. Structuring downstream land use, with this potential

hazard in mind can reduce life and property loss. When a reservoir behind a dam is filled

a series of earthquakes often occurs in the vicinity of the impoundment. The increment

risk these quakes cause above the natural risk is at present uncertain.

VI-33
The objectives of the element are to:

1. Develop methods to verify that an ocean-based earthquake has generated a tsunami.

2. Develop methods to anticipate tsunami run-up in coastal regions.

3. Develop methods to anticipate downstream inundation levels.

4. Determine the incremental seismic risk associated with filling a reservoir and

the degree that this risk should be part of the design criteria for the dam.

7 . Technology Transfer

Within the Earthquake Program, utilization activities have been centralized and formulated as a specific program thrust. The objectives in this element are to:

1. Maintain a technical reference collection of published reports, papers and un- published data available to researchers and professionals, to provide for information

dissemination.

2. Maintain a software center to archive, document, validate and distribute computer

programs developed for earthquake engineering applications.

3. Consolidate the best knowledge regarding design and analytical methods from

current and completed research and professional experience in building codes, design

criteria and standards for use by professionals and regulatory bodies.

4. Conduct regular conferences and workshops to act as foci for dialogues between

the research and professional communities.

5. Conduct post-earthquake inspections to obtain engineering data and to identify

unfulfilled research needs.

8. Implementation Studies

Advances in civil engineering methods are implemented through two types of actions

individual professional applications; and adoption of regulatory or enforcement agencies.

The latter process is primarily political, while for former is educational. The structuring

of the flow of technological developments from research to implementation is a particularly

arduous one, involving a vast complex of institutions, special interest groups and secondary agendas. The way in which this process works is not too well understood by most

people. An understanding of the process is a first step in structuring a methodology to fore-shorten the period from research to practice. The objectives of this program element

are to:

1. Develop an understanding of the implementation process.

2. Develop a methodoJ.ogy to program the education of professionals to the best

earthquake engineering practice.

3. Develop a strategy of program results to be processed through the communi.ty into

regulatory and enforcement agency regulations in a timely fashion.

4. Advocate the accomplishment by public and private groups of the developed plan.

VI-34
9. Social and Behavioral

The mitigation of earthquake effects may depend on a variety of actions not directly

associated with the assurance of safe structures. The possibilities of earthquake control

and prediction present different technical methods that may lead to mitigation. Community

actions during and following the event may significantly affect the consequences of the

quake. Condemnation and redevelopment attitudes, insurance programs and economic assistance may markedly affect the vulnerability of a community to future events. With a view

to filling important gaps in our understanding of earthquakes as they impact man and his

social environment, research in the social and behavioral sciences, and in economics, will

be undertaken to:

1. Develop policy alternatives for the community to provide emergency, rescue, recovery and redevelopment services.

2. Develop policy alternatives for the dissemination of earthquake warnings.

3. Assess the consequences of technical developments and policy alternatives on the

achievement of disaster mitigation.

New areas of research to be undertaken and initiated in the coming fiscal year will

extend the NSF Program to fill several important voids in knowledge which have been identified as essential elements to the research program.

Particular emphasis will be placed on the following general topics:

1. Development of vulnerability methods for assessing the risk posed by structures

not designed and constructed to be earthquake resistant.

2. Extension of experiments begun in 1976 on the three-dimensional non-linear behavior of structural components to include typical building elements. Current design methods

are based on two-dimensional failure models.

3. Research initiations on the design and analysis of earthquake performance of lifeline facilities, utility, transportation, and communication systems which are critical to

the performance of the community.

4. Expansion of FY 1976 activities to develop an understanding of the impediments

of the incorporation of research results into seismic hazard regulations for construction.

5. Research initiations to discover how private vulnerability decisions are made

and may be affected, since business decisions are a major factor in developing the physical

I environment and, thus, affect the seismic vulnerability of entire communities.

In addition to the earthquake forces other environmental loads are closely related to

the design of structures by their dynamic nature. These load effects are: extreme winds,

accidents, explosions, projectiles, wave actions, surges, and floods. These dynamic hazards account for an annual financial loss in excess of $1 Billion. The nature of these

hazards is reasonably well known and the analytical and design methods developed for

seismic resistance can be adopted to these additional forces at a relatively small cost. The research related to these dynamic hazards will; adopt, where appropriate,

seismic and analysis methods to other individual dynamic hazards to buildings and develop

proceedures and methods that integrate the effects of multiple loads sources and types and

that are suitable for professional use and regulatory adoption.

VI-35
Several other Government Agencies are engaged in research problems involving earthquake engineering which effect the successful operation of their day to day mission assignment. Close cooperation between agencies and departments assure that duplicative effort

by academic researchers is eliminated as much as possible.

A review of the program technical elements described above indicates the breadth and

scope of the NSF Earthquake Engineering Program. It is a broad program designed to provide

the necessary research to develop improved building design provisions, design criteria,

evaluation of risks, and preparation for an event should it occur anywhere in the United

States. The main thrust of the program is to mitigate the losses from earthquakes which

may occur anywhere throughout the country. With the assistance of the academic researchers,

engineers, architects, building officials, code writing bodies, contractors and other

future earthquake events may be only simple momentary disruptions rather than devastating

chaos

VI-36
LARGE-SCALE TESTING PROGRAMS

RELATED TO WIND AND SEISMIC EFFECTS

CURRENTLY UNDERWAY IN JAPAN

SEIICHI INABA

Earthquake Engineering Laboratory

National Research Center for Disaster Prevention

Science and Technology Agency

Japan

ABSTRACT

This list has been prepared in order to discuss cooperative research problems in the

area of large-scale testing of structures at the 8th Joint Meeting of U.S. -Japan Panel

on Wind and Seismic Effects. The Japan Panel selected for following four members to work

on the task committee on the large-scale testing program.

Seiichi Inaba, National Research Center for Disaster Prevention

Makoto Watabe, Building Research Institute

Kenkichi Sawada, Public Works Research Institute

Nobuyuki Narita, Public Works Research Institute

In accordance with the exchange of letters between Dr. C. Culver of the National Bureau

of Standards of the U.S. Panel and Mr. S. Inaba of the Japan Panel, this list has been

drafted to inform the U.S. -Japan Panel of work being conducted in this area. The list

includes the large-scale testing programs currently underway in Japan, the organization,

the name of individuals in charge, and a brief description of the objectives, status and

time schedule.

Key Words: Earthquake. Research Programs j Seismic. Testing j Wind.

1. Program ; Dynamic Tests of Structures by Use of A Two-Dimensional Shake Table

Organization : National Research Center for Disaster Prevention

Names in Charge : Seiichi Inaba, Nobuyiiki Ogawa; Earthquake Engineering Laboratory.

Description of the Objectives : It is planned to study the effects of the vertical earthquake shaking of structures using the two-dimensional shake table, which is going to be

built in the National Research Center for Disaster Prevention. The table will be 6 m x 6 m

and have a maximim test weight of 60 tons which can apply simultaneous vertical and horizontal vibrations with a maximum displacements of +100 mm in the horizontal and +50 mm

in the vertical. Maximum velocities of 60 cm/sec in the horizontal and 40 cm/sec in the

vertical, and a maximum accelerations of 1.2 G in the horizontal and 0.8 G in the vertical

direction can be achieved as well as a frequency range between D.C. and 30 Hz. Dynamic test programs have been planned for large-scale models of embankments and

other soil structures, building structures, plant structures, foundation structures and

so forth.

VI-37
status: Soil exploration of the foundation site was performed in 1974. A feasibility

study of the two-dimensional shake table has been conducted. A small-scale model simulator of the table, 1 m x 1 m, was constructed in March, 1976.

Time Schedule : Designs of the shake table, foundation and laboratory building are being

conducted in 1976, and the construction of foundations will start in 1977. Dynamic tests will start in 1980, after completion of the construction.

2. Program : Dynamic Tests of Composite Breakwater

Organization : Port and Harbour Research Institute , Ministry of Transport

Names of Individuals in Charge : Setsuo Noda, Tatsuo Uwabe, Hajime Tsuchida, and Satoshi

Hayashi

Description of the Objectives : Recently, several high composite breakwaters are under

planning in Japan. The composite breakwater consists of a rock-fill or sand-fill mound

and concrete caissons resting on the mound. The largest breakwater of this type in Japan

has been constructed in Ofunato Bay, to protect the coastal area along the bay from hazard

by tsunami; the breakwater is 40 meters in height. The breakwater against tsunami must

have sufficient seismic stability, since it is certain that the breakwater will be shaken

due to earthquake ground motions preceding tsunami arrival.

The Port and Harbour Research Institute (PHRI) is planning to examine the seismic

stability of the composite breakwater by model tests. At present tests are expected to be performed on a shaking table in PHRI, which is an electric-dynamic random shaking

table having frequency range from 0.2 to lOOHz. The size of a container on the table is

5 meters in length in the direction of table motions, and is 1.5 meters in width and 1.5

meters in depth. A wave absorbing channel can be connected with a flexible joint to the

container to eliminate disturbance by waves generated and reflected by an end wall.

Through the tests stability of rock-fill mound, dynamic water pressure to the mound

and the caissons, and relative displacement between the mound and the caissons will be

studied. PHRI has been observing earthquake ground motions at the site of the tsunami breakwater in Ofunato Bay and its earthquake response. The records obtained during the observation will be analysed as a part of the study.

Status : PHRI has some experiences on dynamic tests of such a composite breakwater; in the

tests, however, a small model was used and the model was tested in air.

The earthquake ground motion records at the site of the breakwater and records of its

earthquake responses were once briefly analysed and reported in the following report:

Hajime Tsuchida, Tatsuo Uwabe, Eiichi Kurata, and Satoshi Hayashi: Observation

and Analysis of Earthquake Response of Fill Type Breakwater, Proceedings of the

Japan Earthquake Engineering Symposium 1973 in commemoration of the 50th

anniversary of the Kwanto earthquake, August and September 1973, pp. 375-364.

Time schedule : The tests are expected in late 1976 or in 1977 Japanese fiscal year (from

April to March)

VI-38
3. Program : Experimental Investigation into the Seismic Behavior of Full-size Reinforced

Concrete Structures Utilizing the Large Scale Testing Facilities.

Organization : Building Research Institute, Ministry of Construction

Names of Individuals in Charge : M. Watabe , M. Hirosawa, S. Nakata.

Description of the Objectives : Up to date many experimental studies were made on the

columns, beams and shear walls. However there are many fields still unexplored; such as

the stress distribution among each member of structures at the variable plastic stages,

yielding mechanisms of a structure and ductility of each member during yielding and development of mechanisms in the structure. For example the aseismic tests (static and dynamic)

of a full size structure with slabs, shear walls and foundations need to be conducted.

These test results would make clear the difference between the behavior of a structure and

a unit member, especially the anchorage of main bars or ground effects on walls. Hitherto

laboratory studies on the structural damage due to earthquake were based only on member

tests, and the behavior of an entire structure had to be based on engineering judgement.

Such judgements may lead the serious errors. Therefore, full size structural testing

should be made in order to avoid such errors

.

As a first stage of the testing project, we propose to test a 3- story building with

shear walls. The kind of shear walls will include normal types, precast walls and slit

walls. By performing this test the behavior of shear walls, effects of foundations, soil

conditions and mechanisms during structural yielding can be explored. Various types of

buildings, such as follows, will also be included in the project, i.e., structures without

shear walls with short columns or long columns, long span structures and many types of

ground conditions

.

The Large Scale Loading Facility of the Building Research Institute, can be used for

static and dynamic tests of full size structures by use of hydraulic sero actuators, which

are now under construction and can be used by 1979. All the testing systems, loading, vibration and measurement are directly controlled by a computer system in this facility.

The total cost of this project is expected to be 12 billion yen (40 million dollars)

.

Status : M. Watabe et al. , "A Proposal to the Joint Research Program on the Aseismic

Properties of R.C. Structure by the Full Size Specimens." 7th Joint Meeting U.S. -Japan

Panel on Wind Seismic Effects, UJNR, Tokyo, May 20-23, 1975.

Exchange of Research Personnel : 1976 E. Pfrang (N.B.S.), A. Sozen (Univ. of Illinois),

R. Hanson (Univ. of Michigan) , H. Umemura (Univ. of Tokyo) , H. Aoyama (Univ. of Tokyo)

,

M. Watabe (B.R.I.) . 1977 J. Penzien (E.E.R.C.), R. Clough (E.E.R.C.), P. Jennings

(T.I.C.), T. Okada (Univ. of Tokyo), M. Murakami (chiba Univ.), Y. Yamazaki (B.R.I.)

1978 M. Duke (U.C.L.A.), E. Popov (E.E.R.C.), M. Hirosawa (B.R.I.) , A. Shibata (Tohoku

Univ.), Y. Matsushima (B.R.I.) . 1979 R. Bresler (E.E.R.C.), J. Penzien (E.E.R.C.),

I R. Clough (E.E.R.C.), Y. Ohsawa (E.R.I. , Univ. of Tokyo), S. Nakata (B.R.I.)

VI-39
4. Program : Dynamic Test of Structures for Nuclear Power Plants

Organization : Center for Nuclear Safety Engineering Research

Names in Charge : Resources and Energy Agency, Ministry of International Trade and Industry

Description : Construction of a shake table for testing of nuclear power plant structures

is planned in order to prove the safety of nuclear power plants during and after future

destructive earthquakes. Part of the funds have been obtained from the government since

1975.

The structures that will be dynamically tested will include containers, pressure

vessels, elements of reactors and piping of main-coolant system for both PWR and BWR

nuclear power plants.

The shake table under planning will have dimensions of 15 x 15 m, maximum test

weight of 1000 tons, maximum horizontal stroke of + 200 mm and a vertical stroke of +100

mm. The table can be driven with simultaneous vibrations in the horizontal and vertical

directions up to accelerations of 2 g in the horizontal and 1 g in the vertical with a velocity of 70 cm/sec in the horizontal and 35 cm/sec in the vertical direction.

The total cost of construction, including the shake table, foundation and the laboratory building is expected to be 24.8 billion yen (83 million dollars) and the construction

cost of a test structure will be about 2 billion yen (7 million dollars)

The model scale of the test structures will be 1/4 for a container of 800 NWe PWR and

1/2 for that of a 1100 MWe BWR power plant.

Status : Construction of the table and the research work will be carried out by the Center

for Nuclear Safety Engineering Research, which has been established under the control of

the Resources and Energy Agency in February, 1976. The center is a joint governmentprivate foundation supported by the government and the electric power companies , constructors and manufacturers.

Time Schedule : It will take 5 years to build the shake table and the laboratory and

other testing facilities. The shake tests will begin in 1981.

5. Program: Dynamic Experiments on Earthquake Resistance of Civil Engineering Structures

Organization : Public Works Research Institute, Ministry of Construction

Name in Charge : Ground Vibration Section and Earthquake Engineering Section, Public Works

Research Institute, Ministry of Construction.

Description : Construction of a shake table is planned to prove the earthquake resistance

of engineering structures such as suspension bridges, prefabricated submerged tunnels,

underground pipelines, embankments, etc. There are two kinds of shake-table facilities that have been planned. One will have

the dimension of 6 meter x 8 meter, maximum test weight of 130 tons, maxim\.im stroke of

j;^75 mm, maximum velocity of 60 cm/sec, maximum acceleration of 0.7 g in the horizontal

direction. The other will have 4 shake tables, and each table will have the dimension

of 2 meter x 3 meter, maximum test weight of 30 tons, and other specifications that are almost the same as the above large scale shake table. These two kinds of shake tables

will use 4 common actuators.

VI-40
The total cost of construction including the shake tables, foundations and laboratory

building is expected to be 12. billion yen (4 million dollars).

Status : Dynamic tests of submerged tunnels and underground pipes, using the large scale

shaking table which belongs to the Earthquake Engineering Laboratory, National Research

Center for Disaster Prevention, Science and Technology Agency, had been conducted several

years ago. An experiment on the dynamic behavior of sandy grounds during earthquakes are

now being conducted, using the shake table which have the dimensions of 1.5 meter x 2 meter,

maximum test weight of 5 tons in the horizontal direction.

Time Schedule : The construction of the shake tables and the laboratory and other testing

facilities at Tukuba Science City will start in 1977, and it will take 2 years to complete

a part of those facilities. Dynamic tests are expected to begin in 1979.

6. Program : Dynamic Test of the Suspension Bridge Foundation Model

Organization : Honshu- Shikoku Bridge Authority

Name in Charge : Keiichi Komada, Chief of the Third Design Division

Summary : In order to prove the safety of the bridge foundations at the time of an earthquake, dynamic model tests by use of a large scale shake table is planned in 1976. It is

aimed to clarify the interaction between the foundation and ground. Thus the response

of the foundation and ground will be measured for various depths of the foundation model.

An outline of the test and analysis is as follows

:

1. Model:

Foundation Model: 60 x 80 x 100 cm (Weight: 1 t)

Ground Model: 400 x 900 x 150 cm

2 . Exciting Waves

:

Sinusoidal Wave, Random Wave and Earthquake Wave

3 . Measurements

:

Acceleration Response of the Foundation and Ground

4. Analysis:

Experimental Analysis of Resonance Curves , Vibration Modes and Others

Theoretical Analysis of Response by the Finite Element Method

Status : Honshu- Shikoku Bridge Authority has been ececuting the dynamic test in order to

develop an earthquake proof design for various types of foundations, since 1971. These

details are shown in Table 1.

7. Program : Dynamic Test of Embankment

Organization : Public Works Research Institute

Names in Charge ; Kenkichi Sawada, Yasuyuki Koga: Soil Dynamics Section

Description of the Objectives : The object of our research is to explain the dynamic

behavior of an embankment, using the large-scale shake table established at the National

Research Center for Disaster Prevention.

VI-41
The height of the test embankment is 1.5 m and the depth of the ground base is 1.5 m. The total length of the model is 12 m perpendicularly to the embankment axis and the width

is 2 m. In addition to this basic model, embankments protected with the steel sheet piles

and horizontal sand berms are to be tested. The model embankment is constructed in a water tight steel box on the shake table and the model is perfectly or partially submerged

in water, depending on the test program.

The test of the embankment of a saturated ground soil is a fundamental and important

case. Data on the increase of pore water pressure and acceleration of the model can be

obtained. Analysis using the slice method to evaluate the stability of the embankment

can be applied to explain the slope failure due to the dynamic vibration in which the

effects in the increase of pore water pressure on embankments and ground soil is considered.

Analysis will also be performed in the liquefaction problem under anisotropic stress conditions . Status : Results of this test program have been published in a Technical Report of Public

Works Research Institute and presented at the Annual Conference of the Japan Society of

Soil Mechanics and Foundation Engineering. A paper on the settlement of saturated sands

due to vibration was presented at the UJNR U.S. -Japan Joint Panel on Wind and Seismic

Effects in 1975.

Time Schedule : We have been engaged in the research program for 5 years beginning in 1970.

Our future program is to examine the behavior of vertical vibrations of embankments , soilfoundation interaction, and soil structures constructed with other materials such as steel

piles, concrete slabs, and retaining walls.

8. Program : Aerodynamic Test for Long-Span Bridges and High- Rise Towers

Organization : Public Works Research Institute, Ministry of Construction

Names of Individuals in Charge : N. Narita, Chief, Structure Section and K. Yokoyama

Res. Eng., Structures Section.

Description : Two large wind tunnels are now being constructed. The main feature of these

wind tunnels is as follows:

1. Wind tunnel for flutter testing of bridge and tower structures:

2. Wind tunnel for aerodynamic testing of bridge and tower structures under turbulent

Dimension of test section: 2.5 x 4.0 m (10 m long)

Maximum wind speed

25.0 m (without coarse grid)

flow

Dimension of test section: 6.0 x 3.0 (30 m long)

Maximum wind speed: 15.0 m/s

3. Location of the facilities

Tsukuba Science City in Ibaraki Prefecture about 60 kilometers northeast of

Tokyo

4. Completion

The construction work will be completed in May 1977.

VI-42
The following wind tunnel tests are scheduled to be carried out:

1. Measurement of aerostatic and aerodynamic forces acting on bridge or tower

structures

.

2. Measurement of critical fluttle speed of the structures.

3. Measurement of aerodynamic response of the structures under gusty and turbulent

wind.

Acknowledgments

The following three experts have cooperated with our task committee on the preparation of the list.

Hajime Tsuchida, Port and Harbour Research Institute

Yoshio Komada, Honshu- Shikoku Bridge Authority

Yukiyoshi Ibe, Resources and Energy Agency

The author, serving as co-chairman of the task committee, is responsible for the entire

work in the preparation of this list, contact with the U.S. Panel, and proposals to the

Japan Panel.

Table 1 Details of Experiments

Year Type of Foundation Groxind Katerial Measurement

1971 Rigid Foundation and

Mult i- Column Foimdation Rubter Acceleration and

Displacement

1972 Rigid Foundation end

I-'ulti-Column Foundation

Grouting

Material

Acceleration and

Displacement

1973 Rigid Foundation and

Rigid Foundation witii Piles

Grouting

Llaterial

Acceleration and

Displacement

1974 Rigid Foundation Sand Acceleration,

Displr.cement

,

Soil Pressure and

Pore Pressure

I

!

i

I

VI-43

6i
EARTHQUAKE DAMAGES TO EARTH STRUCTURES

KENKICHI SAWADA

Director

Soil Dynamics Section

Public Works Research Institute

Ministry of Construction

Japan

ABSTRACT

This paper will present the problems associated with the damage to earth structures

during earthquakes, and how some of the problems have been solved. In addition this paper

will present the principle of the design of earth structures, that to date have not been

discussed.

Initially, data has been collected on the amount of damage to earth structures during

earthquakes, how they have failed and relationships to such failures. These data are then used in formulating the mechanics of the failure.

Finally, the details of the repair work required, as listed by field engineers

engaged in this repair work was examined.

Key Words: Construction; Dams; Earthquake; Earth Structures; Roads; Survey.

1. Foreward

We can investigate the damage caused to earth structures by recent earthquakes by

actually visiting the site. However, details of past failures and damage can only be

obtained from literature, but little quantitative data is available.

In this report, details of earthquake damages to soil structures, both past and

present, will be given. The earth structures will include sliding of slope, settlement

of embankment and displacement of retaining walls.

2 . Damage

The expression "There are many cracks in the ground and the land slide dam up river

and overflow has caused a flood" which relates to a geophysical phenomenon given in

literature 1000 years ago. This indicates that the mechanism of the disaster in the

Edo-period resembles todays earthquakes. During this period people constructed river levees

extensively, and many people lived near the river. Therefore the scale of disaster for

these geophysical structures was quite extensive.

Fire was propagated due to earthquakes during this period. This was due in part by

the thickly populated areas with houses and the type of fuel used. "Expanded disaster

was the result of human activity" has generally evolved as a truism, where today we find

the fire and flood associated by earthquakes.

VI-44
The utilization of information, collected during failure of buildings and bridges has

permitted establishment of specifications during the past 30 years. However, these

specifications are limited to only steel and concrete structures.

During this period, several large scale geophysical disasters have occurred, such as

the Kanto earthquake ('23 M=7.9) in Nebukawa along the coast of Sagami Bay where a land

slide (H = 90 m L = 2000 m) occurred, resulting in the soil pushing a standing train into

sea. Also the Kitaizu earthquake ('30 M=7.0) induced failure of an old dike reservoir,

the Nankai earthquake ('46 M=8.0) in Yanokawa along south part of Kii peninsula, created

a large scale slope failure (H = 80.8 m I = 1/25) . In all instances crack and settlement

of road embankments were noted, however, the resulting reports describing these earthquakes, contain no quantitative data.

After the Fukui earthquake ('48 M = 7.3), valuable engineering analysis began,

resulting in procedures in the collection of earthquake data. The epicenter of the Fukui

earthquake was in the Fukui alluvial plain, therefore damage was limited in this plain,

which is one reason why the investigation was simple to perform. Now we know the type of

serious damage which can occur to the earth structures during earthquakes, which include;

failure of roads, railway embankments, river levees, slidings of sand hills and overturning

of retaining walls. The maximum volume of soil, to be removed during these failures was

several thousand m^.

At the Niigata earthquake (-64 M = 7.5) many types of investigations in the earthquake

engineering field were conducted. The main theme of the investigation in soil mechanics,

was liquefaction of the soft sandy ground. The analysis of the settlement and sliding of

the road and railway embankment and river levee were due to the liquefaction of the ground.

The railway embankment (H = 7 m L = 150 m) was damaged and its material shifted 115 m, shows the severety of the damage.

Though similar damage to earth structures have occurred, it was the experience gained

by many engineers who were engaged in soil mechanics, that increased the state of the art. Although they were able to obtain a solution, research work was initiated after the '68 Tokachioki earthquake.

The '68 Tokachioki earthquake ('68 M = 7.9) was an important investigation relative

to damage of an earthquake damage, in which Liquefaction of an embankment occurred.

The precipitation of 100 - 200 mm, two days before the shock occurred was a condition

to soil liquefaction of the relatively uniform graded volcanic soil. The new embankment of

j the National railway suffered serious damages to this embankment which consisted of this

j volcanic soil. The maximum height of the embankment was 15 m. The national route No. 4 which was seriously damaged, was located at Metoki. The entire portion of the embankment

(H = 5 m L = 110 M) was constructed on a swamp, fallured and spread more than 40 m. Another case was damage to a hill side slope (H = 20 m) where the saturated soil spread

1 100 m and the ground of the middle school (H = 12.3 m I = 1/1.5) . Another case involved

an earth dam (H = 8 m L = 200 m) used for irrigation, spread 160 m.

I Even at the relatively small earthquakes, large scale earth structures were damaged.

During the Oitakenchubu earthquake ('75 M = 6.4) , a large slope of an embankment

VI-45
(H = 22 m L = 90 m) slid and stopped traffic flow along a toll road. In addition to that

special case, there are many other cases of land slides which have been increasing in num- bers . Land slides and ground fissure designed as an index Intensity VI of the JMA.

So in the area VI there have to be land slides and cracks. But recently there has been

an opinion that land slides and cracks should have an index of V. As already mentioned, there are many facts relative to damage, but little useful

data. In reporting disaster surveys, there are many ways in describing the frequency,

length and soil volume during failure. It is easy to attribute this reason to the complexibility of the soil composition and field conditions, but we must examine the real cause

more closely. Because of the inaccuracies in soil dynamics describing the behavior of

earth structures during the earthquake, we can not obtain accurate results. At present

it appears that a universal theory to predict the behavior of such structures is not possible.

3. Analysis

Next I will present the results of various analysis of the failure of earth structures,

and their differences. The first is a study by Professor Kawakami, of Tohoku University,

refered to the Miyagikenhokubu earthquake ('62 M = 6.5). In this case an embankment of a

2 road, railway and river levee concentrated in relatively narrow area (40 x 40 Km )

, were

damaged. The investigator attempted to gather stochastic data, which might explain the

influence of the soil property and aging of the soil to the scale of failure. As a result

he states that a seismic area VI existed in which almost all embankments were damaged

seriously, and in seismic area V there wasn't any damage. Further he presents a table

which gives relationships between the ratio of failure and the geophysical condition of

the base, the soil property of the alluvial deposit and the depth of the alluvial deposit.

I think that such arrangements need more data, such as the scale of the sampling

range on the stochastic results and moreover the method in arranging the data gathered

at different earthquakes must be established. These are important points of the stochastic

arrangement especially for the earthquake problems.

The next example is the slope stability analysis of a river levee damaged during the

Niigata earthquake. A slice method is used to analyze the dynamic problem. The river

levee (H = 6 m I = 1/2) as located at the left bank of the Sinano river. In selecting

the values of c and (|), which is the most important procedure in this analysis, a correlation

table between N and c, (J) is used. Supposing that the horizontal and vertical acceleration is 0.14, 0.1 and 0.15, 0.10 respectively, then the safety factor of the sliding was computed as less than 1.0. This static slice method for the dynamic problem, requires numerous solutions, but the vibration test, using the embankment model made of 40 m"^ sand

on the large shake table can clearify the validity of this method. Also the Institute

of National railway has conducted large scale embankment tests, related to the '68 Tokachioki earthquake damage. In order to reproduce the in-situ failure pattern and analyze

the numerical relation and assure the aseimic design of embankment, many tests were conducted.

VI-46
Data froip the Iziihantooki earthquake ('74 M = 6.9), which involved an embankment fail

ure is valuable as its base is hard. In the manual of road construction there is a table

giving the standard slope inclination and critical height for each soil property. Where

the soil property is classified into five categories these classifications are (1) well

graded sand, gravel soil, (2) poorly graded sand, (3) rock or gravel, (4) sandy soil hard

clay, (5) soft clay. Computing the C, and 0 using the slice method for each soil gives;

2 C = 1.0 t/m and 0 = 21° 9° 27° 15° 3° respectively. Also for the other soils gives;

2 2 2 C = 4 t/m and 37°, C = 3 t.m and 0 = 22° and C = 2 t/m and 0=7° respectively for III, IV and V soil. The above values can be considered design values, having a safety surplus.

Most embankments which failed due to the Izuhantooki earthquake, were made of soil I and

the embankment in Nagaturo is the largest, having a 15 m height. Using the above determined C, the 0 stability of the Nagaturo embankment was analyzed. The results give a static state safety factor of 1.66, which became 1.06 under vibration of a 200 gal acceler

ation, when the maximum horizontal shock acceleration is presumed to be 400 gal.

The research work on liquefaction is advanced in the earthquake area. If we limit the

problem to the criteria of liquefaction, we can use various data and get successful result

In the case of perfect liquefaction the soil has no shearing resistance, and thus, the phe nomenon is easy to explain. But there is an increasing rate of pore pressure, than there

is data.

Recent papers have provided various data concerning soil dynamics. However, little

has been done to apply these results to analyze the failure mechanism of earth structures.

The gap between knowledge in the laboratory and field practice is surprisingly wide.

As shown previously, the analysis of earth structures damaged by earthquakes has

advanced slowly, and there is minimal effort these studies have on the reconstruction

and design of new earth structures. The planning and reconstruction of damaged structures

will therefore, not be presented.

4. Reconstruction

In the paper concerning the Kanto earthquake, "the earthquake caused the roads to be

closed, but they were opened the next day" is described. How the field engineers were

able to provide such service, will now be described.

Due to the '68 Tokachioki earthquake, the National route No. 4, was seriously damaged

at Metoki, which is on the border between Aomori and Morioka prefecture. The road

embankment (H = 5 m) was constructed on a swamp and between the road and hill there was

a pond and another hill opposite to the pond beyond 40 m plain. At the time of the shock

movement the body of the embankment crossed the plain and ran toward the opposite hill,

and then the pavement fell. The embankment was constructed on a wood frame, usually used

in river works as the bearing capacity of the swamp was insufficient. Therefore, the

reconstruction work was primarily a problem of drainage. The reconstruction of the road

first required the pond to be drained, then the center line of the road was shifted off the

pond area and a large drain pipe was buried under the toe of the slope. Therefore, the

stability of the embankment could be improved. However, we can not judge the maximum

VI-47
Seismic Intensity that the embankment can withstand. It took 2 days for the temporary

repair design , 10 days for a complete design, and 60 days for the final construction.

The next paragraph will describe some episode as detailed by a bus driver. He was driving the bus on to the embankment when the earthquake occurred , at which point he

went back to the safety area. At that time, the embankment had just failed. At the same time in Hokkaido, one motor car had overturned as the wheel fell into cracks on the road

surface. Description of the later cases are given in an investigation report.

•Due to the Niigata earthquake, the National route No. 7, was damaged and the Ebigase

surface of the road settled and shifted about 4 meters as a result of liquefaction of the

base. Traffic was reopened in the same day. Temporarily, reconstruction consisted of

filling in of the cracks and settled area to maintain drainage and application of an asphalt coat to prevent dust. The design of the reconstructed pavement used the bearing

capacity of the temporary fill as the design value. In such a design, the pavement base

property which was liquefied at the shock, was not improved.

Therefore if an earthquake should occur in this area again, the same damage will

occur

Another example of the strengthening of a structure involves the river levee of

the Mogami River due to the Niigata earthquake. The usual method of repair is to fill

the levee with additional soil to the previous height. However, in this case, the base

soft clay material was replaced with good material. This method of reconstruction was tested by experiments in the laboratory, in order to determine how effective this method

was against future earthquake shocks.

The greatest change in road occurred during the Izuhantooki earthquake. The original

road passed through the bottom of a cliff near the seashore, but during the earthquake this

road was completely burried under the collapsed soil. The road was not reconstructed, but

a tunnel built instead. The opinion that tunnels can serve not only as roads but also as

a shelter against wind and rain is an effective decision. This is because this road and

the people living here were always threatened by land slides caused by rainfall and

earthquakes. There was no requirement that the repair work must be finished early, therefore enough time was available to discuss the plan of reconstruction.

Another example is the reconstruction of a high embankment which failed in the

Ohitachubu earthquake. There were several points to be considered in the planning of

this reconstruction. These are that the additional cutting of a hill must be avoided,

removement of the collapsed soil from the swamp must be avoided, thin over lays on the

slope of the embankment must be avoided and the results of a test boring must be utilized.

The detailed plans are as follows; 1) shift the center line, construct the embankment

with a slope having 1/1.5 inclination and build a buttress retaining wall (H=9 m) at the

bottom, 2) keep the centerline, and construct the embankment having a well crib wall and

construct the bridge. The second proposal was adopted. We can notice here that the plan

of execution was the main interest and there was no special attention to earthquakes and

thus nothing was attained.

VI-48
From examination of the above examples, we can understand the principle of how

earthquake damage to earth structures are treated. Design methods of earth structures,

considering dynamic effects, are not concrete. For field emergencies, the requirement

to repair quickly is an important factor in design. And thus the probability of damage to

this structure again by some earthquake is overlooked. In the case where failure of the

earth structure causes injury to many people and many days repair, it must be designed to withstand greater earthquake shocks.

5 . Final Remarks

Examining many principles, it can be concluded that the design of earth structures is

not established. Many earth structures have been constructed and have been damaged by

earthquakes. Analysis is not sufficient and the requirements for reconstruction have

not been consistent.

VI-49
DYNAMIC TEST OF A CIRCUIT BREAKER

FOR TRANSFORMER SUBSTATION

SEIICHI INABA SHEGIO KINOSHITA

Earthquake Engineering Laboratory

National Research Center for Disaster Prevention

Science and Technology Agency

ABSTRACT

During earthquakes, dependability of electric power supply systems, is required.

However, earthquake damage to the electric power industry has resulted from the structural

failure of porcelain insulators, which are commonly used for electric transmission equipment . Dynamic tests of a circuit breaker of capacity 72/84 kilovolts used for a transformer substation was conduced using the large-scale shaking table of the National Research

Center for Disaster Prevention in 1975. The test was performed under the sponsorship of

Meidensha, a Japanese Manufacturer of electric power equipment. The purpose of the test

was to determine the dynamic characteristics of the prototype structure shielded with the

porcelain insulator. Sinusoidal waves of resonant frequency and earthquake simulated

waves were applied to the test structure. It was found that the failure of the porcelain

insulator governs the seismic resistibility of the circuit breaker and the maximum allowable acceleration at the top of the structure is 8.0 g.

Key Words: Circuit Breaker; Data; Experiment; Shake Table; Testing.; Transformer.

1. The Objectives

The purpose of the dynamic test was to investigate the earthquake effects of a circuit

breaker. The three wave resonance method was applied to the test specimen. The three waves resonance method is a dynamic test procedure which induced three successive waves

at the same frequency as that of the resonant vibration of the test structure, then the

earthquake resistance is evaluated as defined by the first excursion failure.

2 . N-waves Resonance Dynamic Test

Since there are no general input waves during a vibration test, the selection of

appropriate input waves is required for each test. Determinate function waves, such as sinusoidal waves or earthquake simulated waves, are used for the test. Resonant vibration of the sinusoidal waves are also used as an earthquake loading. N-waves resonance

method gives N cycles of sinusoidal waves at resonant frequency of the structure and

provides an observation of the treatment transient state of vibration. The main purpose

of the dynamic test is to evaluate the structural failure defined as the first excursion

failure, where any deflection, strain, or stress of the structure reaches the fracture

VI-50
limit due to the dynamic force. While fatigue failure occurs during a stationary state of

a structure, the first excursion failure is for the transient state, in which N-waves reson- ance method is appropriate as the dynamic test procedure for evaluating the first excursion failure

.

a) N-waves Resonance Vibration of a Spring-Mass System with Damping

We shall consider here a viscously damped spring-mass system excited by harmonic

waves at resonant frequency. The equation of the motion can be written as;

Pi2 \ 2h + w

2 y . _ — 1 dt nndt n dt

u(t):= sin t7 t ; 0-< t <1[T = 2T11I/ tv 2

where w is the natural frequency in radians per second, h is the damping factor, y(t) is n n the relative displacement and u(t) is the foundation displacement. The solution is ex- pressed as

2 2 ^ -X- = cos vr t - -^^^^ sin ( ^il-\^V t + 6 )exp(-h v;^t)-3 dt'^2n n 2 an n u ^

n

where D , and e are written as

n n

4

The maximum response, in terms of acceleration amplification during N-waves of resonance

vibration, is;

n

For the stationary state of resonant vibration.

Computed values of A (h ) are shown in Fig. 1. N n

b) Analysis of the Porcelain Insulator

A circuit breaker is constructed with a frame and a porcelain insulator, which is

divided into two parts the breaker insulator and the supporting insulator. An analytical

model of a continuous system is shown in Fig. 2 where the equation of vibration excited

by the harmonic motion is written as.

VI-51
where u(t) is the motion at x = 0, y(x,t) is the relative motion of the system, k is the

2 inertia radius of the section, and b = E/p. E is Young's modulus and p is the density.

The frequency equation is given as,

1 + COS m cosh m = 0 8

and the natural circular frequency is written in terms of solution m_. as,

w = kb(ni . j=1.2.3,--. ^

J J

the normal function at m = m. is

:

« 0 J

and the solution of equation 7 is

—11

y(x,t) = V.(X)

where q^ (t) is the generalized coordinates. The equation of nth mode is equivalent to

the motion of a one-degree-of-freedom spring-mass system with viscous damping which is

written as

,

2

8 +2hw (5+-CT a =-B tl 12 n n ^ n rn

where is the participation factor and it is written as.

Therefore, the amplification factor due to the N waves resonance excitation is given as,

'n(^)P^^f(^n^ ( 0 ^

< 1 )

Table 1 shows the participation factor p_. ( j=l , 2 , 3 , . . . ) , and the normal function v^ (j=l,

2,3,...) is given in Fig. 3. Considering these results, the bending moment of the system

for the 1st mode of vibration is written as.

VI-52
where is maximiam at x=l, and M(x,t) is the maximum at x=0, then

Therefore M^^^ is proportional to the relative displacement at the top of the porcelain

insulator, and stress and strain will be given as,

6(x,t) = 11(2, t)/ zCx) . 6(s,t)= 6(s,t)/ E

17

3. Dynamic Test of the Circuit Breaker

Vibration tests on the shaking table were conducted in order to observe the dynamic

response of the structure, and also possible leakage of the air-pressure and defects in

the equipment. The equipment was tested before and after the table excitation. No functional deficiency was found even after loading up to the acceleration of 0.3 g, was applied,

a) Test Structure

The test structure was a circuit breaker of 72/64 kilovolts and 1200 ampere

which was constructed with a frame and a porcelain insulator as shown in Fig. 4. The

porcelain insulator is composed of an upper breaker insulator and a lower supporting insulator. The insulator of each phase has dimensions of 2.2 m in height and 900 kg of weight

and composed of porcelain. The lowest end of the supporting insulator has a larger diameter in order to increase its strength and is cemented and bolted to the frame structure.

Therefore, the maximum stress of the insulator is measured at the second fold from the

lowest end, where the critical strain related to the first excursion failure can be observed.

The result of the static test shows that the critical strain for failure is 780 x 10

^

2 (- e ), and the static allowable stress is 5.85 kg/mm since the Young's modulus of

° 2 porcelain is 7500 kg/mm . Insulators of each phase are called A, B, and C phase insulator

respectively. For the test on the shake table, the supporting frame was bolted to a steel

panel and the panel was fixed to the shake table with bolts.

b) Instrumentation and Measurement

Two horizontal excitation directions were chosen for the test as given in Fig. 4,

and are named short axis test and long axis test. Accelerations and strains of the

porcelain insulator were measured using accelerographs and strain gages as shown in Fig.

4, in which points of G18- G are the accelerographs and points of S XX/- S are the strain

gages

.

c) Vibration

Harmonic excitation of sinusoidal waves in the frequency range between O Hz and

10 Hz , which cover the frequency range of earthquake waves , were induced to find the

frequency characteristics of the system including the support frame and to decide the test

frequency during three waves of vibration. Before the harmonic oscilation , the table was driven by an instantaneous step displacement of 0.5 mm in order to induce the free

VI-53
vibration of each phase insulators and to decide the approximate natural frequency and

damping factor. Then the sinusoidal wave, at the acceleration level of 0.03 g, was applied

to the structure until the stationary response was observed at each frequency. This wave

function was changed between 1 hz and 10 hz by the pitch of 0.2 hz near the resonance. These harmonic excitations were applied in two directions as mentioned previously.

The Meidensha, the manufacturer of the circuit breaker, was primarily interested in

the three waves resonance vibration test. Resonant frequency was defined as the frequency

where the maximum amplification was measured in terms of the acceleration at the top of

the porcelain insulator compared to the table acceleration.

Three successive sinusoidal waves at resonance frequency f^, and also at the frequencies of f^ + 0.2 Hz , were applied to the specimen at the acceleration level of 0.3 g. Two directions of excitation were used and the equipment states for breaker-on stage and

for breaker-off stage were tested.

Normal white noise input waves at the level of 0.05 g for 30 seconds and the

simulated earthquake waves of El Centre 1940 N-S direction were also induced for the

dynamic test.

4. Test Result and Discussion

The supporting frame and the porcelain insulators of the circuit breaker are physically

coupled but they are considered to be structurally independent. However, when considering

the earthquake effects on the system the porcelain insulators are more important than the

frame

a) Natural Frequency and Damping Factor

The damping factor and natural frequency of the insulator can be obtained from

the analysis described in Section 2. From the dynamic test results the natural frequency

of first mode is approximately 9 Hz in the long axis direction and 5.8 Hz in the

short axis direction, and a damping factor of approximately 0.02. The natural frequencies

of higher modes are given as

P. = ( m^n^ f _.lo

The natural frequency of the second mode F^ is 56 Hz , therefore, the vibration of first

mode is important as the frequency range agrees with that of earthquake.

b) Acceleration of the Porcelain Insulator

The maximum acceleration of the insulator, due to the N waves resonance vibration

of first mode, is

Gj,(2,h^,U)= v^(x).p,,. ^T(n3^)'U 19

where U is the amplitude of the input acceleration, A^(h^) is obtained from Fig. 1, and

v^(x) is calculated from equation 10. Fig. 5 shows the maximum acceleration at N=3,

VI-54
h =0.02, x=l, and the test results of G and G . Even though the input frequency of the

J. Z o table is not the same as that of the frame , and the initial condition of excitation is not

exactly the same as that of the analysis, results shown on Fig. 5 can be used as the

criteria for the maximum acceleration at the top of the insulator.

c) First Excursion Failure

We consider the first excursion failure in terms of strain, where the strains were

measured during dynamic tests. Maximum strains are observed at the second porcelain fold

from the bottom and proportional to the relative motion at the top of the insulator.

The linear relationship between the maximum strains and the measured acceleration at

the top was found from the dynamic tests, as shown in Fig. 6 and Fig. 7. The equation for

this relationship is written as;

S (maximum strain) = lOOi 10~^i a (nazimum acceleration) 20

Equation 20 is confirmed for the maximum strain of 570 x 10 ^ from the test and the maxi- mum fracture strain determined from the static test is 780 x 10 ^. Assuming that the

equation 20 is valid for the strain of 780 x 10 ^, the first excursion failure will

be induced at the acceleration of 8.0 g at the top of the insulator.

d) Earthquake Resistance of the Porcelain Insulator

We consider that the earthquake resistance of the porcelain insulator governs

the entire system of the circuit breaker. The equation of first excursion failure determined from the N waves resonance vibration of the first mode is written as

Gj^(l.h^,U)=v^(l>p^-A^j(h^).U = __2i

where G is assumed to be 8.0 g from the test results. Since then the safety factor

o at the three waves resonance vibration for the input acceleration of U will be written

as

,

P5 (D) = /( v^(l). . (h^). U ) -22

Fig. 8 and Fig. 9 show the relationships between the safety factor P^CU) and the input

acceleration of U. Fig. 10 shows the relationship between the safety factor Pj^(U) and

the number of waves of resonant frequency of the first mode at the input acceleration

U = 0.3 g, assuming h^ = 0.02.

e) Seismic Effects on the Circuit Breaker

The main purpose of the dynamic test was to evaluate the effects of the earthquake

vibration on the circuit breaker, including the porcelain insulator and supporting frame.

The safety factor of the breaker obtained from the test is given as the following, where

VI-55
the maximum strain of the support insulator at second fold is £^(0),

( U ) = e / (U) 23

The test data of ^'-'^ assuming U = are plotted on the Fig. 8 and Fig. 9. It was determined that the safety factor (P^CU) can be used as a more conservative criteria

than that obtained from test data (U) of the circuit breaker, we can then evaluate

3 the seismic resistance of the circuit breaker on the bases of this criteria given from

three waves resonance vibration test,

f) Earthquake Wave Test

Earthquake simulating waves of El Centre 1940 N-S direction were applied to the test structure. The maximum strains are approximately 40% of the three waves resonance

vibration test in long axis direction, and 70% for the short axis direction. The three waves resonance test seemed to give conservative criteria for practical design compared

to the earthquake simulating wave test.

5. Conclusions

Conclusions of the dynamic test of the circuit breaker are summarized as follows;

1. The allowable acceleration at the top of the porcelain insulater, related to

the first excursion failure, is 8.0 g.

2. The maximum allowable input acceleration to the circuit breaker is greater than

0.5 g, where the safety factor equals unity.

3. The number of successive sinusoidal waves at resonance is more than 9, based on

an assumption that the safety factor equals unity and the input acceleration is 0.3 g.

4. Acceleration of the strain response, due to a simulated earthquake, is less than

the 70% of the maximiom acceleration or strain observed for the three waves resonance

vibration test at an input acceleration of 0.3 g.

Acknowledgments

Considerable assistance was received from other members of the Earthquake Engineering

Laboratory, National Research Center for Disaster Prevention. In particular, Mr. Nobuyuki

Ogawa and Mr. Chikahiro Minowa for arranging the input waves, and Mr. Haruo lida and Mr. Takeshi Kubota in operating the large-scale shake table during the dynamic test. Additional assistance was received from Dr. Takeo Kinoshita, Head of the Large-scale

Testing Division, National Research Center for Disaster Prevention. This project was

a research program sponsored by Meidensha.

VI-56
3C

H=Number of Sin. waves resonance

Damping Factor -/a Fig. 1

Table

Participation Factor p

j m

.

^.1 1 1.8751 -0.87

2 4.6941 -0.43

3 7.8548 -0.25

4 10.9955 -0.18

5 14.1372 -0.14

Pig. 3 Normal Function V.(X)

3

VI-57
Long axis direction Short azis direction

11

1

B ill ^

i V

/25

o

+>

0) /rFig. 4 Circuit Breaker and Instrumentation

i-i uu

A—-Gra- C?i-Tg

Acceleration of the Bottom

Fig. 5

VI-58
VI-59
8.6 Herz A — Go

.1 .2 »3 Input Acceleration

Fig. S

.5 .6

U

Long Azis Direction

X 6,8 Herz

>< oA-

.7 (s)

phase A

phase B

phase C

.1 ,2

Input Acceleration

Fig. 9

.4

Input Acceleration 0.3 g

.3

Gi , U

Short Axis Direction

.5 .6 (g)

Fig. 10 Factor of Safety and Numher of

Waves of Resonance

2 3-^ 5 6 7 8 9 10

N : Kimber of Waves

vi-60
COMPREHENSIVE SEISMIC DESIGN PROVISIONS FOR BUILDINGS

A STATUS REPORT

BY

CHARLES G. CULVER

National Bureau of Standards, Washington, D.C.

A review of the first draft of recently developed U.S. seismic

design provisions for building is presented. The draft includes

regulatory provisions suitable for inclusion as part of a building code

and technical criteria for earthquake resistant design. The provisions

are intended for implementation by standards organizations, model code

groups. Federal agencies and other regulatory groups. Technical

criteria for structural design, architectural and mechanical-electrical

design, and existing buildings are discussed.

KEY WORDS: Buildings; building codes; design; earthquakes;

structural engineering.

VI-61
INTRODUCTION

In 1974 the National Science Foundation and the National Bureau

of Standards, as part of the Cooperative Federal Program on Building

Practices for Disaster Mitigation, undertook a project with the

Applied Technology Council of the Structural Engineers Association

of California to develop comprehensive nationally applicable seismic

design provisions for buildings. It is intended that these design

provisions be national in scope and reflect the current state-of-theart in earthquake engineering in a form adaptable for implementation

by standards organizations, model codes. Federal agencies and other

regulatory groups. Approximately 90 professionals from throughout

the United States representing the design professions, the university

research community, and regulatory groups including the model codes

and state and local building officials are participating in developing

these provisions.

The development and organization of this activity were discussed at

the Sixth Joint UJNR Panel meeting in May 1974 (1) . In January of

1976 a first draft of the provisions (2) was circulated to professional organizations, trade associations, standards groups. Federal

agencies and professionals throughout the United States soliciting

their comments. The purpose of this paper is to briefly review this

first draft. Since many of the provisions are still under study and

specific requirements may change as the work proceeds, the review will

only deal with overall concepts.

VI-62
SCOPE AND ORGANIZATION

The provisions are comprehensive in nature and deal with earthquake resistant design of the structural system, architectural and

non-structural elements and mechanical-electrical systems in buildings.

Both new and existing buildings are included. The design provisions

are presented in two parts: (1) Regulatory Provisions, and

(2) Technical Criteria. The Regulatory Provisions are presented

in the form of a short, legal document written in mandatory nontechnical language suitable for inclusion as part of a building

code. The technical criteria or recommendations for the design and

construction of earthquake resistant buildings represent acceptable

standards of practice and are intended for inclusion in design standards

and materials specifications.

STRUCTURAL DESIGN

Three alternate procedures are presented for determining the

seismic loads which the structural system must be designed to resist.

Moments, shears and axial forces produced by these loads are to be

determined using an elastic analysis. Selection of the procedure for

determining the seismic loads is based on the type of facility or

building use and the expected ground acceleration. For the majority

of buildings in the United States located in areas where the ground

acceleration in a 50 year period would not be expected to exceed

0.05g, the first or least rigorous procedure would be used. Except

for buildings with large differences in stiffness or lateral

VI-63
resistance between adjacent stories or unusual structural features,

seismic forces for the remaining buildings would be determined using

a procedure referred to as the equivalent lateral force procedure.

For the remaining buildings, a modal analysis is required.

The equivalent lateral force procedure is similar to that

currently used for design in the United States (3) . The total lateral

seismic design force or base shear is related to: (1) geographic

location of the building, (2) soil conditions, (3) the type of structural

system, (4) the period of the building, and (5) type of facility or building

use. The design ground motion is described by a smoothed response

spectra normalized to an "effective peak acceleration" or EPA. Values

of EPA corresponding to a 75% to 95% probability of non-exceedence

in a 50 year period are presented in the form of a contour map for

the contiguous United States, Alaska and Hawaii. Using the contour

map, an acceleration for use in design may be determined for a given

geographic location.

The design acceleration is modified by a soil factor. Discrete

values for the soil factor are defined in terms of the soil characteristics. Detailed site investigations are not required to establish

these characteristics.

The design acceleration is further modified by a building response

factor. This factor is defined in terms of the characteristics of the

vertical and lateral load resisting systems and the horizontal system

used to distribute the seismic forces to the lateral load resisting

system. The factor is analogous to the current K factor (3) and is

VI-64
intended to account for the type of performance to be expected from

various systems.

The variation of design force as a function of building stiffness is accounted for through use of the building period. Approximat

formulas, different from those currently specified in design codes,

are given for determining this period. The formulas are expressed

in terms of the dimensions of the building and the material used for

the lateral load resisting system.

The use of the building also affects the design acceleration.

Higher values are used for essential facilities or buildings which

are required to be functional in the immediate post-earthquake period

Numerous additional requirements related to design of the

structural system are included in the provisions. Soil structure

interaction, the distribution of seismic forces through the height

of the building, torsion, overturning, vertical motions, interstory

drifts and second order effects due to large displacements are

included

.

ARCHITECTURAL AND MECHANICAL-ELECTRICAL DESIGN

The provisions dealing with the design of architectural or nonstructural elements and mechanical and electrical equipment and components represent the greatest departure from current practice in

terms of their scope. Under the new provisions, many building comVI-65
ponents and systems not currently considered in the seismic design

will need to be designed for specified seismic forces. These requirements reflect the increased concern for the need to design these items

in view of observations following recent earthquakes.

Formulas for the design forces are given in much the same form as

current requirements for parts or portions of buildings. The force

levels for architectural components and their connections are specified

in terms of performance levels which are in turn related to the type

of facility and the potential hazard associated with failure of the

component. Deformation limits for these components and considerations

related to anticipated interstory drifts are given.

EXISTING BUILDINGS

The provisions for existing buildings consider evaluation of

the potential seismic hazards associated with these buildings and

emergency safety evaluations for post-earthquake use. For existing

buildings, procedures are given for establishing priorities to select

the buildings for evaluation. A two step evaluation procedure employing an initial examination of the design and construction documentation

and an inspection of the building followed by an analytical evaluation

is employed. In the analysis, the current lateral load resisting

capability and the strength of non-structural components are compared

with those required for new construction to provide a basis for determining the need for hazard abatement. The analysis is also used to

VI-66
determine the time period within which any hazard abatement must

be accomplished.

CONCLUSIONS

The first draft of the provisions includes a number of departures

from current practice. Attempts have been made to reflect the current

state-of-the-art. In some cases, such as architectural and nonstructural elements and mechanical-electrical equipment design, they

represent a significant increase in scope over current design requirements .

It is anticipated that the review of this first draft will identify

areas for improvement. The review should insure that the final recommended design provisions to be issued in about a year are practically

viable and represent a significant step toward improved earthquake

resistant design.

REFERENCES

1. Culver, Charles G., "Comprehensive Seismic Design Provisions

for Buildings," Wind and Seismic Effects , NBS Special Publication 444,

(Proceedings of the Sixth Joint UJNR Panel Conference) , April 1976,

2. Working Draft of Recommended Comprehensive Seismic Design Provisions for Buildings, Applied Technology Council, San Francisco,

California, January 31, 1976.

VI-67
3. Uniform Building Code Standards , 1973 Edition, International

Conference of Building Officials, 5360 South Workman Mill Road,

Whittier, California.

VI-68
RETROFITTING OF VULNERABILITY

IN EARTHQUAKE DISASTER MITIGATION PROBLEMS

KAORU ICHIHARA

Director, Public Works Research Institute

EIICHI KURIBAYASHI

Ministry of Construction, Japan

Earthquake Engineering Section

TADAYUKI TAZAKI

Ministry of Construction, Japan

Earthquake Engineering Section

ABSTRACT

This paper discusses a criterion for retrofitting of existing structures vulnerable

to disastrous earthquakes.

It would be ideal if all structural damage could be avoided during earthquakes, however, to completely strengthen structures is not practical because of the limitation of resourses and land-space. On the basis of execution, it is necessary to classify the

structures by functional importance and structural vulnerability. The first category of

classification deals with the structure load. In the category dealing with the location,

the densely inhabited and effectively utilized structure, should have preferential earthquake resistance, and this consideration is the basic philosophy of urban design including

the countermeasure to earthquakes. The second category is classified according to the

purpose of the structure. Life lines, such as traffic, transportation, water and energy

transmission, and comm\inication are indispensable during evacuation and rescue, so that

these structures would be required to retain safety. This is especially true for roads,

which are used not only for a path of evacuation but also for rescue space, fire fighting

etc. immediately after earthquakes. However, it seems impossible that all existing

roads can be modified to such a high standard of earthquake resistance because of the

extreme cost involved. It is therefore more practical and rational to select the important routes which should be modified to meet specific earthquakes.

In this situation the following principles have been chosen in the retrofitting and

the vulnerability for such a decision-making process.

If given certain model routes, retrofitting costs can be obtained for several

methods, for example, perfect, medium and rough, for which the retrofitting cost should

not be 'more than the cost of reconstruction. Next, suitable load-retrofitting relations

ana suitable load-safety relations are examined^ Using these quantitative results

a final decision can be made.

Key Words: Costs; Existing Structures; Optimization; Retrofitting; Seismic.

VI-69
1. Introduction

Due to progressive earthquake resistant design criteria, it is possible to design and

construct various structures resistant to earthquake effects. However, with some exceptions, no one knows multilateral earthquake resistivities. Also immediately after earthquakes, continuity between rescue, transportation, communication, energy supply and circulation of water should be retained.

Concerning the above, this paper discusses the basic principle planning methodology

for appropriate selection of structures for retrofitting corresponding to those of

some vulnerability and for considerable assessment of the difference between reasonable

capacity required and actual capacity of the structure in an earthquake.

Generally greater retrofitting is required for more earthquake resistivity, but perfect earthquake resistivity requires an infinite amount of investment to retrofit. It is

not practical to make such an investment, because of the limitation of resources and land

space. Therefore, an optimum amount of investment should be evaluated relative to the

physical and socio-economic conditions. This of course requires consideration of the loss

of hvunan life and the decrease of the functions of the various structures.

2. Methodology of Selection

The criteria in selecting the structures to be retrofitted, consist of selection of

various physical and socio-economic conditions. The items to be examined in selection of

the structures are shown in Table 1. An example of the relationship between a structure,

durability, and its physical conditions are given in Fig. 1. Initially a method for retrofitting is determined. This is based on the retrofitting

costs by several methods, which are not more than the cost of reconstruction. The capable

load-retrofitting cost and the capable load-safety are examined. Using these quantitative

results a final decision can then be made.

The next chapter described one method to evaluate the optimum amount of investment,

considering only the physical conditions.

3. An Optimization of Initial and Running Costs of Structures Taking into Account

the Probability of Failures Caused by Earthquakes

a) Assumptions

To optimize initial and running costs of structures, taking into account the

probability of failure caused by earthquakes, the following is assumed relative to the

physical conditions with respect to the probability of failure and expenditures;

1. Future earthquake intensities and recurrences can be estimated from trends of

past earthquakes.

2. Probability of failure due to earthquakes can be obtained from existing information .

3. Earthquake resistances of individual structures can be obtained from codes and

regulations

VI-70
4. Total costs consist of the cost of construction, retrofitting and loss of

failure

.

b) Principle in Determining the Safety of Structures in Earthquakes

Let R and S be the resistivity and the intensity of load. The probability P that structures will become limit states is given by Eq. (1)

;

.Pr [ R<S] = jy^M f3(x )dx

where Pr [R<S]

F^(x)

f^(x)

probability of R<S

cumulative distribution function of R random variable of S

(1)

In the case of earthquake, let f , , be a random variable of maximum earthquake

s (x) intensity in a year and , , be a cumulative distribution function of the mean quakeR(x)

resistivity in a year. The cumulative probability of failure of a structure during its

life is therefore;

^ D

= 1 - TT (1 - p) D t = 1

(2)

U^: cumulative probability of survival

T^: durable life

If p is a constant during the life, eq. (2) can be written as;

= 1 - (1 - p)

The assumption (b) given in a) means that;

(3)

VI-71
where U* represents social needs for safety during an earthquake.

Therefore the structures must be designed so satisfy eq. (3). However, existing

structures, with low quake-resistivity and impossible structures to retrofit and thus

satisfy eq. (3) . In these cases (T^ - T) , where T equals the served lives, can be substituted instead of (T^^) into eq. (2) , which considers the fact that the existing structure will fail. This thought can be written as;

'^D-T (4)

u = 1 . TT (1 - p)

(5)

u = u*

In Fig. 3 if the resistivity R of a structure A at time T corresponds to the constant

level of cumulative probability, then structure A is safe. Structure B has also adequate

safety because of a decrease of cumulative probability. Structure C needs to be retrofitted because the resistivity decreases more than A and the cumulative probability exceeds U*. To adopt this criteria, the initial resistivity is not required for retrofitting, but R^ which brings cumulative probability U* during its residual life (T^^ - T)

is sufficient.

Now, let the total cost of a structure during its durable life be defined as follows;

where C

CC^R

C„

total cost of a structure

cost of construction

cost of retrofitting

loss of failure

Supposing that depends only on the cumulative probability, becomes

~- ^co - ^ Uq)

C: constant

U^: cumulative probability assuming that a structure is not retrofitted during its

useful life (a structure without retrofitting = U^^)

Suppose that the cost to retrofit a structure from quake-resistivity R to (R + A) is

proportional to the difference between the cost of construction of R and (R + Ar) , then

becomes

VI-72
C =oc Z {c^(R + aR) -C^(R)}

a: constant

N: number of times of retrofitting

C^(R + Ar) , C^(R) : cost of construction of (R + Ar) and R

The cost of failure (C ) will be divided into an independent part of the cumulative

F probability, such as the indirect losses of failure, and a dependent part on cumulative

probability such as the cost of repair and reconstruction. Therefore, C becomes

F

C„„,m: constant

FO'

Under the conditions of equations (7) through (9) , the total cost C should not be

minimized.

c) Examples

Initially the social needs for safety during earthquake need to be given. Table

2 shows the number of destroyed bridges, after the Kanto earthquake (1923) . In (1964) the

Niigata earthquake it was reported that there were three bridges destroyed and 13 others

(9) damaged bridges, which had lengths longer than 20 m . Accordingly the total number of

bridges destroyed or functionally damaged during the Niigata earthquake can be estimated

as

(3 + 13) X AQ^'^ (Total Number of Bridc^es in Nii^'at a Prcf. in 1964)

60iJ (Number of bridges longer tlian 20 m)

= 107

The total number of the bridges destroyed or functionally damaged the after Kanto

earthquake is estimated as; 22x^.785

The increase in the number of bridges on the principal roads (national highways

,

prefectural roads and principal municipal roads) in Japan after 1948 is shown in Table

3. The increase after the Kanto earthquake is therefore estimated as given in Fig. 4. The average number of bridges after 1923 was 90040. Using the value 0.24, that is the

ratio of the number of bridges on principal roads to that of total roads, the total number

of bridges from 1923 becomes

VI-73
90 040 T 0.24 = 375 000

The cumulative probability converted to a useful life of 5C years, becomes

"^85 X ^

=2.1X10-^

375 000 51

-3

In this example U* is assumed to be 2.1 x 10 . As shown by eq. (1) , the probability

p is dependent on both the quake-resistivity R and the intensity of the load S, however,

it is difficult to evaluate R; therefore it is assumed that R has no deviation but has a certain value and p is dependent only on S. Generally, the relationship between the

earthquake intensity Ijj^^ and the probability p, which means the probability that an earthquake exceeding I occurs in a year, is well known and is as follows

JMA

loliP - 0 - r ij^^ (10)

0 r : constant

JMA

earthquake intensity of Japan Meteorological Agency

Applying the data given in the report of Katayama (12), 6 = 3.24 and y = 0.87 for

Tokyo is determined. (Fig. 5) The relationship between I and response observed

JMA B for road bridges is reported as (13)

B

2 log

I

0. 528 y^-j

0. 2 53

2 log < 0. 264 y

000

0. ROO

1000

X 10°-^

X 10^'^^JMA^^

1000 1

0. 2 53 /

1000 1 0. 800 /

(11)

Where the first equation corresponds to the earthquake of higher mangitude and

the second equation of smaller magnitude. Using the average of both, eq. (11) becomes;

~- "^-^ \jMA 2. 49 (11)'

Inserting p, which corresponds to the cumulative probability U^^ of 2.1 x 10 given

in eq. (2) into eq. (10) and inserting I to eq. (11), I equals 6.87. Thus if a JMA B structure has an initial quake-resistivity of R^ = 6.87, and the quake-resistivity is not

changed during its life, the cumulative probability of 2.1 x 10 is ensured. In general

practice, however, bridge quake resistivities are decreasing because of the decrease of the

stength of materials, corrosion and fatigue. (Fig. 1) In this example the quake-resistivity

at the end of a useful life is assumed as 75% of the initial value. The quake-resistivity

VI-74
R(t) , at time t, is assumed as;

R(l) = 1. 012 log (50 - t) + 5. 15 (12)

based on R „ = 6.87 (Fig. 6) t=0

The coefficients in eq. (7) through (9) can now be estimated. The constant c, in eq.

(7) , is independent of the cumulative probability and the value (c = 0.041) introduced

into the reinforced concrete problems (8) is used herein. For example most typical

2 bridges as of 1973, have a length of 30 m, width of 8. m, area of 240 m . The estimated

cost of such a bridge, based on the assumption that the bridge has a cumulative property

-3

of 2.1 X 10 , will be;

6 0000 (avera.qe cost of gMnerplructure per unit aron for 30 m

CO "

1 - 0. 041 lo^f 2. 1 X 10^^

span) X 240 -r 0. 6 (ratio of cost of superstructure to that of

the total bridpc) n r rJ i \ = 2. 16 X 10 (yen)

Inserting C^(R + AR) and C^{R) calculated by eq. (8) into eq. (7),

C^= aC^^c SdogU^-log Up^^^^)- (13)

where and ^j^^^j^ the cumulative probabilities having quake-resistivity of R and R + Ar during their residual lives. This value is the ratio of the cost of retrofitting to that of construction and the former exceeds the latter because of the scale merit and

temporary conditions, therefore a = 2 will be assumed.

It is assximed that C^^ in eq. (9) consists only of indirect losses, and is determined

according to the ratio of indirect losses to direct losses as reported for the Niigata

earthquake

;

C =

1 billion yen (indirect losses) ^ =2 12x10"^ (yen)

FO 130. 0 billion yen (direct losses) CO

The value m is assumed as 1.0

Substituting R for I in eq. (11) and inserting into eq. (10) gives;

B

p-io'-'' (14)

where R is given by eq. (12) . Thus cumulative probability of the structure, for which

the initial quake-resistivity if R + Ar^, during its durable life can be written as

VI-75
= 1 - Tl (1 - 10 • M . D t = 1

Suppose that the initial quake-resistivity is R + Ar^ and retrofitting is Ar^ at T years after construction. The probability of failure becomes;

= 1- tt'^'ci-io^-^^-^-^^^^^^^^i^^'^^)) Tr

.D t = l Td-T + 1

7. 57 - 1. 74 (R + aRi ) , ^1^^ (1-10 M

Table 4 shows the results for Ar , Ar , C , C , C and C. Taking into account the interest

1 2 C R F rate Y; r - r + (C + C )

Fig. 7, 8, 9 shows the total costs C corresponding to interest rates of 0.00, 0.05

and 0.10 respectively. If Ar^ is less than 0.20 and 0.40, corresponding to the retrofitting of 30 years and 40 years after construction, it is impossible to ensure the cumula-

-3

tive probability of 2.1 x 10 , then the results for that area are missing,

d) Considerations

1. The results noted above give the minimum cost under the condition of U^^ = U*. When Y is 0.00, the minimum cost is realized in case of high initial quake-resistivity

and is not influenced by the time of retrofitting, because the cost of retrofitting is more expensive than that of construction (a=2.0). When Y is 0.05 or 0.10, it is profitable to retrofit and 30 through 40 years (Y=0.05) or 20 through 30 years (Y=0.10) after

construction

2. Using the method described in section 3c, the most economical quake-resistivity

for an initial design can be obtained under a certain condition for the value of the

cumulative probability during the durable lives. Concerning existing structures the

level of retrofitting can be obtained for a certain value of cumulative probability. However, these methods should be applied to the structures of quake-resistivity that ensures

U* at the beginning of construction. Therefore it is different from the methodology

shown in Fig. 3, in that it is sufficient to ensure U* only during the residual lives.

The former methodology should be applied to relatively new structures and the latter to

the old structures.

4. Conclusions

a) To retrofit structures, physical and socio-economic conditions should be considered.

However, it is hard to evaluate the socio-economic conditions quantitatively.

b) Relationships between the cost and the safety can be estimated from the following;

method — cost; cost — capable load; and capable load — safety relations.

VI-76
c) with respect of the application, considering only the physical conditions,

optimization for retrofitting can theoretically be obtained.

d) The methodology can be applied to reasonable retrofitting of vulnerable structures

and also extensive urban renewals which are exposed to high levels of earthquake risk

potentials and the socio-economic conditions are important.

References

1. Japan Urban Center: Investigations on the Planning of Urban Disaster Prevention.

2. City Bureau, Ministry of Construction: Draft of Manuals of Planning to Promote

Countermeasures for Mitigation of Earthquake Disasters, Aug. 1975.

3. J. Cooper: Highway Bridges Seismic Research, Proc. of 7th Joint Meeting of

Panel on Wind and Seismic Effect UJNR, Tokyo, May 1975.

4. Committee of Disaster Prevention, Tokyo Pref . : Investigations on Provision of

Refuge Places for Fires by Earthquakes, Sept. 1973

5. Yokohama City: Risk Energy, 1972.

5. Committee of Disaster Prevention, Nagoya City: Investigations on the Evacuation

Planning from Fires by Earthquakes in Nagoya, June, 1974.

7. Executive Office of the President Office of Emergency Preparedness: Report of the

Congress, Disaster Preparedness, Jan. 1972.

8. Pugsley, A.G. : The Safety of Structures, Edward Arnold Publishers Ltd., 1966.

9. Report of the Public Works Research Institute, Ministry of Construction, Vol. 125,

June 1965.

10. Annual Report of Statistics of Road, 1964.

11. Annual Report of Statistics of Road, 1973.

12. T. Katayama: Fundamentals of Probabilistic Evaluation of Seismic Activity and

Seismic Risk, May 1975.

13. E. Kuribayashi, T. Iwasaki: Observed Earthquake Responses of Bridges, Proc. of

4th WCEE, Jan 1969.

14. Japan Road Association: Annual Report of Road Bridges, 1973.

15. Niigata Pref. : Records on Niigata Earthquake, June 1965.

VI-77
TABLE 1 ITEMS TO I'^CAMINE FOR E^VKTHQUAKE

DISASTER MITIGATION

CLASSIFICATIONS

I'U N CT 1 ON A L I MPORTA NC E DLRVBILITY

REPAIR AND

RECONSTCQNTRIBLTION NUISAIMCE DUMMY UCT ION

VICINITY O o O

STRUCTURE O o O O o

SCALE o o O O o

TAELE 2 DESTKOTOD BRIDGES ALTER

KANTO E.\RTHQUAKE

TABLE 3 TR/VXSITION OF MMBER

OF BRIDGES

EARTHQUAKE DATE

NUMBER OF

DESTROYED

BRl DOES

KANTO 1923. 9. 1 8 NANKAI 1946. 12.21 1 FUKUI 1948. 6.28 1 0 N I I GATA 1 964. 6. 16 3 TOTAL 2 2

YEAR SIDRT SPAiN BRIDGES TOTAL

1 948 1. 730 105. 323 107. 053

1 950 1, 901 113, 621 1 15, 522

1951 1. 934 116, 806 118. 740 1952 1, 994 119, 045 121. 039 1953 2, 0 54 112, 190 1 24. 253

1954 2, 094 123. 237 1 25. 331 1955 2. 1 37 123, 027 1 25, 164 1956 2, 234 124, 473 1 26, 707

1 957 2. 257 124. 523 126, 7 80 1958 2, 282 124, 396 1 26. 678

1959 2, 350 123. 953 1 26. 303 1960 2. 397 1 23, 288 1 25. 685

1961 2, 436 123. 373 1 25, 809 1962 2, 549 123, 84 7 1 26. 396 1963 2, 599 1 23, 971 1 26. 570 1964 2. 701 123, 400 1 26, 101 1965 2. 797 124, 090 126, 887

1 966 2, 883 1 24, 008 1 26, 891

1 967 2, 992 125, 587 128, 579

1968 3, 1 16 124, 726 1 27, 842

1969 3. 344 126, 406 129, 750

1 970 3, 520 126, 677 1 30, 197

1971 3, 725 127. 524 131. 249

1 972 3, 925 1 28. 966 1 32, 891

1 973 4. 107 130. 056 1 34. 163

VI-78
0

T I ME

Fig.-i CHARACTERISTICS OF DURABILITY OF EXISTING STRUCTURES

RECONSTRUCTION

A

8

1

CAPACITY SPEC! RED

3 METHOD OF RETROFITTING ° COS"! 0 CAPABLE LOAD

Fig. -2 INVESTMENTS AND RETROFITTING

VI-79
TABLE 4 TOTAL COSTS OF STRUCTURES ( COST : x lO^YEN)

7" = 0. 00 r = 0. 05 r= 0. 10

Ur Cr Cf c Cr/

[l+rf (l+rf c

Cr/

(l+rT"

Cf/ c WITHOUT RETROFITTING

0. 90 0. 00210 2. 4 0 0. 00 0. 01 2. 41 0. 00 0. 01 2. 41 0. 00 0. 01 2. 41 RETROFITTING AFTER 10 YEARS

0. 0 0 0. 9 7 0. 0749 0. 0744 0. 0016 2. 26 0. 29 0. 01 2. 56 0. 18 0. 00 2. 44 0. 1

0. 00 2. 37

0. 1

0. 84 0. 0502 0. 0409 0. 0018 2. 2 8 0. 2 4 0. 01 2. 5 3 0. 15 0. 00 2. 43 0. 09 0. 00 2. 37

0. 20 0. 72 0. 0338 0. 0336 0. 0019 2. 29 0. 22 0. 01 2. 5 2 0. 14 0. 00 2. 43 0. 09 0. 00 2. 38

0. 30 0. 61 0. 0227 0. 0226 0. 0020 2. 31 0. 19 0. 01 2. 51 0. 11 0. 00 2. 4 2 0. 07 0. 00 2. 38

0. 40 0. 51 0. 0153 0. 0152 0. 0020 2. 32 0. 16 0. 01 2. 49 0. 10 0. 00 2. 42 0. 0 6 0. 00 2. 38

0. 50 0. 41 0. 0103 0. 0102 0. 00 20 2. 34 0. 13 0. 01 2. 4 8 0. 08 0. 00 2. 42 0. 05 0. 00 2. 39

0. 60 0. 31 0. 00688 0. 00683 0. 0020 2. 35 0. 09 0. 01 2. 45 0. 06 0. 00 2. 41 0. 04 0. 00 2. 39

0. 70 0. 21 0. 00461 0. 00458 0. 0020 2. 37 0. 06 0. 01 2. 44 0. 04 0. 00 2. 41 0. 02 0. 00 2. 39

0. 80 0. 11 0. 00310 0. 00308 0. 0020 2. 3 8 0. 0 3 0. 01 2. 4 2 0. 02 0. 00 2. 40 0. 01 0. 00 2. 39

0. 90 0. 01 0. 00207 0. 00205 0. 0020 2. 40 0. 00 0. 01 2. 41 0. 00 0. 00 2. 40 0. 00 0. 00 2. 40 RETROFITTING AFTER 2 0 YEARS

0. 00 1. 14 0. 0749 0. 0737 0. 0008 2. 26 0. 35 0. 01 2. 62 0. 13 0. 00 2. 39 0. 05 0. 00 2. 31

0. 10 0. 91 0. 0502 0. 0494 0. 0013 2. 28 0. 28 0. 01 2. 57 0. 11 0. 00 2. 39 0. 04 0. 00 2. 32

0. 20 0. 78 0. 0338 0. 0332 0. 0015 2. 29 0. 24 0. 01 2. 54 0. 09 0. 00 2. 38 0. 04 0. 00 2. 33

0. 30 0. 6 3 0. 0227 0. 0224 0. 0018 2. 31 0. 19 0, 01 2. 51 0. 07 0. 00 2. 38 0. 03 0. 00 2. 34

0. 40 0. 5 3 0. 0153 0. 0150 0. 0018 2. 32 0. 16 0. 01 2. 49 0. 06 0. 00 2. 38 0. 02 0. 00 2. 34

0. 50 0. 4 3 0. 0103 0. 0101 0. 0018 2. 34 0. 13 0. 01 2. 48 0. 05 0. 00 2. 39 0. 02 0. 00 2. 36

0. 60 0. 33 0. 00688 0. 00676 0. 0018 2. 35 0. 10 0. 01 2. 46 0. 04 0. 00 2. 39 0. 02 0. 00 2. 37

0. 70 0, 23 0. 00461 0. 00454 0. 0018 2. 37 0. 07 0. 01 2. 4 5 0. 03 0. 00 2. 40 0. 01 0. 00 2. 38

0. 80 0. 13 0. 00310 0. 00305 0. 0018 2. 38 0. 04 0. 01 2. 4 3 0. 02 0. 00 2. 40 0. 01 0. 00 2. 39

0. 90 0. 03 0. 00207 0. 00203 0. 0018 2. 40 0. 01 0. 01 2. 42 0. 00 0. 00 2. 40 0. 00 0. 00 2. 40 RETROFITTING AFTER 3 0 YEARS

0. 20 0. 90 0. 0338 0. 0326 0. 0009 2. 29 0. 28 0. 01 2. 58 0. 06 0. 00 2. 35 0. 02 0. 00 2. 31

0. 30 0. 69 0. 0227 0. 0220 0. 0014 2. 31 0. 21 0. 01 2. 53 0. 05 0. 00 2. 36 0. 01 0. 00 2. 32

0. 40 0. 56 0. 0153 0. 0148 0. 0016 2. 32 0. 17 0. 01 2. 50 0. 04 0. 00 2. 36 0. 01 0. 00 2. 33

0. 50 0. 44 0. 0103 0. 0099 0. 0017 2. 34 0. 14 0. 01 2. 49 0. 03 0. 00 2. 37 0. 01 0. 00 2. 35

0. 60 0. 31 0. 00683 0. 00664 0. 0019 2. 35 0. 10 0. 01 2. 46 0. 02 0. 00 2. 37 0. 01 0. 00 2. 36

0. 70 0. 21 0. 00461 0. 00445 0. 0019 2. 37 0. 07 0. 01 2. 4 5 0. 02 0. 00 2. 39 0. 00 0. 00 2. 37

0. 80 0. 11 0. 00310 0. 00299 0. 0019 2. 38 0. 03 0. 01 2. 42 0. 01 0. 00 2. 39 0. 00 0. 00 2. 38

0. 90 0. 01 0. 00207 0. 00199 0. 0019 2. 40 0. 00 0. 01 2. 41 0. 00 0. 00 2. 40 0. 00 0. 00 2. 40 RETROFITTING AFTER 4 0 YEARS

0. 40 0. 72 0. 0153 0. 0140 0. 0008 2. 32 0. 22 0. 01 2. 55 0. 03 0. 00 2. 35 0. 00 0. 00 2. 32

0. 5 0 0. 52 0. 0103 0. 0094 0. 0012 2. 34 0. 16 0. 01 2. 51 0. 02 0. 00 2. 36 0. 00 0. 00 2. 34

0. 60 0. 36 0. 00688 0. 00632 0. 0015 2. 35 0. 11 0. 01 2. 47 0. 0 2 0. 00 2. 37 0. 00 0. 00 2. 35

0. 70 0. 23 0. 00461 0. 00424 0. 0017 2. 37 0. 0 7 0. 01 2. 4 5 0. 01 0. 00 2. 38 0. 00 0. 00 2. 37

0. 80 0. 12 0, 00310 0. 00284 0. 0018 2. 38 0. 04 0. 01 2. 43 0. 00 0. 00 2. 38 0. 00 0. 00 2. 38

0. 90 0. 00207 2. 40 0. 00 0. 01 2. 41 0. 00 0. 00 2. 40 0. 0 0 0. 00 2. 40

VI-80
TD

Fig. -3 EETROFITTING OF EXISTING STRUCTURES

VI-81
op I -

Fig.-5 PROBABILITY OF OCCURRENCE OF EARTHOUAKE TOKYO

VI-82
VI-83
Fig. -9 RELATIONSHIP BETWEEN TOTAL COST AND INITIAL EARTHQUAKE RESISTIVITY, RETROFITTING TIME r = 0.i 0 )

VI-84
DYNAMIC RESPONSE CHARACTERISTICS

OF A MODEL ARCH DAM

CHARLES D. NORMAIC, ROGER D. CROWSON, JIMMY P. BALSARA

Weapons Effects Laboratory

U.S. Army Engineer Waterways Experiment Station

P.O. Box 631, Vicksburg, Miss. 39180

ABSTRACT

The dynamic response characteristics of a model arch dam are given in detail. These

characteristics were determined by siobjecting the model to a vibratory loading. Resulting

defojnnations and velocities are given.

Key Words: Arch Dam; Deflectionsj Dynamic Tests; Model; Structural Analysis; Velocities;

Vibrations

.

1. Introduction

a) In order to develop better design procedures for concrete dams subjected to earthquake forces, an understanding of the significant parameters that influence the dynamic

properties of such structures is necessary. Assumptions regarding geometry, boundary

conditions, and interaction with the foundation and reservoir can significantly affect

earthquake response calculations. Vibration tests provide a means of determining dynamic

properties of dams and evaluating the various parameters that influence these properties.

The investigation of the North Fork Dam was undertaken to verify the use of physical

models and linear three-dimensional (3-D) finite element analyses to determine the dynamic

properties of an arch dam.

b) The North Fork Dam, located near Auburn, California, on the North Fork of the

American River, was built in 1939 to trap sediments from upstream mining operations. It

is a constant-angle arch with a maximum height of 155 ft*, a crest length of 620 ft., and

a crown thickness that varies from 22 ft at the base to 6.8 ft at the crest. The overflow

spillway is a section 200 ft long near the center of the dam and is depressed 3 ft below

the normal crest elevation. The reservoir is very narrow, having a maximum width of

about 700 ft and a length of about 5 miles.

c) Completion of the Auburn Dam 5 miles downstream will inundate the North Fork Dam

completely. Prior to completion, full-scale tests similar to the model tests discussed

in this report will be conducted to correlate model and full-scale test results.

2. Model Construction

a) Details of the first North Fork Dam model, which was destroyed by explosive tests,

are discussed in Reference 1. In the second North Fork Dam model, the foundation, re- action mass, and reservoir of the first model, and a new dam and kicker slab were used.

The kicker slab is a mass of concrete used to transmit the input vibratory load to the

foundation. After complete removal of the original dam and slab, two-way steel reinforceVI-85
ment was welded into position for the new slab. The reinforcement consisted of no. 5 steel bars placed between the foundation and loading plate. Commercially available

concrete having a compressive strength of 4000 psi was used for the new slab. A plan

view of the dam and slab showing reinforcement details is shown in Figure 1.

b) The foundation was slightly altered by cutting a 2-in. by 2-in. groove at the damfoundation interface, thus forming a secure keyway after the dam was cast. The lightweight concrete forms used in the first model were again utilized. A specially prepared

concrete grout with an average 28-day compressive strength of 5000 psi and an average

density of 142.2 pcf was used. The concrete was allowed to cure for 28 days, after which

the forms were removed and a curing compound applied. The new model is shown in Figure 2.

3. Test Procedures

a) Test Equipment: The electromagnetic (EM) and electrohydraulic (EH) vibrators

used in the first test series (Reference 1) were again employed in this test. Instrumentation used for motion measurement consisted of velocity transducers connected by compatible electronics to two analog magnetic tape recorders. During the EM and EH vibrator

tests, nine velocity transducers located along the dam crest and three on the crown contilever section were used to measure both radial and vertical motion. Hydrodynamic pressure

measurements during both EM and EH tests were made with a pressure transducer mounted on

a rod and hand-held at a constant distance of 1 in. from the dam face. Complete descriptions and specifications of the equipment are provided in Reference 1.

b) EM Tests: EM vibrator tests were conducted with the reservoir full (1 in. below

top of spillway crest), 24 in. down, and empty. The two vibrators were placed on the crest

at each end of the spillway and run in phase and 180 deg out of phase. A constant force

level of 15.5 pounds per vibrator, applied in a radial direction, was maintained for all

EM tests. The frequency was swept from 30 to 500 Hz at a rate of 1.0 Hz/sec. Other

sweep rates ranging from 0.25 to 10 Hz/sec resulted in the same response curves. The vibra- tors were held at resonant frequencies for several minutes to allow amplitude buildup.

However, no increase in amplitude was noted for any particular sweep test. Damping tests were conducted by running the EM vibrators at various resonant peaks and abruptly shutting

off the power amplifier The logarithmic signal decay was measured to determine the damping

ratio. Hydrodynamic pressure measurements were made at resonant frequencies by holding a pressure transducer 1 in. from the upstream dam face at various locations. A list of all

EM tests conducted is presented in Table 1, and plan, profile, and section views of the

model and equipment are shown in Figure 3.

c) EH Tests: Base excitation of the dam was provided by the EH vibrator. Vibratory

loads were transferred to the dam foundation through a heavily reinforced concrete slab

(Figure 1) . Frequency sweeps and hydrodynamic pressure measurements were made at dynamic

force input levels of 10,000-40,000 pounds. The frequency was swept from 10 to about

175 Hz at a rate of 0.5 Hz/sec. Hydrodynamic pressure measurements were made at a load

of 20,000 pounds while the vibrator was held at various resonant frequencies.

VI-86
4. Test Results

a) The experimental portion of this study was primarily concerned with determining

natural frequencies, mode shapes, structural damping, and hydrodynamic pressures. Natural

frequencies were obtained by analyzing velocity versus frequency plots, as shown in

Plates 1-3 for the EM tests. Plate 4 shows velocity versus frequency plots for the EH test in which hydrodynamic pressures were measured. Mode shapes determined from EM tests

are presented graphically in Plates 5-7. Damping ratios are given in Table 2, and hydrodynamic pressures are plotted in Plates 8-16. A comparison of natural frequencies obtained

for the first model test (Reference 1) with those reported herein can be made using the

data presented in Table 6. In comparing these data, consideration should be given to the

fact that the first model had been damaged by explosive tests then repaired for the vibration tests. For the study reported herein, the EH tests were aimed at better designing

hydrodynamic pressures and damping ratios. No significant peaks in the velocity data

were noticed during the EH frequency sweeps and therefore natural frequencies could not

be determined.

b) EM Tests : Data from the EM frequency sweep tests were recorded on magnetic tape

and later played back through an X-Y plotter. The records for these tests are shown

in Plates 1-3. During the tests, output from various velocity transducers was recorded on

an X-Y plotter, and acceleration input to the vibrator mass was monitored on an oscilloscope. Oscillograph records were also made at resonant peaks to observe wave shapes and

to provide a crosscheck of the analog- to-digital data used to determine the mode shapes.

c) EH Tests: The EH vibration test data were recorded in the same manner as were

the EM data. The constant load feedback measured by a pressure transducer in the EH vibrator was monitored on an oscilloscope during all tests. After the tests were completed,

all data channels were played back on the X-Y plotter.

d) Mode Shapes: The velocity versus frequency plots were carefully analyzed to determine resonant peaks. The 12 velocity data channels were converted at the resonant

frequencies to digital records, which were then integrated to obtain displacements.

Displacements for the symmetrical and asymmetrical mode shapes for both the radial crest

and cantilever deflections are shown in Plates 5-7. With the reservoir full, the natural

frequencies for the second through the fifth modes were 103, 147, 175, and 243 Hz,

respectively. The first mode excited in the first model (Reference 1) could not be

determined in this test series. With the reservoir empty, the frequencies were 133,

151, 213, and 297 Hz. Values for the second, fourth, and fifth natural frequencies increased 22-29 percent for the empty condition, whereas the third natural frequency increased only 3 percent.

e) Damping: For the DM tests, damping ratios were determined with the reservoir

full and empty. Output from velocity transducers 3 and 6 (location of the vibrators)

was recorded on an oscillograph record as the vibrators were abruptly shut off after being

held at a resonant peak. The relative amplitude versus cycle of oscillation was measured,

and the damping ratio D was determined from the formula;

VI-87
where n = number of cycles of oscillation

= amplitude of 1st cycle

X = amplitude of n^h cycle

n

A tabulation of all damping data is presented in Table 2. The damping ratios are quite uniform, as values ranged from 0.02 to 0.05 for all tests conducted. From the

data collected, the reservoir level, frequency, and phasing of vibrators had no discernable effect on the damping ratios.

f) Hydrodynamic Pressure

1. For the EM tests, hydrodynamic pressure measurements were made at each velocity

transducer location at 1-ft depth intervals. Readings were taken at each point while the

vibrators were held at the second, third, and fourth natural frequencies (105, 147, and

196 Hz, respectively). The EM data are shown in Plates 8-10 as pressure versus depth.

The maximum pressure recorded was 0.275 psi occurring at 175 Hz, 12 in. below velocity

transducer 4.

2. Hydrodynamic pressure measurements for the EH tests were made in the same manner and at the same locations as were those for the EM tests. Readings were taken

with the vibrator held at 80, 98, and 122 Hz. These frequencies were selected to provide

pressure data at frequencies below the dam's fundamental mode and then considerably above

the fundamental mode. The force level was 20,000 pounds. The data plots of pressure

versus depth are shown in Plates 12-14. The maximum pressure recorded was 0.40 psi at

122 Hz, occurring at a depth of ift. below velocity transducer 5.

5. Finite Element Analysis

a) General

1. A 3-D finite element analysis was conducted to determine the vibration characteristics of the model dam. Specifically, mode shapes and corresponding natural frequencies were determined for two finite element models, one in which the foundation was included and one with the base fixed. For each grid, computations were made with the

reservoir full and empty.

2. An eight-node 3-D solid element was used in the calculations. The element has

three translational degrees of freedom (DOF) per node and contains incompatible deformation modes that greatly improve its bending behavior. A detailed description of this

element is presented by Wilson in Reference 2. The small grid consists of the dam alone

and contains 110 elements, 278 nodes, and 570 DOF. The large grid comprises the dam and

the massive concrete foundation of the model. This grid has 916 elements, 1464 nodes, and

2916 DOF. The nodes at the base of the small grid are fixed against translation in all

directions. In order that the effects of deformations in the dam foundation could be

VI-88
studied, the large grid was used with outside foxmdation element nodes fixed against translation. The two grids are shown in Figures 4 and 5.

b) Reservoir Effects

1. The problems of hydrodynamic effects on submerged or partially submerged

structures has received considerable attention in the recent past. One method of accounting for this phenomenon is to construct a large finite element grid modeling the structure

and the water. This method presents problems for many systems since the number of unknowns

can be great, and the lack of symmetry in the mass and the stiffness matrices increases the

amount necessary computational effort.

2. Another approach to the fluid-structure interaction problem is the substructure

concept (Reference 3). In this approach, the structure is discretized by finite elements,

and the reservoir is treated as a continuum boundary value problem. Results of the contin- uum reservoir solution are used as input to the equations of motion, which are expected in

modal coordinates. This method depends heavily on the geometry of the reservoir being

simple, since solutions to the continuum boundary value problem are difficult for complex

boundary geometries. Still another method for handling the hydrodynamic problem is to calculate an effective mass of water and add this mass to the structure. This concept

was proposed by Westergaard (Reference 4) and is based on the idea that the hydrodynamic

force is of an inertia type and can therefore be represented as an effective mass. A

limitation of this technique is observed at frequencies above the fundamental frequency

of the reservoir. This problem is discussed in detail by Chopra (Reference 5) . A comparison of hydrodynamic pressure computed using Westergaard ' s theory and a theory for arch

dam by Perumlaswami and Kar (Reference 6) with experimental data is shown in Plate 15. Westergaard' s theory is based on a rigid vertical plane dam, while Perumalswami and Kar's

theory is based on a rigid, constant-curvature arch dam. Both methods accurately predicted the hydrodynamic pressixres at the deepest section of the dam- reservoir interface at a frequency at which the dam was moving essentially as a rigid body.

3. Due to the complex structure and reservoir geometries of the North Fork Dam,

a finite element analysis of the complete system appeared uneconomical. Also, the complex

reservoir geometry renders the continuum boundary value problem intractable. Based on these considerations and the comparison of Westergaard' s theory with test data, the effective mass concept was used in the finite element analysis. The water mass was distributed

at each vertical cross section of the dam based on the approximate equations presented in

Reference 4, using the height as determined at the midsection of the dam crest. In using

this approximate technique, it is assumed that the motion at a mass point is purely

radial. Based on the model test data, this assumption was felt justified, especially in

the more flexible regions of the dam where water mass effects would be greatest. Since

the frequencies of interest for the dam were in the region of 100-200 Hz, limitation of

the effective mass concept to frequencies below that of the reservoir was not thought to

be a significant problem. As a crude approximation to the fundamental frequency of the

reservoir, the frequency equation for a semi-infinite reservoir was used. For a 5.5-ftdeep reservoir, corresponding to the maximum height of the model dam, this frequency is

VI-89
approximately 1240 Hz, much greater than the 100- to 200Hz band of interest for the model

dam response.

c) Calculations of Mode Shapes and Natural Frequencies

1. Mode shape and natural frequencies of the model dam were determined through

use of the Ritz analysis (Reference 2) and a subspace iteration in the finite element

analysis. The Ritz analysis is based on selecting a set of force patterns with corres- ponding deflected shapes similar to the structure's vibration mode shapes. Obviously,

this method requires some prior understanding of the structure's dynamic response. The

siibspace iteration technique, which is more efficient when the band width of the system

is large, was developed by Bathe and is discussed in detail in Reference 7. Essentially,

this technique relies on repeated application of the Ritz method, each step depending on

the base vectors developed in the previous step until convergence to the desired eigenvectors is obtained. A simultaneous iteration on the mass and stiffness matrices is performed, beginning with a set of trial vectors.

2. In order to determine the sensitivity of the Ritz technique used in this study

two runs were made. In the first run, only the crest elements were subjected to the

static load patterns; in the second run, however, these crest elements plus elements below

the crest sections were loaded. Efficiency in selecting elements to be loaded is discussed

by Wright and Miles in Reference 8. In general, more elements in areas of greater flexibility should be loaded. The results of the two runs are given in Table 3. The average

decrease in frequencies for the first four modes is approximately 19 percent. In Table 4,

frequencies determined using the more accurate Ritz analysis are compared with those calculated using the subspace iteration scheme with 11 working vectors. As expected, the subspace iteration technique gave the lowest values of the calculated frequencies. The calculated results were still appreciably higher than those determined experimentally. The

fixed-base assumption was then investigated by including the massive concrete foundation.

The accuracy of this calculation based on a comparison with experimental data showed con- siderable improvement over the previous results. The results of the calculations made

using the subspace iteration scheme for the dam fixed at the base and with the foundation

included are compared with experimental data in Table 5. The comparisons are shown for

both full and empty reservoirs. The effects of water mass in reducing the computed natural

frequency are consistent with the reduction in frequency measured for the second and

fo'orth modes. However, no significant reduction in frequency was measured for the third

mode. The normalized mode shape calculated for the dam with the fixed base and foundation

are compared with experimental results in Plates 17 and 18 for the symmetrical and

asymmetrical modes, respectively. Results of both calculations agree equally well with

the measured mode shapes.

6. Conclusions

a) Vibration tests conducted on the model dam showed that interaction with the reser- voir decreases the natural of the dam. The frequencies decreased by 20-30 percent for

all but the third mode. The decrease in the third mode was about 3 percent. Apparently,

VI-90
the change in frequency due to reservoir interaction was not uniform for all modes.

b) Hydrodynamic pressures calculated using Westergaard ' s theory and that proposed by

Perumalswami and Kar indicate that the effects of curvature are not great, at least for

the lower natural frequencies. The point of major interest concerning hydrodynamic pressures was the magnification of pressures in the most flexible regions of the dam. This

increase in pressure coupled with the lower moment capacity of this section should be

given special attention in design considerations.

c) Since the load patterns used in the Ritz analysis reflect the manner in which the

inertia distribution of the structure is being approximated (Reference 9) , computational

accuracy increases substantially when the number of loaded elements is increased. Methods

discussed by Wright and Miles (Reference 8) should be considered in selecting the number

and locations of nodes to be loaded. Since the first step of the subspace iteration method

is essentially a Ritz analysis, results from the subspace scheme should yield a lower bound

on computed frequencies. There were no striking dissimilarities in the mode shapes predicted through use of the two schemes . From these considerations , it can be seen that the

Ritz analysis used in the early versions of the Structural Analysis Program can be attractive from an economical standpoint. However, when a higher degree of accuracy is required

in the analysis of geometrically complex structures, the subspace iteration scheme seems very appropriate.

d) When complete fixity of the base is assumed in the finite element analysis, natural

frequencies predicted are higher than those predicted when the fixity associated with the

interaction of the dam and the foundation is used. With the reservoir full, the predicted

frequencies for the case in which the foundation was completely fixed were 20-30 percent

greater than those for the case in which the foundation was included in the analysis

as such. '^he differences in the predicted frequencies were much less for the two cases when the reservoir was empty, which, however, represents a rather academic case.

e) Comparison of natural frequencies determined in the first model study (Reference 1) with those reported herein indicates that structural damage incurred during explosive tests

resulted in the response characteristics of the first model dam being altered. For the

EM tests, these differences are presented in Table 6. Also, a significant peak in the

velocity versus frequency plots was observed in EH tests conducted on the first model.

However, no such peak was noticed in the second model tests. It is felt that the crack

in the first model changed the three-dimensional character of the dam and allowed a type

of cantilever action to result. The altered structure could then respond more efficiently

to the force levels attainable from the EH vibrator and, therefore, some low resonant

peaks were observed. Having no crack, the second model was stiffer and required more

energy from the EH vibrator to develop significant resonant peaks. Due to the amount

of damping, both structural and radiation into the soil, the EH vibrator could not supply

the level of energy required to produce distinguishable resonant peaks in the second model.

VI-91
References

1. Balsara, J. P., Walker, R.E. , and Fowler, J., "Vibration Characteristics of the

North Fork Dam Model," Technical Report N-74-2, Mar 1974, U.S. Army Engineer Waterways

Experiment Station, CE, Vicksburg, Miss.

2. Wilson, E.L., "SAP, A General Structural Analysis Program," Report No. SESM 70-20,

Division of Structural Engineering and Structural Mechanics, University of California,

Berkeley, 1970.

3. Chakrabarti, P., and Chopra, A.K. , "Earthquake Response of Gravity Dams Including

Reservoir Interaction Effects," Report No. EERC 72-6, Earthquake Engineering Research

Center, University of California, Berkeley, 1972.

4. Westergaard, H.M. , "Water Pressures on Dams During Earthquakes," Proceedings,

American Society of Civil Engineers , 57, pp 1303-1318, 1931.

5. Chopra, A.K. , "Hydrodynamic Pressures on Dams During Earthquakes," Journal

,

Engineering Mechanics Division, American Society of Civil Engineers , 93, No, EM6, pp

205-223, 1967.

6. Perumalswami , P.R. and Kar, L. , "Earthquake Hydrodynamic Forces on Arch Dams,"

Journal, Engineering Mechanics Division, American Society of Civil Engineers , 99, No. EM5, pp. 965-977, 1973.

7. Bathe, K.J. , "Solution Methods for Large Generalized Eigenvalue Problems in

Structural Engineering," Report No. SESM 71-20, Division of Structural Engineering and

Structural Mechanics, University of California, Berkeley, 1971.

8. Wright, G.C. and Miles, G.A. , "An Economical Method for Determining the Smallest

Eigenvalues for Large Linear Systems , " International Journal for Numerical Methods in

Engineering , 3, pp 25-33, 1971.

9. Dong, S.B., Wolf, J.A. , Jr., and Peterson, F.E., "On a Direct.j Iterative Eigensolution Technique," International Journal for Numerical Methods in Engineering , 4,

pp 155-161, 1972.

VI-92
-REACTION MASS

5-FT- THICK CONCRETE

Figure 1. Two-way reinforcement of kicker slab

VI-93

DAM CRESTCDG€ or

FOUNDATION APPROX Z' THICK

-AXIS or VIBRATOR

REACTION MASS 5-rT-THICK CONCRETE

PLAN VIEW

VELOCITY

PICKUPS

ELEVATIONS ARE

IN FEET REFERRED

TO MEAN SEA LEVEL DOWNSTREAM ELEVATION SECTION A-A

Figure 3- Plan, elevation, and section of North Fork Dam model

VI-95
Figure k. Small finite element grid

Figure 5 . Large finite element grid including foundation

VI-96
Table 1

EM Vibration Tests Conducted for North Fork Dam Model

Vibration

System

Two em's

Two em's

Two em's Two EM'S

Two EM'S

Two EM'S

Reservoir

Level

Empty

Empty

Full

Full

2h in. down

Full

Remarks

Frequency sweeps at 1 Hz/sec from 30 to

500 Hz, vibrators in phase and out of phase

Damping tests , vibrators shut off at various

resonant peaks

Frequency sweeps

Damping tests

Frequency sweeps

Pressure measurements at various resonant

frequencies

Table 2

Damping Ratios with EM Vibrator and 15.5-Pound Force

Vibrators

Date In Phase (IP) or Pickup Frequency Damping

(19T3) Out of Phase (OP) Location Hz Ratio

Reservoir Empty

21 Feb IP 3 150 0 03

21 Feb IP 3 297 0 02

21 Feb IP 3 323 0 02

21 Feb IP 6 150 0 ou

21 Feb IP 6 297 0 03

21 Feb IP 6 323 0 02

21 Feb OP 3 I3I4 0 03

21 Feb OP 3 150 0 0^

21 Feb OP 3 153 0 02

21 Feb OP 6 13k 0 05

Reservoir Full

22 Feb IP 3 90 0 03

22 Feb IP 3 106 0 oh

22 Feb IP 3

1I+7 0 oh

22 Feb IP 6 90 0 oh

22 Feb IP 6 106 0 oh

22 Feb IP 6 1I+7 0 03

22 Feb OP 3 103 0 05

22 Feb OP 3

10i+ 0 oh

22 Feb OP 3 Ihl 0 03

22 Feb OP 3 175 0 02

22 Feb OP 3 177 0 02

22 Feb OP 6 103 0 05

22 Feb OP 6 10k 0 03

22 Feb OP 6 Ihl 0 02

22 Feb OP 6 175 0 02

22 Feb OP 6 177 0 03

VI-97
Table 3

Sensitivity of Ritz Analysis to Load Pattern

with Reservoir Full and Fixed Base

Natural Frequency, Hz

iMode Load Condition 1 Load Condition 2

1 183 159

2 232 177

3 271 219

h 335 27i+

Table U

Comparison of Natural Frequencies

Determined Using the Ritz Analysis

and the Subspace Iteration Technique

With Reservoir Full and Fixed Base

Natural Frequencies, Hz Mode Ritz Analysis Subspace Iteration

1 159 151

2 177 162

3 219 197

k 27i+ 229

VI-98
Table 5

Comparison of Finite Element Frequency Calcul^itiotu of

Model Dam with Fixed Base and Foundation with

Experimental Data

Natural Frequency, Hz Reservoir Finite El ement Model

Mode Condition Fixed Base Foundat.ioti Experi mental

1 Full 151 12).

1 Empty 171 160

n Full 162 122 103

2 Empty 185 162 133

3 Full 197 151 1)^7

3 Empty 226 196 151

1+ Full 229 182 175

h Empty 26k 231 213

Table 6

Natural Frequencies for 1st and 2nd North Fork Dam Models (Reservoir Full)

Natural Frequencies? HZ Mode 1st Model 2nd Model

1 94

2 102 103

3 118 147

4 150 175

VI-99
NORTH FORK DAM MODEL

VELOCITY VERSUS FREQUENCY

EM TESTS

RESERVOIR FULL

PLATE 1

VI-100
NORTH FORK DAM MODEL

VELOCITY VERSUS FREQUENCY

EM TESTS

RESERVOIR EMPTY

PLATE 2

VI-101
NORTH FORK DAM MODEL

VELOCITY VERSUS FREQUENCY

EM TESTS

RESERVOIR DOWN 24 IN.

PLATE 3

VI-102
RCKUP NO

SECTION A-A

NORTH FORK DAM MODEL

VELOCITY VERSUS FREQUENCY

EH TESTS

RESERVOIR FULL

PLATE 4

VI-103
'o

-2 —

SYMMETRICAL MODES

^ */75 Hx

M 103 Hz

UNSYMMETRICAL MODES

RADIAL CREST

DEFLECTIONS FDR MODES 2-5

EM VIBRATOR TESTS

RESERVOIR FULL

PLATE 5

VI-104
PLATE 6

VI-105
DISPLACEMENT, lO"* IN. 2 1 ° ' i 3 2 1 0 1

I I

I if I I II "I t

SYMMETRICAL

MODES

I r

UNSYMMETRICAL

MODES

RESERVOIR FULL RESERVOIR EMPTY

RADIAL CANTILEVER DEFLECTIONS

NEAR CENTER LINE FOR MODES 2-5

EM VIBRATOR TESTS

PLATE 7

VI-106
PICKUP LOCATIONS

4 5 6

I IN

1 FT

2 FT

q:

inu

(t

a.<

3 FT

"Io -I4 FT

5 FT

5.5 FTHYDRODYNAMIC PRESSURE

FROM EM TESTS WITH

VIBRATOR OUT OF PHASE

FREQUENCY = 1 05 Hz RESERVOlR = 1 IN. BELOW CREST

PLATE 8

VI-107
1234567 PICKUP LOCATIONS 89 I

1 1 \ 1 1 1 1

1

I IN.

u

COI

J 5.5 FTHYDRODYNAMIC PRESSURE

FROM EM TESTS WITH

VIBRATOR IN PHASE

FREQUENCY = 147 Hi

RESERV0IR = 1 IN. BELOW CREST

VELOCITy= 0.36 IPS, PICKUP NO. 3

PLATE 9

VI-108
PICKUP LOCATIONS

4 5 6 T

I IN

U

ITD

(O

Ui

q:

QL<u

0.

2 FT

3 FT

O J. 4 FT

a

UJ

»-I

$

O

_l u

IDI

t0.

5 FT

5.5 FT

HYDRODYNAMIC PRESSURE

FROM EM TESTS WITH

VIBRATOR OUT OF PHASE

FREQUENCY = 1 76 Hi

RESERV0IR = 1 IN. BELOW CREST

VELOCITY= 0.24 IPS, PICKUP NO. 3

PLATE 10

VI-109
0.15 0.1 0.05 0 0.15 0.1 0.05 0 0.15 0.1 0.05 0

PSI

147 H z- IN PHASE 105 Hz-OUT OF PHASE 176 Hz-OUT OF PHASE

VELOCITY TRANSDUCER NO. 6

PSI

147 Hz- IN PHASE 105 H z -OUT OF PHASE 176 Hz-OUT OF PHASE j VELOCITY TRANSDUCER NO. 5

i

I

]

i VARIATION OF

HYDRODYNAMIC PRESSURE [ WITH DEPTH

j EM VIBRATOR TESTS I

1

I PLATE 11

VI-110
1234567 PICKUP LOCATIONS 89

5 FT

5.5 FT

HYDRODYNAMIC PRESSURE

FOR EH TESTS

FREQUENCY = 80 Hi

RESERVOIR = 1-3/8 IN BELOW CREST

VELOCITY= 0.42 I PS, P ICKUP NO5

PLATE 12

VI-111
5.5 FT

HYDRODYNAMIC PRESSURE

FOR EH TESTS

FREQUENCY = 98 Hi

RESERVOIR = 1-3/8 IN BELOW CREST

VELOCITY= 0.45 IPS, PICKUP NO. 5

PLATE 13

VI-112
1234567 PICKUP LOCATIONS 89

1 IN.

2 rr

Ul

ITDm —

3 FT

O

4 FT

>u

-I

(TU

H-Io

5 FT

5.5 FT

HYDRODYNAMIC PRESSURE

FOR EH TESTS

FREQUENCY = 1 22 Hz RESERVOIR = 1-3/8 IN. BELOW CREST

VELOCITY = 0. 80 IPS, PICKUP NO. 5

PLATE 14

VI-113
3 —

1 J L

0.3 0.2 0.1 0

80 Hz

VELOCITY TRANSDUCER NO. 5

VARIATION OF

HYDRODYNAMIC PRESSURE

WITH DEPTH

EH VIBRATOR TESTS

AT 20*^ FORCE

PLATE 15

VI-114
COMPARISON

OF THEORETICAL

AND EXPERIMENTAL

HYDRODYNAMIC PRESSURES

PLATE 16

VI-115
1ST SYMMETRICAL MODE

2ND SYMMETRICAL MODE

COMPARISON OF EXPERIMENTAL

AND ANALYTICAL NORMALIZED

MODE SHAPES

RESERVOIR FULL

SYMMETRICAL MODES

PLATE 17

VI-116
1ST ASYMMETRICAL MODE

2ND ASYMMETRICAL MODE

COMPARISON OF EXPERIMENTAL

AND ANALYTICAL NORMALIZED

MODE SHAPES

RESERVOIR FULL

ASYMMETRICAL MODES

PLATE 18

VI-117
THE MEASUREMENT OF THE DYNAMIC k-VALUE IN SITE

AND ITS APPLICATION TO DESIGN

TETSUO KUNIHIRO

KANAME YAHAGI

MI OHIO OKAHARA

Structure and Bridge Division

Public Works Research Institute

Ministry of Construction

Japan

ABSTRACT

The determination of the dynamic soil k property is deteirmined by a series of bore

hole tests. Results from these tests are discussed and the significance of these data

are presented.

Key Words: Bore Hole ; k Property; Soil ; Stiffness; Proper ty ; Testing.

1. Introduction

The earthquake-proof design of structures including the foundation, is performed

by the so-called seismic-coefficient method in which the dynamic phenomena is substituted

by a staticphenomena. With recent advances in the method of structural analysis and computer systems, calculation of the complicated dynamic phenomena has become

possible. However, in examining the effects of earthquakes relative to actual structures,

it is important to know the nature of the structura-1 materials. This is especially true

concerning the dynamic characteristics of the ground, instead of mere knowledge of the

nature of the dynamic load as an input and the adequate analytical method.

Calculation of horizontal resistance of piles is usually made by using the values

of the modulus of lateral reaction of the ground, k-values, assuming that piles are beams

on an elastic foundation. The k-values are obtained by utilizing the data from bore

holes at the site. The evaluation of the k at the time of the earthquake is obtained

from static tests and then multiplied by a certain conversion coefficient. Scientific

research on the dynamic properties of soil has been conducted, to some extent, but

few have applied the research to practical problem of design of foundations.

In this report, description of the development of a device for measurement of the

dynamic k-values by utilizing boring holes, will be given. This will therefore meet

the urgent need for determining k-values at the site.

2. Basic Problems Concerning k-Value

Definition of the modulus of dynamic lateral reaction force (k-values) of the ground,

is quite difficult because of the following reasons: Complexity of the forms of stratums

that compose the ground; even if the domain of strain is small the stress-strain relation

is non-linear under the loaded conditions: change in the dynamic properties of the

VI-118
ground becomes further complicated due to the extent the strain is sustained in each

stratum.

If the ground is perfectly elastic, no change occurs in the k-values , even though

the working stress may differ in its magnitude and velocity. However, the ground is

not perfectly elastic. Therefore, in order to obtain a reasonable k-value, the effects

of strain, load velocity, number of vibration, etc. should be taken into consideration

in the evaluation of the real k-value.

In order to deal with the above problems, it may be necessary to obtain k-values

by examination of many stratiims and select k-values that are useful in the actual design

from among the acciamulated data during the actual measurements.

3. Basic Problems in Measurement of Static k-Values

The following methods can be used to measure the static k-value at site: Load test

using the steel load plate, 30-50 cm square; calculating back to the results of load

test by means of the Chang's method; horizontal load test utilizing boring hole. Among

these methods, the horizontal load test utilizing boring hole is used most frequently,

because of the ease of performance.

The load test using boring holes as presently performed in this country, consists of

two methods i.e., examining the difference in load and the displacement control. The

first, the uniform load method, measures the mechanical constant and shearing strength of

the ground around the boring walls obtained from the load-displacement correlation as obtained by applying equal direct pressure to the whole circvmiference of the test wall.

The other method, uniform displacement, has similar measurements but controlling the

uniform displacement of the wall of the boring hole in a specified direction.

The preciometer shown in Fig. 1, and L.L.T shown in Fig. 2 is the tester for the

uniform load method. K.K.T. shown in Fig. 3 is a tester for the uniform displacement

method

.

As shown in Fig. 3, the body of the loading part of the tester, K.K.T. for the

uniform displacement method, is a small-sized special jack. Two high-pressure hoses

from this body are connected to the high-pressure hydraulic pump on the ground. The

cylinder of the loading body has 2 holes, one for sending oil to push out the piston

and the other for sending oil to push the piston back into the cylinder after the load

test. When pressed from one direction during the measurement, the piston is pushed out

to give pressure to the load plate. Then, the oil filled aroxind the piston in the

cylinder flows backward to the ground passing through the other hose. The amount of

the oil flowing back is proportional to the displacement of the load plate. Utilizing

this phenomenon, the flow taken into the small-bore stand pipe and the displacement of

the load plate is measured.

The method of the load test in the hole, using the K.K.T. tester, is called the

uniform displacement method because the steel load plate outstretches the hole wall

horizontally by the force of the jack to give rise to a uniform displacement. In short,

the displacement is similar to the horizontal displacement of a pile which occurs in

VI-119
the load plate during the test. Aside from the merits and demerits of the testers , it has been clarified by past test

results by the authors, that there are no great differences in the measured values among

the testers, if the hole walls and ground are under the same conditions. (Literature 1-2)

That the condition of the hole wall during the test, is the greatest factor which

gives rise to errors on the results of load test in holes, has been confirmed. It has

been noted that differences in measured values, caused by the testers, is so small in

comparison with the difference in the degree of irregularity of the boring wall surface.

Basic problems of the bore hole field test, in addition to the measuring of the

static k-value, lie in the fact that the boring technique and the measuring technique

are difficult.

If the boring wall surface became irregular or uneven in the boring process or collapses during the load test, the data obtained thereforewill be erroneous and unusable, even though the test may be conducted carefully. Because no excellent boring technique

can be expected at all times ^ § device is required to develop favorable boring

wall conditions at all the times during the test. This is achieved by attaching a boring

trimmer to the tester in order to obtain the accurate data during measurements.

Fig. 7 shows the dynamic k-value measuring tester. As just described the trimmer is

required and is fitted at the edge of the tester of the bore hole not only for the dynamic

load test but for the static load test. The testing-method of the K.K.T. tester with attached trimmer, is to dig a hole whose

bore is a little smaller than the load plate of the tester. Then enlarge the hole by

cutting with the trimmer until the load plate comes in close contact with the boring hole,

and then insert the load plate. Because of the trimming, the test can be performed with

a boring wall which has no irregularities.

The irregularity of the boring wall creates an tinderestimation of static lateral kvalue. In this regard, Nishigaki (Literature-5) compared the modulus of deformation obtained from various tests of viscous soils and pointed out that the modulus of deformation

obtained from the load test in the boring hole varies greatly under the effect of irregularity of the boring wall.

Fig. 4 compares the results of the measurement of the static k-value for the alluvial

2 clay of Osaka (W^ = 89%, W = 63%, q^ = 0.67 kg/cm ) for the cases when trimming was made

and was not made.

Fig. 5 shows the results of similar comparative test for the diluvial clay (W = 100%

Xj 2 W = 50%, q^ = 5.6 kg/cm ) called Ma-3 that is found in hills of Osaka region.

These figures show that when the boring hole is trimmed, the load displacement curve

shows a sharp start and accordingly a high k-value.

The load displacement curve, in the conventional case where no trimmer is used,

shows no sharp curve at the initial stage and measuring starts at a point on the secondary

ine for the case in which trimming is made.

Fig. 6 shows the measurement results made for the purpose of ascertaining the effects

in the use of trimmer in a single boring hole in diluvial clay of Osaka. This indicates

VI-120
how great the difference of k-values may be dependent on whether the trimmer is or is

not used in hard clay soils, where irregular boring is generally believed not to occur.

4. Dynamic k-Value Tester

The dynamic k-value tester, shown in Fig. 7, is a tester in which the K.K.T.'s

principles have been applied with inclusion of some new devices in its construction.

As shown portions 5, 6, and 7 consist of the same construction of the loading body of

the K.K.T. The load plate 5 is loaded with the boring wall when oil pressure acts on piston 5. Movement of the load plate can not be simple, as in the case of measuring the

static k-value. As mentioned already, the working stress, amplitude and number of vibrations must be interchanged. For instance, when the load plate is charged with sine-wave

type load, the oil pressure on the piston must be varied.

When using the K.K.T. for static tests, the oil pressure is given to the piston from

the hydraulic pump on the ground through the high-pressure hoses (refer, to Fig. 3) . If

the wave type load is applied by attaching a servo-valve etc. to the hydraulic pump,

attenuation of pressure occurs in the process of transmitting pressure to the loading

body in the boring hole through a slender high-pressure hose, and thereby makes it impossibil to transmit the desired load change. This is always the case, when a pressure control

|| is conducted on the ground, even if any other tester (for instance, a rubber tube) is

used.

In this dynamic k-value tester, the mechanism corresponding to the hydraulic pump

on the ground is replaced with the portions given as 10, 11 and 13, and the pressure control isidevised to go directly above the test place. The piston of the hydraulic pump on

the ground is substituted by the piston 10. The cylinder shown by the thick black lines

I in the diagram is filled sufficiently with oil to actuate freely the piston 8. Free

performance of the piston 10 under principles that will be mentd.oned later, causes the

required orderly wave-form oil pressure load to act on the piston 6. Then the load plate

5 gives the required load change to the ground around the boring wall.

The operative power of piston 10 is given from the micro air cylinder 13 through the

coupler 11. The high-pressure air required for the operation of 13 is supplied from the

j high pressure tank of the air compressor on the ground through the air hose and the

j regulator 16. The regulator is capable of regulating the pressure from zero to a maximum pressure

by moving up-and-down the pin projected from the upper part of the regulator.

The pin projected from the regulator 16 moves up-and-down in conformity to the

rotation of the cam circular plate 18, to the height of the cam connected directly to

the DC motor 20 with a reduction gear. The air pressure in the air cylinder can be

changed by adjusting the shape of the cam circular plate to regulate its rotation. Next,

j the air pressure in the air cylinder and the movement of piston 10 and finally the load

I on the load plate can be changed freely to meet the test conditions.

I The stress that acts on the load plate causes a complicated displacement to the

I ground around the boring wall. Displacement is measured indirectly by the linear motion

I VI-121
potentiometer 12 directly connected to the coupler 11 and then registered.

The change in oil pressure in the body, which acts on the piston 6, is measured by

the pressure transducer 4 inserted at the edge of the hydraulic cylinder of the body and

then registered together with the displacement of the load plate.

The notable feature of the dynamic k-value tester lies in the fact that installations on the ground are successfully housed in the body of the tester, which was considered

difficult for conventional type of testers. Thanks to completion of this tester, accurate

dynamic load tests can now be performed easily.

The development of a superior tester can be obtained by remodeling this tester as determined from field experiments.

5. Results of Measurement of Dynamic k-Value

Shown is an example of the analysis made using the results of the measured stressdisplacement from the tester, when a sine-wave type load is given to the load plate.

(Experiment at Urayasu in Chiba Pre)

Experimental results of the dynamic k-value tester are sent to a data recorder as shown in Fig. 8. Fig. 9 shows the results of the stress-displacement measurement. The wave form shown on the upper part, represents the stress and that on the lower part represents the displacement.

As shown in this Fig. , similar strains occur in the ground around the boring wall in

response to the sine-wave type stress. With the increase in number of repetitions of the

load, the displacement of the boring wall gradually increases.

The results of the vibration test of the linear elastic body are shown in Fig. 10. When a sine-wave type vibration stress is given to the linear elastic body and a stationary state is reached, the strain phase of the elastic body lags by 6 from that of the

stress. This is because the stress energy is consumed in the elastic body.

Fig. 11 shows the above phenomenon by a hysteresis loop. In this Fig. , the loop

area, Aw, shows the loss of energy consumed in one cycle of the steady vibration. The

energy given, W, is indicated by the area of Aabc.

The damping factor (Loss tan6) indicated the ratio of Aw to W is defined by the

following equation.

tan6 = AW cd

(When 6 is extremely small)

n W a-b

From the hysteresis loop, the dynamic elastic constant is obtained as follows.

o\ — 02 b

ei - £2 a

VI-122
If the dynamic elastic constant is equal to the static elastic constant, the following

formula is obtained.

El _ Z2

Generally, in the case of vibration under the initial stress, O , the A.B axis increases

o

as compared to the O.A axis, as shown in Fig. 12. As a result:

e 1 Z2

Accordingly, Ed > > E^.

The k-value, can be found by substracting the above mentioned strain quantity, e, by the displacement of the ground, X, and all other elements may be assimied the same,

viz. P^, corresponds to stress and to as given in Fig. 9 and the corresponding

displacements are and X2. The E-values are obtained from Fig. 12, therefore three

k-values are given as follows

:

kah-i = -v'— (max. stress/max. displacement!

1

P kdh-2 = (min. stress/mi n. displacement,

X 2

= ^ ^ = dynamic k-value (kdK value)

Xi - X2

The dynamic k-value, given as the value of the , is obtained from the inclination of the line A-B. For reference, the dynamic k-value is calculated as follows, using the experimental

record of Urayasu in Chiba Pre., GL - 17.5 m.

Stress

Displacement

Pi = 0.367kg/cm2

P2 = 0.231kg/cm2

P3 = 0.136kg/cm2

Xi = 0.0239cm

h

= 0.0186cm

VI-123
= 0.0053cni

(dynamic k-value) = = q^qq^^ =• 25.7kg/cm3

Fig. 13 shows a trend of the gradual decrease in the dynamic k-value with a small

change in the absolute quantities of the stress and displacement as the result of the

repetitions of the sine-wave type loading.

Fig. 14 shows the results of experiments conducted at Urayasu in Chiba Pre., for

a study of the problems concerned with the k-value. The dynamic k-values, plotted in

Fig. 14, show the values for ten repetitions of the load.

The E-value was converted, using the following equation, using k-values measured

by the tester;

E = K • Ip . D(l -/2)

D : Diameter in terms of load plate

Ip : Form factor (0.79 in the case of circular plate) ^ : Poisson's ratio

The load plate used in this experiment was 8.5 cm x 12 cm. Substituting the above

values in to the following equation, 9.95 cm is obtained as the reduced diameter D.

Area of circle : A

Peripheral length : L

Therefore, the E-value can be obtained from k-values using the following equation.

E = 6.60k

••• D = = 9.95cm

= nD

VI-124
The results of a comparative study where all the test values are converted into E- values, are shown on the right colxiitin of Fig. 14. Correlation between the static k-value and the dynamic k-value indicates that no clear answer can be given unless more data is collected. This correlation is shown as based on the following field test results.

When the k-value measured by the conventional method is assumed 1, the static k-value

measured by K.K.T. , using trimmer 3-4 and the dynamic k-value 6-7. The correlation may of course differ according to the conditions of the ground and the test. In any case, however, it is true that the dynamic k-value is higher than the static k-value. The static

k-value, measured with the tester inserted in the hole without trimming by means of the

conventional method, is low compared with the dynamic k-value. Whether or not to use such

a value in the earthquake-proof design has become questionable, and it has been made known

that there is need to make further studies.

6. k-Value and Foundation Design

A series of experiments, conducted by the authors, have revealed that various k-values

are obtainable by changing the test conditions even in the same soil. The question is

then in selecting the k-value for use in foundation design. This is a great problem which

still remains unsolved at present.

The pending problem lies in how to solve the questions arising from the assumed

conditions for Chang's formula, that is generally used in this kind of calculation.

Chang has assumed a constant k-value, which is not consistent with the test conditions,

and utterly disregards the depth and bending. In the Chang's method the tolerable quantity of displacement is determined first, from which the allowable bearing power is calculated. No clear view, however, has been established in the determination of the tolerable

displacement. This poses another important problem which must be solved.

Reasonable determination of the tolerable displacement is as important as the exact

measurement of the k-values, that suit the design conditions. This fact is demonstrated

as follows;

The bearing power, by Chang's formula, while the pile head is set free will be:

6 = 26H

H = 6kD

2b

Therefore

,

In short, the horizontal force of a pile is proportional to the displacement with

the k-value raised to the 3/4th power. To clarify this, tolerable displacement is

VI-125
doubled and the k-value is multiplied by 2, or;

k

3 3 H = 26 •( )'^-T = 1.196-k'^-T )

The horizontal yield strength of the pile increases by 20%, despite having the k-value

multiplied by 1/2. When the displacement is multiplied by 1/2 and the k-value is doubled,

H becomes;

The horizontal yield strength decreases by about 20% despite having the k-value doubled.

In designs, reasonable determination of the tolerable displacement, together with

reasonable and exact k-values that suits the design conditions, are required. The amount

of the elastic strain of the ground is very small and the allowable horizontal yield cer- tainly becomes larger, if a large amo\int of tolerable displacement is incorporated into

the calculation. Should the tolerable displacement exceed the amount of the elastic

strain of the ground, the calculation of the k-value using the modulus of reaction force

of the ground, becomes utterly meaningless. In such a case, introduction of the concept of

passive earth pressure becomes necessary. The behavior of piles in the ground varies

widely, even under the same horizontal load; these variations are due to construction

materials of the foundation, the ratio of the diameter of pile to its length, method of

treating the pile head; number and space of piles. Therefore, in using k-values, examination of the conditions of the design for the foundation is also required.

In performing this kind of calculation, attention should be given as to whether the

load received by the foundation structure is transitory, constant or repetitious in nature.

Needless to say, if the stratum is subjected to long hours of horizontal pressure

that exceeds the preloading value hold by the stratum, a phenomenon of consolidation takes

place which causes vacant space to occur between the structure wall or piles and the ground.

If a horizontal load is applied again after the occurrence of the vacant space, such

a load must be resisted by the stratum located deeper than the initial one, as the resistance value at the vacant space is zero. As a result, stability of the foundation of

such a structure will be reduced, and sometimes it becomes necessary to make further detailed calculations on the stability, even if safety was assured in the calculations using

k-values

In order to make a rational foundation design, there are many problems to be dealt,

besides the question of the k-value. It has been revealed that there is a k-value, the

basic value in calculation, that is more than 6 times higher than others, according to the

selection of the measuring method. It is necessary to reflect on such facts in the

foundation design. The k-value measuring method, using conventional boring holes which

have been proved to afford too small k-value because of large errors, must be corrected.

3 H = 0.84,s -k"^ • T

VI-126
If the tolerable displacement of the foundation caused by the horizontal load is

within the elastic region of the ground and such a load doesn't work constantly, the kvalues measured by use of trimming method should be used in the calculation. When conditions of the foimdation and ground are clearly known, and dynamic k-value is well adaptable

and can be measured with reliable results, such dynamic k-value should be used in the

calculation.

As regard to the amount of the soil elastic strain caused by each structure, it is

considered necessary to elucidate it theoretically through repetitious experiments in

the future. The authors suggest the following, based on the results of experiments, is a tolerable displacement for which dynamic k-value is applicable in the case of comparatively tight sand stratum (N > 10 - 15) or clay (N > 3 - 5) and the pile diameter is

smaller than 100 cm.

Tolerable displacement = Pile diameter x 0.1 x 0.25

The breaking strain of the pile, whose diameter is not large, is about 10% of the

pile diameter. It may not be a great error to consider the elastic region to be about

25% of the diameter.

At present there are many problems associated with the dynamic k-value and the design

of the foundation.

7. Postscript

Despite difficulties in construction work due to the lack of suitable ground in the narrow land of this country, there are many cases in which important structures must be

built. In addition, there is an unfavorable condition of frequent occurrence of earthquake which makes it quite difficult to design a safe and rational foundation.

The authors wish to continue the study of the dynamic k-value problems, because we believe that many problems may be solved in the course of this study which in turn would

contribute to the improvement of design techniques for foundations used in this country.

References

1. Yoshiharu Arie, Tetsuo Okada, Kaname Yahagi (1970) , Horizontal Resistance to Steel Piles in the Arakawa Canal, (1) Difference in k-Value in Boring Holes Resulting from

Difference in Boring Measuring Instruments, Soil and Foundation, Vol. 18, No. 9.

2. Yoshiharu Arie, Tetsuo Okada, Kaname Yahagi (1970), Horizontal resistance to

steel piles in the Arakawa Canal (2) Study on k-Value Obtained in the Horizontal Load Test

3. Harumitsu Tamano, Kaname Yahagi, Kozo Miki (1974) , Basic Study of Dynamic k-Value,

Soil and Foundation, Vol. 22, No. 3

4. Kozo Miki, Hideo Yugami, Matsuo Amamiya (1973) , Measurement by Superhigh-Pressure

K.K.T. tester of strength of Rock Bed, Soil and Foundation, Vol. 21, No. 2.

5. Yoshihiko Nishigaki, Yoichi Okajima (1972) , Modulus of Deformation of Viscous

Soil, Lectures to the 27th Annual Meeting of the Civil Engineering Society.

6. Yoshihiko Nishigaki, Tadayuki Hirobe (1973) , Hysteresis Characteristic of Osaka

Diluvial Clay Announcement at the 8th Meeting of Geotechnique Society.

VI-127
7. Report for Tokyo Expressway Public Corporation Geological Survey at BC-282

Engineering Zone.

VI-128
VI-129
General view of Preciometer

Fig. 1 Block Diagram of Preciometer

(Pressure-Volume Meter)

Muddy water

General view of LLT

Fig. 2 Block Diagram of LLT

VI-130
VI-131
1.0 , 2.0

Amount of displacement Ar (cm)

Fig. 4 Effects of Trimming on k-Value (Osaka diluvial clay)

VI-132
Ar (cm)

Fig. 5 Effects of Trimming on k-Value (Osaka diluvial clay)

VI-133
k-value kg/cm^

t L.L.T . K.K.T

o K. K.T wi th trimmer

E-value (Kj/cne)

1- L.L.T . K.K.T o K.K.T with trimmer

500 1000 1500 2000

Fig, 5 Comparative Test of Static k-Value

VI-134
VI-135
1 o J- V- ^ O) o

c

OJ QJ JZ cn o

t-J CD rv ^ > o

s~ E »-> 1- 3

O E c U- -r- u X

Ol to o > E +-> J Ol

I/) -C

a*

i-o •r- X! s on c

1

o O i/l i.

*-> >, O o c o

OJ

S- O) o

13 i/l cr. 4- e cn

'yi O CD

O) SIZ -t-J -u Q. 03

i- o

3: s-

— +J rtS Q. S_ (—

on OJ Q.

r- >

"O c -o O 03 (DUO

-C .—

OJ OJ

on +-> _c

Ol 03 ^

on -O O

03 E

OJ 03 +J E C

1- <u

i- Ol E

Ol "O QJ

!-> e u

OJ -r- 03 E <— >—

>i Q.

"-J <J l/l C -1- O S_ T3 E '-

CI fO Ol

03 OJ !-> — XI

CL -U O

on +->

OJ

s- DQJ^

-o -c: o 03i

OJ a

o

r

l_

4-'

O)E^o a

u E

03 03 a.

on

OJ 2:

c

o

5

4> k-^

I O

—

E 03

VI-136
Fig. 9 An Example of Dynamic k-Value Measuring by Sine-Wave

Stress Method

Fig. 10 Diagram Showing Results of Vibration Test on Linear Elastic Body

VI-137
0 £2 £o £i

Strain

Fig. 11 Hysteresis Loop of Linear Elastic Body

Vibration Test

I

! Fig. 12 Comparison of Elastic Constant ( Ed > Ei - E2 ) j

I

I

I

i

i' VI-138
0.02

X cm

Fig. 13 Decrease of Dynamic k-Value Resulting from

Increase in Number of Repetition

VI-139

LABORATORY INVESTIGATION OF UNDISTURBED SAMPLING AND

S TANDARD PENETRATION TESTS ON FINE SANDS

W.F. MARCUSON, III

S.S. COOPER

M.A. BIEGANOUSKY

ABSTRACT

Determination of density of sand is presented. Densities determined from using undisturbed samples and from the standard penetration tests are compared. It is shown that the

standard penetration test is not sufficiently accurate to be recommended for final evaluation of the density at a site unless site specific correlations are developed.

Key Words: Density; Liquefaction; Sand; Soils; Tests; Undisturbed Sampling.

1. Introduction

a) Reliable determinations of in situ density of sands below the water table are essential to assess the engineering properties of such materials particularly when liquefaction

may be a problem. This problem has become more critical in recent years because more and more structures are being constructed which would cause catastrophic damage and loss of

life if they failed as the result of stresses attributed to earthquakes. This is particularly true in the case of nuclear power plants and large dams. The critical nature of

this problem, and the difficulties encountered in actually determining the in situ density

has led the Corps of Engineers to extend previous work (1,2,3). The earlier work was conducted at the Waterways Experiment Station (WES) and resulted in the development of an un- disturbed sand sampler, which incorporated a fixed piston and drilling mud to obtain representative samples in situ (1).

b) Less direct methods of determining the in situ density have also been developed

for use in the field. The most common technique for estimating the in situ density of a deposit is the Standard Penetration Test (SPT) (4-8). This method is based on correlations between penetration resistance, overburden pressure, and relative density. The SPT

is widely used because it is expedient and economical. However, the empiricism of the

technique has led many engineers to doubt the validity of the correlations. WES has

evaluated the SPT procedure in the laboratory, the results of which will be presented in

the literature (9), and are summarized here.

c) In conjunction with the SPT investigation, WES also studied the undisturbed fixed

piston sampler. This was accomplished by comparing sampled densities obtained under laboratory conditions, to the actual density of the sampled medium. Results of the comparisons

are also presented. If desired, more detailed documentation will be available in the

literature (10).

VI-141
2. Test Facility

a) The major components of the test facility consist of a 4-ft-(1.22 m) diameter

ring soil container, a massive reinforced concrete foundation, and loading equipment for

applying pressure to the top of the soil specimen (9-11) . Figure 1 shows the circular

stacked ring soil container in which 4-ft- (1.22 m) diam by 6 ft (1.83 m) high specimens were constructed. A schematic of the test facility is shown in Fig. 2.

3. Sampling Equipment

a) The drilling rig used in this study is a commercially available skid mounted Acker

Teredo Mark II Soil Sampling Drill. Low overhead clearance in the laboratory precluded

the use of a mast, so lifting tackle for the drill rods was secured to rafters in the

ceiling. For sampling and drilling operations the rig was elevated on a platform built

level with the top of loader support pedestals. Mud circulation during reaming operations

was accomplished with a portable mud pump which was supplied with mud from a pump atop

the loading head.

b) Undisturbed samples were taken with a Hvorslev (12) 3 in. (7.62 cm) ID thin wall

fixed piston sampler. Nominal dimensions for the 16-gage sample tube included a 2.97 in

(7.54 cm) ID cutting edge, a taper angle of 10 degrees on the cutting edge, and an area

ratio of 11 percent (13). A schematic of this sampler is shown in Fig. 3. The drill rod

used in the sampling operation was 2 in. (5.08 cm) OD "N" size. Reaming was accomplished

with the WES-modified fishtail bit. The bit itself is commercially available, however,

special baffles were added at WES to direct the flow of drilling mud in an upward

direction, away from the bottom of the borehole. Thus, disturbance of the underlying

material by circulating mud is minimized when reaming or drilling to the sampling

depth

c) The split-spoon sampler was driven by a hydraulically operated 140 lb (63.42 kg)

hammer, contained in a perforated cylinder, which was designed by the Vicksburg District,

Corps of Engineers, Fig. 4. This hammer was designed to deliver an identical impulse for

each blow, eliminating the undesirable effects of impulse variations on blow counts. The

hammer is mechanically lifted to a 30 in. (76.2 cm) drop height by two lugs positioned on

a continuous chain. The chain is driven by a hydraulic motor connected to the hydraulic

system of the drill rig.

d) The split-spoon sampler used in the conduct of the test conforms to the specifications outlined in ASTM D1586-67, Penetration Test and Split Barrel Sampling of Soils (14).

4. Test Program

a) The test program consisted of undisturbed sampling and standard penetration testing

on twenty-six 4 ft (1.22 m) diam by 6 ft (1.83 m) high sand specimens built to various

relative densities and subjected to a range of overburden pressures. These specimens

were incrementally constructed in lifts as will be discussed subsequently. Table 1 summarized the test program, and includes information of a general nature regarding the

preparation and testing of the specimens.

VI-142
b) Two \iniform fine sands (SP) were tested during the program. The first sand tested

was Reid-Bedford Model sand, a locally procured material with a coefficient of uniformity

of 1.6. The for Reid-Bedford Model sand is 0.25 mm, and the grain shape is subangular

to subrounded. The second sand tested was Ottawa sand which has a uniformity coefficient

of 1.5, a of 0.21 mm, and a subrounded to rounded grain shape. Fig. 5 depicts typical

grain size distributions for Reid-Bedford Model and Ottawa sands.

c) The maximum and minimum dry densities of the respective sands were determined by

procedures contained in Engineer Manual 1110-2-1906, entitled "Laboratory Soil Manual" (15)

.

The values of maximum and minimum density for Reid-Bedford Model and Ottawa sand are shown in Fig. 5.

d) The samples were constructed in approximately 6 in (15.24 cm) thick using various

raining procedures (9,10,16). Beginning with Test No. 4, density control of the indiviual lifts was monitored using the WES developed box density device (17) . This device is

a 4 in (10.16 cm) by 12 in (30.48 cm) metal form which, including the cutting edges, is

3 in (7.62 cm) deep. A box-shaped scoop and a scraper-type cleanout tool, corresponding

to the desired depth of cleanout, are used to remove the soil sample from within the box.

e) Overburden pressure was applied to every specialty in increments according to a pre2 scribed test schedule. The usual pressures were 10, 40, and 80 psi (69, 2 76, and 552 kN/in , respectively) . There were some departures from the foregoing sequence since some specimens were tested at only one pressure, while others were staged at two or more pressures.

The application of the overburden pressure caused consolidation within the specimen, densification being significant for loose specimens and minimal for dense specimens. A density

correction for overburden pressure was derived from the results on one dimensional con- solidation tests performed on submerged, 3 in (7.62 cm) diam specimens at the same testing

pressures used in the full-scale laboratory test program (9) . The one-dimensional

consolidation test was assumed to be reasonably analogous to conditions in the stacked ring

soil container.

f) As previously mentioned, undisturbed sampling was conducted in conjunction with

standard penetration testing. The sampling was accomplished in stages to conform with

the penetration tests planned (see Table 1) , and all of the undisturbed samples were taken

through the center hole in the loading head, i.e., along the cylindrical axis of the

specimen. Sampling was originally planned at overburden pressures of 10, 40, and 80 psi

2 (69, 2 76, and 552 kN/m , respectively) , however, this could not be accomplished m many

instances, because the drill rig employed did not develop sufficient force to drive

2 the sampler at pressures greater than 10 psi (69 kN/m ) . Consequently, most of the

2 undisturbed 3 in (7.62 cm) QD samples were taken with 10 psi (69 kN/m ) applied to the

specimen. After each sampling operation the center, or sampling, hole was stemmed with

pipe before penetration tests were conducted in the peripheral holes . This practice was adopted to prevent hole closure and concomitant reductions in lateral pressure.

g) The test sequences was generally as follows;

2 1. With 10-psi (69-kN/m ) pressure applied, the first tube was driven 2 ft

(0.61 m) into the specimen through the center hole in the loading head. After withdrawal,

VI-143
the center hole was stemmed with pipe 2 ft (0.61 m) in depth.

2. Split-spoon tests were conducted in a peripheral hole to a total depth of

6 ft (1.83 m) , after which the peripheral hole was stemmed.

3 . The stemming pipe was withdrawn from the center hole and the overburden pres2 sure raised to 40 psi (276 kN/m ). Next, the hole was reamed to a depth of 2 ft (0.61 m)

and the second undisturbed sample was taken from 2 to 4 ft (0.61 m to 1.22 m) in depth

(by temporarily reducing the overburden applied, if necessary). Afterwards, the hole was stemmed to 4 ft (1.22 m) in depth.

4. Penetration tests were conducted in the second peripheral hole as in b,

above.

5. The last undisturbed sample was taken from 4 to 6 ft (1.22 m to 1.83 m) in

depth as described in c except that, conditions permitting, the last sample was driven with

2 80 psi (552 kN/m ) overburden pressure applied to the specimen.

5. Penetration tests were conducted in the third peripheral hole, concluding the

test.

h) Sampling and incremental density determinations were carried out in accordance with

procedures outlined in Ref. 12 and 13. The undisturbed samples were sealed with end

packers and stored overnight in a vertical position. The following day, the tubes were placed in a horizontal position in a wooden carrying rack built for the purpose. The

top surface of each tube was marked so that this orientation could be maintained throughout the density determinations. Each tube was then tapped 50 blows with a rubber hammer

along its top surface to consolidate the sand (25 blows in one direction and 25 blows

in the opposite direction along the tube). The tubes were later cut into 6 in (15.24 cm)

increments for density determinations. Density determinations were also made on shorter

sections if enough additional material remained after the tube was cut into 6 in (15.24 cm)

increments

5. Analyses and Presentation of Results

a) Undisturbed Sampling: It is obvious that an assessment of sampling accuracy is no better than the accuracy with which the specimen conditions can be determined at the time

of sampling. Given the nature of soil materials and the current state-of-the-art of soil

measurements, it is equally obvious that the test results will be accurate only to within

certain limits. Accordingly, it is necessary to evaluate, by the most practical means,

any variables which are believed to influence the tests results significantly. Considering the test procedures employed in this investigation, the most significant variables

to be evaluated were as follows;

Frotable r.a-nre ci" Variables Vari?.tior.. rc: Box density neasurenent error +0.2 (+^3.20 kg/n"*)

3 Specimen nonur.ifcrziixy +2.0 (+^32. CU kg/m / Density change due to overburden

^ pressure

- " -

_

+0.7 (+11.21 kg/3^)

( average

)

VI-144
This table if formulated from data in Ref . 10. accumulated upper bounds for all three vari

ables would thus be + 2.9 pcf (46.45 kg/m^) and the corresponding lower bound would be -1.

3 pcf (-24.03 kg/m ) . The positive (+) bias is assumed to represent overburden pressure

affects

.

b) During the course of this study, data such as shown in Fig. 6 were typically

obtained. The best quality data available from the study was plotted in Fig- 7 as sampled

dry density versus placed dry density. A linear regression analysis was performed on the

data shown and the derived linear fit is indicated by the dashed line on the plot. The

equation of the line is

:

Y = O.olX + 19-5

where: Y = sampled density, pcf

X = placed density, pcf

The correlation coefficient for the linear fit to the data is r =0.83.

sy

c) The placed dry density can be corrected for applied overburden pressure (9,10).

Fig. 8 shows the data plotted as sampled dry density versus corrected placed dry density.

A statistical analysis was also performed on this data and the results are shown in Fig.

3.

d) The corrections for density changes due to overburden pressure are apparently

very significant to the test results as shown by the difference in the regression fits

in Fig. 7 and 8. The random variables associated with placed density determinations are much less susceptable to evaluation and cause some degree of uncertainty in an assess- ment of sampling accuracy. However, an assessment can be made from Plots 7 and 8. Assuming a normal frequency distribution, 95 percent of the sampling data falls within

the range of +2a, where O is the standard deviation. Within this range, sampling accuracy

is indicated to be +3.4 pcf (54.46 kg/m"^) for the density samples taken at relative

densities ranging from 20 to 60 percent. The linear regression fits, to the data presented in Fig. 7 and 8, indicate that sampling slightly densifies the sand at low relative densities and tends to loosen denser sand.

e) Standard Penetration Tests: Details of this analysis are presented elsewhere (9,

11). A summary plot of these data are shown in Fig. 9 .superimposed on this plot are

2 the scatter bands for 10, 40, and 80 psi (69, 276, and 552 kN/m ) data. A regression

analysis was conducted on these data and the following equation derived

:

D = 8.6 + 0.S3

r

N + 10.^ - 3. 2 (OCR) - 0.2h{a 1/2

0,00U5

VI-145
where: D = relative density, percent

rN = standard penetration N values, blows/ft

OCR = overconsolidation ratio

a = vertical overburden stress, psi v

2 This equation fits the data with a coefficient of determination (r ) of 0.78 and has a standard deviation of +7.6 percent. The maximum difference between relative density predicted by Equation 5 and that observed in the laboratory was +18.9 percent (10)

f) The study by Gibbs and Holtz (4) resulted in a recommended family of curves. These recommended curves are superimposed on the WES data spreads in Fig. 10. Good agree2 ment exists at the 10-psi (69 kN/m ) testing pressure. The Gibbs and Holtz correlation

curves were derived from testing two sands, a coarse well graded sand and a fine well

graded sand with 14 percent passing the No. 200 sieve (4) . Reid-Bedford Model and Ottawa

sands are appreciably different with respect to the grain size distributions of the Gibbs

and Holtz sands. Ignoring all other factors, one might conclude that soil type exerts a significant influence on the penetration results, which is in agreement with Kolbuszewski

(18) . WES has extended the SPT study using Platte River sand (the WES Platte River sand

is very similar to the Gibbs and Holtz coarse sand) , and early results display some agreement with the findings of Gibbs and Holtz.

g) Along with the recommended curves, Gibbs and Holtz also published the family of curves which resulted from testing siibmerged sands. These curves are also superimposed

on Fig. 10. In this instance, there appears to be better agreement at the 40-psi (276

2 kN/m ) testing pressure, but overall poor agreement. These curves were not considered

reliable, and were not recommended for use by the authors (5).

h) The correlation curves based on the study performed by Bazaraa (6) are also

superimposed on the WES data in Fig. 10. These curves are based on actual field data

and take into account many different types of sands derived from many sources. The curves predict lower values of relative density for a given N-value than either the Gibbs and

Holtz or WES curves.

6. Conclusions

a) The conclusions drawn from this study directly apply to only the two sands tested.

The equation and data presented herein are not intended for use in predicting the penetration resistance or relative density of sands in general but are presented to describe

the empirical behavior of these two sands under the stipulated laboratory conditions.

b) Undisturbed Sampling: Considerable attention was given to the test variable

which can influence test results (10) , and it was noted that placed density at the time

3 of sampling probably varied from +2.9 to -1.5 pcf (+46.45 to -24.03 kg/m ) from measured

values due to the combined effects of the test variables.

c) The sampled versus placed density comparisons presented suggest that sampling

accuracy using the techniques described is within +3.4 pcf (+54.46 kg/m^) for 95 percent

of the sampling conducted at relative densities ranging from 20 to 60 percent.

VI-146
However, it can also be concluded that a more meaningful assessment of sampling accuracy

could have been made had it been possible to exercise better placed density control during the study; it is probable that in this event the apparent accuracy of sampling would

have shown a corresponding improvement.

d) Despite the uncertainties cited, results of this and earlier work exhibit generally

similar trends. For instance, this and the preceding work indicate that sampling tends

to slightly densify loose sand (D^ < 40 percent) and tends to slightly loosen denser sand

(D^ > 50 percent) . The plots presenting linear regression fits to selected data also lead

to the conclusion that overburdened pressure corrections can significantly influence the

density determinations made with the sampling techniques described. More definitive

conclusions regarding sampling accuracy cannot be advanced because of the uncertainty

associated with the placed density results obtained in this study.

e) Standard Penetration Tests: The penetration resistance (N-value) was observed

to be sensitive to changes in density, overburden pressure, and overconsolidation ratio.

f ) The SPT as performed was fairly repeatable in a nearly homogeneous deposit (+2 or

3 +3 pcf (+32.03 or +48.06 kg/m ) ) ; however, variations in the testing medium produced

scattered N-values. The results obtained do not permit more than a gross estimate of

relative density from N-values. A single value of penetration resistance can represent

a 30 percent spread (+15 percent) in relative density. The heterogeneous conditions of

the field would seem to make the task of estimating relative density even more difficult;

hence correlations between N-values and relative density should be used cautiously and

mainly in a qualitative sense.

g) It has been recently established that different structures and different lateral

stresses are associated with various sand placement techniques (19) . The results of

this testing program (9) showed that the penetration resistance was sensitive to the

various placement techniques. It is concluded from this that the results reflect variations in the structure and lateral stress conditions of a deposit. In practice, this

sensitivity would tend to obscure the actual density condition.

h) The tests performed on specimens having an overconsolidation ratio of 3 produced

N-values which consistently fell near the top of the scatter band for their respective

pressures. This also indicates that the SPT is sensitive to lateral stress conditions.

i) Multiple regression analyses were conducted and relationships, in terms of the

standard penetration N-value and relative density, were developed. These relationships

appear quite good from a statistical point of view, having a coefficient of correlation, r, of approximately 0.9. The relationship for relative density as a function of N, O^,

OCR has a standard deviation of about +7.5 percent. Caution should be exercised in the

use of this equation to predict relative density because it was developed from a limited

data bank and is only applicable to results on Reid-Bedford Model sand.

j) Based on a comparison between the correlations presented by Gibbs and Holtz,

cind WES, it is concluded that the SPT is not sufficiently accurate to be recommended for

final evaluation of the density or relative density at a site unless site specific correlations are developed. However, the SPT does have value in planning the undisturbed

VI-147
sampling phase of the subsurface investigation and for purposes of qualitative comparisons

Acknowledgement

Grateful appreciation is expressed to those colleagues at WES who assisted in the

data acquisition phase of this undertaking and to those who offered constructive criticism

as well as moral support during the development and documentation stages. In addition,

the support of the Office, Chief of Engineers, who funded this project, is gratefully

acknowledged

References

1. Goode, T.B., "Undisturbed Sand Sampling Below the Water Table," Bulletin No. 35,

Jun 1950, U.S. Army Engineer Waterways Experiment Station, CE, Vicksburg, MS.

2. "Density Changes of Sand Caused by Sampling and Testing," WES Potamology Inves- tigations, Report 12-1, June 1952.

3. Potamology Report 18-1, "Rotary Cone Penetrometer Investigations," Jun 1962,

U.S. Army Engineer Waterways Experiment Station, CE, Vicksburg, MS.

4. Gibbs, H.J. and Holtz, W.G. , "Progress Report of Research on the Penetration

Resistance Method of Subsurface Exploration," EM 314, Sep 1952, United States Department

of the Interior. Bureau of Reclamation, Denver, CO.

5. Gibbs, H.J. and Holtz, W.G., "Research on Determining the Density of Sands by

Spoon Penetration Testing," Proceedings , Fourth International Conference on Soil Mechanics

and Foundation Engineering , London, Vol. 1, 1957, pp 35-39.

6. Bazaraa, A.R.S.S., "Use of the Standard Penetration Test for Estimating Settlements of Shallow Foundations on Sands," Ph.D. Thesis, Sep 1967, University of Illinois

at Urbana, University Microfilms, Inc., Ann Arbor, MI.

7. De Mello, V.F.B., "The Standard Penetration Test", Fourth Pan-American Conference

on Soil Mechanics and Foundation Engineering, Vol. 1, 1971, pp 1-86.

8. Peck, R.B., et al., "The Standard Penetration Test, Discussion," Fourth Pan- American Conference on Soil Mechanics and Foundation Engineering , Vol. 3, 1971, pp 59-60,

80-84, 89-103.

9. Bieganousky, W.A. and Marcuson, W.F., III, "Laboratory Standard Penetration Tests

on Reid-Bedford Model and Ottawa Sand," Research Report in pxablication , U.S. Army Engineer

Waterways Experiment Station, CE, Vicksburg, MS.

10. Cooper, S.S., "Laboratory Investigation of Undisturbed Sampling of Cohesionless

Material Below the Water Table," Research Report (in publication), 1976, U.S. Army Engi- neer Waterways Experiment Station, CE, Vicksburg, MS.

11. Marcuson, W.F. , III, and Bieganousky, W.A. , "Laboratory Standard Penetration

Tests on Fine Sands," to be presented at the Annual American Society of Civil Engineers,

meeting in Philadelphia, PA, 1 Oct 1976.

12. Hvorslev, M. Jull, "Subsurface Exploration and Sampling of Soils for Civil

Engineering Purposes," Bulletin No. 35,June 1950, U.S. Army Engineers Waterways Experiment Station, CE, Vicksburg, MS.

13. Office of the Chief of Engineers, "Soil Sampling," EM 1110-2-1907, 31 Mar 1972,

Department of the Army, Washington, D.C.

14. "Penetration Test and Split-Barrel Sampling of Soils," ASTM D1586-67, American

Society of Testing and Materials, 1916 Race Street, Philadelphia, PA.

15. Office of the Chief of Engineers, "Laboratory Soils Testing," EM 1110-2-1906,

30 Nov 1970, Department of the Army, Washington, D.C.

16. Bieganousky, W.A. and Marcuson, W.F. , III, "Uniform Placement of Sand",

Journal of the Geo technical Engineering Division . American Society of Civil Engineers.

VI-148
17. Sloan, R.C. , "Dynamic Bearing Capacity of Soils; Dynamic Loading Machine and

Preliminary Small-Scale Footing Tests," Technical Report No. 3-599, Jun 1962, U.S. Army

Engineer Waterways Experiment Station, CE, Vicksburg, MS.

18. Kolbuszewski , J., Discussions, Fourth International Conference on Soil Mechanics

and Foundation Engineering , London, Vol. Ill, 1957, p 126.

19. Mulilis, J.P. , "The Effects of Method of Sample Preparation on the Cyclic StressStrain Behavior of Sands," 1970, submitted in partial fulfillment of the requirements

for the degree of Doctor of Philosophy, Graduate Division, University of California,

Berkeley.

VI-149
Table 1

Si77r-.°.ry Tacle of Tes-3

Cverb'jxder. d r Pressure

Test

.(axerial Placed Placed

(Percent

)

Preparation MethodSeauence

OOP. Undisturbed Water Table Condition Reoiar'^s —

1 RB>St 95 k UC 9 pvCtating Rainer

liO 1 None Submerged Data considered

Unreliable

2 H3MS 95 k 1;0 9 RR 10-!iC-O0 1 Center Hole 3ubr.erged

3 ?3MS 9U h 35 ?i? 5_:4C-oO 1 Center Hole Drained

1* R3MS 91 6 13 3 RR 10_!i0_30 1 Center Hole Submerged

5 X3MS 98 5 57 0 RR lO-iiO 1 Center Hole Submerged :ro data at cO psi

'later bag r".iptured

6 RSMS 97 7 53 HP lO-Uc-30 3 Center Hole Succeried

7 S3!-«S 9h k 35 2 r3. lO-^C-oO 3 Center Hole Submerged

3 95 5 ^1 3 ?_-< uo - :icne Sub-erged Cb.ecic test for Test '.''Q. 1

9 F3MS 93 3 23 9 ?-?. :rone 3ubt:erg5d ?ained in-o vaoer

10 R2MS 99 9 55 - ?J=. ?.oddei i^O - ZTcte Pained into v=ter Podded

11 93 2 23 2

= 3 -0 - ;::r.e Submerged

12 100 9 7C 5 ?_R ^0 - "cne £uc:nerged —

13 96 0 UU 3 !iO Jione S-bcerged

lU 100 1U. 0( 7 ?B Redded !^0 - "one Subnerg;d Piined into vater Podded

15 ?3VS 93 9 60 2 RR Rodded 1 Ifone Submerged Rained into vater Rodded

16 R3:-!S 92 2 21 9 10 1 N'one Submerged

17 R3M3 101 I; 72 3 ?Pv Tajnped 10 1 None Submerged Manual tariOing vith 1 it square plate

18 101 5 73 5 RF. Tamped 30 1 None Subnerged Manual taoiping

vith 1 ft square plate

19 101 3 72 5 Single Hose Hainer

Uo 1 None Submerged

20 95 k 1*0 3 Circular

Rainer (CR) uo 1 None Submerged

21 P3!'!S 93 7 30 3 CR Uo 1 None Submerged —

22 RH-tS 97 0 1;9 7 CH 80 1 Center Hole Subnerged

23 H5MS 90 5 11 6 CR 30 1 Center Hole Submerged

2U R3MS 101 5 73 3 CR 30 1 Center Hole Submerged

25 Ottava 100 9 53 1 CR 10-UO 1 Center Hole Submerged

26 Ottawa 101 5 56 3 CR 10-Uo 1 Center Hole Submerged

* Pounds per cubic foot (16.02 kg/= )

Pounds per square inch (6.9 V-Ji/:;'). t Reid-3edford 'Model Sa--.d.

VI-150
FIG. 1.—STACKED RING SOIL CONTAIIER

VI-151
S S I "c

u

w o a on I H I

CO I -I K O O M

EH Ej • ?5 O l-l I

<;

Eh10W

EHo

1^oH

EHOw

(/)1I M

VI-152
VI-153
z

UJu

UJ

QCO

(JL. o

CLo

h-$

o

_lUJ

CDI

hQ.U

Q

0

6

I 2

18

24

30 ;•

I

I

I 3 6 :-

42

48

54

60

66

RELATIVE DENSITY , °7o

30 40 50 60 70

72

92

1-1

1-2

1-3

3-1

1

1-3 r\?Si

?. 3-2

3--3 ^

1 4-1

4-2 . •

4-

D

0

0

t.ND DENSiTY 0

n-Ns;TY

. . . . LJ

0

1

94 96 98 100 102 -8 0 8

DRY DENSITY , LB/CU FT

PLACEMENT DENSITY AND DENSITY VARIATION FROM

OF SAMPLED INCREMENTS VS DEPTH PLACED DENSITY

REID-BEDFORD SAND

NOMINAL RELATIVE DENSITY; Dr = 55%

(SPECIMEN 7)

FIG. 6. —TEST RESULTS

VI-154
VI-155
VI-156
LEGEND

O 10 PGI, OCR = I • 10 PSi, OCH = 3 A 40 PSI. OCR = 1 A 10 Pal, OCR = 3 7 UO PSI. OCR = 1 y (10 PSI, OCR - 3 n Oi l AWA SAND OA f A POINTS

NOTE: HIIMQtR tiESIOt DATA POINTS REPRESENIS NUMDER OF DATA POINTS AT SINGLE L UC A I ION.

ITRELATIVE UENSn Y. I'tKctNl

FIG. 10.—COMPARISON OF WES DATA, (;llilt;; ANH llol.T/, ( :oKlil';i A'l'TON CURVES,

AND BAZAHAA CORRL:i,A'L':U,)hl CIlKVl':;;

VI-157
DYNAMIC SOIL PROPERTIES WITH EMPHASIS

ON COMPARISON OF LABORATORY TESTS

AND FIELD MEASUREMENTS

TOSHIO IWASAKI

FUMIO TATSUOKA

Ground Vibration Section

Public Works Research Institute

Ministry of Construction

Japan

ABSTRACT

In order to evaluate analytically the motion of the ground during earthquakes , it

becomes necessary to obtain the dynamic deformation properties of the soil deposits,

expecially the strain amplitude-dependent shear moduli and the damping coefficients. At

two sites, Iruma, Minami-Isu-cho and Ohgi-shima, Kawasaki-shi , insitu seismic surveys were performed. At the former site, a sand embankment was damaged during the Off-Izupeninsula earthquake on May 9, 1974 and at the latter is located reclaimed land where

borehole accelerometers are installed. Furthermore, sand sampled from these sites were tested with the resonant-column apparatus of the Drnevich type in order to obtain

the shear moduli and damping capacities at small strains. Laboratory test results

showed that the two natural sands/ which are well-graded and include fine particles, have

smaller shear moduli than does uniform clean sands such as Toyoura-sand and Ottawa-sand.

A comparison of the shear moduli from shear wave velocities and those from resonant

column tests was performed, giving excellent correlation for sands from both sites.

Key Words: Dynamics ; Field Tests; Laboratory Tests ; Relationships ; Soil Properties.

1. Introduction

In order to analyze the earthquake response of a ground, it becomes necessary to obtain the dynamic deformation properties of the soil for each layer at the site. For engineering application, the earthquake response analyses for horizontally layered sites

have been performed. In these analyses, only the horizontal displacements of the layers

are taken into account and therefore only the density p = (Y^/g) > shear modulus G, and

damping factor T] of each layer is needed. In this paper comparisons of shear moduli G obtained from laboratory tests and those obtained from insitu seismic surveys is presented.

2. Shear Modulus G by Resonant-Column Method

The shear moduli G of sands were obtained by performing dynamic tests using a resonant-column apparatus developed at the Public Works Research Institute. A schematic

diagram of the apparatus is shown in Fig. 1. This resonant-column apparatus is the

Drnevich-type (1) with a hollow cylindrical sample, as shown in Fig. 2. The sample is

VI-158
25 cm in height, 10 cm in outside diameter and 6 cm in inside diameter and is fixed at

the bottom. A pair of oscillators, which supply the system with a torsional vibratory

force, are fastened to the rigid mass on top of the specimen. The confining pressure,

equally applied to the outside and to the hollow of the specimen, is supplied by air

pressure. The axial load can be applied independently of the confining pressure. The

quantities measured during a steady state vibration test are the resonant frequency of the

oscillator-specimen system, the vibration amplitudes at the top of the specimen, the

changes of length and volume in the specimen, the confining pressures and the axial load.

Furthermore, amplitude time decay curves are recorded after shutting off the driving

power at resonance. From these measurements, the shear strain amplitudes, the shear modul

and the logarithmic decrements can be evaluated for the materials examined. Typical test

results are shown in Fig. 10, where the shear moduli in the range of shear strain of 2 x -6 -4

10 to 5 X 10 are presented. Details of the principle and testing procedures have

been presented in previous papers (2,3).

Listed in Table 1 are details of the sands tested in this study. These sands can

be divided into two groups: (1) Clean sands which do not include fine particles smaller

than 0.074 mm in diameter, are poorly graded, and are made by sieving several natural

sands, (2) Natural sands, Irumasand and Ohgi-shima-sand, which are well graded and include

finer particles. Shown in Fig. 2 are gradations of the sands tested, where TO, OS,...

and the abbreviations of the names of sands, listed in Table 1. All of the tests were conducted on air-dry or saturated isotropically consolidated (a /O = 1.0) samples.

—

6

The shear moduli of clean sands, at a single shear strain amplitude (y) of 10 , -5 -4

10 and 10 , are shown in Fig. 4, Fig. 5 and Fig. 6, respectively. The factor p represents the mean principal stress denoted by (O + 2a )/3 where O is the axial stress and

3. IC Si a is the radial stress. These figures show that the shear moduli G of clean sands can

r be represented by the following empirical equations, irrespective of the type of sands;

-6 (2. 17 - e)^ 0. 33 /,^

(7 = 10 ^) G = 900^— P

<1J

2

1 + e

2 2 where G is the shear modulus in kg/cm , p is the mean principal stress in kg/cm and e

is the void ratio. Eq. (3) is identical to the empirical equation for round Ottawa-sand

as proposed by Hardin, et al (4)

.

However, natural sands are generally well graded and to some extent include silty

or clay particles. Test results, for all natural sands, have shown that their shear

moduli are less than those expressed by eqs. (1) through (3) . Therefore, it is apparent

VI-159
that shear moduli of various sands can not be predicted by eqs. (1) through (3)

In the following section, it will be shown that the shear moduli predicted by eqs. (1) through (3) is larger than those measured during in situ seismic surveys. Also, it will be

shown that the shear moduli measured by resonant-column tests, for sand samples taken

from the sites, are almost the same as those obtained during field measurements.

3. Comparison of Laboratory Tests and Field Measurements

Insitu borehole seismic surveys and samplings were conducted at two sites: Iruma,

Minami-Izu-cho and Ohgi-shima, Kawasaki-shi (see Fig. 7) . At the former site, a sandy

embankment was damaged by the Of f-Izu-peninsula earthquake in 1974. The surrounding

slopes of the embankment slipped down and wooden houses on the embankment were severely

damaged. The latter site is reclaimed land, where borehole accelerometers were installed

down to the depth of about 120 m below the ground surface.

At iruma, the borehole seismic surveys were carried out at three locations. The

soil properties are shown in Fig. 8 and Fig. 14, where the soil profile, N-value, wave

velocity Vs and density are presented. The shear wave velocity measurements were performed

at 2 m depth intervals. Shown in Fig. 9 are the relationship between the N values and

the shear moduli G , obtained from the following equation.

where is the shear modulus, is the total unit weight, g is the acceleration of

gravity and vs is the shear wave velocity. Also shown in this figure are three lines,

which represent the empirical relationship between G^ and N-values proposed by three

different groups, Ohsaki, et al. (5), Kajima (6) and Ohyo-chishitsu (7). As seen in Fig.

9 the data is scattered, therefore, empirical relations can not be obtained in predicting

G from N-value. This is due to the fact that the sand deposit of Iruma includes some

Ogravel, which make the N-value much larger in comparison with sands without gravel. It has also been determined that for the Ohgi-shima site, the shear moduli at very small

strain levels can not be predicted by empirical relationships between G^ and the N-value.

Resonant-column tests were performed on sands sampled from the Iruma site. Sands were taken from the deposit just below the ground surface. As the sand deposit is almost

uniform to a depth of about 10 through 20 m, the sand near the ground surface could represent properties of the entire sand deposit. Tests were carried out on dense air-dry and

saturated samples which were completely disturbed. Samples were made by placing air-dry

sand through air into a split mold or by placing saturated sand into a mold filled with

de-aired water. Dens ification was performed by tamping the split mold with a wooden

hammer. Consolidation time was about 1 hr. Considering that the sand deposit is young

and the plasticity index of the sand is almost zero, the effect of cementation and delayed

consolidation is negligible (8). Therefore, the shear modulus of the sand deposit can

be estimated from the test results on completely disturbed specimens. Fig. 10 shows some

O

7t

0

" g

(4)

VI-160
test results on the Iruma sand. Fig. 11 gives the shear moduli at y = 10 / 10 and

-4

10 of the Iruma sand. As seen from Fig. 11, the shear modulus of the Iruma sand does

not depend on wet conditions, and is smaller than those represented by eqs. (1) through

(3) . It should also be noted that the shear moduli of Iruma sand A, Iruma sand C and

Iruma sand Z3, which were made by sieving Original Iruma Sand are identical to those represented by eqs. (1) through (3) as shown in Fig 4 through 6. This means that the shear

moduli of sands are not affected by the properties of the particles but are affected by

the grading or content of fine particles. To obtain the shear moduli of the sand deposit

of the Iruma sand from empirical equations, it is necessary to estimate the effective mean principal stress p' and the void ratio e in the ground. First, the relative density of

every layer 1 m thick, was estimated from the relationship of the effective overburden

pressure O'^ and N-value, (Fig. 12) , as proposed by Gibbs and Holtz (9) . In this estimation extremely high N-values, due to the existence of gravels, was omitted. Second,

both values of the maximum and minimum void ratios were obtained by the method proposed

by Yoshimi and Tohno (10) . In this method, the maximum void ratio is obtained by pouring

air-dry sand into a steel mold with an inner diameter of 6 cm and an inner height of

5.3 cm. The minimum void ratio is obtained by pouring air-dry sand into a mold and by

tamping horizontally the mold with a wooden hammer. Thirdly with the information on saturation rate of sand and the underground water table, the unit weight of sand was

estimated. Then, the effective mean principal stress p' was calculated by using the

values of y , the groundwater table supposing the earthpressure coefficient at rest to be

-6 -4

0.5. Last, the shear moduli at Y

= 10 and y = 10 were obtained by the empirical

equations for the shear moduli of the Iruma sand shown in Fig. 11, using the estimated p'

and e values. Shown in Fig. 13 are the shear moduli, at extreme small strains, obtained

from the shear wave velocities Vs as denoted by solid vertical lines the values of the

-6 -4

shear moduli at Y = 10 and Y = 10 , obtained from the laboratory tests , are represented

-6 -4

by the black circles (G at Y = 10 ) and black triangles (G at Y = 10 ) - The solid

curve given in Fig. 13 indicates that the shear moduli, estimated from these two labora-

-6 -4

tory-test, has values of G at Y

= 10 and Y = 10 , these shear strain amplitudes

occurred in the ground during borehole seismic survey decrease with depth on the order

-7 -5

of about 10 to 10 . Examination of this curve indicates that there is agreement

between the values of the shear moduli predicted by the laboratory tests and those

values obtained by shear wave velocities. Similar results have been obtained for other

location points at Iruma site, as shown in Fig. 15. On the other hand, the shear moduli

obtained from empirical equations for "clean" sands, eqs. (1) through (3), using

estimated values of p' and e are also plotted in Figs. 13 and 15 designated by the white

-6 -4

circles (G at Y = 10 ) and white triangles (G at Y = 10 ) • As seen by Figs. 13 and 15,

the predicted values of eqs. (1) through (3) are much larger than the values from the

shear wave velocities.

A similar survey was also conducted at the Ohgi-shima site, the soil profile of which

is shown in Fig. 16. As the reclaimed sand deposit at this site is also almost uniform

to a depth of about 25 m, resonant- column tests were performed on samples taken jusr

VI-161
below the ground surface, these results were then used to determine the shear moduli in

-6 -5 -4

the ground. The shear moduli at Y = 10 ,10 and 10 are shown in Fig. 17, from which

it can be seen that well-graded Ohgi-shima sand including fine particles, has a smaller

shear moduli than those estimated by eqs. (1) through (3) . It can be seen that the

shear moduli of both air-dry and saturated sands can be represented by the same equations.

Ohgi-shima was reclaimed recently with sands from Sengenyama, located in Chibaprefecture. It should also be noted that Sengenyama sands A,B, and C, which have properties

as shown in Fig. 3 and Table 1, have shear moduli expressed by eqs. (1) through (3) as illustrated in Figs. 4 through 6. This means that the shear moduli of natural sands

are lower than those of clean sands, because the properties of natural sands contain

fine particles. Shown in Fig. 18 is a comparison of the shear modulus estimated from

laboratory tests and those obtained from the shear wave velocity Vs. From this figure

it can be seen that the shear moduli predicted by the laboratory tests are almost identical

to those obtained from field measurements, and that the shear moduli estimated by empirical equations of clean sands are over estimated.

Another important point to be noted is that shear moduli, estimated from resonant

column tests, vary smoothly with respect to depth, but similar variations can not be

obtained by the field seismic survey. Therefore, when conducting earthquake response

analyses of soil deposits, it would be better to determine shear moduli of soil by

considering results from both field seismic surveys and laboratory tests.

4. Conclusions

The paper presented the results of a study in which shear moduli were at very small

strain. The principal conclusions drawn from this study are as follows:

1. With a resonant-column apparatus, the shear moduli of sands were obtained for

-6 -4

the range of strain of Y = 10 to Y = 10 , 2. The shear moduli at each shear strain level of clean sands can be expressed by

an identical empirical equation irrespective of the sand tested,

3. The shear moduli of natural sands, which are well-graded and contain silty and

clay particles, are smaller than those for clean sands.

4. The shear moduli obtained from laboratory tests is almost identical to those

obtained from borehole seismic surveys.

5. The shear moduli in ground, estimated by laboratory tests varies more with

respect to depth, when comparing with the variation of the shear moduli obtained by

seismic surveys.

Acknowledgements

This research was strongly supported by Mr. E. Kurbayashi, Chief of Earthquake

Engineering Section, Public Works Research Institute. Comprehensive tests were conducted

with an assistance of Mr. S. Yoshida, engineer of Ground Vibration Section, Public Works

Research Institute. The authors wish to acknowledge his cooperation.

VI-162
References

1. Drnevich, V.P., Effects of Strain History on the Dynamic Properties of Sand, Ph.D.

Dissertation, University of Michigan, 1967.

2. Kuribayashi, E., T. Iwasaki and F. Tatsuoka, Effects of Stress-Strain Conditions

on Dynamic Properties of Sands, Proc. of JSCE, No. 242, Oct. 1975. pp. 105-114.

3. Kuribayshi, E., T. Iwasaki and F. Tatsuoka, Effects of Particle Characteristics

on Dynamic Deformational Properties of Soils, Proc. of 5th Asian Regional Conf . , Bangalore,

India, Dec. 1975.

4. Hardin, B.O. and W.L. Black, Sand Stiffness Under Various Triaxial Stresses, Jour,

of ASCE, Vol. 92, No. SM2, 1966.

5. Ohsaki, Y. and R. Iwasaki, On Dynamic Shear Moduli and Poissons Ratios of

Soil Deposits, Soils and Foundations, Vol. 13, No. 4, Dec, 1973.

6. Kara, A. et al., Shear Modulus and Shear Strength of Cohesive Soils, Soils and

Foundations, Vol. 14, No. 3, Sept. 1974.

7. Imai, T. , H. Fumoto and K. Yokota, The Relation of Mechanical Properties of Soils

to P- and S-wave Velocities in Japan, Proc. of 4th Japanese Symposium of Earthquake

Engineering, Tokyo, Nov. 1975.

8. Afifi, S.S. and F.E. Richart, Jr., Stress-history Effects on Shear Modulus of

Soils, Soils and Foundations, Vol. 13, No. 1, Mar. 1973.

9. Gibbs, H.J. and W.G. Holtz, Research on Determining the Density of Sand by

Spoon Penetration Test, Proc. of 4th Int. Conf. on Soil Mech. and Found. Eng., Vol. I, 1957.

10. Yoshimi, Y. and I. Tohno, Statistical Significance of the Relative Density, ASTM,

STP, No. 523, 1972.

11. Drnevich, V.P. and F.E. Richart, Dynamic Prestraining of Dry Sand, J. of ASCE,

Vol. 96, No SM2, Mar., 1970.

VI-163
Table 1 List of Materials

Material (^s Dio

(mm) (mm) Uc

V. "^ e °J P

• b)

Toyoura-sand (T0)<^) sub-angular, uniform

1

54 0.12 0.175 1 . 46 0 . 96 0 . 64

Sengenyama-sand A (SA) uniform 2.65 0.85 2.0^^) 0.82 0.61

B (SB) 2.72 0.40 0.85^^) - 0.92 0.67

" C (SC) " 2.68 0.25 0.40^^) 0 .98 0.65

Crushed sand -stone (CS)

angular, uniform

2.73 2.0 'X' 5.0^) 1 .09 0.80

Crushed lime-stone (CL)

sub-angular, uniform

2.74 2.0 ^ 5.0^^) 0.93 0.64

30-50 Ottawa-sand (OS) round, e) uni form

2.66 0.3 0.54 1.8 0.80 0.48

Ohgi-shima-sand (OG)

sub-angular, well-graded

2.69 0.2 0.45 2.25 0.72 0.49

Iruma-sand (IR)

angular, well -graded

2.75 0.28 0.56 2.0 0.84 0.57

Iruma-sand A (IRA)

angular, uniform

2.67 2.0 ^ 5.0d) - 1 .06 0.79

Iruma-sand C (IRC)

angular, uniform

2.83 0.4 0.85^) 0.98 0.66

Iruma-sand Z3 (IRZ3)

angular, uniform

2.79 0.5 0.74 1 .48 0.94 0.68

a) All tests were conducted under 01/03=1.0 - condition,

t") ^max. e^iin. were obtained on air-dry condition.

c) TO, SA, CS, are abbreviation for materials tested.

d) Range of diameter.

e) After Drnevich and Richart (1970)."'*

VI-164
VI-165
axial force

005 0.1 0.2 0.5 I 2 5

Grain Size (mm)

Fig. 3 Gradings of Sands Tested

VI-166
e

2,000

1,000

500

200

100

r--\6r6

0.2

Air- dry

0"|/a"3-- 1.0

10 17-p ) 0.38 G -- 900 ^'

p

I + e

LEGEND

• TO (e --0.69) + CS (e = 0.86)

A SA (e --06 1 ) X XCL (e ^0.61 ) SB (e --0.65 ) IRA (e --0.79) so (e --0.66) V IRC (e --0 68)

0 IRZ3(e = 0.67)

0.5

p (kg/cm^^)

Fig, 4 Shear Moduli at 7 = lo' of Clean Sands

2,000

~p 1,000

500 -

200

00

0.2

r=io"

A i r- d ry cri/a3=i.o j^^^

LEGEND

SC (e = 0 66)

+ C S (e = 0 86)

XCL (e= Q 61)

I RA (e =r 0 7 9)

A SA (e= 0 61) VI RC (e = 0 68)

SB ( e= 0 65) ® I RZ3(e = 0 67)

O 0 S ( e r 0 55) • TO (e = 0 69i

A TO ( e= 0 88)

0.5 I 2

2, P ( kg/cm^)

Fig, 5 Shear Moduli at 7 = 10 of Clean Sands

vi-167
2.000

200 -

10

^1o-4

(2.l7-e)2 0.5 r

LEGEND

Air-dry

(Ji /a3= 10

SC(e^0.66 ) o 0S(e^0.65) +CS(e-0.86) • T0(e^0.69) xCL(e^0.6l ) A TO (e=0.88) IRA (e^0.79)_

A SA(e-0.6l) ^IRC (e-0.68)

o SB (e^ 0.65) ® lRZ3(e-0.67)

.2 0.5 1

p (kg/cm^i

Fig, 6 Shear Moduli at 7 = lo' of Clean Sands

Shizuoka

O Urawa

Tokyo

Kawasaki

Bay of Tokyo

Yokohama ^O^g,.^^^^

Kanagawa

Chiba

0 10 20 30 40 km

I I I I I

Estimated Causative Fault of the

Off-Izu-peninsula earthquake of 1974

Fig, 7 Locations of Iruma and Ohgi-shima

VI-168
Depth

(m)

Soil type

^Gravel P^Silt

EZZZlSand ^^Clay

"1—I r

N- value

Vs ( X 10 m/sec

)

10 20 30 40

~~r- 50 60

4

6

8

10

12

14

16

18

20

22

24

26

28

30

32

Medium

sand

Sand

witti gravel

Medium

sand

Humus

Medium

sand

Sand with

Weathered

tuff

Fig. 8 Soil Profile of Point No, 1 at Iruma

VI- 169
10,000

Fig. 9 Relationship between G from Seismic Survey and N-values

,500

<r 1,000

^ 500

° '—-o . Iruma-sond (Air-dry)

e = 0.64

N> p= 4042

~ o-o—

° p

= 2 033

p

= 1 037

O

"

° —

, P =y ( 0"| + 20"3l

1 1

p

= 0 530

p

= 0 225 kg/cm

o ,

(T, / 0"3 = 1 0

10"^ 2 5 10"" 2 Shear strain amplitude, Y

Fig, 10 Typical Test Result on Iruma Sand

VI-170
(0-1 7(73 ) = 1.0

0.1 0.2 0.5 1 2 5

P = y (0(] + 2(7r) (kg/cm2)

Fig. 11 Shear Moduli at 7 = lo"^, lO'^ and lO"^ of Iruma Sand

VI-171
N-Value

0 10 20 30 40 60 80 100 120

Fig. 12 Relationship between Effective Overburden

Pressure Oy' and N-values

VI-172
G ( kg/cm2 )

1,000 2,000 3.000

LEGEND

Predicted value of G by

experimental equations of clean sands O (y=10-^) G = 900^^n^P°-''

A (^=10-^^) G = 700 '^,'^"1'

Predicted value of G by

experimental equations of Iruma - sand • (y=10-^) G = 580 '^.'I'/' P°^^

r=io

1 +e

-4. 6 = 440^^/^%°"

Fig. 13 Comparison of Shear Modulus by Laboratory Tests and

Field Measurements at Point No. 1 at Iruma

Vl-173
n« Soil type

Depth ^Gravel ^gSilt

(m) EZZZlSand ^^Cloy 10

N- value - Vs ( X 10 m/sec

)

20 30 40 50 60

Density

(t/m^)

4

6

8

10

12

\- 14

16

18

20

22

24

26

28

30

Humus

Weathered

granite

88

188

88

100

100

94

72

88

100

79

7163

750

Fig. 14 Soil Profile of Point No. 3 at Iruma

VI-174
G (kg /cm'

1.000 2,000 3.000

LEGEND

Predicted value of G by

experimental equations of clean sands

O (y-IO-^) G-900^^f^%°^^ + e A ()f = 10-4) G-700 (2.17-e)^

1 + e

0.5

Predicted value of G by

experimental equations of Iruma-sand

» (y-10-6) 0 = 580-^^1^^%°^^

(2.17-e) A [Y^IO-^] G = 440

I + e

P

0.52

Fig, 15 Comparison of Shear Moduli by Laboratory Tests and

Field Measurements at Point No. 3 at Iruma

VI-175
Fig. 16 Soil Profile of Ohgi-shima

VI-176
<MOJ

1 CD

cvj

cr>

CL)

J

csi

CD

2.000

1.000

500

200

100

2,000

1,000

500

200

100

2.000

E 1,000

CP

OJ

1 500

CO

CD

200

100

o Air -dry e=052

• Saturated e = 0.46

(Oi /a3 =1.0)

Saturated e = 046 -

(cr, /a-3 =1.0)

(a-, 7(73=1.0)

0.1 0.2 0.5 1 2

P = y (Oa + 2(7r) (kg/cmM

Fig. 17 Shear Moduli at T = lo"^, 10~^ and 10~^ of Ohgi-shima Sand

Vl-177
0

0

G (kg/cm^

1.000

10

20

m

30

2.000

A O

^ f Vl

1 g = 980 cm/sec^ a\ • A o

A Y

A o

A o

A X o

\ A O

A \ A O

A

A V

^ o

Predicted value of G ^

k A O

by Resonant Column Test i A o

A J A o

LEGEND Predicted value of G by

experimental equations of clean sands

(2.17-e)^ .0.38 o (r=10-^) G = 900^^fVf^P

A (r = 10-^) G = 700^^/^P°'

Predicted value of G by

experimental equations of Ohgi-stiima-sand • (r=10"^) G = 620-^^r^^P°'^^

(r^lO""^) G = 500

iH-e

(2.17- nO,545

I ^-e

Fig. 18 Comparison of Shear Modulus by Laboratory Tests and

Field Measurements at Ohgi-shima

VI-178
DESIGN EARTHQUAKES

E.L. KRINITZSKY and FRANK K. CHANG

Corps of Engineers

Waterways Experiment Station

Vicksburg, Mississippi 39180

ABSTRACT

Relationships between earthquake intensity and epicentral distance are presented. Peak, motion results, from 187 field records, are also given for examination.

Also presented is data on displacements, velocities and intensities.

Key Words: Data; Displacements; Intensity; Seismic Records; Velocity.

1 . Summary

a) A design earthquake is the bedrock ground motion which is specified for a dynamic

analysis. It is based on motions assigned to the maximum credible earthquake (the largest

that is reasonably expected) and attenuated to the site. Given several sources for maxi- mum credible earthquakes, there are several design earthquakes.

b) The size of a maximum credible earthquake and the area in which it occurs is

determined by geological and seismological studies. Attenuation from source to site

is based on seismological interpretation.

c) Since most earthquake data is available only as intensity, it is necessary to specify peak motions in terms of intensity. Fig. 1 relates magnitude to Modified

Mercalli (MM) intensity and to epicentral distance so that instrumentally recorded data

can be related to areas in which there are only observations of intensity.

d) The dispersion of values for acceleration, velocity, and displacement for any

given level of intensity is so large that average or mean lines are meaningless for the

purposes of engineering design. For any important project, peak motions are needed.

Lesser levels of shaking are appropriate if they are accepted in terms of lower levels

of safety.

e) Peak motions, particularly for acceleration, differ greatly in the near to far

field for the same level of intensity. Near field limits for different magnitudes and

epicentral intensities of earthquakes are shown in Table 1. Accelerations, based on 187

strong-motion records uniformly processed at the California Institute of Technology,

are shown in Figures 2 and 3 for near and far fields, respectively. Ground conditions,

whether soil or rock, are not specified as large overlapping variations were found among

them. Figs. 4 and 5 show corresponding curves for velocity and Figs. 6 and 7 for

displacement. The peak curves are reduced by 10 percent increments to a 50 percent level

of the spread of data. Thus, these curves may be used to select appropriate parameters

for peak or lesser motions. The lesser motions can be related to cost-risk analyses based

VI-179
on calculated recurrence rates of earthquakes.

f) These parameters may thus be used to rescale appropriate earthquake records,

those taken from comparable geological settings, for comparable types of faults and re- corded at similar distance from the causative faults. Also, synthetic records can be

rescaled using these values.

g) Durations must also be specified. Figs. 8 and 9 show upper boundaries for duration

in terms of intensity for the near and far fields, respectively. The boundaries are shown separately for soil and rock. Significantly, durations for soil are double those

for rock. Fig. 10 shows the relation of peak values for a magnitude 6.5 earthquake

(San Fernando) in terms of duration and epicentral distance. General statements of

magnitude versus epicentral distance and duration for rock and for soil are shown on

Figs. 11 and 12, respectively. These values may be used to rescale the time functions

for the earthquake records that are selected.

h) Because of uncertainties in the appropriateness for rescaling of any single

earthquake record, several records should be used.

VI-180
Table 1 Limits of the Near Field,

Western United Stat e

s

Radius of

Maximum Near Field

Magnitude Intensity (km)

5.0 VI 5

5.5 VII 15

6.0 VIII 25

6.5 IX 35

7.0 X ko

7.5 XI 1+5

Table 1. Boundaries for field conditions.

500

n m nc z 31 3ni zna. ic i n

MM INTENSITY

Figure 1. Relation of magnitude tu MM intensity and epicentral

distance

.

VI-181
V1-18
0.4

0.3

0)

z

o

i 0.2

_ltuo

o

<

0.1

FAR FIELD

^^^^^

• •

+i

^\• J

•

•

: m 3ZI

MM

3m inn

Figure 3. Acceleration values for MM intensity in the far

field.

VI-183
J

/

/

// /

0 /

/ -/

/

/ —

/

/

/ /

<0// ^/

/

/ —/

NEAR FIELD /// / /

/

/

// / /

/

^—

/

/

^—

/ / /

/ /

/

do

///

L/U-

//

/: r

Figure U. Velocity values for >Eyi intensity in the near field.

V 1-184

VI- 186
RICHTER MAGNITUDE

4.5 5.0 5.5 6.0 6.5 7.0 7.5 S.O 8.5

100

SO

60

50

40

30

20

ID m

O

d

A\O

o

<

o

UJtn

z

o

<

d:3O

10

NEAR FIELD

i

1.0

0.8

0.6

0.5

0.4

0.3

LEGEND

INSTR COMPONENT

0 HORIZONTAL • VERTICAL HORIZONTAL

VERTICAL HORIZONTAL

VERTICAL

SITE CONDITION

SOFT

^ITERMEDIATE

HARD ROCK

0.2

3ZL 3ni •yrn ix x

MM INTENSITY

n xir

Figure 8. Durations for MM intensity in the near x^ield.

VI-187
UPPER BOUND DURATION

FOR SOILFAR FIELD

UPPER BOUND DURATION

FOR HARD ROCK

e

£0

LEGEND

IN5TR COMPONENT

o HORIZONTAL

t> VERTICAL

HORIZONTAL

VERTICAL

HORIZONTAL

VERTICAL

SITE CGNCr^ON

SOFT

iNTERMEDlATE

HARD RC:K

J i L i

MM INTENSITY

XL xn

Figure 9- Duratior.G for ;-C'I intensity in the far field.

V1-188
VI-189
VI-190
VI-191
RELATION BETWEEN SEISMIC COEFFICIENT

AND GROUND ACCELERATION FOR GRAVITY QUAYWALL

SATOSHI HAYASHI

Head, Structures Division

Port and Harbour Research Institute

Ministry of Transport

SETSUO NODA

Chief, Subaqueous Tunnels and Pipe Lines Laboratory

Structures Division

Port and Harbour Research Institute

Ministry of Transport

TATSUO UWABE

Earthquake Resistant Structures Laboratory

Structures Division

Port and Harbour Research Institute

Ministry of Transport

ABSTRACT

The present design standard for port and harbor structures, utilizes seismic coefficients which were obtained from records of 129 gravity quaywalls in 49 ports damaged by

12 earthquakes. The maximum ground accelerations in the ports were estimated by calculating the ground response during the earthquake with reference to the attenuation curves

of the base rock acceleration based on the accelerograms in port area. The seismic coefficients in past earthquakes had upper values of 0.25, and this

upper limit can be related between the coefficient and the maximum ground acceleration by

the following equation.

\

3 ' s'

where eA

a

g

seismic coefficient

maximum ground acceleration (gal)

acceleration of gravity (gal)

Key Words: Acceleration; Accelerogram; Gravity Quaywall; Rock Motion; Seismic Coefficient.

1. Introduction

The present earthquake resistant design for the port and harbor structures is

based on the static theory, so called the seismic coefficient method (1)

The seismic effect on a structure varies with and the stability of the structure will

depend on such factors as the predominant period, amplitude and duration of the earthquake

motion. Therefore, dynamic design can only become an effective method for the port facilities through the use of computer s . However , the seismic coefficient method is still an accepted procedure for common structures, because this method is based on experiments and

VI-192
requires simple calculations for a stability analysis.

Many authors have made an attempt to improve the reliance of the seismic coefficient method. The gravity quaywalls, damaged by earthquakes, were analyzed by the present design procedure, and the seismic coefficients corresponding to the severity of

the seismic effect were obtained. The maximum ground accelerations in the ports were also

estimated by calculating the ground response during the earthquake. These results were

investigated from the aforementioned point of view.

2. Seismic Coefficient

A list of earthquakes and the damaged ports discussed in this paper are shown in

Table 1. The locations of the ports are indicated in Fig. 1. Examples of the damaged

quaywalls are shown in Figs. 2 through 4. In Fig. 5* a flowchard of the estimated

seismic coefficients (represented by e^ in this paper) is explained. As it is seen in

this figure, the stability against sliding, overturning and the bearing capacity at the base of a structure for each port facility are analyzed. The seismic coefficients,

for a safety factor equal to 1, has been obtained. Judging from the characteristics of

damage on quaywalls, the cause of damage has been estimated and the ranges of the seismic

coefficients were deduced for each facility. Finally, only one seismic coefficient or the range of the coefficient, in the particular port, was determined from these ranges (2)

.

Although there are various problems in estimating seismic coefficient e^, use of this

coefficient was consistent with results obtained from the harbor structures due to the

1968-Tokachi-oki Earthquake and 1973-Nemuro-hanto-oki Earthquake (3). Therefore, in

this report the estimation of the seismic coefficient was based on the procedure as shown in Fig. 5.

3. Base Rock Motion

The rock motion at any particular site will depend in large measure on the magnitude

of the earthquake and the distance of the site from the zone of energy release. In general, the amplitude of motion decrease with increasing distance from the zone of

energy release. Methods of determining the maximum accelerations in rock have been proposed by several investigators. As shown in Fig. 6, the site distance differs for each

I investigator. The distance from the zone of energy release, causative fault or the

epicenter of the principal shock may be more reasonable than the epicentral or hypocentral

distance. But it is difficult to determine this distance from the past earthquakes

because the developing process of the source region for a great earthquake is very

complicated. Considering these matters, Mr. Katsumata has proposed the distance from

the edge of source region (the effective distance) which was taken as a sphere of radius

r. The following empirical formula is therefore introduced in this paper (4)

.

logior - 0.5 M - 2.25

*Please note that Fig. 5 appears after Fig. 6. VI-193
where, r: radius of source region (km)

M: magnitude of earthquake

As discussed by Mr. Katsumata, the ground acceleration of the average ground condition in Japan, is related to the base rock acceleration, the magnitude of the earthquake

and the effective distance using strong motion accelerograms.

Source Region: The fault parameters of some past earthquakes were determined by

seismologists. In this study, it was assumed that the source region was equal to the

fault plane. When the fault parameters are unknown, the region is taken as a sphere

which is equal to the aftershock area. If the aftershock area is unknown as in the case

a small earthquake, the region is calculated by the aforementioned equation. According

to reference (4) , it is assumed that the maximum base rock acceleration at an edge of thi source region is 400 gals, regardless of the magnitude of the earthquake.

Strong Motion Accelerogram: The Earthquake Resistant Structures Laboratory of the

Port and Harbour Research Institute has been observing strong motion earthquakes in

port areas of Japan for 14 years. The observation network consists of 65 stations at

44 ports, and at the end of 1975 more than 1058 accelerograms have been obtained (5).

106 accelerograms from 21 stations were selected based on the condition that the maximum

acceleration exceeded 20 gals and the magnitude of the earthquake is known.

Calculation of Base Rock Motion: In order to develop the attenuation curves, the

base rock motions were calculated by the method based on the multiple reflection theory.

The computer program SHAKE developed by Dr. Schnabel and et al. was used, in which the

shear modulus and the damping of soil changed with the strain level. The source regions

were determined from past earthquakes in which the accelerograms were obtained. For

example ,the fault plane and the aftershock area during the 1968-Tokachi-oki earthquake

is as shown in Fig. 7. The relations between the effective distances that were determined from the source region and the base rock accelerations that were calculated, by

the aforementioned method, are shown in Fig. 8. Using the attenuation curves, the base rock accelerations in the damaged ports were

obtained. The source regions of the earthquake, relative to damage of the facilities, were determined in the same manner as described above. For example> the fault plane and

the aftershock area during the Nankai earthquake are shown in Fig. 9. As shown in this

figure, the fault plane agrees with the aftershock area. Additionally, a 63 km radius

of source region, for M = 8.1, is given by the equation previously described.

4. Ground Acceleration

The maximum ground acceleration at the ports damaged by earthquakes were estimated

by calculating the ground response during the earthquake. The incident wave used as an input for the ground response calculation was obtained from the accelerograms at the

nearest station to the damaged port. The ground motions were calculated by the aforementioned computer program.

VI-194
5. Result and Discussion

The severity of ground motion during past earthquakes was estimated, in the form of

a seismic coefficient, on the basis of the stability analysis on the gravity quaywalls by

current design procedure. Using the attenuation curves of the base rock acceleration, the

maximum ground acceleration was also estimated by calculating the earthquake response of

the subsoil layer. The results are shown in Fig. 10. The expression | indicates that

the seismic coefficient exists within the range, and or + means the upper or lower limit

of the range.

According to Fig. 10, 0.25 represents the maximum seismic coefficient during the

past great earthquakes . This fact may be very informative in the re-examination on the

aseismicity of the port facilities.

As shown in Fig. 10, the upper limit of the relation between the seismic coefficient

and the maximum ground acceleration was determined as shown by the solid line. Consequently it can be concluded that the gravity quaywall, for which the design seismic coefficient exists over the line, may have sufficient aseismicity against an earthquake with

the corresponding maximum ground acceleration.

In this figure the seismic coefficient is not proportioned to the ground acceleration.

If the seismic force acting on the structure is considered to be equivalent to the product

of the mass of the structure and maximum ground acceleration a, the coefficient must be

equal to a/g. Experimental studies on the model, using a shaking table, indicates that

the earthpressure under the sinusoidal excitation up to 500 gals is the same as the results

from the current design procedure and the coefficient is in proportion to the acceleration. Therefore, this fact does not explain the solid line curve shown in Fig. 10. Although the gravity quaywall is generally a simple earth structure, its behavior

during the earthquake is not clear. It is supposed from the results that one of the

major reasons the solid line curve exists is the dynamic response effect of the wall

and the backfill, this suggests that seismic force as the earthpressure is not always

increased proportionally with an increase in the ground acceleration during an earthquake.

6. Conclusion

Judging from the stability analysis of gravity quaywalls, by the current design standard, 0.25 is the maximum value of the seismic coefficient e^ corresponding to the severity

of the ground motions in the ports during the past earthquakes.

The maxJ-JRum ground accelerations a in the ports was estimated by calculating the

ground response during the earthquakes with reference to the attenuation curves of the

base rock accelerations based on the accelerograms in the port area. The upper limit of the relation between e^ and a is expressed by the following

equation

:

1^ eA -

3

( 7 )

VI-195
where e,A

a

s

seismic coefficient

maximum ground acceleration (gal)

gravitational acceleration (gal)

Reference

1. Design Manual of Harbour Structures in Japan, The Japan Port and Harbour

Association, April 1967.

2. S. Noda and T. Uwabe; Seismic Disasters of Gravity Quaywalls, Technical Note of

the Port and Harbour Research Institute, No. 227, 1975.

3. I. Mitsuhashi and T. Nakayama; Analysis of the damage to Harbour Structures by

the 197 3 Nemuro-Hanto-oki Earthquake, the same as the above. No. 184, 1974.

4. M. Katsumata; Note on Maximum Amplitude of Acceleration of Earthquake Motion,

Quarterly Journal of Seismology, Vol. 37, No. 3, 1972.

5. E. Kurata, S. lai and H. Tsuchida; Annual Report of Strong-motion Earthquake

Records in Japanese Ports (1975) , Technical Note of Port and Harbour Research Institute,

No. 236, 1975.

6. P.B. Schnable, J. Lysmer and H.B. Seen; SHAKE-A Computer Program for Earthquake

Response Analysis of Horizontally Layered Sites, Report No. EERC 72-12, Col. of Eng.,

Univ. of Calif. Berkeley, 1972.

VI-196
VI-197

Investigation of

Earthquake Resistance of Structural (Shear) Wall Buildings

Carried out at Portland Cement Association

Reported by

Mark Fintel

Portland Cement Association, Skokie, Illinois

In July of 1974, PCA started (with NSF-RANN sponsorship) a comprehensive analytical and

experimental study to investigate the response to earthquake excitation of reinforced concrete

structures containing structural (shear) walls. The aim of the study is to develop design

procedures and reinforcing details for earthquake resistant multistory reinforced concrete

structures containing structural walls.

ANALYTICAL INVESTIGATION

The objective of the analytical study is to investigate (through inelastic dynamic response)

the effects of the following parameters:

1 . Fundamental period of vibration

VI-199
2.- Yield level

3. Post yield stiffness

4. Viscous tamping

5. Character of moment- rotation hysteretic loop

6. Stiffness table

7. Strength table

8. Degree of fixity at base

9. Number of stories or height of building

10. Ground motion intensity

11 . Ground motion frequency characteristics

12. Duration of ground motion

In addition to the major effort of undertaking the dynamic analysis associated with the parametric

studies, preliminary studies of input motion data were undertaken to determine a suitable

basis for normalizing accelerograms with respect to intensity, and also to characterize accelerograms approximately in terms of their frequency content. Also, an initial effort was made

to carry out a parametric study of typical structural wall sections under combined flexure

and axial load.

The analytical work of the initial 18 months has been concentrated on the evaluation of the

results of the dynamic analysis, as well as the experimental data, with the aim of (1) arriving

at guidelines for the design of earthquake-resistant structural walls, with particular reference

to force (moment and shear) levels and deformation demands (ductility and energy dissipation)

which may reasonably be expected under particular ranges of values of both structural and

ground motion parameters. This information, together with data obtained from laboratory

tests, will serve as the basis for the design procedure to be developed as the ultimate result

of the project; (2) synthesizing a "representative" loading history which can be used in the

laboratory testing of isolated wall specimens on the basis of the results of the dynamic analysis;

VI-200
and (3) studying the measured force-deformation curves from the experimental program for

the purpose of determining the influence of shear and examining the range of values of the

forces (moment and shear) as well as the specimen properties when the primary (monotonic)

curve can be adequately predicted on the basis of flexural considerations alone.

A brief discussion of the work carried out during the last six months, and some of the results

obtained is given below:

(A) Force levels and deformation demands corresponding to various ranges

of major structural and ground motion parameters

To determine the range of forces (mainly moments and shears) as well as the deformation

requirements which can be reasonably expected for practical cases, a number of building

configurations using structural walls which could properly be classified as isolated walls

were studied. Sample configurations are shown in Fig. 1 . The variations in plan shown

in Fig. 1 reflect an effort to vary the fundamental period of vibration of the structure by varying

the stiffness-to-mass relationship. Two types of wall sections were considered, i .e. , the

plane rectangular section and a flanged section.

Tests have indicated that the behavior (i.e., stiffness, strength and ductility) of walls subjected

to reversed cycles of inelastic deformation can be significantly affected by the magnitude

of the shear stresses in the critical (hinging) region. For this reason, particular attention

was given to the levels of shear stresses which can be expected in practical cases. This

information will also serve to guide the experimental program in the choice of shear stress

levels to use in tests.

\

'-1

On the basis of the curves shown in Fig. 2, and the assumed plan configurations shown in

Fig. 1, the variation of the maximum shear stress with the width (i.e., the horizontal dimension

VI-201
in the plane of the wall) of the wall, S , was determined for particular values off and p w c corresponding to the different plan configurations considered. A typical plot of the results

is shown in Fig, 3. Also indicated in Fig. 3 are the corresponding rotational ductility requirements

and interstory displacements.

Among others. Fig. 3 indicates that for a particular plan configuration or fundamental period

there is an optimum width of wall with respect to the magnitude of the shear stress below which

a rapid increase in expected shear stress occurs with a decrease in the width of the wall

for the same thickness.

The information typified by Fig. 3, together with the results of the evaluation of the critical

shear/moment ratios expected in isolated walls, as obtained from graphs such as shown in

Fig. 4 and the associated rotational ductility requirements at the bases of walls as indicated

by analyses, will provide the basis for recommendations of force levels and deformation requirements to guide the design of structural walls. When considered together with the experimental

data on strength and deformation capacity of walls of typical sections, this information will

form the basis for the rational design of isolated structural walls for earthquake resistance.

(B) Loading history study

The major aim in undertaking this study is to determine a representative range of earthquake

induced deformations in terms of the amplitude of the maximum response and the number of

cycles of significant amplitude which can be expected in isolated walls under a range of structure

and ground motion parameter combinations. The study was motivated by the belief that the

behavior of laboratory specimens, or actual structures for that matter, can be influenced

significantly by the type of loading (magnitude, number of cycles and sequence) imposed

on the structure.

VI-202
Most, if not all, of the quasi-static reversed loading tests conducted to date liave used more

or less arbitrary loading sequences. The type of loading generally used in laboratory tests

is one characterized by progressively increasing amplitudes of load, sometimes with smaller

intermediate load cycles used just before an increment in the maximum load is applied.

Figures 5(a) , (b) and (c) , for example, show the histories of rotation of a section at the first

floor level (indicative of the total rotation occurring in the first story) of the lumped-mass

model shown in Fig. 6. The ordinates in Fig. 5 have been normalized in each case with respect

to the corresponding yield rotation.

In trying to characterize the response history, as depicted in Fig. 5, the maximum amplitude

as well as the number of "fully reversed cycles", "partially reversed cycles", and the total

number of inelastic cycles of significant amplitude, were noted in each of the cases covered

in the parametric study. Except for two cases in which a 20-second duration of the input

motion were used, all of the cases in the parametric study used 10 seconds of input motion.

A major observation which can be made in relation to Fig. 7 is that for a wide range of values

of structural and ground motion parameters, the number of "fully reversed inelastic cycles"

of response (or loading) is not likely to exceed 5 or 6, under a 10-second duration of strong

ground motion. Of significant interest is the fact that in many cases the maximum amplitude

of response occurs rather early with no significant inelastic cycles preceding it.

EXPERIMENTAL INVESTIGATION

The experimental program is divided into three parts. In Part I, reversing loads are applied

to isolated walls. In Part II, reversing loads are applied to wall-systems. In Part III, elements

of systems are to be tested. Part III is primarily an investigation of the behavior of confined

concrete

.

VI-203
Highlighrts of the program to date

Part I - Isolated Walls . The isolated walls represent an element of a structural wall system.

They are being tested to determine their strength, ductility, and energy dissipation capacity.

The test specimens are approximately 1/3-scale models of actual wails. The model walls are

15-ft. high and have a horizontal length of 6 ft. 3 in. The web thicknesses are 4 in. All

test specimens are subjected to in-plane horizontal reversing loads. The loads are applied

alternately on one side and then on the other. A specimen and the testing apparatus are

shown in Fig. 8.

Controlled variables covered in the program to date include the shape of the wall cross section,

the amount of main fiexural reinforcement, and the amount of hoop reinforcement around the

main fiexural reinforcement.

A summary of results of the entire isolated wall test series through March 1976 is shown in

Table 1

Part II - Systems . Systems containing connecting elements with various proportions will be

constructed and tested at reduced scale. The systems specimens are being designed with the

intent of providing proof tests of structures having proportions and details representative of

typical structures. The process of selecting representative elements for the first systems

specimen is in progress.

Plans have been made to cast the specimens vertically with construction joints at each floor

level. Lateral loads will be introduced into the systems through stubs representing floor

VI-204
slabs. The walls will be constructed with flanges to increase the lateral stability of the systems

An individual wall of the system will be tested prior to the test of the coupled wall system.

Part III - Elements . Tests have been performed on specimens representing the compression

zones of structural walls. These tests are being performed to evaluate the effect of confinement reinforcement and to determine the effective stress-strain curve of confined concrete.

Figure 9 shows a specimen being tested. The controlled variables in the test program include

spacing and size of the confinement reinforcement, concrete strength, amount of longitudinal

reinforcement, and size of test specimen.

Fourteen of the 17 tests planned have been completed. Table 2 summarizes the entire program.

In addition, a series of tests of coupling beam elements for coupled walls is being carried

out. To date, three specimens have been tested to failure under slow load reversals. All

coupling beams had a cross section of 4 x 6-2/3 in. and an overall length of 16-2/3 in., giving

an aspect ratio of 2 . 5

.

VI-205
TABLE 1 TEST RESULTS

Speci- men

Reinforcement ^ 1) Cone. ACI Design Strength Observed Loads

D

{%)

0

{%)

0^n

{%) {%)

f

c (psi)

Flexural Shear Yield Maximum

kips /FT

c kips

c kips

c kipsc Fl 3.89 0.71 0.30 5575 11+5 8.1 ll+O 7.8 150.6 8.I1 187.9 10.5

31 1.11 0.31 0.29 7685 1;6 2.2 82- 3.9 1+5.1 2.1 61.0 2.9

B2 3.67 0.63 0.29 7775 129 6.1 127 6.0 119.7 5.7 152.8 7.2

Rl l.U? 0.31 0,25 61*90 18 0.9 82* k.2 20.1 1.0 26.6 1.1+ B3 1.11 0.31 0.25 1.28 6860 hG 2.3 82 -x- h.l ii5.2 2.3 62.0 3.1

R2 i+.OO 0.31 0.25 2.07 67)40 35 1.8 82* h.2 33.2 1.7 1+8.7 2.5

BI.(2) 1.11 0.31 0.25 1.28 6530 2.1; 82* U.2 1+5.3 2.3 75.3 3.9

B5 3.67 0.63 0.29- 1.35 6570 129 6.6 127 6.5 112.2 5.8 171.3 8.8

B5R^3) 3.67 0.63 0.29 1.35 6170 129 6.8 127 6.7 167.8 8.9

(1) = Ratio of main flexural reinforcement area to gross concrete area of

boundary element

p = Ratio of horizontal shear reinforcement area to gross concrete area

of a vertical section of vail veb

p = Ratio of vertical web reinforcement area to the gross concrete area

n of a horizontal section of the wall web

p^ = Ratio of effective volume of confinement reinforcement to the volume

^ of core in accordance with Eq. (A-1+) of ACI 318-71

(2) Monotonic Loading

(3) Repaired Specimen

Shear reinforcement governed by maximum bar spacing requirements

VI-206
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VI-207
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VI-208
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VI-218
VI- 219
A PHILOSOPHY FOR STRUCTURAL INTEGRITY

OF LARGE PANEL BUILDINGS

MARK FINTEL

Director

Engineering Services Department

Portland Cement Association

Skokie, Illinois

DONALD M. SCHULTZ

Senior Structural Engineer

Engineering Services Department

Portland Cement Association

Skokie, Illinois

ABSTRACT

The paper reviews the various methods to reduce risk from abnormal loads. To limit

the occurrence of progressive collapse in large panel residential structures, a philosophy for establishing General Structural Integrity is developed to assure bridging of local

damage while maintaining overall stability, thus eliminating the need to design for any

particular abnormal load. In this approach, tensile continuity and ductility of the elements, and their connections as well as of the overall structure, is stressed. The

rationale for a minimum tie system consisting of transversal, longitudinal, vertical and

peripheral ties to establish this General Structural Integrity is developed.

The objective of this approach is not to afford absolute safety in regard to any

exceptional event in any part of every building; rather, the intention is to limit and

substantially reduce the general risk of collapse, as compared to that existing if no such measures were taken.

Key Words: Buildings; Collapse; Concrete; Panels; Structural Integrity.

1. Introduction

When a local failure is not confined to the area of initial distress, but spreads

either horizontally or vertically through the structure, it is termed a progressive collapse. More accurately, it may be defined as a phenomenon in which the spread of a local

failure, eventually results in the collapse of a whole building, or a disproportionately

large part of it. Abnormal loading conditions and effects which can be structurally

significant (1-3) include:

- service system (gas) explosions,

- explosive bombings

,

- external explosions from accidents involving transportation of hazardous materials,

- ground vehicular collisions with buildings,

- aircraft collisions with buildings,

V 1-2 20
- tornados

,

- flooding,

- foundation settlements , and

- errors in design and construction.

Traditionally, it was assumed by the profession that if one followed established

practice and codes, the resulting building would automatically have a degree of structural

integrity which would assure an alternate path. for the load when a load-carrying element failed (Fig. 1) . This belief was justified for some traditional forms of construction which have continuity and ductility, however, it was not valid for every form of

construction (4-12)

.

In 1968, attention of the engineering community was focused on the problem of lack

of structural integrity in certain precast structures in a tragic way. In May of that

year a progressive collapse of an apartment building occurred at Ronan Point in London

(Figs. 2, 3) . A gas explosion in an apartment on the 18th floor caused an exterior panel

to be blown out (Fig. 4a) ; this initiated a progressive collapse upwards to the roof,

and then almost down to the ground, as debris fell on succeeding floors (Fig. 4b) . In terms of hazard, the initial damage was of minor consequence. The progressive collapse

which occurred, however, was the result of the inability of the structure to bridge over

the local failure, that is, its lack of integrity. Therefore, it is not merely the

hazard of abnormal loading it is also the susceptibility of a structure to progressive

collapse which presents the real risk.

2. Reducing the Risk of Progressive Collapse

Three alternative approaches, widely varying in concept, can be employed to reduce the

risk of progressive collapse (13)

:

1. Eliminate the hazards which cause local failures (e.g., elimination of gas

installations in multistory buildings, as has been done in France);

2. Design the structure so that the hazard does not cause any local failure (e.g.

5 psi gas explosion load requirement initiated in Great Britain)

;

3. Allow the local failure to occur but design the structure so that progressive

collapse does not occur (ensure an alternate path for the load)

.

a) Elimination of Hazards . Regarding the first approach, it can be assumed that

certain abnormal loadings could actually be eliminated: gas installations could be disallowed in large panel buildings; barriers could make vehicular collision avoidable; and

construction in flood zones could be regulated. However, with the above exceptions,

abnormal loads can hardly be eliminated, and also most of them defy quantitative determination. Since any realistic solution must deal with all abnormal loading conditions to

some extent, this method of eliminating the hazards cannot be deemed an overall solution.

b) Local Resistance . In the second approach, providing local resistance, explicit

loading requirements are established in an effort to provide sufficient strength to re- sist any local failure. In lieu of a probabilistic approach, it has been suggested that

a minimum design load of 5 psi for gas explosions be used. However, the approach of

VI-221
providing resistance to one specific abnormal load has been challenged on a number of

grounds. It is not reasonable to assume that adequate resistance for one abnormal loading

condition will necessarily provide sufficient strength to resist all other abnormal loading

conditions. From an economic viewpoint, additional costs to provide a certain safety

against local failure may not be justifiable in light of the small number of cases where

progressive collapse of large panel structures has occurred

c) Alternate Path through Structural Integrity . Since the nature and magnitude of

most abnormal loads are unpredictable, the third approach of allowing local failure to occur but providing an alternate path within the structure to avoid an overall collapse

appears to be a sound concept. In this approach, a local failure is accepted as an inevitable consequence.

When a structure has the ability to bridge over local failure it has what will be

termed in this article General Structural Integrity (GSI) , the principal elements of

which are continuity and ductility of members and connections and of the structure as a whole

The General Structural Integrity approach can be implemented in two dista.nce manners

the design engineer can be required to apply a rational procedure to establish the necessary integrity; or, alternatively, code writers can develop the necessary minimum detailing

practice (as has been done for other forms of construction) to establish a degree of con- tinuity and ductility. In either case it is necessary to identify the special sensitivity

to abnormal loadings that the particular structural form possesses (4-10)

Minimum detailing as code requirements is recommended as the most viable alternate.*

Although it is sometimes more complex, and by its nature, more general, it has a number of

advantages

1. Code writers and researchers have a general responsibility to evaluate details

to assure the safety of structures.

2. Design engineers should not be required to directly consider the effects of

abnormal loads for one form of construction and not for another.

3. Experience in other building forms has shown that minimum detailing requirements

based on good engineering judgment can establish an adequate degree of structural integrity.

While, by its very nature, LP construction does not lend itself easily to a design

with moment continuity, it is possible to develop force continuity and ductility of the

connections as well as of the overall system through minimum details specified as recom- mended practice. Since such minimum requirements cannot possibly encompass all situations,

substantial reduction in the risk of progressive collapse in all future designs could be

achieved if the detailing requirements are supplemented with an educational program for

the engineering profession.

I

*Most recently this approach has been adopted by the PCI Committee on Precast Bearing Wall

Buildings (Ref . 14)

VI-222
3. Prototype Structure

The cross wall and spine wall type of large panel residential structures (Fig. 5) have

been selected as the representative prototypes for this study since they are the most

popular types used in the United States. The reasons for their popularity are the

economic advantages of a low wall-to-slab ratio, resulting from the comparatively large

slab spans of up to 42 feet utilized until now. The large distance between the concrete

bearing walls permits an open, flexible architectural layout using light partitions. An added advantage of the cross wall type is the architectural freedom in handling the longitudinal elevations. With individual variety for each building made possible, the uniformity inherent in industrial production can be avoided.

4. Rationale for General Structural Integrity

Connections between the precast elements of LP structures are recognized as the weak

link in this structural system. This weakness is attributable to the lack of continuity

through the connections, and to their brittle nature. Friction- type connections can theoretically provide sufficient stability for gravity loads within an LP structure;

however, such connections offer little resistance to the effects of abnormal loadings.

Recent studies and discussions (7-12, 15-25) on the performance of structures under normal

and abnormal loads have concluded that the integrity of a structure , regardless of the

type of construction, depends on its connection characteristics; and that the strength of

the elements cannot be utilized if the strength of the connections is insufficient. Therefore, as stated earlier, a degree of continuity across the connections and ductility

within the connections should be ensured to achieve General Structural Integrity.

Continuity is essential to develop the bridging capabilities needed for transmission

and redistribution of loading (alternate paths) . Ductility is necessary, not only to sustain the deformations that may be associated with conditions of partial stability

(stability of the structure in the damaged state) , but also to establish some measure

of energy absorption under the dynamic effects of both normal and abnormal conditions.

The magnitude of required continuity can be determined by assessing the forces

acting within the connections under various conditions of local failure. Because of the

nature of LP structures, these forces occur only as compression, tension, and shear.*

Adequate compression and shear capacity in LP connections will usually be provided by

design under normal loads. Little tensile capacity, however, is found between elements

in LP structures, and therefore, tensile capacity between elements across the connections

(both horizontally and vertically) must be provided. This required tensile continuity

across and within the connections can be effectively achieved by providing the ties

schematically shown in Fig. 6. The rationale for GSI should encompass the following: DExtent of damage —an assessment of the nature and extent of local damage that is likely to occur under abnormal loadings; 2)Alternate paths —evaluation of alternate structural actions which can develop as a consequence of a local failure to re-establish load flow in the remaining undamaged

uDevelopment of significant moment continuity between individual elements is not considered

essential technically nor easily attainable economically in the usual American-type LP

sys tems

.

VI-223
structure—stability analysis of the partially damaged structure; and

3. Tie requirements—quantitative determination of requirements for transversal,

longitudinal, vertical and peripheral ties to bridge local failures and to assure stability

of the partially damaged structure.

5. Extent of Damage

The extent of local damage which the structure must be capable of bridging affects the

amount of ties needed to establish alternate structural action. Code writing bodies in the

United Kingdom (26) and Sweden (27) have directly acknowledged that a certain amount of

local structural damage is considered inevitable and acceptable by specifying the extent

of local failure which the structure must be able to bridge. The specified amount of

damaged load-bearing wall is to be assumed as "notionally removed" from the structure at

any location within a given building.

The main premise when using the "notional removal" approach is that an element or portion of the structure has been totally removed. In LP buildings with brittle joints

which depend primarily on friction and bond under compressive loadings, the "notional

removal" concept may be a realistic representation of their response to abnormal loadings.

However, as vertical and horizontal ties are introduced within and between the wall

elements and floor components, it becomes less likely that the elements can be literally

removed. Therefore, with ductility and continuity in the form of ties between structural

elements specified as recommended practice in American LP construction, the "notional

removal" approach becomes somewhat questionable.

With LP structures tied together horizontally and vertically it is reasonable to

assvmie that in most instances wall panels become "ineffective" as a result of abnormal

loadings. They will no longer function as load-bearing members as originally intended,

but will remain in place in their damaged condition. Whether wall panels become ineffective or are notionally removed is actually of minor consequence, since in either case

the original load flow is interrupted and an alternate path must be available. In fact,

differentiation between the two concepts is not required at all with regard to the main

alternate paths available, that of cantilever and beam action.

The concept of ineffective behavior is introduced here primarily because of its

effect on debris loading and slab behavior in the damaged state. Specifically, the

requirements for the slabs to "hang together" over the damaged (buckled) wall vary from

the requirement of developing full catenary action over double span when the supporting

wall is notionally removed. As a result, the tie requirements for the slabs using the

"ineffective member" rationale will be different from those required if the "notional

removal" approach is adopted.

When estimating damage extent, the following should be taken into account:

1) probability of occurrence of an abnormal load for each element of the structure,

2) the consequence of failure of a particular element within the structure, and

3) the influence of a particular structural configuration and layout of the walls.

VI-224
6. Element Vulnerability

Since the probable damage caused by abnormal loads differs in various parts of a building (1) , damage criteria which do not delineate the damage location, if applicable

in one area, may not apply in other portions of the building. As a result, it is appropriate to define an element's vcirerability by its location within the structure, considering the consequence of its failure.

a) Exterior (Flank) Wall Panel . The wall panels along the perimeter of an LP building are the most susceptible elements to abnormal loadings. Although the sizes of

typical flank wall elements range from 25 to 45 ft. in length, it is reasonable to assume

that, as an outside limit, an entire unit of a flank wall assembly can become ineffective

(Fig. 7) . This assumption can be somewhat relaxed, however, if measures are taken in the

design to ensure that only a part of the wall panel becomes ineffective. Such measures

may include "weak planes", or additional stiffening elements in the form of bracing walls

or integral columns (Fig. 8).

Under the above conditions, "ineffective" indicates that the wall, although physically

damaged and possibly buckled, is still "strung" in place, not allowing the slab units to drop to the level below to create impact loading.

b) Interior Wall Panel . Interior wall panels, being farther removed from the periphery of the structure than flank wall elements, are subjected to a somewhat lesser degree

of risk of specific abnormal loadings. However, it is assumed as an outside limit that

an entire wall panel can become ineffective as shown in Fig. 7. As in the case of flank

wall assemblies, this assumption can be relaxed if similar measures are taken in the design

to ensure that the entire wall panel cannot become ineffective (Fig. 8)

.

c) Floor/Roof Panel . Floor/roof systems typically consist of side-by-side members

(planks) up to 8 ft. wide. As a result, full continuity of the system perpendicular to

the span is interrupted at each joint. It is reasonable to assume, therefore, that in case of an abnormal loading, only a few floor/roof planks would become ineffective for a full span length. As the floor slab consists of individual planks, there is little

tendency for the failure to spread beyond the planks affected by the abnormal loading.

For floor/roof elements, "ineffective' suggests that they are physically damaged and

substantially weakened, and no longer capable of providing lateral support of wall units

or of participating in the required diaphragm action. However, they are still strung

together and capable of supporting their own dead loads within their original spans.

The most critical condition for floor/roof panels occurs when they lose their

stability due to partial loss of one of their end supports, i.e., a wall panel. The loss

of such support would be due to the ineffectiveness of the wall element below, resulting

in large displacements at the original support. Such conditions require adequate tensile

continuity between adjacent slab spans for the slab elements to hang together in their

deflected shape.

VI-22 3
7. Tie Functions and Alternate Paths

Structural design for "normal" loadings (those specified in codes) begins with an assessment of the loads on the building. These loads should be safely transferred from

the point of application to the final resisting point through a logical load flow. If a panel becomes ineffective as a consequence of an abnormal loading, a new load flow must

be established in the remaining undamaged structure, i.e., an alternate path must be

created.

To avoid progressive collapse which may occur if the structure is unable to bridge

local damage, alternate load paths should be available for any load-bearing wall or portion thereof that might become ineffective. For the slab elements such alternate paths

are not essential. However, to avoid uncontrollable debris loading, provisions should

be made to effectively tie (or string) the floor/roof elements together.

Tying the large panel structure together horizontally and vertically makes it possible to utilize the following structural mechanisms to bridge local failures:

1. cantilever action of wall panels,

2. beam action of wall panels,

3. partial membrane action of successive spans of floor planks,

4. vertical suspension of wall panels, and

5. diaphragm action of the floor planks.

a) Transverse Ties and Cantilever Action . When, at a given floor level, a wall panel

or portion thereof becomes ineffective, the wall panels above that level have lost their

support. The most effective method of transferring the loads is through cantilever action

of the remaining panels above. However, to obtain cantilever action, the means to transfer the vertical load at the supported end of the cantilever must exist. Vertical support

at the cantilever root can be provided by:

1. a vertical connection designed to carry the shear in each story in a wall

assembly consisting of two or more adjacent vertical stacks (Fig. 9)

2. a vertically continuous return wall or a vertically continuous integral column

at the interior edge of the wall assembly consisting of a single vertical stack (Fig. 10)

or

3. the remaining portion of the wall (Fig. 11); special detailing should ensure

that only a portion of the wall becomes ineffective.

Although cantilever support can be assured through proper panel design and layout,

the cantilever moment can be developed only if adequate tensile continuity in the form

of a reinforcing tie exists in the transverse horizontal connection between the wall

panels. To assess these tie force requirements the behavioral mode must be determined.

Depending on the horizontal connection properties , two modes of cantilever action

can be defined as giving upper and lower bounds to the transverse tie force requirements.

In the cantilever action shown in Fig. 12a, no horizontal shear is transmitted between

story-high cantilevers. The strength of this builtup cantilever is equal to the sum of

the strengths of the individual cantilever panels. In practice, this type of action is

unlikely, since the joint is filled with drypack or grout, and some vertical reinforcement

VI-226
(in the form of vertical ties) is used to interconnect successive wall panels. Both the

concrete and the vertical ties act to provide shear resistance across the connection by

shear friction, and consequently, enhance the cantilever strength. This behavioral mode

provides a lower bound in predicting cantilever strength.

If adequate shear capacity can be achieved between successive stories, the entire

wall panel assembly will act as a monolithic cantilever as shown in Fig. 12b. This

behavior can be achieved when there is no relative slip between stories. Such an idealiz

condition is questionable as some slip will most likely occur within the vertical tie

connection details. Therefore, this behavioral mode should be considered an upper bound

in predicting the cantilever strength.

In practice, the behavior will be somewhere between the two modes described above,

and the tensile force requirements for the transverse tie in a multistory cantilever will

depend primarily on the details affecting shear characteristics of the horizontal connections.

The transverse tie within the connection can be in the form of mild steel or unstres

sed prestressing strand. To ensure its effectiveness, the tie should extend the full

width of the structure, and be adequately secured to the peripheral tie system (Fig. 13)

.

b) Transverse Ties and Beam Action . Although cantilever action through the use of

transverse ties is the single most important element in establishing alternate paths in

LP buildings, in some instances of local damage the transverse ties can also be very

effective in developing "beam" action of the wall panels. However, to develop beam

action, a capacity to transfer the vertical load to each of the supported ends must

exist. Such supports are available in the cases shown in Fig. 14. As in the case of cantilever action, the vertical supports for beam action are provided through proper panel design and layout. The beam's flexural resistance can be

developed only if adequate tensile capacity in the form of reinforcing ties exists in

the transverse horizontal connection or within the wall panels themselves. The factors

affecting the overall behavior of this alternate load flow are similar to those examined

under cantilever action. Vertical load-carrying capacity at each end must be assured

and, as before, the shear characteristics of the horizontal connection between successive

wall panels substantially influence the behavior.

The necessary ties to resist the tensile forces from the bending moment will be

less than those required under similar span cantilever action; however, this structural

behavior will usually require more return walls or integral columns to assure vertical

supports. The transverse ties to effect beam action can be either of mild steel reinforcing or unstressed prestressing strand. The ties should be continuous the full width

of the building and anchored to the peripheral tie system (Fig. 13)

.

c) Longitudinal Ties and Partial Membrane Action (Large Deflections of Slabs) . Cantenary action of the continuous slab spans due to loss of supports has been investigated in Europe (28) . Although full catenary action is accepted as an appropriate and

functional means of reestablishing the load flow in European LP structures, it is not

considered in this study a suitable method for American LP construction for the following

VI- 227
1. The typical slabs for European systems span from 10 to 18 ft. , are cast for a particular building, and are joined together by interlacing loops protruding from the

slabs. LP systems used in the United States have floor spans which range from 20 to 40

plus feet. The U.S. slab systems are generally prestressed hollow core slabs produced

in long beds and cut to shorter lengths according to the required span. In this fabrication technique, reinforcement protrusions at the ends are not feasible, and tensile continuity between adjacent spans must be developed by other means. Since the span of the

necessary catenary in U.S. construction (doi±ile the slab span) is about twice that of its

European counterpart, the problems related to physical development of the necessary

longitudinal tensile forces are greatly magnified.

2. To develop the necessary catenary action, the floor systems must undergo large

deflections. With 40 to 80 foot catenary spans in U.S. construction, the necessary deflection may approach a total story height. Such large deflections can be accommodated only

when the panel below is totally removed; this is, in turn, contrary to the concept of walls

becoming ineffective.

The major element of resistance and stability in the damaged wall-type structure is

provided by cantilever and beam action. The fixnction of the slab is to ensure an adequate

degree of partial catenary action to effect a stringing of the elements to inhibit progressive collapse from debris loading should an interior wall become ineffective. The wall's

ineffectiveness is suggested as a limited displacement (Fig. 15) , say up to 24 inches.

Thus, to string the slab elements together in this deflected shape, adequate tensile

continuity must exist between them, above the ineffective wall. Ductility of the connec- tion must be ensured to permit the large deformations and to resist any dynamic effects

that may occur. Continuity can be developed with reinforcing placed in the longitudinal horizontal

keyways between floor panels, or in the topping slab if one is employed. These ties must

provide tensile continuity between adjacent spans and, to ensure integral behavior, the

ties of the end span should be anchored into the peripheral tie system. Since floor

elements are well reinforced, the ties need not be continuous for the full length; the tie

force should be transferred by bond or mechanically from an element in one span to the one

in the next span (Fig. 16)

d) Vertical Ties and Suspension Action . The primary functions of vertical ties

between wall elements in the damaged structure are:

1. to provide vertical suspension of ineffective wall panels and thus avoid debris

loading

2. to provide resistance against "kicking out" of the walls sideways, thus fostering

"ineffective" behavior of wall elements, rather than total removal under abnormal loading;

and

3. to assure clamping and dowel action in the horizontal connections for shear friction to resist horizontal shear and thus develop cantilever and beam action.

VI-228
If an alternate path for the load transfer around an ineffective wall panel has been

established either through cantilever or beam action, the damaged panel should be suspended

from the alternate supporting mechanism. To ensure this suspending action, vertical continuity in the form of tensile ties must exist between and through all wall panels. Tensile

continuity from foundation to the roof can be assured by providing vertical ties proportioned to resist (within each story) the dead loads of the wall plus those loads superimposed by the floor panels. In addition, the vertical tie must resist the stresses from

shear friction clamping in the horizontal connection.

Because of the increased vulnerability of exterior wall panels (as compared to interior wall panels) with regard to abnormal loadings, greater vertical continuity is desirable

in flank wall assemblies in order to afford them increased protection against removal from

the structure. The vertical tie of the flank walls should be placed within the peripheral

tie system to improve its integral connection with the slab system.

The vertical ties may take the form of mild steel, prestressing rods, or prestressing

strand, stresses or unstressed. They should be continuous for the full height of the panel

assembly (Fig. 17) . The ties can also function as reinforcement required for service and

erection loads.

e) Peripheral Ties and Diaphragm Action . The main functions of peripheral ties in

the damaged buildings are:

1. to establish the necessary diaphragm action to resist the effects of wind, torsion

and unequal load distribution throughout the structure;

2. to anchor the longitudinal and transverse tie forces at the periphery; and

3. to help create membrane action in corner slabs when the corner support becomes

ineffective.

These functions can be accomplished by employing a ring beam at the periphery of the

structure at each floor and roof level.

I| To assess the force level required in such a ring beam in a structure is a difficult,

if not impossible, task. In the damaged state the entire slab or roof may act as a membrane^

with its periphery subject to large ring beam forces. It is suggested that the peripheral

tie forces be based on the diaphragm action necessary to resist normal lateral loads,

assiming the customary load factors and capacity reduction factors. However, for purposes

j of continuity and anchorage requirements in the damaged state , the transversal tie forces

required in the flank wall assemblies should be continued in the longitudinal faces (if

larger than those required for diaphragm action) for the design of the complete periphery

(Fig. 18). These continuous ties, which should be located as close to the actual periphery

as possible, may be of mild steel or prestressing strand or mechanical connectors between

ends of planks in adjacent spans.

:| 8. Loads and Safety Factors in the Damaged Structure

j To determine the quantity of ties needed for stability of the damaged structure, the

' necessary loads to be resisted must be established. Consideration should be given only to

, those loads likely to occur before temporary or permanent measures are taken to repair the

VI-229
damaged area. Included in this category are the full deal load and part of the design wind

and live loads. It is sufficient to assume 1/3 of the design live load and 1/3 the design

wind load. Reduced live load is based on the assumption that a building will be evacuated

after an abnormal incident has caused local damage. The reduced wind loading is based on the supposition that the building will have partial stability for a limited period of time.

Since the rate of occurrence of local damage due to abnormal loadings is relatively

low, and since the intent is to assure the stability of partially damaged structures, it

is assumed that the above loads are resisted at a level close to the ultimate strength

of the sections, say 90% of ultimate capacity. Therefore, -for the determination of

minimum tie forces, liberal load factors, Y» a-i^e appropriate. However, the capacity reduction factors, (f), which accoiint for variations in material strengths, workmanship and

dimensional errors combining to result in undercapacity , must still be used. Since the

ties are utilized to resist axial tension and tension due to bending, a factor of 0.9 can be used.

Combining the loads (D, L, W) and load factor, Yf the required strength equation

may be expressed as

U = 1.10 (D + L/3 + W/3)

Incorporating the capacity reduction factor, (f) , the safety factor of the damaged

structure becomes

S F = = 1 22

9. Specifying Tie Forces - Unstressed Strand

Specifying tie requirements, in terms of forces rather than amounts of steel permits

flexibility in the choice of tie type, while fostering a better understanding of the issues

involved. Specifying forces allows the use of unstressed prestressing strand, which has

the advantage of large tensile capacity with a relatively small diameter. Flexible strand

also offers the advantage of easy placement in misaligned connections where a normal mild

steel bar cannot be placed due to its rigidity. While unstressed prestressing strand can- not be used for resistance at service load levels, since it would result in unacceptable

crack widths, the large strains of the strand are not objectionable under catastrophic

conditions

10. Planform

Proper structural planform, i.e., the layout of the structural walls, improves the

overall rigidity and stability of the structure, thereby increasing its structural integrity. Effective planform is best accomplished by arranging walls in both orthogonal directions of the building.

VI-230
since current connection details make frame action between adjacent slab spans and

between slabs and walls almost impossible, lateral resistance of LP buildings is provided

economically only by cantilever action of the stacks of wall panels in their own planes.

Therefore, where possible, longitudinal spine walls or segments should be used to enhance

the stability of the longitudinal direction. They will also improve the stability of individual transversal walls, while decreasing the length of wall likely to be affected by

an abnormal load.

For exterior walls, returns or integral columns are desirable to enhance their resistance to abnormal loadings. If vertical stabilizing cores are the only elements for

lateral stability, it is desirable to distribute them throughout the building (Fig. 10) so that if one is damaged, total stability of the structure under lateral loads is not forfeited. Shrinkage, temperature and distribution steel placed in the topping can also be

used to enable the slab to span in a perpendicular direction and prevent collapse where a support configuration is favorable for two-way action (Fig. 20)

.

11. Building Protection

Protecting the structure from abnormal effects is one of the basic methods to avoid

progressive collapse. When an abnormal load can be controlled, such control is recommeded;

e.g., provision of venting to relieve explosions, use of shock absorbers for crash barriers

against vehicular collision, prohibiting construction in flood plains, etc. In this way,

the forces and their effects due to certain abnormal events can be reduced or eliminated,

thus lowering the overall risk from abnormal events.

12. Concluding Remarks

It is technically very difficult and economically prohibitive to design residentialtype buildings for absolute safety. On the other hand, there is no justification for

constructing buildings which do not afford a certain degree of safety with regard to abnormal loads. To reduce the risk of progressive collapse of large panel residential buildings,

a philosophy is developed to assure bridging of local damage while maintaining stability

of the partially damaged structure by tying the components of large panel structures

i together horizontally and vertically. Based on this philosophy, explicit requirements for

a minimum tie system of transversal, longitudinal, vertical and peripheral ties will be

developed. It is intended that these requirements establish a General Structural Inte- grity for the building to minimize the effects of abnormal loadings.

The provision of General Structural Integrity eliminates the need to design for any

particular abnormal load, since the ability to bridge local damage is provided.

In establishing structural integrity, the need for tensile continuity and ductility

of the elements, as well as of the overall structure, is recognized. This will be

i| accomplished through a rational arrangement of tensile ties and through connection details

I which will assure alternate structural behavior of the damaged building. This combination

of system continuity and ductility should enable the structure to either absorb the

I VI-231
abnormal loads with minimal damage, or bridge localized damage as a result of the abnormal

load. The provision of General Structural Integrity will bring the safety of LP structure

closer to that of the traditional cast-in-place reinforced concrete buildings.

The objective of this approach is not to afford absolute safety in regard to any

exceptional event in any part of every building; rather, the intention is to limit and

substantially reduce the general risk of collapse, as compared to that existing if no

such measures were taken. Based on experimental results performed thus far, it appears

that the amount of ties required to achieve the desired level of General Structural

Integrity will have a minimal effect on the economics of LP structures. With the philosophy established, subsequent experimental tests and analytical studies will focus on development of purposeful details to optimize the effectiveness of the ties in the structure.

Acknowledgement

The material in this paper is excerpted from a larger publication entitled: "Structural Response of Large Panel Buildings to Normal and Abnormal Loads". It is the second

in a series of reports prepared at the Portland Cement Association for the Department

of Housing and Urban Development under Contract H-2131 (Development of Criteria for the

Design and Construction of Large Panel Concrete Structures)

References

1. Schultz, D.M. and Fintel, M. , "Report 1: Loading Conditions," Design and Construction of Large Panel Concrete Structures , Office of Policy Development and Research,

Department of Housing and Urban Development, Washington, D.C., April 1975.

2. Somes, N.F. , Abnormal Loading on Buildings and Progressive Collapse, NTIS Com-73

1113, prepared for the Office of Policy Development and Research, Department of Housing

and Urban Development, Washington, D.C., May 1973.

3. Burnett, E.F.P., "Abnormal Loading and Building Safety," Industrialization in

Concrete Building Construction , SP-48, American Concrete Institute, Detroit, Michigan,

1975.

4. Ferahian, R.H., Design Against Progressive Collapse , Technical Paper 332, Ottawa,

Canada, National Research Coxincil of Canada, Division of Building Research, July 1971.

5. Ferahian, R.H., "Buildings: Design for Prevention of Progressive Collapse,"

Civil Engineering , American Society of Civil Engineers, New York, February 1972.

6. Creasy, L.R. , "Partial Stability," Design Circular , Directorate of Civil Engineering Development, Department of the Environment, London, England, August 1971.

7. Creasy, L.R., "Stability of Modern Buildings," The Structural Engineer , Vol.

50, No. 1, London, England, January 1972.

8. Discussion of the Stability of Modern Buildings, The Structural Engineer , Vol. 50, No. 7, London, England, July 1972.

9. Creasy, L.R. , "The Art of the Possible," A Presidential Address given before

the Institution of Structural Engineers, London, England, October 4, 1973.

10. Stability of Modern Buildings , The Institution of Structural Engineers, London,

England, September 1971.

11. Popoff, A., "State-of-the-Art Report No. 3 - Stability of Precast Concrete

Systems Buildings," Planning and Design of Tall Buildings , Vol. Ill, American Society

of Civil Engineers, New York, 197 3. VI-23 2
12. Popoff, A., "Design Against Progressive Collapse," Journal of the Prestressed

Concrete Institute , Vol. 20, No. 2, March-April 1975.

13. Burnett, E.F.P., The Avoidance of Progressive Collapse ; Regulatory Approaches

to the Problem , prepared for Center for Building Technology, National Bureau of Standards,

Washington, D.C., October 1975.

14. PCI Committee on Precast Bearing Wall Buildings, "Considerations for the Design

of Precast Beaming Wall Buildings to Withstand Abnormal Loads , " Journal of the Prestressed

Concrete Institute , Vol. 21, No. 2, March/April 1976.

15. Granstrom, S., and Carlsson, M. , Building Design for Avoidance of Progressive

Collapse , Part II: Survey of Some Recent Collapses, Library Translation No. 1670,

Building Research Station, Watford, England, April 1972.

16. "Aims of Structural Design", The Institution of Structural Engineers, London,

England, August 1969.

17. Granstrom, S. , "Byggnaders Stabilitet Efter Katastrofskador . Drafter i Elementfogar-Modellforsok, " (Stability of Building After Accidental Damage. Forces in

Element Joints-Model Tests), Statens Institut for Byggnadsforskning , Report R20:1971,

Sweden Symposium at University College of Cardiff, December 1970.

18. The Stability of Precast Concrete Structures , (Proceedings of the discussion

periods following the morning and afternoon sessions at a Seminar at the Europa Hotel,

London, March 27, 1973) , Building Research Establishment, Department of the Environment

Garston, Watford, England, December 1973.

19. "Stability of Buildings and Their Design for Prevention of Progressive Collapse

Due to Local Damage," Report of Subcommittee on Large Concrete Panels , Nordic Concrete

Association Committee on Regulations, Sweden, January 29, 1970.

20. Allen, D.E. and Schriever, W.R. , "Progressive Collapse, Abnormal Loads, and

Building Codes , " Structural Failure: Modes, Causes, Responsibilities , American Society

of Civil Engineers, New York, 1973.

21. Smith, B.S. and Rahman, K.M.K. , "A Theoretical Study of the Sequences of Failure

in Precast Panel Shear Walls," Proceedings of the Institution of Civil Engineers, Vol.

55, Part 2, London, England, September 1973.

22. Armer, G.S.T. and Kumar, S. , Tests on Assemblies of Large Precast Panels , Current

Paper CP 20/72, Building Research Station, Department of the Environment, Garston, Watford,

England, 1972.

23. The Behavior of Large Panel Structures , CIRIA Report No. 45, Construction

Industry Research and Information Association, London, England, 1973.

24. Granstrom, S.A., "The Scandinavian Approach to Structural Safety," Buildings

and the Hazards of Explosion , Proceedings of the Symposium at BRE, Garston, October 18,

1972, Building Research Establishment, Department of the Environment, Garston, Watford,

England, 1974.

25. "Symposixim on Progressive Collapse," ACI Annual Convention, sponsored by the

ad hoc committee on progressive collapse. Technical Activities Committee, American Concrete

Institute, Boston, Massachusetts, April 1975.

26. British Standards Institution, "Part I - Design Materials and Workmanship,"

The Structural Use of Concrete , CPllO: Part 1, BSI, London, England, 1972.

27. Swedish Building Standards, "Swedish Regulations for Prevention of Progressive

Collapse," SBN 22:35, Stockholm, Sweden, 1973.

28. Wilford, M.J. and Yu, C.W., Catenary Action in Damaged Structures-Tests on R.C. Beams and Slabs, Building Research Establishment, Ref . 391/73, Watford, England.

VI-233

VI-235
Failure due to loss of support

Failure due to debris loading

(a) lirmediate Local Damaqe (b) Progressive Collapse

Figure 4 Failure Modes of Ronant Point Collapse

VI-236
Isometric View of Typical Large Panel Concrete Structure

Cross Wall System Spine Wall System

Figure 5 Typical Large Panel Structures

VI-237
VI-2 38
VI-239
VI-240
Continuous vertical connection

11

Ineffective panel

y//////////////////////

Section A - Partial Wall Elevation

Individual elements

Continuous _

"vertical connection

Possible

ineffective panel

Partial Plan

Figure 9 Cantilever Mode—Two Adjacent Wall

Panel Stacks

VI-241
Return wall or

integral column

Section A - Wall Elevation

Possible -

ineffective

panel

Integral

column

Ineffective

panel

Plan Section

of Integral

Column

Return

walls

Partial Plan

Figure 10 Cantilever Mode—Single vertical Wall Panel Stack

VI-242
t

.^^^^ Partial ineffective

panel

Section A - Partial Wall Elevation

Partial Plan

Figure 11 Cantilever Mode—Partial Element Ineffectiveness

VI-243
Vl-244
VI-245
Return wall

or integral

column

Ineffective panel

Section A - Wall Elevation

APossible

ineffective "^p

panel

Plan Section

of Integral

Column

Partial Plan

Figure 14 Beam Mode—Single Vertical Wall Panel Stack

VI-246
VI-247
oc

'X3

'5> cocoooo

(AOo c

laouoou&>

IdaHap

+>H

01CO

Himoo

•rH

4-1 fOuo

1-q 73C

l« +J

(U +)X

1^TJ

0) P

CO

(1) 01

013Wu3

01 •H

P4

VI-248

VI-250
—stabilizing core

(a) Undesirable Core Layout

Longitudinal

stabilizing core

centrally located

within building

Stabilizing cores

Longitudinal

stabilizing cores

distributed

throughout building

(b) Desirable Core Layout

Figure 19 Possible Core Layouts in Typical Cross Wall Structure

Ineffective wall element

i23

/'.'.'••••|A:.':v7?n

Direction of span

before incident

Direction of span

after incident

Figure 20 Change of Slab Span Direction under

the Effects of Abnormal Loadings

VI-251

Wind And Seismic Design of United States

Nuclear Power Plants

LAWRENCE C. SHAO, RICHARD J. STUART and CHARLES H. HOFMAYER

Nuclear Regulatory Commission, Bethesda, Maryland

ABSTRACT

All U.S. Nuclear Power Plants are currently designed to resist the effects of severe and

extreme environmental loads in combination with operating and accident loads (tornado

and accident loads are not combined) . This affords two levels of resistance to wind and

seismic loads i.e., the plant can remain operational during and following the operating

basis earthquake or the design wind, and shall be able to safely shut down following the

design basis tornado or safe shutdown earthquake. The design wind parameters are

normally selected based upon a one hundred year recurrence interval. These parameters

(wind velocity, velocity vertical profiles, and applicable gust factors) are transformed

into applied pressures on exposed surfaces of safety related structures considering the

effects of ground exposure, shape coefficients and pressure profiles at discontinuities.

The design basis tornado is selected as the most severe tornado that can reasonably be

predicted at a site based on the geographical distribution of the frequency of tornado

occurrence. The continental United States is divided into three tornado regions and

for each region the following tornado characteristics are specified: the rotational

wind speed, the translational wind speed, the pressure drop across the tornado, the rate

of pressure drop and the radius of maximum rotational wind speed. The conversion of thes

parameters to design pressures must consider the orientation of the tornado relative to

the structure as well as the other factors considered for the design wind. Various desig

tornado missiles and corresponding velocities are specified for each plant. Safety related structures are designed to resist both local damage (penetration and spalling) and

overall structural response due to these missiles. The SSE*is chosen as the maximum

earthquake expected to occur at the site and normally has a probability of occurrence of

KEY WORDS: Seismic effects; Wind; Tornado; Nuclear Power Plant

* Safe Shotdown Earthquake

VII-1
less than 10 . The OBE is chosen as the earthquake which could reasonably be expected

to affect the plant during its operating life; however, its magnitude cannot be less than

one half that of the SSE. Vertical and horizontal ground acceleration values and broad

band ground response spectra are specified for each plant site. A detailed dynamic

analysis is required for each plant, including the consideration of the effects of damping, soil-structure interaction, three components of earthquake motion, combination of

modal responses, etc. These considerations are detailed in recently published U.S.

Nuclear Regulatory Commission's Standard Review Plans 3.7.1 Seismic Input, 3.7.2 Seismic

System Analysis, 3.7.3 Seismic Subsystem Analysis and 3.7.4 Seismic Instrumentation.

INTRODUCTION

The regulations which govern the wind and seismic design of U.S. nuclear

power plants are 10 CFR, Part 50, Appendix A - "General Design Criteria" and

10 CFR, Part 100, Appendix A - "Seismic and Geologic Siting Criteria". From

these regulations the U.S. Nuclear Regulatory Commission has developed positions

and Standard Review Plans''" which detail approved techniques for satisfying 10 CFR

requirements. The appropriate sections of the Standard Review Plans which cover

wind, tornado and seismic design are listed in Table I.

Before discussing the details of wind and seismic design, it is helpful to

review several cross sections through a reactor containment building. Figures 1

and 2 depict the containments for a typical Boiling Water Reactor (BWR) and a

typical Pressurized Water Reactor (PWR) , respectively. The reactor vessel is

located in the center of both figures. In both of these cases the containment

boundary is protected by a layer of reinforced concrete which is between two and

one half (2-1/2') and five feet (5') thick. Between the reactor vessel and the

containment are concrete walls which function in part, as both radiation and

Standard Review Plan for the Review of Safety Analysis Reports for Nuclear

Power Plants (NUREG-75/087) , September 1975, U.S. Nuclear Regulatory Commission,

Office of Nuclear Reactor Regulation, Washington, D. C. 20555

VII- 2
missile shields. Typical containment dimensions are 220 feet high by 140 feet

in diameter with a 10 foot basemat. The plant embedment varies from thirty (30')

to ninety feet (90') below grade. Although the bearing pressure is extremely

high, power plants are founded on both soil and rock sites. Support buildings

and control buildings are located adjacent to the containment and are more similar

to conventional rectilinear buildings constructed of steel and concrete. However,

the design of all safety related nuclear power plant structures far surpasses conventional structures in the severity of the design loads and in the precision of

the analysis and design techniques. For example, there are two levels of both

wind and seismic design. During the less severe events the plants are designed

to remain fully operational and during the more severe events the plants are designed to safely insert control rods and shut the reactor down.

REQUIREMENTS FOR WIND LOADINGS

Design Wind

The design wind parameters are normally selected based on a one hundred

(100) year recurrence interval. Figure 3 is a plot of the one hundred (100)

year basic wind speed at thirty feet (30') above the ground for the continental

United States.

Having the basic wind speed for a general location one must determine the

vertical velocity profiles and applicable gust factors. These parameters are

dependent upon the site, terrain, and surrounding structures and are generally

determined from site specific meteorological data. These parameters are trans2 formed into design pressures on the structures using ASCE Paper No. 3269 or

3 ANSI A58. 1-1972 . This transformation procedure considers the ground exposure,

structural shape coefficients, gust factors and pressure profiles at discontinuities.

2 ASCE Paper No. 3269, "Wind Forces On Structures," Transactions of the American

Society of Civil Engrs., Vol. 126, Part II, (1961)

3 ANSI A58. 1-1972, "Building Code Requirements for Minimum Design Loads in Building

and Other Structures," Committee A58.1, American National Standards Institute

(1972)

VII-3
As an example of the transformation considerations, Figure 4 illustrates a

qualitative depiction of the pressure profiles of wind impinging on a rectangular structure.

Tornado Loads

Tornado design considerations constitute the second level of wind design of

U.S. Nuclear Power Plants. The continental United States is divided into three

tornado intensity regions as depicted in Figure 5. The most severe tornado

effects occur in Region I. The Design Basis Tornado characteristics differ for

each region and are summarized in Table II.

The qualitative distribution of the rotational wind speed and the pressure

drop relative to the centerline of the tornado is shown in Figure 6. It is apparent from this distribution, that several tornado attitudes and locations must be

investigated in order to determine the worst possible effects upon the structure.

Tables III and IV list typical missiles, missile parameters and corresponding velocities used in the design of nuclear safety related structures. These

missiles vary slightly from plant to plant, but are intended to encompass the

range of credible missiles which might affect plant operation during a tornado.

The missiles listed in Table III are considered to be capable of striking in all

directions. Missiles A, B, C, D, and E are to be considered at all elevations

and missiles F and G at elevations up to thirty (30) feet above all grade levels

within one half (1/2) mile of the facility structures. The "no-tumbling" hori4 zontal missile velocities presented in the Topical Report TVA-TR74-1 are also

acceptable provided that a four thousand pound (4000//) automobile at seventy (70)

mph and elevations up to thirty (30) feet above grade level is added. These

velocities are listed in Table IV. Vertical velocities equal to eighty percent

(80%) of the TVA horizontal velocities listed in Table IV are also acceptable on

an interim basis.

4 "The Generation of Missiles by Tornadoes," TVA-TR74-1, Tennessee Valley Authority

(1974)

vii-4
Given the tornado design parameters listed in Tables II, III and IV, one must

consider the following items in the tornado design of each plant:

1. Transformation Procedures

a. Wind Velocity to Pressure

b. Missile Effective Loads

2. Failure of Non-Safety Related Structures Onto Category I Structures

3. Venting of Structures

4. Combinations of Tornado Design Parameters.

The transformation procedures to convert wind velocity to pressure and to account

for building shape factors, are identical to those outlined under wind loading.

The conversion of the missile parameters to structural effective loads is done by

calculating the kinetic energy of the missile and subsequently converting that

energy to structural reactions. Another consideration in tornado design is to

insure that non-safety related structures will not collapse or generate missiles

which might impact upon safety related (Category I) equipment or structures.

Such structures must be designed to mitigate these failures although significant

deformation may be permitted.

As depicted in Figure 7, venting of structures to reduce the consequences

of the tornado negative pressure may reverse the normal pressure distribution

due to wind. This process becomes even more complex when there are many vented

cells within a structure and detailed air flow computer programs are necessary

to solve such problems.

In order to determine the total tornado load (W^) to be combined with dead

loads, live loads, temperature loads, accident loads, etc, the wind, pressure and

missile loads are combined as follows:

M

0.5 Wp

w„ + w.M

W„ + 0.5 W„ + W, W P 1M

VI 1-5
where : W, Total Tornado Load

W

Tornado Wind Load

Tornado Differential Pressure Load

M

Tornado Missile Load

It should be noted that fifty percent of the negative pressure load is combined

with the wind load as depicted in Figure 6.

In order to mitigate the effects of missile impact, building walls are

strengthened to function as missile barriers, or walls are constructed in the

paths of potential missiles to function solely as missile barriers. These barriers

could be constructed of concrete, steel, or composite construction and multiple

barriers may also be employed. Acceptable methods to determine the depth of penetration for each barrier type are:

1. Concrete - depth of penetration by modified Petry as given by Amirikian^ -

prevent perforation, limit spalling and scabbing

2 . Steel - depth of penetration by Stanford Research Institute Formula

3. Composite and multiple barriers - by Recht and Ipson^

The overall structural damage of a barrier is normally determined by equating the

kinetic energy of the missile to the elasto-plastic strain energy of the structural

component during the impact. The acceptable deformation is limited by restricting the allowable ductility to a small percentage of the ultimate failure ductility

for a particular component and failure mode. Moments, forces, and shears are

A. Amirikian, "Design of Protective Structures," Bureau of Yards and Docks,

Publication No. NAVDOCKS P-51, Department of the Navy, Washington, D. C. (August

1950).

W. B. Cottrell and A. W. Savolainen, "U.S. Reactor Containment Technology,"

ORNL-NSIC-5, Vol. 1, Chapter 6, Oak Ridge National Laboratory.

R. F. Recht and T. W. Ipson, "Ballistic Perforation Dynamics," Journal of Applied

Mechanics, Transactions of the ASME, Vol. 30, Series E, No. 3, September 1963.

VII-6
calculated by developing plastic hinges and using standard formulation for determining their magnitudes. An alternate acceptable method to determine missile loads

on structures is an equivalent static load technique developed by Williamson and

Alvy .

REQUIREMENTS FOR EARTHQUAKE LOADINGS

The design wind and tornado loads are classified, respectively as severe and

extreme environmental conditions. In addition to these loads there is another

set of severe and extreme environmental loads which must also be considered.

They are the Operating Basis Earthquake (OBE) , and the Safe Shutdown Earthquake

(SSE) , respectively. As with the design wind and the tornado, a nuclear power

plant must remain operational during an OBE and must be capable of safely shutting

down during an SSE.

Seismic Input

The design parameters for both design events are the ground acceleration values

determined from site investigations of geology, proximity to faults and the surrounding tectonic provinces, and the design response spectra specified in Regula9 tory Guide (R.G.) 1.60 . A typical R.G. 1.60 spectra is depicted in Figure 8.

This response spectra was determined by subjecting a series of linear oscillators

of varying frequencies to all available earthquake time histories recorded in the

U.S. and enveloping those results. Given the design parameters, the procedure to

determine the building floor response spectra considering soil structure interaction,

damping values and the analytical model will be discussed in more detail.

Variation of soil properties is considered in the analysis procedure as illustrated in Figure 8. The resulting basemat response is then enveloped for use in

determining the floor response.

R. A. Williamson and R. R. Alvy, "Impact Effect of Fragments Striking Structural

Elements," Holmes and Narver, Inc., Revised November 1973.

Regulatory Guide 1.60, "Design Response Spectra for Nuclear Power Plants," USNRC

,

Washington, D. C. 20555

VII-7
The method of analysis for soil supported sites utilizes a finite element grid

as depicted in Figure 9 . Having developed a synthetic time history at the rock

layer which envelops the Regulatory Guide 1.60 spectra in the free field, this input

motion is used to determine the response of the basemat and other major structural

elements. On rock sites a simpler model, depicted in Figure 10, is acceptable.

The spring constants and damping coefficients are normally determined by using an

infinite half space model. The acceptable analytical tools for soil-structure

interaction are listed in Table V as a fimction of the supporting geological medium.

Seismic System Analysis

Given the basemat response, a more detailed model is developed for the remainder

of the structure and major systems and components. These models are normally three

dimensional or axisymmetric . In the development of these models, in the analysis

procedure and in the evaluation of the results, the following considerations must

1. Seismic Analysis Methods

2. Natural Frequencies and Response Loads

3. Procedures Used for Analytical Modeling

4. Soil-Structure Interaction

5. Development of Floor Response Spectra

6. Three Components of Earthquake Motion

7. Combination of Modal Responses

8. Interaction of Non-Category I Structures with Category I Structures

9. Effects of Parameter Variations on Floor Responses

10. Use of Constant Vertical Static Factors

11. Methods Used to Account for Torsional Effects

12. Comparison of Responses

13. Methods for Seismic Analysis of Category I Dams

14. Determination of Category I Structure's Overturning Moments

15. Analysis Procedure for Damping

These items are discussed in great detail in the NRC Standard Review Plan 3.7.2 (and

Vll-8
the reader is referred to that document) . Table VI and Figure 11 depict two of

these considerations. Table VI is a list of acceptable damping values from

Regulatory Guide 1.61^^. Using these values with the appropriate analytical tools

one can develop floor response spectra for use in equipment design and qualification. As depicted in Figure 11, the design peak is broadened fifteen percent (15%)

on either side of the predicted peak in order to account for possible variations in

the analytical technique and the input parameters.

Seismic Subsystem Analysis

Smaller components and piping systems are analyzed in a similar manner. They

are classified as subsystems. In the analysis of such systems, the following items

are considered:

1. Seismic Analysis Methods

2. Determination of Number of Earthquake Cycles

3. Procedures Used for Analytical Modeling

4. Basis for Selection of Frequencies

5. Use of Equivalent Static Load Method of Analysis

6. Three Components of Earthquake Motion

7. Combination of Modal Responses

8. Analytical Procedures for Piping Systems

9. Multiply-Supported Equipment and Components With Distinct Inputs

10. Use of Constant Vertical Static Factors

11. Torsional Effects of Eccentric Masses

12 . Category I Buried Piping Systems and Tunnels

13. Interaction of Other Piping with Category I Piping

14. Seismic Analyses for Reactor Internals

15. Analysis Procedure for Damping

Regulatory Guide 1.61, "Damping Values for Seismic Analysis for Nuclear Power

Plants," USNRC, Washington, D. C. 20555

VII-9
Seismic Instrumentation

In order to monitor the ground acceleration and response spectra during a

seismic event, the NRC requires the following seismic instrumentation require11

ments :

1. Triaxial Time History Accelerograph

2. Triaxial Response Spectrum Recorder

3. Seismic Switch

Such instrumentation is not only used to provide an historical record of the

structural response, but it is also used to provide operator notification for

shut down of the plant when the OBE ground acceleration is exceeded.

The number and location of typical seismic instrumentation is depicted in

Table VII. The recorded data from such instrumentation is used to compare the

predicted response of structures, systems and components with the response

measured during a seismic event. To date, the predicted response has always

enveloped the measured response.

CONCLUSION

A brief discussion of wind and seismic design of U.S. nuclear power plants

has been presented. All plants are currently designed to resist the effects of

severe and extreme environmental loads in combination with operating and accident

loads (tornado and accident loads are not combined) . This affords two levels of

resistance to wind and seismic loads, i.e. the plant can remain operational during and following the operating basis earthquake or the design wind, and shall be

able to safely shut down following the design basis tornado or safe shutdown earthquake. More detailed information with regard to any of the items discussed can

be obtained from the documents referenced in this discussion.

Regulatory Guide 1.12, "Instrumentation for Earthquakes," USNRC, Washington,

D. C. 20555

VII-10
TABLE I - U.S. NUCLEAR REGULATORY COMMISSION

STANDARD REVIEW PLANS

Section Topic

3.3.1 Wind Loadings

3.3.2 Tornado Loadings

3.5.3 Barrier Design Procedures

3.7.1 Seismic Input

3.7.2 Seismic System Analysis

3.7.3 Seismic Subsystem Analysis

3.7.4 Seismic Instrumentation

VII-11
TABLE II - DESIGN BASIS TORNADO CHARACTERISTICS

MAXIMUIVI WIND SPEED^

(mph)

ROTATIONAL

SPEED

(mph)

TRANSLATIONAL

SPEED (mph)

RADIUS OF

MAXIMUM

ROTATIONAL

SPEED

(feet)

PRESSURE

DROP

(psi)

RATE OF

PRESSURE

REGION MAXIMUM minimum'' DROP (psi/sec)

1 360 290 70 5 150 3.0 2.0

II 300 240 60 5 150 2.25 1.2

III 240 190 50 5 150 1.5 0.6

^THE MAXIMUM WIND SPEED IS THE SUM OF THE ROTATIONAL SPEED COMPONENT AND THE MAXIMUM

TRANSLATIONAL SPEED COMPONENT.

^JHE MINIMUM TRANSLATIONAL SPEED, WHICH ALLOWS MAXIMUM TRANSIT TIME OF THE TORNADO

ACROSS EXPOSED PLANT FEATURES, IS TO BE USED WHENEVER LOW TRAVEL SPEEDS (MAXIMUM TRANSIT

TIME) ARE A LIMITING FACTOR IN DESIGN OF THE ULTIMATE HEAT SINK. THE ULTIMATE HEAT SINK

IS THAT COMPLEX OF WATER SOURCES, INCLUDING ASSOCIATED RETAINING STRUCTURES, AND ANY

CANALS OR CONDUITS CONNECTING THE SOURCES WITH; BUT NOT INCLUDING, THE INTAKE STRUCTURES

OF NUCLEAR REACTOR UNITS, REGULATORY GUIDE 1.27 (SAFETY GUIDE 27), "ULTIMATE HEAT SINK." DESCRIBES A BASIS THAT MAY BE USED TO IMPLEMENT GENERAL DESIGN CRITERION 44 OF APPENDIX A TO 10CFR PART50WITH REGARD TO THE ULTIMATE HEATSINK.

VII-12
TABLE III - TYPICAL TORNADO MISSILES

USED IN THE DESIGN OF

NUCLEAR SAFETY RELATED STRUCTURES

Fraction of Total

Tornado Velocity

A. Wood Plank, 4 In. x 12 In. x 12 Ft., Weight 200 Lb. 0.8

B. Steel Pipe, 3 In. Diameter, Schedule 40, 10 Ft. Long,

Weight 78 Lb. 0.4

C. Steel Rod, 1 In. Diameter x 3 Ft. Long, Weights Lb. 0.6

D. Steel Pipe, 6 In. Diameter, Schedule 40, 15 Ft. Long,

Weight 285 Lb. 0.4

E. Steel Pipe, 12 In. Diameter, Schedule 40, 15 Ft. Long,

Weight 743 Lb. 0.4

F. Utility Pole, 1372 In. Diameter, 35 Ft. Long,

Weight 1490 Lb. 0.4

G. Automobile, Frontal Area 20 Ft.^, Weight 4000 Lb. 0.2

I

I VII-13
TABLE IV - TYPICAL TORNADO MISSILES

BASED ON TVA TOPICAL REPORT

Horizontal Velocity

Ft./Sec.

A. Wood Plank, 4 In. x 12 In. x 12 Ft, Weight 200 Lb. 368

B. Steel Pipe, 3 In. Diameter, Schedule 40, 15 Ft. Long,

Weight 115 Lb. 268

C. Steel Rod, 1 In. Diameter x 3 Ft. Long, Weight 8 Lb. 259

D. Steel Pipe, 6 In. Diameter, Schedule 40, 15 Ft. Long,

Weight 285 Lb. 230

E. Steel Pipe, 12 In. Diameter, Schedule 40, 30 Ft. Long,

Weight 1500 Lb. 205

F. Utility Pole, 14 In. Diameter, 35 Ft. Long, Weight

1500 Lb. 241

G. Automobile, Frontal Area 20 Ft.^, Weight 4000 Lb. 100

I

VII-14
TABLE V - ACCEPTABLE METHODS FOR

SOIL-STRUCTURE INTERACTION ANALYSIS

SOIL FOUNDATION * SHALLOWLY EMBEDDED CASE

METHOD OF

SOIL-STRUCTURE

INTERACTION

ANALYSIS

ROCK * * FOUNDATION

DEEPLY

EMBEDDED

CASE t

DEEP SOIL

FOUNDATION

W/UNIFORM

PROPERTIES

DEEP SOIL

FOUNDATION

W/LAYERED

PROPERTIES

SHALLOW

SOIL

FOUNDATION

SINGLE LUMPED

MASS-SPRING

APPROACH

X X

MULTIPLE LUMPED

MASS-SPRING

APPROACH t ^ X X X X

FINITE ELEMENT

APPROACH t + X X X X X

* SOIL FOUNDATION MEANS THE DEPTH OF SOIL BETWEEN THE BOTTOM OF THE FOUNDATION

SLAB AND THE BASE ROCK.

** A MEDIUM FOR WHICH THE SOIL-STRUCTURE INTERACTION EFFECT IS NEGLIGIBLE OR ALTERNATELY, A MEDIUM WITH A SHEAR WAVE VELOCITY GREATER THAN OR EQUAL TO 3500 fps.

t DEEP EMBEDMENT: ACTUAL EMBEDDED DEPTH >15% OF THE LEAST BASE WIDTH OR OTHER

APPROPRIATE VALUE TO BE JUSTIFIED.

tt OR EQUIVALENT.

VII-15
TABLE VI - DAMPING VALUES

(PERCENT OF CRITICAL DAMPING)

STRUCTURE OR COMPONENT

OPERATING BASIS

EARTHQUAKE OR V2 SAFE

SHUTDOWN EARTHQUAKE^

SAFE SHUTDOWN

EARTHQUAKE

Equipment and Large-Diameter Piping Systems

Pipe Diameter Greater Than 12 Inches 2 3 Small-Diameter Piping Systems, Diameter Equal to

or Less Than 12 Inches 1 2 Welded Steel Structures 2 4

Bolted Steel Structures 4 7 Prestressed Concrete Structures 2 5 Reinforced Concrete Structures 4 7

1 . In the dynamic analysis of active components as defined in Regulatory Guide 1 .48, these values should

also be used for SSE. |

! 2. Includes both material and structural damping. If the piping system consists of only one or two spans \ with little structural damping, use values for small diameter piping.

j

I

I

I

I

VII-16 !
TABLE VII - SEISMIC INSTRUMENTATION REQUIREMENTS

INSTRUMENTATION

TRIAXIAL

TIME-HISTORY

ACCELEROGRAPH

TRIAXIAL RESPONSE

SPECTRUM RECORDER

TRIAXIAL PEAK

ACCELEROGRAPH SEISMIC SWITCH

LOCATION SSE

0.3G Or LESS OVER

0.3G

0.3G

Or LESS OVER

0.3G 0.3G Or LESS OVER

0.3G

0.3G Or LESS OVER

0.3G

1. Free Field 1 1

II. Inside Containment

Basement

At Elevation Reactor Equipment Sup. Reactor Piping Sup. Reactor Equipment

Reactor Piping

1* **

1

1* **

1

1*

1***

1*

1

1

1

1

1* 1*

1***

III. Outside Containment

Cat. 1 Structure

Cat. 1 Equip. Sup.

Cat. 1 Piping Sup.

Cat. 1 Equipment

Cat. 1 Piping

1 11***

1

1

1 r**

1

1

* Control Room Readout

** May be Omitted if Soil-Structure Interaction is Negligible. *** Denotes Either of the Two Locations.

VII-17
FIGURE 1 - CONTAINMENT FOR TYPICAL

BOILING WATER REACTOR

VII-18
FIGURE 2 - CONTAINMENT FOR TYPICAL

PRESSURIZED WATER REACTOR

VII-19
80 90 100 O^^Esas—ss 500

FIGURE 3 - BASIC WIND SPEED IN MILES PER HOUR

100 YEAR MEAN RECURRENCE INTERVAL

(ANNUAL EXTREME FASTEST - MILE SPEED 30 FEET ABOVE GROUND)

VII-20
WIND

Actual

Pressure

Average

Pressure

FIGURE 4 - WIND PRESSURE DISTRIBUTION ON WALLS

VII-21
FIGURE 5 - TORNADO INTENSITY REGIONS

VII-22
Plan View

Structure or Element

D

C

FIGURE 6 - DISTRIBUTION OF ROTATIONAL WIND SPEED

AND PRESSURE DROP RELATIVE

TO THE CENTERLINE OF THE TORNADO

VII-23
OPENING IN WINDWARD WALL

FIGURE 7 - WIND PRESSURE DISTRIBUTION

DUE TO OPENINGS IN THE STRUCTURE

VII-24
FIGURE 8 - FREE FIELD R.G.1.60 AND 60% OF R.G.1.60

COMPARED TO BASEMAT RESPONSE

AFTER DECONVOLUTION WITH ± 40 PERCENT

VARIATION OF SOIL PROPERTIES

VII-25
FIGURE 9 - SOIL-STRUCTURE SYSTEM REPRESENTED BY

AXI-SYMMETRIC FINITE ELEMENTS

VII-26
FIGURE 10 - A LUMPED MASS MODEL

OF STRUCTURE-FOUNDATION SYSTEM

VII-27
±15%

0.1 0.2 0.3 0.4 0.5 1 2 3 4 5 10 20 30 40 50 100 FREQ (CPS)

FIGURE 11 - FLOOR RESPONSE SPECTRUM CURVE

VII-28
OUTLINE OF BASIC PHILOSOPHY AND PRACTICES

OF ASEISMATIC DESIGN

FOR NUCLEAR FACILITIES IN JAPAN

MAKOTO WATABE

YORIHIKO OHSAKI

Chief, USEE

Building Research Institute

Ministry of Construction

Japan

ABSTRACT

In Japan, consistent standard specifications or guide lines for the aseismatic design

of nuclear facilities has not officially been established. Only recently, in March 1975,

has the Regulatory Guide of Design Earthquake Ground Motions for Nuclear Power Facilities

been proposed. The proposed Regulatory Guide has already had an effect in a practical

sense in spite of its interim nature. It is the purpose of this paper to briefly summarize

the design philosophies and procedures of this aseismatic design currently being used in

practice in Japan.

Key Words: Design Guides, Nuclear Facilities.

1 . Foreward

The Aseismic Design Committee was established on September 25, 1974, at the 129th meeting of the Examination Committee on Reactor Safeguards for the purpose of reviewing the

fundcimental concept underlying the articles related to aseismic design of nuclear power

facilities in "the Regulatory Guide for Safety Design of Light Water Reactors", of the

Japan Atomic Energy Commission, which has been in effect since April, 1970.

Since the first meeting on October 30, 1974, five meetings have since been held, and

a tentative conclusion has been arrived at recently on the regulatory guide of earthquake

ground motions for aseismic design of nuclear power facilities as is shown in this Interim

Report

.

The Aseismic Design Committee is presently continuing to investigate matters to be

attended to in the practical application of this regulatory guide, and the results will

be appended as a commentary to the regulatory guide afterwards

.

Further, "Aseismic Design Classification of Nuclear Power Facilities" and "Load

Combinations and Evaluation of Stresses" will be investigated and, upon completion,

the final report will be published as "Regulatory Guide for Aseismic Design of Nuclear

Power Facilities" (tentative title) including the "Regulatory Guide of Design Earthquake

Ground Motions for Nuclear Power Facilities".

The design earthquake ground motions [1] for aseismatic design of nuclear power

facilities shall be determined based on earthquake motions at the surface of the base

VII-29
stratum (1,2) in the proposed site. Where the surface of the base stratum remarkably undulates, is excavated, or is overlain by surface layers (1,3), the effects thereof shall be

taken separately into account.

The design earthquake ground motions at the free surface (1,4) of the base stratum

(hereinafter referred to as "the basic design earthquake ground motions") in the proposed

site shall be determined in accordance with fundamental concepts as described in the

following items

1. The basic design earthquake ground motions are classified into SI and S2 depending

upon the intensity.

(a) The basic design earthquake gro\md motions Si are such ground motions as considered to be caused at the free surface of the base stratum in the proposed site by

the maximum design earthquakes.

The maximum design, ear thquake is referred to that which causes the free surface of

the base stratum, in the proposed site, the most unfavorable ground motions among earthquakes which are expected to occur.

(b) The basic design earthquake ground motions S2 are such ground motions as considered to be caused at the free surface of the base stratum in the proposed site by

the extreme design earthquakes.

The extreme design earthquakes, is referred to that which causes the free surface of

the base stratum, in the proposed site, the most unfavorable ground motions among earthquakes which are expected to occur.

2. The earthquakes causing the basic design earthquake motions, near to distant

earthquakes shall be considered.

3. The basic design earthquake ground motions shall be specified in terms of the

following items

(a) Maximum velocity amplitude

(b) Wave shape

(c) Duration

4. The maximum velocity amplitude (5) , the wave shape (6) , and the duration (7) of

the basic design earthquake ground motions shall be evaluated taking the following items

into consideration:

(a) Magnitudes, epicenters, hypocenters, after-shock areas, after-shock volumes

(8), the highest intensities (or estimated values thereof) of ground motions, and situations of damage (including damage to structures, overturning of tombstones, etc.) of

historic earthquakes which have affected the proposed site and neighboring areas.

(b) Statistical expectation (9) of the intensity of ground motion based on the

past destructive earthquakes' data.

(c) Magnitudes and distances from the centers of energy release to the proposed

site of the design earthquakes.

i) The magnitude of the maximum design earthquakes shall be determined

primarily upon activities of past earthquakes which have influenced the proposed site,

also considering the state (10,11) of active faults (12) existing nearby and the seismoVII-30
tectonical structures (13) . The magnitudes of the extreme design earthquakes shall be determined based upon

maximum values (13) potentially pertaining to the seismo-tectonical structures and the

maximum scales of active faults existing nearby.

ii) The distance from the centers of energy release of the maximum or extreme design earthquakes shall be determined in consideration of the locations of nearby

active faults, the centers of energy release of past earthquakes, and the seismo-tectonical

structures

.

(d) Data from instrumental observations during past earthquakes, and the results

of instrumental observations at the proposed site and lithological surveys of strata underlying the proposed site.

5. The basic design earthquake motions (14,15) shall be expressed as time histories

and/or smoothed design response spectra, collectively considering the above paragraphs a,

b, c, and d. As the design time histories,

(a) records of past earthquake ground motions at the proposed site or adequate

modifications therof,

(b) records of past strong-motion earthquakes under similar conditions or

adequate modifications therof, or

(c) simulated (artifically synthesized) ground motions may be used. In any case,

however, careful attention should be paid to the compatibility (16) between the design

time histories and the design response spectra.

Definitions and Reference Data

a) Fig. 1 DEFINITION'S AMO PxSZPJESCil

Fig.l

FREE SURFACE OF BASESTRATi:y

JRE SURFACE D LAYER

BUILD r; ^STRUCTU...

BASE STRATUM BASE strati:'!

« DESIGN EQ. GROUriD MOTIONS A BASIC DESIGN EQ. GROUND YOTIOfiS

b) Base Stratum: a firm, intact (not significantly fissured nor weathered) rock

which was formed in general in the Tertiary or earlier era.

c) Surface Layer: geologically newer soil layer or formation, or weathered rock

overlying the base stratum

d) Free Surface of Base Stratvmi: a horizontal, flat surface of the base stratum

extending over a considerable area, above which neither surface layers nor structures

VII-31
are assumed to be present. Where such surface cannot really be identified within or around the proposed site, it should reasonably be assumed.

e) The maximum velocity amplitude of the basic design earthquake ground motions may

be evaluated by

^j^0.6LM-Plog,oX-3 (1)

where vmaxM

P

Q

X

max. velocity amplitude (cm/sec.) at the free surface of the base stratum

magnitude in Gutenberg-Richter scale

1.66 + 3.60/X

0.631 + 1.83/X

hypocentral distance (Km) f ) Envelope fianction defining the wave shape may be assumed as shown in Fig. 2

Fig. 2

0.04

002

5.5 65 7.5 Magnitude

8i

g) The duration of the basic design earthquake ground motions may be estimated by

t^= 10 (M-2. 5)/3. 23

(2)

where t^: duration (sec.)

M: magnitude in Gutenberg-Richter scale

h) When the after-shock volume is approximately assumed as a sphere, its radius may

be estimated by

D = 10 0.353M-1.134

(3)

VII-32
where D: radius of spherical after-shock volume (km)

M:' magnitude in Gutenberg- Richter scale

i) Knowing the magnitude M and the epicentral distance R or the hypocentral distance

X from a site with respect to an earthquake, it is possible to approximately estimate the

intensity scale I, for instance, by an empirical formula

I = 2(M-lnR)-0.00183R-0.307 R > 100 kn

for > I = 2M-0.86861nX-0.01668X-3.9916 R < 100 km

Therefore, from earthquakes' data for the past T years, the probability distribution

fQI = FtD/T T = 0,1,2, ,7 (5)

may be evaluated, where F(I) represents the number that the site experienced the earthquakes of intensity scale I. Usually, as shown in Fig. 3 (a) , the probability distribution F (I) approximately takes an exponential form. Thus, the integral

( fCDdi

represents the probability per one year that the site will be influenced by an earthquake

of intensity scale larger than I. From Fig. 3(b), where the integral values are plotted

against I, the statistical expectancy of the intensity scale, I^, can be found corresponding to any return period of r years. Since the relation between intensity scale I and acceleration a on the surface of medium-staff soil deposit may be expressed approximately by

1/2

a = 0.45 X 10 gal (6)

the statistical expectancy of the acceleration can be obtained by substituting for I

in eq. (5) . In a similar way, the statistical expectancy of the velocity of ground

motions can be determined based on the same past earthquakes' data by using eq. (1) instead

of eqs. (4) and (6) .

j) Status of Active Fault: 1) length, 2) activity (average displacement velocity),

3) pattern of movement and displacement at earthquake occurrence, and 4) the year of the

last movement (blank period) of the active fault

k) If, with respect to an active fault.

average displacement velocity (m/year)

displacement at earthquake occurrence (m) length (km)

magnitude of caused earthquake

period of earthquake occurrence (year)

VII-33
there are the relations

log D = 0.60M-3.91 (7) log L = 0.60M-2.91 (8) R = D/S (9) Eqs. (7) and (8) have been obtained from the experience of Japanese island earthquakes,

while from world-wide data they become

log D = (0.55 + 0.07)M - (5.71 + 0.50) (10)

log L = 1.32M - 7.99 (11) It is frequently recognized that there is the specific magnitude pertaining to a specific active fault or its localized segment.

1) Active Fault: a fault which evidently moved in the Quarternary era (within last

1,000,000 years) and also is capable of causing earthquake in the future .-

H.ZO-'.Ok.m

VII-34
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1

VII-35
TORNADO-BORNE MISSILES

by

James F. Costello

U.S. Nuclear Regulatory Commission, Washington, D.C.

ABSTRACT: Nuclear power plants are designed to withstand the effects

of severe environmental events, including tornadoes, without endangering public health and safety. Along with high wind velocities and

drops in pressure, an effect which must be considered is the fact that

objects of varying sizes are displaced by tornadoes. Predictions of

what objects are likely to fly and at what velocities are difficult.

Knowledge of the actual windfields present in a tornado is sketchy as

is information about the aerodynamics of bluff bodies. This paper

outlines methods which have been developed in order to make conservative

estimates of the effects of tornado-generated missiles.

i

I

KEY WORDS: Structural Engineering; Dynamics Tornadoes ; Missiles;

Trajectories; Power Plants.

li

VII-36
INTRODUCTION

To ensure the safety of nuclear power plants in the event of a

tornado strike it is required that, in addition to the direct action of

the wind and of the moving ambient pressure field, the designer consider

the impact of tornado-borne missiles, i.e., of objects moving under the

action of aerodynamic forces induced by the tornado wind. It is, therefore, necessary that estimates be made of the speeds attained by potential

missiles under tornado wind conditions specified for the design of nuclear

power plants.

In order to make these estimates, some assumptions are necessary.

These assumptions include:

a) the initial conditions of the problem, i.e., the initial position

of the object with respect to the ground and to the tornado center,

and the initial velocity of the object, b) the detailed features of

the wind flow field, and c) the aerodynamic characteristics of the

object (which, in most cases of practical interest, is a bluff body).

Differences between the various sets of basic assumptions used in estimating tornado-borne missile velocities may be ascribed, in part, to the

probabilistic nature of the problem. For any given tornado wind field,

the initial conditions constitute a set of random variables. More importantly, however, such differences are a consequence of serious uncertainties regarding both the structure of the tornado flow and the

aerodynamic behavior of the potential missile.

VII-37
MODEL FOR THE TORNADO-BORNE MISSILE MOTION

The motion of an object may be described in general by solving a

system consisting of three equations of balance of momenta and three

equations of balance of moments of momenta. In the case of a bluff body,

one major difficulty in writing these six equations is that the aerodynamic forcing functions are not known.

It is possible to measure in the wind tunnel aerodynamic forces and

moments acting on a bluff body under static conditions for a sufficient

number of positions of the body with respect to the mean direction of the

flow. On the basis of such measurements, the dependence of the forces and

moments on position, and corresponding aerodynamic derivatives, can be

obtained. Aerodynamic forces and moments can then be calculated following

the well-known pattern used in airfoil theory; for example, if an airfoil

has a time-dependent vertical motion h(t) in a uniform flow with velocity U,

and if the angle of attack is a = const, the lift coefficient is

This procedure for calculating aerodynamic forces and moments is valid

if the quasi -steady assumption is acceptable and if the body concerned

behaves aerodynamical ly like an airfoil - i.e., if the body is streamlined and if no flow separation occurs. However, in the case of unconstrained bluff bodies moving in a wind flow the validity of such a procedure

remains to be demonstrated.

In the absence of a satisfactory model for the aerodynamic description

of the missile as a rigid (six-degrees-of -freedom) body, it is customary

VII-38
to resort to the alternative of describing the missile as a material point

acted upon a drag force

D=l/2pCpA 17^-7^ I (2)

where p = air density, = wind velocity, Vj^ = missile velocity, A is a

suitably chosen area and is the corresponding drag coefficient.

This model is reasonable if, during its motion, the missile either (a)

maintains a constant or almost constant attitude with respect to the relative

velocity vector v^ - w^, or (b) has a tumbling motion such that, with no

significant errors, some mean value of the quantity C^A can be used in the

expression for the drag D. The assumption of a constant body attitude with

respect to the flow would be credible if the aerodynamic force were applied

at all times exactly at the center of mass of the body--which is highly

unlikely in the case of a bluff body in a tornado flow--, or if the body

rotation induced by a non-zero aerodynamic moment with respect to the center

of mass were inhibited by aerodynamic damping forces intrinsic in the body--

fluid system.

Wind tunnel test results suggest that no stabilizing effect by the

flow can generally be expected to inhibit the rotation of bluff bodies.

The assumption that potential tornado-borne missiles will tumble during

their motion appears therefore to be reasonable.

Assuming that Eq. 2 is valid and that the average lift force vanishes

under tumbling conditions, the motion of the missile viewed as a one-degreeof-freedom system is governed by the relation:

VII-39
dv.M

= - 1/2 p (Vk. - V ) - gk (3) dt

V,M

- V, m w

where g

= acceleration of gravity, k = unit vector along the vertical axis

and m = mass of missile. It follows from Eq. 3 that for a given flow field

and initial conditions, the motion depends only upon the value of the

parameter C^A/m. Experimental studies aimed at finding reasonable ways to

estimate an "effective" C^A value are now being carried out at Colorado

State University.

The subject of tornado-borne missiles is still under active investigation by the U.S. Nuclear Regulatory Commission and the Electric Power

Research Institute as well as by a number of architectural and engineering

firms. The following section gives a list of references. The list is by

no means exhaustive. But it includes, as well as some more recent work,

some papers of historical interest.

VII-40
REFERENCES

F. C. Bates, A. E. Swanson (1967), "Tornado Considerations for Nuclear

Power Plants," Transactions , American Nuclear Society, Vol. 10,

pp. 712-713.

A. K. Bhattacharyya, R. C. Boritz and P. K. Niyogi (1975), "A Consistent

Method of Determining Effects of Tornado Missiles on Nuclear Power Plant

Design," Proceedings, Second ASCE Conference on Structural Design of

Nuclear Plant Facilities.

J. R. Eagleman, V. U. Muirhead, and N. Willems (1975) Thunderstorms , Tornadoes and Building Damage , Lexington Books, Lexington, Mass.

W. H. Hoecker, "Wind Speed and Air Flow in the Dallas Tornado of

April 2, 1975," Monthly Weather Review , Vol. 88, No. 5.

W. Huang and J. M McLaughlin (1975), "Tornado Generated Missiles,"

Proceedings, Second ASCE Conference on Structural Design of Nuclear

Plant Facilities.

R. A. James, E. G. Burdette, and C. N. Sun (1975), "Generation of Missiles

by Tornadoes," Proceedings, Second ASCE Conference on Structural Design of

Nuclear Plant Facilities.

A.J.H. Lee (1975), "Trajectory of Tornado Missiles and Their Design

Parameters" Proceedings, Second ASCE Conference on Structural Design

of Nuclear Plant Facilities.

E. H. Markee, J. G. Beckerley, and K. E. Sanders (1974), Technical Basis

for Interim Regional Tornado Criteria , WASH-1300 (UC-11), U.S. Atomic

Energy Commission, Office of Regulation.

J. R. McDonald, R. C. Mehta, and J. E. Minor (1974), "Tornado-Resistant

Design of Nuclear Power-Plant Structures," Nuclear Safety , Vol. 15, No. 4,

D. F. Paddleford (1969) Characteristics of Tornado Generated Missiles , Report WCAP - 6897, Westinghouse Electric Corp., Pittsburgh, Pennsylvania.

G. H. Redman (1976), "Wind Field and Trajectory Models for TornadoPropelled Objects," EPRI-308, Electric Power Research Institute, Palo

Alto, CA.

E. Simiu and M. R. Cordes (1976), "Tornado-Borne Missile Speeds," NBS

ID-1050, National Bureau of Standards, Washington, D.C.

VII-41

Structural Damage to Bridges

Resulting From

the Guatemala Earthquake

James D. Cooper

Federal Highway Administration, Washington, D. C.

ABSTRACT

The Guatemala earthquakes of February 4 and 6, 1976, caused

severe economic hardships because of highway bridge failures and

damage. The damage to three major bridges, Agua Caliente, LaAsuncion,

and Incienso is described. A general discussion of damage to bridges

and the roadv^ay along a major highway, the Atlantic Highway, (Route

CA9) , is also presented.

KEYWORDS: Earthquake damage; Bridges; Guatemala earthquake

VIII-1
INTRODUCTION

The Guatemala earthquakes of February 4 and 6, 1976, caused extensive damage to two

major highway bridges, LaAsuncion located in the northeast part of Guatemala City, and Agua

Caliente located approximately 17 miles (27 km) northeast of Guatemala City on Route CA9.

A third major bridge, Incienso, located in the northwest section of Guatemala City suffered

some damage- Bridges within Guatemala City, with the exception of LaAsuncion and Incienso,

performed well, suffering little or no damage. Some evidence of displaced bearing pads,

settled aprons on abutment slopes and minor impacting at abutments was evident on a few of

these structures. Bridge damage outside Guatemala City was generally confined to the area

of approximate surface faulting, Figure 1, along the Motagua River northeast of the city.

Eleven bridges along the 184 miles (297 km) of Route CA9 between Guatemala City and Puerto

Barrios on the east coast experienced varying degrees of damage, mostly to bearings.

The earthquake caused additional extensive damage to mostly non-engineered structures

situated along the Motagua River Valley. The approximate location of surface faulting

(left lateral), the shaded area shown in Figure 1, follows the Motagua River and Route CA9

,

and extends approximately 149 miles (240 km) from just east of Gualan, to northwest of

Guatemala City. A maximum 55 inches (140 cm) of fault displacement was reported north of

Guatemala City.

The National Earthquake Information Service reported the February 4 earthquake as

being 7.5 on the Richter scale with the epicenter at latitude 15.27°N, longitude 89.25°W,

99 miles (160 km) northeast of Guatemala City between the towns of Gualan and Los Amates.

Reportedly, ground shaking in Guatemala City lasted between 20 and 30 seconds. Two seismoscope records were obtained and no strong-motion accelerograph records were recorded. The

February 6 earthquake is reported as being a 5-3/4 to 6 magnitude event with a preliminary

epicentral location of 14.3°N, 90.5°W. Most damage was caused by the February 4 event.

Specific discussion of damage to the Agua Caliente, LaAsuncion, and Incienso bridges

is presented, followed by a brief general discussion of damage to the bridges and roadway

along Route CA9 , the Atlantic Highway. The bridge inspections were made in cooperation with

the second reconnaissance team fielded by the Earthquake Engineering Research Institute

March 1-6, 1976, one month following the earthquakes.

VII 1-2
RIO AGUA CALIENTE BRIDGE

The most severely damaged bridge was the important Rio Agua Caliente Bridge, located

approximately 17 miles (27 km) northeast of Guatemala City on Route CA9, the Atlantic Highway. This highway serves as the major transportation link between Guatemala City and the

economically strategic port facility at Puerto Barrios, on the east coast of Guatemala.

A simple plan and elevation view of the five-span, simply supported steel plate girder

bridge is shown in Figure 2. The bridge was designed in accordance with the 1953 AASHO

"Standard Specification for Highway Bridges" using an earthquake factor of 10% dead load.

The span lengths from the west abutment are 148 feet (45 m) , 98 feet (30 m) , 148 feet (45 m

98 feet (30 m) , 98 feet (30 m)

.

The straight steel superstructure spans were designed using two 8-foot (2.4 m) deep,

built-up exterior plate girders, two interior 18 WF 50 stringers with spirals for composite

deck action, 30 WF 108 floor beams, and bottom lateral bracing. The concrete deck slab,

curved in plan with 8.5% superelevation, is 6.5 inch (16.5 cm) thick with a 2 inch (5.1 cm

asphalt concrete overlay. Total deck width, including parapets, is 33.7 feet (10.25 m)

.

The superstructure sits on four reinforced concrete pier bents which vary in height

between 72 feet (21.9 m) and 76 feet (23.2 m) above maximum high water level. The pier

bents are tied into pedestals which sit on spread footings. Total depth of pedestals and

spread footings vary between 10 feet (3.1 m) and 33 feet (10 m) below maximum high water

level. The end spans sit on abutment walls on spread footings.

Figure 3 shows a general view of the bridge looking south. The earthquake caused the

three center spans to fall off the bearing supports. Pier bents and abutments appeared to

be undamaged with the exception of minor spalling where the falling girders had impacted

against the columns and tie beams. Extremely minor hairline cracks were noted around the

base of the columns of pier #2.

The fallen second and fourth spans appear to have initially lost support on the south

side of the bent; i.e., along the outside radius of the deck curvature. As the south side

of the superstructure fell, the bearings on the north side of the bent gave way, thus allow

ing the spans to rotate down to their final resting position.

Significant longitudinal pier motion occurred, propelling the girder of span #2

approximately 7 feet (2 m) beyond the centerline of pier bent #2, as seen in Figure 4.

VIIl-3
Additional evidence of pier rocking can be seen in Figure 5. Note the soil at the base of

pier #2 which has been compressed leaving approximately 1/2 inch (1.2 cm) void between soil

and foundation. This could account for approximately 1.5 inches (3.8 cm) of longitudinal

displacement about the centerline at the top of the pier bent, assuming simple rocking about

the base of the foundation. Figure 6 shows the collapsed center span and roadway which slid

straight down pier #3. Note the only visible damage to pier #3 is in the form of spalled

concrete from girder impact above the lower tie beam on the left column and on the upper

right side of the lower tie beam.

A view looking east from the base of pier #3 through twisted girders, stringers, floor

beams, and lateral bracing is shown in Figure 7.

The bridge collapsed, principally because of the bearing details used and the lack of

longitudinal and transverse superstructure restraint at the supports. Figure 8 shows a

fallen expansion rocker at the base of pier #3. The rocker. which supported a 148 foot (45 m)

span, is 11 inches (28 cm) by 17 inches (43 cm) at the base and 13.5 inches (34 cm) high.

Figure 9 shows the fixed bearings in place atop the free standing piers. An expansion

rocker can be seen resting on the tie beam in the foreground. The north girder appears to

have walked approximately 5 inches (13 cm) to the south on the expansion bearing. It is

speculated that had appropriate hinge restrainers been in place, thus tying the structure

together, collapse probably could have been avoided.

Figure 10 shows the north girder and fixed bearing detail at the east abutment. Minor

extension of anchor bolts was noted, otherwise no significant damage was visible at the east

abutment. There was no indication of loose soil or stones having been expelled from weep

holes in the abutment or stone wall and no cracking was observed in the stone wall.

Figure 11 shows the north girder fixed bearing detail at the west abutment. The bearing

support is inclined approximately 8°. Note the anchor bolt extension and the space between

the abutment bearing ledge and bearing plate. Figure 12 shows the north girder resting on

the expansion bearing which is inclined at approximately 30°. A recommendation was made to

place temporary shoring under the girders at the west abutment and pier #1 in view of the

extremely unstable position of the span.

The pier bents and abutments appear to be structurally undamaged and plans are being

made to reuse them when the superstructure is rebuilt. Particular attention will be devoted

VIII-4
to the bearings in the design of the replacement structure,

Iramediately following the inspection of the collapsed Agua Caliente Bridge, the U. S.

Army Corps of Engineers moved in to clear the site and start construction of a temporary

Bailey Bridge.

It is interesting to note that a curved railroad steel truss bridge located approximately 1 mile (1.6 km) north of the Agua Caliente, oriented in the same direction as Agua

Caliente, suffered no apparent damage. Although the railroad structure was inaccessible

at the time of the Agua Caliente inspection, it was operational as evidenced by the passage

of a freight train. It is believed that the rails probably provided adequate structural

continuity to resist the earthquake motions.

LA ASUNCION BRIDGE

The LaAsuncion Bridge, located in the northeast section of Guatemala City, is a threespan, continuous plate girder bridge that was severely damaged during the earthquake. The

bridge is under the jurisdiction of the local city government. No plans are available;

however, personnel from the Guatemala Department of Roads, Ministry of Communications and

Public Works provided the general plan and elevation views of the bridge, shown in Figure 13.

The span lengths are 148 feet (45.1 m) , 196 feet (59.9 m) , and 148 feet (45.1 m) . The

deck is 48 feet (14.6 m) wide. The piers are single, reinforced concrete columns with cap

beams. The west pier is 86 feet (26 m) above grade while the east pier is 82 feet (25 m)

above grade. Foundation type is unknown.

During the earthquake, the superstructure rotated, in plan, about the east pier causing

the bearing supports to dislodge. The west end of the bridge deck displaced 39 inches

(100 cm) laterally to the south while the east end displaced 20.5 inches (52 cm) laterally

to the north, as shown by the dashed line in the plan view of Figure 13. The lateral offset

at the west abutment is shown in Figure 14 between the once continuous median delineator

on the approach roadway and deck.

The approach roadway itself was badly cracked, Figure 15, and damage to the west

abutment and wing walls indicated the presence of extremely high backfill forces. Figure 16

shows abutment wall failure. The pipe in the foreground is a drain scupper. Figure 17

VIII-5
shows a lateral foundation offset of 2.5 inches (6.4 cm) relative to the south wing wall.

The view looks down along the south wing wall which is shown at the right of the photograph.

The evidence indicates the complete failure of the west abutment. There was, however, no

evidence of foundation failure at the east abutment, and the piers, as viewed from both

abutments, appeared undamaged.

The bridge deck dropped 8.3 inches (21 cm) as it displaced and dislodged from the rocker

bars at the west abutment, Figure 18. Figure 19 shows the bottom chord of the end diaphragm

at the west abutment resting on a bearing plate with shear key and a dislodged rocker bar to

the right. A view of a failed bearing detail at the east abutment ledge is shown in

Figure 20.

Figure 21 shows a general view of the bridge. Note a missing bearing plate under the

north (right) girder at the east pier in the foreground. The only noticeable damage to the

steel superstructure was at the west abutment, as shown in Figure 22. The bottom lateral

bracing and end diaphragm were slightly buckled. Figure 23 shows the elbow of a 10-inch

(25 dm) water pipe at the southeast corner of the bridge with no apparent damage. The pipe

could not be traced and there was no evidence of seeping water along the embankment or

apron at the east abutment.

The LaAsuncion Bridge appears to be structurally sound and can be jacked back into place

once the damage to the west abutment has been corrected.

INCIENSO BRIDGE

The Incienso Bridge, Figure 24, located in the northwest section of Guatemala City,

was constructed in 1972. The bridge is of interest because it incorporates seismic hinge

and abutment restrainers and although minor damage occurred, remained in service following

the earthquake. The performance of the eight-span concrete bridge provided an indication

that the incorporation of new seismic design techniques in new construction can provide

adequate structural resistance against major earthquakes.

The three main spans were twin segmental, precast, prestressed concrete box girders,

203 feet (62 m) , 400 feet (122 m) , 203 feet (62 m) in length constructed by the balanced

cantilever method. The main span is supported on reinforced concrete piers 262 feet (80 m)

and 230 feet (70 m) in height. The five side spans, each 92 feet (28 m) in length, are

VIII-6
constructed with seven prestressed concrete I-beams supported on reinforced concrete pier

bents. The total deck width is 82 feet (25 m)

.

The main three spans suffered no visible damage. Figure 25 shows evidence of compressional failure of the pavement at the west abutment due to impacting. Although not quite as

visible in the photograph, there is evidence of a 1 inch (2.5 cm) vertical displacement of

the deck. There is also evidence of minor deck impacting between the prestressed I girder

spans

.

The generally good performance of the bridge is attributed to the use of seismic

restraint mechanisms ; i.e., hinge restrainers, which tied the prestressed I girders together.

Figure 26. The restrainer caps, which pass through the concrete end diaphragms of the

girders, can be seen in Figure 27. A total of 12 restrainers are placed across each joint.

Each restrainer cable is prestressed to 45 metric tons and has a 63 metric ton tensile

strength. The only restrainer failure occurred across the west abutment, where all 12

cables failed. Figure 28 shows a "seismic block" i.e., a concrete block tied to an abutment

or pier through which the restrainer cable passes, at the west abutment between the end

diaphragm on the left and abutment wall on the right. The restrainer cable failed in

tension and shear at the point which is flush against the girder end diaphragm. (The cable

in the photograph is not the restrainer cable.)

The failure of the restrainer cables at the west abutment resulted in the walking of

bearing pads including the dislodging of one pad, shown in Figure 29. The temporarily

blocked girder on the concrete bearing plate where the bearing pad was removed is shown in

Figure 30.

Additional restraint was provided through the use of abutment tie back cables, shown

in Figure 31. It was reported that there are seven tie back cables across each abutment,

although only five cable caps were noted across the abutment wall. Each cable is reported

to have a tensile strength of 180 metric tons. There was no evidence of abutment tie back

cable failure.

THE ATLANTIC HIGHWAY - CA9

The Atlantic Highway is one of the most economically strategic highways in Guatemala,

extending from Guatemala City to the port facility at Puerto Barrios, 184 miles (297 km) to

VII 1-7
the north east. The road was closed following the earthquake due to severe landslides,

damaged bridges and road sloughing, thus interrupting the flow of commercial traffic.

Bridges

There are approximately 42 bridges along the Atlantic Highway, including the Agua

Caliente. Eleven suffered varying degrees of damage. Typical types of structures include

deck, through, and pony trusses; steel girders with composite acting decks; and prestressed

and reinforced concrete girder bridges. The predominant types are simple span trusses

supported on rocker bearings. Excluding the Agua Caliente bridge, most damage included

lateral displacement of the decks, impacting of deck against abutment walls causing cracking and spalling, and tipping or falling of rocker bearings.

Figure 32 shows the damaged Guastatoya Bridge, 44 miles (71 km) northeast of Guatemala

City. The structure, a three-span deck truss with approximately 120-foot (37 m) spans, was

closed to traffic. Rocker bearings tipped, note the differential vertical displacement at

the right pier, and the right span at the abutment displaced laterally, almost falling off

its bearing.

Figure 33 shows the Teculutan Bridge, which is typical of many that experienced damage

along the Atlantic Highway. Minor abutment pounding was noted at this site. Typical damage

to truss-structures is shown in Figure 34. Keeper plates on flange bearing shoes provided

little lateral resistance to earthquake forces, as evidenced in the photograph. In spite of

the widespread damage, all bridges along the highway were open to traffic except the Agua

Caliente and Guastatoya structures.

Roads

The Atlantic Highway is a two-lane asphalt road 22 feet (6,6 m) wide. Significant road

sloughing occurred between Guatemala City and the Agua Caliente Bridge site, typically on

fill areas across steep erosional cuts. Figures 35 and 36 show typical sloughing along the

route. Damage was held to a minimum where retaining walls were constructed at the sides

of the road, Figure 37.

The most severe landslides, which closed the highway, occurred approximately 31 miles

(50 km) northeast of Guatemala City and extended over a distance of approximately 6 miles

VIII-8
(10 km), Figures 38 and 39. The U. S. Army Corps of Engineers was in the process of

clearing the roads and allowed limited passage over the tops of the slides. Future earthquakes can be expected to cause similar traffic disruption along CA9 because of the

unfavorable terrain.

Figure 40 shows the Motagua Fault Zone passing across the Atlantic Highway approximately 2 miles (3 km) northeast of El Progresso. The roadway across the fault zone was

totally pulverized. Note the 39.4 inch (100 cm) left lateral displacement at the fault as

measured from the projection of the white centerline in the background to white centerline

in the foreground. Figure 41 shows the fault zone passing through the cut slope at the side

of the road. There was no indication of vertical fault displacement at this site.

CONCLUSIONS

The Guatemala earthquakes vividly demonstrated the vulnerability of bearing assemblies

commonly used to support typical highway bridges, not only in Guatemala, but also in the

United States. The types of bearings which failed have not been designed to withstand the

large dynamic forces and motions imposed by earthquake. Lessons learned from this and

previous earthquakes will preclude the use of this type detail in future construction.

Thus, the critical issue is what can or will be done to improve the seismic resistance of

existing bridges which have this vulnerable detail. Retrofitting techniques have been

identified which can reduce susceptibility to damage. The most difficult problem, however,

is to convince appropriate responsible authorities that for a relatively small investment,

future major damage can be averted.

The Guatemala earthquakes provided the first full-scale test of a structure whose

design had apparently been modified to incorporate some of the lessons learned from bridge

damage in the 1971 San Fernando, California, earthquake. The performance of the Incienso

Bridge demonstrated that bridges which are properly designed, and, in particular, tied

together can withstand major earthquakes.

An attempt should be made to identify the multitude of existing bridges which are

located in seismically vulnerable areas of the United States. Importance and seismic

vulnerability factors can be developed for each so that as funding becomes available, the

structures can be retrofitted.

VIII-9
It is obviously impossible to totally prevent landslides from occurring in areas of

unfavorable terrain. However, critically important routes in rough terrain should be

identified and plans made to provide either quick opening or alternate emergency routes

following an earthquake.

REFERENCES

1, Newsletter of the Earthquake Engineering Research Institute, David J. Leeds, Ed.,

"The Guatemala Earthquake of February 4, 1976 - A Preliminary Report by the EERI

Reconnaissance Team," Volume 10, No. 2B, March 1976.

2. M. A. Sozen and J. Roesset, "Preliminary Notes on Structural Damage Caused by

Guatemala Earthquakes of 4 and 6 February 1976." A report to the Panel on Earthquakes, Committee on National Disasters of the Commission on Sociotechnical Systems,

National Research Council, February 15, 1976.

Acknowledgements

The author wishes to express his appreciation and sincere thanks

to Ing. Ernesto Argueta Pineda and his staff at the Guatemala Departmento da Puentes, Direccion General de Caminos, who provided valuable

technical data and assistance during the inspection of damaged

bridges. Official transportation, provided by Ing. Argueta, assured

prompt access to points of interest along the Atlantic Highway.

The author is also greatly indebted to the Earthquake Engineering

Research Institute (EERI) and Mr. Henry J. Degenkolb for providing

names of key contact persons while in Guatemala,

VIII-10
VIII-11
VIII-12
VIII-13
Figure 6 Agua Callente

Collapsed Center Span and Roadway

Figure 8 Agua Callente

Fallen Expansion Rocker , Base of Pier 3

Figure 7 Agua Callente - Detail of Collapsed Span at Base of Pier 3

VIII-14
Figure 9 Agua Caliente - View from East Abutment

Figure 10 Agua Caliente

East Abutment, North Girder and Bearing

VIII-15
Figure 12 Agua Caliente - Tipped Rocker, Pier 1

f

Figure 15 LaAsuncion - Cracked Approach at West Abutment

VIII-16
VIII-17
Figure 17 LaAsuncion - West Abutment Foundation Damage

Figure 18 LaAsuncion - Vertical Deck Displacement at West Abutment

j

VIII-18

I

I'
VIII-19
Figure 22 LaAsuncion - Bottom Lateral Bracing and End Diaphragm Damage

Figure 23 LaAsuncion - Undamaged Water Pipe at East Approach

I

e 24 Incienso - General View of Bridge j

i

VIII-20
Figure 25 Incienso - Pavement Cracking at West Abutment

igure 28 Incienso - Seismic Block, West Abutment

VIII-21
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VIII-22
Figure 30 Incienso' - Missing Neoprene Bearing Pad at South Girder, West

Abutment

Figure 33 CA9-Teculutan - Typical Through Truss

VIII-23
VIII-24
VIII-25
Figure 37 CA9-Roadway Retaining Wall

Figure 39 CA9-Landslides

VIII-26
VIII-27
NBS-1 14A (REV. 7-73)

U.S. DEPT. OF COMM. BIBLIOGRAPHIC DATA

SHEET

1. PUBLICATION OR REPORT NO. NBS SP-477

2. Gov't Accession

No.

3. Recipient's Accession No.

4. TITLE AND SUBTITLE ^jjjj) SEISMIC EFFECTS

Proceedings of the Eighth Joint Panel Conference of the U.S. - Japan Cooperative Program in Natural Resources, May 18-21, 1976,

National Bureau of Standards, Gaithersburg, Maryland

5, Publication Date May 1977

6. Performing Organization Code

7. AUTHOR(S)

H. S. Lew, Editor

8. Performing Organ. Report No.

9. PERFORMING ORGANIZATION NAME AND ADDRESS

NATIONAL BUREAU OF STANDARDS

DEPARTMENT OF COMMERCE

WASHINGTON, D.C. 20234

10. Project/Task/Work Unit No.

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12. Sponsoring Organization Name and Complete Address (Street, City, State, ZIP)

Same as Item 9

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15. SUPPLEMENTARY NOTES

Library of Congress Catalog Card Number; 77-600015

16. ABSTRACT (A 200-word or less factual summary of most sigriificant information. If document includes a significant

bibliography or literature survey, mention it here.)

The Eighth Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects

was held in Gaithersburg, Maryland on May 18-21, 1976. The proceedings of the

Joint Meeting include the program, the formal resolutions, and the technical

papers. The subject matter covered in the papers includes wind effects on structures and design criteria; extreme winds for structural design; earthquake

ground motions and instrumentation; seismicity and earthquake risk; seismic

effects on structures and design criteria; lessons learned from recent natural

disasters; design of nuclear reactor facilities.

17. KEY WORDS (six to twelve entries; alphabetical order; capitalize only the first letter of the first key word unless a proper name; separated by semicolons)

Accelerograph; Bridges; Buildings; Codes; Disaster; Dynamic Analysis; Earthquakes; Ground Failures; Nuclear Facilities; Seismicity; Soils; Standards;

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