FOREWORD 

This report presents the results of a laboratory and field study of composite piles for use as 

foundation elements for bridges. Two types of composite piles were studied, a fiber-reinforced 

polymer (FRP) concrete-filled shell, and a plastic pile reinforced with a welded steel cage. Both 

axial and lateral short-term load displacement behavior was studied, as well as interfaced 

mechanical properties. Axial and lateral load-displacement behavior was similar for the concrete 

control pile and the FRP pile, and the plastic pile was significantly less stiff in both loading 

modes. 

 Gary Henderson 

 Director, Office of Infrastructure Research 

and Development 

NOTICE 

This document is disseminated under the sponsorship of the U.S. Department of Transportation 

in the interest of information exchange. The U.S. Government assumes no liability for the 

contents or the use thereof. The contents of this report reflect the views of the contractor, who is 

responsible for the accuracy of the data presented herein. The contents do not necessarily reflect 

the official policy of the U.S. Department of Transportation. The report does not constitute a 

standard, specification, or regulation. 

The U.S. Government does not endorse products or manufacturers. Trademarks and 

manufacturers’ names appear in this report only because they are considered essential to the 

object of this document. 

QUALITY ASSURANCE STATEMENT 

The Federal Highway Administration (FHWA) provides high-quality information to serve 

Government, industry, and the public in a manner that promotes public understanding. Standards 

and policies are used to ensure and maximize the quality, objectivity, utility, and integrity of its 

information. FHWA periodically reviews quality issues and adjusts its programs and processes to 

ensure continuous quality improvement. 
1. Report No 2. Government Accession No. 3. Recipient's Catalog No. 

FHWA-HRT-04-043 N/A N/A 

4. Title and Subtitle 5. Report Date 

 March 2006 

 

6. Performing Organization Code 

A Laboratory and Field Study of Composite Piles for Bridge Substructures 

N/A 

7. Authors(s) 8. Performing Organization Report No. 

Miguel A. Pando, Carl D. Ealy, George M. Filz, J.J. Lesko, and E.J. Hoppe N/A 

9. Performing Organization Name and Address 10. Work Unit No. (TRAIS) 

N/A 

11. Contract or Grant No. 

DTFH61-99-X-00023 

13. Type of Report and Period Covered 

Virginia Transportation Research Council 

530 Edgemont Road 

Charlottesville, VA 22903

Final Report 

July 1999–May 2003

12. Sponsoring Agency Name and Address 14. Sponsoring Agency Code 

Office of Infrastructure R&D 

Federal Highway Administration 

6300 Georgetown Pike 

McLean, VA 22101-2296 

 

15. Supplementary Notes 

Contracting Officer’s Technical Representative: Carl Ealy, HRDI-06

 

16. Abstract 

The most commonly used pile materials are steel, concrete, and wood. These materials can degrade, and the degradation rate can be relatively 

rapid in harsh marine environments. It has been estimated that the U.S. spends over $1 billion annually for repair and replacement of 

waterfront piling systems. This high cost has spurred interest in alternative composite pile materials such as fiber-reinforced polymers 

(FRPs), recycled plastics, and hybrid materials. Because only minimal performance data have been collected for composite piles, a research 

project was undertaken to investigate (1) soil-pile interface behavior of composite piles, (2) the long-term durability of concrete-filled FRP 

shell composite piles, and (3) the driveability and axial and lateral load response of concrete-filled FRP composite piles and steel-reinforced 

recycled plastic piles by means of field tests and analyses. In addition, a long-term monitoring program was implemented at a bridge over the 

Hampton River in Virginia. 

According to laboratory rest results, values of residual interface friction angle between three pile surfaces and a subrounded to rounded sand 

were 27, 25, and 28 degrees for a FRP composite pile, the recycled plastic pile, and the prestressed concrete pile respectively, while the 

values of residual interface friction angle between these piles and a subangular to angular sand were 29, 29, and 28 degrees for the FRP 

composite pile, the recycled plastic pile, and the prestressed concrete pile, respectively. Regarding durability of FRP composite piles, it was 

found that moisture absorption caused degradation of strength and stiffness of the FRP shells, but that freeze-thaw cycles had little effect. 

Analyses indicate that FRP degradation due to moisture absorption should have minimal impact on axial capacity of the FRP composite piles 

because most of the axial capacity is provided by the concrete infill; however, FRP degradation has a larger effect on lateral capacity because 

the FRP shell provides the capacity on the tension side of the pile. The field tests demonstrated that there were not major differences in 

driveability of the FRP composite pile, the recycled pile, and the prestressed concrete pile. In static load tests, the FRP composite pile and 

prestressed concrete pile exhibited similar axial and lateral stiffness, and the plastic pile was significantly less stiff. Conventional static 

analyses of axial load capacity, axial load versus settlement, and lateral load versus deflection provided reasonable predictions for the 

composite piles, at least to the levels of accuracy that can be achieved for more common pile materials. The long-term monitoring program 

has been implemented for an FRP composite pile and a prestressed concrete pile so that their load-transfer performance can be compared over 

time. The long-term monitoring is being done by Virginia DOT.

17. Key Words 18. Distribution Statement 

Composite piles, driven piles, dynamic analysis 

of piles, FRP piles, recycled plastic piles, 

CAPWAP

No restrictions. This document is available to the 

Public through the National Technical Information Service; Springfield, VA 

22161 

19. Security Classif. (of this report) 20. Security Classif. (of this page) 21. No. of Pages 22. Price 

 Unclassified Unclassified 384 N/A 

Form DOT F 1700.7 (8-72) Reproduction of completed page authorized (art. 5/94)
ii 

SI* (MODERN METRIC) CONVERSION FACTORS 

APPROXIMATE CONVERSIONS TO SI UNITS

Symbol When You Know Multiply By To Find Symbol 

LENGTH 

in inches 25.4 millimeters mm 

ft feet 0.305 meters m 

yd yards 0.914 meters m 

mi miles 1.61 kilometers km 

AREA 

in2 square inches 645.2 square millimeters mm2

ft2 square feet 0.093 square meters m2

yd2 square yard 0.836 square meters m2

ac acres 0.405 hectares ha 

mi2 square miles 2.59 square kilometers km2

VOLUME 

fl oz fluid ounces 29.57 milliliters mL 

gal gallons 3.785 liters L 

ft3 cubic feet 0.028 cubic meters m3 

yd3 cubic yards 0.765 cubic meters m3 

NOTE: volumes greater than 1000 L shall be shown in m3

MASS 

oz ounces 28.35 grams g

lb pounds 0.454 kilograms kg

T short tons (2000 lb) 0.907 megagrams (or "metric ton") Mg (or "t") 

TEMPERATURE (exact degrees) o

F Fahrenheit 5 (F-32)/9 Celsius oC 

or (F-32)/1.8 

ILLUMINATION 

fc foot-candles 10.76 lux lx 

fl foot-Lamberts 3.426 candela/m2 cd/m2

FORCE and PRESSURE or STRESS 

lbf poundforce 4.45 newtons N 

lbf/in2 poundforce per square inch 6.89 kilopascals kPa 

APPROXIMATE CONVERSIONS FROM SI UNITS 

Symbol When You Know Multiply By To Find Symbol 

LENGTH

mm millimeters 0.039 inches in 

m meters 3.28 feet ft 

m meters 1.09 yards yd 

km kilometers 0.621 miles mi 

AREA mm2

 square millimeters 0.0016 square inches in2 

m2 square meters 10.764 square feet ft2 

m2 square meters 1.195 square yards yd2 

ha hectares 2.47 acres ac 

km2 square kilometers 0.386 square miles mi2 

VOLUME 

mL milliliters 0.034 fluid ounces fl oz 

L liters 0.264 gallons gal 

m3 cubic meters 35.314 cubic feet ft3 

m3 cubic meters 1.307 cubic yards yd3 

MASS 

g grams 0.035 ounces oz

kg kilograms 2.202 pounds lb

Mg (or "t") megagrams (or "metric ton") 1.103 short tons (2000 lb) T 

TEMPERATURE (exact degrees) o

C Celsius 1.8C+32 Fahrenheit oF 

ILLUMINATION 

lx lux 0.0929 foot-candles fc 

cd/m2 candela/m2 0.2919 foot-Lamberts fl

FORCE and PRESSURE or STRESS 

N newtons 0.225 poundforce lbf 

kPa kilopascals 0.145 poundforce per square inch lbf/in2

*SI is the symbol for th International System of Units. Appropriate rounding should be made to comply with Section 4 of ASTM E e 380. 

(Revised March 2003)
iii 

TABLE OF CONTENTS 

CHAPTER 1. INTRODUCTION...........................................................................1

1.1 GENERAL INTRODUCTION ................................................................................... 1

1.2 OBJECTIVES............................................................................................................... 2

1.3 ORGANIZATION OF REPORT................................................................................ 2

CHAPTER 2. BACKGROUND AND LITERATURE REVIEW.......................5

2.1 INTRODUCTION ........................................................................................................ 5

2.2 COMPOSITE PILE BACKGROUND....................................................................... 5

2.2.1 Types of Composite Piles...................................................................................... 5

2.3 LITERATURE REVIEW FOR THE COMPOSITE PILES SELECTED............. 9

2.3.1 Structural Behavior .............................................................................................. 9

2.3.2 Long-Term Durability Behavior........................................................................ 17

2.3.3 Geotechnical Behavior........................................................................................ 19

2.4 SUMMARY................................................................................................................. 23

CHAPTER 3. EXPERIMENTAL STUDY OF INTERFACE BEHAVIOR 

BETWEEN COMPOSITE PILES AND TWO SANDS ......................25

3.1 INTRODUCTION ...................................................................................................... 25

3.2 SOIL MATERIALS ................................................................................................... 28

3.2.1 Index Properties .................................................................................................. 28

3.2.2 Direct Shear Tests of Sands................................................................................ 30

3.3 PILE SURFACES....................................................................................................... 34

3.3.1 Introduction......................................................................................................... 34

3.3.2 Surface Topography Characterization ............................................................. 35

3.3.3 Interface Hardness.............................................................................................. 40

3.4 INTERFACE SHEAR TESTS .................................................................................. 41

3.5 DISCUSSION OF RESULTS.................................................................................... 51

3.5.1 Multiple Linear Regression for Density Sand tan δpeak Values....................... 52

3.5.2 Multiple Linear Regression for Density Sand tan δcv Values.......................... 53

3.5.3 Multiple Linear Regression for Model Sand tan δpeak Values......................... 54

3.5.4 Multiple Linear Regression for Model Sand tan δcv Values............................ 55

3.5.5 Observations from the Linear Regression Analyses Results .......................... 56

3.5.6 Influence of Angularity of Sand......................................................................... 56

3.6 SUMMARY................................................................................................................. 57

CHAPTER 4. EXPERIMENTAL DURABILITY STUDY OF FRP 

COMPOSITE PILES ..............................................................................59

4.1 INTRODUCTION ...................................................................................................... 59

4.2 BACKGROUND ON DEGRADATION OF GLASS FRP COMPOSITES ......... 59

4.3 LABORATORY STUDY OF DURABILITY.......................................................... 60

4.3.1 Description of Test Specimens ........................................................................... 61

4.3.2 Test Equipment and Procedures ....................................................................... 65
iv 

4.3.3 Baseline Mechanical Properties......................................................................... 70

4.3.4 Properties as a Function of Submergence Time and Moisture....................... 72

4.3.5 Freeze-Thaw Degradation of Saturated FRP Samples.................................... 93

4.3.6 SEM Imaging....................................................................................................... 95

4.4 LONG-TERM STRUCTURAL CAPACITY OF FRP COMPOSITE PILES..... 97

4.4.1 Long-Term Structural Capacity of Composite Pile......................................... 97

4.4.2 Comments Related to the Axial Strain Levels in Piles .................................... 99

4.5 SUMMARY............................................................................................................... 100

CHAPTER 5. FIELD LOAD TESTS AT THE ROUTE 40 BRIDGE .......... 103

5.1 INTRODUCTION .................................................................................................... 103

5.2 DESCRIPTION OF THE BRIDGE ....................................................................... 103

5.2.1 The Former Bridge ........................................................................................... 103

5.2.2 The New Bridge................................................................................................. 105

5.3 DESCRIPTION OF TEST PILES.......................................................................... 107

5.3.1 Composite Test Pile........................................................................................... 107

5.3.2 Prestressed Concrete Test Pile......................................................................... 109

5.4 SOIL CONDITIONS AT THE TEST SITE .......................................................... 110

5.5 PILE FABRICATION AND INSTALLATION.................................................... 111

5.5.1 Fabrication of the Piles..................................................................................... 111

5.5.2 Pile Installation.................................................................................................. 114

5.5.2 Pile Installation.................................................................................................. 114

5.6 STATNAMIC FIELD TESTING OF TEST PILES ............................................. 117

5.6.1 Instrumentation................................................................................................. 118

5.6.2 Axial Load Tests................................................................................................ 119

5.6.3 Lateral Load Tests ............................................................................................ 122

5.7 COMPOSITE PILES IN THE NEW ROUTE 40 BRIDGE................................. 130

5.7.1 Connection of Piles with Pile Cap Beam......................................................... 130

5.8 SUMMARY AND CONCLUSIONS....................................................................... 133

CHAPTER 6. FIELD LOAD TESTS AT THE ROUTE 351 BRIDGE ........ 135

6.1 INTRODUCTION .................................................................................................... 135

6.2 DESCRIPTION OF THE BRIDGE ....................................................................... 135

6.2.1 The Original Bridge.......................................................................................... 135

6.2.2 The New Bridge................................................................................................. 138

6.3 DESCRIPTION OF TEST PILES.......................................................................... 140

6.3.1 Prestressed Concrete Test Pile......................................................................... 140

6.3.2 FRP Composite Test Pile.................................................................................. 140

6.3.3 Polyethylene Composite Test Pile.................................................................... 143

6.4 SOIL CONDITIONS AT THE TEST SITE .......................................................... 146

6.4.1 Geology............................................................................................................... 146

6.4.2 Subsurface Conditions...................................................................................... 147

6.4.3 Hampton River.................................................................................................. 148

6.5 TEST PILE INSTRUMENTATION AND FABRICATION ............................... 150

6.5.1 Test Pile Instrumentation................................................................................. 150

6.5.2 Fabrication of Prestressed Concrete Test Pile ............................................... 155
v 

6.5.3 Fabrication of FRP Composite Test Pile......................................................... 156

6.5.3 Fabrication of FRP Composite Test Pile......................................................... 156

6.5.4 Fabrication of Plastic Composite Test Pile..................................................... 160

6.6 PILE INSTALLATION AND DYNAMIC TESTING.......................................... 162

6.6.1 Pile Driving ........................................................................................................ 162

6.6.2 Dynamic Testing................................................................................................ 166

6.7 PILE INTEGRITY TESTING OF TEST PILES.................................................. 169

6.8 FIELD TESTING OF TEST PILES....................................................................... 174

6.8.1 Axial Load Tests................................................................................................ 174

6.8.2 Lateral Load Tests ............................................................................................ 182

6.9 SUMMARY AND CONCLUSIONS....................................................................... 183

CHAPTER 7. ANALYSES OF THE AXIAL LOAD TESTS AT THE 

ROUTE 351 BRIDGE .......................................................................... 185

7.1 INTRODUCTION .................................................................................................... 185

7.2 AXIAL PILE CAPACITY....................................................................................... 185

7.2.1 Methods to Estimate Axial Load Capacity of Driven Piles in Sand............. 185

7.2.2 Predicted Axial Capacities ............................................................................... 195

7.2.3 Summary of Axial Pile Capacity Predictions ................................................. 202

7.3 LOAD-SETTLEMENT BEHAVIOR, AXIALLY LOADED SINGLE PILES.. 204

7.3.1 Introduction....................................................................................................... 204

7.3.2 Background, Load-Transfer Method for Pile Settlement Predictions......... 205

7.3.3 Predictions Using Empirical Load-Transfer Curves..................................... 206

7.3.4 Predictions Using Theoretical Load-Transfer Curves .................................. 217

7.3.5 Residual Stresses ............................................................................................... 235

7.3.5 Residual Stresses ............................................................................................... 235

7.4 CONCLUSIONS....................................................................................................... 235

CHAPTER 8. ANALYSES OF THE LATERAL LOAD TESTS AT THE 

ROUTE 351 BRIDGE .......................................................................... 237

8.1 INTRODUCTION .................................................................................................... 237

8.2 GOVERNING DIFFERENTIAL EQUATION FOR THE LATERALLY 

LOADED PILE PROBLEM ................................................................................... 237

8.2.1 Assumptions and Limitations of the Governing Differential Equation....... 241

8.3 METHODOLOGY USED TO ANALYZE THE LATERALLY LOADED 

TEST PILES ............................................................................................................. 242

8.3.1 P-y Curves.......................................................................................................... 242

8.3.2 P-y Method of Analysis..................................................................................... 247

8.4 NUMERICAL ANALYSES RESULTS ................................................................. 248

8.4.1 General Input Information .............................................................................. 249

8.4.2 P-y Analyses Results ......................................................................................... 252

8.4.2 P-y Analyses Results ......................................................................................... 252

8.4.3 Comparison of the Initial p-y Modulus Curves for the Three Test Piles..... 266

8.5 LIMITATIONS OF P-Y ANALYSES.................................................................... 266

8.6 SUMMARY............................................................................................................... 267
vi 

CHAPTER 9. LONG-TERM MONITORING AT THE ROUTE 351 

BRIDGE................................................................................................. 269

9.1 INTRODUCTION .................................................................................................... 269

9.2 INSTRUMENTED PRODUCTION PILES .......................................................... 269

9.2.1 Prestressed Concrete Production Pile ............................................................. 270

9.2.2 FRP Composite Production Pile ...................................................................... 272

9.3 SOIL CONDITIONS AT THE INSTRUMENTED PRODUCTION PILES ..... 275

9.4 PILE INSTALLATION AND DYNAMIC TESTING.......................................... 278

9.4 PILE INSTALLATION AND DYNAMIC TESTING.......................................... 278

9.4.1 Pile Driving ........................................................................................................ 278

9.4.2 Dynamic Testing................................................................................................ 279

9.5 PILE INTEGRITY TESTING OF TEST PILES.................................................. 281

9.6 MONITORING DATA GATHERED TO DATE ................................................. 284

9.7 SUMMARY............................................................................................................... 287

CHAPTER 10. COST INFORMATION FOR COMPOSITE PILES .......... 289

10.1 INTRODUCTION .................................................................................................... 289

10.2 COST INFORMATION FOR THE ROUTE 40 BRIDGE PROJECT............... 289

10.3 COST INFORMATION FOR THE ROUTE 351 BRIDGE PROJECT............. 289

10.3.1 Hardcore FRP Composite Pile......................................................................... 290

10.3.2 Plastic Pile.......................................................................................................... 290

10.3.3 Prestressed Concrete Pile ................................................................................. 290

10.3.4 Summary of the Route 351 Bridge Cost Information.................................... 290

10.4 SUMMARY............................................................................................................... 290

CHAPTER 11. SUMMARY AND CONCLUSIONS ...................................... 293

11.1 INTRODUCTION ................................................................................................... 293

11.2 SUMMARY OF ACTIVITIES AND CONCLUSIONS........................................ 294

11.2.1 Literature Review ............................................................................................. 295

11.2.2 Interface Study .................................................................................................. 295

11.2.3 Durability Study................................................................................................ 296

11.2.4 Field Tests at the Route 40 Bridge................................................................... 297

11.2.5 Field Tests at the Route 351 Bridge................................................................. 298

11.2.6 Axial Analyses ................................................................................................... 299

11.2.7 Lateral Analyses................................................................................................ 300

11.2.8 Long-Term Monitoring .................................................................................... 301

11.2.9 Cost Information for Composite Piles............................................................. 302

11.3 RECOMMENDATIONS FOR FUTURE WORK ................................................ 302

11.3.1 Geotechnical Studies......................................................................................... 302

11.3.2 Durability Studies ............................................................................................. 303

11.3.3 Structural Tests................................................................................................. 303

11.3.4 Cost Analyses..................................................................................................... 304

APPENDIX A. INTERFACE TEST RESULTS.............................................. 305
vii 

APPENDIX B. MOISTURE DIFFUSION INTO A CYLINDRICAL FRP 

COMPOSITE........................................................................................ 321

APPENDIX C. STRUCTURAL TESTS RESULTS FROM COMPOSITE 

PILE CUTOFF SECTIONS FROM THE ROUTE 40 BRIDGE 

PROJECT.............................................................................................. 325

C.1 PUSHOUT TEST RESULTS .................................................................................. 326

C.2 CREEP BENDING TEST........................................................................................ 328

APPENDIX D. GEOTECHNICAL FIELD INVESTIGATIONS AT THE 

ROUTE 351 BRIDGE TEST SITE..................................................... 331

REFERENCES.................................................................................................... 347
viii 

LIST OF FIGURES 

Figure 1. Photos. Degradation of conventional piles (Iskander and Hassan 1998). ....................... 5 

Figure 2. Illustration. Common types of composite piles............................................................... 6 

Figure 3. Graph and Photos. Confinement effect of FRB tube on concrete (Fam and Rizkalla 

2001a, b). .............................................................................................................................. 11 

Figure 4. Graph. Experimental versus predicted load-strain behavior using Fam and 

Rizkalla’s model. .................................................................................................................. 13 

Figure 5. Illustration. Strip elements for sectional analysis (Mirmiran and Shahawy 1996). ..... 14 

Figure 6. Graphs. Experimental versus analytical moment-curvature response (adapted from 

Fam and Rizkalla 2002)........................................................................................................ 15 

Figure 7. Graph. Interaction diagrams, concrete-filled FRP tubes (Mirmiran 1999). .................. 16 

Figure 8. Graphs. Moisture absorption-related durability model. ................................................ 18 

Figure 9. Images. SEM images showing FRP damage (McBagonluri, et al., 2000). ................... 19 

Figure 10. Illustration. Influence of soil-pile friction on pile capacity......................................... 26 

Figure 11. Graph. Grain size curves to test sands......................................................................... 29 

Figure 12. Photos. Microscopic views of the test sands. .............................................................. 29 

Figure 13. Graphs. Direct shear test results for Density sand (average Dr = 70 %). .................... 31 

Figure 14. Graphs. Direct shear test results for Density sand (average Dr = 100 %). .................. 32 

Figure 15. Graphs. Direct shear test results for Model sand (average Dr = 75%). ....................... 33 

Figure 16. Illustration. Stylus profilometer sketch (Johnson 2000).............................................. 35 

Figure 17. Chart. Graphical representation of roughness parameters Rt, Sm, and Ra. .................. 36 

Figure 18. Photo and Graph. Surface characteristics of Lancaster FRP composite pile. ............ 37 

Figure 19. Photo and Graph. Surface characteristics of Hardcore FRP composite pile. ............. 37 

Figure 20. Photo and Graph. Surface characteristics of Hardcore FRP plate............................... 38 

Figure 21. Photo and Graph. Surface characteristics of Hardcore surface-treated FRP plate. .... 38 

Figure 22. Photo and Graph. Surface characteristics of Plastic Piling plastic composite pile. ... 39 

Figure 23. Photo and Graph. Surface characteristics of prestressed concrete pile. ...................... 39 

Figure 24. Photo and Graph. Surface characteristics of steel sheet pile....................................... 40 

Figure 25. Illustration. Sketch of modified interface shear test setup. ......................................... 42 

Figure 26. Graphs. Typical interface shear test results for Density sand (σ'n ≈ 100 kPa). .......... 45 

Figure 27. Graphs. Typical interface shear test results for Model sand (σ'n ≈ 100 kPa). ............ 46 

Figure 28. Graph. Interface shear strength envelopes for Lancaster Composite FRP shell. ....... 47 

Figure 29. Graph. Interface shear strength envelopes for Hardcore Composites FRP shell. ...... 47 

Figure 30. Graph. Interface shear strength envelopes for untreated Hardcore FRP plate. .......... 48 

Figure 31. Graph. Interface shear strength envelopes for treated Hardcore FRP plate. .............. 48 

Figure 32. Graph. Interface shear strength envelopes for PPI plastic........................................... 49 

Figure 33. Graph. Interface shear strength envelopes for concrete. ............................................. 49 

Figure 34. Graph. Interface shear strength envelopes for steel..................................................... 50 

Figure 35. Graph. Multiple linear regression on Density sand tan δpeak values............................ 52 

Figure 36. Graph. Multiple linear regression on Density sand tan δcv values. ............................. 53 

Figure 37. Graph. Multiple linear regression on Model sand tan δpeak values.............................. 54 

Figure 38. Graph. Multiple linear regression on Model sand tan δcv values. ............................... 55 

Figure 39. Photos. Burnoff testing................................................................................................ 66 

Figure 40. Photo. Typical tension test setup................................................................................. 67 
ix 

Figure 41. Photos. Typical split disk test setup. ........................................................................... 68 

Figure 42. Photo. Freeze-thaw chamber. ...................................................................................... 68 

Figure 43. Photo and Illustration. Freeze-thaw fixture................................................................. 69 

Figure 44. Graph. Average freeze-thaw cycle undergone by FRP samples. ................................ 69 

Figure 45. Graph. Representative baseline longitudinal tension stress-strain curves................... 70 

Figure 46. Graph. Representative baseline hoop tension stress-strain curves. ............................. 71 

Figure 47. Graph. Absorption curves for Lancaster 12-inch FRP tube. ....................................... 72 

Figure 48. Graph. Absorption curves for Lancaster 24-inch FRP tube. ....................................... 73 

Figure 49. Graph. Absorption curves for Hardcore 12-inch FRP tube......................................... 73 

Figure 50. Graph. Absorption curves for Hardcore 24-inch FRP tube......................................... 74 

Figure 51. Graphs. Selected diffusion analyses for Lancaster 12-inch FRP samples................... 76 

Figure 52. Graphs. Selected diffusion analyses for Lancaster 24-inch FRP samples................... 77 

Figure 53. Graphs. Selected diffusion analyses for Hardcore 12-inch FRP samples. .................. 78 

Figure 54. Graphs. Selected diffusion analyses for Hardcore 24-inch FRP samples. .................. 79 

Figure 55. Graph. Longitudinal tensile properties versus submergence time for Lancaster 

24-inch FRP tube. ................................................................................................................. 82 

Figure 56. Graph. Hoop tensile properties versus submergence time for Lancaster 24-inch 

 FRP tube............................................................................................................................... 82 

Figure 57. Graph. Longitudinal tensile properties versus submergence time for Lancaster 

12-inch FRP tube. ................................................................................................................. 83 

Figure 58. Graph. Hoop tensile properties versus submergence time for Lancaster 12-inch 

FRP tube................................................................................................................................ 84 

Figure 59. Charts. Longitudinal tensile properties versus moisture content for Lancaster 

12-inch tube. ......................................................................................................................... 85 

Figure 60. Charts. Hoop tensile properties versus moisture content for Lancaster 12-inch 

FRP tube................................................................................................................................ 86 

Figure 61. Graph. Longitudinal tensile properties versus submergence time for Hardcore 

24-inch FRP tube. ................................................................................................................. 87 

Figure 62. Charts. Longitudinal tensile properties versus moisture content for Hardcore 

24-inch FRP tube. ................................................................................................................. 88 

Figure 63. Graph. Longitudinal tensile properties versus submergence time for Hardcore 

12-inch FRP tube. ................................................................................................................. 89 

Figure 64. Graph. Hoop tensile properties versus submergence time for Hardcore 12-inch 

FRP tube................................................................................................................................ 90 

Figure 65. Charts. Longitudinal tensile properties versus moisture content for Hardcore 

12-inch FRP tube. ................................................................................................................. 91 

Figure 66. Charts. Hoop tensile properties versus moisture content for Hardcore 12-inch 

FRP tube................................................................................................................................ 92 

Figure 67. Graph. Influence of freeze-thaw cycling on the longitudinal tensile properties 

for the Lancaster 24-inch FRP tube. ..................................................................................... 93 

Figure 68. Graph. Influence of freeze-thaw cycling on the longitudinal tensile properties 

for the Lancaster 12-inch FRP tube. ..................................................................................... 94 

Figure 69. Graph. Influence of freeze-thaw cycling on the longitudinal tensile properties 

for the Hardcore 24-inch FRP tube....................................................................................... 94 

Figure 70. Graph. Influence of freeze-thaw cycling on the longitudinal tensile properties 

for the Hardcore 12-inch FRP tube....................................................................................... 95 
x 

Figure 71. Photo. SEM images of Lancaster 24-inch FRP tube. .................................................. 96 

Figure 72. Photo. SEM images of Lancaster 12-inch FRP tube. .................................................. 96 

Figure 73. Photo. SEM images of Hardcore 24-inch FRP tube.................................................... 96 

Figure 74. Photo. SEM images of Hardcore 12-inch FRP tube.................................................... 97 

Figure 75. Graph. Estimated long-term axial capacity of the 12-inch Lancaster pile. ................. 98 

Figure 76. Graph. Estimated long-term flexural capacity of the 12-inch Lancaster pile............. 99 

Figure 77. Map. Location map of the Route 40 Bridge project in Sussex County, VA 

(Fam, et al., 2003)............................................................................................................... 104 

Figure 78. Photo. Former Route 40 Bridge................................................................................. 104 

Figure 79. Photos. Signs of deterioration of the former Route 40 Bridge (Fam, et al., 2003). .. 105 

Figure 80. Illustration. Schematic of the new Route 40 Bridge.................................................. 106 

Figure 81. Photo and Illustration. Concrete-filled tubular piles. ................................................ 108 

Figure 82. Graph. Stress-strain response of concrete used in the composite pile....................... 109 

Figure 83. Illustration. Reinforcement details of prestressed concrete pile................................ 110 

Figure 84. Illustration. Simplified soil stratigraphy near test pile area....................................... 111 

Figure 85. Photo. Fabrication of prestressed concrete pile......................................................... 112 

Figure 86. Photos. Fabrication of concrete-filled FRP Piles. ..................................................... 113 

Figure 87. Photos. Driving of test piles. ..................................................................................... 114 

Figure 88. Graph. Driving records for test piles. ........................................................................ 115 

Figure 89. Graphs. End-of-driving PDA recordings................................................................... 116 

Figure 90. Illustration. Test Pile Instrumentation. ...................................................................... 119 

Figure 91. Photo. Axial load test using the Statnamic device. ................................................... 120 

Figure 92. Graph. Pile head displacement versus equivalent axial static load. .......................... 121 

Figure 93. Graph. Axial load—axial strain behavior of test piles. ............................................. 121 

Figure 94. Illustration and Graph. Variation of axial strain along pile length for three 

Statnamic loads. .................................................................................................................. 122 

Figure 95. Photos. Lateral Statnamic setup at the Route 40 project........................................... 123 

Figure 96. Graphs. Displacement time histories at the loading point for both piles 

(Pando, et al., 2004). ........................................................................................................... 124 

Figure 97. Graphs. Peak lateral displacement profiles for both test piles at different cycles of 

Statnamic load..................................................................................................................... 124 

Figure 98. Graph. Calculated static and dynamic (static + damping) resistances for both test 

piles..................................................................................................................................... 127 

Figure 99. Graph. Moment-curvature responses for composite and pretressed 

concrete piles. ..................................................................................................................... 129 

Figure 100. Graph. Computed and measured lateral loan-displacement response for both test 

piles (Pando, et al., 2004).................................................................................................... 130 

Figure 101. Illustration. Details of pile head showing the bars used to connect the pile to cap 

beam.................................................................................................................................... 131 

Figure 102. Ilustration. Connection of composite piles to cap beam at Pier No. 2. ................... 132 

Figure 103. Photos. Pier No. 2 including the composite piles and reinforced concrete 

cap beam. ............................................................................................................................ 132 

Figure 104. Photo. The new Route 40 Bridge over the Nottoway River in Virginia. ................ 133 

Figure 105. Map. Location map of the Route 351 Bridge in Hampton, VA. ............................. 136 

Figure 106. Photo. Aerial view of the Route 351 Bridge in Hampton, VA................................ 136 

Figure 107. Photos. Wide-angle views of the original Route 351 Bridge. ................................. 137 
xi 

Figure 108. Photo. Signs of deterioration of the original Route 351 Bridge.............................. 138 

Figure 109. Illustration. Schematic of the new Route 351 Bridge.............................................. 139 

Figure 110. Illustration. Test pile cross section details............................................................... 140 

Figure 111. Graphs. Test pile material properties....................................................................... 142 

Figure 112. Graph. Axial load-axial strain behavior of test piles............................................... 143 

Figure 113. Graphs. Flexural characteristics for the three test piles........................................... 145 

Figure 114. Map. Location of test pile site at the Route 351 Bridge.......................................... 146 

Figure 115. Charts. Simplified soil stratigraphy near test pile area............................................ 148 

Figure 116. Illustration. Pile load test layout.............................................................................. 149 

Figure 117. Illustration. Instrumentation layout for prestressed concrete test pile..................... 151 

Figure 118. Illustration. Instrumentation layout for FRP composite test pile. ........................... 153 

Figure 119. Illustration. Instrumentation layout for plastic composite test pile. ........................ 154 

Figure 120. Photos. Fabrication of prestressed concrete test pile............................................... 155 

Figure 121. Photos. Fabrication of concrete-filled FRP piles..................................................... 157 

Figure 122. Photos. Setup used for concrete filling of FRP composite piles. ............................ 158 

Figure 123. Photos. Concrete filling of FRP composite piles..................................................... 159 

Figure 124. Photos. Rebar cage of the plastic composite test pile.............................................. 160 

Figure 125. Photos. Manufacturing process for the plastic composite test pile. ........................ 161 

Figure 126. Graph. Driving records for test piles. ...................................................................... 162 

Figure 127. Photos. Installation of prestressed concrete test pile............................................... 163 

Figure 128. Photos. Installation of FRP composite test pile....................................................... 164 

Figure 129. Photos. Installation of plastic composite test pile. .................................................. 165 

Figure 130. Graphs. PDA recordings during restrike. ................................................................ 167 

Figure 131. Photos. PIT tests on test piles.................................................................................. 170 

Figure 132. Graph. PIT sounding on the prestressed concrete test pile before installation....... 171 

Figure 133. Graph. PIT sounding on the prestressed concrete test pile after installation.......... 171 

Figure 134. Graph. PIT sounding on the FRP composite test pile before installation. .............. 172 

Figure 135. Graph. PIT sounding on the FRP composite test pile after installation. ................. 172 

Figure 136. Graph. PIT sounding on the plastic composite test pile before installation. .......... 173 

Figure 137. Graph. PIT sounding on the plastic composite test pile after installation............... 173 

Figure 138. Photos. Axial load test of prestressed concrete pile. ............................................... 175 

Figure 139. Graph. Axial load test results. ................................................................................. 176 

Figure 140. Graph. Distribution of residual loads. ..................................................................... 178 

Figure 141. Graph. Distribution of residual stresses................................................................... 178 

Figure 142. Graph. Load distribution for the three test piles at the Davisson failure loads. ..... 179 

Figure 143. Graph. Mobilized average unit shaft resistance—toe resistance relationships 

for the three test piles.......................................................................................................... 181 

Figure 144. Graph. Apparent strength gain with time in the three test piles. ............................. 181 

Figure 145. Graphs. Deformed shapes of piles at different lateral loads.................................... 182 

Figure 146. Graph. Measured lateral deflections at ground surface for the three test piles. ..... 183 

Figure 147. Illustration. Load transfer in an axially loaded pile................................................. 186 

Figure 148. Graphs. Interpreted average CPT and SPT design profiles for Route 351 

test site. ............................................................................................................................... 195 

Figure 149. Graph. Accuracy of Nordlund’s method predictions using δ values from 

Nordlund’s charts................................................................................................................ 196 
xii 

Figure 150. Graph. Accuracy of Nordlund’s method predictions using δ values from 

interface shear tests............................................................................................................. 197 

Figure 151. Graph. Accuracy of API method predictions using δ values from table 36........... 198 

Figure 152. Graph. Accuracy of API method predictions using δ values from interface 

shear tests............................................................................................................................ 199 

Figure 153. Graph. Accuracy of LCPC method predictions using steel pile assumption. ........ 200 

Figure 154. Graph. Accuracy of LCPC method predictions using concrete pile assumption. .. 201 

Figure 155. Graph. Accuracy of IC method predictions using δ values from interface 

shear tests............................................................................................................................ 202 

Figure 156. Illustration. Idealized model used in T-Z load-transfer analyses. ........................... 205 

Figure 157. Graph. Pile tip load—pile tip displacement curve (Q-Z) (API 1993). .................... 206 

Figure 158. Graph. Maximum shear stress distribution along pile shaft, according to 

API (1993). ......................................................................................................................... 207 

Figure 159. Graph. Settlement predictions for the prestressed concrete pile using 

API (1993). ......................................................................................................................... 208 

Figure 160. Graph. Settlement predictions for the FRP pile using API (1993).......................... 209 

Figure 161. Graph. Settlement predictions for the plastic pile using API (1993). ..................... 210 

Figure 162. Graph. Normalized T-Z curves according to API (1993) and 

Vijayvergiya (1977)............................................................................................................ 211 

Figure 163. Graph. Normalized Q-Z curves according to Vijayvergiya (1977) and 

API (1993). ......................................................................................................................... 212 

Figure 164. Graph. Maximum shear stress distributions used in predictions using 

Vijayvergiya (1977)............................................................................................................ 213 

Figure 165. Graph. Settlement predictions for the concrete pile using Vijayvergiya (1977).... 214 

Figure 166. Graph. Settlement predictions for the FRP pile using Vijayvergiya (1977). ......... 215 

Figure 167. Graph. Settlement predictions for the plastic pile using Vijayvergiya (1977)....... 216 

Figure 168. Illustrations. Concentric cylinder model for settlement analysis of axially loaded 

piles (adapted from Randolph and Wroth 1978)................................................................. 218 

Figure 169. Graph. Linear T-Z curve obtained using Randolph and Wroth (1978)................... 221 

Figure 170. Graph. Linear Qb-Z curve obtained using Boussinesq’s theory.............................. 222 

Figure 171. Graph. Hyperbolic T-Z curve.................................................................................. 226 

Figure 172. Graph. Variation of secant shear modulus for different hyperbolic-type 

models................................................................................................................................. 227 

Figure 173. Graph. Theoretically derived T-Z curve using concentric cylinders and the 

modified hyperbola from Fahey and Carter (1993). ........................................................... 228 

Figure 174. Graph. Theoretically derived Q-Z curve using Boussinesq’s theory and the 

modified hyperbola from Fahey and Carter (1993). ........................................................... 229 

Figure 175. Graph. Route 351 initial shear modulus profile from CPT correlations. ................ 230 

Figure 176. Graph. Settlement predictions for the concrete pile using theoretically derived 

transfer curves..................................................................................................................... 232 

Figure 177. Graph. Settlement prediction for the FRP pile using theoretically derived transfer 

curves. ................................................................................................................................. 233 

Figure 178. Graph. Settlement predictions for the plastic pile using theoretically derived 

transfer curves..................................................................................................................... 234 

Figure 179. Illustration. Laterally loaded pile problem. ............................................................. 239 
xiii 

Figure 180. Illustration. Distribution of stresses against a pile before and after lateral loading 

(adapted from Reese and Van Impe 2001). ........................................................................ 243 

Figure 181. Graphs. Typical p-y curve and resulting p-y modulus (Reese and 

Van Impe 2001). ................................................................................................................. 244 

Figure 182. Illustration. Schematic showing the influence of shape of cross section of pile 

on the soil reaction p (adapted from Reese and Van Impe 2001)....................................... 245 

Figure 183. Graph. Elements of a characteristic p-y curve for sand based on recommendations 

by Reese, et al. (1974)......................................................................................................... 246 

Figure 184. Illustration. Schematic showing p-y model used for analysis of laterally 

loaded piles. ........................................................................................................................ 248 

Figure 185. Graphs. In situ test data for the upper soils at the northern end of the test 

pile site................................................................................................................................ 250 

Figure 186. Graphs. In situ test data for the upper soils at the southern end of the test 

pile site................................................................................................................................ 251 

Figure 187. Graph. Initial p-y modulus profile used to define p-y default curves for LPILE 

analyses on the prestressed concrete pile............................................................................ 253 

Figure 188. Graphs. Predicted versus measured lateral displacement profile for prestressed 

concrete pile (low lateral loads).......................................................................................... 254 

Figure 189. Graphs. Predicted versus measured lateral displacement profile for prestressed 

concrete pile (medium lateral loads)................................................................................... 254 

Figure 190. Graphs. Predicted versus measured lateral displacement profile for prestressed 

concrete pile (high lateral loads)......................................................................................... 255 

Figure 191. Graph. Calculated load-deflection curve for the prestressed concrete pile............ 256 

Figure 192. Graph. Calculated load-slope curve for the prestressed concrete pile..................... 256 

Figure 193. Graph. Initial p-y modulus profile used to define default p-y curves for LPILE 

analyses on the FRP pile..................................................................................................... 258 

Figure 194. Graphs. Predicted versus measured lateral displacement profile for FRP pile 

(low lateral loads). .............................................................................................................. 258 

Figure 195. Graphs. Predicted versus measured lateral displacement profile for FRP pile 

(medium lateral loads). ....................................................................................................... 259 

Figure 196. Graphs. Predicted versus measured lateral displacement profile for FRP pile 

(high lateral loads). ............................................................................................................. 259 

Figure 197. Graph. Calculated load-deflection curve for the FRP pile. ..................................... 260 

Figure 198. Graph. Calculated load-slope curve for the FRP pile.............................................. 261 

Figure 199. Graph. Initial p-y modulus profile used to define default p-y curves for LPILE 

analyses on the plastic pile.................................................................................................. 262 

Figure 200. Graphs. Predicted versus measured lateral displacement profile for plastic pile 

(low lateral loads). .............................................................................................................. 263 

Figure 201. Graphs. Predicted versus measured lateral displacement profile for plastic pile 

(medium lateral loads). ....................................................................................................... 263 

Figure 202. Graphs. Predicted versus measured lateral displacement profile for plastic pile 

(high lateral loads). ............................................................................................................. 264 

Figure 203. Graph. Calculated load-deflection curve for the plastic pile................................... 265 

Figure 204. Graph. Calculated load-slope curve for the plastic pile. ......................................... 265 

Figure 205. Illustration. Location of instrumented production piles at the 

Route 351 Bridge. ............................................................................................................... 270 
xiv 

Figure 206. Illustration. Load transfer instrumentation layout for prestressed concrete 

production pile. ................................................................................................................... 271 

Figure 207. Illustration. Instrumentation layout for FRP composite production pile................. 273 

Figure 208. Illustration. Layout of external durability instrumentation used in the FRP 

composite production pile................................................................................................... 274 

Figure 209. Graph. Simplified stratigraphy along the Route 351 Bridge alignment.................. 275 

Figure 210. Graph. Simplified stratigraphy in the vicinity of the instrumented prestressed 

concrete pile installed at Pier 10. ........................................................................................ 276 

Figure 211. Graph. Simplified stratigraphy in the vicinity of the instrumented FRP composite 

pile installed at Pier 11........................................................................................................ 277 

Figure 212. Graph. Driving records for instrumented production piles. .................................... 278 

Figure 213. Graph. PDA recordings during restrike (Spiro and Pais 2002b)............................. 280 

Figure 214. Graph. PIT sounding on the prestressed concrete production pile before 

installation........................................................................................................................... 282 

Figure 215. Graph. PIT sounding on the prestressed concrete production pile after 

installation and restrike....................................................................................................... 282 

Figure 216. Graph. PIT sounding on the FRP composite production pile before 

installation........................................................................................................................... 283 

Figure 217. Graph. PIT sounding on the FRP composite production pile before 

installation........................................................................................................................... 283 

Figure 218. Photo. Route 351 Bridge under construction on December 30, 2002..................... 285 

Figure 219. Graph. Load distributions on November 7, 2002, and December 30, 2002, 

for the prestressed concrete production pile at Pier 10....................................................... 286 

Figure 220. Graph. Load distributions on November 7, 2002, and December 30, 2002, for the 

FRP composite production pile at Pier 11. ......................................................................... 287 

Figure 221. Graphs. Interface shear test results, Density sand-to-Lancaster FRP shell 

interface............................................................................................................................... 306 

Figure 222. Graphs. Interface shear test results, Density sand-to-Hardcore FRP shell 

interface............................................................................................................................... 307 

Figure 223. Graphs. Interface shear test results, Density sand-to-Hardcore FRP plate 

(untreated) interface............................................................................................................ 308 

Figure 224. Graphs. Interface shear test results, Density sand-to-treated Hardcore FRP plate 

interface............................................................................................................................... 309 

Figure 225. Graphs. Interface shear test results, Density sand-to-plastic interface.................... 310 

Figure 226. Graphs. Interface shear test results, Density sand-to-concrete interface................. 311 

Figure 227. Graphs. Interface shear test results, Density sand-to-steel interface....................... 312 

Figure 228. Graphs. Interface shear test results, Model sand-to-Lancaster FRP shell 

interface............................................................................................................................... 313 

Figure 229. Graphs. Interface shear test results, Model sand-to-Hardcore FRP shell 

interface............................................................................................................................... 314 

Figure 230. Graphs. Interface shear test results, Model sand-to-Hardcore FRP plate 

(untreated) interface............................................................................................................ 315 

Figure 231. Graphs. Interface shear test results, Model sand-to-treated Hardcore FRP plate 

interface............................................................................................................................... 316 

Figure 232. Graphs. Interface shear test results, Model sand-to-plastic interface...................... 317 

Figure 233. Graphs. Interface shear test results, Model sand-to-concrete interface................... 318 
xv 

Figure 234. Graphs. Interface shear test results, Model sand-to-steel interface. ........................ 319 

Figure 235. Graph. FRP moisture concentration profile, inner radius dry, outer 

radius saturated. .................................................................................................................. 324 

Figure 236. Graph. FRP moisture concentration profile, inner and outer radii saturated.......... 324 

Figure 237. Photo. VTRC pushout test setup. ............................................................................ 326 

Figure 238. Photo. VT pushout test setup................................................................................... 327 

Figure 239. Photo. Creep bending test setup. ............................................................................. 329 

Figure 240. Graph. Creep deflection test results......................................................................... 329 

Figure 241. Map. Location of field tests..................................................................................... 332 

Figure 242. Chart. Boring B-1 (SPT-1). ..................................................................................... 333 

Figure 243. Chart. Boring B-2 (SPT-2). ..................................................................................... 335 

Figure 244. Chart. CPT-1............................................................................................................ 337 

Figure 245. Chart. CPT-2............................................................................................................ 338 

Figure 246. Chart. CPT033......................................................................................................... 339 

Figure 247. Chart. CPT034......................................................................................................... 343 

Figure 248. Chart. DMT. ............................................................................................................ 346 
xvi 

LIST OF TABLES 

Table 1. Selected projects involving installation of composite piles.............................................. 7 

Table 2. Expressions for the strength of confined concrete and maximum axial strain 

(adapted from Shehata, et al., 2002). .................................................................................... 10 

Table 3. Available structural information for steel-reinforced plastic composite piles............... 17 

Table 4. Comparison of pile impedance. ...................................................................................... 21 

Table 5. Interface behavior test matrix. ........................................................................................ 27 

Table 6. Index parameter values of the sands used in this study. ................................................. 28 

Table 7. Internal friction angles obtained from direct shear tests. (1) ............................................ 30 

Table 8. Summary of surface roughness measurements............................................................... 40 

Table 9. Surface hardness. ............................................................................................................ 41 

Table 10. Summary of interface shear test results on Density sand. ............................................ 43 

Table 11. Summary of interface shear test results on Model sand. .............................................. 44 

Table 12. Summary of interface friction angles............................................................................ 50 

Table 13. Multiple linear regression results for tan δpeak of Density sand.................................... 52 

Table 14. Multiple linear regression results for tan δcv of Density sand. ..................................... 53 

Table 15. Multiple linear regression results for tan δpeak of Model sand...................................... 54 

Table 16. Multiple linear regression results for tan δcv of Model sand. ....................................... 55 

Table 17. Characterization data for Lancaster Composite 12-inch FRP tube............................... 62 

Table 18. Characterization data for Lancaster Composite 24-inch FRP tube............................... 62 

Table 19. Characterization data for Hardcore Composites 12-inch FRP tube.............................. 64 

Table 20. Characterization data for Hardcore Composites 24-inch FRP tube.............................. 64 

Table 21. As-received longitudinal tensile properties. ................................................................. 70 

Table 22. As-received hoop tensile properties.............................................................................. 71 

Table 23. Fickian diffusion parameters for the Lancaster 12-inch FRP. ...................................... 80 

Table 24. Fickian diffusion parameters for the Lancaster 24-inch FRP. ...................................... 80 

Table 25. Fickian diffusion parameters for the Hardcore 12-inch FRP........................................ 81 

Table 26. Fickian diffusion parameters for the Hardcore 24-inch FRP........................................ 81 

Table 27. Mechanical properties of FRP shell of composite pile............................................... 108 

Table 28. Pile-driving measurements for the prestressed and composite piles. ......................... 117 

Table 29. Summary of signal-matching analyses at end of driving (from Muchard, 

et al., 1999). ........................................................................................................................ 117 

Table 30. Summary of pile-driving measurements for the prestressed and composite piles 

(after Spiro and Pais 2002a and b)...................................................................................... 168 

Table 31. Summary of CASE and CAPWAP analyses results (Spiro and Pais 2002a and b).... 169 

Table 32. Test dates for the test pile program at the Route 351 Bridge project. ........................ 174 

Table 33. Comparison of failure loads for the three test piles.................................................... 177 

Table 34. Load distributions from static load tests at the Davisson failure loads. ..................... 180 

Table 35. Unit shaft and toe resistances from static axial load tests........................................... 180 

Table 36. API recommendations for side friction in siliceous soil (API 1993).......................... 189 

Table 37. API recommendations for tip resistance in siliceous soil (API 1993)........................ 190 

Table 38. LCPC friction coefficient αLCPC for driven piles in sands (Bustamante and 

Gianeselli 1982).................................................................................................................. 191 
xvii 

Table 39. LCPC bearing capacity factors for driven piles in sands (Bustamante and 

Gianeselli 1982).................................................................................................................. 191 

Table 40. Predicted axial capacities for the test piles at Route 351............................................ 203 

Table 41. Relationships commonly used for elastic piles in flexion. ......................................... 238 

Table 42. Recommended criteria for p-y curves in different soils (adapted from Reese, 

et al., 1997). ........................................................................................................................ 244 

Table 43. Properties of test piles................................................................................................. 249 

Table 44. Parameters used to define default p-y curves in LPILE for the prestressed 

concrete pile........................................................................................................................ 253 

Table 45. Parameters used to define default p-y curves in LPILE for the FRP pile................... 257 

Table 46. Parameters used to define default p-y curves in LPILE for the plastic pile. .............. 262 

Table 47. Summary of pile-driving measurements for the prestressed and FRP production 

piles (after Spiro and Pais 2002b)....................................................................................... 280 

Table 48. Summary of CASE and CAPWAP analyses results (Spiro and Pais 2002b). ............ 281 

Table 49. Test dates for the instrumented production pile program at the Route 351 Bridge.... 284 

Table 50. Monitoring dates for prestressed concrete pile at the Route 351 Bridge.................... 284 

Table 51. Monitoring dates for FRP composite pile at the Route 351 Bridge............................ 284 

Table 52. Detailed project objectives.......................................................................................... 294 

Table 53. Organization of the figures in appendix A.................................................................. 305 

Table 54. Summary of pushout test results................................................................................. 327 
1 

CHAPTER 1. INTRODUCTION 

1.1 GENERAL INTRODUCTION 

This research project was undertaken to investigate the potential of using composite piles 

for load-bearing applications, specifically bridge substructures. Composite piles refer to 

alternative pile foundations composed of fiber reinforced polymers (FRPs), recycled 

plastics, or hybrid materials that are placed in the ground to support axial and/or lateral 

loads. Traditionally, piles are made of materials such as steel, concrete, and timber. These 

pile materials have limited service life and high maintenance costs when used in harsh 

marine environments (Lampo, et al., 1998). Degradation problems of conventional piles 

include chloride attack on concrete, steel corrosion, and marine borer attack on timber 

piles. It has been estimated that repair and replacement of piling systems costs the United 

States more than $1 billion annually (Lampo, et al., 1997). High repair and replacement 

costs drive the need to investigate the feasibility of using FRP composite materials for 

pile foundations (Lampo, et al., 1998, Iskander and Hassan 1998). FRP composite 

materials are considered an attractive alternative for marine and other harsh environments 

because they are resistant to the degradation mechanisms mentioned above. 

Composite piles have been available in the North American market since the late 1980s, 

but to date their use has been limited mainly to marine fender piles, load-bearing piles for 

light structures, and experimental test piles (Iskander, et al., 2001). Composite piles have 

not yet gained wide acceptance in the civil engineering industry, primarily due to the lack 

of a long track record of performance, and the scarcity of well-documented field load 

tests. However, FRP composite piles may exhibit longer life cycle and improved 

durability in harsh marine environments, thereby presenting the potential for substantially 

reduced costs. Potential disadvantages of using composite piles are related to cost and 

performance. At present, composite piles are generally more expensive than traditional 

piles (Hoy 1995, Iskander and Hassan 1998). Driveability may be less efficient with these 

piles. Structural properties, including low bending and axial stiffness, could result in 

large deformations. Piles with composite exteriors may have low surface friction. The 

scope and extent of potential problems with composite piles are uncertain due to the lack 

of a long-term track record of their use. 

Some of these potential disadvantages may be relative. The apparent higher cost is 

expected to decrease as composite pilings gain wider penetration in the civil engineering 

industry (Iskander and Hassan 1998). Ballinger (1994) pointed out that, although the cost 

of FRP composite materials may be higher, the cost of labor and use of equipment 

necessary for construction work may be lower due to their lighter weight. Furthermore, 

the author suggested that not only should costs be compared on a total installed first-cost 

basis but also on a reasonable total life cycle cost basis. Most composite piling 

manufacturers believe their products may be competitive when compared to the life cycle 

cost of traditional piles in some applications (Ballinger 1994). Some manufacturers claim 

their composite piles may last twice as long as treated wooden piles (Iskander and Hassan 
2 

1998). However, to confidently design facilities using composite piles, engineers need 

more information about composite piles. 

1.2 OBJECTIVES 

The main focus of this research project is to determine the potential of composite piles 

for use in load-bearing applications such as bridge substructures. To confidently establish 

the feasibility of using composite piles for load-bearing structures, more information and 

performance data were gathered in critical areas, such as long-term durability, 

geotechnical behavior (including driveability), and soil-pile load transfer interactions. 

The overall objective of this research project is to establish the feasibility of using 

composite piles in bridge substructures. More specific objectives are to: 

1. Evaluate the long-term durability of concrete-filled FRP composite pipe piles. 

2. Perform laboratory tests to evaluate the soil-pile interface behavior of concrete-filled 

FRP composite pipe piles and compare them with tests carried out on conventional 

prestressed concrete piles. 

3. Design and perform a field test pile program to evaluate the driveability and axial and 

lateral behavior and capacity of a concrete-filled FRP composite pile, and a steelreinforced recycled plastic composite pile, and to compare them to a conventional 

prestressed concrete pile. 

4. Design and perform a production pile testing and monitoring program in a real bridge 

project to compare long-term performance, load transfer, and durability of concretefilled FRP composite piles and conventional prestressed concrete piles. 

1.3 ORGANIZATION OF REPORT 

This report is organized into 11 chapters and 4 appendices. Chapter 2 presents 

background information on composite piles in general, and a literature review with 

specific focus on the two composite piles investigated in this research (i.e., concretefilled FRP composite piles and steel-reinforced recycled plastic composite piles). 

Chapter 3 presents the results of the experimental study on interface behavior between 

composite piles and two sands. A summary of the interface shear test results is given for 

seven pile material types (five composite pile materials and two conventional pile 

materials) against two types of sands. The soil materials, pile materials, and test 

procedures employed are also described in chapter 3. 

Chapter 4 presents the methods and results of the experimental durability study of FRP 

composite piles. Overview and background information related to durability of FRP 

composite materials also is presented in this chapter. 
3 

Chapters 5 and 6 present the findings from the field load tests at the Route 40 and Route 

351 Bridges, respectively. For each field load test program, information is provided 

pertaining to the description of the bridges, the description of the test piles, the soil 

conditions at the test sites, pile fabrication, instrumentation, and installation. The results 

of axial and lateral load tests are presented and discussed for each field load test program. 

Chapter 5 also presents details of the finished Route 40 Bridge, including a description of 

Pier No. 2 of the bridge, which is solely supported by concrete-filled FRP composite 

piles. 

The results from axial and lateral analyses for the field load tests carried out at the Route 

351 Bridge are presented in chapters 7 and 8, respectively. 

In chapter 9, the long-term monitoring program implemented for two production piles at 

the Route 351 Bridge is described. Only preliminary load transfer instrumentation data 

are presented because the bridge is still under construction. The long-term monitoring 

will be carried out by the Virginia Department of Transportation. 

Chapter 10 provides cost information for the composite piles and prestressed concrete 

piles used in this research project. 

Summary, conclusions, and recommendations for future work are presented in chapter 11. 

The results of the laboratory tests performed for the interface behavior study are 

presented in appendix A. The solution for the moisture diffusion into a cylindrical FRP 

composite is presented in appendix B. The results of the structural tests carried out on 

pile cutoff pieces from the Route 40 project are presented in appendix C. These tests 

included a creep bending test and pushout tests to investigate the bond strength between 

the concrete core and FRP shell of the composite piles. Appendix D includes the results 

of geotechnical field investigations carried out at the test site at the Route 351 Bridge 

project.

5 

CHAPTER 2. BACKGROUND AND LITERATURE REVIEW 

2.1 INTRODUCTION 

This chapter presents the general background of composite piles and a literature review, 

with specific focus on the two types of composite piles selected for this research: 

concrete-filled FRP composite piles and steel-reinforced recycled plastic composite piles. 

2.2 COMPOSITE PILE BACKGROUND 

Composite piles refer to alternative pile foundations which make use of FRPs, recycled 

plastics, and other materials to replace and/or protect steel or concrete, with the intent to 

produce piles that have lower maintenance costs and longer service lives than 

conventional piles, especially when used in marine applications and other corrosive 

environments. Examples of deteriorated conventional piles are shown in figure 1. 

Figure 1. Photos. Degradation of conventional piles (Iskander and Hassan 1998). 

Composite piles were first used in the United States in the late 1980s as replacements for 

timber fender piles at the Port of Los Angeles (Heinz 1993). The first composite pile 

prototype was driven in 1987 and consisted of a composite steel pipe encased by recycled 

plastic (Horeczko 1995). Since the first composite pile application in 1987, several other 

projects have used composite piles. A list of selected projects where composite piles have 

been used is presented in table 1. As shown in the table, and reported by Iskander, et al., 

(2001), composite piles have been primarily used as marine fender piles, as load-bearing 

piles for light structures, and as experimental test piles. 

 

2.2.1 Types of Composite Piles 

Several composite pile products are available in the market today, such as steel pipe core 

piles, structurally reinforced plastic matrix piles, concrete-filled FRP pipe piles, 

fiberglass pultruded piles, and plastic lumber piles. Of these five pile types, the first three 
6 

are considered to be better suited for load-bearing applications (Lampo, et al., 1998). 

These three pile types are shown in figure 2. 

Figure 2. Illustration. Common types of composite piles. 

The following subsections describe these three load-bearing composite pile candidates in 

more detail. 

2.2.1.1 Steel pipe core pile 

This pile consists of a recycled plastic shell with a steel pipe core interior, as shown in 

figure 2a. This pile may be obtained in diameters ranging from 8 to 24 inches (20.3 to 

61.0 cm) and up to 75 feet (22.9 m) in length (Plastic Pilings 2001). The steel pipe core 

ranges from 4 to 16 inches (10.2 to 40.6 cm) outer diameter, with pipe wall thicknesses 

between 6 to 40 millimeters (0.236 to 1.57 inches) (Plastic Pilings 2001). 
Table 1. Selected projects involving installation of composite piles. 

Site Year Application Pile Manufacturer Pile Type 

1987 Fender piles Plastic Pilings, Inc. Steel pipe with thick plastic shell 

Port of Los Angeles, CA 1991–5 Fender piles Plastic Pilings, Inc.; Seaward International, Inc.; Hammer’s 

Plastic Recycling 

Plastic piling with steel core 

Port of NJ, Newark, NJ 1991 Fender piles Plastic Pilings, Inc. Steel pipe with thick plastic shell Naval Station Roosevelt Roads, Puerto Rico 1991 Trial fender piles Plastic Pilings, Inc. Steel pipe with thick plastic shell Port of Grays Harbor, Aberdeen, WA 1992–3 Fender piles Plastic Pilings, Inc. Steel pipe with thick plastic shell Port of Seattle, WA 1993 Fender piles Plastic Pilings, Inc. Steel pipe with thick plastic shell Port of Oakland, CA 1993 Fender piles Plastic Pilings, Inc. Steel pipe with thick plastic shell Pearl Harbor, HI 1994 Fender piles Plastic Pilings, Inc. Steel pipe with thick plastic shell 

Seaward International, Inc. Recycled plastic with fiberglass reinforcing Port of NY/NJ 1994 Fender piles Creative Pultrusions, Inc. Ultra high molecular weight polyethylene 

Pier Bravo, NAS North Island, San Diego, CA 1995 Fender piles Plastic Pilings, Inc. Recycled plastic reinforced with welded steel rebar cage Delaware Bay, DE 1996 Fender piles Hardcore Composites Fiberglass shell filled with concrete Port of New Orleans, LA 1996 Fender piles Seaward International, Inc. Recycled plastic with fiberglass reinforcing Pier 16—Naval Amphibious Base Coronado, San Diego, CA 1996 Fender piles Plastic Pilings, Inc. Recycled plastic reinforced with welded steel rebar cage US Navy EMR Facility Pier Ingleside, TX 1997 Pier piles Lancaster Composites FRP shell with concrete core Pier 23, Norfolk, VA 1997 Fender piles Lancaster Composites FRP shell with concrete core 

7
8

The steel core provides the structural strength while the plastic shell protects the pile 

from degradation. The plastic shell can be omitted below the portion of the pile exposed 

to water (Plastic Pilings 2001). If the plastic shell is used only in the upper portion of the 

pile that is exposed to water, the design procedure for this pile would be essentially the 

same as for a conventional steel pipe pile. The plastic shell does not come into play 

structurally, since its only function is to protect the steel pipe along the exposed portion 

of the pile. Since the design procedure of steel pipe piles is well established, it was 

decided that there was relatively little need for further research on this kind of pile. 

2.2.1.2 Structurally reinforced plastic matrix pile 

This composite pile consists of a recycled plastic matrix structurally reinforced with FRP 

rods or a welded steel rebar cage. The typical configuration of this type of pile is shown 

in figure 2b. An attractive feature of this pile is that when reinforced with FRP rods, the 

pile is nonmagnetic and 100 percent recyclable (Lindsay 1996). This environmentally 

friendly composite pile type uses approximately 240 recycled 1-gallon (3.79-L) milk jugs 

per linear foot (0.305 m) of a 12-inch (0.305-m) nominal diameter pile (Taylor 1995). 

The piles are commercially available in diameters ranging from 8 to 24 inches (0.203 to 

0.610 m), and are reinforced with 6 to 24 FRP or steel reinforcing rods, which range 

between 0.75 to 1.41 inches (1.91 to 3.58 cm) in diameter (Plastic Pilings 2001, Seaward 

International 2001). 

These piles are produced by a continuous extrusion process that enables the production of 

continuous pile of virtually any length (Lindsay 1996, Taylor 1995). This continuous 

length precludes design problems inherent in segmented piles (Horeczko 1995). To 

enhance durability, the plastic matrix is chemically treated with antioxidants and 

ultraviolet (UV) inhibitors that retard the effects of UV light on plastic (Taylor 1995, 

Iskander and Hassan 1998). 

To date, these piles have been used primarily as fender piles in several naval facilities 

(Hoy 1995). Lindsay (1996) reported preliminary tests on 12-inch (0.305 m) diameter, 

60-ft (18.3 m) long piles. The results indicated the piles could carry axial working loads 

of up to 10 tons (89 kN) and 20 tons (178 kN) per pile when reinforced with FRP and 

steel rods, respectively. Some potential problems associated with these piles include the 

possibility of debonding of the reinforcing FRP rods and the potential for high creep rates 

associated with the high polymeric content of these piles. One version of this pile is 

structurally reinforced with a reinforcement cage with the rebars welded to a continuous 

steel spiral. This version of the pile is reported not to have the aforementioned problems 

(Plastic Pilings 2001). 

Although this type of composite pile is considered to have potential for load-bearing 

applications, it was not initially selected for the second phase of the project due to budget 

limitations. The late withdrawal of one of the pile manufacturers allowed inclusion of this 

pile at the last minute. 
9

2.2.1.3 Concrete-filled FRP tube pile 

In a concrete-filled FRP composite pile there are two main structural components: an 

FRP shell or tube, and a concrete infill without steel reinforcement. The FRP shell 

provides, among other things, a stay-in-place concrete form, confinement to the concrete, 

tensile reinforcement, and corrosion protection (Fam and Rizkalla 2001a, b). The 

concrete infill provides compressive load capacity. The two main manufacturers of this 

type of pile in the United States are Hardcore Composites and Lancaster Composite 

(Iskander, et al., 2001). 

Hardcore piles can be installed by driving the empty FRP shell and then filling it with 

concrete, although they are also installed by filling with concrete and then driving after 

the concrete has cured (Hardcore Composites 2001). Lancaster Composite’s piles are sold 

under the commercial name CP40, and are filled with concrete and allowed to cure prior 

to driving (Stapleman 1997, Lancaster Composite 2000). 

The FRP shells used by Hardcore Composites are fabricated using a vacuum-assisted 

resin transfer molding process (VARTM), while Lancaster Composite’s FRP shells are 

made using a filament winding technique. Typically, both piles are available in diameters 

ranging from 8 to 24 inches (0.203 to 0.610 m), with wall thicknesses ranging between 

0.18 to 0.36 inches (0.46 to 0.91 cm). Both manufacturers have the flexibility to modify 

the FRP composite shell laminate architecture (including number of layers, fiber 

orientation, and resin and fiber types) to better suit the load demands of a specific project. 

These piles can be produced in virtually any length (Lancaster Composite 2000, Hardcore 

Composites 2001). 

This type of composite pile was considered to have good potential for load-bearing 

applications, hence it was selected for the second phase of the research project. 

 

2.3 LITERATURE REVIEW FOR THE COMPOSITE PILES SELECTED 

The initial research plan entailed focusing the study on concrete-filled composite piles, 

which were selected after the initial pile screening phase. The piles were to be 

manufactured by Lancaster Composite, Inc., and Hardcore Composites. The late 

withdrawal of Lancaster Composite, Inc. allowed the inclusion of a recycled plastic 

composite pile manufactured by Plastic Pilings, Inc. Information available for concretefilled FRP composite piles and steel-reinforced recycled plastic composite piles is 

presented in the following subsections. 

2.3.1 Structural Behavior 

 

2.3.1.1 Concrete-filled FRP composite piles 

Considerable effort has been expended to research the structural behavior of concretefilled FRP composite tube piles (e.g., Lampo, et al., 1998, Samaan, et al., 1998, 

Mirmiran, et al., 1999, Fam 2000, Fam and Rizkalla 2001a, b, Mirmiran, et al., 2001, 
10

Moran and Pantelides 2002a, Becque, et al., 2003). Relevant findings from selected 

research studies are described in the following subsections. 

 

2.3.1.1.1 Behavior under axial loading 

The structural behavior of concrete-filled FRP composite tubes under axial compression 

has been studied extensively and several analytical models have been proposed to predict 

their response. Table 2 lists these models. 

 

Table 2. Expressions for the strength of confined concrete and maximum axial 

strain (adapted from Shehata, et al., 2002). 

Author Type of 

Confinement Ultimate Strength fcc Ultimate Axial Strain εcc 

Fardis and 

Khalili 

(1981) 

GFRPencased 

concrete 

Karbahari 

and Eckel 

(1993) 

FRP-encased 

concrete 

Mirmiran 

and 

Shahawy 

(1997)* 

GFRPencased 

concrete 

— 

Miyauchi, 

et al. 

(1997)** 

CFRPwrapped 

concrete 

Samaan, et 

al. (1998)* 

GFRPencased 

concrete 

Saafi, et al. 

(1999) 

CFRP and 

CFRP encased 

concrete 

Toutanji 

(1999) 

CFRP and 

GFRP 

wrapped 

concrete 

Spoelstra 

and Monti 

(1999) 

CFRP and 

GFRPwrapped and 

encased 

concrete 

Notes: * units are in MPa, ** Equation for εcc is valid for fco = 30 MPa 

CFRP = Carbon fiber-reinforced polymer, GFRP = Glass fiber-reinforced polymer, 

fco = compressive strength of unconfined concrete, εco = ultimate strain of unconfined concrete, 

Ef = Hoop modulus of elasticity of FRP tube, ff = Hoop tensile strength of FRP tube, 

fl = lateral confining stress, tf = thickness of FRP tube, D = diameter of concrete core. 
11

Recently, other models have been proposed by Fam (2000), Fam and Rizkalla (2001a, b), 

Moran and Pantelides (2002b), Shehata, et al. (2002), and Becque, et al. (2003). Most of 

these models build upon the simple observation that the typical stress-strain curve of 

concrete-filled FRP composite columns has an approximately bilinear shape, as shown in 

figure 3. 

Figure 3. Graph and Photos. Confinement effect of FRB tube on concrete 

(Fam and Rizkalla 2001a, b). 

As shown in figure 3, the FRP tube of a composite pile contributes structurally to the pile 

by resisting some of the axial load, and by providing confinement to the concrete core. 

The beneficial effect of confinement on the total load-carrying capacity of a short, 

concrete-filled FRP tubular element was studied by Fam and Rizkalla (2001a). As 

illustrated in figure 3, the capacity of the composite stub significantly exceeds the loadsharing capacity of the two individual materials (Fam and Rizkalla 2001a). The loadstrain curve starts to depart from the unconfined concrete curve in the vicinity of the 

unconfined concrete strength. As this stress level is approached, the concrete core starts 

to experience significant microcracking as well as increased lateral expansion. In 

response to the lateral expansion of the concrete, the FRP shell applies a radial confining 

pressure, which continuously increases due to its linear elastic properties (Fam 2000). 

The second slope of the load-strain curve is a function of the hoop tensile stiffness of the 

FRP shell, and the ultimate peak strength is governed by the hoop tensile strength of the 

FRP shell. A more detailed description of the behavior of concrete-filled FRP tubes under 

axial compression loading can be found in Fam and Rizkalla (2001a, b). 
12

The short-term axial capacity of a concrete-filled FRP tube can be predicted using a 

confinement model such as the one proposed by Fam and Rizkalla (2001b). This model is 

an incremental variable confinement model that satisfies equilibrium and radial 

displacement compatibility between the concrete core and the FRP tube, and it makes use 

of the constant confinement model of Mander, et al. (1988). 

The radial confinement pressure applied by the FRP shell can be obtained from 

equilibrium and by imposing radial displacement compatibility between the concrete core 

and the FRP shell. When both the concrete core and FRP shell are axially loaded with the 

same strain level (εcc), the radial confinement pressure (σR) can be estimated as follows 

(Fam 2000): 

 (1) 

where 

 

• υc is the Poisson ratio of concrete (can be varied by using the secant Poisson’s ratio) 

• υFRP is the Poisson ratio of the FRP shell (-hoop strain /axial strain under axial 

loading) 

• RFRP and tFRP are the radius and wall thickness of the FRP tube, respectively 

• EFRP-Hoop is the elastic modulus of the FRP shell in the hoop direction (hoop tension) 

• Ec is the elastic modulus of the concrete core (can be varied by using the secant 

modulus) 

• εcc is the axial strain of the concrete-filled FRP tube stub 

The variable radial confining pressure is obtained from equation 1 by using the secant 

elastic modulus and the secant Poisson’s ratio of the concrete as a function of axial strain 

level. 

The model relates the axial stress of the confined concrete core (fcc), at any given axial 

strain level (εcc), to the peak confined strength of the concrete (f′cc) as follows: 

 (2) 

where 

 (3) 

and where 

• ε′cc is the strain at peak strength f ′cc

• r = Eco ÷ ( Eco–Esec) 
13

• Eco is the initial tangent elastic modulus of unconfined concrete, and can be estimated 

as Eco(MPa) ≈ 5000*(square root of f ′c ), where f ′c is the strength of the unconfined 

concrete in MPa

• Esec is the secant elastic modulus of confined concrete (≈ f ′cc / ε′cc) 

The peak confined strength (f ′cc) can be estimated using the following expression 

proposed by Mander, et al. (1988): 

 (4) 

The load-strain response, predicted using the model shown above, and the experimental 

load-strain behavior obtained for Test Stub No. 1 (from Fam and Rizkalla 2001a), are 

shown in figure 4. The figure shows good agreement between the model and the 

experimental results. 

Figure 4. Graph. Experimental versus predicted load-strain behavior 

using Fam and Rizkalla’s model. 

This model will be used later in chapter 4 to illustrate the influence that the degradation 

of the hoop properties (stiffness and strength) of FRP tube has on the long-term axial 

capacity of concrete-filled tubular FRP piles. 

2.3.1.1.2 Flexural behavior 

Research on the flexural behavior of concrete-filled FRP circular tubes includes studies 

by Mirmiran (1999), Fam (2000), Mirmiran, et al. (2000), Fam and Rizkalla (2000), 

Mirmiran, et al. (2001), and Fam and Rizkalla (2002). These structural elements can be 

used to resist bending moments, but the benefits of concrete confinement are less in 

bending than the purely axial loading case (Fam 2000). In flexure, the FRP tube acts as a 

noncorrosive reinforcement, while the concrete provides the internal resistance force in 
14

the compression zone and increases the stiffness of the member (Fam and Rizkalla 2002). 

The concrete core also prevents local buckling of the FRP tube. 

A comprehensive experimental program to study the short-term flexural behavior and 

failure modes of concrete-filled FRP tubes was recently completed by Fam and Rizkalla 

(2002). Their study showed that the flexural behavior is highly dependent on the stiffness 

and diameter-to-thickness ratio of the FRP tube, and to a lesser extent on the concrete 

strength. The study also showed that, in general, the cracking moment resistance is 

relatively small compared to the ultimate moment capacity 

The short-term flexural capacity of a concrete-filled FRP tube can be predicted using a 

strain compatibility/equilibrium model such as the one proposed by Mirmiran and 

Shahawy (1996) or Fam (2000). The methodology assumes that sections normal to the 

neutral axis remain plane after bending, and that the FRP shell is perfectly bonded to the 

concrete interior (i.e., a linear strain distribution through the cross section). The method 

consists in discretizing the pile cross section into a series of strip elements for integrating 

the normal stresses over the cross-sectional area. The discretization results in both FRP 

and concrete strip elements, as shown in figure 5. The stress integration is done by 

assigning the appropriate constitutive model for each material. Details of the 

development of the model and its validation through experimental results can be found in 

Fam (2000) and Fam and Rizkalla (2002). 

Figure 5. Illustration. Strip elements for sectional analysis 

(Mirmiran and Shahawy 1996). 

 

The load-curvature response, predicted using the model just described, and the 

experimental load-strain behavior for Beams 4 and 13 from Fam and Rizkalla (2002), are 

shown in figure 6. The figure shows good agreement between the model and the 

experimental results, especially when accounting for tension stiffening. 
15

Figure 6. Graphs. Experimental versus analytical moment-curvature response 

(adapted from Fam and Rizkalla 2002). 

This model will be used later in chapter 4 to illustrate the influence that the degradation 

of the FRP tube properties (stiffness and strength) has on the long-term axial capacity of 

concrete-filled tubular FRP piles. 

2.3.1.1.3 Behavior under combined bending and axial loading 

Mirmiran and Shahawy (1999) performed a detailed study on concrete-filled FRP tubes 

under various combinations of axial and flexural loads. Two types of FRP tubes were 

used to simulate the conditions of overreinforcement (where compression failure 

governs) and underreinforcement (where tension failure governs). The overreinforced 

specimens were prepared using tubes 348 mm (13.7 inches) in diameter with a wall 

thickness of 14 mm (0.551 inches), while the underreinforced specimens used 369-mm 

(14.5-inch) tubes with a 6-mm (0.236-inch) wall thickness. The reinforcement ratios 

(ratio of area of FRP shell to area of concrete core) for the over-and underreinforced 

concrete-filled tubes were 18.27 and 7.56 percent, respectively. Since the strength of the 

FRP tubes differed in these specimens, the authors proposed the use of a reinforcement 

index, defined as the reinforcement ratio multiplied by the ratio of the axial tensile 

strength of the FRP tube to the concrete compressive strength. The reinforcement indices 

of the over- and underreinforced sections were 3.39 and 0.19, respectively. The 

interaction diagrams for the two types of specimens are shown in figure 7. 
16

Figure 7. Graph. Interaction diagrams, concrete-filled FRP tubes (Mirmiran 1999). 

The overreinforced specimens were found to behave better as beam columns. They 

deflected to a lesser extent (ultimate deflections of the overreinforced specimens were 

about 25 to 50 percent lower than the underreinforced specimens), and failed at much 

higher bending moments. Failure of the overreinforced specimens while in compression 

was considered to be gradual or ductile. The underreinforced failure mode was brittle and 

sudden. Based on this study, the authors concluded that concrete-filled FRP tubes could 

be used for beam-column applications, and recommended the use of overreinforced 

specimens. Test observations also indicated that bond failure or slippage in beam 

columns is not as significant as in beam specimens (pure flexure), as long as the end 

connections are designed properly. For beam specimens, shear transfer mechanisms such 

as internal ribs or treatments of the inner surface of the tubes were recommended to 

enhance the composite action between the FRP shell and the concrete core. 

Design of concrete-filled FRP tubes for this type of loading can be done using the same 

sectional analysis as described in section 2.3.1.1.2 for pure flexure. This analysis 

methodology can be easily programmed and used to generate moment-curvature and 

interaction diagrams. A more detailed description of the methodology can be found in 

Mirmiran and Shahawy (1996), Mirmiran (1999), and Fam (2000). 

 

2.3.1.2 Steel-reinforced recycled plastic composite piles 

Research on the structural behavior of steel-reinforced recycled plastic piles is notably 

absent in published literature. Information about the performance of these piles is limited 

to reports commissioned by pile manufacturers. Table 3 lists the available structural 
17

information from the manufacturer. This information consists of test results, and it does 

not include design methods. 

Table 3. Available structural information for steel-reinforced 

plastic composite piles. 

Type of test Source 

Three-point bend test Asaro (2000) 

Limited flexural tests Warren (1996) 

Axial compression test Stokes (2002) 

Iskander, et al. (2003), studied the compressive strength of the plastic material used for 

these kinds of piles. The results showed a marked nonlinear behavior and significant 

spatial distribution of strength and density within the specimens. However, the results 

from tests shown in table 3 indicate that the majority of the structural capacity and 

stiffness comes from the welded steel rebar cage. 

 

2.3.2 Long-Term Durability Behavior 

The lack of long-term performance and durability data on composite piles is a concern 

(Lampo, et al., 1998, Iskander and Hassan 1998, Iskander, et al., 2001). There is an 

urgent need for research in this area. 

2.3.2.1 Concrete-filled FRP composite piles 

The principal mechanisms for degradation of the FRP composites used in piles are 

moisture absorption over time and freeze-thaw cycles. Chemical degradation and UV 

radiation also affect long-term durability, but to a lesser extent. Furthermore, UV 

degradation resistance of most composites is improved by applying protective coats and 

additives during the manufacturing process. The degradation mechanisms related to 

moisture absorption and freeze-thaw cycles are discussed below. 

Degradation due to moisture absorption may significantly reduce the life of FRP 

composites (Garcia, et al., 1998). Absorbed moisture can cause pronounced changes in 

modulus, strength, and strain to failure (Springer, et al., 1980). The moisture content of 

submerged FRP composites increases through diffusion. The absorbed moisture can act 

as a plasticizer of the composite resin, and can cause matrix cracking, fiber-matrix 

debonding, and corrosion of glass fibers (stress corrosion) (Garcia, et al., 1998). These 

effects result in a reduction of strength and stiffness of the FRP composite. For example, 

Schulheisz, et al. (1997), recorded strength and stiffness reductions on the order of 20 

percent and 5 percent, respectively, for E-glass/vinyl ester composites submerged in 
18

25 o

C (77 °F) water for a period of 200 days. The implications of such strength and 

stiffness reductions on the design of composite piles can be significant. 

In addition to submergence time, temperature and stress level also influence the amount 

of moisture that FRP will absorb while submerged. An FRP composite subject to tensile 

stresses is expected to have a faster rate of moisture absorption. Figure 8 shows 

schematically how time and temperature influence moisture absorption, how moisture 

absorption degrades property values, and how exposure time thus produces a reduction in 

property values. Evidence of the existence of damage to FRP composites after 

submergence in water is illustrated in the scanning electron microscope (SEM) images 

shown in figure 9. The SEM image of the submerged sample shows fiber damage and 

crack formation at the fiber-matrix interface (McBagonluri, et al., 2000). 

Figure 8. Graphs. Moisture absorption-related durability model. 
19

Figure 9. Images. SEM images showing FRP damage (McBagonluri, et al., 2000). 

Freeze-thaw cycles can also degrade FRP composites. Freeze-thaw cycles cause the 

volume of absorbed water to increase. Such volume changes lead to crack propagation 

and damage to FRP (Verghese, et al., 1999). 

 

2.3.2.2 Steel-reinforced recycled plastic composite piles 

Iskander and Hassan (2001) carried out an accelerated degradation study on the recycled 

plastic material used in the plastic piling manufactured by Seaward International. Their 

study involved compression tests on coupon samples submerged in aqueous solutions 

with pH ranging between 2 and 12. A 25 percent loss in compressive resistance was 

predicted to take place after a 21-year exposure to the types of aqueous solutions used. 

The durability of steel-reinforced plastic piles installed at Pier Bravo in NAS North 

Island, San Diego, CA was evaluated in 1999. The piles were manufactured by Plastic 

Pilings, Inc. and were installed in December 1995. Inspections carried out 3.5 years after 

installation showed no significant degradation. 

2.3.3 Geotechnical Behavior 

There has been little research to date on geotechnical issues related to either concretefilled FRP composite or steel-reinforced recycled plastic piles. Some important issues are 

pile driveability, lateral load capacity, and axial load capacity. These issues are discussed 

in the following subsections. 
20

2.3.3.1 Pile driveability

To date, very few case histories are available with monitored driving information of 

composite piles. Parametric or analytical driveability studies have been recently 

performed (e.g., Ashford and Jakrapiyanun 2001, Iskander, et al., 2001, Iskander and 

Stachula 2002) using wave equation analysis (e.g., computer code GRLWEAPTM (GRL 

Wave Equation Analysis Program)) to investigate the feasibility of driving FRP 

composite piles efficiently and without damage, and to compare their performance to 

conventional steel and concrete piles. These analytical studies found that composite piles 

can attain equivalent design capacity of conventional piles, but the low impedance of 

some composite piles may limit their driveability and pile capacity. These studies did not 

involve any field testing. The sparse field information available has been mostly 

generated by pile manufacturers. 

A recent study involving field data not generated by pile manufacturers was carried out 

by Mirmiran, et al. (2002). This study involved driving four FRP composite piles and one 

prestressed concrete pile. The FRP tubes used for the piles had an outside diameter of 

0.348 m (1.14 ft) and a wall thickness of 14 mm (0.546 inches). Two of the composite 

piles were driven with concrete infill and the other two were driven empty with a steel 

cone tip. The total length of the FRP piles was between 7.9 and 9.1 m (25.9 and 29.8 ft). 

The prestressed concrete pile was 0.356 m (1.17 ft) square and 9.1 m (29.8 ft) long. The 

filled FRP piles and the prestressed concrete pile were driven to depths of about 7.3 m 

(23.9 ft). The authors found that driving stresses in filled tubes were comparable to those 

for the prestressed concrete pile. The empty tubes were found to be susceptible to 

buckling and damage during driving, unless driven to shallow depths in soft soils or with 

a steel mandrel. No geotechnical information was provided about the site where the field 

tests were carried out. 

It is clear that a significant amount of research is needed in this area. Particularly, there is 

an urgent need for more field tests carried out in a controlled fashion at wellcharacterized sites to carefully assess and verify the driveability of these new piles. 

In general, pile driveability depends on the energy delivered to the pile by the piledriving hammer, the resistance to driving offered by the soil, the ability of the pile to 

transfer driving stresses to the pile tip, and the strength of the pile to resist driving 

stresses. Good driving occurs when the hammer effectively transmits energy to the pile 

and the induced stress wave develops a force in the pile sufficient to overcome the soil 

resistance. Poor driving occurs when the transmitted energy is too low to develop a force 

capable of overcoming the soil resistance. If the stresses induced during driving are 

higher than the strength of the pile, the pile will be damaged. 

The ability of the pile to transfer the energy imparted by the driving hammer into a force 

in the pile is related to the pile impedance (or dynamic stiffness). Impedance is defined as 

Z = EA/c, where E is the elastic modulus of the pile, A is the pile cross-sectional area, 

and c is the wave propagation speed in the pile. The greater the impedance of the pile, the 

greater is the force that will be transmitted by the pile into the ground. Table 4 compares 
21

typical impedance values of four selected composite piles with impedances of three 

conventional piles of similar cross sections. 

Table 4. Comparison of pile impedance. 

Pile Description 

Impedance 

E A MN s

c m

⋅ ⎛ ⎞ ⋅ ⎜ ⎟ ⎝ ⎠

Impedance 

Ratio1 Source 

12 inch square prestressed concrete 750 8 Iskander and Stachula 2002, 

Ashford and Jakrapiyanun 2001

12.8 inch diameter concrete-filled 

FRP 691 7.4 Ashford and Jakrapiyanun 2001

10.6 inch diameter plastic encased 

steel pipe 412 4.4 Iskander and Stachula 2002 

13.4 inch diameter steel pipe pile 

(0.374 inch wall thickness) 391 4.2 Ashford and Jakrapiyanun 2001

12.4 inch diameter timber pile 266 2.8 Iskander and Stachula 2002 

15.7 inch diameter recycled plastic 

pile (reinforced with FRP tendons) 215 2.3 Iskander and Stachula 2002 

14 inch diameter hollow FRP shell 

(0.51 inch wall thickness) 93.7 1 Ashford and Jakrapiyanun 2001

Note: (1) Impedance ratio = Pile impedance/Impedance of hollow FRP pile. 

1 inch = 25.4 mm. 

Table 4 shows the impedance ratio of each pile compared to the pile of lowest impedance 

(the empty FRP shell). It can be seen that the prestressed concrete pile and the concretefilled FRP pile have similar impedance values, and they have the highest values. The 

plastic-encased steel pipe pile and the FRP-reinforced recycled plastic pile impedance 

values are about 55 percent and 29 percent of the impedance of the prestressed concrete 

pile, respectively. The lowest impedance corresponds to the empty FRP shell with a value 

equal to about 12.5 percent of the prestressed concrete pile. 

2.3.3.2 Lateral load capacity 

The design of piles for use against lateral loads is usually governed by the maximum 

tolerable deflection (Poulos and Davis 1990). Lateral deflections of single piles depend 

on the lateral load, the bending stiffness (EI) of the pile, and the soil resistance to lateral 

movement (characterized by soil strength and stiffness). Based on the available structural 

studies of concrete-filled FRP composite pipe piles mentioned earlier, it is expected that 

these composite piles will generally have lower bending stiffness than the conventional 

prestressed concrete piles commonly used for bridge substructures. Therefore, lateral 

deflections of composite piles are expected to be greater than for conventional piles 

subjected to the same lateral loads. The higher flexibility of these composite piles could 

prove to be very important in cases where lateral deflections control design. The reduced 

bending stiffness will also have an important impact on the load transfer mechanism and 

the depth to which lateral soil resistance is mobilized. 
22

For laterally loaded “conventional” piles, it is common practice to analyze the loaddeflection response by using analytical methods such as the Winkler Method (subgrade 

reaction method), elastic continuum theory, p-y method, and finite element-based 

methods. Detailed descriptions of these methods can be found elsewhere (e.g., Reese 

1984, Poulos and Davis 1990). All of these methods tend to model the pile as an elastic 

beam. However, for composite piles, this assumption may no longer be acceptable. Han 

(1997) and Han and Frost (1997) pointed out that to reasonably predict the loaddeflection response of a laterally loaded composite pile, the shear deformation effects 

should be taken into account. This issue arises due to the fact that composite materials 

have considerably lower shear modulus (G) than conventional materials (Scott, et al., 

1998). Therefore, strictly speaking, the classical Bernoulli-Euler beam theory, which 

ignores shear deformation, is not applicable (Bank 1989, Han and Frost 1997). For 

example, a study by dos Reis and Goldman (1987) on FRP composite tubular shells 

indicated that deflections predicted using classical beam theory yielded deflections about 

30 percent lower than predictions using finite element analysis that incorporated the 

effect of shear deformations. Han and Frost (1997) did a theoretical study that extended 

the existing elastic continuum solution to include shear deformation effects and pile-soil 

slip. Their solution, from the theoretical point of view, offers a reasonable design 

approach for composite piles. However, their model is quite complex and requires 

considerable computational effort. Also, their model has not yet been confirmed by 

model or full-scale tests of composite piles. Certainly more research is required in this 

area. Further research should aim not only to improve understanding of the loaddeflection response of composite piles, but also to develop reliable and easy-to-use 

design procedures that can be readily implemented by practitioners. 

 

2.3.3.3 Axial load capacity 

The behavior of piles during axial loading depends on the axial stiffness (AE) of the pile, 

residual stresses left in the pile and the soil after pile installation, and the resistance of the 

soil to pile downward movement. 

The axial stiffness of concrete-filled FRP composite pipe piles is expected to be 

comparable to that of conventional prestressed concrete piles, since the composite axial 

stiffness of these piles is largely determined by the stiffness of the concrete infill. 

An important potential difference between composite piles and conventional piles is that 

the shear strength at the interface between the pile and the soil may be different due to 

unique interface properties such as lower surface hardness and different surface 

roughness. Such differences could have a significant influence on the applicability of 

standard procedures for estimating pile skin friction. To date, most laboratory studies 

published in this area involve interface shear tests on traditional pile materials. Relatively 

few studies are available on interface shear behavior between FRP composite pile 

materials and soils. 

Han (1997) carried out an experimental study that compared the interface friction 

between an E-glass/polyester composite sheet pile and sand with the interface friction 
23

between mild steel (A36) and sand. In general, Han found that the FRP specimen was 

relatively rougher than the steel specimen, which resulted in higher peak interface friction 

angles for the FRP composite than for the steel. However, issues like surface roughness 

degradation were not explored. Interface creep deformations were not studied either. 

Interface shear test data for other types of composite piles are not currently available. 

Clearly, this area requires a considerable amount of study. 

 

2.4 SUMMARY 

This chapter provides background information on composite piles and reviewed literature 

related to the two pile types used in the research project: concrete-filled FRP piles and 

steel-reinforced recycled plastic piles. Research on structural behavior, long-term 

durability behavior, and geotechnical behavior for each pile type was reviewed. The 

literature review indicates that the following knowledge gaps exist. 

1. While research on the structural behavior of concrete-filled FRP piles has received a 

fair amount of attention, similar research on steel-reinforced recycled plastic piles is 

restricted to reports commissioned by pile manufacturers. 

2. Long-term durability research for both pile types is limited, and further study is 

needed to understand the effects of moisture absorption and freeze-thaw cycles. 

3. There is only a small body of research on geotechnical behavior of composite piles. 

There is a need for more controlled field tests at well-characterized sites to carefully 

assess and verify pile driveability. Future research should aim to improve 

understanding of the load-deflection response for composite piles, and also to develop 

reliable and easy-to-use design procedures for practitioners.

25

CHAPTER 3. EXPERIMENTAL STUDY OF INTERFACE 

BEHAVIOR BETWEEN COMPOSITE PILES AND TWO SANDS 

3.1 INTRODUCTION 

The properties of materials used in composite piles can differ significantly from those of 

conventional pile materials (e.g., steel and concrete), including differences in anisotropy, 

stiffness, surface hardness, and surface roughness. These differences may produce 

differences in the behavior of interfaces between the piles and the surrounding soil. 

Interface behavior plays an important role in the ultimate capacity and load transfer 

characteristics of the pile, as shown schematically in figure 10. For example, the axial 

pile capacities of a composite pile and a conventional pile, with identical dimensions and 

soil conditions, may be quite different due to differences in the developed shaft capacity. 

To date, most laboratory studies published in this area involve interface shear tests on 

traditional pile materials. Relatively few studies are available on interface shear behavior 

between composite pile materials and soils. The principal focus of this chapter is on the 

soil-pile interface behavior characteristics of composite piles against sand. This chapter 

summarizes interface shear tests results for seven pile material types (five composite pile 

materials and two conventional pile materials) against two types of sands. The interface 

behavior of the different piles is studied within the geotribology framework proposed by 

Dove and Jarrett (2002). 

This chapter is divided into five sections: 

 

1. Soil materials. Describes the gradation and strength properties of each of the soils 

used for interface testing. 

2. Pile surfaces. Describes the surface texture and hardness of the different interface 

materials tested. 

3. Interface shear tests. Describes the equipment and procedures used for interface 

shear testing, and summarizes the results obtained. 

4. Discussion of results. 

5. Summary. 
26

Figure 10. Illustration. Influence of soil-pile friction on pile capacity. 
27

A matrix summarizing the tests carried out for the interface behavior study is provided in 

table 5. 

Table 5. Interface behavior test matrix. 

Type of Test Type of Soil Type of Pile 

Surface Comments 

Direct shear test Density sand Not applicable Average Dr = 70% 

Direct shear test Density sand Not applicable Average Dr = 100% 

Direct shear test Model sand Not applicable Average Dr = 75% 

Interface shear test Density sand Lancaster CP40 Curved specimen, 24-inch 

nominal diameter 

Interface shear test Model sand Lancaster CP40 Curved specimen, 24-inch 

nominal diameter 

Interface shear test Density sand Hardcore 24-4 Curved specimen, 24-inch 

nominal diameter 

Interface shear test Model sand Hardcore 24-4 Curved specimen, 24-inch 

nominal diameter 

Interface shear test Density sand Hardcore FRP, 

no bonded sand 

Flat plate, surface untreated 

with no acrylic coating 

Interface shear test Model sand Hardcore FRP, 

no bonded sand 

Flat plate, surface untreated 

with no acrylic coating 

Interface shear test Density sand Hardcore FRP, 

with bonded sand 

Flat plate, textured surface 

with bonded sand 

Interface shear test Model sand Hardcore FRP, 

with bonded sand 

Flat plate, textured surface 

with bonded sand 

Interface shear test Density sand Plastic Piling 

recycled plastic 

Flat coupons from prismatic 

rectangular beams 

Interface shear test Model sand Plastic Piling 

recycled plastic 

Flat coupons from prismatic 

rectangular beams 

Interface shear test Density sand Concrete Coupon from prestressed 

concrete pile 

Interface shear test Model sand Concrete Coupon from prestressed 

concrete pile 

Interface shear test Density sand Steel Coupon from sheet pile 

Interface shear test Model sand Steel Coupon from sheet pile 

1 inch = 25.4 mm. 
28

3.2 SOIL MATERIALS 

Two sands were used in this study: Density sand and a Model sand. The Density sand is a 

fine- to medium-grained, silica sand with subrounded to rounded grains, and the Model 

sand consists of a blend of Light Castle sand with Bedding sand. The Model sand consists 

predominantly of fine-grained sand with subangular to angular grains. The Density sand 

was used in a previous study by Gómez, et al. (2000). The Model sand was prepared to 

mimic the foundation soils of the pile test site at the Route 351 Bridge. 

3.2.1 Index Properties 

Index parameter values for both sands used in this study are listed in table 6. Grain size 

distribution curves for these sands are shown in figure 11. The Density sand has grain 

sizes ranging from 0.2 to 0.9 mm (0.0078 to 0.035 inches) with no fines. The Model sand 

has grain sizes ranging from 0.1 to 2 mm (0.0039 to 0.078 inches) with about 5 percent 

fines. Microscopic views of the sands are shown in figure 12. 

 

Table 6. Index parameter values of the sands used in this study. 

Parameter Density Sand (1) Model Sand ASTM Standard 

D10 (mm) 0.3 0.08 

D50 (mm) 0.5 0.18 

Cu 1.8 2.6 

Cc 1.1 0.8 

D2487 

γmax (kN/m3) 17.5 17.0 D4253 

γmin (kN/m3) 15.1 14.1 D4254 

Gs 2.65 2.65 D854 

 Note: (1) After Gómez, et al. (2000); 1 mm = 0.039 inches; 1 kN/m3 = 6.36 lbf/ft3. 
29

Figure 11. Graph. Grain size curves to test sands. 

Figure 12. Photos. Microscopic views of the test sands. 
30

3.2.2 Direct Shear Tests of Sands 

Direct shear tests were performed to determine the internal friction angles of the Density 

and Model sands. Displacement controlled direct shear tests with a 101.6 mm (4-inch) 

square shear box were performed. All tests were carried out at a horizontal displacement 

rate of 0.9 mm/minute (min) (0.036 inch/min). The direct shear tests were performed in 

accordance with American Society for Testing and Materials (ASTM) Standard D3080-

98 (ASTM 1998). Samples were prepared by means of air pluviation and tamping. Dry 

sand was poured through a 6.5 mm (0.254 inch) funnel positioned at 10 mm (0.39 inch) 

above the top of the sand surface. Increased density was achieved by means of a small 

tamping device. To minimize the friction between the sand and the soil box, the inside of 

the walls was coated with a thin film of vacuum grease. 

Direct shear test results, in terms of shear stress versus horizontal displacement curves, 

are presented in figures 13 through 15 for the medium-dense to dense Density and Model 

sands, respectively. For the entire range of relative densities and normal pressures tested, 

all of the direct shear tests exhibited a peak shear stress followed by a stress reduction 

toward a residual shear stress with a nearly constant volume state. The peak and residual 

(nearly constant volume state) friction angles obtained from the direct shear tests are 

summarized in table 7. 

Table 7. Internal friction angles obtained from direct shear tests. (1)

Parameter Density Sand 

(Average Dr = 

70%) 

Density Sand 

(Average Dr = 10%) 

Model Sand 

(Average Dr = 

75%) 

Peak φ′p 34.7 o

 39.3 o 43.4 o

Residual φ′cv 29.8 o

 29 o 36.2 o

Notes: (1) Four normal stresses were used: 25, 50, 100, and 200 kPa (3.63, 7.25, 14.5, and 29 lbf/inch2). 
31

Figure 13. Graphs. Direct shear test results for Density sand (average Dr = 70 %). 
32

Figure 14. Graphs. Direct shear test results for Density sand (average Dr = 100 %). 
33

Figure 15. Graphs. Direct shear test results for Model sand (average Dr = 75%). 
34

In table 7, Dr refers to the sample’s relative density, which can be calculated as follows: 

 (5)

where 

• γmax and γmin = are the maximum and minimum dry unit weights for each sand (values 

given in table 6) 

• γdry = dry unit weight of the samples in the direct shear tests 

Both linear and curved-peak shear strength envelopes are shown in figures 13 through 15. 

The nonlinear shear strength envelopes were obtained using the following expression 

proposed by Duncan, et al. (1980): 

 (6)

where 

• φo = peak secant friction angle at a normal stress equal to the atmospheric pressure (Pa

= 101.4 kPa) 

• Δφ = reduction in peak secant friction angle for a tenfold increase in normal stress, σ′n

The values of φo and Δφ for both sands are shown in figures 13 through 15. 

The shearing area of the soil specimens decreases during direct shear testing. The direct 

shear test stress paths, shown in figures 13 through 15 in τ-σ space, have been corrected 

for this effect. 

3.3 PILE SURFACES 

3.3.1 Introduction 

The interface testing program included determining interface friction angles between the 

two sands described above and the following seven pile surfaces: Lancaster FRP shell 

(curved), HardcoreTM FRP shell (curved), Hardcore FRP plate without bonded sand (flat), 

Hardcore FRP plate with bonded sand (flat), Plastic Piling recycled plastic (flat), steel 

sheet pile (flat), and prestressed concrete (flat). The following subsections describe these 

pile surfaces within a geotribology framework. 
35

3.3.2 Surface Topography Characterization

The importance of surface topography, and the degree of surface roughness in interface 

behavior and soil-structure interaction response, have been noted by many investigators 

(e.g., Yoshimi and Kishidi 1982, Kishida and Uesugi 1987, Lehane, et al., 1993, Dove, et 

al., 1997, Han 1997, Dove and Frost 1999, Dove and Jarrett 2002). 

The surface roughness of each pile material investigated was measured with a TaylorHobson Talysurf stylus profilometer. A schematic of the profilometer used in this study is 

shown in figure 16. As the figure illustrates, the profilometer comes equipped with a 

diamond stylus tip that traverses the surface to be characterized. Transducers in the 

profilometer help record the vertical and horizontal coordinates of the stylus tip. 

Therefore, the raw surface profile data consists of a series of discrete coordinate pairs 

(x, z) that can be postprocessed and analyzed to obtain the surface topography 

characterization of the material investigated (Dove and Jarrett 2002). 

Surface topography was quantified by three commonly used roughness parameters: the 

maximum peak-to-valley height (Rt), the average mean line spacing (Sm), and the 

arithmetic average roughness (Ra). Sm is twice the mean distance between locations at 

which the profiles cross the centerline drawn through the centroid of the profile. Ra is the 

arithmetic average value of the profile departure from the mean line along the profile 

length. These three roughness parameters are illustrated in figure 17. More detailed 

discussion of these parameters is given in International Organization for Standardization 

(ISO) Standard 4287 (ISO 1997). 

Figure 16. Illustration. Stylus profilometer sketch (Johnson 2000). 
36

Figure 17. Chart. Graphical representation of roughness parameters Rt, Sm, and Ra. 

Coupons approximately 25 mm (0.975 inch) by 200 mm ( 7.8 inches) were cut from each 

pile material. A series of six profiles, each 49 mm (1.91 inches) in length, were made on 

the surface of each coupon. All profiles were made in the longitudinal direction of the 

piles. Photos of the 7 pile surfaces and corresponding representative surface profiles are 

shown in figures 18 to 24. 

The roughness parameters for the seven interface materials are summarized in table 8. 

The parameter values reported and subsequently used are averages of the six individual 

values obtained for each coupon. 
37

Figure 18. Photo and Graph. Surface characteristics of 

Lancaster FRP composite pile. 

Figure 19. Photo and Graph. Surface characteristics of 

Hardcore FRP composite pile. 
38

Figure 20. Photo and Graph. Surface characteristics of Hardcore FRP plate. 

Figure 21. Photo and Graph. Surface characteristics of 

Hardcore surface-treated FRP plate. 
39

Figure 22. Photo and Graph. Surface characteristics of 

Plastic Piling plastic composite pile. 

Figure 23. Photo and Graph. Surface characteristics of prestressed concrete pile. 
40

Figure 24. Photo and Graph. Surface characteristics of steel sheet pile. 

Table 8. Summary of surface roughness measurements. 

Roughness Parameters 

Rt (μm) Sm (mm) Ra (μm) Interface Material 

Type Average S. D. Average S. D. Average S. D. 

LC—CP40 93.5 29.7 7.49 1.13 16.3 3.6 

HC—24–4 177.5 52.8 5.08 0.90 33.3 12.3 

HC untreated FRP plate 133.8 13.0 1.55 0.57 19.9 2.7 

HC treated FRP plate 341.2 83.7 0.49 0.03 31.4 2.4 

PPI plastic 365.9 73.2 1.86 0.21 43.1 9.1 

Concrete 335.3 90.9 1.94 0.59 28.0 5.0 

Steel 66.7 3.2 0.43 0.03 5.7 0.3 

Notes: LC = Lancaster Composite, Inc., HC = Hardcore Composites, PPI = Plastic Pilings, Inc., Rt = maximum 

peak-to-valley height, Sm = average mean line spacing, Ra = arithmetic average roughness, SD = 

standard deviation; 1 μm = 0.000039 inch; 1 mm = 0.039 mm. 

3.3.3 Interface Hardness 

The hardness of the interface material is another important factor for consideration when 

studying interface behavior. As the surface hardness of the construction material 

decreases, it becomes possible for the soil grains to penetrate the construction material 

and shear failure may occur if soil grains plow along the material surface (Pando, et al., 

2002a). In their study of sand-polymer interfaces, O’Rourke, et al. (1990), found that the 

interface frictional strength decreased with increasing polymer hardness. The authors 
41

found that relatively hard and smooth polymer surfaces did not show significant 

dilatancy, and the sand grains tended to slide along the polymer interface; whereas the 

particle movement on relatively soft and smooth polymer surfaces involved 

predominantly rolling and temporary grain indentations that resulted in higher interface 

friction values. The findings by O’Rourke, et al. (1990), are mainly applicable to smooth 

polymeric materials since, as the authors pointed out, surface roughness of rough 

polymers will promote dilation and result in increased shear resistance. 

The surface hardness of the different pile/interface materials was investigated by means 

of the Vickers hardness test. The Vickers hardness test has the flexibility to allow 

application of small loads, which is convenient for testing softer materials such as 

recycled plastic and FRP composites, and consists of pressing a standard diamond 

pyramid into the sample. The Vickers hardness number (HV) is related to the load 

applied and the area of the diamond pyramid indentation. Vickers hardness tests were 

performed in accordance with ASTM Standard E384-99 (ASTM 1999). The results of the 

hardness tests are summarized in table 9. 

 

Table 9. Surface hardness. 

Vickers Hardness (HV) Interface Material Mean SD N 

LC—CP40 24.0 1.83 20 

HC—24–4 17.7 2.14 20 

HC untreated FRP flat plate 35.7 10.38 20 

HC treated FRP flat plate Not measurable (1)

PPI plastic 8.5 5.05 20 

Concrete 63.7 36 36 

Steel 184.2 10.28 20 

Notes: (1): Not measurable because it is difficult to clearly measure the diamond indentation mark due 

to presence of the bonded sand particles. 

 HV = Vickers hardness number (ASTM Standard E384-99) (carried out at room temperature). 

 LC = Lancaster Composite, Inc., HC = Hardcore Composites. 

 PPI = Plastic Pilings, Inc. SD = standard deviation, n = number of tests performed. 

3.4 INTERFACE SHEAR TESTS 

Representative coupons were obtained from each of the seven different pile materials 

investigated. Careful consideration was given to selecting pile coupons that were 

representative of the average surface texture of the piles. The coupons from the plastic, 

concrete, and steel piles, and from the two Hardcore FRP plates, were all flat surfaced. 

Therefore, it was possible to use a flat shear box for interface shear testing of these 

materials. 

The coupons from the cylindrical FRP shells (i.e., Lancaster CP40 and Hardcore 24-4) 

were cut with a water-cooled saw to dimensions that fit into the bottom half of the shear 

box. However, because both composite FRP shells were fabricated with thermoset resins, 

their curvature could not be removed. To permit testing of these curved-pile specimens, 
42

the top half of the shear box was modified such that it could conform to the outside 

curvature of the piles, as shown in figure 25. The interface shear tests carried out on the 

other pile materials involved the use of the unmodified top half of the direct shear box. 

The interface shear tests were performed in accordance with ASTM Standard D5321-97 

(ASTM 1997), and the horizontal displacement rate was 0.9 mm/min (0.35 inch/min). 

Figure 25. Illustration. Sketch of modified interface shear test setup. 

The interface tests performed using the Density and Model sands are summarized in 

tables 10 and 11, respectively. Interface shear tests were carried out at constant normal 

stresses ranging from 23 to 200 kPa (3.3 to 29 lbf/inch2

) (the contact area of the soil 

sample remained constant throughout the test). Typical interface shear stress versus 

interface displacement curves for the seven pile material types are shown in figures 26 

and 27 for the Density and Model sands, respectively. Additional details of the interface 

tests performed for this investigation are presented in appendix A. 

Interface strength envelopes for the seven pile materials and for the Density and Model 

sands are shown in figures 28 through 34. Curvature of the strength envelopes with 

increasing normal stresses was generally not significant. The interface friction angles 

obtained from linear fits to the test data are summarized in table 12.
Table 10. Summary of interface shear test results on Density sand. 

Pile/Material Dr (%) σ'n (kPa) Sm (mm) Rt (μm) Hardness (HV) Sm/D50 Rt/D50 τp (kPa) dp (mm) 

57.5 38.2 7.49 93.5 24 14.97 0.19 14.5 0.45 64.3 63.2 7.49 93.5 24 14.97 0.19 22.8 0.35 62.4 113.2 7.49 93.5 24 14.97 0.19 40.0 0.42 

Lancaster Composite FRP shell (CP40) 65.7 200 7.49 93.5 24 14.97 0.19 70.2 0.60 

61.4 22.5 5.08 177.5 17.7 10.16 0.36 13.1 2.14 66.0 50 5.08 177.5 17.7 10.16 0.36 28.7 2.27 66.3 100 5.08 177.5 17.7 10.16 0.36 57 1.77 

Hardcore Composites FRP shell (24–4) 68.3 200 5.08 177.5 17.7 10.16 0.36 111.1 1.99 

64.6 26.15 1.55 133.8 35.7 3.10 0.27 15.02 0.96 61.5 50.06 1.55 133.8 35.7 3.10 0.27 29.81 1.17 63.4 107.3 1.55 133.8 35.7 3.10 0.27 57.93 1.61 

Hardcore FRP plate; untreated surface 66.2 199.73 1.55 133.8 35.7 3.10 0.27 107.55 1.59 

62.6 26.15 0.49 341.2 See note (1) 0.98 0.68 18.48 1.22 64.7 50.06 0.49 341.2 See note (1) 0.98 0.68 30.97 1.91 63.7 107.17 0.49 341.2 See note (1) 0.98 0.68 63.34 1.74 

Hardcore FRP plate; treated surface 64.9 183.5 0.49 341.2 See note (1) 0.98 0.68 116.03 2.26 

65.9 29.11 1.86 365.9 8.5 3.73 0.73 17.10 1.26 61.1 56.77 1.86 365.9 8.5 3.73 0.73 32.63 1.28 60.8 105.37 1.86 365.9 8.5 3.73 0.73 56.05 2.20 61.9 180.37 1.86 365.9 8.5 3.73 0.73 93.91 1.82 

PPI plastic 

61.1 199.72 1.86 365.9 8.5 3.73 0.73 103.39 1.66 

63.2 23.7 1.94 335.3 63.7 3.88 0.67 18.1 1.71 61.9 51.5 1.94 335.3 63.7 3.88 0.67 35.4 0.93 63.3 101.9 1.94 335.3 63.7 3.88 0.67 68.5 1.70 

Prestressed concrete pile 

64.0 175.4 1.94 335.3 63.7 3.88 0.67 111.5 2.55 

67.0 25.57 0.43 66.7 184.2 0.86 0.13 13.48 0.73 65.9 51.27 0.43 66.7 184.2 0.86 0.13 31.38 1.10 64.1 101.85 0.43 66.7 184.2 0.86 0.13 59.11 0.79 

Steel from sheet pile 

65.4 202.33 0.43 66.7 184.2 0.86 0.13 105.11 1.06 

Notes: (1) Not measurable because it is difficult to clearly measure the diamond indentation mark due to presence of the bonded sand particles. 

D50 = 0.5 mm for Density sand, Dr =Relative density after sample consolidation (as per equation 5), σ′n= Normal pressure, Sm = Average mean line sp

Rt = Maximum peak–to-valley height, Hardness (HV) = Average Vickers hardness number, 

τp= Peak interface shear stress, dp = Horizontal displacement at peak, 

τcv = Residual interface shear stress, dCV = Horizontal displacement at residual. 

1 mm = 0.039 inches; 1 μm = 0.000039 inch; 1 kPa = 0.145 poundforce per square inch. 

43 
Table 11. Summary of interface shear test results on Model sand. 

Pile/Material Dr (%) σ'n (kPa) Sm (mm) Rt (μm) Hardness (HV) Sm/D50 Rt/D50 τp (kPa) dp (mm) 

57.8 22.5 7.49 93.5 24 41.59 0.52 10.3 2.48 58.7 50 7.49 93.5 24 41.59 0.52 27.5 8.81 63.2 100 7.49 93.5 24 41.59 0.52 52.3 2.11 

Lancaster Composite FRP shell (CP40) 64.9 200 7.49 93.5 24 41.59 0.52 102.4 2.51 

56.1 22.5 5.08 177.5 17.7 28.22 0.99 14.5 0.31 64.4 50 5.08 177.5 17.7 28.22 0.99 28.6 12.3 60.9 100 5.08 177.5 17.7 28.22 0.99 56.5 9.26 

Hardcore Composite FRP shell (24–4) 67.9 200 5.08 177.5 17.7 28.22 0.99 113.2 12.36 

65.4 26.15 1.55 133.8 35.7 8.61 0.74 16.44 0.98 59.1 50.06 1.55 133.8 35.7 8.61 0.74 33.55 1.15 64.7 107.17 1.55 133.8 35.7 8.61 0.74 66.38 1.71 

Hardcore FRP plate; untreated surface 66.4 183.52 1.55 133.8 35.7 8.61 0.74 112.68 1.92 

63 26.15 0.49 341.2 See note (1) 2.71 1.90 23.26 0.91 65 50.08 0.49 341.2 See note (1) 2.71 1.90 39.00 1.61 

60.1 107.3 0.49 341.2 See note (1) 2.71 1.90 80.45 1.97 

Hardcore FRP plate; treated surface 62 152.2 0.49 341.2 See note (1) 2.71 1.90 115.75 2.26 

64.4 26.15 1.86 365.9 8.5 10.36 2.03 14.42 0.91 58.1 50.06 1.86 365.9 8.5 10.36 2.03 28.63 0.80 65.1 107.25 1.86 365.9 8.5 10.36 2.03 74.64 1.00 66.4 152.2 1.86 365.9 8.5 10.36 2.03 105.69 1.47 

PPI plastic 

62.4 199.7 1.86 365.9 8.5 10.36 2.03 126.73 1.64 

58.9 23.7 1.94 335.3 63.7 10.79 1.86 19.8 1.00 58.7 51.5 1.94 335.3 63.7 10.79 1.86 37.8 1.35 61.9 101.9 1.94 335.3 63.7 10.79 1.86 72.5 1.12 

Prestressed concrete pile 

68.2 175.4 1.94 335.3 63.7 10.79 1.86 116.7 1.42 

63.5 26.08 0.43 66.7 184.2 2.39 0.37 17.42 0.46 62.5 51.27 0.43 66.7 184.2 2.39 0.37 31.84 0.79 64.3 101.85 0.43 66.7 184.2 2.39 0.37 63.33 0.97 

Steel from sheet pile 

67.0 202.33 0.43 66.7 184.2 2.39 0.37 121.47 1.39 

Notes: (1) Not measurable because it is difficult to clearly measure the diamond indentation mark due to presence of the bonded sand particles. 

D50 = 0.18 mm for Model sand, Dr= Relative density after sample consolidation (as per equation 5), σ′n = Normal pressure, Sm = Average mean line sp

Rt = Maximum peak-to-valley height, Hardness (HV) = Average Vickers hardness number, τp= Peak interface shear stress, dp = Horizontal displaceme

τcv = Residual interface shear stress, dCV = Horizontal displacement at residual. 

1 mm = 0.039 inches; 1 μm = 0.000039 inch; 1 kPa = 0.145 poundforce per square inch. 

44 
45 

Figure 26. Graphs. Typical interface shear test results for 

Density sand (σ'n ≈ 100 kPa). 
46 

Figure 27. Graphs. Typical interface shear test results for 

Model sand (σ'n ≈ 100 kPa). 
47 

Figure 28. Graph. Interface shear strength envelopes for 

Lancaster Composite FRP shell. 

Figure 29. Graph. Interface shear strength envelopes for 

Hardcore Composites FRP shell. 
48 

Figure 30. Graph. Interface shear strength envelopes for 

untreated Hardcore FRP plate. 

Figure 31. Graph. Interface shear strength envelopes for 

treated Hardcore FRP plate. 
49 

Figure 32. Graph. Interface shear strength envelopes for PPI plastic. 

Figure 33. Graph. Interface shear strength envelopes for concrete. 
50 

Figure 34. Graph. Interface shear strength envelopes for steel. 

Table 12. Summary of interface friction angles. 

Soil properties Relative 

Roughness 

Interface 

Angles Soil Pile Average

Dr (%) 

D50

(mm)

Surface 

Hardness 

(HV) Rt/D50 Sm/D50

δp

(

o

) 

δcv

(

o

) 

LC FRP 62.5 24.0 0.19 14.97 19.7 16.6 

HC FRP 65.5 17.7 0.36 10.16 29.2 27.3 

Untreated FRP 

plate 63.9 35.7 0.27 3.10 28.4 25.7 

Treated FRP plate 64.0 N/T 0.68 0.98 31.9 27.8 

PPI plastic 62.2 8.5 0.73 3.73 27.6 24.9 

Concrete 63.1 63.7 0.67 3.88 33.0 27.7 

Density Sand 

Steel 65.6 

0.50 

184.2 0.13 0.86 28.2 25.1 

LC FRP 61.2 24.0 0.52 41.59 27.3 26.0 

HC FRP 62.3 17.7 0.99 28.22 29.5 29.3 

Untreated FRP 

plate 63.9 35.7 0.74 8.61 31.7 28.0 

Treated FRP plate 62.5 N/T 1.90 2.71 37.3 32.6 

PPI plastic 63.3 8.5 2.03 10.36 33.4 28.8 

Concrete 61.9 63.7 1.86 10.79 34.3 28.0 

Model Sand 

Steel 64.3 

0.18 

184.2 0.37 2.39 31.2 28.6 

Note: N/T = not testable. 1 mm = 0.039 inch 
51 

3.5 DISCUSSION OF RESULTS 

It has been demonstrated previously that the strength and volume change behaviors of an 

interface system composed of a granular soil and a construction material are largely 

controlled by three factors: the relative size of the sand grains with respect to the surface 

asperity height and spacing, the hardness of the materials, and the soil grain shape 

(Kishida and Uesugi 1987, O'Rourke, et al., 1990, Irsyam and Hryciw 1991, Hryciw and 

Irsyam 1993, Dove and Harpring 1999, Dove and Frost 1999, Frost and Han 1999, and 

Dove and Jarrett 2002). The interface shear test program conducted in this study was not 

designed to systematically investigate the different factors that influence the interface 

shear behavior of sands and FRP materials. Nevertheless, the interface friction angle 

values obtained in this study (shown in table 12), were obtained from interface shear tests 

having important differences in the following basic variables: 

1. Mean grain size of sand, D50. 

2. Angularity of sand. 

3. Maximum peak-to-valley height of pile surface, Rt. 

4. Average mean line spacing of pile surface, Sm. 

5. Pile surface Vickers hardness, HV. 

The first two variables are related to soil characteristics, and the remaining three are 

related to pile surface characteristics. All of these variables, except angularity, were 

measured quantitatively. Therefore, the influence of the quantitative variables on the 

interface friction coefficients (peak and residual tan δ) can be investigated by performing 

multiple linear regressions. Linear regression analyses for both sand types and for peak 

and residual interface friction coefficients are presented in the following subsections. The 

regressions were carried out with the interface friction coefficient as the dependent 

variable, and Rt/D50, Sm/D50, and HV as independent variables. Due to small sample size 

(N = 7), statistically significant results (at the 5 percent level) were not expected. 

The linear regression analysis results for both sand types and for peak and residual 

interface friction coefficients are presented in sections 3.5.1 through 3.5.4. The general 

discussion of the regression results is provided in section 3.5.5. 
52 

3.5.1 Multiple Linear Regression for Density Sand tan δpeak Values 

Results of the regression analysis for the tan δpeak values from tests using the Density 

sand are summarized in table 13. 

Table 13. Multiple linear regression results for tan δpeak of Density sand. 

Regression Equation ( ) t 3 4 m peak

50 50

R S tan 0.504 0.174 8.93 10 1.919 10 HV

D D

− − ⎛⎞ ⎛⎞ δ= + ⋅ − ⋅ ⋅ + ⋅ ⋅ ⎜⎟ ⎜⎟ ⎝⎠ ⎝⎠

R2 (N=7) 0.662 

R2 change due to Rt/D50 0.355 

R2 change due to Sm/D50 0.302 

R2 change due to HV 0.005 

The coefficient of determination (R2) for this fit was 0.66, shown graphically in figure 35. 

The contributions of Rt/D50, Sm/D50, and HV to R2

 were 53.6 percent, 45.6 percent, and 

0.8 percent, respectively. This indicates that hardness (HV) is the least significant 

predictor of the tan δpeak values compared to Rt/D50 and Sm/D50. 

Figure 35. Graph. Multiple linear regression on Density sand tan δpeak values. 
53 

3.5.2 Multiple Linear Regression for Density Sand tan δcv Values

Results of the regression analysis for the tan δcv values from tests using the Density sand 

are summarized in table 14. 

Table 14. Multiple linear regression results for tan δcv of Density sand. 

Regression Equation ( ) 224 t m cv

50 50

R S tan 0.502 6.839 10 1.03 10 1.48 10 HV

D D

−−− ⎛⎞ ⎛⎞ δ= + ⋅ ⋅ − ⋅ ⋅ − ⋅ ⋅ ⎜⎟ ⎜⎟ ⎝⎠ ⎝⎠

R2 (N=7) 0.581 

R2 change due to Rt/D50 0.264 

R2 change due to Sm/D50 0.313 

R2 change due to HV 0.004 

The coefficient of determination (R2) for this fit was 0.58, and is shown graphically in 

figure 36. The contributions of Rt/D50, Sm/D50, and HV to R2

 were 45.4 percent, 53.9 

percent, and 0.7 percent, respectively. As in the previous analyses, this indicates that the 

hardness (HV) has considerably less influence on the tan δcv values than Rt/D50 and 

Sm/D50. 

Figure 36. Graph. Multiple linear regression on Density sand tan δcv values. 
54 

3.5.3 Multiple Linear Regression for Model Sand tan δpeak Values

Results of the regression analysis for the tan δpeak values from tests using the Model sand 

are summarized in table 15. 

Table 15. Multiple linear regression results for tan δpeak of Model sand. 

Regression equation: ( ) 235 t m peak

50 50

R S tan 0.608 6.160 10 3.31 10 5.81 10 HV

D D

−−− ⎛⎞ ⎛⎞ δ= + ⋅ ⋅ − ⋅ ⋅ − ⋅ ⋅ ⎜⎟ ⎜⎟ ⎝⎠ ⎝⎠

R2 (N=7) 0.892 

R2 change due to Rt/D50 0.612 

R2 change due to Sm/D50 0.279 

R2 change due to HV 0.001 

The coefficient of determination (R2) for this fit was 0.89, and is shown graphically in 

figure 37. The contributions of Rt/D50, Sm/D50, and HV to R2

 were 68.6 percent, 31.3 

percent, and 0.1 percent, respectively. As in the previous analysis, this indicates that 

hardness (HV) has considerably less influence on the tan δpeak values than Rt/D50 and 

Sm/D50. 

Figure 37. Graph. Multiple linear regression on Model sand tan δpeak values. 
55 

3.5.4 Multiple Linear Regression for Model Sand tan δcv Values

Results of the regression analysis for the tan δcv values from tests using the Model sand 

are summarized in table 16. 

 

Table 16. Multiple linear regression results for tan δcv of Model sand. 

Regression equation ( ) 234 t m cv

50 50

R S tan 0.574 1.236 10 1.97 10 1.79 10 HV

D D

−−− ⎛⎞ ⎛⎞ δ= + ⋅ ⋅ − ⋅ ⋅ + ⋅ ⋅ ⎜⎟ ⎜⎟ ⎝⎠ ⎝⎠

R2 (N=7) 0.474 

R2 change due to Rt/D50 0.275 

R2 change due to Sm/D50 0.179 

R2 change due to HV 0.020 

The coefficient of determination (R2) for this fit was 0.47, and is shown graphically in 

figure 38. The contributions of Rt/D50, Sm/D50, and HV to R2

 were 58 percent, 37.8 

percent, and 4.2 percent, respectively. As in the previous analysis, this indicates that 

hardness (HV) has considerably less influence on the tan δcv values than Rt/D50 and 

Sm/D50. 

 

Figure 38. Graph. Multiple linear regression on Model sand tan δcv values. 
56 

3.5.5 Observations from the Linear Regression Analyses Results 

Based on the four linear regression analyses presented above, the following observations 

and comments are made. 

1. The limited size of the samples used for the regression analyses diminishes the 

statistical significance of the results of the regression analyses. 

2. All four regression equations showed a positive coefficient for the Rt/D50 term. This 

indicates a general trend that the interface friction coefficients (peak and residual) 

increase with increasing relative asperity height, Rt/D50. This finding concurs with the 

observations noted by Uesugi (1987) and Frost and Han (1999). 

3. All four regression equations showed a negative coefficient for the relative spacing 

term (Sm/D50), indicating that the interface friction coefficients (peak and residual) 

decrease with increasing relative spacing, Sm/D50, which is reasonable over the range 

of Sm/D50 values tested. 

4. The four regression equations showed both positive and negative coefficients for the 

Vickers hardness (HV) term. This finding, and the low contribution of this variable to 

the overall R2

 coefficient of the fit, indicate that hardness (HV) is not a significant 

predictor of the interface friction coefficients values. 

5. The regressions show only moderate fit strengths, with coefficients of determination 

(R2) ranging from 0.47 to 0.89. This indicates that other factors besides Rt/D50, 

Sm/D50, and HV also have important influences on the values of the interface friction 

coefficient. 

6. Although soil angularity could not be included in the regression analyses, it can be 

seen that inclusion of an angularity variable alone will not be enough to improve the 

strength of the fits. This can be seen in figure 34, where the goodness of the fit is poor 

despite the fact that all points correspond to the same sand with the same angularity. 

Adding an angularity term to the regression equation would not increase the strength 

of this fit. This indicates that factors other than soil angularity influence the values of 

the interface friction coefficient. 

3.5.6 Influence of Angularity of Sand 

The regression analyses presented in the preceding sections did not include the influence 

of soil angularity. As explained earlier, this was because the angularity of the sands used 

in this study was only described qualitatively. The sand angularity for the Density sand 

was described as subrounded to rounded, and for the Model sand as subangular to 

angular. In general, angular sands have values of higher interface friction angle than 

rounded sands (Han and Frost 1999). This is consistent with results for the Model and 

Density sands, as listed in table 12. Due to the higher contact stresses at sharp grain 

edges, angular soil grains are able to penetrate into the surfaces of some of the pile 
57 

materials. For this reason, and because angular particles may interlock better with the pile 

surface roughness, interface friction angles tend to be larger for angular soil grains than 

for rounded grains. 

3.6 SUMMARY 

The following laboratory activities were performed for this investigation: 

1. Selection of sand specimens for interface testing. 

2. Soil characterization tests to determine grain size distribution, minimum/maximum 

density, and specific gravity testing. 

3. Direct shear tests on Density and Model sands to determine internal friction angles. 

4. Selection of pile surface specimens for interface testing. 

5. Surface topography characterization and surface hardness determination of seven pile 

surfaces. The seven pile surfaces included four FRP composites (commercially 

available from two FRP composite pile manufacturers), one recycled plastic pile, and 

two conventional pile materials (a prestressed concrete pile and a steel sheet pile). 

6. Design and construction of a modified top half shear box to permit testing of curved 

pile surfaces obtained from the FRP composite tubes. 

7. Interface shear tests for the seven piles and two sand types to determine the interface 

behavior and interface friction angles of these interfaces. 

Interface shear tests were performed on 14 types of soil-pile interfaces. Tests were carried 

out for two types of sands: Density sand (D50 = 0.5 mm (0.019 inch), subrounded to 

rounded particle shape) and Model sand (D50 = 0.18 mm (0.007 inch), subangular to 

angular particle shape). Seven pile surfaces were tested: Lancaster FRP composite 

(curved), Hardcore FRP composite (curved), Hardcore FRP composite plate (flat), 

Hardcore FRP composite plate with bonded sand treatment (flat), PPI recycled plastic 

coupon (flat), prestressed concrete pile coupon (flat), and a steel sheet pile coupon (flat). 

The results of these tests are summarized in appendix A. 

The peak and residual interface friction angles for the 14 sand-to-pile interfaces tested are 

summarized in tables 10 through 12. The following points and observations can be made 

regarding the experimental data. 

1. The interface friction angle values obtained for the Density sand tested against the 

Lancaster FRP composite pile were much lower than the values obtained for the other 

pile surfaces. The peak and residual interface friction angles obtained for the 

Lancaster FRP composite pile were 19.7o

 and 16.6o, respectively. In contrast, the 

interface friction angle values obtained for the Density sand tested against the other 
58 

pile surface types ranged from 27.6o

 to 33o, and from 24.9o to 27.8o, for the peak and 

residual conditions, respectively. 

2. The Lancaster FRP composite pile was measured to have the largest average mean 

line spacing (Sm) and the second lowest maximum peak to valley height (Rt) as 

reported in table 8. These values make this pile surface the smoothest of all seven 

surfaces tested. 

3. In general, the subangular to angular Model sand gives slightly higher interface 

friction angles than the subrounded to rounded Density sand. However, for the 

Lancaster FRP composite pile, the interface friction angles obtained with the Model 

sand are much higher than the Density sand values. 

 

4. In general, the interface friction angles, both peak and residual, were found to 

increase with increasing relative asperity height, Rt/D50. 

5. In general, the interface friction angles, both peak and residual, were found to 

increase with decreasing relative spacing, Sm/D50, which is reasonable over the range 

of Sm/D50 tested. 

6. Linear regression analyses between the interface friction coefficients and the 

variables Rt/D50, Sm/D50, and HV showed moderate fit strengths. The regression 

analyses suggest other factors besides Rt/D50, Sm/D50, and HV also have important 

influences on the values of interface friction coefficeint. 

7. The bonded sand surface treatment used for the Hardcore FRP plate was successful in 

increasing the interface friction angle values.
59 

CHAPTER 4. EXPERIMENTAL DURABILITY STUDY OF FRP 

COMPOSITE PILES 

4.1 INTRODUCTION 

The lack of long-term performance and durability data of composite piles is an important 

concern that must be addressed to permit confidence in the use of these piles for longterm load-bearing applications. Consequently, a durability study was undertaken to assess 

the rate degradation of mechanical properties (strength and stiffness) of commercially 

available glass FRP (GFRP) tubes used to fabricate concrete-filled FRP composite piles. 

The study focused on the degradation effects of moisture absorption and exposure to 

freeze-thaw cycles. 

The durability study was carried out on four FRP tubes supplied by two manufacturers of 

FRP composite piles. The laboratory degradation study included FRP shell 

characterization, determination of baseline mechanical properties, measurement of 

moisture absorption as a function of time, measurement of mechanical properties as a 

function of moisture absorption, and mechanical properties as a function of exposure to 

freeze-thaw cycles. The methodology and results of this durability study are presented in 

this chapter. The durability study did not include the steel-reinforced plastic pile because 

this pile type was initially not included in the scope of this project, and was only included 

for field testing after the late withdrawal of one of the FRP composite pile manufacturers. 

4.2 BACKGROUND ON DEGRADATION OF GLASS FRP COMPOSITES 

Some of the factors that contribute to the degradation of the GFRP composites over time 

include temperature, sunshine (UV rays), moisture absorption, freeze-thaw cycles, and 

load (Garcia, et al., 1998, Nishizaki and Meiarashi 2002). Among these factors, the 

influence of moisture absorption and exposure to freeze-thaw cycles are considered to be 

the most critical factors. Exposure to alkaline environments and UV radiation also affect 

long-term durability, but to a lesser extent. Furthermore, UV degradation resistance of 

most composites is being improved by applying protective coats and additives during the 

manufacturing process. The degradation mechanisms related to moisture absorption and 

freeze-thaw cycles are discussed below. 

Degradation due to moisture absorption may significantly reduce the life of FRP 

composites. The influence that moisture absorption has on the mechanical properties of 

GFRP composites is well documented (e.g., Shen and Springer 1976, Garcia, et al., 1998, 

Prian and Barkatt 1999, Verghese, et al., 1999, Shao and Kouadio 2002, Phifer 2003). 

Absorbed moisture can cause pronounced changes in modulus, strength, and strain to 

failure (Springer, et al., 1980). Moisture content of submerged FRP composites increases 

through diffusion. The absorbed moisture can act as a plasticizer of the composite resin, 

and can cause matrix cracking, fiber-matrix debonding, and corrosion of glass fibers 

(stress corrosion) (Garcia, et al., 1998). These effects result in a reduction of strength and 

stiffness of the FRP composite. For example, Phifer (2003) recorded tensile strength and 
60 

stifofness reductions on the order of 60 percent and 10 percent, respectively, for Eglass/vinyl ester composites submerged in fresh water for a period of about 2 years. The 

author projected FRP tensile strength loss of about 80 percent after 50 years’ 

submergence. The implications of such strength reductions on the design of composite 

piles can be significant. 

In addition to submergence time, temperature and stress level also influence the amount 

of moisture that FRP will absorb while submerged. The moisture absorption increases 

with increasing water temperature and stress level. Evidence of the existence of damage 

to FRP composites after submergence in water has been found in SEM images (e.g., 

figure 9), where fiber damage and crack formation at the fiber-matrix interface were 

observed (McBagonluri, et al., 2000). 

Freeze-thaw cycles can also degrade FRP composite properties. The degradation due to 

freeze-thaw cycling is primarily related to microcracking caused by the volume increase 

of the absorbed water in the composite (Verghese, et al., 1999). The mechanical 

properties of the composite are also affected by cold temperatures and by the cycling 

process (Karbhari 2002). 

The FRP shell will potentially be exposed to chemical agents as a result of the alkaline 

pore water from the concrete inside the FRP shell. Karbahari, et al. (2002), studied the 

effect of concrete-based alkali solutions on the durability of vinylester GFRP composites. 

The authors found evidence of significant degradation and strength loss due not only to 

moisture absorption but also to the exposure to the high pH levels and characteristic 

solution chemistry of concrete pore fluid. The experimental durability program carried 

out for this study did not include investigation of possible degradation effects due to 

exposure to concrete pore fluid solutions. 

A concrete-filled FRP composite pile installed in tidal and marine environments would be 

exposed to salt water. However, McBagonluri, et al. (2000), found that the rate of 

strength loss was not significantly affected by salt content, but rather by absorption of 

moisture. Therefore, the durability experiments carried out for this study were performed 

using fresh water. 

 

4.3 LABORATORY STUDY OF DURABILITY 

The general approach used in this study was to track moisture diffusion into the material 

and to evaluate the degradation of the mechanical properties of the different FRP shells as 

a function of submergence time and moisture absorption. The loss of material strength 

and stiffness is then related to the long-term structural capacity of the pile using 

conventional structural models for FRP composite piles available in the literature (e.g. 

Fam and Rizkalla 2000, Mirmiran 1999), as summarized in chapter 2. The study 

described herein does not include effects of creep or chemical attack as a result of the 

presence of the concrete infill in the composite pile. Changes in the concrete properties 

with time (e.g., possible increases or decreases in strength and/or stiffness) were not 

included in this study. 
61 

4.3.1 Description of Test Specimens 

Commercially available FRP tubes from two manufacturers that specialize in concretefilled FRP composite piles were used for the durability experimental program. Tubes 

with 0.305-m (12-inch) and 0.610-m (24-inch) nominal diameter were tested. The results 

presented here are based on tests carried out on their standard FRP tube designs. It should 

be noted, however, that both pile manufacturers have the flexibility to modify the FRP 

composite shell laminate architecture (including number of layers, fiber orientation, and 

resin and fiber types) to better suit the load demands of a specific project. 

The following subsections provide detailed descriptions of the tubes from each 

manufacturer, including dimensions, material, number of layers, laminate structure, fiber 

volume fraction, and manufacturing technique. 

4.3.1.1 Specimens from Lancaster Composites, Inc. 

The two sizes of tubes supplied by Lancaster Composites, Inc. were fabricated by 

Ameron International Corporation. The FRP tubes were manufactured using the filamentwinding technique, in which E-Glass continuous fiber rovings are impregnated with resin 

(epoxy or polyester) and wound over a rotating steel mandrel, following a predetermined 

winding pattern. 

Characterization of the FRP tubes was based on visual inspection, information supplied 

by the manufacturers, and burnoff tests on coupon samples. The burnoff tests were 

performed in general accordance with ASTM Standard D2584. The burnoff test results 

and other characterization information are provided in tables 17 and 18 for the 12- and 

24-inch (0.31- and 0.61-m) diameter tubes, respectively. 
62 

Table 17. Characterization data for Lancaster Composite 12-inch FRP tube. 

Manufacturer pile designation CP40 12-inch 

Nominal diameter 305 mm (12 in) 

Measured outside diameter 324.7 ± 1.9 mm (1)

Total wall thickness 6.05 ± 0.14 mm (1)

Liner thickness 0.59 ± 0.05 mm (1)

Liner material Epoxy resin sandwiched between industrial paper 

Unit weight of tube 17.45 kN/m3

Fiber type Owen Corning Type 30 

113 Yield E-glass 

Matrix resin type Epoxy anhydrite hardener/Dow Der 331 (83/100 ratio) 

Fiber layup angles [-88o/+8 o /-88 o /+8 o /-88/+8 o /-88 o/+8 o /-88]T

(2)

Fiber layer average thicknesses [0.54/0.69/0.54/0.69/0.54/0.69/0.54/0.69/0.54] (mm) 

Total resin fraction by weight 38.6% 

Total fiber fraction by volume 45.8 ± 1.0% (1), (3)

Fiber volume fraction 88o layers 38.4 ± 1.0% (1)

Fiber volume fraction 8o layers 55.8 ± 0.3% (1)

Percentage of fiber at 88o 40.1 ± 0.6% (1) (by weight) 

Percentage of fiber at 8o 59.9 ± 0.6% (1) (by weight) 

Notes: (1) Mean ± standard deviation. 1 mm = 0.039 inch 

(2) Angles are measured with respect to pile longitudinal axis. Subscript T indicates orientation for 

the layers of the total laminate are presented within the brackets. 

 (3) Based on volume of FRP laminate only; liner is excluded from volume. 

Table 18. Characterization data for Lancaster Composite 24-inch FRP tube. 

Manufacturer pile designation CP40 24-inch 

Nominal diameter 610 mm (24 in) 

Measured outside diameter 625 ± 1.8 mm (1)

Total wall thickness 7.35 ± 0.23 mm (1)

Liner thickness 1.68 ± 0.09 mm (1)

Liner material C-veil, a chopped fiber mat and resin 

Unit weight of tube 17.36 kN/m3

Fiber type Owen Corning Type 30 

113 Yield E-glass 

Matrix resin type Ashland Chemical Apropol 7241 

isophtalic acid polyester resin 

Fiber layup angles [+35o/-35 o /+85 o /+35 o /-35 o]T

(2)

Fiber layer average thicknesses [1.13/1.13/1.13/1.13/1.13] (mm) 

Total resin fraction by weight 39.3% 

Total fiber fraction by volume 51.2 ± 2.4% (1), (3)

Fiber volume fraction ± 35o layers 52.7 ± 2.9% (1)

Fiber volume fraction 85o layers 45.5 ± 1.0% (1)

Notes: (1) Mean ± standard deviation. 1 mm = 0.039 inch 

(2) Angles are measured with respect to pile longitudinal axis. Subscript T indicates orientation for 

the layers of the total laminate are presented within the brackets. 

 (3) Based on volume of FRP laminate only; liner is excluded from volume. 
63 

The 0.305-m (12-inch) nominal diameter tube has an outer diameter of 325 mm (12.8 

inches) and an average total wall thickness of 6.05 mm (0.238 inches). The tube has a 

0.59-mm (0.023-inch) thick liner at the inner surface, which results in a net structural 

wall thickness of 5.46 mm (0.215 inches). The wall structure of the tube consists of nine 

layers of fibers oriented at -88 degrees and +8 degrees with respect to the longitudinal 

axis of the tube. The stacking sequence of the layers is [-88o

/+8o /-88o /+8o /-88°/+8o / 

-88o

/+8o /-88o]T (subscript T indicates the stacking sequence for the total FRP laminate is 

presented within the brackets). The layers with fibers oriented at -88 degrees with respect 

to the longitudinal axis are 0.54-mm (0.021-inch) thick, and the layers with fibers 

oriented at +8 degrees with the respect to the longitudinal axis are 0.69 mm (0.027 

inches). The FRP tube liner is composed of overlapped industrial paper impregnated with 

epoxy resin, and is about 0.59 mm (0.023 inches) thick. The average overall fiber volume 

fraction of this tube is 45.8 percent. The matrix resin for this FRP tube consisted of Dow 

D.E.R. 331 epoxy mixed with anhydrite hardener. The fiber type was Owens Corning 

Type 30 113 Yield E-Glass rovings. 

The 0.610-m (24-inch) nominal diameter tube has an outer diameter of 625 mm (24.6 

inches) and an average total wall thickness of 7.35 mm (0.289 inches). The tube has a 

1.68-mm (0.066-inch) thick liner at the inner surface, which results in a net structural 

wall thickness of 5.67 mm (0.223 inches). The wall structure of the tube consists of five 

layers of fibers oriented at [+35o/-35o/+85o/+35o/-35 o]T, measured with respect to the 

longitudinal axis of the tube. The fiber layers are about 1.13 mm (0.044 inches) thick. 

The FRP tube liner is composed of C-veil and a mat of chopped fiber impregnated with 

resin, and is about 1.68 mm (0.066 inches) thick. The average overall fiber volume 

fraction of this tube is 51.2 percent. The matrix resin for the 0.610-m (24-inch) FRP shell 

was isophthalic acid polyester resin (Ashland Chemical Apropol 7241). The fiber type 

was Owens Corning Type 30 113 Yield E-Glass rovings. 

4.3.1.2 Specimens from Hardcore Composites 

Two FRP tube sizes were supplied by Hardcore Composites. The tubes had 12-inch and 

24-inch (0.31-m and 0.61-m) nominal diameters and were fabricated at its manufacturing 

plant in New Castle, DE. Both FRP tubes sizes were fabricated using a vacuum assisted 

resin transfer molding (VARTM) process. The matrix resin used in the tubes is Dow 

vinyl ester Derakane resin, and the fiber reinforcing consists of plies of stitch-bonded 

multiaxis glass fiber fabric. Each ply includes four layers of fibers oriented at 0, +45, -45, 

and 90 degrees with respect to the longitudinal axis of the tube. 

 

Characterization of the FRP tubes was based on visual inspection, information supplied 

from the manufacturer, and burnoff tests performed on coupon samples. The burnoff tests 

were carried out in general accordance with ASTM Standard D2584. Characterization 

information, including burnoff test results, are provided in tables 19 and 20 for the 12- 

and 24-inch (0.31- and 0.61-m) diameter tubes, respectively. 

 
64 

Table 19. Characterization data for Hardcore Composites 12-inch FRP tube. 

Manufacturer pile designation 12-3 

Nominal diameter 305 mm (12 in) 

Measured outside diameter 321.7 ± 3.8 mm (1)

Total wall thickness 5.66 ± 0.28 mm (1)

Liner thickness 0.68 ± 0.11 mm (1)

Liner material Screen 

Unit weight of tube 18.6 kN/m3

Fiber type E-glass multiaxis fabric type Quad Q-9100. 

4 stitched layers oriented at [0o

/+45o/-45o/90o]

 (2)

Matrix resin type Dow vinyl ester Derakane resin 

Number of multiaxis plies 3 x [0o

/+45o/-45o/90o]

 (2)

Multi-axis fiber layer thicknesses 1.66 ± 0.11 mm (1)

Total resin fraction by weight 28.2% 

Total fiber fraction by volume 52.8 ± 0.1% (1), (3)

Fiber volume fraction 0o layers 60.6 ± 2.6% (1)

Fiber volume fraction ±45o layers 41.7 ± 6.7% (1)

Fiber volume fraction 90o layers 34.2 ± 8.3% (1)

Percentage of fiber at 0o 56.1 ± 1.4% (1) (by weight) 

Percentage of fiber at ±45o 37.1 ± 1.1 % (1) (by weight) 

Percentage of fiber at 90o 6.8 ± 0.3 % (1) (by weight) 

Notes: (1) Mean ± standard deviation. (2) Angles are measured with respect to pile longitudinal axis. 

 (3) Based on volume of FRP laminate only; liner is excluded from volume. 1 mm = 0.039 inch 

Table 20. Characterization data for Hardcore Composites 24-inch FRP tube. 

Manufacturer pile designation 24-4 

Nominal diameter 610 mm (24 in) outside 

Measured outside diameter 622 ± 1.6 mm (1)

Total wall thickness 8.86 ± 0.68 mm (1)

Liner thickness 0.68 ± 0.15 mm (1)

Liner material Screen 

Unit weight of tube 18.67 kN/m3

Fiber type E-glass multiaxis fabric type Quad Q-9100. 

4 stitched layers oriented at [0o

/+45o/-45o/90o]

 (2)

Matrix resin type Dow vinyl ester Derakane resin 

Number of multi-axis plies 4 x [0o

/+45o/-45o/90o]

 (2)

Multiaxis fiber layer thicknesses 2.05 ± 0.14 mm (1) (mm) 

Total resin fraction by weight 28.6% 

Total fiber fraction by volume 52.1 ± 0.5% (1), (3)

Fiber volume fraction 0o layers 61.2 ± 1.4% (1)

Fiber volume fraction ±45o layers 45.2 ± 0.6% (1)

Fiber volume fraction 90o layers 40.7 ± 1.1% (1)

Percentage of fiber at 0o 53.9 ± 0.7% (1) (by weight) 

Percentage of fiber at ±45o 38.9 ± 0.8% (1) (by weight) 

Percentage of fiber at 90o 7.3 ± 0.1% (1) (by weight) 

Notes: (1) Mean ± standard deviation. (2) Angles are measured with respect to pile longitudinal axis. 

 (3) Based on volume of FRP laminate only; liner is excluded from volume. 1 mm = 0.039 inch 
65 

The 305-mm (12-inch) nominal diameter tube has an outer diameter of 322 mm (12.7 

inches) and a total wall thickness of 5.66 mm (0.223 inches). The tube has a 0.68-mm 

(0.027-inch) thick liner at the inner surface, which results in a net structural wall 

thickness of 4.98 mm (0.196 inches). The wall structure of the tube consists of three 

layers of multiaxis fabric. Each fabric layer consists of four layers of fibers oriented at 

[0o

/90o /+45o/-45o]. The symbol typically used to designate this fiber lay-up is [0o/90o

/+45o

/-45o]3T (where the subscript 3T indicates the FRP laminate is composed of three 

plies each with four layers oriented at 0o/90o /+45o/-45o). The average thickness of the 

multiaxis fabric layers is 1.66 mm (0.065 inches). The average thicknesses of the layers 

with fibers oriented at ± 45, 0, and 90 degrees with respect to the longitudinal axis of the 

tube are 0.74 mm (0.029 inches), 0.76 mm (0.030 inches), and 0.16 mm (0.006 inches), 

respectively. The FRP tube liner is composed of a plastic screen material used to 

facilitate the injection of resin during the manufacturing process. The average overall 

fiber volume fraction of these tubes was measured to be 52.8 percent. 

The 610-mm (24-inch) nominal diameter tube has an outer diameter of 622 mm (24.5 

inches) and a total wall thickness of 8.86 mm (0.349 inches). The tube has a 0.68-mm 

(0.027-inch) thick liner at the inner surface, which results in a net structural wall 

thickness of 8.18 mm (0.322 inches). The wall structure of the tube consists of four layers 

of multiaxis fabric. Each fabric layer consists of four layers of fibers oriented at [0o/90o

/+45o

/-45o]. The symbol typically used to designate this type of fiber layup is [0o/90o

/+45o/-45o]4T (where the subscript 4T indicates the FRP laminate is composed of four 

plies each with four layers oriented at 0o

/90o /+45o/-45o). The average thickness of the 

multiaxis fabric layers is 2.05 mm (0.081 inches). The average thicknesses of the layers 

with fibers oriented at ±45, 0, and 90 degrees with respect to the longitudinal axis of the 

tube are 0.92 mm (0.036 inches), 0.94 mm (0.037 inches), and 0.19 mm (0.007 inches), 

respectively. The average overall fiber volume fraction of these tubes was measured to be 

52.1 percent. 

 

4.3.2 Test Equipment and Procedures 

The test procedures and apparatus used in this durability study are described in this 

subsection. 

4.3.2.1 Burnoff tests 

As indicated earlier, coupons from the different FRP tubes investigated were subjected to 

burnoff tests to determine information such as fiber volume content, resin content, the 

number of fiber layers, and orientations. The tests were carried out in general accordance 

with ASTM Standard D2584. The tests consist of heating coupon samples from the FRP 

tubes in a furnace oven to a temperature of about 550 o

C (1022 °F). The heat from the 

oven is used to burn the resin matrix, leaving the fiber layers exposed. The oven used for 

burnoff testing is shown in figure 39a. An FRP coupon sample after resin burnoff is 

shown in figure 39b. The number of fiber layers and orientation can be determined from 

this test, as shown in figure 39b. 
66 

Figure 39. Photos. Burnoff testing. 

4.3.2.2 Axial tension tests 

Axial tensile properties were evaluated using tensile tests performed on FRP samples in 

accordance with ASTM Test Method for Tensile Properties of Polymer Matrix Composite 

Materials (D3039). Test specimen strips were cut in the longitudinal direction using a 

water-cooled saw. The average test specimen dimension was 25 mm by 200 mm (0.98 inch 

by 7.8 inch). The tests were carried out using an INSTRON® test frame operated at a 

constant rate of displacement of 1.27 mm/min (0.05 inches/min). Axial strain was 

measured primarily with the use of extensometers; however, selected samples were also 

instrumented with strain gages. A typical tension test setup is shown in figure 40. 
67 

Figure 40. Photo. Typical tension test setup. 

4.3.2.3 Hoop tension tests 

Tensile properties in the hoop direction were evaluated using the split disk test method. 

The split disk fixtures were custom made to fit the two FRP tube diameter sizes 

investigated in this study. The split disk tests were performed in accordance with ASTM 

Test Method for Apparent Hoop Tensile Strength of Plastic or Reinforced Plastic Pipe by 

Split Disk Method (D2290). Split disk specimens were precision cut from the FRP shells 

with a nominal width of 40 mm (1.6 inches). The tests were carried out using a 534 kN 

Tinius-Olsen machine and an INSTRON test frame at a constant rate of displacement of 

2.5 mm/min (0.1 inches/min). The typical split disk test setup is shown in figure 41. 
68 

Figure 41. Photos. Typical split disk test setup. 

4.3.2.4 Freeze-thaw testing 

Saturated FRP specimens were subjected to freeze-thaw cycles using a programmable 

freeze-thaw chamber shown in figure 42. 

Figure 42. Photo. Freeze-thaw chamber. 

The FRP specimens were fitted into special freeze-thaw fixtures that loaded the 

specimens in four-point bending. This arrangement was chosen to approximate the 

bending moment present in piles near the pile cap. The fixture used is shown in figure 43. 

Use of this fixture resulted in tensile strains in the outer extreme fiber of the 24-inch 
69 

diameter tube specimens ranging from 2,800 to 5,500 microstrains. The specimens from 

the 12-inch tubes were subjected to freeze-thaw cycles in an unloaded state, i.e., 

unstrained. The chamber was programmed to produce approximately 10 freeze-thaw 

cycles per day. A typical freeze-thaw cycle is shown in figure 44. Specimens were 

subjected to 100, 300, and 500 freeze-thaw cycles. As a point of interest, Southwest 

Virginia experiences about 50 freeze-thaw cycles per year (i.e., 500 freeze-thaw cycles in 

about 10 years). 

Figure 43. Photo and Illustration. Freeze-thaw fixture. 

Figure 44. Graph. Average freeze-thaw cycle undergone by FRP samples. 

More details on the freeze-thaw study are presented later in this chapter. 
70 

4.3.3 Baseline Mechanical Properties

To study the durability and long-term performance of FRP composite piles, it was 

necessary to first evaluate the baseline mechanical properties of the FRP shell in the 

longitudinal (axial) and hoop directions. 

 

4.3.3.1 Longitudinal tension 

The baseline properties of each pile type were obtained from a series of longitudinal 

tension tests, and the results are summarized in table 21. Typical stress-strain curves 

obtained from these tests for each FRP type are shown in figure 45. 

Table 21. As-received longitudinal tensile properties. 

Property Lancaster FRP Hardcore FRP tube 

24-inch 12-inch 24-inch 12-inch 

Tensile strength (MPa) Mean 113.1 249.3 432.2 494.3 (1)

SD (3) 7.3 17.7 46.9 60.6 

Peak strain (%) Mean 1.02 1.70 1.86 1.78 

SD 0.1 0.25 0.27 0.12 

Initial modulus (Gpa) Mean 14.8 23.2 27.3 29.6 (2)

SD 1.9 2.1 3.15 0.9 

Number of specimens N 29 9 9 5 

Notes: (1) Strength calculated using total thickness of FRP tube (includes liner). 

(2) Initial modulus calculated between 0 and 4,000 microstrains. 

(3) SD = Standard deviation 

(4) 1 inch = 25.4 mm; 1 MPa = 145 lbf/inch2

; 1 GPa = 145,000 lbf/inch2 

Figure 45. Graph. Representative baseline longitudinal tension stress-strain curves. 
71 

4.3.3.2 Hoop tension 

The baseline hoop tensile properties for the composite piles are summarized in table 22. 

Typical stress-strain curves obtained from split disks tests are shown in figure 46. 

 

Table 22. As-received hoop tensile properties. 

Property Lancaster FRP Hardcore FRP 

24-inch 12-inch 12-inch 

Mean 121.4 195.2 (2) Tensile strength (MPa) 93.5 (1)

SD (5) 5.9 15.9 7.9 

Mean 1.26 (3) Peak strain (%) 1.28 1.45 

SD 0.34 0.10 0.40 

Initial modulus (Gpa) Mean 19.1 16.5 11.9 (4)

SD 2.3 2.2 3.1 

Number of specimens N 5 11 5 

Notes: (1) Strength calculated using total thickness of FRP tube (includes liner). 

(2) The strain and modulus values are based only on the second batch of tests that used inner and 

outer strain gages (5 tests). 

(3) No inner gages were used for this test batch; hence strain values are from outer gages only. 

(4) Initial modulus calculated between 0 and 4,000 microstrain. 

(5) SD = Standard deviation. 

(6) 1 inch = 25.4 mm; 1 MPa = 145 lbf/inch2

; 1 GPa = 145,000 lbf/inch2

Figure 46. Graph. Representative baseline hoop tension stress-strain curves. 

It was not possible to perform split disk tests on the 24-inch (0.61 m) Hardcore FRP shell 

specimens due to the noncircular shape of the inside of these shell. Hardcore shells have 

several resin transfer tubes attached to the inner wall (running longitudinally) that are 

used to inject the resin into the composite during manufacturing of the FRP shell, and at 
72 

the end of the process, the tubes are left in place filled with resin. The size and number of 

these tubes made it impractical to test these specimens, as the internal clearance is not 

circular. 

4.3.4 Properties as a Function of Submergence Time and Moisture 

4.3.4.1 Moisture absorption tests 

Moisture absorption isothermal curves were obtained by immersing samples in fresh 

water tanks at the following temperatures: 22 o

C (72 °F), 35 oC (95 °F), 45 oC (113 °F), 

55 oC (131 °F), 65 oC (149 °F), and 80 oC (176 °F). The submerged FRP coupons were 

about 100 mm (3.9 inches) long and 25 mm (0.98 inch) wide, with their edges epoxycoated to promote one-dimension diffusion. Plots of moisture absorption versus time for 

the 12-inch and 24-inch (0.31-m and 0.61-m) diameter Lancaster FRP shells are shown in 

figures 47 and 48, respectively. Moisture absorption plots for the 12-inch and 24-inch 

(0.31-m and 0.61-m) diameter Hardcore FRP shells are shown in figures 49 and 50, 

respectively. The moisture content shown in these figures corresponds to the percentage 

of weight gain with respect to the initial weight of the samples before submergence. 

Figure 47. Graph. Absorption curves for Lancaster 12-inch FRP tube. 
73 

Figure 48. Graph. Absorption curves for Lancaster 24-inch FRP tube. 

Figure 49. Graph. Absorption curves for Hardcore 12-inch FRP tube. 
74 

Figure 50. Graph. Absorption curves for Hardcore 24-inch FRP tube. 

All moisture absorption tests showed a rapid gain of weight at the start of the 

experiments, with the rate of increase being the greatest at the highest temperature 

(80 o

C (176 °F)). With time, most of the tests below 65 oC (149 °F) stabilized, reaching a 

saturation or equilibrium moisture content. For the tests at 80 oC (176 °F), the weights 

either continued to increase gradually, or in some cases like the Lancaster 24-inch 

(0.61-m) at 80 oC (176 °F), the samples started to decrease sharply in weight after 

reaching a peak. Nishizaki and Meiarashi (2002) reported similar decreases in moisture 

content for samples submerged at high temperatures. The authors attributed this behavior 

to loss of material from the composite matrix. The Lancaster 24-inch (0.61-m) samples 

with water temperatures between 55 and 65 o

C (131 and 149 °F) seem to have a nonFickian or two-stage absorption behavior. Phifer (2003) found that, in some cases, 

Langmuirian diffusion predicts the behavior of E-glass vinyl ester composites better than 

Fickian diffusion. Diffusion analyses using Fickian and Langmuirian models are 

presented in the following subsection. 

4.3.4.2 Diffusion properties 

Two types of diffusion models were used to fit the experimental absorption data: Fickian 

and Langmuirian. These models are described briefly below. 

The Fickian diffusion model is commonly used to predict moisture absorption of 

composite materials. This model has the convenience of being simple. The model 

assumes a constant diffusivity, D, and a saturation moisture, M∞, to which the sample 
75 

tends to as the submergence time increases. The expression for Fickian diffusion of a thin 

plate of thickness, h, is given by the following expression (Shen and Springer 1976): 

 

 (7) 

where 

• M(t) = moisture content of the FRP specimen at submergence time, t

• M∞ = moisture content at saturation 

• t = time of submergence 

• D = diffusivity through the thickness of the FRP specimen 

• h = thickness of FRP specimen 

Gurtin and Yatomi (1979) proposed using a Langmuirian diffusion model for FRP 

composites exhibiting non-Fickian absorption behavior. The following expression is 

given to track the moisture content of specimens following Langmuirian diffusion: 

(8) 

where 

• M(t) = moisture content of the FRP specimen at submergence time, t

• M∞ = the Fickian moisture content at saturation 

• t = time of submergence 

• D = the Fickian diffusivity through the thickness of the FRP specimen 

• α and β = constants usually obtained from curve fitting 

• h = thickness of the FRP specimen 

The similarity of this model to the Fickian is evident from comparing equations 7 and 8. 

The first non-unity term in equation 8 is indicative of a two-stage diffusion process. More 

details on the Langmuirian model can be found in Gurtin and Yatomi (1979). 

Curve fits using the Fickian and Langmuirian models were performed on the 

experimental moisture absorption data gathered for the different FRP shells evaluated in 

this study. The curve fits for water temperatures 22 o

C (72 °F), 45 oC (113 °F), and 65 oC 

(149 °F) are shown in figures 51 and 52 for the 12-inch and 24-inch (0.31-m and 0.61-m) 

diameter Lancaster FRP shells, respectively. Fits for these same water temperatures for 

the 12-inch and 24-inch (0.31-m and 0.61-m) diameter Hardcore FRP shells are shown in 

figures 53 and 54, respectively. 
76 

(a) Water at 22 °C

(b) Water at 45 °C

(c) Water at 65 °C

Figure 51. Graphs. Selected diffusion analyses for Lancaster 12-inch FRP samples. 
77 

(a) Water at 22 °C 

(b) Water at 45 °C

(c) Water at 65 °C

Figure 52. Graphs. Selected diffusion analyses for Lancaster 24-inch FRP samples. 
78 

(a) Water at 22 °C

(b) Water at 45 °C

(c) Water at 65 °C

Figure 53. Graphs. Selected diffusion analyses for Hardcore 12-inch FRP samples. 
79 

(a) Water at 22 °C

(b) Water at 45 °C

(c) Water at 65 °C

Figure 54. Graphs. Selected diffusion analyses for Hardcore 24-inch FRP samples. 
80 

These figures indicate that Langmuirian diffusion, which has four curve-fitting 

parameters (D, M∞, α, β), is better able to track the measured moisture absorption than 

the Fickian diffusion, which has two curve-fitting parameters (D, M∞), for samples 

submerged at temperatures between 55 and 80 o

C (131 and 176 °F). 

The diffusivity (or diffusion coefficient D) and the saturation moisture content (M∞) 

obtained from Fickian diffusion analyses are summarized in tables 23 through 26 for the 

Lancaster 12-inch (0.31-m), Lancaster 24-inch (0.61-m), Hardcore 12-inch (0.31-m), and 

Hardcore 24-inch (0.61-m) samples, respectively. Although there is much scatter in the 

values, it can be seen in the tables that the diffusivity and the saturation moisture tend to 

increase with temperature. Even though Langmuirian diffusion provided a better fit to the 

data, the Fickian model and parameter values are used for predicting the long-term 

composite pile strength, as described later in this chapter, because of the simplicity of the 

Fickian model. 

Table 23. Fickian diffusion parameters for the Lancaster 12-inch FRP. 

Water Temperature (oC) Diffusivity, D (cm2/hr) Saturation Moisture, M∞ (%) 

22 1.25 E-4 0.36 

35 1.53 E-4 0.36 

45 8.57 E-5 0.40 

55 1.21 E-4 0.42 

65 9.22 E-5 0.43 

80 3.38 E-5 0.87 (1)

Note: (1) Did not stabilize at the end of test; value is only used for curve fitting Fickian curves to data. 

 (2) 1 cm2

 = 0.15 inch2

Table 24. Fickian diffusion parameters for the Lancaster 24-inch FRP. 

Water Temperature (oC) Diffusivity, D (cm2/hr) Saturation Moisture, M∞ (%) 

22 1.23 E-5 0.48 

35 2.04 E-5 0.45 

45 4.89 E-4 0.45 

55 1.48 E-4 0.40 (1)

65 3.08 E-4 0.43 (1)

80 4.59 E-4 0.45 (1)

Note: (1) Did not stabilize at the end of test; value is only used for curve fitting Fickian curves to data. 

 (2) 1 cm2

 = 0.15 inch2
81 

Table 25. Fickian diffusion parameters for the Hardcore 12-inch FRP. 

Water Temperature (o C) Diffusivity, D (cm2/hr) Saturation Moisture, M∞ (%) 

22 8.17 E-6 0.19 

35 1.99 E-5 0.19 

45 1.52 E-5 0.27 

55 4.81 E-5 0.22 

65 6.08 E-5 0.18 

80 1.17 E-5 0.50 (1)

Note: (1) Did not stabilize at the end of test; value is only used for curve fitting Fickian curves to data. 

 (2) 1 cm2

 = 0.15 inch2

Table 26. Fickian diffusion parameters for the Hardcore 24-inch FRP. 

Water Temperature (oC) Diffusivity, D (cm2/hr) Saturation Moisture, M∞ (%) 

22 1.16 E-4 0.21 

35 7.85 E-5 0.26 

45 4.42 E-5 0.26 

55 9.54 E-5 0.24 

65 2.30 E-4 0.26 

80 1.70 E-4 0.26 (1)

Note: (1) Did not stabilize at the end of test, value is only used for curve fitting Fickian curves to data. 

(2) 1 cm2 = 0.15 inch2

4.3.4.3 Measured properties as a function of submergence time and moisture 

The effect of time of submergence and moisture on the mechanical properties of the FRP 

tubes was evaluated by performing longitudinal and hoop tension tests on specimens aged 

in 22 oC (72 °F) water for submergence times ranging between 4 and 787 days. Typically, 

five or more specimens were tested at each submergence period. The tests were carried 

out using the same procedure used to determine the baseline properties. 

Variations of the mechanical properties of the aged FRP samples are presented in the 

following subsections. The results presented are normalized with respect to the baseline 

property values. 

 

4.3.4.3.1 Aged properties for Lancaster 24-inch FRP tube 

Variations of the longitudinal and hoop properties versus submergence time for this FRP 

shell are shown in figures 55 and 56, respectively. The error bars shown in these figures 

correspond to ± one standard deviation. Longitudinal tensile test results on this particular 

FRP composite shell showed no evidence of significant degradation of property values. 

This is likely related in part to the fiber orientation of this composite (±35o

/85o/±35o)T, 

which makes it a matrix-dominated composite in the longitudinal direction, and also 

possibly due to a gain of ductility of the matrix upon immersion in water. Tensile 

properties in the hoop direction showed some degradation, but it was not very significant. 

The hoop tensile strength decreased by about 8 percent. The error bars shown in these 

figures indicate considerable scatter. 
82 

Figure 55. Graph. Longitudinal tensile properties versus submergence time 

for Lancaster 24-inch FRP tube. 

Figure 56. Graph. Hoop tensile properties versus submergence time 

for Lancaster 24-inch FRP tube. 
83 

4.3.4.3.2 Aged properties for Lancaster 12-inch FRP tube 

The longitudinal and hoop properties versus submergence time for this FRP shell are 

shown in figures 57 and 58, respectively. 

Figure 57. Graph. Longitudinal tensile properties versus submergence time 

for Lancaster 12-inch FRP tube. 

Figure 57 shows that after approximately 400 days of submergence in fresh water at 

22 oC (72 °F), the longitudinal tensile strength and strain to failure decreased about 9 and 

18 percent, respectively. The initial stiffness (in the 0 to 4,000 microstrains range) 

remained almost constant. 
84 

Figure 58. Graph. Hoop tensile properties versus submergence time 

for Lancaster 12-inch FRP tube. 

Figure 58 shows that the hoop tensile strength and initial stiffness decreased about 17 

percent and 18 percent, respectively. Degradation levels for strength and stiffness are 

more significant for the 12-inch (0.31-m) sample compared to the 24-inch (0.61-m) 

sample, because the fiber layup of the 12-inch (0.31-m) composite is more fiber 

dominated. However, the strain to failure did not change much. 

Strength and stiffness versus percentage of moisture absorption are shown in figures 59 

and 60 for the longitudinal and hoop directions, respectively. Both figures show tensile 

strength decreases with increasing moisture content, but the strength values tend to 

stabilize at higher moisture contents. The hoop tensile stiffness displayed a similar 

behavior, as shown in figure 60b. These plots are useful for predicting mechanical 

properties versus time, as shown later in this chapter. 
85 

Figure 59. Charts. Longitudinal tensile properties versus moisture content 

for Lancaster 12-inch tube. 

Moisture content (%)

0.0 0.1 0.2 0.3 0.4 0.5 0.6

Normalized longitudinal tensile modulus ET/E0

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

Moisture content (%)

0.0 0.1 0.2 0.3 0.4 0.5 0.6

Normalized longitudinal tensile strength XT/X0

0.0

0.2

0.4

0.6

0.8

1.0

1.2

≈ 26 days

submerge

≈ 420 

days

a) Variation of strength 

b) Variation of stiffness 
86 

Figure 60. Charts. Hoop tensile properties versus moisture content 

for Lancaster 12-inch FRP tube. 

Moisture content (%)

0.0 0.2 0.4 0.6 0.8

Normalized hoop tensile strength XT/X0

0.0

0.2

0.4

0.6

0.8

1.0

1.2

Moisture content (%)

0.0 0.2 0.4 0.6 0.8

Normalized hoop modulus ET/E0

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

a) Variation of strength 

b) Variation of stiffness 
87 

4.3.4.3.3 Aged properties for Hardcore 24-inch FRP tube 

The longitudinal properties versus submergence time for this FRP shell are shown in 

figure 61. 

Figure 61. Graph. Longitudinal tensile properties versus submergence time 

for Hardcore 24-inch FRP tube. 

Figure 61 shows that after approximately 636 days of submergence in fresh water at 

22 oC (72 °F), the longitudinal tensile strength decreased about 13 percent. The strain to 

failure decreased by about 20 percent. The initial stiffness values (0 to 4,000 

microstrains) did not show a clear trend with respect to the baseline values. The error bars 

shown indicate that the strain and stiffness values had considerable scatter. 
88 

The variation of the longitudinal tensile strength and stiffness versus percentage of 

moisture absorption are shown in figure 62. This figure shows a tensile strength loss of 

about 16 percent at the highest moisture content value of 0.33 percent. The initial 

stiffness values show considerable scatter, and no clear trend can be determined. 

Figure 62. Charts. Longitudinal tensile properties versus moisture content 

for Hardcore 24-inch FRP tube. 

a) Variation of strength 

b) Variation of stiffness 

Moisture content (%)

0.0 0.1 0.2 0.3

Normalized longitudinal tensile strength XT/X0

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

Moisture content (%)

0.0 0.1 0.2 0.3

Normalized longitudinal tensile modulus ET/E0

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

1.6
89 

4.3.4.3.4 Aged properties for Hardcore 12-inch FRP tube 

The longitudinal and hoop tensile properties versus submergence time for this FRP shell 

are shown in figures 63 and 64, respectively. 

 

Figure 63. Graph. Longitudinal tensile properties versus submergence time 

for Hardcore 12-inch FRP tube. 

Figure 63 shows that after approximately 398 days of submergence in fresh water at 

22 oC (72 °F), the longitudinal tensile strength and strain to failure decreased by about 24 

percent and 25 percent, respectively. The initial stiffness increased by about 10 percent. 
90 

Figure 64. Graph. Hoop tensile properties versus submergence time 

for Hardcore 12-inch FRP tube. 

Figure 64 shows that after approximately 464 days of submergence in fresh water at 

22 oC (72 °F), the hoop tensile strength decreased by about 20 percent. The stiffness and 

strain-to-failure values showed very high scatter. This could be related to material 

variability. Another possible reason could be related to the infusion tubes used to inject 

the resin during manufacturing of the FRP tubes. These tubes were removed prior to hoop 

split disk testing; however, at these locations the FRP shell wall thickness increased and 

decreased sharply within a small distance. It is possible that high stress concentrations 

occurred at these locations and affected the results. 

The variation of tensile strength and stiffness versus percentage of moisture absorption 

for the longitudinal and hoop directions are shown in figures 65 and 66, respectively. 
91 

Figure 65. Charts. Longitudinal tensile properties versus moisture content 

for Hardcore 12-inch FRP tube. 

As shown in figure 65a, the longitudinal tensile strength decreases with increasing 

moisture content. The maximum tensile strength degradation was about 25 percent at a 

moisture content of about 0.25 percent. No clear trend was observed between longitudinal 

tensile modulus and moisture content. 

a) Variation of strength 

b) Variation of stiffness 

Moisture content (%)

0.00 0.05 0.10 0.15 0.20 0.25 0.30 0.35

Normalized longitudinal tensile strength XT/X0

0.0

0.2

0.4

0.6

0.8

1.0

1.2

Moisture content (%)

0.0 0.1 0.2 0.3

Normalized longitudinal tensile modulus ET/E0

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

1.6

1.8
92 

Figure 66. Charts. Hoop tensile properties versus moisture content for 

Hardcore 12-inch FRP tube. 

Figure 66 shows very high scatter for hoop tensile strength and stiffness, hence no clear 

relationship was observed with moisture content. As mentioned before, the high scatter in 

the hoop tensile properties for the 12-inch (0.31-m) Hardcore shell could be related to 

material variability, and the sharp variations of FRP shell wall thickness near the location 

of the resin infusion tubes may have created high stress concentrations and affected the 

results. 

a) Variation of strength 

Moisture content (%)

0.0 0.2 0.4 0.6 0.8

Normalized hoop tensile strength XT/X0

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

Moisture content (%)

0.0 0.2 0.4 0.6 0.8

Normalized hoop tensile modulus ET/E0

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

b) Variation of stiffness 
93 

4.3.5 Freeze-Thaw Degradation of Saturated FRP Samples

Longitudinal tensile tests were performed on saturated coupon samples after 0, 100, 300, 

and 500 freeze-thaw cycles. The saturated coupons refer to specimens that were soaked in 

22 o

C (72 °F) water until they reached moisture content at saturation, that is, M∞ (see 

figures 47 through 50 for the time of soaking required to reach M∞). Three or more 

samples were tested at each stage. The typical temperature cycle undergone by the 

specimens is shown in figure 44. The specimens were cycled between -17 oC (1.4 °F) and 

+6 oC (43 °F), and the duration of each cycle was about 2.4 hours. The test setup is 

described in section 4.3.2.4. 

The results are presented in figures 67 through 70. These figures are normalized with 

respect to the properties obtained from samples submerged in 22 o

C (72 °F) water at or 

near the saturation moisture content (M∞). 

Figure 67. Graph. Influence of freeze-thaw cycling on the longitudinal tensile 

properties for the Lancaster 24-inch FRP tube. 
94 

Figure 68. Graph. Influence of freeze-thaw cycling on the longitudinal tensile 

properties for the Lancaster 12-inch FRP tube. 

Figure 69. Graph. Influence of freeze-thaw cycling on the longitudinal tensile 

properties for the Hardcore 24-inch FRP tube. 
95 

Figure 70. Graph. Influence of freeze-thaw cycling on the longitudinal tensile 

properties for the Hardcore 12-inch FRP tube. 

The freeze-thaw data indicates little change in the longitudinal tensile properties of the 

FRP tubes except for the Lancaster 24-inch (0.61-m), where a strength decrease of about 

20 percent was recorded after exposure to 500 freeze-thaw cycles. 

 

4.3.6 SEM Imaging 

Scanning electron microscopy was used to investigate the damage mechanisms for the 

FRP composite tubes. The SEM images allowed examination of the state of the fibers and 

the surrounding matrix at different exposure conditions used in the experimental 

program. For each FRP tube type, SEM micrographs were obtained for the following 

conditions: as-received, at the end of moisture absorption in 22 o

C (72 °F) water (i.e., 

saturation), and after freeze-thaw cycling. SEM images for the four types of FRP tubes 

investigated are shown in figures 71 through 74. 
96 

Figure 71. Photo. SEM images of Lancaster 24-inch FRP tube. 

Figure 72. Photo. SEM images of Lancaster 12-inch FRP tube. 

Figure 73. Photo. SEM images of Hardcore 24-inch FRP tube. 
97 

Figure 74. Photo. SEM images of Hardcore 12-inch FRP tube. 

Some damage was noticed in the SEM images taken at the saturation moisture state. The 

signs of degradation included damage at the fiber and at the interface between the fiber 

and matrix. 

 

4.4 LONG-TERM STRUCTURAL CAPACITY OF FRP COMPOSITE PILES 

The durability experimental program described above was based on testing FRP coupons 

submerged in 22 o

C (72 °F) water with their edges coated to approximate 1-D diffusion 

conditions. For a concrete-filled FRP pile, the diffusion will be predominantly in the 

radial direction. Furthermore, the boundary conditions on the inner and outer surfaces of 

the FRP shell of the pile will be different from the boundary conditions of the 

experimental program. A more realistic estimation of the FRP strength as a function of 

time can be modeled by calculating the average moisture content of the FRP shell using 

the approach described in Appendix B and in Pando, et al. (2001). This approach 

involves calculating the moisture content of the FRP shell of the pile versus time, using a 

radial moisture diffusion model and incorporating more appropriate boundary conditions. 

Once the relationship between the FRP shell moisture content and time is computed, one 

can calculate the FRP strength-versus-time relationships using the experimental data 

presented in section 4.3. Although relationships of FRP properties versus time are useful, 

for the purpose of estimating the long-term capacity we only need the residual values for 

the FRP mechanical properties (i.e., the estimated long-term FRP properties). 

 

4.4.1 Long-Term Structural Capacity of Composite Pile 

This section presents a simplified example, intended for illustrative purposes only, to 

demonstrate the importance of considering evaluation of moisture aging effects on the 

FRP strength and stiffness (especially for bending capacity). The long-term structural 

capacity of concrete-filled FRP piles can be estimated using the short-term models for 

axial and flexural structural behavior by Fam and Rizkalla (2001a and b, 2002), coupled 

with the long-term mechanical properties of the FRP tube after degradation has occurred. 

The long-term mechanical properties of the FRP tube can be estimated using 

experimental relationships between FRP properties and moisture content, such as those 
98 

presented earlier in this chapter. The moisture content used to estimate the long-term FRP 

properties should correspond to the long-term average moisture concentration expected in 

the field for the FRP tube in the pile. A procedure to estimate the long-term moisture 

content for a cylindrical FRP is described in Appendix B and in Pando, et al. (2002b). 

 

As an example, the long-term axial capacity for the 12-inch (0.31-m) Lancaster pile is 

estimated. With the FRP dimensions and as-received properties presented in Section 

4.3.3, and using the model of Fam and Rizkalla (2001a and b), we obtain the axial loadstrain response curve shown as a solid line in figure 75. If we now assume a 25 percent 

and 40 percent long-term reduction for the FRP hoop stiffness and strength, respectively, 

we obtain the long-term load-strain curve shown as the dashed line in this same figure 

(note these degradation values are higher than the measured degradation levels in this 

study). This represents approximately a 5 percent ultimate (structural) axial capacity 

reduction. 

Figure 75. Graph. Estimated long-term axial capacity of the 12-inch Lancaster pile. 

As shown in figure 75, the impact of degradation of the FRP mechanical properties on the 

long-term axial structural capacity of concrete-filled FRP composite piles is small due to 

the fact that the majority of the capacity contribution is from the concrete infill. 

Similarly, the long-term flexural capacity (structural) for the same Lancaster pile can be 

predicted using the model proposed by Fam and Rizkalla (2002). The short-term 

moment-curvature response is shown as the solid line in figure 76. Assuming similar 

stiffness and strength reduction levels as before, but in the longitudinal direction, we 
99 

obtain the dashed moment-curvature line shown in figure 76. This represents a 24 percent 

reduction in the long-term structural flexural capacity. 

Figure 76. Graph. Estimated long-term flexural capacity 

of the 12-inch Lancaster pile. 

Figure 76 shows that the impact of FRP degradation is more significant for the flexural 

capacity because the FRP shell provides most of the capacity contribution on the tension 

side of the pile in flexion. 

4.4.2 Comments Related to the Axial Strain Levels in Piles 

The structural ultimate limit state for axially loaded concrete-filled FRP columns or piles 

occurs at axial strain levels that exceed the normal range of ultimate strains typically used 

for normal unconfined concrete (e.g., 0.001 to 0.003). As shown in chapter 2, and as 

reported by Fam and Rizkalla (2002) and others, the presence of the FRP shell allows this 

type of structural element to undergo much larger axial strains (e.g., to about 0.009 to 

0.013). It is important to compare these strain levels to the strain levels that can be 

allowed to develop in a pile before serviceability limits are exceeded. 

The strain distribution along the pile depends on the soil conditions. A pile that develops 

all or most of its resistance from end bearing, such as piles with their tips founded on 

hard soil or rock, have almost a constant axial strain distribution with depth. For piles that 

develop their resistance from side friction, the strain decreases with depth at a rate 

dependent on the side friction distribution along the pile length. The maximum strain in a 

pile will occur within the portion of the pile above the ground surface. The maximum 

axial strains in an axially loaded pile will be influenced by the allowable pile settlement 
100 

determined from serviceability requirements. Typically the allowable settlement is 

between 25 and 50 mm (0.975 and 1.95 inches). For the case of an end-bearing pile with 

an allowable settlement of 37 mm (1.44 inches), the length of the pile would have to be 

less than 12 m (39 ft) to develop axial strains in the pile above 0.003. For piles carrying 

their load purely in side friction, and with an allowable settlement of 37 mm (1.44 

inches), the length of the pile would have to be less than 24 m (79 ft) to start developing 

axial strains above 0.003. The pile lengths would have to be shorter if the allowable pile 

settlement is below 37 mm (1.44 inches). This simple example illustrates how the 

additional strength and ductility provided by the FRP shell may not be utilized in the field 

for piles with serviceability requirements, as described above. Thus, serviceability may 

control design of FRP piles in many practical applications. 

4.5 SUMMARY 

A laboratory testing program was completed to study the long-term performance of FRP 

composite pipe piles. This durability study, which addressed the FRP shells of Lancaster 

and Hardcore composite piles, included FRP shell characterization, determination of 

baseline mechanical properties, measurement of moisture absorption as a function of time 

and temperature, measurement of mechanical properties as a function of moisture 

absorption, and measurement of mechanical properties as a function of freeze-thaw 

cycles. 

From the results of the durability study, the following conclusions and observations were 

made. 

1. Moisture absorption has a small impact on the strength of some FRP composites, and 

a larger impact on others. 

 

2. Curve fits using the Fickian and Langmuirian models were performed on the 

experimental moisture absorption data gathered for the different FRP shells evaluated 

in this study. The results indicate that Langmuirian diffusion, which has four curvefitting parameters (D, M∞, α, β), is better able to track the measured moisture 

absorption than the Fickian diffusion, which has two-curve fitting parameters (D, 

M∞). This finding is particularly relevant for samples submerged at temperatures 

between 55 and 80 o

C (131 and 176 °F). 

 

3. The levels of strength degradation corresponding to steady state moisture contents 

reached after about 2.5 years of submergence were as follows: 

 

- Longitudinal tensile strength reductions on saturated FRP specimens were 

measured to be 0.5 , 7.1, 12.7, and 24.9 percent for the Lancaster 24-inch 

(0.61-m), Lancaster 12-inch (0.31-m), Hardcore 24-inch (0.61-m), and 

Hardcore 12-inch (0.31-m) samples, respectively. 

 

- Measured hoop tensile strength reductions due to moisture were 7.6, 17.3, and 

19.6 percent for the Lancaster 24-inch (0.61-m), Lancaster 12-inch (0.31-m), 
101 

Hardcore 24-inch (0.61-m), and Hardcore 12-inch (0.31-m) samples, 

respectively. 

 

4. Degradation of longitudinal mechanical properties for the 24-inch (0.61-m) Lancaster 

pile was not as significant as for the Hardcore pile because of the different fiber 

layups. The Lancaster pile was matrix dominated in the longitudinal direction because 

the fibers with closest alignment to the longitudinal axis were 35 degrees off 

alignment. 

5. The impact of FRP degradation on the long-term structural capacity of the piles was 

investigated using models by Fam and Rizkalla (2001a and b, and 2002). This 

approach showed that, for a 12-inch (0.31-m) diameter FRP pile, the axial structural 

capacity will decrease about 5 percent if the FRP tube hoop properties degrade 25 

percent and 40 percent in stiffness and strength, respectively. The small impact that 

the FRP degradation has on the axial pile capacity is due to the fact that majority of 

the capacity contribution is coming from the concrete infill. The impact on larger 

diameter piles is expected to be even smaller. For the flexural capacity of a 12-inch 

(0.31-m) pile, the results show that a 24 percent reduction in flexural capacity can be 

expected if the FRP tube longitudinal tensile properties degrade 25 percent and 40 

percent in stiffness and strength, respectively. This result is as expected, because the 

concrete is weak in tension and the FRP tube provides the tensile reinforcement to the 

pile during bending. 

6. Exposure to freeze-thaw cycles was found to have little effect on the longitudinal 

tensile properties of the saturated FRP tubes. This is based on longitudinal tensile 

tests on samples exposed to 100, 300, and 500 freeze-thaw cycles.

103 

CHAPTER 5. FIELD LOAD TESTS AT THE ROUTE 40 BRIDGE 

5.1 INTRODUCTION 

In 2000, the Virginia Department of Transportation (VDOT) employed concrete-filled 

FRP composite piles for an entire bent of the new Route 40 Bridge over the Nottoway 

River in Sussex County, VA. As part of this project, a test pile program was performed at 

the bridge site to investigate and compare the driveability, axial load capacity, and lateral 

load response of a composite pile and a standard prestressed concrete pile. The same type 

of composite pile was used in the test pile program as was used in the bridge bent. This 

project is believed to be the first to use this type of composite pile in a long-term, loadbearing application such as a bridge. Previously, these piles have been primarily used in 

marine fendering projects, in light load applications, or for experimental research. A 

summary of the overall project and the main findings of the field load test program are 

presented in this chapter. Results from this project have also been published by Pando, et 

al. (2000), and Fam, et al. (2003). 

5.2 DESCRIPTION OF THE BRIDGE 

5.2.1 The Former Bridge 

VDOT decided to replace the existing Route 40 Bridge (Structure No. 1006) over the 

Nottoway River in Sussex County, VA, at the location shown in figure 77. The old 

bridge, which consisted of a steel truss supported by concrete piers as shown in figure 78, 

suffered from excessive deterioration in both superstructure and substructure. Full-height 

vertical cracks existed in the concrete columns of the piers, as shown in figure 79(a). 

Spalling and cracking of the concrete bearing seats were observed, as shown in figure 

79(b). Map cracking, up to 1.6 mm (1/16 inch) width, was also observed in the abutment 

bearing seats, as shown in figure 79(c). In addition to the deterioration of the 

substructure, excessive corrosion with large section losses was observed in the roller 

bearing devices of the truss supports. Additionally, the floor beams suffered from section 

losses up to 3.2 mm (1/8 inch) in both the web and bottom flanges. The bridge was 

functionally obsolete with a roadway clearance of only 7.3 m (24 ft) curb to curb, and 

was rated as scour critical. 

 
104 

Figure 77. Map. Location map of the Route 40 Bridge project in 

Sussex County, VA (Fam, et al., 2003). 

Figure 78. Photo. Former Route 40 Bridge. 
105 

Figure 79. Photos. Signs of deterioration of the former Route 40 

Bridge (Fam, et al., 2003). 

5.2.2 The New Bridge 

The new bridge is 85.3 m (280 ft) long and 15.2 m (50 ft) wide, and consists of a fivespan slab-type superstructure supported by four piers and two end abutments, as shown in 

figure 80. Each pier consists of a reinforced concrete beam-type elevated pile cap 

supported by ten piles. Based on the design requirements, VDOT specified a 508 mm 

(20-inch) prestressed solid square pile, as shown in figure 83. The design axial load of 

each pile was 667 kN (150 kips). As a part of the composite piles study sponsored by the 

Federal Highway Administration (FHWA), VDOT took the initiative to use an alternative 

composite pile system consisting of concrete-filled FRP tubes for the entire group of piles 

supporting Pier No. 2 (see figure 80). The objectives of the VDOT initiative were to 

examine the feasibility of using composite piles for bridge substructures, particularly to 

examine the driveability of the piles using conventional pile drivers, to assess 

performance under axial and lateral loading conditions using full-scale field tests, and to 

compare composite piles to conventional prestressed concrete piles. The following 
106 

section provides a brief description of both the prestressed concrete test pile and the 

precast composite test pile. 

Figure 80. Illustration. Schematic of the new Route 40 Bridge. 
107 

5.3 DESCRIPTION OF TEST PILES 

5.3.1 Composite Test Pile 

The composite test pile used for this project consists of a concrete-filled tubular FRP pile, 

in which the FRP tube serves as both a permanent lightweight noncorrosive formwork for 

the concrete and a reinforcement element. Figure 81(a) is a photograph of the composite 

piles installed at bridge bent No. 2. The FRP tube is composed of several layers of fibers 

embedded in a polymeric resin. The layers of fibers are oriented in several different 

directions with respect to the longitudinal axis of the tube in order to provide strength and 

stiffness in both the axial and circumferential directions. The circumferential stiffness and 

strength of the FRP tube provides confinement to the concrete core under axial stress. 

This confinement increases the strength and ductility of the pile. The axial strength of the 

FRP tube contributes to the flexural strength of the pile. 

A 13.1 m (43 ft) long GFRP tube, type CP40, was fabricated using a filament-winding 

technique wherein E-glass continuous fiber rovings were impregnated with polyester 

resin and wound over a rotating steel mandrel, following a predetermined winding 

pattern. The average measured fiber volume fraction of the FRP tube was 51.2 percent. 

The FRP tube has an average outer diameter of 625 mm (24.6 inches) and a total wall 

thickness of 7.35 mm (0.289 inches). The tube has a 1.68-mm- (0.066-inch-) thick liner at 

the inner surface, which results in a net structural wall thickness of 5.67 mm (0.223 

inches). The wall structure of the tube consists of five layers, each 1.13 mm (0.044 

inches) thick. The two inner and outer layers contain fibers oriented at θ = ±35 degrees, 

where θ is the inclination of the fibers with respect to the longitudinal axis of the tube. 

The middle layer, which is sandwiched between the inner and outer layers, contains 

fibers oriented at θ = 85 degrees, which means that the middle fibers are oriented nearly 

circumferentially. The laminate structure of the FRP tube is shown in figure 81(b). The 

mechanical properties of the FRP tube are provided in table 27. 

The FRP tube was filled with a concrete mix that includes an expansive additive to 

reduce the effect of shrinkage and thereby enhance the bond between the concrete and the 

FRP tube. The specified compressive strength of the concrete fill at 28 days is 41.4 MPa 

(6 ksi). Compressive strength tests carried out on concrete core samples from the 

composite test pile are shown in figure 82. The measured compressive strength is about 

60 MPa. The measured modulus of elasticity is about 15 GPa, which is smaller than the 

typical values expected for normal concrete of the same compressive strength. The reason 

for the small modulus of elasticity value for the concrete in the composite pile was not 

investigated, but it could be related to the expansive additive. 
108 

Figure 81. Photo and Illustration. Concrete-filled tubular piles. 

Table 27. Mechanical properties of FRP shell of composite pile. 

Mechanical Property* Longitudinal Axial Direction Hoop Direction 

Tensile strength 113 MPa 138 MPa 

Tensile modulus** 15 GPa 22.1 GPa 

Strain to failure 1–1.4% 1.25% 

Notes: * From coupon tests (see chapter 4 for more details on mechanical properties of FRP shell) 

** Elastic modulus for the strain range between 0 to 4,000 microstrains. 

1 MPa = 145 lbf/inch2

; 1 GPa = 145,000 lbf/inch2 
109 

Figure 82. Graph. Stress-strain response of concrete used in the composite pile. 

5.3.2 Prestressed Concrete Test Pile 

The performance of the composite test pile was compared with that of a prestressed 

concrete test pile, which is the pile type commonly used in VDOT bridge projects. The 

prestressed concrete test pile is a 508-mm (20-inch) square pile with a length of about 

13.1 m (43 ft). Figure 83 shows the pile details. The pile is prestressed using a total of 

fourteen, 12.7-mm (0.5-inch) diameter, 7-wire strands of 1861 MPa (270 ksi) ultimate 

strength, pretensioned to produce a prestress level of 5.6 MPa (0.809 ksi) in accordance 

with VDOT standards. The specified concrete compressive strength at 28 days is 40 MPa 

(5.8 ksi). The ties consist of No. 16M (No. 5) gage spiral wire, as shown in figure 83. 

Concrete test results indicated a 28-day compressive strength of about 43 MPa (6.2 ksi) 

for this pile. 
110 

Figure 83. Illustration. Reinforcement details of prestressed concrete pile. 

5.4 SOIL CONDITIONS AT THE TEST SITE 

In general, the soil conditions in the vicinity of the test piles consist of about 7.9 m 

(25.9 ft) of loose to medium-dense silty sand with some gravel, underlain by very stiff to 

hard silty clay, as shown in figure 84. The clay was encountered to the maximum depth 

investigated, which was 15.7 m (51.5 ft). The ground water level was observed at a depth 

of about 8.1 m (26.6 ft). 

The sand in the upper 4 m (13.1 ft) was brown, medium fine, with some fine gravel and 

traces of silt. Standard penetration resistance (SPT) blow counts (N-values) in the upper 

sand were five blows per 0.3 m (0.98 ft), which is indicative of a loose relative density. 

The sand below 4 m (13.1 ft) and extending to a depth of 7.9 m (25.9 ft) was brown to 

gray, fine-grained, silty with some gravel. SPT blow counts in the lower sand ranged 

from 17 to 25 blows per 0.3 m (0.98 ft), which corresponds to a medium density. 

Underlying the sand was green, very stiff to hard, silty clay, with SPT blow counts 

ranging from 41 to 88. The average SPT N-value below 10 m (32.8 ft) depth was 75 

blows per 0.3 m (0.98 ft). The near-surface stratum of loose sand comprises the soil of 

most significance to the lateral response of the piles. 

 
111 

Figure 84. Illustration. Simplified soil stratigraphy near test pile area. 

5.5 PILE FABRICATION AND INSTALLATION 

5.5.1 Fabrication of the Piles 

The prestressed concrete piles were fabricated at the precasting plant in accordance with 

VDOT standards, as shown in figure 85. After fabricating and curing, the piles were 

shipped to the bridge site. The FRP tubes for the composite piles were shipped to the 

precast plant and filled with concrete while held in an inclined position and supported by 

a steel beam along the full length of the tubes, as shown in figure 86(a and b). Wooden 

end plugs were used to seal both ends of the tube. The plugs were secured in position 

using metal straps connected to both the plug and the composite tube in radial directions, 

as shown in figure 86(b). Concrete was pumped into the upper end through a hole in the 

wooden plug. The concrete-filled composite tubes were handled using eight-point lifting 

devices along the length of the pile, as shown in figure 86(c). 
112 

Figure 85. Photo. Fabrication of prestressed concrete pile. 
113 

 

Figure 86. Photos. Fabrication of concrete-filled FRP Piles. 
114 

5.5.2 Pile Installation 

Both the composite and prestressed test piles were driven using a hydraulic impact 

hammer, type ICE model 160S, which was also used to drive the production piles for the 

Route 40 Bridge. The ram weight is 71.2 kN (16,000 lbs), and the manufacturer’s rated 

energy at the maximum stroke of 1.2 m (3.9 ft) is 85.4 kN-m (63,025 lbf-ft). A 190-mm- 

(7.5-inch-) thick plywood pile cushion was used during pile driving. Figure 87(a and b) 

shows the composite pile during and after driving, as well as the prestressed pile during 

driving. 

Figure 87. Photos. Driving of test piles. 

First, the composite pile was driven to a depth of about 8.5 m (27.9 ft), which was 

slightly above the originally planned embedment depth. The composite pile driving was 

stopped sooner than planned based on the driving monitoring information, which 

suggested that the 4,000 kN Statnamic device might not have been sufficient to test the 

pile if the pile were driven deeper. Next, the prestressed pile was driven to a 9.4 m (30.8 

ft) depth. This depth produced the same pile tip elevation for both test piles. Figure 88 

shows the driving records for both piles expressed in terms of number of blows per 

0.25 m (0.82 ft) versus the depth of the pile tip below the driving template. At the end of 

driving, the blow count was 6 blows per 25 mm (1 inch) of pile penetration for the 

composite pile and 4 blows per 25 mm (1 inch) for the prestressed concrete pile. 
115 

 

Figure 88. Graph. Driving records for test piles. 

During pile driving, two sets of independent dynamic measurements were taken: Pile 

Driving Analyzer (PDA) measurements recorded by Federal Highway Administration 

(FHWA) personnel, and Foundation Pile Diagnostic System (FPDS) measurements. A 

summary of the most relevant dynamic testing results is presented herein, and a more 

detailed discussion can be found in Muchard, et al. (1999). Dynamic testing was 

accomplished by attaching strain transducers and accelerometers at about 2 diameters 

from the top of the piles in pairs spaced 180 degrees apart. To attach the transducers to 

the composite pile, small windows were cut out of the FRP shell, which allowed the 

instrumentation to be directly attached to the concrete infill. PDA records for both types 

of piles at the end of initial driving are shown in figure 89. Both pile types show similar 

dynamic behavior. The PDA records for both piles show two force peaks: the first at 

about 0 and the second at about L/c (where L is the pile length, and c the wave speed). 

This type of response is produced by the natural frequency of the hammer and driving 

train used. A velocity peak can also be seen at about 2L/c. This peak is characteristic of 

reflected tensile waves from the toe. The recorded driving tensile stresses were relatively 

small. This is partly explained by the relatively heavy hammer, which generates larger 

compressive waves than the reflected tensile waves coming up from the toe. Table 28 

provides some of the measurements obtained during pile driving. It can be seen that the 

wave speeds, maximum compressive stresses, and maximum tensile stresses are each 

similar in the different pile types. The measured stress levels are lower than the allowable 

stresses recommended for prestressed piles (Hannigan, et al., 1996). No standards are 
116 

currently available for allowable driving stresses of composite piles. Estimates of pile 

capacity from the dynamic strain and acceleration measurements were obtained using the 

Case method (Muchard, et al., 1999). Using a Case damping value of 0.4, Muchard, et al., 

(1999) estimated end-of-driving pile capacities of 3390 kN (762 kips) and 3160 kN (710 

kips) for the composite and prestressed test piles, respectively. 

Figure 89. Graphs. End-of-driving PDA recordings. 
117 

Table 28. Pile-driving measurements for the prestressed and composite piles.

Pile Type 

Measurement 

Prestressed Composite 

Wave speed 3705 m/s 3612 m/s 

Maximum compression stress 

measured during driving 17.6 MPa 19.2 MPa 

Maximum tensile stress 

measured during driving 4.8 MPa 2.9 MPa 

Allowable stresses Tension < 7 MPa 

Comp. < 31 MPa 

No standards 

available 

1 m = 3.28 ft; 1 MPa = 145 lbf/inch2 

Using the TNOWAVE wave equation software, Muchard, et al. (1999), performed 

signal-matching analyses on strain and acceleration data for selected blows near the end 

of each pile driving. The signal matching was performed using a two-layer soil model 

consisting of a 7.5 m (24.6 ft) sand layer overlaying stiff clay. The predicted capacities 

from signal-matching analyses are summarized in table 29. 

Table 29. Summary of signal-matching analyses at end of driving 

(from Muchard, et al., 1999). 

Signal-Matching Analyses Predicted Capacity Pile Type Total Skin End Bearing 

Prestressed 3070 kN 1780 kN 1290 kN 

Composite 3140 kN 1750 kN 1390 kN 

1 kN = 225 lbf 

The signal-matching analyses predictions suggest that about 57 percent of both pile 

capacities is due to skin friction contribution. When comparing the pile capacities 

described above, it is important to keep in mind the differences in end bearing and skin 

friction areas of the piles. Due to differences in cross section and embedment depths, the 

composite pile has a skin area 12.6 percent smaller, and a tip area 18.9 percent larger, 

than the prestressed pile. 

5.6 STATNAMIC FIELD TESTING OF TEST PILES 

Prior to installation of the production piles for the new Route 40 Bridge, the two test piles 

were subjected to axial and lateral loading using a Statnamic testing device. The 

Statnamic testing system is a patented system. This system applies the load to the pile by 
118 

combustion of fuel within a closed chamber, in which gas pressure builds and “launches” 

a heavy reaction mass upward (or laterally) at a very high acceleration close to 20 g’s. 

When the reaction mass is launched, an equal and opposite reaction force acts on the test 

pile. The Statnamic setup is such that the load is applied in a somewhat gradual manner, 

as opposed to an impact load like the one produced by a driving hammer. The applied 

force is a time-dependent load with a duration typically between 0.2 to 0.3 seconds. From 

the recorded Statnamic force, one can estimate the equivalent static load by subtracting 

the inertia and damping forces. This equivalent static force can be derived by using the 

Unloading Point Method (UPM) proposed by Middendrop, et al. (1992). The Statnamic 

test methodology and data reduction procedure have been described by Middendrop, et 

al. (1992), and Brown (1994). 

 

The following sections provide a description of the instrumentation used for Statnamic 

testing, and a summary of the most relevant Statnamic test results carried out at the Route 

40 Bridge. The main goal of this discussion is to provide some insight on the relative 

performance of the composite pile compared to a prestressed concrete pile. Detailed 

coverage and discussion of the Statnamic load test results carried out for this project are 

presented in Muchard, et al. (1999). 

 

5.6.1 Instrumentation 

Both the composite and the prestressed concrete test piles were instrumented with strain 

gages in the axial direction at three levels along the length of the piles. At each level, two 

strain gages were placed near the two opposite faces of the pile, as shown in figure 90. 

The strain gages were mounted on 914-mm- (36-inch-) long No. 13M steel bars, 

embedded in the concrete core. After completion of axial load testing, each test pile was 

also instrumented with a string of eight lateral motion sensors to measure the lateral 

displacement profile of the pile during lateral Statnamic load testing. The sensors were 

lowered inside a slope inclinometer casing embedded in the test piles. The locations of 

the lateral motion sensors are shown in figure 90. One embedded accelerometer was 

installed near the pile toe to measure the pile toe movement during Statnamic axial load 

tests. 
119 

Figure 90. Illustration. Test Pile Instrumentation. 

5.6.2 Axial Load Tests 

 

Axial Statnamic load tests were carried out on both test piles seven days after pile 

driving. The axial load setup using the Statnamic testing system is shown in figure 91. 

During the Statnamic test, several measurements are taken, including the applied load, 

the pile head displacement, the toe accelerometer readings, and strains. Pile head 

displacement is measured using a laser beam device shown in figure 91. A detailed 

description of the Statnamic technique can be found in Brown (1994). Each test pile was 

subjected to three loading cycles with an increasing magnitude of applied load. 
120 

Figure 91. Photo. Axial load test using the Statnamic device. 

The equivalent static load versus pile head axial displacement for the prestressed concrete 

and composite piles for the three cycles are shown in figure 92. In both piles, the 

behavior during the last cycle showed that the geotechnical capacity of the piles was fully 

mobilized, as evident from the rapid increase of the pile displacement near the end of the 

test. Davisson’s failure criterion (Davisson 1972) was applied considering the failure 

envelope encompassed by all three load cycles. Based on this criterion, the composite 

pile has an ultimate load capacity of 4,359 kN (980 kips) at a head displacement of about 

18 mm (0.70 inch). The Davisson criterion was not satisfied for the prestressed concrete 

pile. However, the maximum load applied was 4,190 kN (942 kips) at a displacement of 

about 13 mm (0.51 inch). Both piles exhibited similar axial capacities, even though their 

shaft areas and end areas are slightly different. 
121 

Figure 92. Graph. Pile head displacement versus equivalent axial static load. 

Figure 93 shows the equivalent axial static load versus axial strain behavior of both the 

composite and prestressed piles based on the peak load and corresponding strain during 

each of the three load cycles. The axial strain is based on the average strain of the two 

uppermost strain gages (closest to the loading end), since they measured the maximum 

axial strain induced in the pile. Figure 93 shows that the behavior of the two piles was 

relatively similar, with the composite pile being slightly stiffer than the prestressed pile. 

The figure also shows that the design load of the piles, 667 kN (150 kips), is significantly 

lower than the ultimate load. 

Figure 93. Graph. Axial load—axial strain behavior of test piles. 
122 

The variation of the axial strains along the length of both the prestressed and composite 

piles is shown in figure 94. It can be seen that the measured strains along the depth are 

similar in both piles. The figure also shows that the strain level is reduced with increasing 

depth; this is attributed to the gradual transfer of axial load from the pile to the soil 

through skin friction along the pile shaft. 

Figure 94. Illustration and Graph. Variation of axial strain along 

pile length for three Statnamic loads. 

5.6.3 Lateral Load Tests 

 

Following the axial load tests, lateral load tests were conducted on both the composite 

and prestressed piles using the Statnamic Testing System shown in figure 95. The loading 

system is similar to that used for the axial load test; however, the setup was placed 

horizontally in order to produce a lateral load. The thrust against the pile is transferred 

through a calibrated load cell and hemispherical bearing. During lateral Statnamic testing, 

several measurements are taken, including the applied load and the lateral displacement 

of the pile at different depths. The two test piles were subjected to four progressively 

increasing cycles of load, with measured peak forces ranging up to about 290 kN 

(65,250 lbf) for the composite pile and about 350 kN (78,750 lbf) for the prestressed 

concrete pile. The Statnamic device operated on a sled resting upon falsework, which 

supported the device so that the loading was applied at the same elevation for both piles. 

Based on the falsework elevation, the point of lateral load application was approximately 

3 m (9.84 ft) and 2 m (6.56 ft) above the ground surface for the composite and 

prestressed pile, respectively. As noted earlier, the ground surface elevation at the 

prestressed concrete pile is about 1 m (3.28 ft) higher than at the composite pile. 
123 

 

Figure 95. Photos. Lateral Statnamic setup at the Route 40 project. 

The responses of the test piles were measured using multiple displacement transducers 

and subsurface accelerometers. Redundant measurements of displacement at the point of 

loading were made using both types of instruments. For subsurface displacements, eight 

down-hole lateral motion accelerometers were used to track the lateral movement profiles 

of the piles (see figure 90). The accelerometers were placed inside inclinometer casings at 

depths ranging from near the top of the pile to 8.8 m (28.9 ft) below the loading point. 

The time-dependent displacement at the location of each accelerometer was derived by 

double integration of the recorded acceleration versus time record. All of the data were 

monitored using a data acquisition system with a sampling rate of 5,000 samples per 

second for each channel. Data were recorded with a pretrigger of about 0.3 seconds prior 

to firing the Statnamic device and a duration after trigger of 1 to 2 seconds. 

The plots in figure 96(a and b) are the displacement versus time records at the loading 

point for each of the four loads applied to the prestressed concrete pile and the composite 

pile, respectively. It can be seen that the period of oscillation increased with increasing 

displacement amplitude, indicating nonlinearity in the pile response. Compared to the 

prestressed concrete pile, the larger displacement amplitude (at a smaller maximum load) 

and lower frequency (longer period) of the composite pile indicates a less stiff lateral 

response. 

The peak lateral displacements obtained from the string of subsurface lateral motion 

accelerometers provide the displaced shapes shown in figure 97 for each of the four 

cycles of Statnamic load applied to each pile. Both piles appeared to form a hinge at a 

depth of about 5 m (16.4 ft) below the point of loading. For the last two load cycles, the 

deflection profile along the depth of the piles was almost bilinear, with a sudden change 

in slope occurring at a depth of about 4.8 m (16 ft) measured from the loading point. This 

suggests that failure may have been initiated in both piles at the location of the sudden 

change in slope during the last two cycles. 
124 

Figure 96. Graphs. Displacement time histories at the loading point 

for both piles (Pando, et al., 2004). 

Figure 97. Graphs. Peak lateral displacement profiles for both test piles 

at different cycles of Statnamic load. 
125 

A simple, single-degree-of-freedom system model, as described by Brown (unpublished 

paper), can be used to derive the equivalent static pile response from the Statnamic lateral 

test measurements. Interpretation of the Route 40 lateral Statnamic data was originally 

presented by Muchard, et al. (1999). More recently, Pando, et al. (2004), reanalyzed the 

data incorporating the composite pile nonlinearity. A summary of the methodology for 

lateral Statnamic data reduction is presented below. 

The single-degree-of-freedom model includes inertia of the pile rotating about a hinge 

point below the groundline, a viscous damping component, and a nonlinear static spring 

resistance. 

 (9) 

where 

Fstatnamic = measured force on the Statnamic load cell 

Finertia = inertial resistance from effective mass of the foundation 

Fdamping = effective viscous damping resistance 

Fstatic = effective static soil resistance 

The inertial resistance is approximated as that of a cylinder rotating about its base, with a 

diameter equal to that of the pile and a height taken as approximately 5.5 meters (18.0 ft). 

The height is based on the observed depth of about 5 m (16.4 ft) to a hinge, as shown in 

figure 97, plus the 0.5 m (1.6 ft) of pile that extends above the loading point. For a round 

pile of radius (r), height (h), and mass (m), the mass moment of inertia about the base 

(Iy), is 

 (10) 

and for a square pile of width b, Iy is 

 (11) 

The rotational acceleration of such a cylinder in relation to a displacement x at the 

loading point z (approximately 5 m (16.4 ft) above the base of the rotating cylinder in this 

case) would be &

x&/z. Equating expressions for the inertial moment about the base, 

 (12) 

Therefore, 

 (13)

where me = Iy/z

2

. 

The effective mass (me) may be thought of as a lumped mass, which if placed at the 

loading point z, would produce the same inertial reaction as the rotating cylinder. For the 
126 

tests at the Route 40 site, me is calculated to be 1.36 kN-s2

/m (93.3 lbf-s2/ft) for the 

prestressed concrete pile and 1.6 kN-s2

/m (109.8 lbf-s2/ft) for the composite pile. In order 

to include the effect of some accelerating mass from the soil surrounding the piles, this 

effective mass is increased slightly and is estimated as approximately 1.5 and 1.75 kNs

2

/m (102.9 and 120.1 lbf-s2/ft)for the prestressed concrete and composite piles, 

respectively. 

The damping resistance is represented by a viscous damper in which the force Fdamping is 

proportional to the velocity, x& , by a constant, c (which is in units of force-time/length). 

In order to relate this more meaningfully to a system damping parameter, the damping 

constant, D, is expressed as a percentage of the critical damping, cc, by 

 (14)

where k is the static stiffness. Thus, 

 (15) 

The static resistance, Fstatic, is modeled as a function of displacement, x, using a spring 

with stiffness, k, that varies with each load level. The spring stiffness, k, represents the 

effect of the entire lateral soil reaction along the length of the pile. The soil response for 

lateral loading at large strains is known to be highly nonlinear. Pando, et al. (2004), took 

into account the nonlinear stiffness of the soil by using a linear stiffness for each load, but 

decreasing the value of k as the load magnitude increases. 

The model uses equations 9, 13, and 15, and the parameters are me, D, and k. Because me

was evaluated as described above for the test piles, the only unknown parameters are D 

and k. Values of D and k were obtained by fitting the model to the data shown in figure 

96. The parameter values that produced the best fit with the data for the prestressed 

concrete pile model are: an effective mass, me, of 1.5 kN-s2/m (102.9 lbf-s2/ft) an 

effective damping ratio, D, ranging from 25 percent at small loads to 18 percent at the 

largest load; and a stiffness, k, that varies from 2,000 kN/m (137,200 lbf/ft) at small 

displacements (less than 20 mm (0.78 inch)) to 500 kN/m (34,300 lbf/ft) at the maximum 

displacement. The parameter values that produced the best fit for the composite pile 

model are an effective mass (me) of 1.75 kN-s2

/m (120.1 lbf-s2/ft); an effective damping 

ratio (D) ranging from 25 percent at small loads to 18 percent at the largest load; and a 

stiffness (k) that varies from 2,000 kN/m (137,200 lbf/ft) at small displacements (less 

than 20 mm (0.78 inch)) to 360 kN/m (24,696 lbf/ft) at the maximum displacement. The 

lower stiffness of both piles at large amplitudes, and of the composite pile relative to the 

prestressed pile, are indicated by the lower frequencies of motion. 

Using these parameter values, both the static soil resistance, Fstatic, and the total soil 

resistance, Fstatic + Fdamping, can be calculated. The peak values of Fstatic and Fstatic + Fdamping

versus displacement at the loading point for each cycle of Statnamic loading are plotted 

in figure 98. It can be seen that the resistance contributed by damping is not large. The 

figure also demonstrates that there is a pronounced change in the slope of the load versus 

deflection curve for the composite pile at a lateral load of about 50 kN (11,250 lbf), 
127 

indicating that first cracking of the composite pile may have occurred at this load. 

However, the composite pile exhibits a continuing capacity to sustain additional lateral 

load, albeit at substantially larger displacements. The prestressed concrete pile, on the 

other hand, exhibits a change in slope of its load versus deflection curve at a lateral load 

of about 125 kN (28,125 lbf), with comparatively little ability to sustain additional load 

beyond that point. 

 Figure 98. Graph. Calculated static and dynamic (static + damping) 

resistances for both test piles. 

Pando, et al. (2004), performed p-y lateral pile analyses to predict the lateral loaddeflection behavior of the Route 40 test piles. The analyses were carried out using the 

program LPILE (Reese, et al., 2000) and incorporated nonlinearity of both the soil 

resistance, P, and the pile flexural stiffness, EI. 

 

The nonlinear flexural stiffness was obtained from the moment-curvature relationship 

derived for both piles. For the concrete-filled FRP tubes, the moment-curvature 

relationship was estimated using a sectional analysis as described by Mirmiran, et al. 

(2000), and Fam and Rizkalla (2002). This methodology assumes that sections normal to 

the neutral axis remain plane after bending, and that the FRP shell is perfectly bonded to 

the concrete interior (i.e., a linear strain distribution through the entire cross section). The 

method consists of discretizing the pile cross section into a series of strip elements so that 

the normal stresses can be integrated over the cross-sectional area. The discretization 

results in strip elements that include both FRP and concrete components. The stress 

integration is done by assigning the appropriate constitutive model for each material. The 
128 

FRP shell can be modeled accurately as linear elastic within the limiting tensile and 

compressive strains. The concrete can be modeled using relationships of the type shown 

in figure 82. This analysis methodology can be easily programmed and can be used to 

generate moment-curvature and interaction diagrams. This methodology was used by 

Fam (2000) to calculate the moment-curvature response of two concrete-filled FRP tubes 

of the same type as used at the Route 40 Bridge, except that the concrete strength was 

lower. The good agreement between the measured and calculated moment-curvature 

behavior for the two piles tested by Fam (2000) is shown in figure 99. 

Using the procedure described above, the moment-curvature relationships were 

calculated for the composite piles used at the Route 40 Bridge project, and the results are 

shown in figure 99. It can be seen that the response is initially stiff until a moment of 

about 110 kN-m (80,850 lbf-ft) is reached, at which point the first cracking of the 

concrete occurs. Beyond this point, the composite pile can sustain additional moment, 

although the response is much less stiff than initially. Failure is predicted to occur at a 

bending moment of about 630 kN-m (463,050 lbf-ft), when the FRP shell ruptures in 

tension. 

The moment-curvature behavior of the prestressed concrete piles was predicted 

analytically using equilibrium and strain compatibility. A modified Ramberg-Osgood 

function was used to model the stress-strain relationship of the prestressing steel strands. 

A generalized expression, given by Popovics, was used to model the nonlinear stressstrain relationship of concrete (Collins and Mitchell, 1997). The moment-curvature 

response based on the analysis of the prestressed concrete pile is shown in figure 99. The 

moment-curvature calculations were terminated when the compressive strain in the 

concrete reached 0.003. 

It can be seen that the derived static load versus displacement curves from figure 98 have 

a similar nature to the moment-curvature response shown in figure 99, for both the 

prestressed concrete and composite piles. For the composite pile, the sudden change in 

stiffness at a lateral load of 48.9 kN (11 kips) reflects the cracking point. For the 

prestressed pile, the initial stiffness is high due to the prestressing effect, as well as the 

higher elastic modulus of steel. 
129 

Figure 99. Graph. Moment-curvature responses for composite 

and pretressed concrete piles. 

Using the moment-curvature relationships in figure 99 and the soil property values 

proposed in the standard Reese, et al. (1974), p-y formulation, the predicted lateral 

response for the test piles is shown in comparison to the measured response in figure 100. 

It can be seen that the calculated and measured responses for the composite pile are in 

good agreement. For the prestressed concrete pile, the calculated and measured 

stiffnesses are in good agreement initially, but the calculated minimum load is less than 

the measured maximum load. 
130 

Figure 100. Graph. Computed and measured lateral loan-displacement response 

for both test piles (Pando, et al., 2004). 

5.7 COMPOSITE PILES IN THE NEW ROUTE 40 BRIDGE 

Pier No. 2 (see figure 80) of the Route 40 Bridge consists of a reinforced concrete cap 

beam supported by 10 composite piles. The composite production piles are identical to 

the composite test pile in terms of size, laminate structure of the tube, and the methods of 

construction and driving. The installed composite piles of Pier No.2 are shown in figure 

81(a). 

5.7.1 Connection of Piles with Pile Cap Beam 

After driving both the prestressed concrete and composite piles, and prior to casting the 

cap beams, special preparation of the pile heads was necessary to facilitate connecting the 

piles to the cap beams. Eight, 25-mm- (1-inch-) diameter holes were drilled through the 

top flat surface of each pile, using a regular rock drill. The holes were 460 mm (18 

inches) deep, parallel to the longitudinal axis of the piles. In the prestressed concrete 

piles, the holes were equally spaced in a 330-by-330-mm (13-by-13-inch) square pattern. 

In the composite pile, the holes were equally spaced in a 447-mm (17.6-inch) diameter 

circular pattern as shown in figure 101. These arrangements allow for a 76 mm (3-inch) 

concrete cover for each hole. Eight, 1,219-mm (48-inch) long, No. 22M (No. 7) steel 

rebars were inserted in the holes and epoxy resin was used to secure the bars inside the 

concrete. Figure 101(a and b) shows the details of the pile head and the steel dowels in 
131 

both the prestressed concrete and composite piles. The formwork arrangement of the cap 

beams were placed such that the bottom surface of the cap beam is 152 mm (6 inches) 

below the upper surfaces of the piles to allow for embedment of the piles inside the cap 

beam. Figure 102(a) shows elevation views of Pier No. 2 through and in between the 

composite piles. Figure 102(b) shows the No. 22M (No. 7) steel bars, which are 

embedded 762 mm (30 inches) inside the cap beam. Figure 103 shows closeup views of 

Pier No. 2 and the connection between the composite pile and cap beam. A general view 

of the new completed bridge is shown in figure 104. The bridge was opened for traffic in 

the year 2000. 

Figure 101. Illustration. Details of pile head showing the bars used 

to connect the pile to cap beam. 
132 

Figure 102. Ilustration. Connection of composite piles to cap beam at Pier No. 2. 

Figure 103. Photos. Pier No. 2 including the composite piles 

and reinforced concrete cap beam. 
133 

Figure 104. Photo. The new Route 40 Bridge over the Nottoway River in Virginia. 

5.8 SUMMARY AND CONCLUSIONS 

In 2000, VDOT replaced the old Route 40 Bridge (Structure No. 1006) over the 

Nottoway River. New precast composite piles, consisting of concrete-filled GFRP tubes, 

were used to support one of the cast-in-place reinforced concrete cap beams that directly 

support the superstructure. Prior to construction, full-scale field tests were undertaken to 

investigate the feasibility of construction, handling, and drivability, as well as the 

structural performance of the new composite piles in comparison to conventional 

prestressed concrete piles. Based on the testing, analysis, and construction experience, the 

following conclusions can be drawn. 

1. Both the composite and prestressed concrete piles performed similarly during pile 

driving, as demonstrated by the driving records, PDA traces, measured wave speed 

generated by the driving hammer, and measured compressive and tensile stresses 

tests. 

2. Both the composite and prestressed concrete piles performed similarly under the axial 

load tests. Full geotechnical capacity was mobilized in both cases before structural 

failure of the piles. The axial loads at failure were significantly higher than the design 

pile load. 

3. Initially, the composite test pile exhibited a lateral stiffness similar to that of the 

prestressed concrete test pile up to a load of about 40 to 50 kN (9,000 to 11,250 lbf). 

This load level was found to correspond to first cracking of the composite pile. 

Beyond this load, the composite pile exhibited a much lower stiffness than the 
134 

prestressed concrete pile, but the composite pile did demonstrate a continuing ability 

to sustain lateral load with additional relatively large deformations. 

4. The flexural strength of the 625-mm- (24.6-inch-) diameter composite pile with a 

5.85-mm- (0.230-inch-) thick GFRP tube was calculated to be higher than the flexural 

strength of the 508-mm- (20-inch-) square concrete pile prestressed with fourteen, 

12.7-mm (0.5-inch) strands, as shown in figure 99. 

5. Calculations show the composite pile fails in bending by fracture of the GFRP tube 

on the tension side, while the prestressed concrete pile fails by yielding of the strands 

in tension, followed by crushing of the concrete in compression. 

6. The lateral load field tests on both the composite and prestressed piles showed similar 

behavior to that obtained from laboratory flexural test and analysis. 

7. The lateral load-versus-deflection response for the composite pile was predicted with 

good accuracy using conventional procedures typically used for prestressed concrete 

piles. 

8. Similar pile-to-cap beam connections were used for the composite and prestressed 

concrete piles, including eight No. 7 steel dowels embedded 457 mm (18 inches) 

inside the piles from one end and extending 762 mm (30 inches) into the cap beam. 

The piles themselves were embedded 152 mm (6 inches) inside the cap beam. 

9. Concrete-filled FRP tubes were successfully installed as piling for a bridge pier at the 

Route 40 Bridge. To date, no indications of unsatisfactory performance have been 

reported. 
135 

CHAPTER 6. FIELD LOAD TESTS AT THE ROUTE 351 BRIDGE 

6.1 INTRODUCTION 

This chapter summarizes the results and findings of field load tests carried out at the 

Route 351 Bridge project in Hampton, VA. The test pile program described here involved 

one prestressed concrete pile and two different types of composite piles that were 

subjected to axial and lateral load testing. One of the composite piles consisted of an FRP 

tube that was filled with concrete and reinforced with steel bars. The other composite pile 

was formed of polyethylene with steel reinforcing bars. In addition to the test pile 

program, two instrumented piles, consisting of a prestressed concrete pile and an FRP 

pile, have been installed in the bridge to study long-term performance and load-transfer 

behavior. The focus of this chapter is the test pile program, including descriptions of the 

test piles, subsurface conditions, pile driving, dynamic testing, static axial load testing, 

and static lateral load testing. The instrumented production piles and the long-term 

performance study are described in chapter 9. Results from the Route 351 tests also have 

been published by Pando et al. (2003). 

6.2 DESCRIPTION OF THE BRIDGE 

6.2.1 The Original Bridge 

In 2001, VDOT started a bridge replacement project at the Route 351 Bridge crossing of 

the Hampton River in Hampton, VA. A general location map and an aerial view of the 

original bridge are shown in figures 105 and 106, respectively. Wide-angle photographic 

views of the original bridge are shown in figure 107. The original bridge, built in the 

early 1940s, was 228.45 m (749.5 ft) long and had a roadway 10.35 m (34 ft) wide. The 

rated capacity of the bridge was for 133.5-kN (15-ton) trucks. The bridge was composed 

of 16 equal spans supported on steel beams. The superstructure was supported by 15 piers 

and 2 end abutments. The foundations of the piers and abutments consisted of groups of 

timber piles with reinforced concrete pile cap beams. 
136 

Figure 105. Map. Location map of the Route 351 Bridge in Hampton, VA. 

Figure 106. Photo. Aerial view of the Route 351 Bridge in Hampton, VA. 
137 

Figure 107. Photos. Wide-angle views of the original Route 351 Bridge. 

The bridge suffered from excessive deterioration in the superstructure and substructure, 

which led to the decision of replacement. Some signs of deterioration in the original 

bridge can be seen in figure 108. 
138 

Figure 108. Photo. Signs of deterioration of the original Route 351 Bridge. 

6.2.2 The New Bridge 

The new bridge is 228.5 m (749.7 ft) long and 12.8 m (42 ft) wide, and consists of a 13-

span superstructure supported by 12 piers and 2 end abutments, as shown in figure 109. 

The bridge deck is supported on American Association of State Highway and 

Transportation Officials (AASHTO) Type II prestressed concrete beams for most of the 

bridge; however, on the first four spans on the west side of the bridge, the deck is 

supported on steel plate girders. The steel plate girders are used in the portion of the 

bridge underneath the Highway I-64 overpass. As shown in figure 109, this area 

corresponds to Piers 1, 2, and 3. These spans are longer than on the east side of the bridge 

and the orientation of the piers had to be aligned with Highway I-64 due to workspace 

limitations imposed by the I-64 overpass. The bridge piers, except for Pier 2 located 

under highway I-64, consist of an elevated reinforced concrete beam-type pile cap 

supported by a group of seven prestressed concrete piles. The bridge rated capacity is for 

MS18 loading and alternate military loading. Based on the design requirements, VDOT 

specified use of 610-mm (24-inch) prestressed solid square piles for the pier foundations. 

The design axial load for the piles in the piers is 890 kN (200 kips). The new bridge was 

completed and opened to traffic in 2003. 

 

As part of this study, an instrumented FRP composite pile was installed at the center of 

the group of piles supporting Pier 11 (see figure 109), and an instrumented prestressed 

concrete pile was installed in the center of Pier 10. Through monitoring of these piles, the 

long-term performance and durability of the two pile types can be assessed and 

compared. The long-term study is described in chapter 9. 

The following sections provide a brief description of both the conventional prestressed 

piles and the composite piles. 
139 

Figure 109. Illustration. Schematic of the new Route 351 Bridge. 
140 

6.3 DESCRIPTION OF TEST PILES 

The three test piles, whose cross sections are shown in figure 110, are described in the 

following subsections. 

Figure 110. Illustration. Test pile cross section details. 

6.3.1 Prestressed Concrete Test Pile 

The prestressed concrete pile is a 610 mm (24 inch) square pile, 18 m (59.1 ft) long. The 

pile is prestressed with 16, 12.7 mm (0.5 inch)-diameter, 7-wire strands of 1.86 GPa (270 

ksi) yield strength steel. The strands were strained to produce an effective prestress of 5.5 

MPa (0.803 ksi) after losses. The spiral ties consist of No. 16M (No. 5) wire with a pitch 

of 150 mm (6 inches). The 28-day strength of the concrete was specified to be at least 35 

MPa (5,080 psi), and the test results show that the strength of concrete was 55 MPa 

(7,980 psi) at the time of the load tests. 

Figure 111(a) shows the stress-strain diagrams for the concrete and steel components of 

the prestressed concrete pile. The concrete stress-strain diagrams were obtained from 

laboratory tests, and the steel diagram shows the minimum yield strength specified for 

1.86 GPa steel strands. Based on these stress-strain diagrams, the calculated axial 

stiffness, EA (where E is the Young’s modulus and A is the cross-sectional area) of the 

pile is 8.2 x 106

 kN (1.84 x 106 kips), as shown in figure 112. Figure 113 shows the 

moment-curvature and flexural stiffness diagrams calculated using the procedures 

described in Collins and Mitchell (1997). 

6.3.2 FRP Composite Test Pile 

The FRP composite test pile consists of an 18.3-m (60-ft) long FRP tube filled with 

concrete and reinforced with steel bars prior to driving. The FRP tube consists of a tube 

with an outside diameter of 622 mm (24.3 inch) and a wall thickness of 10.7 mm (0.417 
141 

inch). The polymer used is Dow vinyl ester Derakane resin, and the fiber reinforcing 

consists of four plies of stitch-bonded glass fiber fabric. Each ply includes four layers of 

fibers oriented at 0, +45, -45, and 90 degrees with respect to the longitudinal axis of the 

tube. The average ply thickness is 1.93 mm (0.0753 inch). The minimum fiber volume 

fraction is 50 percent. 

Prior to driving, the tube was filled with nonshrink concrete with 19 mm (0.75 inch) 

aggregate and Sika Viscocrete additive to produce a fluid, self-compacting concrete. The 

7-day strength of the concrete was 27.4 MPa (3,970 psi), and the strength at the time of 

driving was about 34 MPa (4,930 psi). The pile is reinforced with 14 No. 25M steel bars, 

grade 420 (No. 8 steel bars, grade 60). The spiral ties consist of No. 10M (No.3) gage 

wire with a pitch of 150 mm (6 inches). 

Figure 111(b) shows the stress-strain diagrams for the FRP, concrete, and steel 

components of the FRP pile. The FRP and concrete stress-strain diagrams were obtained 

from laboratory tests, and the steel diagram shows the minimum yield strength specified 

for Grade 420 steel. Based on these stress-strain diagrams, the calculated axial stiffness, 

EA, of the pile is 7.36 x 106

 kN (1.65 x 106 kips), as shown in figure 112. Figure 113 

shows the moment-curvature and flexural stiffness diagrams calculated using the 

procedures described in Fam 2000. 
142 

Figure 111. Graphs. Test pile material properties. 
143 

Figure 112. Graph. Axial load-axial strain behavior of test piles. 

6.3.3 Polyethylene Composite Test Pile 

The polyethylene composite pile consists of a plastic matrix and a steel reinforcing cage. 

The plastic matrix is a proprietary mixture of recycled postconsumer and/or postindustrial 

polyethylene thermoplastics, with added colorants, UV inhibitors, and antioxidants. The 

final density of the plastic ranges from 7.1 to 9.4 kN/m3

 (45 to 60 pcf). The Young’s 

modulus of the plastic is specified by the manufacturer to be at least 275 MPa (40 ksi). 

The pile is reinforced with 24 No. 25M steel bars, grade 420 (No. 8 steel bars, grade 60). 

The spiral ties, which consist of No. 10M (No. 3) gage wire with a pitch of 230 mm (9 

inches), were welded to the longitudinal bars. The pile is fabricated by forcing molten 

plastic into a mold under a 3.4 MPa (500 psi) pressure at 260 oC (500 oF). The plastic pile 

is 592 mm (23 inches) in diameter and 18.3 m (60 ft) long. A steel pipe with an outside 

diameter of 141 mm (5.5 inches) and a wall thickness of 6.35 mm (0.25 inches) was cast 

in the plastic pile to provide a place for installing an inclinometer casing after the pile 

was cast. The steel pipe was used because of concerns that standard inclinometer casings 

might not survive the high temperature and high pressure of the plastic molding process. 

Figure 111(c) shows the stress-strain diagrams for the polyethylene and steel components 

of the polyethylene pile. The polyethylene stress-strain diagrams were obtained from 

laboratory tests, and the steel diagram shows the minimum yield strength specified for 

Grade 420 steel. Based on these stress-strain diagrams, the calculated axial stiffness, EA, 
144 

of the pile is 3.2 x 106

 kN (720 x 106 poundforce), as shown in figure 112. Figure 113 

shows the moment-curvature and flexural stiffness diagrams calculated using the same 

equilibrium and strain compatibility concepts used for the other two piles (Collins and 

Mitchell 1997, Fam 2000). 
145 

Figure 113. Graphs. Flexural characteristics for the three test piles. 

Curvature (rad/km)

0 5 10 15 20 25 30

Moment (kN-m)

0

200

400

600

800

1000

1200

Plastic pile

Prestressed concrete pile

FRP pile

Moment (kN-m)

0 200 400 600 800 1000 1200

Flexural stiffness, EI, (kN-m2

)

0

100000

200000

300000

400000

Plastic pile

Prestressed concrete pile

FRP pile

a) Moment—curvature 

b) Flexural stiffness as a function of applied moment 

1 kN-m = 737.57 lbf-ft 

1 rad/km = 0.621 rad/mile 

1 kN-m2 = 2421 lbf-ft2 
146 

6.4 SOIL CONDITIONS AT THE TEST SITE 

The test pile program was conducted on a spit of land projecting into the Hampton River 

near the Route 351 crossing of the Hampton River in the Chesapeake Bay area of 

Virginia. The location of the test pile program is shown in figure 114. 

 

Figure 114. Map. Location of test pile site at the Route 351 Bridge. 

6.4.1 Geology 

The test site is located in the area of Chesapeake Bay that falls in the Coastal Plain 

geologic province of Virginia. The Chesapeake Bay is about 10,000 years old, and 

originated during the early Pleistocene epoch when the rising of the Atlantic Ocean sea 

level resulted in flooding and deposition of marine sediments. This flooding and 

deposition alternated with sea level lowering and erosion of older sedimentary deposits 

(Frye 1986, Meadors 1977). The numerous transgressions and regressions of the Atlantic 

Ocean created the present-day stratigraphy in this area (Martin, et al., 1987). 

The east and west banks of the Hampton River consist of Tabb Formation deposits 

underlain by the Yorktown Formation (Johnson 1976, Frye 1986). The Tabb Formation 

deposits generally consist of nearshore marine clayey and silty sands with occasional 

gravels (Johnson 1976). The Yorktown Formation has experienced numerous cycles of 

erosion, desiccation, and deposition. Soils from this formation are generally silty sands 

containing varying amounts of minor shell fragments (Martin, et al., 1987, Johnson 

1976). 
147 

6.4.2 Subsurface Conditions 

The soil stratigraphy at the test site is shown in figure 115. The uppermost layer at the 

test pile site is silty fine sand fill, approximately 1.0 m (3 ft) thick. The fill is underlain by 

loose to medium dense silty fine sand to a depth of 13.0 m (42.5 ft). The silty fine sand is 

underlain by stiff sandy clay to a depth of 15.5 m (51 ft). Beneath the clay, medium dense 

to dense silty and clayey sand was encountered to the bottom of the borings at a depth of 

30.6 m (100 ft). The ground water level, which is controlled by the adjacent Hampton 

River, is at 0.9 to 1.5 m (3 to 5 ft) below the ground surface. 

The test site was investigated with two hollow stem auger borings, four cone 

penetrometer probes, and one dilatometer probe. The locations of the borings and probes 

are shown in relation to the test pile layout in figure 116. The borings were advanced 

using hollow-stem auger techniques, and standard penetration tests (SPT) were 

performed in the borings using a safety hammer. The field SPT N-values are shown in 

figure 115, together with descriptions of the materials encountered in the borings. Figure 

115 also shows the tip resistance and the sleeve friction from the four cone penetrometer 

test (CPT) probes. Additional subsurface investigation information for the pile load test 

area is included in appendix C. 
148 

Figure 115. Charts. Simplified soil stratigraphy near test pile area. 

 

6.4.3 Hampton River 

The water from the Hampton River is salty or brackish. Water samples taken from the 

bridge site were tested at Virginia Tech. Anion tests showed concentrations of chloride, 

bromide, and sulfate of 11,720, 28.3, and 1,498 mg/L (1.56, 0.00376, and 0.199 oz/gal), 

respectively. Cation tests showed concentrations of sodium, potassium, magnesium, and 

calcium of 6,151, 226, 558, and 259 mg/L (0.818, 0.030, 0.0742, and 0.0344 oz/gal), 

respectively. 

The water level of this river is tidal. The mean low and high water level are -0.4 m (-1.3 

ft) and +0.5 m (1.6 ft) with respect to mean sea level, respectively. The extreme high 

water level event for this area was recorded in 1932 at +2.180 m (7 ft) above mean sea 

level. 
149 

Figure 116. Illustration. Pile load test layout. 
150 

6.5 TEST PILE INSTRUMENTATION AND FABRICATION 

6.5.1 Test Pile Instrumentation 

All three test piles were instrumented with 16 sister bar strain gages located at 6 levels. A 

pair of gages was located at each level to provide gages on opposite sides of the pile. Two 

extra gages were installed at the uppermost and lowermost levels for redundancy. Each 

strain gage consisted of a No. 12M, grade 420 steel bar (No. 4 steel bar, grade 60), 0.9 m 

(2.9 ft) long, with four foil resistance strain gages that were precision mounted on each 

bar. The foil gages were mounted and wired to configure a full Wheatstone bridge that 

would cancel out any bending strains induced in the steel bar. For the polyethylene 

plastic pile, the foil gages and lead wires were made of materials capable of withstanding 

the high temperatures involved in manufacturing this pile. The sister bar strain gage 

survival rates, after pile manufacturing, driving, and restrike, were 50 percent, 56 percent, 

and 81 percent for the prestressed concrete, FRP, and plastic piles, respectively. An 

inclinometer casing was also installed in each test pile to permit better resolution of the 

deflected shapes during lateral load testing. 

The as-built instrumentation layout for the prestressed concrete test pile is shown in 

figure 117. 
151 

Figure 117. Illustration. Instrumentation layout for prestressed concrete test pile. 
152 

In addition to the 16 sister bar gages, the FRP test pile had the following instrumentation 

to monitor strains in the FRP tube: 

• Eighteen foil gages bonded to the internal wall of the FRP tube. The foil gages were 

type CEA-13-500UW-350, temperature compensating, underwater, resistance strain 

gages. The gages were configured in a one-quarter active bridge. A protective coating 

system was used to cover these gages to protect them from possible moisture and 

chemical attack during filling of the tubes with concrete. The protective coating 

consisted of a Teflon film, overlain by a layer of polymer compound, overlain by 

aluminum foil, overlain by another layer of polymer compound, overlain by another 

layer of aluminum foil, with the edges sealed with a rubber coating. 

• Fiber optic sensors embedded in the FRP tube. To make it possible to embed the 

sensors, the fiber optic cable and sensors were installed during the manufacturing of 

the FRP tubes. The installation, data collection, and interpretation of the fiber optic 

sensors were carried out by an FHWA contractor. 

The as-built instrumentation layout for the FRP test pile is shown in figure 118. 

The instrumentation of the plastic composite pile was similar to that used for the 

prestressed concrete pile. The as-built instrumentation layout for the plastic composite 

test pile is shown in figure 119. 
153 

Figure 118. Illustration. Instrumentation layout for FRP composite test pile. 
154 

Figure 119. Illustration. Instrumentation layout for plastic composite test pile. 
155 

6.5.2 Fabrication of Prestressed Concrete Test Pile 

On August 14, 2001, the prestressed concrete test pile was fabricated in general 

accordance with VDOT standards. Photos of test pile fabrication are shown in figure 120. 

After fabricating and curing, the piles were shipped to the bridge site. 

Figure 120. Photos. Fabrication of prestressed concrete test pile. 
156 

6.5.3 Fabrication of FRP Composite Test Pile 

In January 2002, two FRP tubes were shipped to a casting yard in Virginia Beach, VA. 

The two FRP tubes were 18.3 m (60 ft) and 21.3 m (70 ft) long for the test pile and 

production pile, respectively. The tubes are shown in figure 121(a). The instrumentation 

of the FRP piles was carried out prior to concrete pouring between January 24 and 

February 2, 2002. Figure 121(b) shows installation of the foil gages to the inside wall of 

the FRP tube. 

Prior to concrete pouring, the steel rebar cage was inserted into the FRP tubes as shown 

in figure 121(c). The piles were then raised with a crane and secured in a vertical position 

with the aid of steel sheet piles and steel bracing as shown in figures 122(a) and (b). The 

bottoms of the FRP tubes were fitted into 76-mm (3-inch) grooves in a concrete slab. The 

bottoms of the FRP tubes were also bolted to the concrete slab by means of L-shaped 

angles, as shown in figure 122(c). An elephant trunk, shown in figure 122(d), was 

lowered into the pile with a hopper located at the top of the pile. The concrete was then 

poured through the hopper using a 2-cubic-yard (1.53-cubic-meter) bucket and a crane. 

Photos of the concrete pour are shown in figure 123. Concrete filling of the FRP 

composite piles took place on February 6, 2002. 

After fabrication, the piles were left in a vertical position for 2 weeks. The piles were 

shipped to the bridge site on February 20, 2002. 
157 

Figure 121. Photos. Fabrication of concrete-filled FRP piles. 
158 

Figure 122. Photos. Setup used for concrete filling of FRP composite piles. 
159 

Figure 123. Photos. Concrete filling of FRP composite piles. 
160 

6.5.4 Fabrication of Plastic Composite Test Pile 

The polyethylene composite test pile was instrumented with special sister bar strain gages 

made of materials capable of withstanding the high temperatures and pressures involved 

in manufacturing of the pile. The instrumentation was installed between August 20 and 

August 22, 2001. 

Photos of test pile fabrication are shown in figures 124 and 125. After fabricating, the 

pile was shipped to the bridge site. 

Figure 124. Photos. Rebar cage of the plastic composite test pile. 
161 

Figure 125. Photos. Manufacturing process for the plastic composite test pile. 
162 

6.6 PILE INSTALLATION AND DYNAMIC TESTING 

6.6.1 Pile Driving 

The test piles were driven at the locations shown in figure 116. Prior to driving, all three 

piles were placed in pits approximately 1.0 m (3 ft) deep, thus removing the man-made 

fill materials. An ICE model 80S single-acting diesel hammer was used to drive the piles. 

The maximum rated energy of the hammer is 108.6 kN-m (80,000 ft-lbs). All piles were 

driven with 230-mm- (9-inch-) thick plywood pile cushions. After driving, the pile 

cushions were about 140 mm (5.5 inches) thick. The prestressed concrete, FRP, and 

plastic piles were driven to depths of 16.76, 17.22, and 17.27 m (55.0, 56.5, and 56.6 ft), 

respectively, below the original ground surface. The pile-driving records are shown in 

figure 126. Pile installation photos for the prestressed concrete, FRP, and plastic piles are 

shown in figures 127 through 129, respectively. 

Figure 126. Graph. Driving records for test piles. 
163 

Figure 127. Photos. Installation of prestressed concrete test pile. 
164 

Figure 128. Photos. Installation of FRP composite test pile. 
165 

Figure 129. Photos. Installation of plastic composite test pile. 
166 

6.6.2 Dynamic Testing 

Dynamic monitoring was performed using a Pile Driving Analyzer (PDA). Monitoring 

was performed during initial driving and during restrike, which occurred 5 days after 

initial driving. Dynamic testing was accomplished by attaching strain transducers and 

accelerometers at about 2 diameters from the top of the piles in pairs spaced 180 degrees 

apart. To attach the transducers to the FRP composite pile, small windows were cut out of 

the FRP shell, which allowed the instrumentation to be directly attached to the concrete 

infill. For the plastic pile, the transducers were bolted to the recycled plastic. 

PDA records for the three types of piles at restrike are shown in figure 130. The PDA 

records show a force peak at about time 0 (measured with respect to the horizontal bars 

that are shown in figure 130, which represent the wave travel time for two pile lengths for 

each pile). A velocity peak can also be seen at about 2L/c (where L is the pile length, and 

c the wave speed). This peak is characteristic of a reflected tensile wave from the toe. The 

prestressed concrete and FRP composite piles show similar dynamic behavior. A large 

separation was observed between the force and velocity traces, beginning at about L/c for 

both of these piles. For the plastic pile, a gradual separation between the force and 

velocity traces starts to occur shortly after 0. Spiro and Pais (2002a and b) encountered 

problems during interpretation of the dynamic data from the plastic test pile. The stressstrain nonlinearity of the plastic (to which the transducers were attached), the strain rate 

dependency of the elastic properties, and the presence of voids in the plastic were all 

considered by the authors as possible reasons for poor quality of the dynamic data for the 

plastic pile. The velocity peak at 2L/c, which is due to tensile wave reflection at the toe, 

was not as pronounced for the plastic pile as for the other two piles. 
167 

Figure 130. Graphs. PDA recordings during restrike. 
168 

Table 30 provides some of the measurements obtained during pile driving. It can be seen 

that the prestressed and FRP piles had similar wave speeds, and the wave speed recorded 

in the plastic pile was about 18 percent lower than the prestressed pile. The maximum 

compressive and tensile stresses were lower than the allowable stresses recommended for 

prestressed piles (Hannigan, et al., 1996). The highest driving stresses were recorded in 

the FRP composite pile; however, no standards are currently available for allowable 

driving stresses for composite piles. 

Table 30. Summary of pile-driving measurements for the prestressed 

and composite piles (after Spiro and Pais 2002a and b).

Pile Type 

Measurement 

Prestressed FRP Plastic 

Wave speed 3,800 m/s 3,782 m/s 3,100 m/s 

Maximum compression stress 

measured during driving 11.0 MPa 16.2 MPa 9.9 MPa 

Maximum tensile stress 

measured during driving 5.6 MPa 8.5 MPa 3.3 MPa 

Allowable stresses Comp. < 24.5 MPa 

Tension < 6.7 MPa 

No standards 

available 

No standards 

available 

1 MPa = 145 lbf/inch2 

Estimates of pile capacity from the dynamic strain and acceleration measurements were 

obtained using the Case method (Spiro and Pais 2002a and b). The strain and acceleration 

data from restrike were subjected to CAPWAP analyses (Spiro and Pais 2002a and b). 

Both the Case method and CAPWAP capacities are listed in table 31. For all three test 

piles, substantial increases in the Case method-derived capacity occurred between the end 

of initial driving and restrike 5 days later. According to the CAPWAP analyses, the 

prestressed concrete pile has the highest capacity at 1,960 kN (440 kips), and the FRP 

pile and the plastic pile have lower capacities equal to 96 percent and 77 percent, 

respectively, of the prestressed concrete pile capacity. The CAPWAP analyses also 

indicate that the shaft capacities, as percentages of the total capacities, are 46 percent, 43 

percent, and 41 percent for the prestressed concrete, FRP, and plastic piles, respectively. 
169 

Table 31. Summary of CASE and CAPWAP analyses results 

(Spiro and Pais 2002a and b). 

Method Capacity 

Component

Prestressed 

Concrete Pile FRP Pile Plastic Pile 

Case (1) (end-of-driving) Total 200 kN 1,600 kN 520 kN 

Case (1) (restrike (2)) Total 1,800 kN 2,300 kN 1,300 kN 

Shaft 894 kN 815 kN 618 kN 

CAPWAP Toe 1,066 kN 1,069 kN 894 kN (3) (restrike (2)) 

Total 1,960 kN 1,884 kN 1,512 kN 

Notes: (1) Case method using a Case damping coefficient of 0.60. 1 kN = 225 lbf 

 (2) Restrike of piles performed 5 days after initial driving. 

 (3) CAPWAP equivalent damping coefficient = 0.35. 

6.7 PILE INTEGRITY TESTING OF TEST PILES 

Pile integrity tests (PIT) were performed on the test piles before and after pile 

installation. These tests consisted of applying a small impact load to the top of the pile 

with a hand-held hammer. The impact from the hammer generates a small downward 

compression wave (low strain) that travels along the pile with a wave speed, c. The PIT 

tests prior to installation were performed by tapping the piles while they were lying 

horizontally on the ground. Figure 131 shows selected photos of PIT tests before and 

after installation. 

For a uniform pile, supported without any resistance to deformation and without any 

imperfections, voids, or damage, the compression wave will travel unchanged to the 

bottom of the pile and will be reflected upward as a tension wave. However, if the 

compression wave encounters a void or damage before reaching the bottom of the pile, a 

tension wave will be generated sooner than the reflection from the pile toe. Damage of 

the pile is assessed based on the velocity curve, which is investigated for wave reflections 

that may indicate change in pile properties due to defects. However, it is important to 

point out that reflections also occur due to soil resistance, particularly for piles with high 

shaft friction. For long piles with high shaft friction, the reflection from the toe may be 

very small. To help improve identification of the reflections, the records are amplified 

exponentially with time. 
170 

Figure 131. Photos. PIT tests on test piles. 
171 

PIT results for the prestressed concrete test pile before and after driving are shown in 

figures 132 and 133, respectively. No discernable damage was noted based on 

comparison of the pre- and postdriving PIT soundings. The soil resistance in the pile after 

installation results in a smaller toe reflection, as shown in figure 133. 

Figure 132. Graph. PIT sounding on the prestressed concrete 

test pile before installation. 

Figure 133. Graph. PIT sounding on the prestressed concrete 

test pile after installation. 
172 

PIT results for the FRP test pile are shown in figures 134 and 135. No significant 

difference between the pre- and postdrive PIT traces was observed for the FRP pile, and 

similar velocity changes were noted at about 17 and 35 ft (5.19 and 10.7 m) from the pile 

top. Toe reflection in the postdrive PIT is almost not discernible, probably due to high 

soil resistance. No discernable damage was noted based on comparison of the pre- and 

postdriving PIT soundings. 

Figure 134. Graph. PIT sounding on the FRP composite test pile before installation. 

Figure 135. Graph. PIT sounding on the FRP composite test pile after installation. 
173 

PIT results for the plastic composite test pile are shown in figures 136 and 137. 

Figure 136. Graph. PIT sounding on the plastic composite 

test pile before installation. 

Figure 137. Graph. PIT sounding on the plastic composite test pile after installation. 

Differences can be seen between the pre- and postdrive PIT traces from the plastic pile. 

Some of the new velocity peaks observed in the postdriving PIT may be due to soil 

resistance. However, these peaks may also be related to damage that may have occurred 

in the pile during pile driving. The observed differences in PIT records are not conclusive 

to indicate whether the pile was damaged during pile installation. 
174 

6.8 FIELD TESTING OF TEST PILES 

The axial and lateral load test results for the three test piles are summarized in the 

following sections. Important date information for the Route 351 test pile program is 

summarized in table 32. 

Table 32. Test dates for the test pile program at the Route 351 Bridge project. 

Pile Type Cross 

Section 

Length

(m) 

Fabrication

Date 

Driving

Date 

Restrike

Date 

Axial Load 

Test Date 

Lateral 

Load Test 

Date 

Prestressed 

concrete 

610 mm 

square 

18 8/14/01 12/5/01 12/10/01 12/17/01 12/17/01

Plastic 592 mm 

circular 18.3 8/23/01 12/5/01 12/10/01 12/21/01 12/21/01

FRP 622 mm 

circular 18.3 2/6/02 2/21/02 2/26/02 3/4/02 3/5/02

 1 mm = 0.039 inches; 1 m = 3.28 ft 

6.8.1 Axial Load Tests 

Static axial load tests were performed on the three test piles in accordance with ASTM 

D1143. The compressive test loads were applied by jacking against a steel frame 

anchored by eight, HP 14x89 reaction piles placed as shown in figure 116. The reaction 

system was designed for a maximum load of 4,448 kN (1,000 kips). The load was applied 

by three calibrated hydraulic jacks connected to a common manifold and pressure gage 

and operated by a single hydraulic pump. The load applied by this jacking system was 

measured using a 4,448 kN (1,000 kips) calibrated load cell. The deflections were 

monitored using dial gages, LVDTs (linear variable deformation transducers), and a wire 

and mirror system, with support provided by an isolated reference beam. Figure 138 

show photos of the axial load test setup. 

The axial load test was carried out by increasing the axial load using constant load 

increments of 89 kN (20 kips). After each load increment, the load was held constant for 

a time interval of 2.5 minutes. This type of loading sequence is called Quick MaintainedLoad Test (Fellenius 1990) and is in general accordance with the Quick Load procedure 

included in ASTM D1143. 

The data from the dial gages and wire and mirror system were collected by an FHWA 

contractor. The data from the LVTDs, sister bar strain gages, foil gages used in the FRP 

pile, and load were acquired using a MEGADAC® data acquisition system. FHWA 

operated the MEGADAC system. 
175 

Figure 138. Photos. Axial load test of prestressed concrete pile. 

Figure 139 shows the measured axial load versus pile head displacement diagrams for all 

three test piles. It can be seen in figure 139 that the prestressed concrete pile and the FRP 

pile exhibit stiffer initial response than the plastic pile. 
176 

Figure 139. Graph. Axial load test results. 

The pile capacities according to Davisson’s criterion (Davisson 1972) are also shown in 

figure 139. The pile failure loads obtained using this criterion and other commonly used 

methods described by Fellenius (1990) are listed in table 33. 

Based on Davisson’s criterion, the prestressed concrete pile exhibited the highest capacity 

at 3,095 kN (696 kips), and the FRP pile and the plastic pile have lower capacities equal 

to 73 percent and 69 percent, respectively, of the prestressed concrete pile capacity. 
177 

Table 33. Comparison of failure loads for the three test piles. 

Failure Criterion (1)

Prestressed Pile 

Failure Load (kN) 

FRP Pile 

Failure Load (kN) 

Plastic Pile 

Failure Load (kN) 

De Beer’s 3,095 2,070 2,040 

Vander Veen’s 3,180 2,600 2,250 

Brinch-Hansen’s 

(90% criterion) 3,080 2,113 2,163 

Chin’s 3,080 2,815 2,615 

Brinch-Hansen’s 

(80% criterion) 3,103 2,576 2,459 

Davisson’s 3,095 2,260 2,130 

Note: (1) Description of failure criteria can be found in Fellenius (1990). 1 kN = 225 lbf 

6.8.1.1 Residual stresses 

The strain gages that were installed in the test piles permitted calculation of the residual 

strains and stresses locked in the pile after pile driving. 

The residual strains were calculated by comparing the strain gage readings before and 

after pile installation. The residual load distributions along the piles were then obtained 

by multiplying the recorded residual strains by the respective pile axial stiffness (EA). 

The residual load distributions are shown in figure 140. 

The average residual stresses acting along the pile shaft can be estimated by considering 

the vertical equilibrium of the different pile segments between the levels of sister bar 

strain gages. Using this procedure, the estimated residual stress distributions for the three 

test piles are shown in figure 141. The residual stresses for the prestressed concrete and 

FRP piles are estimated to range between ±9 kPa (1.305 lbf/inch2

). The estimated 

residual stresses for the plastic pile do not seem very realistic; nevertheless, their values 

ranged between +28 kPa (4.06 lbf/inch2

) and -40 kPa (5.8 lbf/inch2). 
178 

Figure 140. Graph. Distribution of residual loads. 

Figure 141. Graph. Distribution of residual stresses. 
179 

6.8.1.2 Load transfer 

The three test piles have different embedded shaft areas and different toe areas. 

Consequently, it is informative to separate the shaft and end resistances, and to normalize 

these resistances by the shaft and toe areas. The strain gages that were installed in the test 

piles permitted calculations of load transfer from the piles to the ground. These 

calculations were performed with consideration of residual stresses after driving, using 

the procedures described in Fellenius (2002). The resulting load transfer diagrams are 

shown in figure 142. From these diagrams, the shaft and toe components of pile 

capacities can be determined at the Davisson failure loads, and these are listed in table 

34. The shaft capacities, as percentages of the total capacities, are 78 percent, 65 percent, 

and 70 percent for the prestressed concrete, FRP, and plastic piles, respectively. 

Figure 142. Graph. Load distribution for the three test piles 

at the Davisson failure loads. 
180 

Table 34. Load distributions from static load tests at the Davisson failure loads. 

Prestressed Concrete 

Pile FRP Pile Plastic Pile 

Capacity 

Component Capacity 

(kN) 

Days 

after 

EOID* 

Capacity 

(kN) 

Days 

after 

EOID 

Capacity 

(kN) 

Days 

after 

EOID 

Shaft 2,406 1,480 1,487 

Toe 689 780 643 

Total 3,095 

12 

2,260 

11 

2,130 

16 

* EOID = the end of initial driving 1 kN = 225 lbf 

Using the shaft and toe capacities from the static load tests, together with the embedded 

pile shaft areas and the pile end areas, the average unit shaft resistances (fs) and the unit 

toe resistances (qt) were determined for conditions at the Davisson failure loads. The 

resulting values are listed in table 35, which shows that the average unit shaft resistances 

decrease in order, from the prestressed concrete pile to the plastic pile to the FRP pile, 

and that the unit toe resistances increase in the same order from the prestressed concrete 

pile to the plastic pile to the FRP pile. The observed relationship between mobilized 

average shaft stress and toe stress for each of the three piles is shown in figure 143. 

Table 35. Unit shaft and toe resistances from static axial load tests. 

Pile Type Average Unit Shaft Resistance, 

fs (kPa) 

Unit Toe Resistance, 

qt (kPa) 

Prestressed concrete pile 61.8 1,854 

FRP pile 46.9 2,564 

Plastic pile 48.9 2,339 

Note: (1) Resistance values are for Davisson failure loads. 1 kPa = 0.145 lbf/inch2 

There are notable differences between the results of the static load tests and the 

CAPWAP analyses on restrike hammer blows, as shown in table 31. The capacities from 

the static load tests are higher than the capacities from the CAPWAP analyses by 58 

percent, 20 percent, and 41 percent for the prestressed concrete, FRP, and plastic piles, 

respectively. Strength gain during the 6 to 11 days that passed between restrike and the 

static load tests may be due to continued setup during this time period. This apparent 

strength gain versus elapsed time after end of initial driving (EOID) is illustrated in figure 

144 by plotting the capacities determined using the Case method, CAPWAP analyses, 

and static load test results. It is also noted that the shaft capacities from the static load 

tests are much higher percentages of total capacity than are the shaft capacities from the 

CAPWAP analyses. Again, this may be due (at least in part) to changes in capacity 

distribution occurring during setup. It is also recognized that CAPWAP analyses and 

static load tests are different methods of determining pile capacity, and thus may produce 

different total capacities and different capacity distributions. 
181 

 

Figure 143. Graph. Mobilized average unit shaft resistance— 

toe resistance relationships for the three test piles. 

Figure 144. Graph. Apparent strength gain with time in the three test piles. 
182 

6.8.2 Lateral Load Tests 

For each test pile, a static lateral load test was performed on the same day as, or the first 

day following, the static axial load test. The lateral load tests were performed in 

accordance with ASTM D3966, using the same reaction system as used for the axial load 

tests. A calibrated jack and a calibrated load cell were used. Displacements near the pile 

head were monitored with dial gages and LVDTs, with support provided by an isolated 

reference beam. 

Deformed shapes of the piles at several different lateral load magnitudes were obtained 

from standard inclinometer probe surveys along the installed casings. These shapes are 

shown in figure 145, together with the elevations of the original ground surface, the 

bottom of the pits at the test pile locations, and the points of load application. The piles 

exhibit negligible lateral deformation below depths of 6, 3.3, and 3.5 m (19.7, 10.8, and 

11.5 ft) below the bottom of the pits at the test pile locations for the prestressed concrete, 

FRP, and plastic piles, respectively. 

Figure 145. Graphs. Deformed shapes of piles at different lateral loads. 

The measured load is plotted against pile deflection at the original ground surface in 

figure 146, which shows that the prestressed concrete pile and the FRP pile exhibited 

similar load-deflection response and that the plastic pile deformed the most. At a lateral 
183 

load of 200 kN (45 kips), the deflections were approximately 35, 33, and 70 mm (1.37, 

1.29, and 2.73 inches) for the prestressed concrete, FRP, and plastic piles, respectively. 

Figure 146. Graph. Measured lateral deflections at ground surface 

for the three test piles. 

6.9 SUMMARY AND CONCLUSIONS 

A test pile program was conducted to permit direct comparison of the axial and lateral 

load behavior of three test piles: (1) a conventional, 610-mm (23.8-inch) square, 

prestressed concrete pile; (2) a 622-mm (24.3-inch) diameter composite pile formed of an 

FRP shell and filled with concrete and steel reinforcing bars; and (3) a 592-mm 

(23.1-inch) diameter composite pile formed of a polyethylene plastic matrix with steel 

reinforcing bars. Laboratory tests were performed on the pile materials, and the axial and 

bending stiffnesses of the piles were calculated. The piles are all about 18 m (59.0 ft) 

long. The soils at the test site were investigated, and they consist primarily of mediumdense silty sand with a 2.5-m (8.20-ft) thick layer of stiff sandy clay located about 1 to 

3.5 m (3.28 to 11.5 ft) above the pile toes. The field testing program included dynamic 

measurements with CAPWAP analyses, static axial load tests, and static lateral load tests. 

The following conclusions can be drawn from the test program. 
184 

1. The axial stiffness (EA) of the prestressed concrete pile and the FRP pile are similar 

to each other and about 2.5 times the axial stiffness of the plastic pile. 

2. Over a working range of bending moments, the flexural stiffness (EI) decreases in 

order from the prestressed concrete pile to the FRP pile to the plastic pile. 

3. All three piles exhibited substantial setup after installation based on comparisons of 

Case method capacities at the end of initial driving and restrike 5 days later, as well 

as comparisons of CAPWAP analyses of restrike and static axial load tests several 

days after restrike. 

4. Applying Davisson's failure criterion to the static axial load tests, the axial capacities 

were 3,090, 2,260, and 2,130 kN (695,250, 508,500, and 479,250 lbf) for the 

prestressed concrete pile, the FRP pile, and the plastic pile, respectively. 

5. When evaluated at the Davisson failure load, the average unit shaft resistances are 

61.8, 46.9, and 48.9 kPa (8.96, 6.80, and 7.09 lbf/inch2

) for the prestressed concrete 

pile, the FRP pile, and the plastic pile, respectively, and the corresponding unit toe 

resistances are 1,854, 2,564, and 2,339 kPa (268.8, 371.8, and 339.2 lbf/inch2). 

6. The prestressed concrete pile and the FRP pile exhibited similar load-deflection 

response in static lateral load tests. The plastic pile exhibited much larger deflections 

at the same lateral loads. 
 

185 

CHAPTER 7. ANALYSES OF THE AXIAL LOAD TESTS AT THE 

ROUTE 351 BRIDGE 

7.1 INTRODUCTION 

In this chapter, calculations using methods commonly employed in practice are presented 

for the pile axial load capacity, and for the load-deflection behavior, of the three test piles 

at the Route 351 Bridge project. The calculations are compared with the recorded pile 

capacities and the measured load-deflection responses. Background information for the 

methodologies used to perform the analyses is also presented and discussed in this 

chapter. 

7.2 AXIAL PILE CAPACITY 

7.2.1 Methods to Estimate Axial Load Capacity of Driven Piles in Sand 

The axial capacity of driven piles in sand is possibly the area of greatest uncertainty in 

foundation design (Randolph, et al., 1994), and many methods are available for 

predicting pile capacity. Despite the extensive amount of research in this area, the 

appropriateness and accuracy of current design methods are often questioned (Randolph, 

et al., 1994, Olson 2002). In practice it is common to use factors of safety of two, three, 

or more, since measured pile capacities of driven piles have been found to differ from the 

calculated capacities by more than three times (Olson 2002). It is expected that these 

uncertainties will be applicable to driven composite piles. With this in mind, the 

predictions for the composite piles presented in this chapter are expected to show at best 

a level of accuracy similar to the predictions for conventional piles. 

The predictions presented herein are based primarily on design methods commonly used 

by U.S. highway agencies, with the exception of a relatively new method proposed by 

Jardine and others at the Imperial College in England; that method seems to provide a 

rational approach for assessing axial load capacity of driven piles (Jardine and Chow 

1996). 

A sketch of the axially loaded pile problem is shown in figure 147. Generally, most 

methods estimate the ultimate pile capacity based on consideration of vertical equilibrium 

of the pile at the state of incipient bearing capacity failure (i.e., when the settlement of the 

pile increases rapidly with little further increase in load at the pile head). From figure 

147, the ultimate axial capacity of a pile is given by the equation: 

 (16) 

where QT is the estimated ultimate axial capacity, Qs is the ultimate shaft capacity 

resulting from the surrounding soil in side shear, Qb is the total ultimate tip load at the 

base or tip of the pile (total indicates the weight of the displaced volume of soil is 
 

186 

included), and Wp is the weight of the pile. In practice, the term Wp is usually neglected 

since it is considered small compared to the accuracy of the prediction of QT. 

Figure 147. Illustration. Load transfer in an axially loaded pile. 

The load capacity contributed by the shaft side shear (Qs) is calculated by integrating the 

side shear stresses along the pile’s embedded length: 

 (17) 

where fs(z) is the shear stress between the pile and the soil at a depth z, p(z) is the pile 

perimeter at depth z, and L is the embedment depth of the pile. 

The expression used for the tip or end-bearing of the pile is usually of the form: 

 (18) 

where Ab is the tip or end area of the pile, and qb is the total ultimate end-bearing or tip 

stress. 
 

187 

The different pile capacity determination methods deal with ways of estimating the side 

friction (fs) and the ultimate end-bearing stress (qb). The following sections summarize 

the most common methods used in practice to estimate fs and qb for piles driven in sands. 

7.2.1.1 Nordlund’s method 

Nordlund’s method is often used by U.S. highway agencies and is advocated by the 

FHWA (Olson 2002). Nordlund’s method is a semiempirical method based on a database 

of pile load tests (Nordlund 1963, 1979). The method accounts for differences in soil-pile 

interface for different piles. For the sake of brevity, only the most pertinent details of the 

method are presented. A complete description of this method can be found in Hannigan, 

et al. (1996). 

According to the Nordlund’s method, the shaft shear stress (fs) of a uniform cross section 

of pile embedded at a length L can be estimated as follows (Hannigan, et al., 1996): 

 (19) 

where 

Kδ = coefficient of lateral earth pressure 

CF = correction factor for Kδ when δ ≠ φ

σ′v(z) = free-field effective overburden pressure at depth z

δ = interface friction angle between pile and soil 

As pointed out by Olson (2002), the term σ′v(z) in equation 19 should, rigorously 

speaking, correspond to the effective vertical stress at depth z, right next to the pile shaft. 

Because this stress is unknown, it is common to use the free-field vertical stress that 

corresponds to the in situ value far away from the influence of the pile. 

The expression proposed by Nordlund for the ultimate end-bearing stress (qb) is as 

follows: 

 (20) 

where 

αT = dimensionless factor dependent on pile depth-width ratio 

N′q = bearing capacity factor (function of φ′) 

σ′v(z = L) = free-field effective overburden pressure at the pile tip (z = L) 

Values for the coefficients Kδ ,CF , N'q, and αT, and for the interface friction angle δ, are 

typically read from design charts found in the technical literature, such as the FHWA 

manual for driven piles (Hannigan, et al., 1996). 

 
 

188 

7.2.1.2 API method 

The American Petroleum Institute (API) provides design recommendations for axially 

loaded piles in the API RP 2A publication titled Recommended Practice for Planning, 

Designing, and Constructing Fixed Offshore Platforms (API 1993). Although this 

method is not as commonly used by highway agencies as is Nordlund’s method, it is 

worth presenting here because the API recommendations are based on a large database of 

axial pile load tests that is continually evaluated and updated (Pelletier, et al., 1993). 

According to API (1993), the shaft shear for piles in sands can be estimated using: 

 (21) 

where 

σ′v(z) = free-field effective overburden pressure at depth z

K = earth pressure coefficient 

δ = interface friction angle between pile and soil (as per table 36) 

API (1993) recommends using a value of 1.0 for the earth pressure coefficient (K) for full 

displacement piles, such as the piles tested in this project. Equation 21 implies that the 

shaft friction can increase indefinitely with depth (i.e., with increasing effective 

overburden pressure). However, API recommends using limiting values for fs, as shown 

in table 36. In the absence of interface friction angle test data, API recommends using the 

values of soil-pile interface friction angles (δ) listed in table 36. 
 

189 

Table 36. API recommendations for side friction in siliceous soil (API 1993). 

Limiting Shaft Friction, 

fs Density Soil Description 

Soil-Pile Interface 

Friction Angle, δ 2

(degrees) (kips/ft2

) (kPa) 

Very loose Sand 

Loose Sand-silt 3

Medium Silt 

15 1.0 47.8 

Loose Sand 

Medium Sand-silt 

Dense Silt 

20 1.4 67.0 

Medium Sand 

Dense Sand-silt 

25 1.7 81.3 

Dense Sand 

Very dense Sand-silt 

30 2.0 95.7 

Dense Sand 

Very dense Sand-silt 

35 2.4 114.8 

Notes: (1) These values provided by API are intended as guidelines only, and where detailed information is available 

other values may be justified. 

(2) Suggested values are for steel pipe piles driven in cohesionless soils. 

(3) Sand-silt is described as soils with significant fraction of both sand and silt. 

API recommends using the following expression to estimate the ultimate end-bearing or 

tip resistance for piles driven in sands: 

 (22) 

where 

σ′v(z = L) = free-field effective overburden pressure at the pile tip (z = L) 

Nq = bearing capacity factor (as per table 37) 

Similarly to fs, API recommends limiting the value of the tip resistance (qb). The 

recommended values of Nq and limiting qb are shown in table 37. 
 

190 

Table 37. API recommendations for tip resistance in siliceous soil (API 1993). 

Limiting Tip Resistance, qb

Density Soil Description Nq

(kips/ft2

) (MPa) 

Very loose Sand 

Loose Sand-silt 

Medium Silt 

8 40 1.9 

Loose Sand 

Medium Sand-silt 

Dense Silt 

12 60 2.9 

Medium Sand 

Dense Sand-silt 

20 100 4.8 

Dense Sand 

Very dense Sand-silt 

40 200 9.6 

Dense Sand 

Very dense Sand-silt 

50 250 12.0 

7.2.1.3 LCPC method 

The Laboratoire Central des Ponts et Chaussees (LCPC) method estimates pile capacities 

based on CPT tip resistance (qc) values (Bustamante and Gianeselli 1982). A detailed 

description of the method can be found in Lunne, et al., 1997. 

The unit shaft resistance (fs) is determined based on the pile type, installation method, 

and the value of the CPT tip resistance (qc). The recommended pile shaft resistance is 

obtained by dividing the measured qc by a friction coefficient (αLCPC). A limiting shaft 

friction is recommended based on pile and soil type. The values of recommended friction 

coefficients, αLCPC, and maximum limit of fs for driven piles in sands are summarized in 

table 38. 
 

191 

Table 38. LCPC friction coefficient αLCPC for driven piles in sands 

(Bustamante and Gianeselli 1982). 

Pile Category 

Coefficient, αLCPC

Maximum Limit of fs

(MPa) State of Sand qc 

(MPa) 

Concrete Steel Concrete Steel 

Loose ≤ 5 60 120 0.035 0.035 

Medium dense 5 to 12 100 200 0.08 0.08 

Dense to very dense > 12 150 200 0.12 0.12 

1 MPa = 145 lbf/inch2 

The LCPC method correlates the pile unit tip resistance (qb) to an equivalent average 

cone resistance (qca) calculated using the qc values within 1.5 pile diameters (or widths) 

below and above the pile tip elevation, as described by Lunne, et al. (1997). The LCPC 

pile tip resistance (qb) is calculated by multiplying the equivalent average cone resistance 

(qca) by an end bearing coefficient (kc). The recommended values for end-bearing factors 

for driven piles in sands are summarized in table 39. 

Table 39. LCPC bearing capacity factors for driven piles in sands 

(Bustamante and Gianeselli 1982).

State of Sand qc (MPa) Factor kc

Loose ≤ 5 0.5 

Medium dense 5 to 12 0.5 

Dense to very dense > 12 0.4 

1 MPa = 145 lbf/inch2 

7.2.1.4 Imperial College method 

This method was developed at the Imperial College in London, England, by Jardine and 

his colleagues based on a series of investigations using instrumented field test piles 

(Jardine 1985, Bond 1989, Lehane 1992, Chow 1996). The instrumented piles were 

tubular closed and open-ended steel piles driven in sands for supporting offshore 

structures used in the oil industry. 

The major findings of the Imperial College experiments, related to piles driven in sands, 

are summarized below (Jardine and Lehane 1993, Jardine and Chow 1996): 

• The peak local shaft friction (fs) is related to the local radial effective stress at failure 

(σ′rf) by the following simple Mohr-Coulomb effective stress criterion: 

 (23) 
 

192 

where 

σ′rf = the local radial effective stress at failure 

δf = δcv = failure or constant volume interface friction angle 

• The radial effective stresses developed around the pile shaft after installation (σ′rc) 

were found to depend strongly on the initial relative density of the sand (Dr) and the 

distance to the pile tip from the point where the radial stress is being evaluated. The 

following empirical expression, based on statistical analyses of over 60 datasets of 

radial stress, is recommended to estimate the local radial effective stress (σ′rc) of 

close-ended piles: 

 (24) 

where 

σ′rc(z) = local radial effective stress after installation 

qc = CPT tip resistance 

z = depth below ground surface 

L = pile embedded length 

h = L-z = distance to pile tip 

σ′v(z) = free-field effective overburden pressure at depth z 

Pa = atmospheric pressure (= 100 kPa) 

R = pile radius 

h/R = 8 for points close to the tip where h < 8R 

• Jardine and Lehane (1993) found that the radial stresses could rise considerably (by 

about 50 percent) due to localized dilation at the pile-soil interface during axial 

compression loading of the pile. The radial stresses were found to reach a maximum 

near the mobilization of the peak shear stresses. The increase in radial effective 

stresses was found to vary in inverse proportion to the pile radius, and was found to 

be predicted reasonably well using the simple elastic cylindrical cavity expansion 

theory proposed by Boulon and Foray (1986). 

• Constant volume or residual interface friction angles (δcv) were found to control the 

developed peak local shear stresses in pile tests. The values determined from 

laboratory interface shear tests were found to correlate well with their field test 

measurements. The interface friction angle was found to depend on sand grain size, 

shape, and mineral type. The pile material hardness and surface roughness were also 

found to influence δ. Jardine and coworkers pointed out that the level of radial 

effective stress may also influence δ, but they indicated that more research is needed 

to investigate this further. 
 

193 

The Imperial College (IC) method provides a simple, practical, and rational approach to 

predict pile capacity of piles driven in sands. A complete description of the IC method 

can be found in Jardine and Chow (1996), and a brief description is presented here. 

The basic steps involved in the IC method to estimate the shaft capacity of closed-ended 

driven piles in sands are: 

1. Calculate the local radial effective stress around the pile after pile installation, 

σ′rc using equation 24. 

2. Calculate the increase in radial effective stresses (∆σ′rd) due to dilation during 

pile loading. Jardine and Chow (1996) recommend using the following 

expression based on cylindrical cavity expansion theory: 

 (25) 

where 

G = operational shear modulus of sand (see note below) 

δh = average radial displacement (see note below) 

R = pile radius 

Notes: 

G: Since the shear stiffness of sands is nonlinear, pressure dependent, and 

anisotropic, the method recommends using reliable and appropriate 

measurements of the shear modulus. If these measurements are not 

available, the following expression (after Chow 1996) is recommended: 

 (26) 

where 

qc = CPT tip resistance 

Α = 0.0203 

B = 0.00125 

C = 1.216 x 10-6

σ′v = free-field effective overburden pressure 

Pa = atmospheric pressure (100 kPa) 

δh: The term δh/R, in equation 25, comes from cylindrical cavity 

expansion theory, and corresponds to the cavity strain. The term δh refers 

to radial sand dilation, and can be estimated as being equal to the average 
 

194 

peak-to-trough height obtained from surface roughness measurements of 

the pile (Chow 1996). The average peak-to-trough height (Ri) is equal to 

twice the centerline average roughness parameter (Rcla). Rcla is a surface 

roughness parameter commonly used by material scientists, and it is 

defined as the average distance of peaks and troughs to the surface profile 

centerline taken over an 8-mm (0.312-inch) gage length. 

 

3. Select the interface friction angle (δ) based on laboratory interface shear tests. 

Alternatively, δ can be estimated using published correlations, preferably if 

developed considering pile material characteristics (roughness and hardness) 

and sand characteristics (shape, D50, mineralogy). 

Calculate the local radial effective stress at failure, as follows: 

 (27) 

where 

σ′rc from step 1 using equation 24 

Δσ′rd from step 2 using equation 25 

4. Calculate the local shear stresses using equation 23. 

5. Determine the shaft capacity by integrating the local shear stresses (obtained 

in step 4) along the embedded shaft length. 

The IC method defines the tip resistance as the total utilizable tip resistance at a pile head 

displacement equal to 10 percent of the pile diameter. The recommended pile base 

resistance for closed-ended, driven piles in sand is related to the CPT tip resistance (qc,) 

averaged near the pile tip, according to the following empirical equation: 

 (28) 

where 

q bar subscript c = the CPT tip resistance averaged over 1.5 pile diameters above and 

below the pile tip 

D = the pile diameter 

DCPT = 0.036 m = the CPT diameter 

The equation shown above is based on IC pile load test data that showed that the 

measured profiles of pile tip resistance (qb) fell close to the CPT qc traces. The ratio qb/qc

was found to be less than unity and to decrease with pile diameter (Jardine and Chow 
 

195 

1996). Equation 28 tends to zero at D = 3.6 m (11.8 ft); therefore, a lower bound value of 

b c q 0.13 q = ⋅ is recommended for D > 2 m (6.56 ft). 

7.2.2 Predicted Axial Capacities 

Predictions for the three piles tested at the Route 351 Bridge project were made using the 

four methods described in section 7.2.1. The predictions presented in the following 

sections were based on the geotechnical investigation results presented in chapter 6 and 

the interpreted average CPT and SPT design profiles, shown in figure 148. The 

predictions for each pile type included the specific information for each pile capacity 

prediction method. Pile-specific information included cross-sectional shape and 

dimensions, embedment depths, surface roughness, and interface friction values. 

Figure 148. Graphs. Interpreted average CPT and SPT design profiles 

for Route 351 test site. 

7.2.2.1 Nordlund’s method predictions 

Two sets of axial capacity predictions were made using Nordlund’s method. One set of 

predictions was carried out using the recommended values of δ read directly from the 

Nordlund’s design charts. The curve for concrete was used for the prestressed concrete 
 

196 

pile, and the curve for steel was used for the composite piles. The second set of Nordlund 

method predictions used the constant volume interface friction angles (δcv) obtained in 

this research, as presented in chapter 3. The two sets of axial pile capacity predictions 

using Nordlund’s method were compared to the measured pile capacities evaluated from 

the load test results using the Davisson’s failure criterion; these predictions are depicted 

in figures 149 and 150. 

 Figure 149. Graph. Accuracy of Nordlund’s method predictions 

using δ values from Nordlund’s charts. 
 

197 

Figure 150. Graph. Accuracy of Nordlund’s method predictions 

using δ values from interface shear tests. 

Figures 149 and 150 show that the tip capacities (Qb) were underpredicted by about 35 

percent for the composite piles and 15 percent for the prestressed concrete pile. The shaft 

capacities were overpredicted in both sets of predictions. The shaft capacities predicted 

using Nordlund’s recommended interface friction angles, using the steel curve for the 

composite piles, were 35 to 43 percent higher than those measured for the composite 

piles and 23 percent higher for the prestressed concrete pile. The shaft capacities 

calculated using the δ values from interface shear tests were about 65 percent higher for 

the composite piles and 20 percent higher for the prestressed concrete pile. The average 

coefficient of lateral earth pressure (Kδ) from Nordlund’s method was about 1.67. This 

value of Kδ may be over predicting the effective radial normal stresses acting at the pilesoil interface, thereby producing higher shaft capacities than measured. To match the 

shaft capacities from the pile load tests, the coefficient of lateral earth pressure (Kδ) 

would have to be around 1.0 for the composite piles and about 1.4 for the prestressed 

concrete pile, when using the δ values from the interface shear tests. 

7.2.2.2 API method predictions 

Two sets of axial capacity predictions were made using the API (1993) method. One set 

of predictions was carried out using API’s recommended value of δ from table 36. The 

second set of API method predictions used the constant volume interface friction angles 
 

198 

(δcv) from the interface shear tests presented in chapter 3. Comparisons of the two API 

predictions and the measured pile capacities are shown in figures 151 and 152. 

Figure 151. Graph. Accuracy of API method predictions 

using δ values from table 36. 
 

199 

Figure 152. Graph. Accuracy of API method predictions 

using δ values from interface shear tests. 

The tip capacities (Qb) predicted for the three piles, using the API method, were within 

±15 percent of the measured values based on the Davisson failure criterion. The shaft 

capacities estimated using the API recommended interface friction angles, which are for 

steel pipe piles, were about 23 percent lower than measured for the composite piles and 

43 percent lower for the prestressed concrete pile. The shaft capacities estimated using 

the δ values from the interface shear tests were between 0.3 and 6 percent higher for the 

composite piles and 23.5 percent lower for the prestressed concrete pile. The average 

coefficient of lateral earth pressure (K) from the API method is 1.0. This value of K, 

coupled with the δ values from the interface shear tests, resulted in very reasonable 

predictions of the pile shaft capacities for the composite piles. 
 

200 

7.2.2.3 LCPC method predictions 

Axial capacity predictions using the LCPC method (Bustamante and Gianeselli 1982) 

were based on the average obtained from applying the method to the CPT tip resistance 

(qc) data available at the test site, except that probe CPT-2 was excluded because it 

produced anomalous data in the depth range of 3 to 8 m (9.84 to 26.2 ft). For each pile 

type, two LCPC capacity calculations were performed: one assumes the particular pile 

under consideration is a steel pile, and the other assumes the pile is a concrete pile. 

Comparisons of the measured pile capacities and the LCPC predictions using the steel 

and concrete pile assumptions are shown in figures 153 and 154, respectively. 

Figure 153. Graph. Accuracy of LCPC method predictions 

using steel pile assumption. 
 

201 

Figure 154. Graph. Accuracy of LCPC method predictions 

using concrete pile assumption. 

The tip capacities (Qb) estimated using the LCPC method over predicted the test results 

by 25 and 38 percent for the composite piles and by 71 percent for the prestressed 

concrete pile. The shaft capacities estimated using the LCPC steel pile approach were 24 

and 28 percent lower than those measured for the composite piles and 43 percent lower 

for the prestressed concrete pile. The shaft capacities estimated using LCPC concrete pile 

approach were 19 and 25 percent higher than those measured for the composite piles and 

6 percent lower for the prestressed concrete pile. 

 

7.2.2.4 Imperial College method predictions 

The IC method (Jardine and Chow 1996) was used to predict the axial capacities of the 

three test piles using the average interpreted design CPT tip resistance (qc) profile shown 

in figure 148, and the interface friction angle and surface roughness data presented in 

chapter 3. Comparisons of the measured pile capacities and the IC method predictions are 

shown in figure 155. 
 

202 

Figure 155. Graph. Accuracy of IC method predictions 

using δ values from interface shear tests. 

The predicted tip capacities (Qb) using the IC method were 8 percent below those 

measured for the FRP composite pile, 5 percent above those measured for the plastic 

composite pile, and 5 percent above those measured for the prestressed concrete pile. The 

shaft capacities estimated using the IC approach were 10 and 12 percent higher than those 

measured for the composite piles, and 26 percent lower for the prestressed concrete pile. 

The level of agreement between measured and calculated capacities is quite good for all 

cases except the total and shaft capacities of the prestressed concrete pile. 

7.2.3 Summary of Axial Pile Capacity Predictions 

The predicted axial pile capacities presented in the preceding sections are summarized in 

table 40. 
 

203 

Table 40. Predicted axial capacities for the test piles at Route 351. 

Total Capacity Shaft Capacity Toe Capacity Pile Prediction Method 

Qc (kN) Qc/Qm Qc (kN) Qc/Qm Qc (kN) Qc/Qm

Nordlund (w/ IST δ's) 3,480 1.12 2,890 1.2 590 0.85 

Nordlund (Concrete) 3,530 1.14 2,950 1.23 590 0.85 

API (w/ recommended δ's) 2,180 0.7 1,380 0.57 800 1.16 

API (w/ IST δ's) 2,640 0.85 1,840 0.76 800 1.16 

LCPC—Steel 2,560 0.83 1,380 0.57 1180 1.71 

LCPC—Concrete 3,440 1.11 2,260 0.94 1180 1.71 

PSC 

Imperial College 2,500 0.81 1,770 0.74 720 1.05 

Nordlund (w/ IST δ's) 2,920 1.29 2,440 1.65 480 0.62 

Nordlund (Steel) 2,600 1.15 2,120 1.43 480 0.62 

API (w/ recommended δ's) 1,830 0.81 1,160 0.79 670 0.86 

API (w/ IST δ's) 2,240 0.99 1,570 1.06 670 0.86 

LCPC—Steel 2,090 0.93 1,120 0.76 970 1.25 

LCPC—Concrete 2,820 1.25 1,850 1.25 970 1.25 

FRP 

Imperial College 2,340 1.04 1,630 1.1 710 0.92 

Nordlund (w/ IST δ's) 2,830 1.33 2,400 1.61 430 0.68 

Nordlund (Steel) 2,450 1.15 2,010 1.35 430 0.68 

API (w/ recommended δ's) 1,720 0.81 1,120 0.75 610 0.94 

API (w/ IST δ's) 2,100 0.99 1,490 1 610 0.94 

LCPC—Steel 1,960 0.92 1,070 0.72 890 1.38 

LCPC—Concrete 2,660 1.25 1,770 1.19 890 1.38 

Plastic 

Imperial College 2,340 1.1 1,670 1.12 670 1.05 

Notes: IST δ's refers to interface friction angles from interface shear tests. Qc = calculated value (shown with three 

significant digits), Qm = measured value corresponding to Davisson’s failure criterion. Qc/Qm = the ratio of calculated 

to measured. Conversion factor: 1 kN = 225 lbf. 

The various methods used to predict axial pile capacity led to ratios of calculated-tomeasured pile capacities (Qc/Qm) ranging from 0.70 to 1.14 for the prestressed concrete 

pile, and from 0.81 to 1.33 for the composite piles. If only the predictions from the IC 

method and the API method using the δ-values from the interface shear tests are 

considered, the Qc/Qm ratios ranged from 0.81 to 0.85 for the prestressed concrete pile 

and from 0.99 to 1.10 for the composite piles. In other words, these two methods 

underpredict the capacity of the prestressed concrete pile by about 20 percent, and 

overpredict the capacities for the composite piles by about 10 percent. In routine deepfoundation design practice, factors of safety as low as 1.5 are sometimes used, but 

typically they are 2.0 or higher. Even a lower bound factor of safety of 1.5 would ensure 

that the ultimate geotechnical capacity of the composite piles is not exceeded, but it may 

not be high enough to ensure acceptable deformations of the foundation system. 
 

204 

In summary, it can be seen from the prediction results that the level of accuracy of the 

predictions is comparable for all three types of piles. In terms of axial pile capacity 

predictions, the results presented in this chapter seem to indicate that conventional static 

analysis methods are applicable to composite piles. However, additional case histories 

would be needed to corroborate and extend this conclusion to other composite pile types 

and to different soil conditions. 

 

7.3 LOAD-SETTLEMENT BEHAVIOR, AXIALLY LOADED SINGLE PILES 

7.3.1 Introduction 

The settlement of a single pile in soil under compressive axial loading is a function of 

several factors, such as: 

• Soil stratigraphy and characteristics (stiffness, density, compressibility, nonlinearity). 

• Pile shape and dimensions. 

• Pile stiffness. 

• Pile-soil interaction characteristics. 

Several methods are reported in the literature for computation of load-settlement behavior 

of a single pile under axial compression loading. The methods are typically divided in the 

following main categories: 

• Elastic methods. 

• Load-transfer (T-Z) methods. 

• Modified hyperbolic methods. 

• Numerical techniques (including finite element and finite difference methods). 

A good overview of the different techniques available can be found in Poulos (1989). For 

this study we focus mainly on the load-transfer (T-Z) methodology, which is commonly 

used by FHWA and U.S. State transportation agencies. This methodology has recently 

become more accessible to U.S. highway agencies with the release of the FB-Pier 

program, which uses this technique for settlement calculations of single and pile groups 

(Hoit, et al., 2000). The development of this program was sponsored by FHWA and is 

distributed freely to U.S. transportation agencies. Other commonly used programs that 

employ the T-Z methodology to analyze settlement of axially loaded single piles include 

A-Pile (Ensoft 1998) and CAXPILE (Dawkins 1982). 

A brief overview of the load-transfer T-Z methodology is provided in the following 

section. 
 

205 

7.3.2 Background, Load-Transfer Method for Pile Settlement Predictions 

The load-transfer (T-Z) method is probably the most widely used technique to study the 

problem of single axially loaded piles, and is particularly useful when the soil behavior is 

clearly nonlinear and/or when the soil surrounding the pile is stratified. This method 

involves modeling the pile as a series of elements supported by discrete nonlinear 

springs, which represent the resistance of the soil in skin friction (T-Z springs), and a 

nonlinear spring at the pile tip representing the end-bearing (Qb-Z) spring. The soil 

springs are nonlinear representations of the soil reaction, T (or Qb for the pile tip), versus 

displacement (Z) as shown schematically in figure 156. Assuming the T-Z and Qb-Z 

curves are available, the axial load-settlement response can be obtained with the aid of a 

computer program such as FB-Pier. 

Figure 156. Illustration. Idealized model used in T-Z load-transfer analyses. 

Adequate T-Z and Qb-Z curves are essential in this method to obtain reliable settlement 

and load transfer calculations for axially loaded single piles. Such load-transfer curves 

were originally obtained empirically; one example is the work of Coyle and Sulaiman 

(1967), who published empirical T-Z curves based on model and full-scale pile load tests 

in sand. Building on this work, and based on additional empirical results, general 

recommendations for estimating T-Z and Qb-Z curves for driven piles in sands have been 

proposed by Vijayvergiya (1977) and API (1993). Load-transfer curves can also be 

constructed satisfactorily using a theoretical approach related to the shear stiffness of the 

soil surrounding the pile (Poulos 2001). Several methodologies to develop theoretically 
 

206 

based load transfer curves have been proposed (e.g., Kraft, et al., 1981, Chow (1986), 

McVay, et al., 1989, and Randolph 1994). 

 

Section 7.3.3 presents settlement predictions for the test piles using empirical loadtransfer curves (API 1993, Vijayvergiya 1977), while section 7.3.4 presents prediction 

results using new theoretical load-transfer curves. 

7.3.3 Predictions Using Empirical Load-Transfer Curves 

7.3.3.1 Predictions using load-transfer curves recommended by API (1993) 

7.3.3.1.1 Load-transfer curves recommended by API (1993) 

In the absence of more precise information, API (1993) recommends the use of the 

following T-Z curve expression for noncarbonate sands: 

 

 (29) 

where 

fs = the unit side shear stress mobilized along a pile segment at movement, z 

fs, max = the maximum unit side shear stress 

z = movement of pile segment 

zc = movement required to mobilize fs, max (0.1 inches for sands) 

In the absence of more definitive criteria, API (1993) recommends using the pile tip load 

versus tip settlement curve shown in figure 157. 

Figure 157. Graph. Pile tip load—pile tip displacement curve (Q-Z) (API 1993). 
 

207 

In figure 157, Q is the mobilized end-bearing capacity at an axial tip displacement z, and 

Qb is the ultimate end-bearing load (computed using qb from equation 22 times the pile 

tip area). As shown in figure 157, Qb is mobilized at a pile tip displacement equal to 

10 percent of the pile diameter, D. 

7.3.3.1.2 Maximum shaft shear stress distributions recommended by API (1993) 

As described previously in section 7.2.1.2, API recommends using maximum shear 

stresses for the pile shaft that increase linearly with depth until they reach a prescribed 

limiting value. Graphical representations of the maximum shaft shear stresses for the 

three test piles, as recommended by API (1993), are shown in figure 158. In section 7.2.3 

it was indicated that better pile capacity estimates were obtained using the API method 

with δ values from interface shear tests. Despite this finding, the shear stresses shown in 

figure 158 correspond to the δ values recommended by API (1993). This was done to be 

able to assess the pile settlement predictions using the original method without major 

modifications. 

 

Figure 158. Graph. Maximum shear stress distribution 

along pile shaft, according to API (1993). 

7.3.3.1.3 Prediction for the prestressed concrete test pile 

The load-settlement responses predicted for the concrete test pile, using the API (1993) 

T-Z and Qb-Z curves, are shown in figure 159. The curve showing a lower predicted 

capacity is based on the API load-transfer curves using a limiting shear stress. The 

second prediction was made using the API curves but without a limiting shaft shear 

stress. 

Figure 159 shows that the predicted settlement curves have relatively good agreement in 

the initial portion of the load settlement curve. However, they both underpredict the 
 

208 

measured pile capacity. This is consistent with the axial pile capacity predictions 

presented in section 7.2.2.2, where the predicted API pile capacity (using the API 

recommended δ values) was shown to be about 30 percent lower than measured. The 

predicted load-settlement curve using no limiting fs shows a slightly better agreement 

with the field measurements, but also underpredicts the pile capacity by about 18 percent. 

The extent of this underprediction is due to the low shaft capacity estimated using the 

API (1993) method, which (as indicated in section 7.2.2.2), was 43 percent lower than 

measured. 

Figure 159. Graph. Settlement predictions for the prestressed 

concrete pile using API (1993). 

7.3.3.1.4 Prediction for the FRP test pile 

The predicted load-settlement curves for the FRP test pile, using the API (1993) loadtransfer curves with and without a limiting shaft shear stress, are shown in figure 160. 

Figure 160 shows a good agreement for the initial portions of the predicted settlement 

curves and the measured load-settlement curve. However, similar to the prestressed 

concrete pile predictions, the API prediction using an fs cutoff underpredicts the pile 

capacity, and the onset of large settlements occurs at lower load than observed in the field 

measurements. The prediction with no fs cutoff shows a much better correspondence with 

the field measurements, but also shows the onset of large settlements occurring at a load 

slightly below that measured in the field. 
 

209 

Figure 160. Graph. Settlement predictions for the FRP pile using API (1993). 

7.3.3.1.5 Prediction for plastic test pile 

The predicted load-settlement curves for the plastic test pile, using the API (1993) loadtransfer curves with and without a limiting shaft shear stress, are shown in figure 161. 

Figure 161 shows that the predicted load-settlement curve, using the API (1993) method 

with a limiting fs, agrees well with the initial portion of the field measurements, but large 

settlements start to occur sooner than observed at about 1,700 kN (382,000 lbf). On the 

other hand, the API prediction using no fs cutoff shows very good agreement in both the 

initial and final portions of the settlement curve from the pile load test. 
 

210 

Figure 161. Graph. Settlement predictions for the plastic pile using API (1993). 

7.3.3.2 Predictions using Vijayvergiya (1977) load-transfer curves 

7.3.3.2.1 Load-transfer curves recommended by Vijayvergiya (1977) 

Vijayvergiya (1977) recommends the following load-transfer expression of mobilized 

unit side shear (fs) as a function of shaft movement (Z): 

 (30) 

where 

fs = the unit side shear stress mobilized along a pile segment at movement, z 

fs, max = the maximum unit side shear stress 

z = movement of pile segment 

zc = movement required to mobilize fs, max (0.3 inches for sands) 

The normalized T-Z curve obtained from equation 30 is shown in figure 162. For 

comparison purposes, the API (1993) T-Z curve is also shown in this figure. 
 

211 

Figure 162. Graph. Normalized T-Z curves according 

to API (1993) and Vijayvergiya (1977). 

Similarly, the following expression for pile tip load versus tip settlement for piles in 

sands is proposed by Vijayvergiya (1977): 

 (31) 

where 

q = the pile tip resistance mobilized at displacement z 

q max = the maximum tip resistance (end bearing) 

z = movement of pile segment 

zc = movement required to mobilize qmax (0.25 inch for sands) 

The normalized Q-Z curve proposed by Vijayvergiya (1977) is shown in figure 163. For 

comparison purposes, the corresponding API (1993) curve for a 24-inch diameter pile is 

also shown. 
 

212 

Figure 163. Graph. Normalized Q-Z curves according 

to Vijayvergiya (1977) and API (1993). 

 

It is clear from figure 163 that the Q-Z curve proposed by Vijayvergiya mobilizes the full 

end-bearing capacity at much smaller displacement than the API curve. Vijayvergiya 

(1977) estimates that about 6.35 mm (0.25 inches) of tip displacement are required to 

mobilize the ultimate pile end-bearing capacity, while API (1993) estimates that a pile tip 

displacement equal to 10 percent of the pile diameter is required. 

 

7.3.3.2.2 Maximum shaft shear stress distributions 

Vijayvergiya (1977) recommended using an expression similar to API (1993) to estimate 

the maximum shear stresses along the pile (see equation 21). The author indicated that 

the coefficient of lateral earth pressures (K) can be as high as 1.25 for driven piles in 

medium-dense to dense sand. 

The predictions presented in the following subsections are based on shaft shear stresses 

calculated using equation 21, with the δ values from the interface shear tests presented in 

chapter 3, and K values of 1.0 and 1.25. The shear stress distributions used are shown in 

figure 164. No limiting value of shear stress was used, although it should be pointed out 

that Vijayvergiya (1977) does mention the existence of a limiting value in the 1971 

version of the API guidelines. 
 

213 

Figure 164. Graph. Maximum shear stress distributions used 

in predictions using Vijayvergiya (1977). 

7.3.3.2.3 Tip resistances 

Vijayvergiya (1977) recommends estimating the tip resistance for piles driven in sands as 

the product of a bearing capacity factor (Nq) and the free-field effective overburden stress 

at the base of the pile (σ'v). 

The values recommended by API (1993) were used in the predictions presented in the 

following sections. 

7.3.3.2.4 Prediction for the prestressed concrete test pile 

The predicted load-settlement curves for the prestressed concrete test pile, using the 

recommendations proposed by Vijayvergiya (1977) for load-transfer curves, are shown in 

figure 165. 
 

214 

Figure 165. Graph. Settlement predictions for the concrete pile 

using Vijayvergiya (1977). 

Figure 165 shows that the predicted load-settlement curve, using Vijayvergiya’s (1977) 

recommendations, agrees well with the initial portion of the field measurements, but 

shows a stiffer response for loads higher than 1,000 kN (225,000 lbf). The ultimate pile 

capacities, predicted using coefficients of lateral earth pressures (K) of 1.0 and 1.25, were 

about 2,885 and 3,125 kN (649,125 and 703,125 lbf), respectively. These capacities, as 

percentages of the pile capacity measured in the field, are about 93 percent and 101 

percent for K equal to 1.0 and 1.25, respectively. 

7.3.3.2.5 Prediction for the FRP test pile 

The predicted load-settlement curve for the FRP test pile, using the Vijayvergiya (1977) 

load-transfer curves with a coefficient of lateral earth pressure (K) equal to 1.0, is shown 

in figure 166. 
 

215 

Figure 166. Graph. Settlement predictions for the FRP pile 

using Vijayvergiya (1977). 

Figure 166 shows that the predicted load-settlement curve, using Vijayvergiya’s (1977) 

method with a lateral earth pressure coefficient (K) equal to 1.0, agrees reasonably well 

with the field measurements. 

7.3.3.2.6 Prediction for the plastic test pile 

The predicted load-settlement curve for the FRP test pile, using the Vijayvergiya (1977) 

load transfer curves and with a lateral earth pressure coefficient (K) equal to 1.0, is 

shown in figure 167. 
 

216 

Figure 167. Graph. Settlement predictions for the plastic pile 

using Vijayvergiya (1977). 

Figure 167 shows that the predicted load-settlement curve, using Vijayvergiya’s (1977) 

recommendations and with a lateral earth pressure coefficient (K) equal to 1.0, is stiffer 

than observed in the field. 

7.3.3.3 Comments on the predictions obtained using empirical load-transfer curves 

The exactness of load-settlement behavior predicted using the load-transfer (T-Z) method 

depends to a great extent on the accuracy of the predicted ultimate shaft and tip capacities 

(e.g., as presented in section 7.2.2 for the API method), and on the displacement 

characteristics (e.g. stiffness, linear or nonlinear, displacement required for full-strength 

mobilization) chosen to model the soil-pile interface and the soil-pile tip during load 

transfer. For example, the results for the prestressed concrete pile, presented in the 

previous section and calculated using the API method, showed a significant 

underprediction of the pile capacity. This result is related to the low ultimate shaft 

resistance predicted compared to the actual ultimate shaft capacity recorded in the field. 

The prediction using Vijayvergiya’s (1977) recommendations, with a coefficient of 

lateral earth pressure (K) equal to 1.25, was reasonably good. The predictions for the 

composite piles were more accurate because the calculated ultimate shaft and tip 

capacities were much closer to the actual values measured. 
 

217 

7.3.4 Predictions Using Theoretical Load-Transfer Curves 

The following sections present the background and describe the methodology used in this 

research to develop theoretically based load-transfer curves. Most of the work in this area 

is based on the pioneering work by Kraft, et al. (1981). 

7.3.4.1 Theoretical load-transfer curves for a linear elastic soil 

The soil deformation around a pile shaft can be idealized as concentric cylinders in shear 

(Frank 1974, Randolph and Wroth 1978). This idealization assumes the soil displacement 

due to the pile loads is predominantly vertical, and that radial displacements are 

negligible. Studies by Frank (1974) and Baguelin, et al. (1975), have shown that the 

concentric cylinder approach is a good approximation of the deformation patterns 

obtained using more rigorous analyses such as the finite element method. 

Randolph and Wroth (1978) derived an expression for pile settlement in a linear elastic 

soil based on the concentric cylinder assumption. The author’s derivation starts by first 

considering the vertical equilibrium of the annular-shaped soil element (ring of soil 

between two concentric cylinders) shown in figure 168: 

 (32) 

where the term definitions are shown in figure 168. The vertical stress (σy) shown in this 

figure refers to net change in vertical stress due to the axial load (P) applied to the pile. 
 

218 

Figure 168. Illustrations. Concentric cylinder model for settlement analysis 

of axially loaded piles (adapted from Randolph and Wroth 1978). 

Simplifying and neglecting second-order terms, equation 32 reduces to: 

 (33) 

However, according to Randolph and Wroth (1978), the rate of change of vertical stress 

with respect to depth is much less than the rate of change of shear stress with respect to 

radial distance during axial loading of a pile. Therefore the second term of equation 33 

can be neglected, and the equilibrium equation shown above can be approximated as: 

 (34) 
 

219 

Integrating this last expression, we obtain: 

 (35) 

where 

τ(r) = shear stress acting at a radial distance (r) from the centerline of the pile 

το = τ(ro) = shear stress at the pile/soil interface (i.e., at a radial distance ro) 

ro = pile radius (or equivalent radius for noncircular piles) 

Equation 35 indicates that the shear stresses in the vicinity of the pile are inversely 

proportional to the radial distance to the pile. 

The concentric cylinder model assumes that simple shear conditions prevail. This model 

also assumes that the radial displacements are negligible compared to the vertical 

displacements of the soil. Therefore the shear strain (γ) can be approximated as: 

 (36) 

where 

u = radial displacement (neglected) 

w = vertical displacement 

For linear elastic soils, the shear strain is related to the shear stress as follows: 

dw (r) (r) dr G(r)τ

γ= = (37) 

By substituting equation 35 into equation 37, we obtain: 

 (38) 
 

220 

The settlement of the pile shaft (ws) as originally proposed by Randolph and Wroth 

(1978), is obtained by integrating equation 38 and assuming a linear elastic soil with no 

radial variation of G (i.e., G(r) = G), as follows: 

 (39) 

where rm is the radial distance at which shear stresses in the soil become negligible 

Randolph and Wroth (1978) suggested estimating rm using the following empirical 

expression: 

 (40) 

where 

l = pile embedment depth 

ρ = factor of vertical homogeneity of soil stiffness = G(at pile middepth)/G(at pile tip) 

υ = Poisson’s ratio of the soil. 

Equation 39 results in an equivalent linear T-Z load-transfer curve, as shown in figure 

169. The stiffness or slope of this T-Z curve is: 

 (41) 
 

221 

Figure 169. Graph. Linear T-Z curve obtained using Randolph and Wroth (1978). 

The value selected for rm has a large influence on the resulting pile settlement. A recent 

study by Guo and Randolph (1999) used finite element and fast Lagrangian analysis of 

continua (FLAC) analyses to determine the value of rm more precisely. As a first 

approximation, equation 40 provides reasonable values for rm (Randolph and Wroth 

1978). 

The load-transfer curve for the pile tip can be approximated using Boussinesq’s solution 

for a rigid footing resting on an elastic half-space (Poulos and Davis 1990): 

 (42) 

where 

Qb = the pile tip load 

Es, ν = elastic modulus and Poisson’s ratio of soil beneath pile tip 

ro = pile radius 

 

Equation 42 can be expressed in terms of the shear modulus by substituting 

Es = 2G(1+ν) as follows: 

 (43) 

The resulting Qb-Z load transfer curve is shown in figure 170. 
 

222 

Figure 170. Graph. Linear Qb-Z curve obtained using Boussinesq’s theory. 

The slope of the linear Qb-Z corresponds to the stiffness coefficient of the pile tip spring, 

and is given by: 

 (44) 

In the preceding derivations it was assumed that the soil is linear and elastic, with the 

parameters G and ν constant with both strain level and with radial distance from the pile. 

However, most soils exhibit stress-strain behavior that is highly nonlinear and their 

stiffness can decay rapidly with strains (Atkinson 2000). If settlement predictions are 

made using the linear elastic load-transfer curves shown above, the appropriate elastic 

properties should be used. Poulos (2001) recommends using a secant shear modulus 

corresponding to the load range applied to the pile. For the tip of a pile in sand, Poulos 

(2001) recommends increasing the shear modulus to account for soil densification below 

the pile tip due to driving. Successful application of a linear theory to a nonlinear 

problem requires judicious selection of elastic property values. 

Nonlinear load-transfer curves that include soil nonlinearity are derived in the following 

subsections. 
 

223 

7.3.4.2 Theoretical load-transfer curves for a nonlinear soil 

7.3.4.2.1 Initial shear modulus 

As shown in section 7.3.4.1, the shear stiffness of the soil deposit is a key property 

required in calculations of pile settlements. Sands typically only behave as linear elastic 

at very small shear strains of less than about 10 microstrains (γ ≤ 10-3 percent). The shear 

modulus in this range of very small shear strains is often referred to as the initial 

maximum or elastic shear modulus. Here it will be referred to as the initial shear modulus 

(Go). To perform nonlinear pile settlement analyses, the initial shear modulus profile 

along the length of the pile must be estimated. 

Elastic theory relates Go to the shear wave speed (Vs) as follows: 

 (45) 

where ρ is the soil mass density 

Therefore, the initial shear modulus could be obtained from in situ measurements of the 

shear wave speed, such as from seismic CPT soundings. If seismic tests are not available, 

as is the case for the Route 351 site, Go can be estimated using empirical correlations 

with in situ CPT tests (Lunne, et al., 1997). 

Several equations have been proposed to estimate Go from CPT qc; some of the most 

important ones are: 

• Baldi, et al. (1989), from calibration chamber tests on uncemented, quartzitic sands: 

 (46) 

where σ'vo is the in situ effective overburden pressure at the depth of qc, and the units 

of Go, qc, and σ'vo are in kPa. 

 

• Rix and Stokoe (1991) from calibration chamber tests on uncemented, quartzitic 

sands: 

 (47) 

 
 

224 

where σ'vo is the in situ effective overburden pressure at the depth of qc, and where 

Go, qc, and σ'vo are in kPa. 

 

• Equation 48, originally proposed by Chow (1996). Equation 48 is based on a 

reinterpretation of the calibration chamber tests run by Baldi, et al. (1989): 

 (48) 

where 

qc = CPT tip resistance 

A = 0.0203 

B = 0.00125 

C = 1.216 x 10-6

σ′v = free-field effective overburden pressure 

Pa = atmospheric pressure (100 kPa) 

7.3.4.2.2 Modeling soil nonlinearity using hyperbolic models 

Hyperbolic stress-strain models have been found to adequately represent the nonlinear 

behavior of most soils (Kondner 1963, Kondner and Zelasko 1963, Duncan and Chang 

1970). Use of hyperbolic functions to model nonlinear load-transfer functions for piles 

has been proposed by several researchers (e.g., Kraft, et al., 1981, Chow 1986, McVay, et 

al., 1989). The hyperbolic form of the shear stress-shear strain relationship is given by: 

 (49) 

where 

γ = shear strain 

Go = initial tangent shear modulus 

τult = shear stress value that the hyperbola tends asymptotically (not to be confused with 

τmax) 

τ = shear stress value corresponding to strain, γ

Duncan and Chang (1970) related the asymptotic value, τult, to the maximum shear stress, 

τmax, by: 

 (50) 
 

225 

where Rf is the failure ratio, and is typically a value less than or equal to one; for values 

of Rf less than 1.0 the failure stress (τmax) is reached at a finite shear strain; for Rf equal to 

1.0, the failure stress would be reached asymptotically at a shear strain of infinity. 

By substituting equation 50 and τ = Gsecγ into equation 49, we obtain the following 

expression for the secant shear modulus: 

 (51) 

If we include the radial variation of the shear stress (τ) obtained earlier using the 

concentric cylinder assumption (i.e., by substituting equation 35 into equation 51), we 

obtain: 

 (52) 

Substituting the secant modulus Gsec from equation 52 into equation 38, and evaluating 

the integral from radial distances ro to rm, we obtain the following nonlinear load-transfer 

(T-Z) relationship: 

 (53) 

Equation 53 results in the T-Z load-transfer curve shown in figure 171. For comparison, 

the linear T-Z curve derived earlier is also shown. It can be seen that the initial stiffness 

or slope of the nonlinear T-Z curve is the same as the linear curve (equation 41). The 

equation shown above becomes the linear elastic T-Z curve of equation 39 when Rf = 0. 
 

226 

Figure 171. Graph. Hyperbolic T-Z curve. 

For the hyperbolic model, the relationship between the normalized secant shear modulus 

(Gsec/Go) and the normalized shear stress (τ/τmax) plots as a straight line with slope -Rf, as 

shown in figure 172 (as given from equation 51). Fahey and Carter (1993) and Fahey, et 

al. (1994), showed that the hyperbolic model appropriately represented the shear modulus 

degradation observed in cyclic test data, but found that for monotonic tests on both 

normally consolidated and overconsolidated sands, the shear modulus degraded at a much 

faster rate than suggested by the hyperbolic model. Randolph (1994) also pointed this 

out, and indicated that in practice, real soils often show a more rapid decay of the secant 

shear modulus with shear stress. This rapid reduction in Gsec is observed when the very 

low strain initial shear modulus, obtained from dynamic tests, is used as initial reference 

(Randolph 1994, Atkinson 2000). The value of the initial shear modulus measured using 

conventional laboratory tests is smaller than the “true” initial shear modulus at very small 

strains measured using dynamic measurements (Atkinson 2000). The stiffness measured 

using conventional laboratory tests is lower because it is affected by factors such as the 

quality of the soil sample, small disturbances during sample setup in the test apparatus, 

seating of loading platen, and small misalignments of loading ram (Atkinson 2000). The 

conventional hyperbolic model may be more appropriate to capture the soil nonlinearity 

when the initial modulus used is from conventional laboratory tests. However, for initial 

soil modulus based on dynamic test measurements, the modified hyperbolic model such 

as the one proposed by Fahey and Carter (1993) may be more suitable (Randolph 1994). 
 

227 

 

Figure 172. Graph. Variation of secant shear modulus 

for different hyperbolic-type models. 

To allow for a faster degradation of the secant shear modulus degradation, with respect to 

the “true” initial very low strain shear modulus, Fahey and Carter (1993) proposed using 

the following modified form of the hyperbolic expression: 

 (54) 

where 

Go = initial shear modulus (for shear strains below 10-5) 

Gsec = secant shear modulus 

τ = current shear stress 

τmax = maximum shear stress 

f, g = empirical curve fitting parameters 

Equation 54 becomes the standard hyperbolic model for f = Rf and g = 1. Fahey and 

Carter (1993) found that the values of f of 0.98 and g of 0.25 gave a good fit for normally 

consolidated Toyoura sand. Toyoura sand is the Japanese standard sand, and it consists of 

a subangular fine sand with a mean particle size (D50) of 0.17 mm (Ishihara 1996). This 

sand has characteristics similar to the in situ sand at the Route 351 project. The nonlinear 
 

228 

T-Z curve, corresponding to the Fahey and Carter (1993) modified hyperbolic Gsec curve, 

can be easily derived using the same procedure as before. This expression is as follows: 

 (55) 

Equation 55 becomes equation 53 (which corresponds to the conventional hyperbolic 

curve) when f = Rf and g = 1. The T-Z curve obtained using equation 55, and f = 0.98 and 

g = 0.25, is shown in figure 173. As shown in figure 173, the initial stiffness (Ks) of this 

T-Z curve is the same as the other two (since Ks depends only on Go, ro, and rm), but this 

curve degrades at a much faster rate than the one based on the conventional hyperbolic 

equation. 

Figure 173. Graph. Theoretically derived T-Z curve using concentric cylinders 

and the modified hyperbola from Fahey and Carter (1993). 

So far we have only dealt with the side shear load-transfer curves (T-Z curves). Because 

the nature of the shear strain distribution below a pile tip is more complex, it is not as 

straightforward to model soil nonlinearity for Q-Z curves as it was for the T-Z curves. An 

initial approximation consists of assuming a hyperbolic variation of the stiffness 

coefficient of the pile tip spring such as proposed by Chow (1986): 

 (56) 
 

229 

where Ko

base is equal to Kbase from equation 44 (linear elastic soil case) 

The expression for the load-transfer curve is: 

 (57) 

The Qb-Z load-transfer curve obtained using equation 57 is shown in figure 174. For 

comparison purposes the linear elastic and conventional hyperbola curves are also shown. 

Figure 174. Graph. Theoretically derived Q-Z curve using Boussinesq’s theory 

and the modified hyperbola from Fahey and Carter (1993). 

 

As shown in figure 174, the ultimate tip resistance is developed at tip displacements of 

about 0.22 percent, 4.4 percent, and 10 percent, for the load-transfer curves based on the 

linear elastic, conventional hyperbolic (Rf = 0.95), and modified hyperbolic (f = 0.98, g = 

0.25) models, respectively. The very small peak tip displacement in the linear elastic 

load-transfer curve is related to the use of the very low strain initial shear modulus. As 

mentioned earlier, use of a secant modulus is recommended to obtain more realistic 

displacements when using a linear elastic theory. 
 

230 

7.3.4.3 Input parameters using modified hyperbola theoretical load-transfer curves 

To analyze the piles tested at the Route 351 Bridge project site using the modified 

hyperbolic load-transfer curves derived above, adequate values for the initial shear 

modulus (for shaft and base), and the maximum loads for the load-transfer curves (i.e., 

the maximum shear stress for the shaft T-Z, and maximum end-bearing pressure for the 

pile tip Q-Z) must be selected. The following subsections describe this selection process 

in detail. 

 

7.3.4.3.1 Hyperbola parameters 

Based on the characteristics of the sands encountered at the test site, modified hyperbola 

parameters of f = 0.98 and g = 0.25 were used. 

 

7.3.4.3.2 Initial shear modulus 

The profiles of initial shear modulus obtained using the equations 26, 46, and 47 shown 

above, and the design CPT profile for the Route 351 test site (figure 148), are shown in 

figure 175. 

Figure 175. Graph. Route 351 initial shear modulus profile from CPT correlations. 
 

231 

Both the magnitude and distribution (radially and with depth) are important. Poulos 

(2001) distinguished the following different values of the shear modulus required for 

settlement calculations of single piles: 

• The value of Go in the vicinity of the pile shaft (this value will strongly influence the 

settlement of the pile). 

• The value of Go immediately below the pile tip. 

Pile installation will greatly affect the shear modulus values in a zone near the pile 

(Poulos 1988). For driven piles in sands, the stiffness may actually be higher than the 

in situ value of the soil mass prior to pile installation (Randolph 1994). During pile 

installation the increase of radial stresses and associated densification will tend to 

increase the soil shear modulus; however, a competing effect which tends to reduce the 

shear modulus is the induced shear strains. The effects of pile installation on the 

magnitude and distribution of Go are difficult to quantify. If the installation effects are 

known, Kraft, et al. (1981), and Randolph (1994) have presented procedures that can be 

used to include the radial variation of shear modulus in the theoretical load-transfer 

curves. However, both considered the installation effects less important than the 

nonlinear effects and generally minor compared to the uncertainties involved in the 

estimated soil properties. Therefore, the analyses presented in the following sections are 

based on the Go profile presented in figure 175 with one exception: for the pile tip the Go

value selected was based on an average of the in situ Go values measured immediately 

below the pile tip, multiplied by a factor of 4. This factor of 4 is used to account for sand 

densification below the pile tip. Poulos (2001) recommends increasing the in situ 

stiffness immediately below the tip of driven piles in sands by a factor between 3 and 5. 

7.3.4.3.3 Maximum shear stresses 

The maximum shear stresses are estimated using equation 21, using the interface friction 

angles obtained from interface friction tests. The coefficient of horizontal earth pressure 

(K) is assumed to range between 1.0 and 1.25. The range selected for K is consistent with 

the values recommended in the literature for displacement piles driven in sands (Bowles 

1996, Poulos 2001). 

7.3.4.3.4 Ultimate tip resistance 

The theoretical Qb-Z curves require as input, in addition to the initial shear modulus and 

the Poisson’s ratio for the soil, the ultimate tip resistance, Qb-max. For the predictions 

presented in the following sections, the Qb-max values were selected from the Imperial 

College method estimates presented earlier (table 40). Qb-max values from other prediction 

methods could have been used, but the IC method had the closest toe capacity 

predictions. 
 

232 

7.3.4.4 Prediction for the concrete test pile 

The predicted load-settlement curves for the prestressed concrete test pile, using the 

proposed theoretically derived load-transfer curves, are shown in figure 176. 

Figure 176. Graph. Settlement predictions for the concrete pile 

using theoretically derived transfer curves. 

Using the maximum shear stresses calculated with a coefficient of horizontal earth 

pressures (K) of 1.25, the predicted load curve is in reasonably good agreement with the 

measured response. The analysis with K = 1.25 predicts the initial slope of the loadsettlement curve and ultimate pile capacity quite well, but shows a slightly stiffer 

response for loads between 1,500 and 2,900 kN (337,500 and 652,500 lbf). 

7.3.4.5 Prediction for the FRP test pile 

Using the proposed theoretically derived load-transfer curves, and a value for K of 1.0, 

the load-settlement curve for the FRP test pile was predicted, as shown in figure 177. 
 

233 

Figure 177. Graph. Settlement prediction for the FRP pile 

using theoretically derived transfer curves. 

Figure 177 shows a good agreement between the predicted load-settlement curve and the 

measured response. 

7.3.4.6 Prediction for the plastic test pile

The predicted load-settlement curves for the plastic test pile, using the proposed 

theoretically derived load-transfer curves, are shown in figure 178. 
 

234 

Figure 178. Graph. Settlement predictions for the plastic pile 

using theoretically derived transfer curves. 

Using the maximum shear stresses calculated with values of the coefficient of horizontal 

earth pressures (K) of 0.90 and 1.00, the predicted load curves are, up to an applied axial 

load of approximately 2,250 kN (506,250 lbf), in reasonably good agreement with the 

measured response. The analysis with K = 0.9 predicts well the load at which the steep 

change of slope occurs. However, neither of the analyses captures the slope of the second 

portion of the curve. 

 

7.3.4.7 Comments on the predictions using theoretical load-transfer curves 

The predictions obtained using the theoretically derived load-transfer curves were in very 

good agreement with the measured field behavior. The best agreements were found for 

values of the coefficient of horizontal earth pressures (K) of 1.25, 1.0, and 0.9 for the 

prestressed concrete, FRP, and plastic piles, respectively. These differences in the K 

coefficient values are contrary to the expected outcome. The expected outcome was to 

have similar K values for the three test piles since they were installed in similar soil 

conditions, and all are full-displacement piles with similar cross sectional dimensions. It 

is possible that this result is again related to the δ values measured in the laboratory not 

being representative of field conditions, perhaps due to the small displacements (only 0.5 

inches (12.7 mm)) used in the lab. 
 

235 

7.3.5 Residual Stresses 

The analyses presented in this chapter did not include the effects of residual stresses, 

although their importance in obtaining more accurate predictions is recognized. Based on 

the small level of residual stresses measured in the field (see chapter 6), the effects of 

residual stresses were not considered. The impact of not incorporating residual stress 

effects would likely affect stiffness in the pile response rather than ultimate capacity. 

7.4 CONCLUSIONS 

The axial capacities of the three test piles were predicted using static analysis methods 

commonly employed in practice. The methods employed were the Nordlund method, the 

API method, the LCPC method, and the IC method. The various methods used to predict 

axial pile capacity yielded ratios of calculated to measured pile capacities ranging from 

0.70 to 1.14 for the prestressed concrete pile, and from 0.81 to 1.33 for the composite 

piles. It was found that the level of accuracy of the predictions was comparable for all 

three test piles. 

The prediction results seem to indicate that conventional static analysis methods are 

applicable to composite piles. However, it is the author’s opinion that additional case 

histories are needed to corroborate and extend this conclusion to other composite pile 

types and to different soil conditions. 

The load-settlement behavior was predicted using T-Z analyses. Several of these analyses 

were carried out to determine the adequacy of this method to analyze axially loaded 

composite pile types such as those studied in this research project. The analyses were 

completed using empirical load-transfer curves available in the literature (e.g., API 1993, 

Vijayvergiya 1977) and theoretically derived load-transfer curves. 

The predictions using the empirical load transfer curves showed reasonable agreement 

with the field measurements for the composite piles. The predictions for the prestressed 

concrete pile showed less agreement, and tended to underestimate the measured pile 

capacity. This result could be related to the value of the interface friction angle used. The 

calculations were based on the interface friction angle measured in the lab, which may be 

inadequate to represent the field conditions experience by the pile (e.g., in the field the 

pile interface experienced much larger relative displacements compared to the 0.5 inches 

(12.7 mm) used in the laboratory). 

The predictions obtained using the theoretically derived load-transfer curves were in very 

good agreement with the measured field behavior. However, the values of the coefficient 

of horizontal earth pressures (K) that resulted in the best matches with field behavior 

were 1.25, 1.0, and 0.9 for the prestressed concrete, FRP, and plastic piles, respectively. 

The differences in the K coefficient values are contrary to the expected outcome. The 

expected outcome was to have similar K values for the three test piles since they were 

installed in similar soil conditions, and all are full-displacement piles with similar cross 

sectional dimensions. It is possible that this result is again related to the δ values 
 

236 

measured in the laboratory not being representative of field conditions, perhaps due to the 

small displacements (only 0.5 inches (12.7 mm)) used in the lab. 

The prediction results seem to indicate that conventional T-Z analyses are applicable to 

composite piles. However, it is believed that additional case histories are needed to 

corroborate and extend this conclusion to other composite pile types and to different soil 

conditions. 
 

237 

CHAPTER 8. ANALYSES OF THE LATERAL LOAD TESTS AT 

THE ROUTE 351 BRIDGE 

8.1 INTRODUCTION 

An important objective of this research is to determine whether accurate analyses of the 

lateral load-deflection behavior of composite piles can be performed using the same 

procedures typically used for prestressed concrete piles and other conventional pile types. 

If they can, this would remove one of the impediments to using composite piles by 

verifying that established procedures can be employed for design to resist lateral loads, at 

least for the type of composite piles studied in this research. 

This chapter describes the procedure used for analyzing the lateral load tests of the three 

test piles at the Route 351 Bridge project. The results of the analyses are compared to the 

measured responses presented in chapter 6. A brief overview of the laterally loaded pile 

problem and a description of the methodology used in lateral pile analyses also are 

presented in this chapter. 

 

8.2 GOVERNING DIFFERENTIAL EQUATION FOR THE LATERALLY 

LOADED PILE PROBLEM 

A laterally loaded single pile presents a soil-structure interaction problem. The soil 

reaction is dependent on the pile movement, and the pile movement is dependent on the 

soil reaction. The solution must satisfy a nonlinear differential equation as well as 

equilibrium and compatibility conditions. The solution usually requires several iterations. 

Elastic beam relationships that are used commonly in analysis of laterally loaded piles are 

summarized in table 41. These quantities are obtained from differentiating deflection y 

with respect to the distance along the pile (x). 
 

238 

Table 41. Relationships commonly used for elastic piles in flexion. 

Variable Formula Units 

Distance along the length of the pile 

(measured from pile head) X [L] 

Distance to neutral axis within pile 

cross section Z [L] 

Deflection Y [L] 

Slope or rotation of pile section [Dimensionless] 

Curvature [Radians/L] 

Bending moment [F x L] 

Shear force [F] 

Axial load Q [F] 

Soil reaction (or load intensity) [F/L] 

Notes: EpIp = flexural stiffness of pile, where Ep = elastic modulus of pile material, and Ip = moment of inertia of pile 

cross section with respect to the neutral axis. 

Figure 179 shows a loaded pile and typical profiles of net soil reaction, deflection, slope, 

and moment. 
 

239 

Figure 179. Illustration. Laterally loaded pile problem. 

The governing differential equation for the problem of a laterally loaded pile was derived 

by Hetenyi (1946). The differential equation can be obtained by considering moment 

equilibrium of the infinitesimal element of length (dx) as shown in figure 179. 

 (58) 

Neglecting quadratic terms, and differentiating twice with respect to x, we obtain: 

 (59) 

The term involving the axial load, Q, can be ignored for the test piles investigated in this 

research since the vertical load present during testing was mainly from self-weight and 

can be considered negligible. The magnitude of the bending moment acting at a given 

xx x x x

F yp y φ M

x

dx

pright

p (soil resistance)

pleft

a) Pile loading b) Net soil reaction c) Pile deflection d) Slope e) Bending moment

xx x x x

F yp y φ M

x

dx

pright

p (soil resistance)

pleft

a) Pile loading b) Net soil reaction c) Pile deflection d) Slope e) Bending moment

p 

M + dM 

V + dV 

V 

M 

dx 
 

240 

section of a pile can be calculated by integrating the normal stresses, σ(z), acting within 

the cross section of area, A, as follows: 

 (60) 

If we assume that plane sections of the pile remain plane after loading, we can calculate 

the strains across the pile cross section if we know the rotation of the section, φ = dy/dx, 

and the position of the neutral axis. For a given rotation, φ, we have the following: 

 (61) 

where 

u(x,z) = is the displacement in the x-direction across the pile cross section 

ε(z) = strains in the x-direction across the pile cross section 

z = distance to the neutral plane 

 

Substituting the expression for σ(z) from equation 61 into equation 60, we obtain: 

 (62) 

If the pile material is linear elastic with a constant Young modulus, Ep, we would have: 

 (63) 

Substituting equation 63 into equation 59 and ignoring the axial load term, Q, we obtain: 

 (64) 

From consideration of the horizontal force equilibrium of the infinitesimal element of the 

pile shown in figure 179, we obtain: 

 (65) 
 

241 

Substituting equation 65 into equation 59, we obtain the following governing differential 

equation, which is commonly used to analyze piles under lateral loads: 

 (66) 

The variable, p(x), in equation 66 corresponds to the resultant soil resistance force per 

unit length of pile that occurs when the unit length of pile is displaced a lateral distance, 

y, into the soil. A crucial point for solution of the differential equation shown above is 

adequate representation of the soil reaction, p. If the soil reaction, p, has a linear 

relationship with lateral pile deflection, y, the equation has a closed-form solution. 

However, the relationship between the soil reaction (p) and the pile deflection (y) is 

nonlinear and also varies along the pile depth. In practice, it is common to solve this 

differential equation (equation 66) using numerical methods such as the finite difference 

method, and by modeling the soil reaction using nonlinear p-y curves. The analyses 

presented in this chapter were carried out using this approach. The p-y method used to 

model the soil reaction is discussed in section 8.3. 

8.2.1 Assumptions and Limitations of the Governing Differential Equation 

Implicit in the derivation of equation 66 is the fact that the pile is made of a 

homogeneous, isotropic, linear elastic material with the same modulus of elasticity in 

tension and compression. Hence, the flexural stiffness of the pile, EpIp, is assumed to 

remain constant during bending. As shown in chapter 6, this assumption is not valid for 

the piles tested in this research. The nonlinearity of the elastic properties of the piles can 

be dealt with during the numerical solution of the differential equation by means of 

successive iterations to account for the nonlinear properties of the structural materials of 

the pile (Reese and Van Impe 2001). 

Another important assumption used in the derivation of equation 66 is that shear strains 

(or deformations) are small; that is, normals are assumed to remain normal to the neutral 

axis. This is a common assumption in classical beam theory or Bernoulli-Euler theory. 

The Mindlin beam theory, on the other hand, assumes normals to the neutral axis remain 

straight and undeformed, but not necessarily perpendicular to the neutral axis (Holzer, 

2001). If we denote with φ the angle of rotation of the normal, we have the following for 

both beam theories: 

 Bernoulli-Euler theory: φ = dy/dx

 Mindlin theory: φ = dy/dx + γ

If the shear strain, γ, is zero, the two theories are equal. The Mindlin theory assumes the 

shear strain is constant over the cross section, that is, independent of z. In reality, the 

shear strain (or stress) varies over the cross section; γ varies parabolically over a 

rectangular cross section made of a uniform material. The effect of shear deformations on 

beam deflections has been studied by Stippes, et al. (1961). They found that the total tip 
 

242 

deflection of a cantilever beam with uniformly distributed load and rectangular cross 

section is as follows: 

 (67) 

where the additional deflection due to shear deformations is given by the second term in 

the brackets. 

The effect of shear deformations increases with increasing E/G ratios and decreasing 

slenderness ratios (L/D). For the piles tested at the Route 351 Bridge, the slenderness 

ratio (L/D) is about 15 (considering only the length of the pile where lateral deflections 

are significant), and the E/2G ratio is estimated to be about 1.3 for the prestessed concrete 

and FRP piles, and about 5 for the plastic pile. Therefore, the error associated with 

neglecting shear deformations is estimated to be less than 0.6 percent for the prestressed 

concrete and FRP piles, and less than 2.2 percent for the plastic pile. However, Han 

(1997) pointed out the importance of considering shear deformations when studying 

laterally loaded FRP composite piles. This is believed to be especially important when 

dealing with thin-walled or hollow FRP beams. Han (1997) reported values of E/2G 

between 4 and 15 for typical FRP composites materials (with no concrete infill). These 

E/2G ratios would result in deflection errors between 1.8 percent to 6.7 percent for a 

slenderness ratio of 15. 

 

8.3 METHODOLOGY USED TO ANALYZE THE LATERALLY LOADED 

TEST PILES 

The p-y method is widely used for design of laterally loaded piles. This method replaces 

the soil reaction with a series of independent nonlinear springs. The p-y curves represent 

the nonlinear behavior of the soil by relating the soil reaction and pile deflection at points 

along the pile length. A review of the p-y method is presented in sections 8.3.1 and 8.3.2. 

8.3.1 P-y Curves 

8.3.1.1 Introduction 

The prediction of the soil resistance at any point along the pile as a function of pile 

deflection is perhaps one of the most critical factors in solving the problem of a laterally 

loaded pile. The distribution of stresses against a cylindrical pile before installation is 

shown in figure 180(a). At a given depth, the stresses will be uniform and normal to the 

pile wall (assuming the pile is installed vertically and without inducing bending) (Reese 

and Van Impe 2001). After the pile is subjected to lateral loading, the pile will deflect and 

the soil stresses acting on the pile would have a distribution similar to that shown in 

figure 180(b). It is important to point out that some of the stresses will not be 

perpendicular to the pile wall due to development of shear stresses at the interface 
 

243 

between the pile and the soil. The net soil reaction, p(x), is obtained by integrating the 

stresses around the pile cross section. The units of p(x) are force per unit length. 

Figure 180. Illustration. Distribution of stresses against a pile before and after 

lateral loading (adapted from Reese and Van Impe 2001). 

In general, p-y curves are nonlinear and they are a function of depth, soil type, and pile 

dimensions and properties. A typical p-y curve is shown in figure 181. Important 

elements of the p-y curve include the initial slope, Epy-max, and the ultimate soil resistance 

value, Pult. At any point of the p-y curve the soil reaction, p, is related to the pile 

deflection, y, through the p-y modulus, Epy (Reese and Van Impe 2001). The p-y modulus 

is also known as the reaction modulus and it has units of force/length2

. Reese and Van 

Impe (2001) propose using the nomenclature given above instead of the modulus of 

subgrade reaction, which was originally developed to describe settlement of footings and 

relates the footing pressure (units of force/length2

) to the footing settlement (units of 

length). These authors also point out that, although the subgrade modulus and Epy are 
 

244 

related to the values of the Young modulus of the soil, Es, they are not only a function of 

the soil, but rather a result of the soil-structure interaction process between the soil and 

the footing and pile, respectively. 

Figure 181. Graphs. Typical p-y curve and resulting p-y modulus 

(Reese and Van Impe 2001). 

Ideally p-y curves should be generated from full-scale lateral load tests on instrumented 

test piles. In the absence of experimentally derived p-y curves, it is possible to use 

empirical p-y formulations that have been proposed in the literature for different types of 

soils. Table 42 lists the sources for some of the p-y expressions commonly used in 

practice. 

Table 42. Recommended criteria for p-y curves in different soils 

(adapted from Reese, et al., 1997). 

Soil Type and Condition Reference 

Soft clay below the water table Matlock (1970) 

Stiff clay below the water table Reese, et al. (1975) 

Stiff clay above the water table Welch and Reese (1972), Reese and Welch (1975)

Sands Reese, et al. (1974) 

Sands API (1993) 

Soils with cohesion and friction Evans and Duncan (1982) 

Weak rock Reese (1997) 

Strong rock Nyman (1980) 
 

245 

The p-y curves are not defined solely by soil characteristics (Ashour and Norris 2000). In 

addition to the properties of the soil surrounding the pile, the p-y curves are influenced by 

several other factors, such as: pile cross-sectional shape and dimensions, interface friction 

angle between soil and pile, pile bending stiffness, and pile head conditions (Ashour and 

Norris 2000, Reese and Van Impe 2001). Ashour and Norris (2000) used the strain wedge 

model to study analytically the influence of some of these factors on p-y curves. They 

found that for uniform sand deposits, a stiffer pile results in stiffer p-y curves. They also 

found that if two piles have the same width, but one has a circular cross section and the 

other has a square cross section, the resulting p-y curves will be different. The square pile 

in sand showed a soil-pile resistance higher than the circular pile. The findings from 

Ashour and Norris are based on analytical studies, and to the best of our knowledge no 

full-scale experiments have been reported to confirm their findings. Reese and Van Impe 

(2001) also pointed out the influence of the shape of the pile cross section on the soil 

resistance, p, as illustrated in figure 182. 

Figure 182. Illustration. Schematic showing the influence of shape of cross section 

of pile on the soil reaction p (adapted from Reese and Van Impe 2001). 

The majority of the methods listed in table 42 consider only the influence of the soil 

properties and the pile width. If other factors such as pile shape and surface texture must 

be considered, p-y curves should be obtained experimentally based on full-scale tests. 

The p-y analyses carried out in this research used published recommendations for p-y 

curves. The recommendations by Reese, et al. (1974), were used for the sandy soils at the 

test site. A brief description of these recommendations is provided below. 

8.3.1.2 P-y curves for sands based on recommendations by Reese, et al., 1974 

The typical shape of a p-y curve for sands, as recommended by Reese, et al., 1974, is 

shown in figure 183. 
 

246 

Figure 183. Graph. Elements of a characteristic p-y curve for sand 

based on recommendations by Reese, et al. (1974). 

As shown in figure 183, the main elements that define the p-y curves for sands 

recommended by Reese, et al. (1974), are: 

• Initial p-y modulus, Epy-max, that defines the initial portion of the curve up to point A. 

• Ultimate soil resistance, pult, which defines the curve at point C and beyond. 

• Transition zone between points A and C. 

The coordinates of point C are y = 3b/80, and p = pult, where b is the pile width. The 

transition zone consists of two parts: a parabolic section between points A and B, and a 

straight line portion between points B and C. The coordinates of point B are defined as: 

 (68) 

where A bar subscript s and Bs are coefficients obtained from charts provided by Reese, 

et al., 1974 

The equation of the parabola is obtained knowing that it passes through point B and that 

it must be tangent to the straight line between points B and C. The coordinates of point A 

are obtained by finding the intersection point of the initial straight portion, with slope 

Epy-max, and the parabola. 

The LPILE 4.0M program includes this type of p-y curve as a default p-y curve for sands. 

The p-y curves are generated automatically by the program. The user needs to specify the 
 

247 

initial slope of the curve (Epy-max) and the soil properties of the sand (effective unit weight 

and friction angle) to define the ultimate soil resistance (pult). 

Reese, et al. (1974), recommend using a variation of Epy-max that increases linearly with 

depth, according to: 

 (69) 

where 

k = a constant giving the variation of Epy-max with depth 

x = depth below ground surface 

Typical k values for loose and medium-dense sands below the water table are 5.4 and 

16.3 MN/m3 (34,403 and 103,847 lbf/ft3), respectively (Reese, et al., 1974). 

Before presenting the analyses results, a brief description of the p-y method of analysis is 

presented in the following section. 

8.3.2 P-y Method of Analysis 

The p-y method of analysis of laterally loaded piles is analogous to the T-Z method 

described in chapter 7 to analyze the axially loaded piles. In essence, the method consists 

of dividing the pile into a series of increments of equal length. The governing differential 

equation (equation 66) is solved using the finite difference technique. The soil is 

idealized as a series of independent nonlinear springs whose characteristics are 

represented by the p-y curves described in the previous section. The idealization used in 

the p-y method is shown in figure 184. 
 

248 

Figure 184. Illustration. Schematic showing p-y model used for analysis of laterally 

loaded piles. 

The remainder of this chapter will deal with lateral load analyses carried out for the test 

piles at the Route 351 project. 

8.4 NUMERICAL ANALYSES RESULTS

The analyses performed in this research used the p-y method and the computer program 

LPILE Plus 4.0M (2000). This program employs a finite difference formulation to solve 

the differential equation presented in the previous section, with nonlinear p-y curves to 

model the soil. LPILE Plus 4.0M allows computation of the pile response with userspecified nonlinear pile flexural stiffness, EpIp. 

The program contains default p-y curves that can be used for different types of soils. As 

an alternative, the program also allows the user to input p-y curves developed using other 

formulations. For the analyses carried out in this research, the piles were discretized into 

300 elements—the maximum number of increments allowed in the LPILE 4.0M 

program. 
 

249 

8.4.1 General Input Information 

8.4.1.1 Pile information 

Table 43 summarizes the pile properties used in the analyses. The nonlinear flexural 

stiffness relationships presented in chapter 6 were used for the three test piles. 

 

Table 43. Properties of test piles. 

Property Prestressed Concrete 

Pile FRP Pile Plastic Pile 

Width/diameter (m) 0.610 0.622 0.592 

Perimeter (m) 2.44 1.95 1.86 

Area (m2) 0.372 0.304 0.275 

Length (m) 18.0 18.3 18.3 

Initial EpIp (kN-m2) 335,610 186,510 71,705 

EpIp versus M See figure 113(b) See figure 113(b) See figure 113(b) 

Notes: EpIp = flexural stiffness of pile, M = bending moment. 1 m = 3.28 ft 

 1 kN-m2

 = 2,421 lbf-ft2

8.4.1.2 Upper soil stratigraphy information for p-y analyses 

The soil stratigraphy near the top of the pile is the most important when studying laterally 

loaded piles (Duncan, et al., 1994). Typically, the significant lateral deflections of piles 

occur within the upper 8 to 10 diameters. The upper soil stratigraphy of the test pile site 

was found to be somewhat different at the north and south ends of the site. 

Figure 185 shows representative in situ test information for the upper 10.0 m (32 ft) of 

stratigraphy at the northern end of the test pile site. The uppermost layer at the northern 

end of the test pile site is silty fine sand fill approximately 1.0 m (3.2 ft) thick. The fill is 

underlain by loose to medium-dense, silty fine sand to a depth of 13.0 m (42.6 ft). 

The stratigraphy of the upper 10.0 m (32 ft) at the south end of the test pile site is shown 

in figure 186. The stratigraphy at the south end consists of 0.5 m (1.6 ft) of silty sand fill, 

which overlies a medium-stiff sandy, silty, clay layer that extends to about 1.8 m (5.9 ft) 

depth, which overlies loose to medium-dense silty sand. The extent of the clay layer was 

determined primarily based on the visual classification of the retrieved split spoon 

samples from boring SPT-2. The presence of the clay layer was confirmed indirectly via 

interpretation of the CPT and the flat dilatometer (DMT) test information available for 

the south end of the test site. 
 

250 

Figure 185. Graphs. In situ test data for the upper soils 

at the northern end of the test pile site. 
 

251 

Figure 186. Graphs. In situ test data for the upper soils 

at the southern end of the test pile site. 
 

252 

8.4.2 P-y Analyses Results 

A series of p-y analyses were carried out using the default p-y curves embedded in the 

LPILE program for different types of soils. As shown in section 8.4.1.2, the predominant 

type of soil at the test site consisted of silty sands. The default p-y curves recommended 

by Reese, et al. (1974), were selected to model these soils. 

The uppermost soil layer at the test site consists of manmade fills, with an average 

thickness of about 1 m and 1.8 m (3.3 ft and 5.9 ft) at the north and south ends of the test 

site, respectively. Since all three test piles were installed in pits approximately 1.0 m (3.3 

ft) deep, the majority of the fill materials were considered to be removed. Therefore, the 

p-y analyses were carried out using a soil model consisting of predominantly silty sands. 

However, it is recognized that the clayey fill layer present in the southern end extends 

about 0.9 m (2.9 ft) beyond the pit bottom of the prestressed concrete pile. This remnant 

fill was not specifically incorporated in the LPILE model for the test pile in the south 

side; instead, it was assumed to be part of the underlying silty sand deposit. This 

approximation was considered reasonable given the sandy and silty nature, and the 

thickness, of the low plastic clay fill layer. The results of the p-y analyses for the three 

test piles are summarized in the following sections. 

8.4.2.1 Analyses for the prestressed concrete test pile 

The prestressed concrete pile is located at the southern end of the test pile site. The 

surface stratigraphy for this pile is shown in figure 186. The pit excavated for this pile is 

0.79 m (2.6 ft) deep. The original ground surface and the point of load application are 

1.24 m and 1.34 m (4 ft and 4.4 ft) below the top of the pile, respectively. 

The LPILE analyses were carried out using a two-layer model consisting of a layer of 

loose to medium-dense sand approximately 10 m (32.8 ft) thick, underlain by medium to 

dense sand. The LPILE model was constructed to take into account the 0.79 m- (2.5 ft-) 

deep pit excavated before installation of this pile. As mentioned earlier, the silty sands 

were modeled using the p-y curves recommended for sands by Reese, et al. (1974). The 

main input information required to define these curves is the initial p-y modulus and the 

friction angle of the sand. The initial p-y modulus for sands can be adequately modeled as 

increasing linearly with depth (Reese, et al., 1974). The rate of increase of the p-y 

modulus was selected based on trial-and-error until a best fit was obtained between the 

LPILE results and the field measurements. The p-y parameters that provided the best 

match are summarized in table 44. The initial p-y modulus values used in the analyses are 

shown in figure 187. 
 

253 

Table 44. Parameters used to define default p-y curves 

in LPILE for the prestressed concrete pile. 

Parameter Loose Sand Medium-Dense Sand 

Default p-y curve Reese, et al., 1974 Reese, et al., 1974 

γ′, Submerged unit weight (kN/m3) 10 11 

c, Cohesion (kPa) 0 0 

φ, Friction angle (degrees) 33 35 

Epy-max, Initial modulus of p-y curve See figure 187 See figure 187 

1 kN/m3 = 6.36 lbf/ft3 

1 kPa = 0.145 lbf/inch2

Figure 187. Graph. Initial p-y modulus profile used to define p-y default 

curves for LPILE analyses on the prestressed concrete pile. 

The predicted deflected pile shapes are compared to the measured shapes in figures 188 

through 190. 
 

254 

Figure 188. Graphs. Predicted versus measured lateral displacement profile 

for prestressed concrete pile (low lateral loads). 

Figure 189. Graphs. Predicted versus measured lateral displacement profile 

for prestressed concrete pile (medium lateral loads). 
 

255 

Figure 190. Graphs. Predicted versus measured lateral displacement profile 

for prestressed concrete pile (high lateral loads). 

Figures 188–190 show that the deflected shapes of the pile are predicted reasonably well 

using the p-y modulus (Epy-max) values from figure 187 and the p-y curve equations 

recommended by Reese, et al., (1974). However, the predictions overestimate the lateral 

deflections for the two first loads of 51.2 and 97.3 kN (11,520 and 21,893 poundforce), 

and slightly underpredict the lateral deflections for the higher load levels. This could be 

related to the characteristics of the Reese, et al., (1974) p-y curves that may be on the 

“soft” side at low load levels and “stiff” at high load levels. As illustrated in figure 183, 

the Reese, et al., (1974) curves use a parabola and a straight section to connect the Pult

with the initial p-y modulus line. Since there are several possible ways to connect these 

two states, it is conceivable that other shapes of transition zone will result in different 

prediction results. A transition zone shape that is stiffer at low load levels and softer at 

high load levels may be better suited to capture the behavior of the soils encountered at 

this site. However, the approach was to use the default p-y model that would produce the 

best level of prediction possible. 

Using the LPILE soil model described above, the pile lateral deflections and head 

rotations at ground surface were calculated. The results are shown in figures 191 and 192. 
 

256 

Figure 191. Graph. Calculated load-deflection curve 

for the prestressed concrete pile. 

Figure 192. Graph. Calculated load-slope curve for the prestressed concrete pile. 

The predicted values of lateral deflection and pile head rotation at the ground surface 

show reasonably good agreement with the field measurements. These figures also show 
 

257 

the predicted values if the flexural stiffness of the pile is modeled as being constant, that 

is, independent of the level of applied moment. The predicted deflections using constant 

pile flexural stiffness are approximately 7 percent lower than measured at the pile under 

large lateral loads (> 200 kN (45,000 lbf)), and the agreement is closer at lower lateral 

loads. This finding is reasonable since the flexural stiffness of the prestressed pile is 

approximately constant up to a moment of about 400 kN-m (295,028 lbf-ft) (see figure 

113(b), chapter 6). Beyond this moment, the flexural stiffness of this pile decreases 

almost linearly with increasing applied moment. A similar behavior was observed for the 

head rotations. 

8.4.2.2 Analyses for the FRP pile 

The FRP pile is located at the northern end of the test pile site. The stratigraphy 

information for this pile is shown in figure 185. The pit excavated for this pile is 1.06 m 

(3.5 ft) deep. The original ground surface and the point of load application are 1.09 m 

(3.6 ft) and 0.79 m (2.6 ft) below the top of the pile, respectively. 

For this pile, the LPILE model used in the analyses was constructed using two layers: an 

upper layer of loose to medium-dense sand, and a lower layer of medium to dense sand. 

The p-y parameters that provided the best match are summarized in table 45. Both sand 

layers were modeled using the p-y curves of the type recommended by Reese, et al. 

(1974). 

Table 45. Parameters used to define default p-y curves in LPILE for the FRP pile. 

Parameter Loose Sand Medium Dense Sand 

Default p-y curve Reese, et al., 1974 Reese, et al., 1974 

γ′, Submerged unit weight (kN/m3) 10 11 

C, Cohesion (kPa) 0 0 

φ, Friction angle (degrees) 33 35 

Epy-max, Initial modulus of p-y curve See figure 193 See figure 193 

1 kN/m3 = 6.36 lbf/ft3 

1 kPa = 0.145 lbf/inch2 

The initial p-y modulus values that provided the best fit with the field measurements are 

shown in figure 193. The predicted deflected pile shapes are compared to the measured 

shapes in figures 194 through 196. 
 

258 

Figure 193. Graph. Initial p-y modulus profile used to define default p-y curves 

for LPILE analyses on the FRP pile. 

Figure 194. Graphs. Predicted versus measured lateral displacement profile 

for FRP pile (low lateral loads). 
 

259 

Figure 195. Graphs. Predicted versus measured lateral displacement profile 

for FRP pile (medium lateral loads). 

Figure 196. Graphs. Predicted versus measured lateral displacement profile 

for FRP pile (high lateral loads). 
 

260 

These figures show that the pile deflections are predicted reasonably well using the p-y 

curves recommended by Reese, et al. (1974), for sands and the p-y modulus (Epy-max) 

values from figure 194. For the first lateral load of 51.6 kN (11,610 lbf), the prediction 

overestimates the lateral deflections. This could be related to the p-y curve shape 

characteristics, as for the prestressed concrete pile. 

Using the p-y curves given above to model the soil, lateral deflections and pile head 

rotations were computed for the FRP pile at the ground surface. The results are shown in 

figures 197 and 198. 

Figure 197. Graph. Calculated load-deflection curve for the FRP pile. 
 

261 

Figure 198. Graph. Calculated load-slope curve for the FRP pile. 

The predicted values of lateral deflection and pile head rotation at the ground surface 

show good agreement with the field measurements. These figures also show the 

calculated values assuming that the flexural stiffness of the pile is constant, that is, 

independent of the level of applied moment. The predicted deflections using constant pile 

flexural stiffness are approximately 30 percent lower than measured under large lateral 

loads (> 150 kN (33,750 lbf)), and the agreement is closer at lower lateral loads. This 

finding is reasonable since the flexural stiffness of the FRP pile is approximately constant 

up to a moment of about 200 kN-m (147,514 lbf-ft) (as shown in figure 113(b), chapter 

6). Beyond this moment, the flexural stiffness of this pile decreases with increasing 

applied moment. A similar behavior was observed for the pile head rotations. 

8.4.2.3 Analyses for the plastic pile 

The plastic pile is located at the center of the test pile site. The extent of the surficial fill 

at this location was not determined, but is expected to be similar to the conditions found 

at the south and north ends of the site. The pit excavated for this pile is 0.91 m (3 ft) 

deep. The original ground surface and the point of load application are 1.02 m and 1.21 m 

(3.3 ft and 4 ft) below the top of the pile, respectively. 

For this pile, like the other piles, the two-layer LPILE model consisted of a layer of loose 

sand extending 10 m (32.8 ft) deep and an underlying layer of medium to dense sand. The 

p-y parameters that provided the best match are summarized in table 46. Both sand layers 

were modeled using the p-y curve types recommended by Reese, et al. (1974). 
 

262 

Table 46. Parameters used to define default p-y curves in LPILE for the plastic pile. 

Parameter Loose Sand Medium Dense Sand 

Default p-y curve Reese, et al., 1974 Reese, et al., 1974 

γ′, Submerged unit weight (kN/m3) 10 11 

c, Cohesion (kPa) 0 0 

φ, Friction angle (degrees) 33 35 

Epy-max, Initial modulus of p-y curve See figure 199 See figure 199 

1 kN/m3 = 6.36 lbf/ft3 

1 kPa = 0.145 lbf/inch2

The profile of initial p-y modulus with the depth that provided the best match is shown in 

figure 199.The predicted deflected pile shapes are compared to the measured shapes in 

figures 200 through 202. 

Figure 199. Graph. Initial p-y modulus profile used to define default p-y curves 

for LPILE analyses on the plastic pile. 
 

263 

Figure 200. Graphs. Predicted versus measured lateral displacement profile 

for plastic pile (low lateral loads). 

Figure 201. Graphs. Predicted versus measured lateral displacement profile 

for plastic pile (medium lateral loads). 
 

264 

Figure 202. Graphs. Predicted versus measured lateral displacement profile 

for plastic pile (high lateral loads). 

Figures 200–202 show that the pile deflections are predicted reasonably well using the 

p-y curves recommended by Reese, et al. (1974), for sands and the p-y modulus (Epy-max) 

values from figure 199. The level of agreement between the calculations and 

measurements was similar for the different load levels, suggesting that the Reese, et al., 

(1974) p-y curves adequately capture the response measured in the field. 

Using the LPILE soil model described above, pile lateral deflections and head rotations at 

the ground surface were computed. The results are shown in figures 203 and 204. 
 

265 

Figure 203. Graph. Calculated load-deflection curve for the plastic pile. 

Figure 204. Graph. Calculated load-slope curve for the plastic pile. 

 
 

266 

The calculated values of lateral deflection and pile head rotation at the ground surface 

show good agreement with the field measurements. For this pile, the predicted values 

using variable and constant flexural stiffness of the pile are the same. This finding is 

reasonable because the flexural stiffness of the plastic pile is primarily the result of 

contributions from the steel rebar cage. Therefore, as shown in figure 113(b) (chapter 6), 

the flexural stiffness for this pile is approximately constant up to a moment of about 650 

kN-m (479,421 lbf-ft). This moment was not exceeded during field load testing. 

8.4.3 Comparison of the Initial p-y Modulus Curves for the Three Test Piles 

The initial p-y modulus for the three test piles was assumed to increase linearly with 

depth. This assumption was considered reasonable for sand deposits such as those 

encountered at the test site. The rate of modulus increase with depth was selected to 

provide the best match between the analytical predictions and the field measurements. 

Using this approach, we obtained rates of modulus increase with depth of 1.7, 5.5, and 

2.2 MN/m3

 (10,831, 35,041, and 14,016 lbf/ft3) for the prestressed, FRP, and plastic piles, 

respectively. These variations are not in agreement with the trend expected based on the 

results of the axial load tests, for which the average unit shaft capacities were 61.8, 46.9, 

and 48.9 kPa (8.96, 6.80, and 7.09 (lbf/inch2) for the prestressed concrete pile, the FRP 

pile, and the plastic pile, respectively. 

The differences in the backcalculated rate of increase of the initial p-y modulus with 

depth can be due to several factors, such as: 

• Differences in soil stratigraphy at the location of each test pile. For example, a thin 

clay layer was encountered at the southern end of the test pile area near the 

prestressed concrete test pile. This layer was found to extend about 0.9 m (2.9 ft) 

below the pit at the prestressed concrete pile location. 

• Differences in pile properties such as cross-sectional shape, pile width, pile stiffness, 

surface roughness, and interface friction. 

Although the methodology based on double derivation of the bending moment versus 

depth curves (Reese and Van Impe 2001) is more appropriate for detailed backcalculation 

of p-y curves, even this method would not have produced p-y curves in accordance with 

expectations based on the pile characteristics. It seems most likely that the differences in 

p-y curves are due largely to differences in subsurface conditions. 

8.5 LIMITATIONS OF P-Y ANALYSES 

The p-y method was selected to analyze the lateral load tests carried out at the Route 351 

Bridge. This methodology was selected because it is commonly employed in practice. For 

this research project, it was desirable to verify whether established procedures, such as 

the p-y method, could be employed to analyze composite piles. Despite the popularity of 

this method, it has limitations as described below: 
 

267 

• The soil is idealized as a series of independent nonlinear springs represented by 

p-y curves. Therefore, the continuous nature of the soil is not explicitly modeled. 

• The results are very sensitive to the p-y curves used. The selection of adequate 

p-y curves is the most crucial problem when using this methodology to analyze 

laterally loaded piles (Reese and Van Impe 2001). 

• Selecting appropriate p-y modulus and p-y curves is a difficult task. While the 

selection of values of the initial p-y modulus (Epy-max) is related to the soil modulus, it 

is also related to the interaction between the pile and the soil. Reese and Van Impe 

(2001) point out that p-y curves and modulus are influenced by several pile-related 

factors, such as: 

o Pile type and flexural stiffness. 

o Type of loading (monotonic or cyclic). 

o Pile geometry. 

o Pile cap conditions. 

o Pile installation conditions. 

8.6 SUMMARY 

Numerous p-y analyses were carried out to determine the adequacy of this method to 

analyze laterally loaded composite pile types such as the ones studied in this research 

project. 

A derivation of the governing differential equation for the laterally loaded pile problem 

was presented, and possible limitations in composite piles analysis were discussed. 

The importance of considering shear deformations in lateral pile analyses was discussed. 

The impact of shear deformations increases with increasing E/G ratios, and decreases 

with increasing slenderness ratios (L/D). For the test piles assessed in this project, the 

error associated with neglecting shear deformations is estimated to be less than 2.5 

percent. 

The importance of including the nonlinearity of the flexural stiffness was discussed and 

illustrated with the analyses results. 

The results of the p-y analyses using published p-y curves that are embedded in the 

LPILE 4.0M computer program showed reasonably good agreement with the field 

measurements. 

The initial modulus of the p-y curves was found to increase with depth at the highest rate 

for the FRP pile, at an intermediate rate for the plastic pile, and at the lowest rate for the 

prestressed concrete pile. 

 

269 

CHAPTER 9. LONG-TERM MONITORING AT THE 

ROUTE 351 BRIDGE 

9.1 INTRODUCTION 

An important reason that composite piles have not gained wide acceptance in civil 

engineering practice is the lack of a long-term track record of performance. To help fill 

this void, a long-term monitoring program was implemented at the Route 351 Bridge 

project. The initial plan was to compare the performance of three adjacent bridge piers, 

each supported by different pile types. Coordination and logistical difficulties, however, 

prevented implementation of the initial plan. Consequently, the original long-term 

monitoring program was reduced to monitoring two instrumented piles installed at the 

center of two adjacent piers. 

The instrumented production piles will be monitored for several years. This reduced 

program is not ideal, particularly because the bent where the FRP composite pile is 

located will have two types of piles, each with a different stiffness, and this may affect 

load distribution among the piles in the pier. In addition, the pile arrangements in the two 

piers are not the same. The pier with the FRP pile has five central piles installed 

vertically, and the two outer piles are installed with 4-to-1 batter in the direction 

perpendicular to the bridge axis. The pier with the instrumented prestressed concrete pile 

has three central piles installed vertically, and the two outer piles on each end of the pier 

are installed at 4-to-1 batter oriented in the direction of the bridge axis. Despite these 

shortcomings, this monitoring program is expected to provide some insight on the longterm durability and performance of the FRP composite pile. This chapter describes the 

instrumented production piles; the instrumentation used; the soil conditions at the 

locations of the instrumented piles; the installation information including driving records, 

PDA, CAPWAP, and PIT tests; and the monitoring data gathered to date. 

9.2 INSTRUMENTED PRODUCTION PILES 

Long-term instrumentation was installed in a Hardcore FRP composite pile and a 

prestressed concrete pile, and the piles were installed as the center piles of Piers 11 and 

10, respectively. Figure 205 shows the locations of the instrumented production piles. 
 

270 

Figure 205. Illustration. Location of instrumented production piles 

at the Route 351 Bridge. 

9.2.1 Prestressed Concrete Production Pile 

The instrumented prestressed concrete production pile has characteristics similar to the 

test pile described in chapter 6. However, the production pile is longer, with a length of 

22 m (72.2 ft). The 28-day strength of the concrete used in this pile is 61 MPa (8,845 

lbf/inch2

). 

The prestressed concrete pile was instrumented with eighteen vibrating-wire sister bar 

strain gages located at seven levels, as shown in figure 206. These gages will be used to 

assess the load transfer from pile to soil. 

As shown in figure 206, a pair of gages was located at each level to provide gages on 

opposite sides of the pile. Two extra gages were installed at the uppermost and lowermost 

levels for redundancy. The gages are Geokon Model 4911 rebar strain meter gages. These 

gages consist of two lengths of No. 12M, grade 420 steel bar (No. 4 steel bar, grade 60) 

welded onto each end of a miniature vibrating-wire strain gage element. The entire 

assemblage is 1.3 m (4.3 ft) long. The instrumentation for the prestressed concrete 

production pile was installed on November 6, 2001. The pile was cast on November 7, 

2001. The vibrating-wire sister bar strain gage survival rate, after pile manufacturing, 

driving, and restrike, was 100 percent. 
 

271 

Figure 206. Illustration. Load transfer instrumentation layout 

for prestressed concrete production pile. 
 

272 

9.2.2 FRP Composite Production Pile 

The instrumented FRP composite production pile has the same characteristics as the FRP 

composite test pile described in chapter 6, but the length of this pile is 21.2 m (69.6 ft). 

Similar to the prestressed concrete pile, the FRP composite production pile was 

instrumented with 18 vibrating-wire sister bar strain gages located at 7 levels, as shown 

in figure 207. These gages will be used to assess the load transfer from pile to soil. 

 

The FRP pile was also instrumented with gages installed to study the performance and 

durability of the FRP tube. The following durability instrumentation was installed to 

monitor strains in the FRP tube: 

• Two foil gages at each of six levels, labeled IG-1 through IG-6, as shown in figure 

207, for a total of twelve gages bonded to the internal wall of the FRP tube. The foil 

gages are type CEA-13-500UW-350, temperature compensating, underwater, 

resistance strain gages. The gages are configured in a one-quarter active bridge. A 

coating system was used to cover these gages and protect them from possible 

moisture and chemical attack during filling of the tubes with concrete. The protective 

coating consisted of a Teflon film, overlain by a layer of polymer compound, overlain 

by aluminum foil, overlain by another layer of polymer compound, overlain by 

aluminum foil, with the edges sealed with a rubber coating. 

• Four vibrating-wire strain gages were installed on the outside surface of the FRP pile 

at the locations shown in figure 208. These gages are Geokon Model VK4100 

vibrating-wire strain gages. These gages have a 50 mm- (2 inch-) gage length. The 

gages were installed by epoxy bonding their mounting tabs to the pile exterior. In 

addition to providing strain data, these gages can measure temperature since they 

come with a thermistor incorporated. 

The vibrating-wire sister bars and the internal foil resistance strain gages for the FRP 

composite production pile were installed between January 24 and February 1, 2002. The 

concrete was poured for this pile on February 6, 2002. The vibrating-wire sister bar strain 

gage survival rate, after pile manufacturing, driving, and restrike, was 100 percent. The 

survival rate of the internal foil gages was 30 percent. The majority of the internal foil 

gages started to drift after the concrete was poured. This could be due to moisture from 

the concrete. The external vibrating-wire strain gages were installed after pile installation 

on November 7, 2002. As of December 30, 2002, all external gages were functioning. 
 

273 

Figure 207. Illustration. Instrumentation layout for FRP composite production pile. 
 

274 

Figure 208. Illustration. Layout of external durability instrumentation used 

in the FRP composite production pile. 
 

275 

9.3 SOIL CONDITIONS AT THE INSTRUMENTED PRODUCTION PILES 

VDOT investigated the soil conditions along the bridge alignment. The general soil 

stratigraphy found along the Route 351 Bridge alignment is shown in figure 209. The 

uppermost layer of river sediments consists of very soft to soft silty clay of varying 

thicknesses. The soft river sediments are underlain by loose to medium-dense, silty fine 

sand extending to variable depths depending on the location along the alignment. In 

general, the silty fine sand was found to increase in density with increasing depth, 

becoming dense to very dense at a depth of about 30 m (98.4 ft). 

The specific soil conditions found at Piers 10 and 11 are shown in figures 210 and 211, 

respectively. As shown, a sandy, silty, clay interbed was found at Piers 10 and 11, 

between elevations -18.7 and -20 m (61.3 and 65.6 ft), and -11.6 to -20 m (38.0 and 65.6 

ft), respectively. For reference, the elevations of the installed piles are also shown in 

these figures. 

Figure 209. Graph. Simplified stratigraphy along the Route 351 Bridge alignment. 

 
 

276 

Figure 210. Graph. Simplified stratigraphy in the vicinity of the instrumented 

prestressed concrete pile installed at Pier 10. 
 

277 

Figure 211. Graph. Simplified stratigraphy in the vicinity of the instrumented 

FRP composite pile installed at Pier 11. 
 

278 

9.4 PILE INSTALLATION AND DYNAMIC TESTING 

9.4.1 Pile Driving 

The instrumented production piles were driven with an ICE model 80S single-acting 

diesel hammer. The maximum rated energy of the hammer is 108.6 kN-m (80,000 ft-lbs). 

Both instrumented production piles were driven with 230 mm- (9 inch-) thick plywood 

pile cushions. The pile-driving records are shown in figure 212. The prestressed concrete 

pile and FRP pile were driven on June 13 and June 14, 2002, respectively. 

Figure 212. Graph. Driving records for instrumented production piles. 
 

279 

9.4.2 Dynamic Testing 

Dynamic monitoring was performed using a Pile Driving Analyzer (PDA). The PDA 

monitoring and data processing were performed by an FHWA contractor. Monitoring was 

performed during both the initial driving and during restrike. Restrike occurred on June 

18, 2002, which represents a setup period of 4 and 5 days after initial driving for the FRP 

composite pile and prestressed concrete pile, respectively. PDA records for the two 

instrumented production piles at restrike are shown in figure 213. The PDA records show 

a force peak at about time 0, measured with respect to the horizontal bars in figure 213 

that represent the wave travel time for two pile lengths for each pile. The prestressed 

concrete and FRP composite piles show similar dynamic behavior. Both piles show a 

large increase in the velocity record and corresponding low force at about 2L/c (where L 

is pile length, and c is wave speed). This type of PDA record shape is characteristic of 

piles with a small toe resistance. Very little separation between the force and velocity 

traces appears between time 0 and about 0.5L/c. This corresponds to the zone of zero to 

minimal shaft resistance; that is, the portion of the pile above the ground surface and the 

portion through the very soft river sediments where shaft resistance is small. 

Table 47 provides some of the measurements obtained during pile driving. It can be seen 

that the wave speed (c) and impedance values (EA/c) recorded in the FRP composite pile 

were about 5.3 percent and 6.1 percent lower, respectively, than the prestressed pile. The 

maximum compressive and tensile stresses in the prestressed concrete pile were lower 

than the allowable stresses recommended for prestressed piles (Hannigan, et al., 1996). 

The highest driving stresses were recorded in the FRP composite pile; however, no 

standards are currently available for allowable driving stresses of FRP composite piles. 
 

280 

Figure 213. Graph. PDA recordings during restrike (Spiro and Pais 2002b). 

Table 47. Summary of pile-driving measurements for the prestressed 

and FRP production piles (after Spiro and Pais 2002b).

Pile Type 

Measurement 

Prestressed FRP 

Wave speed (c) 4054 m/s 3840 m/s 

Impedance (EA/c) 3589 kN/m/s 3370 kN/m/s 

Maximum compression stress 

measured during driving 11.7 MPa 13 MPa 

Maximum tensile stress 

measured during driving 7 MPa 8.8 MPa 

Allowable stresses Comp. < 24.3 MPa 

Tension < 7 MPa 

No standards 

available 

1 m = 3.28 ft 

 1 kN/m = 68.6 lbf/ft 

 1 MPa = 145 lbf/inch2 
 

281 

Estimates of pile capacity from the dynamic strain and acceleration measurements were 

obtained using the Case method and CAPWAP analyses (Spiro and Pais 2002b). Both the 

Case method and CAPWAP capacities are listed in table 48. The FRP composite 

production pile showed an increase in the Case method capacity of about 125 percent 

between the end of initial driving and restrike 4 days later. According to the CAPWAP 

analyses, the prestressed concrete and FRP composite piles have capacities of 1,980 and 

2,046 kN (445 and 460 kips), respectively. These CAPWAP capacities are over two times 

larger than the axial design load of 890 kN (200 kips) for the piles in the piers. The 

CAPWAP analyses also indicate that the shaft capacities, as percentages of the total 

capacities, are 82 percent and 71 percent for the prestressed concrete and FRP piles, 

respectively. These high shaft capacity percentage values agree with the observations 

made earlier regarding the shape of the PDA records being characteristic of piles with 

low toe resistance. 

Table 48. Summary of CASE and CAPWAP analyses results 

(Spiro and Pais 2002b). 

Method Capacity 

Component 

Prestressed 

Concrete Pile FRP Pile 

Case (1) (End-of-Driving) Total Not available (2) 890 kN 

Case (1) (Restrike (3)) Total 2091 kN 2002 kN 

Shaft 1624 kN 1459 kN 

CAPWAP (Restrike Toe 356 kN 587 kN (3)) 

Total 1980 kN 2046 kN 

Notes: (1) Case method using a Case damping coefficient of 0.60. 1 kN = 225 lbf, or 0.225 kip 

(2) Not available due to signal acquisition problems. 

(3) Restrike of piles performed 4 and 5 days after initial driving, for the FRP and prestressed concrete piles. 

9.5 PILE INTEGRITY TESTING OF TEST PILES 

Pile integrity tests (PIT) were performed on the instrumented production piles before and 

after pile installation. These tests were carried out using the same procedure used for the 

test piles described in chapter 6. 

PIT results for the prestressed concrete production pile installed at Pier 10, performed 

before and after pile driving, are shown in figures 214 and 215, respectively. No 

discernable damage was noted based on comparison of the pre- and postdriving PIT 

soundings. The soil resistance in the pile after installation results in a smaller toe 

reflection, as shown in figure 215. 
 

282 

Figure 214. Graph. PIT sounding on the prestressed concrete 

production pile before installation. 

Figure 215. Graph. PIT sounding on the prestressed concrete 

production pile after installation and restrike. 
 

283 

PIT results for the FRP composite pile installed at Pier 11 are shown in figures 216 and 

217. No significant difference between the pre- and postdrive PIT traces was observed for 

the FRP pile. Toe reflection in the postdrive PIT is almost not discernible, probably due 

to high soil resistance. No discernable damage could be detected based on comparison of 

the pre- and postdriving PIT soundings. 

 

Figure 216. Graph. PIT sounding on the FRP composite 

production pile before installation. 

Figure 217. Graph. PIT sounding on the FRP composite 

production pile before installation. 
 

284 

9.6 MONITORING DATA GATHERED TO DATE 

Important date information for these two piles is summarized in table 49. 

Table 49. Test dates for the instrumented production pile 

program at the Route 351 Bridge. 

Pile Type Cross Section Length 

(m) 

Fabrication 

Date 

Driving 

Date 

Restrike 

Date 

Prestressed 

concrete 610 mm square 22 11/7/01 6/13/02 6/18/02 

FRP 622 mm circular 21.2 2/6/02 6/14/02 6/18/02 

1 mm = 0.039 inch 

1 m = 3.28 ft 

Due to delays in the construction schedule of the bridge, only a few readings have been 

made on the instrumented production piles. The instrumentation for the prestressed and 

FRP composite piles have been read on the dates shown in tables 50 and 51, respectively. 

On December 30, 2002, the estimated loads acting on the prestressed concrete and FRP 

composite piles were 24.1 percent and 30.9 percent of the axial design load, respectively. 

 

Table 50. Monitoring dates for prestressed concrete pile at the Route 351 Bridge. 

Date Description of Pile Loading Condition Estimated Load on Pile 

6/18/02 After pile installation and restrike 0 kN 

11/7/02 Pile cap and girders from east span installed 110 kN 

12/30/02 Pile cap and girders from adjacent spans 

installed. Deck slab for east span in place 227 kN 

1 kN = 225 lbf 

Table 51. Monitoring dates for FRP composite pile at the Route 351 Bridge. 

Date Description of Pile Loading Condition Estimated Load on Pile 

6/18/02 After pile installation and restrike 0 kN 

11/7/02 Pile cap and girders from east and west spans 

installed 155 kN 

12/30/02 Pile cap, girders, and deck slab from adjacent 

spans in place 290 kN 

1 kN = 225 lbf 

The loads acting on each pile were estimated based on the information reported in the 

bridge design drawings. The weight of the pile cap is based on unit weight for the 

reinforced concrete of 23.6 kN/m3

 (150 lbf/ft3) and the pile cap dimensions of 1.22 by 
 

285 

1.27 by 12.5 m (4.00 by 4.17 by 41.0 ft). The load from the AASHTO Type II girders (7 

girders per each span) is calculated based on a span length of 15.5 m (50.8 ft) and a girder 

weight per unit length of about 5.6 kN/m (384 lbf/ft) (Nilson 1987). The load due to the 

reinforced concrete bridge deck is based on a slab thickness of 0.2 m (0.656 ft) and a 

deck width of 12.8 m (42.0 ft). As a first approximation, the loads acting on the pile were 

estimated by dividing the total pier load by the number of piles in the pier. 

A photo of the bridge taken on December 30, 2002, is shown in figure 218. This photo 

shows that Piers 7 through 12 have been installed and the bridge deck has been poured up 

to Pier 10. The photo also shows the formwork for the deck between Piers 9 and 10. 

Figure 218. Photo. Route 351 Bridge under construction on December 30, 2002. 

The load distributions estimated for the prestressed concrete and FRP composite piles are 

shown in figures 219 and 220, respectively. These load distributions were estimated by 

multiplying the average strain at each level of vibrating-wire sister bars by the values of 

axial stiffness (EA) of each pile (as reported in chapter 6). The load distributions were 

calculated considering the residual stresses after driving using the procedure described by 

Fellenius (2002). 

Long-term monitoring and data collection responsibility are being transferred to VDOT 

and the Virginia Transportation Research Council (VTRC). 
 

286 

Figure 219. Graph. Load distributions on November 7, 2002, and December 30, 

2002, for the prestressed concrete production pile at Pier 10. 
 

287 

Figure 220. Graph. Load distributions on November 7, 2002, and December 30, 

2002, for the FRP composite production pile at Pier 11. 

9.7 SUMMARY 

A monitoring program has been implemented at the Route 351 Bridge project to compare 

the long-term performance, load-transfer, and durability characteristics of an FRP 

composite pile and a standard prestressed concrete pile. The two production piles selected 

for monitoring were installed at the centers of Piers 10 and 11. The piles consist of (1) a 

conventional, 610-mm (23.8-inch) square, prestressed concrete pile installed at Pier 10, 

and (2) a 622-mm (24.3-inch) diameter composite pile formed of an FRP shell and filled 

with concrete and steel reinforcing bars installed at Pier 11. The instrumented production 

piles are about 22 m (72 ft) long. 

Load-transfer instrumentation, consisting of 18 vibrating-wire sister bars, was installed at 

7 levels in both piles. Durability instrumentation, consisting of resistance foil gages and 

vibrating-wire gages, was installed on the FRP composite pile to monitor the inside and 

outside of the FRP tube, respectively. 

The soils at the production pile locations were investigated by VDOT, and consisted of 

very soft to soft silty clay underlain by loose to medium-dense silty sand. A sandy, silty, 

clay interbed was found at Pier 10 between elevations -18.7 and -20 m (61.3 and 65.6 ft), 

and at Pier 11 between elevations -11.6 to -20 m (38.0 and 65.6 ft). 

 
 

288 

The field testing of the instrumented production piles included PDA, CAPWAP, and PIT 

tests. The following conclusions can be drawn from this test program. 

1. The FRP composite pile and prestressed concrete pile performed similarly during pile 

driving, as demonstrated by the driving records, PDA traces, measured wave speed 

generated by the driving hammer, and measured compressive and tensile stresses. 

2. Using the Case method, instrumented production piles showed similar estimated pile 

capacities on pile restrike blows. 

3. CAPWAP analyses at pile restrike showed similar estimated pile capacities and high 

shaft capacity components. The shaft capacities were estimated as being 82 percent 

and 71 percent of the total capacity for the prestressed concrete and FRP composite 

piles, respectively. Based on the CAPWAP analyses, the pile capacities were 

estimated to be more than twice the axial design load (890 kN (200,250 lbf)) of the 

pier piles. 

4. No discernable damage was detected in the piles after installation, based on PIT 

testing performed on the piles before and after pile driving. 

5. Monitoring data of the piles is limited to date. On December 30, 2002, the loads 

acting on the piles were estimated as 215 and 275 kN (48,375 and 61,875 lbf) for the 

prestressed concrete pile and FRP composite pile, respectively. 

Durability instrumentation was installed on the FRP composite pile to monitor the inside 

and outside strains of the FRP tube. The durability instrumentation had to be installed 

after the construction activities near Pier 11 were completed. In addition to providing the 

strain data for the FRP shells, the durability strain gages will also provide records of 

temperature because vibrating-wire strain gages incorporate thermocouples to provide 

information necessary for reducing the strain gage data. The strain and temperature data 

will be used to interpret the performance of the FRP shells, as follows: knowledge of the 

exposure time and temperature can be used to estimate the FRP properties of modulus 

and strength from the laboratory durability study described in chapter 4; knowledge of 

the strain in the production piles can be combined with the modulus to give the stress in 

the FRP shell; and comparing the stress with the strength of the FRP shell indicates how 

close the shell is to failure.
 

289 

CHAPTER 10. COST INFORMATION FOR COMPOSITE PILES 

10.1 INTRODUCTION 

Composite materials have attractive features such as high resistance to corrosion, high 

strength-to-weight ratios, and the ability to be used in infrastructure projects with little or 

no maintenance requirements (Meiarashi, et al., 2002). Recent studies have found that the 

lower weight of bridge decks made of FRP composite materials results in greater cost 

effectiveness due to shorter construction time and decreased cost of the bridge 

superstructure (Ehlen 1999). The economical competitiveness of composite piles needs to 

be assessed. To help provide data for such assessments, this chapter provides cost 

information for the composite piles and prestressed concrete piles used in this research 

project. 

 

10.2 COST INFORMATION FOR THE ROUTE 40 BRIDGE PROJECT 

The cost per unit length of the concrete-filled FRP composite piles used at the Route 40 

Bridge project was $312/m ($95/ft). This unit cost includes the materials (GFRP tube and 

concrete infill), the manufacturing costs (including FRP tube and concrete casting), and 

shipping to the job site. The cost of driving the composite piles was $66/m ($20/ft). This 

results in a total cost of $378/m ($115/ft) for the installed composite pile. On the other 

hand, the total cost of the installed 508-mm (20-inch) square prestressed concrete pile 

was $213/m ($65/ft). Therefore, the initial cost comparison indicates about 77 percent 

higher unit cost for the composite pile. No instrumentation was installed in these piles. 

No significant time differences were observed in the installation time required for these 

two types of piles. The unit weights of the two piles were similar due to the concrete 

infill; therefore, the high strength-to-weight ratio of the composite material was not 

utilized. The composite piles were not easy to lift or handle, and required special 

arrangements and equipment. For example, the slickness of the FRP pile’s outer surface 

made it difficult to use conventional slings to lift the piles with the crane. In addition, the 

number of pick-up points had to be increased due to the lower cracking moment of these 

piles compared to the prestressed concrete pile. The difficulties encountered by the 

contractor in handling these piles also may have been associated with lack of experience 

in dealing with these piles. However, such problems are expected to occur in other 

projects until the use of composite piles becomes more common. 

10.3 COST INFORMATION FOR THE ROUTE 351 BRIDGE PROJECT 

The cost information for the three types of piles tested at the Route 351 Bridge project is 

provided below. These test piles were heavily instrumented, and the costs provided below 

do not include the instrumentation or its installation. However, pile fabrication and 

handling costs may have been increased due to the presence of the instrumentation. 
 

290 

10.3.1 Hardcore FRP Composite Pile 

The cost for the 24-inch FRP tubes was $165/m ($50/ft). This cost is freight-on-board 

(FOB) at the Hardcore plant in New Castle, DE. The cost associated with steel 

reinforcement, concrete infill, and transportation was about $541/m ($165/ft). The 

resulting total unit cost for the pile delivered on site is $705/m ($215/ft). The cost of 

installation of the test piles at this project was about $82/m ($25/ft). Therefore, the initial 

cost of the installed FRP composite pile is $787/m ($240/ft). 

10.3.2 Plastic Pile 

The ordinary unit cost for this type of pile FOB at the Plastic Piling, Inc., (PPI) plant in 

Rialto, CA is $410/m ($125/ft). The cost for shipment of this test pile from Rialto, CA, to 

Hampton, VA, was $209/m ($64/ft). The resulting total unit cost for the pile delivered on 

site was $619/m ($189/ft). The cost of installation for the test piles at this project was 

about $82/m ($25/ft). Therefore, the initial cost of the installed PPI composite pile was 

$701/m ($214/ft). 

10.3.3 Prestressed Concrete Pile 

The total cost of the installed 610-mm (24-inch) square prestressed concrete pile was 

about $180/m ($55/ft). 

10.3.4 Summary of the Route 351 Bridge Cost Information 

The initial cost information given above indicates that the unit costs for the PPI plastic 

composite pile and the Hardcore composite pile are 289 percent and 337 percent higher 

than for the prestressed concrete pile, respectively. 

No significant installation time differences were observed during handling and 

installation of the composite piles compared to the prestressed concrete pile. The 

handling and lifting of the FRP composite piles was the most difficult due to the slickness 

of the outer surface of the FRP shell. Special arrangements were required to handle this 

pile, including using steel anchors to keep the lifting slings from sliding and rolling. The 

location and number of pickup points for handling this pile had to be modified with 

respect to conventional practice due to the lower cracking moment of these piles 

compared to the prestressed concrete piles. These kinds of complications were not 

encountered during the handling and installation of the plastic composite pile. 

10.4 SUMMARY 

The initial costs of the composite piles studied in this project were higher than the initial 

unit costs for prestressed concrete piles. The initial unit costs of the installed composite 

piles at the Route 40 Bridge were about 77 percent higher than the unit costs for the 

prestressed concrete piles. The initial unit costs for the composite piles installed at the 
 

291 

Route 351 Bridge were higher than the cost of the prestressed concrete piles by about 289 

percent and 337 percent for the plastic and FRP piles, respectively. 

The cost effectiveness of the composite piles is expected to improve with economies of 

scale as production volume increases. The low maintenance requirements of these 

composite piles also increase their cost effectiveness. Life cycle cost analyses should 

consider the pile lifespan, the annualized maintenance costs, and the replacement costs of 

composite piles compared to prestressed concrete piles. The number of years of use that 

should be factored into the life cycle cost analyses may not be related only to the lifespan 

of the piles, but may also be governed by the actual lifespan of the bridge superstructure, 

which may be related to deterioration or increased traffic demands. A life cycle cost 

analysis was not performed for this study due to lack of maintenance cost and frequency 

information.

 

293 

CHAPTER 11. SUMMARY AND CONCLUSIONS 

11.1 INTRODUCTION 

This study investigated the potential use of composite piles for load-bearing applications, 

specifically bridge substructures. The term “composite piles” refers to alternative pile 

foundations composed of FRPs, recycled plastics, or hybrid materials that are placed in 

the ground to support axial and/or lateral loads. Conventional piles are made of steel, 

concrete, and timber; these materials have limited service life and high maintenance costs 

when used in harsh marine environments (Lampo, et al., 1998). Degradation problems of 

conventional piles include chloride attack on concrete, steel corrosion, and marine borer 

attack on timber piles. It has been estimated that repair and replacement of piling systems 

costs the United States more than $1 billion annually (Lampo, et al., 1997). High repair 

and replacement costs drive the need to investigate the feasibility of using FRP composite 

materials for pile foundations (Lampo, et al., 1998; Iskander and Hassan 1998). FRP 

composite materials are considered an attractive alternative for marine and other harsh 

environments because they are resistant to the degradation mechanisms mentioned above. 

Composite piles have been available in the North American market since the mid 1980s, 

but their use to date has been limited mainly to marine fender piles, load-bearing piles for 

light structures, and experimental test piles (Iskander, et al., 2001). Composite piles have 

not yet gained wide acceptance in the civil engineering industry, primarily due to the lack 

of a long track record of performance. However, FRP composite piles may have a longer 

life cycle and improved durability in harsh marine environments, which presents the 

potential for substantially reduced costs. 

To confidently assess the feasibility of using composite piles for load-bearing structures, 

information and performance data were gathered in critical areas of structural behavior 

and performance; long-term durability; and geotechnical behavior, including driveability 

and soil-pile load-transfer interactions. The overall objective of this research project was 

to evaluate the feasibility of using composite piles in bridge substructures. Table 52 

presents the four detailed objectives of the project and indicates how each objective was 

met. 
 

294 

Table 52. Detailed project objectives. 

Objective Evidence of Objective Completion 

Evaluation of the long-term durability of 

concrete-filled FRP composite pipe piles. 

An experimental study was designed and 

implemented to investigate the degradation 

of the mechanical properties of glass FRP 

composites used to fabricate composite piles. 

The study investigated the effects of moisture 

absorption and exposure to freeze-thaw 

cycles. 

Performance of laboratory tests to 

evaluate the soil-pile interface behavior 

of concrete-filled FRP composite piles 

and compare them with tests carried out 

on conventional prestressed concrete 

piles. 

Interface shear tests were carried out on four 

FRP composite piles, one plastic pile, and 

two conventional piles. 

Design and performance of a test pile 

program to evaluate the driveability, axial 

and lateral behavior, and capacity of 

concrete-filled FRP composite piles 

compared to conventional prestressed 

concrete piles. 

Full-scale test pile programs were performed 

at two bridge sites in Virginia. Field data 

were gathered, analyzed, and disseminated to 

contribute to the current state of knowledge 

of composite piles. 

Design and performance of a production 

pile testing and monitoring program in a 

real bridge project to compare long-term 

performance, load-transfer capability, and 

durability of concrete-filled FRP 

composite piles and conventional 

prestressed concrete piles. 

A long-term monitoring program was 

designed and implemented at the Route 351 

Bridge. The program consisted of 

instrumenting two production piles of the 

bridge: a concrete-filled FRP pile and a 

conventional prestressed concrete pile. The 

long-term monitoring of the instrumented 

piles will provide useful information to 

compare the performance of these two pile 

types. The long-term monitoring will be 

carried out by VDOT. 

The remainder of this chapter provides summaries of the projects’ activities and 

conclusions. The chapter concludes with a discussion of the practical implications of this 

research, and recommendations for future research. 

11.2 SUMMARY OF ACTIVITIES AND CONCLUSIONS 

This section summarizes the activities and conclusions for the literature review, the 

interface laboratory study, the durability study of concrete-filled FRP tubular piles, the 

field tests at the Route 40 and Route 351 Bridge projects, the analytical studies on axial 

and lateral behavior, the long-term monitoring program, and the composite pile cost 

information. 
 

295 

11.2.1 Literature Review 

A literature review was performed to: (1) identify suitable composite pile candidates to 

be used in high load-bearing projects such as bridge substructures, (2) identify the 

research needs, and (3) help design the work plan for this research project. 

11.2.2 Interface Study 

The following laboratory activities were performed for the interface study: 

• Selection of sand specimens for interface testing. 

• Soil characterization tests to determine grain size distribution, minimum/maximum 

density, and specific gravity testing. 

• Direct shear tests on Density and Model sands to determine internal friction angles. 

• Selection of pile surface specimens for interface testing. 

• Surface topography characterization and surface hardness determination of seven pile 

surfaces. The seven pile surfaces included four FRP composites (commercially 

available from two FRP composite pile manufacturers), one recycled plastic pile, and 

two conventional pile materials (a prestressed concrete pile and a steel sheet pile). 

• Design and construction of a modified top half shear box to permit testing of curved 

pile surfaces obtained from the FRP composite tubes. 

• Interface shear tests for the seven piles and two sand types to determine the interface 

behavior and interface friction angles of these interfaces. 

Interface shear tests were performed on 14 types of soil-pile interfaces. Tests were carried 

out for two types of sands: Density sand (D50 = 0.5 mm (0.0195 inch), subrounded to 

rounded particle shape) and Model sand (D50 = 0.18 mm (0.00702 inch), subangular to 

angular particle shape). Seven pile surfaces were tested: Lancaster FRP composite 

(curved), Hardcore FRP composite (curved), Hardcore FRP composite plate (flat), 

Hardcore FRP composite plate with bonded sand treatment (flat), PPI recycled plastic 

coupon (flat), prestressed concrete pile coupon (flat), and a steel sheet pile coupon (flat). 

The peak and residual interface friction angles for the 14 sand-to-pile interfaces tested are 

summarized in tables 10 through 12 (chapter 3). 

From the results of the interface study, the following conclusions and observations were 

made: 

1. The interface friction angle values obtained for the Density sand tested against the 

Lancaster FRP composite pile were much lower than the values obtained for the other 
 

296 

pile surfaces. The peak and residual interface friction angles obtained for the 

Lancaster FRP composite pile were 19.7o and 16.6o, respectively. In contrast, the 

interface friction angle values obtained for the Density sand tested against the other 

pile surface types ranged from 27.6o

 to 33o, and from 24.9o to 27.8o, for the peak and 

residual conditions, respectively. 

2. The Lancaster FRP composite pile was found to have the largest average mean line 

spacing (Sm) and the second-lowest maximum peak-to-valley height (Rt ) as reported 

in table 8 (chapter 3). These values make this pile surface the smoothest of all seven 

surfaces tested. 

3. In general, the subangular to angular Model sand gives slightly higher interface 

friction angles than the subrounded to rounded Density sand. However, for the 

Lancaster FRP composite pile, the interface friction angles obtained with the Model 

sand are much higher than the Density sand values. 

4. In general, the interface friction angles, both peak and residual, were found to 

increase with increasing relative asperity height (Rt/D50). 

5. In general, the interface friction angles, both peak and residual, were found to 

increase with decreasing relative spacing (Sm/D50), which is reasonable over the range 

of Sm/D50 tested. 

6. Linear regression analyses between the interface friction coefficients and the 

variables Rt/D50, Sm/D50, and HV showed moderate fit strengths (where Rt and Sm are 

surface roughness parameters and HV is the Vickers hardness number). The 

regression analyses suggest other factors besides Rt/D50, Sm/D50, and HV also have 

important influences on the values of interface friction coefficient. 

7. The bonded sand surface treatment used for the Hardcore FRP plate was successful in 

increasing the interface friction angle values. 

11.2.3 Durability Study 

A laboratory testing program was completed to study the long-term performance of FRP 

composite pipe piles. This durability study addresses the FRP shells of the Lancaster and 

Hardcore composite piles, and included FRP shell characterization, determination of 

baseline mechanical properties, measurement of moisture absorption as a function of time 

and temperature, measurement of mechanical properties as a function of moisture 

absorption, and measurement of mechanical properties as a function of freeze-thaw 

cycles. 

From the results of the durability study, the following conclusions and observations were 

made: 
 

297 

1. The experimental program revealed strength and stiffness degradation due to moisture 

absorption at room temperature of up to 25 percent for the E-glass composites tested. 

These levels of degradation correspond to steady-state moisture contents reached after 

about 2.5 years of submergence. 

2. Degradation of longitudinal mechanical properties of the 24-inch (0.61-m) Lancaster 

pile was not as significant as for the Hardcore pile because of the different fiber layups. The Lancaster pile was matrix dominated in the longitudinal direction, since the 

fibers with closest alignment to the longitudinal axis were 35 degrees off alignment. 

3. For the composite piles with fiber-dominated composite, the levels of degradation of 

mechanical properties were similar in hoop and axial directions. 

4. The impact of FRP degradation on the long-term structural capacity of the piles was 

investigated using models recommended by Fam and Rizkalla (2001a and b, and 

2002). This approach showed that, for a 12-inch (0.305-m) diameter FRP pile, the 

axial structural capacity will decrease about 5 percent if the FRP tube hoop properties 

degrade 25 percent and 40 percent in stiffness and strength, respectively. The small 

impact FRP degradation on the axial pile capacity is due to the fact that the majority 

of the capacity contribution comes from the concrete infill. The impact on larger 

diameter piles is even smaller. A similar analysis was carried out for the flexural 

capacity of a 12-inch (0.305-m) pile. The results show that a 24 percent reduction in 

flexural capacity can be expected if the FRP tube longitudinal tensile properties 

degrade 25 percent and 40 percent in stiffness and strength, respectively. 

5. Exposure to freeze-thaw cycles was found to have little effect on the longitudinal 

tensile properties of the saturated FRP tubes. This conclusion is based on longitudinal 

tensile tests of samples exposed to 100, 300, and 500 freeze-thaw cycles. 

11.2.4 Field Tests at the Route 40 Bridge 

In 2000, VDOT replaced the old Route 40 Bridge (Structure No. 1006) over the 

Nottoway River. New precast composite piles, consisting of concrete-filled GFRP tubes, 

were used to support one of the cast-in-place reinforced concrete cap beams that directly 

support the superstructure. Before construction, full-scale field tests were undertaken to 

investigate the feasibility of construction, handling, and driveability, as well as the 

structural performance of the new composite piles in comparison to conventional 

prestressed concrete piles. 

Based on the testing, analysis, and construction experience at the Route 40 Bridge, the 

following conclusions and observations were made. 

1. Both the composite and prestressed concrete piles performed similarly in the axial 

load tests. Full geotechnical capacity was mobilized in both cases without structural 

failure of the piles. The axial loads at geotechnical capacity were significantly higher 

than the design pile load. 
 

298 

2. Initially, the composite test pile exhibited a lateral stiffness similar to that of the 

prestressed concrete test pile up to a load of about 40 to 50 kN (9,000 to 11,250 lbf). 

This load level was found to correspond to first cracking of the composite pile. 

Beyond this load, the composite pile exhibited a much lower stiffness than the 

prestressed concrete pile, but the composite pile did demonstrate a continuing ability 

to sustain lateral load with additional relatively large deformations. 

3. The flexural strength of the 625-mm (24.6-inch) diameter composite pile with a 5.85-

mm (0.230-inch) thick GFRP tube was calculated to be higher than the flexural 

strength of the 508-mm (20-inch) square concrete pile prestressed with fourteen, 

12.7-mm (0.5-inch) strands, as shown in figure 99 (chapter 5). 

4. Calculations show that the composite pile fails in bending by fracture of the GFRP 

tube on the tension side, while the prestressed concrete pile fails by yielding of the 

strands in tension, followed by crushing of the concrete in compression. 

5. The lateral load field tests on both the composite and prestressed piles showed similar 

behavior to that obtained from laboratory flexural tests and analysis. 

6. The lateral load-versus-deflection response for the composite pile was predicted with 

good accuracy using conventional procedures typically used for prestressed concrete 

piles. 

7. Similar pile-to-cap beam connections were used for the composite and prestressed 

concrete piles, including eight No.7 steel dowels embedded 457 mm (18 inches) 

inside the piles from one end and extending 762 mm (30 inches) into the cap beam. 

The piles themselves were embedded 152 mm (6 inches) inside the cap beam. 

8. Concrete-filled FRP tubes were successfully installed as piling for a bridge pier at the 

Route 40 Bridge. To date, no indications of unsatisfactory performance have been 

reported. 

11.2.5 Field Tests at the Route 351 Bridge 

A test pile program was conducted to permit direct comparison of the axial and lateral 

load behavior of three test piles: (1) a conventional, 610-mm (23.8 inch) square 

prestressed concrete pile; (2) a 622-mm (24.3-inch) diameter composite pile formed of an 

FRP shell and filled with concrete and steel reinforcing bars; and (3) a 592-mm (23.1-

inch) diameter composite pile formed of a polyethylene plastic matrix with steel 

reinforcing bars. Laboratory tests were performed on the pile materials, and the axial and 

bending stiffnesses of the piles were calculated. The piles are all about 18 m (59.0 ft) 

long. The soil at the test site consists primarily of medium-dense silty sand with a layer of 

stiff sandy clay located about 1 to 3.5 m (3.28 to 11.5 ft) above the pile toes. The field 

testing program included dynamic measurements with CAPWAP analyses, static axial 

load tests, and static lateral load tests. 
 

299 

Based on the testing, analysis, and construction experience, the following conclusions 

and observations were made. 

1. The axial stiffness (EA) of the prestressed concrete pile and the FRP pile are similar 

to each other and about 2.5 times the axial stiffness of the plastic pile. 

2. Over a working range of bending moments, the flexural stiffness (EI) decreases in 

order from the prestressed concrete pile to the FRP pile to the plastic pile. 

3. All three piles exhibited substantial setup after installation based on comparisons of 

Case method capacities at the end of initial driving and restrike 5 days later, as well 

as comparisons of CAPWAP analyses of restrike and static axial load tests several 

days after restrike. 

4. Applying Davisson’s failure criterion to the static axial load tests, the axial capacities 

were 3,090, 2,260, and 2,130 kN (695,250, 508,500, and 479,250 lbf) for the 

prestressed concrete pile, the FRP pile, and the plastic pile, respectively. 

5. When evaluated at the Davisson failure load, the average unit shaft resistances are 

61.8, 46.9, and 48.9 kPa (8.96, 6.80, and 7.09 lbf/inch2

). for the prestressed concrete 

pile, the FRP pile, and the plastic pile, respectively, and the corresponding unit toe 

resistances are 1,854, 2,564, and 2,339 kPa (268.8, 371.8, and 339.2 lbf/inch2). 

6. The prestressed concrete pile exhibited a geotechnical capacity slightly above the 

midpoint between the capacities calculated using the LCPC methods for core 

penetrometer testing of steel and concrete piles of the same geometry as the test pile. 

The FRP and the plastic piles exhibited geotechnical capacities much closer to the 

LCPC-derived calculations for steel piles than for concrete piles of the same 

geometries as the test piles. 

 

11.2.6 Axial Analyses 

The axial capacities of the three test piles were predicted using static analysis methods 

commonly employed in practice. The methods employed were the Nordlund method, the 

API method, the LCPC method, and the IC method. The various methods used to predict 

axial pile capacity yielded ratios of calculated to measured pile capacities ranging from 

0.70 to 1.14 for the prestressed concrete pile, and from 0.81 to 1.33 for the composite 

piles. It was found that the level of accuracy of the predictions was comparable for all 

three test piles. 

The prediction results seem to indicate that conventional static analysis methods are 

applicable to composite piles. However, it is the author’s opinion that additional case 

histories are needed to corroborate and extend this conclusion to other composite pile 

types and to different soil conditions. 
 

300 

The load-settlement behavior was predicted using T-Z analyses. Several analyses were 

carried out to determine the adequacy of this method to analyze axially loaded composite 

pile types such as those studied in this research project. The analyses were completed 

using empirical load-transfer curves available in the literature (e.g., API 1993, 

Vijayvergiya 1977) and using theoretically derived load-transfer curves. 

The predictions using the empirical load-transfer curves showed reasonable agreement 

with the field measurements for the composite piles. The predictions for the prestressed 

concrete pile showed less agreement, and tended to underestimate the measured pile 

capacity. This result could be related to the value of the interface friction angle used. The 

calculations were based on the interface friction angle measured in the lab, which may be 

inadequate to represent the field conditions experience by the pile (e.g., in the field the 

pile interface experienced much larger relative displacements compared to the 0.5 inches 

(12.7 mm) used in the laboratory). 

The predictions obtained using the theoretically derived load-transfer curves were in very 

good agreement with the measured field behavior. However, the values of the coefficient 

of horizontal earth pressures (K) that resulted in the best matches with field behavior 

were 1.25, 1.0, and 0.9 for the prestressed concrete, FRP, and plastic piles, respectively. 

The differences in the K coefficient values are contrary to the expected outcome. The 

expected outcome was to have similar K values for the three test piles since they were 

installed in similar soil conditions, and all are full-displacement piles with similar crosssectional dimensions. It is possible that this result is again related to the δ values 

measured in the laboratory not being representative of field conditions, perhaps due to the 

small displacements (only 0.5 inches (12.7 mm)) used in the lab. 

The prediction results seem to indicate that conventional T-Z analyses are applicable to 

composite piles. However, it is believed that additional case histories are needed to 

corroborate and extend this conclusion to other composite pile types and to different soil 

conditions. 

11.2.7 Lateral Analyses 

Numerous p-y analyses were carried out to determine the adequacy of this method to 

analyze laterally loaded composite piles, such as the ones studied in this research project. 

A derivation of the governing differential equation for the laterally loaded pile problem 

was presented, and possible limitations in analyzing composite piles were analysis 

discussed. 

The importance of considering shear deformations in lateral pile analyses was discussed. 

The impact of shear deformations increases with increasing E/G ratios, and decreases 

with increasing slenderness ratios (L/D). For the test piles assessed in this project, the 

error associated with neglecting shear deformations is estimated to be less than 2.5 

percent. 
 

301 

The importance of including nonlinearity of the flexural stiffness was discussed and 

illustrated with analyses results. 

The results of the p-y analyses using published p-y curves that are embedded in the 

LPILE 4.0M computer program showed reasonably good agreement with the field 

measurements. 

The initial modulus of the p-y curves was found to increase with depth at the highest rate 

for the FRP pile, at an intermediate rate for the plastic pile, and at the lowest rate for the 

prestressed concrete pile. This outcome is not as expected, and the reason for the 

difference in initial modulus values is not known. 

Although the analysis results seem to indicate that conventional p-y analyses are 

applicable to composite piles, additional case histories are needed to determine what 

factors govern the appropriate value of the initial p-y modulus for different pile types in 

various soils. 

11.2.8 Long-Term Monitoring 

A monitoring program has been implemented at the Route 351 Bridge project to compare 

the long-term performance, load-transfer, and durability characteristics of an FRP 

composite pile and a standard prestressed concrete pile. The two production piles selected 

for monitoring were installed at the centers of Piers 10 and 11. 

The field testing of the instrumented production piles included PDA, CAPWAP, and PIT 

tests. The following conclusions can be drawn from this test program. 

1. The FRP composite pile and prestressed concrete pile performed similarly during pile 

driving, as demonstrated by the driving records, PDA traces, measured wave speed 

generated by the driving hammer, and measured compressive and tensile stresses. 

2. Using the Case method, both instrumented production piles showed similar estimated 

pile capacities on pile restrike blows. 

3. CAPWAP analyses at pile restrike showed similar estimated pile capacities and high 

shaft capacity components. The shaft capacities were estimated as being 82 percent 

and 71 percent of the total capacity for the prestressed concrete and FRP composite 

piles, respectively. Based on the CAPWAP analyses, the pile capacities were 

estimated to be more than twice the axial design load of 890 kN (200,250 lbf). 

4. No discernable damage was detected in the piles after installation, based on PIT 

testing performed on the piles before and after pile driving. 

Load-transfer instrumentation was installed in both piles. Durability instrumentation was 

installed on the FRP composite pile to monitor the inside and outside strains of the FRP 

tube. Monitoring data on the piles is limited to date; therefore, no conclusions can be 
 

302 

drawn at this time regarding the long-term performance of concrete-filled FRP composite 

piles. 

11.2.9 Cost Information for Composite Piles 

Composite materials have attractive features such as high resistance to corrosion, high 

strength-to-weight ratios, and the ability to be used in infrastructure projects with little or 

no maintenance requirements (Meiarashi, et al., 2002). Recent studies have found that the 

lower weight of bridge decks made of FRP composite materials results in greater cost 

effectiveness due to shorter construction time and decreased cost of the bridge 

superstructure (Ehlen 1999). The economical competitiveness of composite piles needs to 

be assessed. To help provide data for such assessments, cost information for the 

composite piles and prestressed concrete piles used in this research project was compiled 

and presented. 

The initial costs of the composite piles studied in this project were found to be higher 

than the initial unit costs for prestressed concrete piles. The initial unit costs of the 

installed composite piles at the Route 40 Bridge were about 77 percent higher than the 

unit costs for the prestressed concrete piles. The initial unit costs for the composite piles 

installed at the Route 351 Bridge were higher than costs of the prestressed concrete piles 

by, for the plastic piles, about 289 percent and, for the FRP piles, about 337 percent. 

The cost effectiveness of composite piles is expected to improve with economies of scale 

as production volume increases. The low maintenance requirements of these composite 

piles also increase their cost effectiveness. Life cycle cost analyses should consider the 

pile lifespan, the annualized maintenance costs, and the replacement costs of composite 

piles compared to prestressed concrete piles. The number of years of use that should be 

factored into the life cycle cost analyses may not be related only to the lifespan of the 

piles but may also be governed by the actual lifespan of the bridge superstructure, which 

depends on the rate of superstructure deterioration and changing traffic demands. Life 

cycle cost analyses were not performed for this study due to lack of maintenance cost and 

frequency information. 

11.3 RECOMMENDATIONS FOR FUTURE WORK 

Based on the findings from this investigation, the following recommendations are made 

for future work on composite piles. 

11.3.1 Geotechnical Studies 

Additional full-scale field load test programs are recommended to extend the conclusions 

of this study to other composite piles and soil types. Well-documented field loading tests 

of composite piles are scarce, and this lack of reliable data may be one reason that 

composite piles are not in widespread use for load-bearing applications. Therefore, it is 

important to increase the database of instrumented load tests of composite piles. 
 

303 

Finite element analyses are recommended to model the field tests of the test piles. This 

type of analysis takes into account the continuous nature of the soil deposits, as opposed 

to the T-Z and p-y approaches used in this study. Comparison of the results from the 

different analysis approaches would be useful. 

Laboratory studies involving lateral load tests on model or small-diameter piles installed 

in uniform and controlled soil conditions may be useful to study the influence that pile 

characteristics and flexural stiffness have on p-y curves. 

Pile damage should be assessed after pile driving by excavating the piles. Visual 

inspections of the retrieved piles will permit assessment of changes in the pile surface due 

to scratching and scraping of the outer surface. The conclusions drawn from the 

laboratory interface study regarding the influence of surface roughness and hardness 

characteristics on interface friction behavior at relatively small displacements may not be 

applicable to real piles that may experience severe surface wear during pile installation. 

11.3.2 Durability Studies 

Studies of small-diameter concrete-filled FRP piles may be useful to examine the 

durability of axial and flexural structural capacity. Smaller diameter tubes could allow a 

larger number of tests at a reasonable cost and within a reasonable time frame. A larger 

number of tests would permit interpreting the data scatter associated with inherent 

variability of the material properties. 

A durability study that includes piles exposed to salt water and other aqueous solutions 

representing groundwater conditions in industrial sites is recommended. 

Other recommended studies include assessment of degradation due to UV radiation and 

exposure to high temperatures, resistance to fire, and corrosion of the steel cage in PPI 

piles. 

11.3.3 Structural Tests 

Test involving compression, bending, and combined axial and bending loading are 

recommended to determine the full interaction diagrams of composite piles. To date, 

limited data is available for the combined loading condition. 

The influence of temperature on structural behavior should be determined, and the effects 

that differences in coefficients of thermal expansion of the different pile constituents have 

on pile performance also require study. For example, in concrete-filled FRP piles, the 

coefficient of thermal expansion of the FRP tube will be a function of factors such as 

fiber and resin type, number of fiber layers, fiber lay-up, fiber volume fraction, etc. If the 

resulting thermal coefficient of the FRP tube is significantly different from that of the 

concrete core, it is conceivable that the two materials would debond. 
 

304 

Resistance to fire exposure should be investigated. Although the majority of the pile 

length is embedded in the ground or submerged in water, there is usually a portion of the 

pile that is exposed. Although the probability of exposure to fire is low, fire resistance 

properties could be important in locations where the piles form part of priority bridge 

structures that have to survive more severe conditions. 

Additional recommended studies include tests to assess the behavior under cyclic loading 

of composite piles, studies of creep under axial compression, and the influence of rate of 

loading on the pile structural behavior. 

11.3.4 Cost Analyses 

It is recommended that detailed life cycle cost analyses be performed for different 

composite pile types and conventional piles.
 

305 

APPENDIX A. INTERFACE TEST RESULTS 

This appendix contains the results of all the interface tests performed for the interface 

study described in chapter 3. 

Table 53. Organization of the figures in appendix A. 

Pile/Material Density Sand (1) Model Sand (2)

Lancaster composite FRP shell (CP40) Figure 221 Figure 228 

Hardcore composite FRP shell (24-4) Figure 222 Figure 229 

Hardcore FRP plate—untreated surface Figure 223 Figure 230 

Hardcore FRP plate—treated surface Figure 224 Figure 231 

PPI plastic pile Figure 225 Figure 232 

Prestressed concrete pile Figure 226 Figure 233 

Steel sheet pile Figure 227 Figure 234 

Notes: (1) See table 10, chapter 3, for a summary of interface shear test results on Density sand. 

 (2) See table 11, chapter 3, for a summary of interface shear test results on Model sand. 
 

306 

Figure 221. Graphs. Interface shear test results, Density 

sand-to-Lancaster FRP shell interface. 
 

307 

Figure 222. Graphs. Interface shear test results, Density 

sand-to-Hardcore FRP shell interface. 
 

308 

Figure 223. Graphs. Interface shear test results, Density 

sand-to-Hardcore FRP plate (untreated) interface. 
 

309 

Figure 224. Graphs. Interface shear test results, Density 

sand-to-treated Hardcore FRP plate interface. 
 

310 

Figure 225. Graphs. Interface shear test results, Density sand-to-plastic interface. 
 

311 

Figure 226. Graphs. Interface shear test results, Density sand-to-concrete interface. 
 

312 

Figure 227. Graphs. Interface shear test results, Density sand-to-steel interface. 
 

313 

Figure 228. Graphs. Interface shear test results, Model 

sand-to-Lancaster FRP shell interface. 
 

314 

Figure 229. Graphs. Interface shear test results, Model 

sand-to-Hardcore FRP shell interface. 
 

315 

Figure 230. Graphs. Interface shear test results, Model 

sand-to-Hardcore FRP plate (untreated) interface. 
 

316 

Figure 231. Graphs. Interface shear test results, Model 

sand-to-treated Hardcore FRP plate interface. 
 

317 

Figure 232. Graphs. Interface shear test results, Model sand-to-plastic interface. 
 

318 

Figure 233. Graphs. Interface shear test results, Model sand-to-concrete interface. 
 

319 

Figure 234. Graphs. Interface shear test results, Model sand-to-steel interface.

 

321 

APPENDIX B. MOISTURE DIFFUSION INTO A CYLINDRICAL 

FRP COMPOSITE 

The procedure for calculating moisture diffusion into the FRP of a composite pile is 

presented in this appendix. The diffusion is assumed to be Fickian, and based on the 

geometry of the FRP pile, it is further assumed to be predominantly in the radial 

direction. For these conditions, the governing partial differential equation for the 

moisture concentration is given by 

 (70) 

where C is the moisture concentration (also C = M(t)) and Dr is the diffusivity for radial 

diffusion. 

If we assume the FRP shell is initially dry, the initial condition for the moisture 

concentration is given by 

 (71) 

The last requirement for the solution of the diffusion problem is that of the boundary 

conditions on the inner and outer radii of the FRP shell. As a first approximation, we 

specify that these concentrations are constant; that is, 

 (72) 

where ri and ro are the inner and outer radii, respectively, and Ci and Co are the 

corresponding concentrations. 

To solve equation 70 subject to the initial condition given by equation 71 and the 

boundary conditions in equation 72, the separation-of-variables technique can be used. 

This technique requires homogeneous boundary conditions rather than inhomogeneous 

conditions. Therefore, a solution of the following form is sought: 

 (73) 

where ( ) ~

C r satisfies the time-independent form as follows: 

 (74) 
 

322 

with boundary conditions given by 

 

 (75) 

This solution is given by 

 (76) 

By combining equation 76 with equations 70 through 72, we find that c(r.t) is the 

solution of equation 70 with boundary conditions given by 

 (77) 

and an initial condition given by 

 (78) 

Problems of this type have been studied in detail by a number of researchers. In 

particular, using the results of Necati Őzişik (1989), the solution is given by 

 (79) 

where 

 

 (80) 
 

323 

and βm represents the positive root of the transcendental 

 (81) 

where J0 and Y0 are Bessel functions of the first and second kind (order zero), 

respectively. 

The moisture content for the FRP shell can be bound by two extreme cases that should 

bound the FRP moisture absorption behavior. In the first case, we assume that the 

composite is saturated on the outer radius due to its immersion in water, and that it 

remains completely moisture-free on the inner surface. The corresponding boundary 

conditions are given by 

 (82) 

where M∞ is the saturation concentration from the experimental data in chapter 4. 

Assuming that C(ri,t) = 0 should give the lowest values for the moisture content in the 

pile, and hence the highest FRP strength (since the strength is reduced by an increase of 

moisture). 

In the second case, we assume that the FRP shell is saturated on both the inner and outer 

radii. The corresponding boundary conditions are given by 

 (83) 

This case should give us the highest values for the moisture content in the FRP shell of 

the pile, and hence the lowest FRP strength. 

The procedure described above provides a means to estimate the bounds for the moisture 

content distribution within the FRP shell of a composite pile. To do so, the values of M∞

and diffusivity (Dr) must be known. For the piles studied in this project, these values can 

be taken from tables 23-26. 

 

The results of this approach for the Lancaster 24-inch FRP shell are shown in figures 235 

and 236. These figures correspond to the predicted moisture concentration profiles for the 

two boundary condition cases described above. The value of D used corresponds to 22 o

C 

(72 °F) temperature. 
 

324 

Figure 235. Graph. FRP moisture concentration profile, 

inner radius dry, outer radius saturated. 

Figure 236. Graph. FRP moisture concentration profile, 

inner and outer radii saturated.
 

325 

APPENDIX C. STRUCTURAL TESTS RESULTS FROM 

COMPOSITE PILE CUTOFF SECTIONS FROM 

THE ROUTE 40 BRIDGE PROJECT 

This appendix contains results from structural tests carried out on cutoff Lancaster 

composite pile samples obtained for the Route 40 bridge project. The structural tests 

include: 

C.1. Pushout tests to assess the bond strength between the concrete core and the inner 

surface of the FRP tubes. 

C.2. A creep bending test carried out on an 18-ft (5.49-m) long pile piece. 

The results are summarized in the following pages. 
 

326 

C.1 PUSHOUT TEST RESULTS 

Pushout tests were performed to investigate the strength of the bond between the concrete 

and the FRP shell of Lancaster composite piles. The pushout test specimens were 

prepared by cutting slices from composite pile cutoff sections from the Route 40 Bridge 

project. The pushout specimens had a nominal diameter of 0.6 m (24 inches), and ranged 

from 75 mm to 150 mm (3 to 6 inches) in length. 

Some of these tests were performed by VTRC, and some were performed by Virginia 

Tech (VT). Each institution used a different test setup. The VTRC tests were performed 

by setting the test specimens on a block with a carefully fabricated hole that permits the 

block to support the FRP shell but allows the concrete core to pass through the hole. The 

load was applied at a rate of 3,000 lbs/min (1,362 kg/min). The VTRC test setup is shown 

in figure 237. 

Figure 237. Photo. VTRC pushout test setup. 
 

327 

The tests at VT were performed by bonding the outside of the FRP shell to a slightly 

larger steel pipe section, supporting the steel pipe section, and pushing the concrete core 

so it begins to move out of the FRP shell. Tests were carried out at a displacement rate of 

0.002 inches/min (0.0508 mm/min). The VT test setup for the pushout tests is shown in 

figure 238. The test results are summarized in table 54. 

Figure 238. Photo. VT pushout test setup. 

Table 54. Summary of pushout test results.

VTRC (1) VT (2) Total 

Number of tests 13 6 19 

Average peak 

bond strength (psi) 29.95 14.1 24.9 

Standard deviation (psi) 9.75 20.1 15.3 

Notes: (1) Load controlled test at 3,000 lb/min (1,362 kg/min). 1 psi = 6.89 kPa 

(2) Displacement controlled tests. Bond strengths were corrected for voids in the bond area. 
 

328 

As shown in table 54, regardless of the test setup, the bond strength values obtained are 

fairly low, with an overall bond strength average of 170 kPa (24.9 psi). Test results also 

show a wide scatter of bond strength values. This wide variability of results could be 

related to factors such as the presence of voids between the concrete core and the interior 

of the FRP shell; possible damage to bond during cutting, shipping, and handling; 

temperature effects (expansion and contraction); and scale or size effects. Smaller bond 

strengths were obtained using the VT test setup compared to the VTRC test results. The 

bond strength magnitude differences are likely related to inherent differences in the test 

setup and load application procedure, as well as possible additional specimen disturbance 

of the VT specimens during the extra shipping and handling required to move the 

specimens from VTRC to VT. 

 

C.2 CREEP BENDING TEST 

A four-point creep bending test setup was designed and built to perform a creep test on an 

18-ft (5.49-m) long section of 24-inch (0.61-m) diameter Lancaster composite pile. This 

pile was left over from VDOT’s Route 40 Bridge project, where Lancaster composite 

piles were used for one bent. The test facility was built at VT’s Kentland Farms, 

approximately 16 km (10 miles) west of Blacksburg, VA. 

A photograph of the creep bending test setup is shown in figure 239. As shown in this 

photo, two dead weights of about 32 kN (7.2 kips) are hung at the middle thirds of the 

beam. These loads induce a maximum moment of about 58 kN-m (43 kip-ft). This 

moment is not a large percentage of the expected ultimate moment, but it is a moment 

that can be applied economically. 

The centerline deflection curve obtained from the test is shown in figure 240. 
 

329 

 

Figure 239. Photo. Creep bending test setup. 

Figure 240. Graph. Creep deflection test results. 

 

331 

APPENDIX D. GEOTECHNICAL FIELD INVESTIGATIONS AT 

THE ROUTE 351 BRIDGE TEST SITE 

This appendix contains the results of geotechnical field investigations completed at the 

test site of the Route 351 Bridge project. 
 

332 

Figure 241. Map. Location of field tests. 
 

333 

Figure 242. Chart. Boring B-1 (SPT-1). 
 

334 

Figure 242. Chart. Boring B-1 (SPT-1)—Continued 
 

335 

Figure 243. Chart. Boring B-2 (SPT-2). 
 

336 

Figure 243. Chart. Boring B-2 (SPT-2)—Continued 
 

337 

Figure 244. Chart. CPT-1. 
 

338 

Figure 245. Chart. CPT-2. 
 

339 

Figure 246. Chart. CPT033. 
 

340 

Figure 246. Chart. CPT033—Continued 
 

341 

Figure 246. Chart. CPT033—Continued
 

342 

Figure 246. Chart. CPT033—Continued
 

343 

Figure 247. Chart. CPT034. 
 

344 

Figure 247. Chart. CPT034—Continued 
 

345 

Figure 247. Chart. CPT034—Continued
 

346 

Figure 248. Chart. DMT. 
 

347 

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